U.S. Department of Transportation 

Federal Highway Administration

Publication No. FHWA-NHI-16-027

FHWA GEC 013 

April 2017

NHI Course No. 132034

Ground Modification Methods

Reference Manual – Volume I
NOTICE 

The contents of this document reflect the views of the authors, who are responsible for the facts and 

accuracy of the data presented herein. The contents do not necessarily reflect policy of the Department 

of Transportation. This document does not constitute a standard, specification, or regulation. The United 

States Government does not endorse products or manufacturers. Trade or manufacturer’s names appear 

herein only because they are considered essential to the objective of this document. 

COVER PHOTO CREDITS

Upper left: Massachusetts Department of Transportation

Upper middle: MixOnSite USA, Inc.

Upper right: Bob Lukas, Ground Engineering Consultants, Inc.

Lower left: Menard Group USA

Lower middle: Subsurface Constructors 

Lower right: Hayward Baker
Technical Report Documentation Page

1. Report No.

FHWA-NHI-16-027

2. Government Accession No. 3. Recipient's Catalog No.

4. Title and Subtitle

GEOTECHNICAL ENGINEERING CIRCULAR NO. 13 

GROUND MODIFICATION METHODS - REFERENCE MANUAL

VOLUME I

5. Report Date

December 2016 

6. Performing Organization Code

7. Author(s)

Vernon R. Schaefer, Ryan R. Berg, James G. Collin, Barry R. 

Christopher, Jerome A. DiMaggio, George M. Filz, Donald A. Bruce, 

and Dinesh Ayala

8. Performing Organization Report No.

9. Performing Organization Name and Address

Ryan R. Berg & Associates, Inc.

2190 Leyland Alcove 

Woodbury, MN 55125 

10. Work Unit No. (TRAIS)

11. Contract or Grant No.

DTFH61-11-D-00049/0009

12. Sponsoring Agency Name and Address

National Highway Institute

U.S. Department of Transportation 

Federal Highway Administration, Washington, DC 20590

13. Type of Report and Period Covered

14. Sponsoring Agency Code

15. Supplementary Notes

FHWA COTR: Heather Shelsta

FHWA Technical Working Group Leader: Barry Siel, PE; Silas Nichols, PE; Scott Anderson, PhD, PE; and 

Brian Lawrence, PE. 

Contractor Technical Consultants: Jie Han, PhD, PE

This manual is the updated version of FHWA NHI-06-019/20, prepared by Ryan R. Berg & Associates, Inc.;

authored by V. Elias, J. Welsh, J. Warren, R. Lukas, J. Collin, and R. Berg; FHWA Technical 

Consultants J. DiMaggio and S. Nichols.

16. Abstract

This FHWA Geotechnical Engineering Circular No. 13 provides guidance on Ground Modification 

Methods, and also serves as the reference manual for FHWA NHI courses No. 132034, 132034A, and 132034B 

on Ground Modification Methods. The purpose of this manual is to introduce available ground modification 

methods and applications to design generalists, design specialists, construction engineers, and specification and 

contracting specialists involved with projects having problematic site conditions. 

An introductory chapter provides a description, history, functions, and categories of ground modification. 

A description of the web-based GeoTechTools (http://www.geotechtools.org) technology selection guidance 

system and geotechnology catalog is also provided in the first chapter. The introductory chapter is followed by 

stand-alone technical category chapters. Each category chapter includes a broad introduction to the technical 

category including typical applications, a listing of common technologies used in the U.S., and summaries for 

specific technologies in the category. Each technology summary includes: description; advantages and 

limitations; applicability; complementary technologies; construction methods and materials; photographs; design 

guidance; quality assurance methods; costs; specifications; and reference list. Each technical category and the 

technology summaries therein reflect current practice in design, construction, contracting methods, and quality 

procedures. This publication was prepared with the practicing transportation specialist in mind and with the 

benefit of extensive industry review.

17. Key Words compaction, deep and mass soil mixing,

dynamic column supported embankments, grouting, 

lightweight fills, pavement subgrade stabilization, 

reinforced soil structures, stone columns, vertical 

drains, vibro-compaction

18. Distribution Statement

No restrictions.

19. Security Classification (of this report)

Unclassified

20. Security Classification (of this page)

Unclassified

21. No. of Pages

386

22. Price

Form DOT F 1700.7 Reproduction of completed page authorized
* SI is the symbol for the International System of Units. Appropriate rounding should be made to comply with Section 4 of ASTM E380.

(Revised March 2003)

SI* (MODERN METRIC) CONVERSION FACTORS 

 
I-i 

PREFACE

One of the major tasks within geotechnical engineering is to design, implement and evaluate 

ground modification schemes for infrastructure projects. During the last forty years 

significant new technologies and methods have been developed and implemented to assist the 

geotechnical specialist in providing cost-effective solutions for construction on marginal or 

difficult sites.

The impetus for ground modification has been both the increasing need to use marginal sites 

for new construction purposes and to mitigate risk of failure or of poor performance. During 

the past several decades, ground modification has come of age and reached a high level of 

acceptance in the geotechnical community. Its use is now routinely considered on most 

projects where poor or unstable soils are encountered. From the geotechnical engineer's point 

of view, ground modification means the modification of one or more of the relevant design 

engineering properties (e.g., increase in soil shear strength, reduction of soil compressibility, 

and reduction of soil permeability) – or the transfer of load to more competent support layers. 

From the contractor’s point of view, ground modification may mean a reduction in 

construction time and/or a reduction in construction costs. Both points of view are valid 

reasons to consider the use of ground modification techniques and are often mutually 

inclusive. 

Herein, ground modification is defined as the alteration of site foundation conditions or 

project earth structures to provide better performance under design and/or operational 

loading conditions. Ground modification objectives can be achieved using a large variety of 

geotechnical construction methods or technologies that alter and improve poor ground 

conditions where traditional over-excavation and replacement is not feasible for 

environmental, technical or economic reasons. Ground modification has one or more of the 

following primary functions, to: 

• increase shear strength and bearing resistance,

• increase density,

• decrease permeability,

• control deformations (settlement, heave, distortions),

• improve drainage,

• accelerate consolidation,

• decrease imposed loads,

• provide lateral stability,
I-ii

• increase resistance to liquefaction, and/or

• transfer embankment loads to more competent subsurface layers.

The purpose of GEC 13 is to introduce available ground modification methods and 

applications to design generalists (i.e., project planners, roadway designers, consultant 

reviewers, etc.), design specialists (i.e., geotechnical, structural, pavement, etc.), construction 

engineers, specification writers, and contracting specialists involved with projects having 

problematic site conditions. This publication was prepared with practicing transportation 

specialists and generalists in mind. 

The introductory chapter provides a description, history, functions, and categories of ground 

modification. Additionally, the role of ground modification in addressing project risks and 

constraints and risk mitigation, and contracting mechanisms and their impact on selection of 

ground modification technologies are described. The chapter also includes description of the 

web-based GeoTechTools (http://www.geotechtools.org) technology selection guidance 

system, and its use for the initial screening process of developing a short-list of technologies 

applicable to a given project. The GeoTechTools geotechnology catalog, of over 50 

technologies, and the engineering tools provided for each technology are described. A 

discussion of final project-specific technology selection that extends beyond the initial 

screening that can be developed within GeoTechTools is included in Chapter 1. Through 

incorporation of technology and project specific factors, a 12-step process is presented that 

leads to selection of a preferred, specific technology for a given project. 

The introductory chapter is followed by stand-alone technical category chapters. Each 

category chapter includes a broad introduction to the technical category including typical 

applications, a listing of common technologies used in the United States, and summaries for 

specific technologies in the category. Each technology summary includes: description; 

advantages and limitations; applicability; complementary technologies; construction methods 

and materials; design guidance; quality assurance methods; costs; specifications; and 

reference list. Each technical category and the technology summaries therein reflect current 

practice in design, construction, contracting methods, and quality assurance procedures. 

Transportation focused case histories are included for select technologies. 

This 2016 GEC 13 reference manual on Ground Modification Methods is an update to the 

2006 FHWA-NHI-06-019/020 Ground Improvement Methods reference manual. Lead author 

of the 2006 manual was Victor Elias, PE, and is his last major work. Mr. Elias had a 

distinguished professional career and provided significant contributions to the design and 

construction of safe, cost-effective geotechnical works in transportation works. He had been 

the Principal Investigator for several major research and/or implementation projects focused 
I-iii

on durability of soil reinforcement materials, design guidance and specifications for retaining 

walls foundations and, and ground improvement methods. 
I-iv

Chapters and technology categories contained in this Volume I of the FHWA Ground 

Modification reference manual set:

Chapter 1 Introduction to Ground Modification Technologies

Chapter 2 Vertical Drains and Accelerated Consolidation

Chapter 3 Lightweight Fills

Chapter 4 Deep Compaction

Chapter 5 Aggregate Columns 

Chapters and technology categories contained in the companion Volume II of the FHWA 

Ground Modification reference manual set:

Chapter 6 Column-Supported Embankments

Chapter 7 Deep Mixing and Mass Mixing

Chapter 8 Grouting 

Chapter 9 Pavement Support Stabilization Technologies 

Chapter 10 Reinforced Soil Structures
Chapter 1 

INTRODUCTION TO 

GROUND 

MODIFICATION 

TECHNOLOGIES 

1-i 

CONTENTS

1.0 DESCRIPTION AND HISTORY ........................................................................... 1-1

1.1 Description ............................................................................................................ 1-1

1.2 Historical Overview ............................................................................................. 1-2

1.3 Focus and Scope ................................................................................................... 1-4

2.0 BASIC FUNCTIONS OF GROUND MODIFICATION ..................................... 1-6

2.1 Typical Functions and Typical Applications ..................................................... 1-6

2.1.1 Functions ............................................................................................................ 1-6

2.1.2 Applications ....................................................................................................... 1-8

2.2 Applicability Limits ............................................................................................. 1-9

2.3 Feasibility Evaluations....................................................................................... 1-11

2.3.1 Project Constraints ........................................................................................... 1-12

2.3.2 Geotechnical Performance Criteria/Indicators ................................................. 1-12

2.3.3 Environmental and Space Considerations ....................................................... 1-13

2.3.4 Site Conditions ................................................................................................. 1-13

2.4 Limitations .......................................................................................................... 1-13

2.5 Alternative Solutions ......................................................................................... 1-14

3.0 TECHNOLOGY CLASSIFICATION AND ELEMENTS ................................ 1-15

3.1 Classification by Function ................................................................................. 1-15

3.2 Elements of Construction .................................................................................. 1-17

4.0 CONSTRAINTS AND RISK MANAGEMENT ................................................. 1-20

4.1 Types of Constraints .......................................................................................... 1-20

4.1.1 General ............................................................................................................. 1-20

4.1.2 Geotechnical .................................................................................................... 1-20
1-ii

4.2 Types of Risks ..................................................................................................... 1-21

4.3 Risk Management Process ................................................................................ 1-24

5.0 CONTRACTING ALTERNATIVES, SPECIFICATIONS, AND 

QUALITY ASSURANCE ................................................................................................. 1-27

5.1 Contracting Mechanisms Used in Project Delivery ........................................ 1-27

5.2 Specification Development ................................................................................ 1-29

5.3 Quality Assurance .............................................................................................. 1-30

5.4 Construction Control and Instrumentation Monitoring ................................ 1-31

5.5 Considerations of QA in Ground Modification .............................................. 1-33

6.0 COST ANALYSIS ................................................................................................. 1-35

6.1 General Cost Components ................................................................................ 1-35

6.2 Factors That Influence Ground Modification Costs....................................... 1-35

6.3 Preliminary Cost Estimation ............................................................................ 1-37

7.0 GEOTECHTOOLS................................................................................................ 1-38

7.1 Background, Development, Audience, and Use .............................................. 1-38

7.2 Catalog of Technologies ..................................................................................... 1-39

7.3 Technology Selection Guidance ........................................................................ 1-40

7.4 Products/Tools .................................................................................................... 1-41

7.5 Summary ............................................................................................................. 1-42

8.0 PROJECT EVALUATION AND GEOTECHNOLOGY SELECTION .......... 1-43

8.1 Introduction ........................................................................................................ 1-43

8.2 Process to Identify Potential Poor Ground Conditions and Need for 

Ground Modification ..................................................................................................... 1-43

8.2.1 Step 1: Identify Potential Poor Ground Conditions and Need for Ground 

Modification ................................................................................................................. 1-44

8.2.2 Step 2: Identify or Establish Performance Requirements ................................ 1-44

8.2.3 Step 3: Identify and Assess General Site Conditions....................................... 1-44

8.2.4 Step 4: Assessment of Subsurface Conditions ................................................. 1-44
1-iii

8.2.5 Step 5: Develop a Short-List of Geotechnologies Applicable to Site 

Conditions .................................................................................................................... 1-45

8.2.6 Step 6: Consider Project Constraints ............................................................... 1-45

8.2.7 Step 7: Consider Project Risks ......................................................................... 1-46

8.2.8 Step 8: Prepare Preliminary Designs ............................................................... 1-46

8.2.9 Step 9: Identify Alternative Solutions (Bridge, Re-route, Deep 

Foundations, etc.) ......................................................................................................... 1-46

8.2.10 Step 10: Evaluate Project Requirements, Constraints, and Risks Against 

Factors Affecting Geotechnology Selection ................................................................ 1-47

8.2.11 Step 11: Compare Short-List of Geotechnology Alternatives with 

Geotechnology Selection Factors................................................................................. 1-48

8.2.12 Step 12: Select a Preferred Geotechnology ..................................................... 1-48

8.3 Additional Considerations – Detailed Subsurface Investigation, Design, 

and Cost Estimate .......................................................................................................... 1-49

8.4 Geotechnology Selection Example .................................................................... 1-49

8.5 Combination of Geotechnologies ...................................................................... 1-51

9.0 REFERENCES ....................................................................................................... 1-53
1-iv

LIST OF FIGURES

Figure 1-1. Elements of construction. .................................................................................. 1-18

Figure 1-2. SHRP 2 R09 iterative risk management process............................................... 1-25

Figure 1-3. Federal Lands Highway project development work process. ........................... 1-28

LIST OF TABLES

Table 1-1. Technical Categories and Technology Summaries .............................................. 1-5

Table 1-2. General Applicability of Technologies .............................................................. 1-10

Table 1-3. Technologies Classified by Function ................................................................. 1-15

Table 1-4. Design-Bid-Build (D-B-B) versus Design-Build (D-B) Risk Profiles ............... 1-23

Table 1-5. Devices to Monitor Geotechnical Performance ................................................. 1-33

Table 1-6. Comparative Unit Costs by Ground Modification Technology, November 

2016................................................................................................................................ 1-36

Table 1-7. Ground Modification Technology Selection Steps ............................................ 1-43

Table 1-8. Geotechnology Selection Factors ....................................................................... 1-48

Table 1-9. Sample Project Selection Matrix ........................................................................ 1-50

Table 1-10. Technology Combinations Found in Literature Review .................................. 1-52
1-1 

1.0 DESCRIPTION AND HISTORY

1.1 Description

When difficult ground conditions are encountered there are a number of alternatives that can 

be employed to achieve project objectives. These alternatives include: (1) bypassing the poor 

ground through relocation of the project to a more suitable site or through the use of a deep 

foundation; (2) removing and replacing the unsuitable soils; (3) designing the planned 

structure to accommodate the poor/marginal ground; or (4) modifying (improving) the 

existing soils, either in-place or by removal, treatment and replacement of the existing soils; 

(ASCE 1978; Mitchell 1981). Through a wide-variety of modern ground improvement and 

geoconstruction technologies, marginal sites and unsuitable in-situ soils can be improved to 

meet demanding project requirements, making the latter alternative an economically 

preferred solution in many cases. In essence, the modern builder has the option to “fix” the 

poor ground conditions and to make them suitable for the project’s needs (Munfakh and 

Wyllie 2000). A variety of terms are used to describe this “fixing the ground”: ‘soil 

improvement’, ‘ground improvement’, ‘ground treatment’, or ‘ground modification’. Charles 

(2002) notes that the process of altering the ground is ground treatment, while the purpose of 

the process is ground improvement, and the result of the process is ground modification. For 

better or worse the treatment has modified the ground’s support conditions. 

Herein, ground modification is defined as the alteration of site foundation conditions or 

project earth structures to provide better performance under design and/or operational 

loading conditions (USACE 1999). Ground modification objectives can be achieved using a 

large variety of geotechnical construction methods or technologies that alter and improve 

poor ground conditions where replacement is not feasible for environmental, technical or 

economic reasons. Ground modification has one or more of the following primary functions: 

• Increase shear strength and bearing resistance

• Increase density 

• Decrease permeability

• Control deformations (settlement, heave, distortions) 

• Increase drainage

• Accelerate consolidation

• Decrease imposed loads

• Provide lateral stability

• Increase resistance to liquefaction
1-2 

• Transfer embankment loads to more competent subsurface layers

There are over four million miles of highways in the United States, including over 164,000 

miles on the National Highway System that form the backbone of the public road network 

(Richard Weingroff, personal communication). The American Society of Civil Engineers 

2013 Report Card for America’s Infrastructure noted that 32% of America’s major roads are 

in poor or mediocre condition, costing U.S. motorists who are traveling on deficient 

pavement and bridges $67 billion a year (ASCE 2013). Many miles of these roadways need

to be reconstructed, rehabilitated, or upgraded in areas of difficult ground conditions. These 

efforts increasingly must be done under severe time constraints while minimizing disruption 

to existing traffic, and with the goal of producing long-lived facilities. The selected course of 

action must often avoid destruction of or harmful effects to existing, adjacent pavement 

systems or structural facilities such as bridges, retaining walls, and embankments that still 

have remaining useful life. The selection of an appropriate geoconstruction technology to use 

in a transportation project is a complex undertaking that depends upon integration of 

available knowledge and a number of problem-specific and site-specific constraints and 

requirements. 

1.2 Historical Overview

An early report on soil improvement was that of the ASCE Committee on Placement and 

Improvement of Soils in 1978 in which it was noted: 

Soil, nature’s most abundant construction material, has been used by man for his 

engineering works since prior to the beginnings of recorded history. Virtually all 

construction is done on, in, or with soil, but not always are the natural soil conditions 

adequate to accomplish the work at hand. The basic concepts of soil improvement—

densification, cementation, reinforcement, drainage, drying, and heating—were 

developed hundreds or thousands of years ago and remain unchanged today (ASCE 

1978). 

While roadway foundations and fortifications have been constructed out of soils for 

centuries, the development of machines greatly increased the efficiency of such construction. 

It was the invention of machines in the Industrial Revolution and the 19th century that 

allowed very significant improvements in the quality and quantity of work undertaken. 

Drainage methods to improve road performance on poor ground conditions, including 

crowning transverse grades to drain water away from roadbeds and the use of clean, free 

draining aggregate to permit the free drainage of water began to be used. In the 20th century, 

the development of soil mechanics as a discipline provided the basis for understanding soil 

behavior. This development of improved understanding combined with the development of 
1-3 

equipment led to a number of improvement techniques including densification, soil mixing, 

grouting, and reinforcing methods to mitigate problem ground conditions (ASCE 1978). 

Many of the ground modification techniques originated in Europe and the Far East and were 

subsequently brought to the United States. Often contractors led the development of the 

techniques as they wrestled with poor ground conditions and made improvements in 

equipment to make their efforts more efficient and cost-effective. The contractor led 

development often meant that the techniques were experience-based and sometimes 

proprietary. FHWA (DiMillio 1999) noted that throughout their short history, commercial 

and technological innovations [by contractors] in ground modification technologies have 

almost always preceded research studies of fundamental performance and the development of 

engineering guidelines. 

FHWA (DiMillio 1999), in summarizing the results of Demonstration Project No. 116, 

Ground Improvement Methods (Elias et al. 1999), noted that ground improvement techniques 

were found to provide benefits in the following five major areas:

• Utilization of less costly foundation systems 

• Reduction in right-of-way acquisitions

• Less environmental disturbance 

• Reduction in construction time

• Improved traffic control through construction zones 

The impetus for ground modification has been both the increasing need to use marginal sites 

for new construction purposes and to mitigate risk of failure or potential poor performance. 

During the past several decades, ground modification has come of age and reached a high 

level of acceptance in the geotechnical community. Its use is now routinely considered on 

most projects where poor or unstable soils are encountered, especially on sites underlain by 

suspect or uncontrolled fills. From the geotechnical design engineer's point of view, ground 

modification means the modification of the relevant engineering property (e.g., increase in 

soil shear strength, reduction of soil compressibility, and reduction of soil permeability) – or 

the transfer of load to more competent support layers. From the contractor’s point of view, 

ground modification may mean a reduction in construction time and/or a reduction in 

construction costs. Both points of view are valid reasons to consider the use of ground 

modification techniques and are often mutually inclusive. 
1-4 

1.3 Focus and Scope

One of the major tasks within geotechnical engineering is to design, implement and evaluate 

ground modification schemes for infrastructure projects. During the last forty years 

significant new technologies and methods have been developed and implemented to assist the 

geotechnical specialist in providing cost-effective solutions for construction on marginal or 

difficult sites. The purpose of this manual is to introduce available ground modification 

methods and applications to design generalists (i.e., project planners, roadway designers, 

consultant reviewers, etc.), design specialists (i.e., geotechnical, structural, and pavement), 

construction engineers, and specification and contracting specialists involved with projects 

having problematic site conditions. This publication was prepared with the practicing 

transportation specialist in mind and with the benefit of extensive industry review. 

This chapter provides a description, history, functions, and categories of ground 

modification. Additionally, the role of ground modification in addressing project risks and 

constraints and risk mitigation, and contracting mechanisms and their impact on selection of 

ground modification technologies are described. Typical unit costs are provided. This chapter 

also includes description of the web-based GeoTechTools (http://www.geotechtools.org) 

technology selection guidance system and geotechnology catalog, and its use for the initial 

screening process of developing a short-list of technologies applicable to projects. A 

discussion of final project-specific technology selection that extends beyond the initial 

screening that can be developed within GeoTechTools is included. Through incorporation of 

technology and project specific factors, a 12-step process is described that leads to a final 

selection of a preferred project-specific technology. 

The introductory chapter is followed by stand-alone technical category chapters. Each 

category chapter includes a broad introduction to the technical category including typical 

applications, a listing of common technologies used in the U.S., and summaries for specific 

technologies in the category. Each technology summary includes: description; advantages 

and limitations; applicability; complementary technologies; construction methods and 

materials; photographs; design guidance; quality assurance methods; costs; specifications; 

and reference list. Each technical category and the technology summaries therein reflect 

current practice in design, construction, contracting methods, and quality procedures. Current 

transportation case histories are included for selected technologies. The nine technical 

categories and the individual technologies included in each category are shown in Table 1-1. 
1-5 

Table 1-1. Technical Categories and Technology Summaries

Chapter Category Technologies

1 Introduction • All

2 Vertical Drains and 

Accelerated Consolidation

• Prefabricated Vertical Drains (PVDs), with 

and without fill preloading

3 Lightweight Fills

• Granular Fills: Wood Fiber; Blast Furnace 

Slag; Fly Ash; Boiler Slag; Expanded Shale, 

Clay and Slate, Shredded Tires

• Compressive Strength Fills: Geofoam; 

Foamed Concrete

4 Deep Compaction • Deep Dynamic Compaction

• Vibro-Compaction 

5 Aggregate Columns • Stone Columns 

• Rammed Aggregate Piers

6 Column Supported 

Embankments 

• Column Supported Embankments 

• Reinforced Soil Load Transfer Platform 

• Columns: Non-compressible 

• Columns: Compressible

7 Soil Mixing • Deep Mixing

• Mass Mixing

8 Grouting 

• Chemical (Permeation) Grouting

• Compaction Grouting 

• Bulk Void Filling

• Slabjacking

• Jet Grouting 

• Rock Fissure Grouting 

9 Pavement Support 

Stabilization

• Mechanical Stabilization 

• Chemical Stabilization 

• Moisture Control 

10 Reinforced Soil Structures 

• Reinforced Embankments 

• Reinforced Soil Walls 

• Reinforced Soil Slopes 

• Soil Nailing

Considerations that are essential in the selection, design, construction, validation, and 

monitoring of technologies on any successful ground modification project are listed and 

discussed in the following sections. 
1-6 

2.0 BASIC FUNCTIONS OF GROUND MODIFICATION 

2.1 Typical Functions and Typical Applications 

Many ground modification and geoconstruction technologies are available to improve the 

properties of soils, and the methods can be categorized in a number of ways. Mitchell (1981) 

provided the following categories in his State-of-the-Art paper: compaction, with emphasis 

on in-situ deep densification of cohesionless soils; consolidation by preloading and/or 

vertical drains and electro-osmosis; grouting; soil stabilization using admixtures and by ion 

exchange; thermal stabilization; and reinforcement of soil. More recently, Munfakh and 

Wyllie (2000) suggested eight main categories: Densification, Consolidation, Weight 

Reduction, Reinforcement, Chemical Treatment, Thermal Stabilization, Electrotreatment, 

and Biotechnical stabilization. The current International Society for Soil Mechanics and 

Geotechnical Engineering (ISSMGE) technical committee on ground improvement (TC 211, 

formerly TC17) lists five categories: improvement without admixtures in non-cohesive soils; 

improvement without admixtures in cohesive soils; improvement with admixtures or 

inclusions; improvement with grouting type admixtures; and earth reinforcement (Chu et al.

2009). Herein ground modification technologies are categorized by the functions introduced 

in Section 1.1 and are discussed in more detail below. 

Within transportation infrastructure a number of applications can benefit from the use of 

ground modification. The applications include bridge support, embankments, embankment 

widening, pavement support, and construction working platforms. Each of these applications 

is discussed in more detail below. 

2.1.1 Functions

2.1.1.1 Increase Shear Strength and Bearing Resistance

Here the function of the ground modification is to increase the soil’s strength, which in turn 

increases the bearing resistance for foundations and embankments. Increases in soil strength 

and bearing resistance can be accomplished by densifying loose cohesionless soils, 

consolidating soft clay soils, or the addition of cementing agents to the soil. 

2.1.1.2 Increase density

This function generally applies to the densification of loose sands through technologies that 

add energy to the soil through a vibration or dynamic process. The imparted energy changes 

the loose sand into a more dense state. The more dense soil has increased strength and 

bearing resistance and increased resistance to liquefaction. In cohesive soils increasing the 

density is accomplished through consolidation processes that remove water from the void 
1-7 

spaces thus reducing the amount of settlement that will occur when loads are applied to the 

soil. 

2.1.1.3 Decrease Permeability

Here the function is to decrease the amount of water flowing through the soil. This can be 

accomplished by increasing the density of the soil or through the addition of grouts or 

binders that make the soil relatively impermeable or fill fissures.

2.1.1.4 Control Deformation

Controlling deformation includes reducing total settlement, heave, and distortion caused by 

differential settlement. Methods include those that densify or consolidate the foundations 

soils, or strengthen the soils through grouts or binders to control deformations. Deformation 

control can also be accomplished through the use of columns to transfer loads to more 

competent materials. Expansive soil heave can be treated using binders that mitigate the 

effects of water. 

2.1.1.5 Increase Drainage 

Increasing drainage allows for more efficient removal of water from foundation soils, 

subgrades, and base and subbase courses. Almost all soils are improved in their strength and 

stiffness properties with reduction in water. Increased drainage can also be used to reduce 

liquefaction susceptibility of cohesionless materials. 

2.1.1.6 Accelerate Consolidation 

Accelerating consolidation reduces the time involved for settlement in foundation soils to 

occur. Consolidation can be accelerated by reducing the drainage path length for cohesive 

soils in combination with embankment loading or fill preloading. This can be accomplished 

through the use of prefabricated vertical drains or other columns that allow water an easier 

flow path. 

2.1.1.7 Decrease Imposed Loads 

Decreasing imposed loads through the use of lighter weight fill materials reduces loads on 

weak soils reducing settlement and stability issues. 
1-8 

2.1.1.8 Provide Lateral Stability

Change in grade requirements can be accomplished by use of a number of earth retaining 

systems that provide lateral support and stability to site soils, in both cut and fill situations. 

Such support can be provided for both vertical and sloping cases. 

2.1.1.9 Increase Resistance to Liquefaction

The resistance of cohesionless soils to liquefaction can be accomplished by densifying the 

soils through vibratory or dynamic methods that increase the density of the cohesionless 

materials. Other means of increasing resistance to liquefaction include the addition of grouts 

and binders to the soil matrix, increased drainage of the soils, and isolation of the potentially 

liquefiable soils.

2.1.1.10 Transfer Vertical Loads to More Competent Soil or Rock Layers

Here, vertical loads – typically embankments or fill retaining structures – are transferred 

through loose or weak soils by columns that transfer the embankment loads to more 

competent layers. This technique helps control settlement, particularly differential settlement, 

and stability of the highway feature on the unstable soils. 

2.1.2 Applications 

Ground modification and geoconstruction technologies can be used in a number of highway 

and transportation infrastructure applications. Common applications are discussed below. 

2.1.2.1 Structure Support 

Bridges are used to cross water and also to provide grade separation pathways over other 

highways, railroads and other infrastructure. At water crossings, the bridge is often situated 

in an alluvial environment in which nature has left a variety of soil deposits ranging from soft 

clays to loose sands to dense sands. The deposits are often interlayered and non-uniform. 

While deep foundations are often used to support bridge abutments and piers, ground 

modification technologies are alternatives that can be used to improve the site conditions, 

allowing less expensive shallow or intermediate foundations to be used. Retaining structures 

can also be used for abutments and approach embankments for bridges for both water 

crossings and grade changes. 

2.1.2.2 Embankments 

Embankments are used to support highways. They are used to change the grade along an 

alignment to provide better vertical position for the roadway and as approaches to bridges. 
1-9 

Embankments are often constructed across unstable soils. Ground modification technologies 

can be used to improve the unstable foundation soils to reduce settlement and stability issues, 

while avoiding excavation and disposal of the unsuitable soil. 

2.1.2.3 Embankment Widening 

The need to increase capacity of roadways often means adding lanes of traffic to existing 

roads. In locations where the existing roadway is constructed on compressible unstable soils, 

widening the road by adding embankment can lead to differential settlement between the 

existing and the new embankments, global instability, etc. Ground modification can stabilize 

the compressible or unstable soils, reducing the potential for the unwanted movements. This 

can be accomplished using lightweight fill materials, column supported embankments, and 

methods that increase the bearing resistance of the underlying soils. 

2.1.2.4 Pavement Support 

The pavement section is supported by the subgrade soil, which in some cases is poor, 

requiring very thick structural sections, and may not even support construction equipment. 

Support for the pavement section can be increased in several ways including stabilizing the 

subgrade, and using alternative or recycled materials. The subgrade and base layers can be 

improved through mechanical and chemical means to improve strength, and also by drainage 

efforts to reduce the adverse effects of water.

2.1.2.5 Construction Working Platform

Construction platforms are almost always needed to support ground modification equipment 

on poor soils that are being stabilized. They are also often needed for temporary roadways to 

allow construction to proceed and for storage of equipment and materials during 

construction. The means of support for working platforms is similar to that for pavement 

support, but is considered temporary in nature. 

2.2 Applicability Limits

All ground modification technologies have limits on their applicability. Limitations may be 

defined as soil type applicability, depth of treatment, etc. General applicability of the 

technologies covered within this manual is summarized in Table 1-2, by technology category. 

The advantages and potential disadvantages of each technology are listed and discussed 

under their respective category chapter. 
1-10 

Table 1-2. General Applicability of Technologies 

Category Technologies Applicability

Vertical Drains 

and Accelerated 

Consolidation 

PVDs, with and without fill 

preloading 

Compressible clays, saturated low 

strength clays

Lightweight Fills Compressive Strength Fills:

Geofoam; Foamed Concrete

Broad applicability; no geologic or 

geometric limitations 

Lightweight Fills 

Granular Fills: Wood Fiber; 

Blast Furnace Slag; Fly 

Ash; Boiler Slag; Expanded 

Shale, Clay, and Slate; Tire 

Shreds 

Broad applicability; no geologic or 

geometric limitations

Deep Compaction Deep Dynamic Compaction

Loose pervious and semi-pervious 

soils with fines contents less than 

15%, materials containing large 

voids, spoils and waste areas 

Deep Compaction Vibro-Compaction 

Cohesionless soils, clean sands with 

less than 15% silts and/or less than 

2% clay 

Aggregate 

Columns Stone Columns Clays, silts, loose silty sands, and uncompacted fill

Aggregate 

Columns Rammed Aggregate Piers Clays, silts, loose silty sands, uncompacted fill 

Column 

Supported 

Embankments 

Column Supported 

Embankments 

Soft compressible clay, peats, and

organic soils where settlement and 

global stability are concerns

Column 

Supported 

Embankments 

Reinforced Soil Load 

Transfer Platform

Soft compressible clay, peats, and

organic soils where settlement and 

global stability are concerns

Column 

Supported 

Embankments 

Columns: Non-compressible 

All soil types, in particular weak 

soils that cannot support surface 

loads 

Column 

Supported 

Embankments 

Columns: Compressible All soil types except very soft soils 

low undrained shear strength 

Soil Mixing Deep Mixing

Suitable in large range of soils, ones 

that can be stabilized with cement, 

lime, slag, or other binders 

Soil Mixing Mass Mixing Peat, soft clay, dredged soil, soft silt, 

sludges, contaminated soils 

Grouting Chemical (Permeation) 

Grouting 

Wide range of soil types including 

weakly cemented rock-fill materials
1-11 

Category Technologies Applicability

Grouting Compaction Grouting 

Cohesionless granular soils, 

collapsible soils, and unsaturated fine 

grained soils; may be used to fill 

voids in sinkholes or abandoned 

mine shafts; can arrest settlement 

under a structure and lift foundations 

that have settled

Grouting Jet Grouting Wide range of soil types and 

groundwater conditions

Grouting Rock Fissure Grouting Structural stability and groundwater 

control 

Grouting Bulk Void Filling and 

Slabjacking

All soil types were voids develop 

under pavements 

Pavement Support 

Stabilization Mechanical Stabilization 

Weak subgrades, loose sands, and to 

stabilize thin aggregate layers on 

subgrades with CBR<8 

Pavement Support 

Stabilization Chemical Stabilization 

Portland cement and lime: high 

plasticity clays

Fly ash: soils with little or no plastic 

fines

Asphalt: silty, sandy and granular 

soils

Cement Kiln Dust: expansive soils 

Pavement Support 

Stabilization 

Moisture Control All soil types

Reinforced Soil 

Structures

Reinforced Embankments 

Over Soft Soils 

Soft soil foundations, with no 

limitation on depth of soft soils

Reinforced Soil 

Structures Reinforced Soil Walls

Well suited in fill embankments, 

steep-sided terrain, ground subject to 

soil instability and where foundations 

soils are poor 

Reinforced Soil 

Structures Reinforced Soil Slopes Can be constructed over any firm foundation 

Reinforced Soil 

Structures Soil Nail Walls

Dense to very dense granular soils 

with apparent cohesion, weathered 

rock, stiff to hard fine-grained soils, 

engineered fill, residual soils, glacial 

till

2.3 Feasibility Evaluations

The feasibility of a ground modification method for a particular project need depends upon 

the function(s) of the modification and the method(s) selected to carry out the function. 
1-12 

Feasibility evaluation includes the identification and evaluation of: technical issues, project 

development/delivery methods, performance criteria and quality assurance procedures, and 

non-technical issues that affect the utilization of ground improvement and geoconstruction 

technologies. A generalized summary of the evaluation process for use with the technologies 

in this reference manual can be summarized as:

1. Identify project conditions which could require ground modification or 

geoconstruction technologies, such as projects that encounter: 

a. Poor ground conditions which will not provide adequate support for the 

transportation related structure. Poor ground conditions are typically 

characterized by soft or loose foundation soils, which, under load, would 

cause long-term vertical and/or lateral deformations, or cause construction or 

post-construction instability.

b. Project constraints which require retaining walls or steep slopes.

c. Pavement foundations which require improvement. 

d. Need for of a working platform or access road. 

2. Identify or establish performance requirements. 

3. Identify and assess any time, space or environmental constraints. 

4. Assess the site conditions. 

5. Assess project constraints. 

6. Identify limitations on the use of ground modification technologies. 

7. Consider alternatives to the use of ground modification technologies. 

These items are discussed in more detail below. 

2.3.1 Project Constraints 

Project constraints are generally such items as the project schedule and time, phasing 

requirements; budget and cost; project conditions such as right-of-way limits, geometry, 

scale, utilities and sequence; traffic flow/interruption and congestion; weather; and 

environmental factors. Schedule acceleration is of key importance on almost all projects 

today. Therefore speed of ground modification construction is often a key selection factor. 

2.3.2 Geotechnical Performance Criteria/Indicators

For each project, performance requirements need to be established. These can be thought of 

as the operational criteria for the facility. The performance criteria/indicators are typically 
1-13 

defined in terms of geotechnical performance criteria such as stability requirements, 

allowable long-term deformations (total or differential settlement), rate of settlement, 

seepage (quantity and flow rate), and durability (service life) and maintenance activities. 

These criteria establish the level of improvement required in terms of soil properties such as 

strength, density, modulus, compressibility, and hydraulic conductivity. 

2.3.3 Environmental and Space Considerations

Environmental and space constraints must be identified and assessed. Environmental 

constraints may include the disposal of spoils from the particular ground modification 

technology, the disposal of waste materials encountered on the site, protection of the site 

from erosion, protection of surface and ground waters from pollution, and the effects of 

construction vibrations, noise and dust. Space constraints typically refer to site accessibility, 

sufficient area for construction equipment to operate safety, overhead clearance, and adjacent 

structures and utilities.

2.3.4 Site Conditions

Site conditions, first and foremost, means assessment of the subsurface conditions. The level 

of detail regarding the assessment of subsurface conditions will vary significantly across the 

wide range of transportation related projects and the type of ground modification selected. 

Regardless of the project type, the soils which will affect the performance requirements must 

be identified and the necessary engineering properties established to perform a preliminary 

design for the project. At a minimum, the type, depth, and extent of needed treatment must be 

determined, as well as the location of the groundwater table. For sites with poor ground 

conditions, it is also valuable to have at least a preliminary assessment of the shear strength, 

compressibility, and organic content of the identified poor soils. Additionally, assessment of 

subsurface obstructions in terms of cobbles, boulders, or construction debris, water bearing 

sands, organic layers, and very stiff surface deposits can affect the selection of appropriate 

technologies. The availability of materials for construction such as sand, gravel, and water 

are also important site considerations.

2.4 Limitations

Limitations that can affect the use of ground modification technologies include those that are 

general in nature, those that are related to the site, and those that are technology dependent.

A number of non-technical and project specific parameters can limit the use of ground 

modification technologies. General limitations include items such as lack of knowledge about 

technologies; lack of organizational structure and policies to encourage use of innovations; 

absence of simple, comprehensive, reliable analysis and design procedures; lack of effective 
1-14 

quality assurance procedures; undefined established engineering parameters and/or 

performance criteria; long-term performance uncertainty; and availability of equipment, 

experienced personnel, and appropriate materials. These limitations can be overcome in a 

number of ways, including promotion of technologies, collaboration, education efforts, 

development of agency ground modification technology champions, documentation of 

agency case histories, and hosting demonstration projects. 

Limitations related to the site include a lack of site characterization information; 

environmental constraints; obstructions, both above the ground and below the ground; 

development within the immediate vicinity; unusual construction loads and vibrations; and 

environmental conditions. These limitations will vary for each site and for specific 

technologies and need to be assessed on a project-specific basis.

A general, potential limitation on the use of ground modification methods is that most 

techniques have been traditionally designed using allowable stress design (ASD) rather than 

the recently implemented load and resistance factor design (LRFD). Currently, most ground 

modification methods have not been calibrated as part of the LRFD conversion process that 

was applied to structural foundations and selected (i.e., MSE walls, soil nail walls) earth 

retention methods. Many methods do not lend themselves to an LRFD approach. 

Additionally, some ground modification methods are proprietary and/or have black box 

design methods. 

In addition to general and site limitations, each technology has limitations on its use such as 

headroom requirements, depth limitations, design and quality assurance requirements, time to 

construct and to be effective, and so on. The limitations for specific technologies are 

discussed in more detail in the technology summaries. 

2.5 Alternative Solutions 

The use of a ground modification technology is not the only feasible means of addressing 

poor ground conditions and alternative solutions should always be considered. Alternatives 

include bypassing the poor ground conditions through relocation of the project to a more 

suitable site or through the use of a deep foundation; removing and replacing the unsuitable 

soils; designing the planned structure to accommodate the poor/marginal ground conditions; 

or using a structural solution such as a bridge. 
1-15 

3.0 TECHNOLOGY CLASSIFICATION AND ELEMENTS

3.1 Classification by Function

As noted in Section 1.1, ground modification solutions can provide a number of functions 

and ground modification technologies can be classified in a variety of ways. Herein, 

technologies are classified by the functions defined in Sections 1.1 and 2.1.1. Table 1-3 

shows the technologies included in this manual classified by function. As can be seen, there 

is considerable overlap as many technologies fit into several functional categories.

Table 1-3. Technologies Classified by Function

Function Technologies Comment

Increase shear 

strength and bearing 

resistance

• Vibro-Compaction

• Dynamic Compaction

• Compaction Grouting

• Mixing Methods

• PVDs

• Stone Columns

• Rammed Aggregate Piers

• Chemical Stabilization

• Mechanical Stabilization

Some technologies will 

work in all soil types; 

others are limited to 

cohesive or cohesionless 

soils. 

Increase density

• Vibro-Compaction

• Dynamic Compaction

• Blasting Compaction

• Compaction Grouting

• Mixing Methods

• PVDs

Some technologies will 

work on all soil types; 

others are limited to 

cohesive or to 

cohesionless soils. 

Decrease 

permeability

• Bulk-infill Grouting

• Chemical Grouting

• Jet Grouting

• Deep Mixing Methods

Type of grouting 

dependent upon soils, 

depths, geology, and 

design requirements. 

Control 

deformations 

(settlement, heave, 

distortions) 

• Columns Supported Embankments

• Reinforced Load Transfer Platforms

• Non-compressible Columns

• Mixing Methods

• Vibro-Compaction

• Dynamic Compaction

• Stone Columns

• Rammed Aggregate Piers

• Chemical Stabilization

• Mechanical Stabilization

• Encapsulation

Technologies generally 

used to bypass or isolate 

soft ground or to modify 

and improve the soft 

ground. 
1-16 

Function Technologies Comment

Increase drainage

• PVDs

• Aggregate Columns

• Geotextile Encased columns

• Electro-osmosis

• Geosynthetics in Pavement

Drainage

Generally increase 

drainage by inserting a 

drainage path within the 

soil to be drained. 

Accelerate 

consolidation 

• PVDs

• Aggregate Columns

• Geotextile Encased Columns

Acceleration due to 

decreased length of flow 

path to dissipate excess 

pore water pressure.

Decrease imposed 

loads 

• Granular Fills (Wood Fiber; Blast

Furnace Slag; Fly Ash; Boiler Slag;

Expanded Shale, Clay & Slate; Tire

Shreds)

• Compressive Strength Fills:

(Geofoam, Foamed Concrete)

Densities vary from 1 to 

90 pcf. Granular fill usage 

subject to local 

availability.

Provide lateral 

stability

• Mechanically Stabilized Earth

(MSE) Walls

• Reinforced Soil Slopes

• Soil Nailing

Techniques based on 

internal reinforcement of 

soils.

Increase resistance 

to liquefaction

• Aggregate Columns

• Deep Dynamic Compaction

• Deep Mixing

• Jet Grouting

• Vibro-Compaction

Techniques generally to 

modify and improve soil 

susceptible to 

liquefaction. Aggregate 

columns also provide 

drainage path for excess 

pore water pressure.

Transfer 

embankment loads 

to more competent 

layers

• Column Supported Embankments

• Reinforced Soil Load Transfer

Platforms

• Non-compressible Columns

• Compressible Columns

A column supported 

embankment may be 

constructed on a load 

transfer platform that is 

supported on columns. 

Within the GeoTechTools system, technologies are classified by applications for a variety of 

geotechnical solutions. Thus within GeoTechTools, lists of technologies are provided under 

the following geotechnical solutions: earthwork construction, soft ground drainage and 

consolidation, densification of cohesionless soils, construction of vertical support elements, 

embankments over soft soils, lateral earth support, cutoff walls, liquefaction mitigation, 

increased pavement performance, void filling, and sustainability. 
1-17 

3.2 Elements of Construction

Difficult, unstable ground conditions requiring ground modification solutions are typically 

characterized by soft or loose foundation soils, which under load, would cause excessive 

short- or long-term settlement or cause construction or post-construction instability, The 

technologies discussed in this manual are generally applicable to elements of construction 

over , construction over stable or stabilized soils, geotechnical pavement components, and 

working platforms. These elements are shown in Figure 1-1 and are further discussed below. 

These elements are often used in conjunction with structures such as bridges. 
1-18 

Figure 1-1. Elements of construction.
1-19 

Construction over Unstable Soils focuses on methods to support embankment and 

embankment widening on the weak or loose foundation soils, i.e., typically below-grade 

technologies. Methods include ground modification technologies that improve the unstable 

soils so that the embankments can be safely constructed without unacceptable deformation 

issues. Although the ground modification is often below-grade, some at-grade technologies 

are also applicable to this element. Applications include bridge abutment support, approach 

embankments, roadway widening, and support of structural elements. 

Construction over Stable or Stabilized Soils focuses on methods for embankment and/or 

embankment widening construction, i.e., above-grade technologies. Methods include fill 

placement and compaction procedures, reduction of embankment width/volume, fill earth 

retention systems, and slope stabilization systems. The ground modification methods 

strengthen the embankment materials, and allow for geometric constraints such as retaining 

walls, or stabilizing cut slopes. Applications include retaining systems for bridge abutments 

and to reduce right-of-way needs.

Geotechnical Pavement Components focuses on methods to improve the supporting elements 

of pavements: the subgrade, and base and subbase courses. Methods include fill placement, 

stabilization, grouting, and reinforcement technologies. Recycling/reuse of materials in the 

geotechnical pavement support sections are also included.

Working Platforms focuses on methods to provide working platforms on either stable or 

unstable subgrades. Methods include fill placement, stabilization, and reinforcement 

technologies. Recycling/reuse of materials in the geotechnical pavement support sections are 

also included. Working platforms are also applicable to Construction over Unstable Soils and 

Geotechnical Pavement Components. Working Platforms are generally assumed to be 

temporary structures. 
1-20 

4.0 CONSTRAINTS AND RISK MANAGEMENT 

Constraints are limitations or restrictions on activities and can affect project delivery and 

performance. Constraints are often project specific, and can also pose risks to the successful 

completion and performance of a project. Such risks can be identified and managed through a 

risk management process. Ground modification methods can be used to mitigate some 

specific project risks, and can be selected to work within a given project set of constraints. A 

discussion of constraints and risk management follows. 

4.1 Types of Constraints 

Constraints are often driven by factors external to the project owner. Project constraints vary 

by project and could include general or specific issues and restrictions in the following areas: 

economic, legal, environmental, social, management, time, and technical. Each of these areas 

might include a related list of subtopics which would be addressed collectively or separately. 

Herein, with the emphasis on the use and application of ground modification technologies, 

constraints are defined in two categories, general constraints and geotechnical constraints. 

The key to constraint mitigation is the early identification and clear definition of the 

constraint. 

4.1.1 General

General constraints include all those items not explicitly connected with the technical (or 

geotechnical) aspects of the project. For transportation infrastructure projects key general 

constraints include the project schedule and time; budget and cost; project conditions such as 

right-of-way limits, geometry, scale, utilities and sequence; traffic flow/interruption; weather; 

environmental; availability of agency personnel, agency organization and structure, project 

management philosophy; contracting processes; public perceptions; and liability. Some of 

these constraints influence the project during its entire life while others become important at 

some point during project development or during and following construction. In general it is 

better to identify constraints as early in a project as possible so that a proactive approach can 

be taken to address and resolve the constraint. Many constraints can be complex, for 

example, time can include the time consideration related to planning, for design, for 

procurement, and the time available for construction. Depending on the overall project 

schedule and critical milestones a delay in one project phase may affect the time of the 

overall project. 

4.1.2 Geotechnical

The focus of the geotechnical constraints are those specific to geotechnical aspects of the 

project and the ground modification technologies that might be used. Such constraints 
1-21 

include subsurface conditions, including archaeological remains, contaminated groundwater, 

obstructions, and buried utilities; proprietary products/processes; knowledge and experience 

with specific technologies; allowable movements-both vertical and lateral; stability 

requirements; noise and vibrations due to construction equipment; time for modification to be 

effective (some improvement methods occur over a defined time period); spoils 

development, volume, and disposal; environmental impacts of the technology; availability of 

qualified contractors, personnel, materials and specialty equipment; and constructability of 

the selected technology.

4.2 Types of Risks 

In engineering contexts, risk is the product of probability of an event and its consequences 

(Baecher and Christian 2003). The expectation is that the consequence is an adverse 

outcome. As applied to geotechnical construction, risk might be considered as the probability 

that an adverse event will occur and that the event will impact achievement of project 

objectives. Hence the geotechnical risks associated with a project are those in which an 

adverse geotechnical outcome occurs that has an impact on the project, whether that impact 

is simply a delay in completion or perhaps the redesign of a geotechnical component (e.g., 

foundation) to accommodate unforeseen conditions. A key geotechnical risk is the failure to 

identify potential problems and design for their consequences. Possible geotechnical risks 

may relate to the following: 

• Slope Instability – soil and rock slopes (natural and man-made)

• Deformations (vertical and lateral)

• Bearing capacity

• Liquefaction

• Subsidence (sub‐surface voids or change in subsurface stresses)

• Groundwater levels and fluctuations

• Chemically reactive ground

• Contamination

• Unforeseen ground conditions (extent, type) [or from a contractor’s point of view:

differing geotechnical site conditions]

• Problem ground conditions – expansive, collapsible, organic/peat, and dispersive

soils, fills, landfills

• Inadequate geotechnical investigation
1-22 

• Inadequate design

• Poor construction or inadequate construction monitoring 

Additionally, one could consider risks associated with material characterization (properties), 

geotechnical models, service conditions, load calculations, resistance quantification, as well 

as risks related to project development, planning, and execution. The importance of 

considering geotechnical risks is highlighted by the fact that contract disputes in highway 

works are often related to geotechnical issues.

Geotechnical risks can be accepted, managed, and mitigated (Chapman 2012). Indeed an 

entire specialty area has developed around risk management, as discussed in the next section. 

Sound geotechnical practices related to design, construction, and monitoring, such as hazard 

recognition during site investigations can substantially reduce risks. 

The magnitude of transportation infrastructure construction is very large and increasingly 

complex. A number of delivery/contracting methods are being used to construct 

transportation facilities including Design-Bid-Build, Design-Build, Public Private 

Partnership (P3), and Construction Manager-General Contractor (CMGC). Geotechnical risks 

can occur in all these types of construction/delivery contracting methods; the risks remain but 

are shifted among the participants. Geotechnical risks in Design-Build (D-B) construction 

contracts was the subject of a recent TRB Synthesis (Gransberg and Loulakis 2012) due to 

the large increase in D-B contracts in transportation work. The study looked at the 

differences in geotechnical scope risk, geotechnical cost risk, and geotechnical schedule risk 

between design-bid-build (D-B-B) and design-build (D-B) contracts. Risks in scope, cost, 

and schedule are key factors in a contractor’s successful completion of a project. Table 1-4 

summarizes the results of the study.
1-23 

Table 1-4. Design-Bid-Build (D-B-B) versus Design-Build (D-B) Risk Profiles

Contract

Type 

Risk 

Type Contractor Team Owner

D-B-B Scope 

• Warranties and Guarantees 

• Latent Defects - Workmanship 

• Competent Geotechnical Construction 

Personnel Available 

• Design Error and Omissions 

• Latent Defects - Design 

• Direct and Tacit Approval of Constructive Changes to 

Design 

D-B Scope 

• Design Errors and Omissions 

• Warranties and Guarantees 

• Latent Defects – Design and Workmanship 

• Competent Geotechnical Design Personnel 

Available

• Clear Geotechnical Scope Definition 

• Direct and Tacit Approval of Constructive Changes to 

Geotechnical Design 

• Geotechnical Design Review Comments and Directives 

• Technical Review Capability 

D-B-B Cost 

• Rework 

• Subcontractor Default 

• Market Fluctuation after Award 

• Redesign and Resultant Rework 

• Construction Contract Amount 

• Market Fluctuation During Design - Material and Labor 

D-B Cost 

• Rework 

• Redesign 

• Subcontractor Default 

• Market Fluctuation During Design - 

Material and Labor 

• Design-Build Contract Amount 

• Prompt Payment 

• Design-Builder Default 

D-B-B Schedule 

• Contract Completion 

• Date 

• Liquidated Damages 

• Timely Design Completion 

• Owner Furnished Property Delivery

D-B Schedule 

• Delivery on Approved Schedule 

• Fast-Track Geotechnical Rework 

• Liquidated Damages 

• Unrealistic Schedule 

• Timely Geotechnical Design Approvals on Fast-Track 

Project 

• Owner Furnished Property Delivery

Source: Gransberg and Loukakis 2012
1-24 

Gransberg and Loulakis (2012) make clear that the risks don’t go away, but rather 

responsibility for some risks may shift. For example, they note how the items design errors 

and omissions, latent defects in design, and market fluctuations during design for materials 

and labor move from the Owner column in D-B-B to the Design-Builder column in D-B.

4.3 Risk Management Process

The risk management process can be defined as “The systematic application of management 

policies, procedures and practices to the activities of communicating, consulting, establishing 

the context, and identifying, analyzing, evaluating, treating, monitoring and reviewing risk.” 

Risk management is viewed as a way to focus limited resources, strengthen the ability to 

efficiently manage program delivery, and improve communication and manage risk 

corporately.

The first part of the risk management process is the identification of the context of the risk, 

identifying project objectives, the criteria on which to assess the risks and who will do the 

assessment. A major portion of this activity is risk assessment in which the risks are 

identified, analyzed and assessed, and prioritized in a risk register. Risk response plans are 

then developed. A risk tracking plan is then developed which allows monitoring, evaluation, 

and adjustment as responses to risk are made and new risks are included. This process can be 

applied at many levels including at project, program and enterprise levels. Unfortunately, 

decision making for ground modification technologies has often been reactive rather than 

proactive. Use of the risk management process applied to geotechnical aspects of a project 

can help identify potential events that might affect a project and allow for proactive use of 

the technologies discussed herein. Note that communication and consultation occur at each 

and every step, starting at the very beginning with identification and engagement of 

stakeholders. 

A recent SHRP2 study on the Process for Managing Risk on Rapid Renewal Projects (R09) 

is a significant resource on risk management for all types of projects, with particular 

emphasis on geotechnical issues (Golder Associates Inc., 2011). As noted within this project 

report:

The primary objective of the risk management process, whether at the individual project 

level or for a “program” of individual projects, is to optimize project performance (e.g., 

minimize cost, minimize disruption, etc.). Problems can arise during a project that leads 

to undesirable performance. Anticipating the problems upfront can lead to management 

strategies that minimize undesirable performance. An appropriate formal risk 

management approach also needs to be efficient and defensible, as well as adequate (as 
1-25 

opposed to perfect), in achieving the objective of optimizing project performance. The 

process also must be compatible with the DOT organization and their projects. 

The risk management process can be conducted in a number of ways. The SHPR 2 R09 

process is shown in Figure 1-2 and consists of six steps which form a continuous loop, 

indicating a continual process of communication and updating. Each step is briefly 

summarized below. 

After Golder Associates Inc. 2011

Figure 1-2. SHRP 2 R09 iterative risk management process.

To start the process, the project scope, strategy and conditions must be identified and 

enumerated. The first step in the risk management process is Structuring, which is where 

base project performance is identified and includes the planned project scope, strategy, and 

key conditions and assumptions. A base project description is developed and includes all the 

relevant elements of the project. The second step is Risk Identification where the risks and 

opportunities relative to the base project performance are enumerated. The objectives of risk 

identification are to: identify, categorize, and document all risks and opportunities that could 

significantly affect the project’s base performance measures; start a risk register, which is a 

comprehensive set of non-overlapping risks and opportunities; and set the stage for 

subsequent steps in the risk-management process. The third step is Risk Assessment where 

the relative severity of the risks and opportunities are assessed so they can be prioritized for 

subsequent management. As the project develops and conditions change, the risk factors for 
1-26 

previously identified risks might change and need to be reassessed, while the factors for any 

new risks must be assessed. The updated assessments of factors describing the severity of 

each risk are maintained in the project risk register. 

The fourth step is Risk Analysis where a more quantitative determination of project 

performance is assessed using uncertainty and probability distributions. This step is 

particularly valuable on complex projects where large amounts of information must be 

processed. In this step the uncertainty in project performance measures are quantified and 

sensitivity analyses conducted to assess individual risks to the base performance 

requirements. The fifth step is Risk Management Planning. Here proactive ways to manage 

the identified risks are identified and evaluated for possible use should the risk develop. As 

the project proceeds and risks change, these plans must be reviewed and updated as 

necessary to optimize project performance. Step six, Risk Management Implementation, 

consists of implementing and monitoring the developed Risk Management Plan. This step 

insures that the periodic review and updating of all steps is done on a timely and appropriate 

basis. This step is tied back to the project scope, strategy and conditions identified at the start 

of the process. 
1-27 

5.0 CONTRACTING ALTERNATIVES, SPECIFICATIONS, AND QUALITY 

ASSURANCE 

Project planning is concerned with understanding project constraints and identifying risks. 

Constraints and risks will always be present on a project, however, the discernment of 

constraints and the management of risks can be affected by the contracting methods used. 

The sections below discuss contract mechanisms, specifications, quality assurance, and 

instrumentation monitoring and construction control methods to deal with constraints and 

risks.

5.1 Contracting Mechanisms Used in Project Delivery 

The development and construction of transportation projects involves a number of processes 

and requires a high level of coordination among many stakeholders. Several contracting 

mechanisms are available for delivery of projects; however, project delivery extends beyond 

the procurement phase and approach of a particular project. Herein, project delivery refers to 

the entire process from initial planning and programming to close-out of construction while 

contracting or contracting method refers to the roles and responsibilities for conducting 

design and construction activities. Transportation agencies play a significant role in the 

implementation of a project from its inception to its construction, with responsibilities that 

include: estimating and controlling costs; ensuring the fulfillment of environmental and 

Federal requirements; obtaining adequate financing and the overall managing of the various 

parties involved in bringing the project to a successful completion. 

While nomenclature may vary from state to state, and activities may overlap, the following 

items related to project delivery are generally contained somewhere within each state’s 

process. 

• Planning and programming 

• Conceptual studies 

• Environmental stewardship 

• Right-of-way

• Preliminary design

• Final design and PS&E (plans, specification, and estimate)

• Construction 
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Figure 1-3 from the Federal Lands Highway Project Development and Design Manual 

(PDDM) provides a generalized concept of the project development work process (FHWA 

2011). 

After FHWA 2011

Figure 1-3. Federal Lands Highway project development work process.

From this figure it can be seen that projects begin from needs studies and outside requests for 

projects. The projects are then considered and placed in a long-range transportation plan with 

programming prioritization occurring, leading to the development of project concepts 

(planning). The boxes on the left hand side of this figure collectively constitute what is 

generally called planning and programming. What follows next is project scoping which 

requires the collection of data, adherence to agency standards and criteria, and preliminary 

engineering studies. The next step is preliminary design. The environmental studies and 

documentation may occur during project scoping or preliminary design, depending upon the 

project and the transportation agency. Once preliminary design is completed and approvals 

are secured, the project can move into final design and development of plans, specifications, 

and estimates. Next follows construction. 

Traditionally, the design and construction phases of this process were conducted using the 

design-bid-build process wherein the agency (or its consultants) designed the project, the 

design was let out for bidding, with the lowest responsible bidder awarded the construction 

contract. In recent decades several other contracting mechanisms have been developed 

including Design-Build (DB), and Construction Manager-General Contractor (CMGC). Each 

of these methods has a slightly different approach to the project work process on a project 

and each has advantages and challenges. 
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In D-B, the contracting method combines two, usually separate project phases into a single 

contract, and one entity – the design-build team – works under a single contract with the 

project owner to provide design and construction services. This provides unified work flow 

from conceptualization through completion – thereby integrating the roles of designer and 

constructor. In this method, owners generally execute a single, fixed-fee contract for both 

engineering services and construction. The D-B entity may be a single firm, a consortium, 

joint venture or other organization assembled for a particular project. 

Public-Private-Partnerships (P3) is contractual agreements formed between a public agency 

and a private sector entity that allow for greater private sector participation in the delivery 

and financing of transportation projects. P3 projects also routinely include an operations 

component for a defined time period and tolls. Following the operating period the P3 project 

is returned to the original owner in a defined state of good condition. In a P3, a government 

entity contracts with a private venture to develop, construct, and operate a public project. 

Often the private entity assumes substantial financial, technical, and operational risk in the 

project. There are many different P3 structures, and the degree to which the private sector 

assumes responsibility – including financial risk – differs from one application to another. 

Additionally, different types of P3s lend themselves to the development of new facilities and 

others to the operation or expansion of existing assets. 

The CMGC process involves the project owner hiring a contractor to provide feedback 

during the design phase before the start of construction. The CMGC process involves two 

contract phases. The first contract phase, the design phase, allows the contractor to work with 

the designer and the project owner to identify risks, provide costs projections and refine the 

project schedule. Once the design phase is complete, the contractor and project owner 

negotiate on the price for the construction contract. If all parties are in agreement with costs 

then the second contract phase, the construction phase, is kicked off and construction begins. 

This method is often used to accelerate project delivery. 

As discussed in Section 4.2, these various contracting methods do not in and of themselves 

eliminate risks, the risks are simply shifted among the various entities involved. 

5.2 Specification Development 

Preparation of specifications is an integral part of the design process for the use of any 

ground modification technology. Specifications are the written instructions describing the 

work that is to be undertaken. Specifications are part of the contract documents, which also 

include the drawings, bid or proposal documents, agreement forms, and contract 

modifications. Specifications communicate to bidders prior to contract award, and to the 

selected contractor thereafter, the definitive directions, procedures, and material and 
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equipment requirements the owner considers necessary for completing the contract work. In 

this context, specifications directly affect the quality of design and construction of every 

highway product, as well as the cost of construction and maintenance (FHWA 2010 and 

Scott et al. 2014). 

Often state DOTs do not have specifications or special provisions prepared for ground 

modification technologies. Or if they do, it is only for a limited number of ground 

modification technologies that have been previously used by that agency. Hence, 

specifications will often have to be developed to use ground modification methods on a 

particular project. Guidance on specification development for particular technologies is 

covered in the category chapters for individual technologies in this manual. Guide 

specifications for specific technologies can be found in GeoTechTools system in the 

individual technology information pages (see Section 7). These guide specifications reflect a 

summary of existing specifications for technologies and, in many cases, provide a preferred 

guide specification for the specific technology that can be used in specification development 

for application to specific projects. 

5.3 Quality Assurance 

TRB Circular E-C037 (TRB 2002) defines Quality Assurance (QA) as “All those planned 

and systematic actions necessary to provide confidence that a product or facility will perform 

satisfactorily in service.” The circular goes on to state the following:

QA addresses the overall problem of obtaining the quality of a service, product, or 

facility in the most efficient, economical, and satisfactory manner possible. Within this 

broad context, QA involves continued evaluation of the activities of planning, design, 

development of plans and specifications, advertising and awarding of contracts, 

construction, and maintenance, and the interactions of these activities. 

For many years the term QC/QA was used to describe quality activities associated with 

construction where Quality Control (QC) referred to quality activities conducted by the 

contractor and Quality Assurance (QA) referred to quality activities conducted by the owner. 

More recently, the term Quality Assurance is being used as the umbrella term that includes 

the contractor’s QC activities and the acceptance functions of the owner-agency. AASHTO 

(2006), FHWA (2008), and Code of Federal Regulations (23 CFR 637) all define the core 

elements of a construction Quality Assurance Program to include the following: 

1. Contractor Quality Control (QC) 

2. Agency Acceptance

3. Agency Independent Assurance (IA)
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4. Dispute Resolution 

5. Laboratory Accreditation and Qualification

6. Personnel Qualification/Certification

All six elements are deemed necessary to have a complete and effective QA Program. A QA 

program missing any one or more of the elements is not sufficient and should not be 

construed as being “substantially compliant” with the intent of the AASHTO guidelines or 

the federal regulation. These core elements are more fully explained in AASHTO (2006) and 

FHWA (2008). The contractor is responsible for performing all the QC activities to monitor, 

assess, and adjust their production or placement processes to ensure that the final product 

will meet the specified level of quality (AASHTO 2006). Acceptance is all the factors used 

by the Agency (i.e., certification, sampling, testing, inspection) to evaluate the degree of 

compliance with contract requirements and to determine the corresponding value for a given 

product (FHWA 2008). The acceptance activity includes monitoring of the contractor’s QC 

activities. The other four elements ensure that appropriate processes, accredited laboratories, 

and qualified personnel are conducting the QC and/or QA activities. Importantly, the overall 

Construction Quality Assurance Program is unaffected by the use of different contract 

methods such as Design-Bid-Build or Design-Build. 

5.4 Construction Control and Instrumentation Monitoring 

Monitoring is used to provide quality assurance on methods and materials of construction 

and to reduce risks associated with material and property variations of the in situ soils. Often 

times, ground modification work is associated with construction over or through problematic 

ground conditions. The sampling, testing, and quantifying properties of foundation materials 

can be challenging; and quantifying spatial variation of properties across a site can be even 

more challenging. Therefore, instrumentation and monitoring during ground modification 

construction is routinely performed. 

Field observations, including quantitative measurements obtained by field instrumentation, 

provide the means by which the geotechnical engineer, in spite of inherent limitations, can 

design a project to be safe and efficient, and the contractor can execute the work with safety 

and economy (Dunnicliff 1988). The primary reason to instrument and monitor the 

construction and performance of a geotechnical feature is to help manage risk (Marr 2013). 

Historically, monitoring of geotechnical features has been performed during construction 

activities for verification of design (e.g., staged construction, test sections) and after 

construction for performance monitoring. There is much overlap between the two phases. 

Performance monitoring is undertaken primarily to give warning of an impending failure so 

that mitigation measures could be undertaken, to monitor a feature that had already moved as 
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indicated by surface cracks, or to perform research on methods to analyze and improve the 

performance of the feature. Over the past twenty years or so, the reasons for performance 

monitoring have expanded considerably. The following are the principal reasons to monitor 

geotechnical performance in today’s environment (Marr 2013): 

1. Show that performance is safe and acceptable

2. Provide a warning of an impending failure or unsafe condition

3. Reveal unknowns and aid use of the Observational Method

4. Evaluate contractor’s means and methods

5. Control construction or operations

6. Minimize damage to adjacent structures

7. Inform stakeholders

8. Satisfy regulators

9. Document performance for assessing damage

10. Improve performance and advance state-of-knowledge

Performance monitoring has long been a key tool of geotechnical engineers. During 

construction, activities 4 and 5 above are conducted to ensure that the contractor’s work is 

performed according to the contract requirements, safely, and that adjacent structures are not 

being damaged. Further, the outcome and performance of most geotechnical features depends 

on the construction sequence and the contractor’s means and methods. Increasingly project

requirements may be in the form of a performance specification where the contractor 

provides the details of design, completes the work, and demonstrates, by a specifically 

defined requirement, the end product performs as intended. Geotechnical instrumentation can 

be used to determine whether the contractor’s means and methods meet the specified 

performance requirements. The instrumentation program results, additionally, can provide 

sufficient data to show the potential of undesirable performance early enough to take 

corrective action. The instrumentation data may show why the contractor’s means and 

methods are not working which allows them to be adjusted to reduce their impact on project 

performance. Hence, instrumentation can save money by helping to reduce the consequences 

of undesirable performance. Ineffective or inefficient aspects of the contractor’s means or 

methods can also be identified from instrumentation data (Marr 2007a). 

As Marr (2007b) indicated, today performance monitoring has exploded to provide us 

capabilities to measure just about anything, anywhere, and in real time. The declining cost of 

monitoring hardware and today’s ubiquitous communications systems make real time 

monitoring a cost-effective construction tool. Table 1-5 summarizes the types of sensors that 
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are frequently used in geotechnical monitoring and can be used to monitoring ground 

modification performance. 

Table 1-5. Devices to Monitor Geotechnical Performance

Instrument Application

Observation Well Measure depth to ground water 

Piezometer Measure total head at a specific location

Earth Pressure Cell Measure total normal stress in soil

Contact Pressure 

Cell

Measure normal stress between soil and contact with more rigid 

material like rock or concrete

Load Cell Measure force in a structural member such as a strut or tieback

Settlement Plate Measure vertical deflection at a specific point

Settlement Gage Measure vertical deflection of one point relative to another 

Deformation 

Monitoring Point Measure Δx, Δy, Δz, ΔL

Flow Meter Measure flow through a collector pipe 

Flow Weir Measure depth of flow over a weir

Crack Meter Measure change in dimension between two points on opposite 

sides of a crack in 1, 2, or 3 planes 

Strain Gage Measure change in length over a known short distance 

Tilt Meter, 

Inclinometer Measure deviation from vertical as indicated by the pull of gravity 

Borehole 

Extensometer

Measure change in distance between two or more points in a 

borehole 

Geophone Measure velocity of motion in 1 to 400 Hz range 

Accelerometer Measure acceleration of motion in 1 to 4,000 Hz range 

Temperature Measure temperature at the location of the sensor

Barometer Measure atmospheric pressure 

Automated Total 

Station

Measure position of multiple targets relative to fixed targets to 

about 0.05 inches 

Global Positioning 

System

Measure position of one or more point relative to global reference 

system to about 0.05 inches 

Seismographs Measure dynamic motions resulting from shock loads such as 

blasting, pile driving and operation of heavy equipment 

Source: Marr 2007b

5.5 Considerations of QA in Ground Modification 

There is complexity in the selection and application of quality assurance methods in Ground 

Modification technologies. First, the contracting method for the work can affect the 

responsibility for the quality assurance methods. Clear communication is necessary in all 

contacting methods to ensure that appropriate QA methods are used and to ensure that the 

appropriate parties are conducting their respective responsibilities. Second, there are 
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differences in the complexity of the QA methods used for particular ground modification 

technologies. The QA procedures for some technologies are simple in nature while others are 

more complex. Knowledge of the QA methods used in particular ground modification 

technologies may affect the ultimate selection or desirability of using of a technology. 
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6.0 COST ANALYSIS

6.1 General Cost Components

The purpose of this section is to present an overall approach that may be used to develop a 

cost estimate for any ground modification project. The cost estimate method described is 

similar to that commonly used in engineering to develop a conceptual cost estimate. That is, 

sufficient detail is used to identify the cost components which may have the largest effect on 

total project cost, however specific bid quantities and potential bidders are not necessarily 

known. 

The method requires that a baseline cost for the particular ground modification technology be 

known. Over the past several decades, numerous ground modification projects have been 

completed in the United States and typical costs have varied during this time period primarily 

due to the business cycle, the experience of specialty ground modification contractors, 

advances in ground modification equipment, and the amount of competition from more 

conventional solutions. Due to the wide variety in types of ground modification technologies, 

there is no typical “average” price for “ground modification.” Prior to describing a suggested 

cost estimating method for ground modification technologies, a discussion on the key factors 

which affect ground improvement costs is presented. 

6.2 Factors That Influence Ground Modification Costs 

The primary cost items for any ground modification or geo-construction technology consist 

of costs for mobilization, materials, labor, and the equipment for the particular technology. 

Added to this must be the cost of quality assurance including instrumentation and/or load 

tests that might be conducted as part of the QA process. There are many factors which can 

affect the cost of a specific project including project type, application, geoconstruction 

technology, soil type, labor rates, utility conflicts, location, weather, competition, etc. 

Identifying and understanding how these variables impact cost can be beneficial when 

evaluating the applicability of a ground modification or geo-construction solution. 

Often the costs of the primary cost items described above are rolled into prices that are 

quoted in lineal feet or square feet of installed product. The role and influence of factors 

specific to individual technologies are discussed within the respective technical summaries 

and the technology cost documents within GeoTechTools. Typical unit costs for technologies 

are shown in Table 1-6. 
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Table 1-6. Comparative Unit Costs by Ground Modification Technology, November 

2016 

Category Technology Unit Cost

Vertical Drains and 

Accelerated Consolidation PVDs, with and without fill preloading $0.50–$4/lft

Lightweight Fills Compressive Strength Fills: Geofoam; 

Foamed Concrete $75–$150/yd3

Lightweight Fills

Granular Fills: Wood Fiber; Blast 

Furnace Slag; Fly Ash; Boiler Slag; 

Expanded Shale, Clay and Slate; Tire 

Shreds 

$3–$15/yd3

Deep Compaction Deep Dynamic Compaction $10–$30/yd2

Deep Compaction Vibro-Compaction $5–$9/lft

Aggregate Columns Stone Columns and Rammed Aggregate 

Piers $15–$60/lft 

Column Supported 

Embankments Column Supported Embankments $9/ft2 + cost of the column 

Column Supported 

Embankments Columns: Non-compressible $30–$80/lft

Column Supported 

Embankments Columns: Compressible $20–$100/lft 

Soil Mixing Deep Mixing (dry) $60–$125/lft 

Soil Mixing Mass Mixing $15–$75/yd3

Grouting Technologies Chemical Grouting $20/ft + $0.65/qt 

Grouting Technologies Compaction Grouting $75–$750/yd3

Grouting Technologies Bulk Void Filling $50–$150/yd3

Grouting Technologies Slabjacking $6.50–$9.30/ft2

Grouting Technologies Jet Grouting $250–$750/yd3

Grouting Technologies Rock Fissure Grouting $25–$80/ft2

Pavement Support 

Stabilization Technologies Mechanical Stabilization $1–$5/yd2

Pavement Support 

Stabilization Technologies Chemical Stabilization $2–$5/yd2

Pavement Support 

Stabilization Technologies Moisture Control $3–$12/lft 

Reinforced Soil Reinforced Embankments $2–$12/yd2

Reinforced Soil MSE Walls $30–$65/ft2

Reinforced Soil Reinforced Soil Slopes $3–$25/ft2

Reinforced Soil Soil Nailing $20–$50/lft 

Note that mobilization and demobilization of equipment to the project site is generally a 

lump sum item and can vary from a few hundred dollars to more than $100,000 depending on 
1-37 

the technology. The costs for the original site investigation, quality assurance testing, and 

instrumentation are not included in the technology unit prices or mobilization costs. 

6.3 Preliminary Cost Estimation 

Selection of a specific geotechnical solution should be based first on sound engineering. 

Routinely, two or more technologies may be identified which are potential technical 

solutions; when this occurs, engineers typically consider the initial cost of a solution as part 

of the selection process. It is important to note that while initial cost is a consideration when 

selecting a solution, it should not be the driving force; performance, construction time, life 

cycle costs, risks and safety should be factored into the evaluation of alternative geotechnical 

solutions. 

There are many factors which can affect cost for a specific project (i.e. soil type, labor rates, 

utility conflicts, etc.); identifying and understanding how these variables impact cost can be 

beneficial when evaluating the applicability of a geotechnical solution. 

To develop a preliminary cost estimate for the use of a ground modification technology 

information about the site conditions must be known. This information also is necessary for 

the preliminary design of the particular ground modification technique. This preliminary 

design information can then be used to develop specific quantities of materials, equipment, 

etc. that can used to compute a preliminary cost estimate. Cost estimating guidance for

specific technologies is provided in the respective technical summaries. GeoTechTools (see 

Section 7) provides Excel spreadsheets that aid in the development of preliminary cost 

estimates for all of the technologies addressed within this reference manual. 
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7.0 GEOTECHTOOLS

GeoTechTools is a web-based geoconstruction technologies guidance and selection system. 

GeoTechTools is accessible through the website at http://www.geotechtools.org.Within this 

section, the development and content of the system are reviewed and summarized. The 

Ground Modification Reference Manual and GeoTechTools are complementary to one 

another. A Users Guide for GeoTechTools is available on the website and can be accessed 

from the home page. 

7.1 Background, Development, Audience, and Use

GeoTechTools was developed under the auspices of the second Strategic Highway Research 

Program (SHRP2), which was created by the U.S. Congress in 2006 to address challenges of 

moving people and goods efficiently and safely on the nation’s highways. SHRP2 had four 

main focus areas: Safety, Renewal, Reliability, and Capacity, with a number of projects 

under each area. Geotechnical transportation issues were addressed under the SHRP2 

Renewal Focus Area, in which the goal was to develop a consistent, systematic approach to 

the conduct of highway renewal that is (1) rapid, (2) causes minimal disruption, and (3) 

produces long-lived facilities. The SHRP2 R02 project was aimed at identifying geotechnical 

solutions for three elements: (1) construction of new embankments and roadways over 

unstable soils, (2) widening and expansion of existing roadways and embankments, and (3) 

stabilization of the working platform. The R02 project titled: Geotechnical Solutions for Soil 

Improvement, Rapid Embankment Construction, and Stabilization of the Pavement Working 

Platform. 

The project consisted of identification of existing and emerging geotechnical materials and 

systems for ground modification; the technical issues and project development/delivery 

methods necessary to encourage their widespread implementation; performance criteria and 

QA/QC procedures; and the non-geotechnical project-specific obstacles constraining full 

utilization of the identified systems. A catalog of materials and systems for rapid renewal 

projects was developed and the current state of the practice of design, QC/QA, costs, and 

specifications for each technology was assessed. The resulting information was cataloged in 

a database and made accessible through a web-based system. 

The web-based information, guidance, and selection system for geoconstruction and ground 

modification solutions is called GeoTechTools and available at http://www.geotechtools.org. 

The value of the system is that it collects, synthesizes, integrates, and organizes a vast 

amount of critically important information about geotechnical solutions in a system that 

makes the information readily accessible to the user. The target audience for the system is 

primarily public agency geotechnical engineering personnel at local, state, and federal levels. 
1-39 

However, civil/structural, construction, pavement, and construction engineers in consulting, 

contracting, and academia have also found the system useful, as well as transportation 

managers and decision makers. Although developed for the transportation industry, the 

GeoTechTools system can be equally as valuable to other civil construction industries, and 

should have broad appeal to the overall geotechnical community both nationally and 

internationally (Schaefer et al. 2012). 

The system was developed along the lines of the three elements described above; however, 

the final applications were divided into four areas, as shown in Figure 1-1. The system was 

developed with input from the research team members, the project Advisory Board, an 

Expert Contact Group, Federal Highway Administration, and SHRP2. Meetings were 

conducted throughout the project to bring together state agency transportation personnel, 

practitioners, contractors, and academics who work with the relevant geotechnical materials, 

systems, and technology areas. These meetings provided valuable information which 

addressed technical and non-technical obstacles limiting widespread effective use of these 

technologies; available best opportunities to advance the state of practice of the technologies; 

and future directions of these technologies in transportation work. These meetings also 

served to provide valuable refinements and additional information to the final system. The 

goal of the GeoTechTools system is to provide a comprehensive tool that provides guidance 

for applying these geoconstruction solutions to all types of transportation infrastructure. 

7.2 Catalog of Technologies 

The Catalog of Technologies webpage provides a listing of the ~50 geoconstruction 

technologies in the system. Each technology name is directly linked to a Technology 

Information webpage for that technology. The Technology Information page represents the 

technology transfer for each geoconstruction technology included in the system. Included on 

each Technology Information page is a series of ratings. These ratings were developed 

through the completion of a qualitative assessment to rate the technologies according to 

Degree of Technology Establishment in the U.S., Potential Contribution to Rapid Renewal of 

Transportation Facilities, Potential Contribution to Minimal Disruption of Traffic, and 

Potential Contribution to Production of Long-Lived Facilities.

From the individual Technology Information page, the user can access the following 

documents which are generally provided as Portable Document Format (PDF) files: 

Technology Fact Sheet, Photographs, Case Histories, Design Guidance, Quality 

Control/Quality Assurance Guidance, Cost Information, Specifications, and Bibliography. 

These documents were developed based on comprehensive analysis and evaluation of each 

technology to produce an in-depth overview that included advantages, potential 

disadvantages, applicable soil types, depth/height limits, groundwater conditions, material 
1-40 

properties, project specific constraints, equipment needs, and environmental considerations. 

Assessments were completed for design, QC/QA, and specifications to identify key material 

for each technology. The downloadable documents available on the Technology Information 

pages result from the completion of these assessments and evaluations for each technology. 

7.3 Technology Selection Guidance 

The Technology Selection section was developed to aid in identifying a targeted list of 

potential geoconstruction technologies for a user defined set of project conditions (Douglas et 

al., 2012, 2014). The Technology Selection section provides both a listing of the technologies 

sorted by classification and a dynamic, Interactive Selection Tool. After the user identifies 

potential technologies, the Technology Information pages can be accessed which includes 

information necessary for additional screening (i.e., depth limits, applicability to different 

soil types, acceptable groundwater conditions, applicability to different project types, ability 

to deal with project-specific constraints, general advantages/disadvantages, etc.). 

An experienced engineer can access solutions according to particular classifications or 

categories of problems. The technologies are grouped by the following classifications: 

Earthwork Construction, Soft Ground Drainage & Consolidation, Densification of 

Cohesionless Soils, Construction of Vertical Support Elements, Embankments Over Soft 

Soils, Lateral Earth Support, Cutoff Walls, Liquefaction Mitigation, Increased Pavement 

Performance, Void Filling, and Sustainability.

The Interactive Selection Tool allows the user to assess technologies based on several 

applications. The uniqueness of the Interactive Selection Tool is the approach of assigning a 

geoconstruction technology on the basis of application. The first decision in the tool is to 

select one of the four listed applications, which are: Construction over Unstable Soils; 

Construction over Stable or Stabilized Soils; Geotechnical Pavement Components including 

Base, Subbase, and Subgrade; and Working Platforms (shown in Figure 1-1). The Interactive 

Selection Tool is a knowledge based system. Special programming formed the logic and the 

knowledge is contained in a series of tables within the database. Each selection queries a 

database column and utilizes a nested if...then statement to sort the appropriate technologies. 

A significant benefit of the rule-based approach is the sharing of knowledge, especially when 

the knowledge is not the type of knowledge typically published in scholarly publications. 

Similar to most geotechnical analytical solutions, the results of the analysis must be 

measured against the opinion of an experienced geotechnical engineer practicing in the local 

area of the project. The Interactive Selection Tool does not replace the project Geotechnical 

Engineer. The Geotechnical Engineer’s “engineering judgment” should be part of the final 

selection process, which takes into consideration the following project specific items: 
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construction cost, maintenance cost, design and quality control issues, performance and 

safety (pavement smoothness; hazards caused by maintenance operations; potential failures), 

inconvenience (a tangible factor, especially for heavily traveled roadways or long detours); 

environmental aspects, and aesthetic aspects (appearance of completed work with respect to 

its surroundings). 

7.4 Products/Tools 

Each technology has an individual Technology Information webpage that provides access to 

eight technology specific products/tools: Technology Fact Sheets, Photographs, Case 

Histories, Design Guidance, Quality Control/Quality Assurance (QC/QA), Cost Information, 

Specifications, and Bibliography. 

The Technology Fact Sheet is a two-page, summary information sheet that provides basic 

information on the technology, including basic function, general description, geologic 

applicability, construction methods, transportation applications, complementary technologies, 

alternate technologies, potential disadvantages, example successful applications, and key 

references. The Photos show pictorially the equipment or methods used in the technology and 

can be valuable to provide a visual introduction. The Case Histories provide two-page 

summaries of completed projects where the technologies were used, preferably conducted in 

the U.S. by a State Transportation Agency (STA). These summaries provide information on 

project location, owner, a project summary, performance, and contact information. For some 

technologies relatively new to the U.S., the initial case histories were developed from 

projects outside of the U.S. Transportation personnel are encouraged to add Case Histories 

from their work to this website.

The Design Guidance summarizes the recommended design procedures for the technology. 

In cases where a well-established procedure (such as a U.S. Federal Highway Administration 

[FHWA] manual) exists, that procedure is recommended. In cases of technologies with 

multiple design procedures but where there is no preferred procedure, the assessment led to a

recommendation of which procedure(s) to use. The Design Guidance product also identifies 

typical considerations which should be considered including: performance criteria/indicators,

subsurface conditions, loading conditions, material characteristics, and construction 

techniques. The QC/QA Procedures document provides a summary of recommended 

monitoring practices during construction for each technology. The recommended QC/QA 

procedures result from an assessment of the current state of the practice of each technology. 

For a few technologies, design and/or QC/QA procedures were refined and improved during 

the system development.
1-42 

For most technologies, two documents are available to assist with estimating costs. The first, 

a downloadable document from the Technology Information webpage titled Cost 

Information, provides an explanation of the cost items specific to the technology, generally 

emanating from the pay items contained in identified specifications. Project specific 

conditions and their impact on cost are discussed in the explanation. The Cost Information 

compiles available regional and cost numbers from STA bid tabs or national data bases when 

available. For technologies with limited or no STA cost history, the Cost Information 

provides a discussion of important considerations for the technology when estimating costs. 

The second document consists of a spreadsheet developed to assist in preparing preliminarily 

cost estimates for a technology. The spreadsheet can only be accessed through a link in the 

Cost Information document in order to force the user to access the cost discussion prior to 

developing a preliminary estimate. The Excel spreadsheet could not be prepared for some 

technologies due to lack of available cost information. The spreadsheet can be modified by 

the user to estimate specific project cost based on either a preliminary or final design. Many 

decisions in transportation are cost driven. In order to avoid quick elimination of 

technologies based solely on cost, simplified “rule of thumb” costs were avoided in the Cost 

Information documents. The cost spreadsheets require that a preliminary assessment or 

design be completed prior to estimating costs. A valid comparison of technology costs can 

only be completed after a preliminary design has been developed. The information system 

provides the user with the tools to complete a preliminary design and subsequent cost 

analysis which captures the technology-specific costs of implementation and construction. 

A Specifications document is provided for each technology. The information presented in the 

Specifications document varies from identification of an existing specification which can be 

used for future projects, to a specification developed during system development, to a 

description of topics for consideration when developing a specification for a specific 

technology. The final document available for each technology is a Bibliography compiled 

during the research project. In order to assist the user in sorting the references in the 

Bibliography document, a reference matrix with 22 categories is provided to highlight the 

information in the reference, such as technology overview, design procedure, construction 

methods, cost, specification, QC/QA, and case history to name a few.

7.5 Summary 

GeoTechTools is a web-based geoconstruction technologies guidance and selection system. 

Information on over 50 geoconstruction and ground modification technologies has been 

collected, synthesized, integrated, and organized in a single location. While GeoTechTools

was originally developed for the transportation industry, GeoTechTools can be equally as 

valuable to other civil construction industries. 
1-43 

8.0 PROJECT EVALUATION AND GEOTECHNOLOGY SELECTION

8.1 Introduction

The selection of a suitable ground modification method and optimization of its design and 

construction to meet specific project requirements requires extensive knowledge. This 

knowledge must include available and technically feasible ground treatment technologies and 

careful evaluation of several factors. These factors include understanding the functions of the 

methods, utilization of several selection criteria, use of appropriate design procedures, 

selection and implementation of the appropriate methods for quality assurance, and 

consideration of all relevant cost components and environmental factors. 

Evaluation and selection of an appropriate ground modification technology for a specific 

project should be done through a logical, sequential process. The steps in the process include 

a sequence of steps which proceed from simple to more detailed, allowing a best method to 

emerge. The process is described in the sections below. 

8.2 Process to Identify Potential Poor Ground Conditions and Need for Ground 

Modification 

The process to identify potential poor ground conditions, the need for ground modification 

and the selection of appropriate technologies to use follows a logical sequence. The steps 

involved are summarized in Table 1-7 and are explained in more detail below. 

Table 1-7. Ground Modification Technology Selection Steps

Step Description

Step 1 Identify potential poor ground conditions and need for ground modification 

Step 2 Identify or establish performance requirements

Step 3 Identify and assess general site conditions 

Step 4 Assessment of subsurface conditions 

Step 5 Develop a short-list of geotechnologies applicable to site conditions 

Step 6 Consider project constraints 

Step 7 Consider project risks 

Step 8 Prepare a preliminary design

Step 9 Identify alternative solutions (bridge, re-route, deep foundations, etc.) 

Step 10 Evaluate project requirements, constraints, and risks against factors affecting 

geotechnology selection

Step 11 Compare short-list of geotechnology alternatives with geotechnology selection 

factors

Step 12 Select a preferred geotechnology
1-44 

8.2.1 Step 1: Identify Potential Poor Ground Conditions and Need for Ground 

Modification 

Identify project conditions which could require ground modification or geoconstruction 

technologies, such as projects that encounter: 

• Poor ground conditions which will not provide adequate support for a transportation

related structure. Poor ground conditions are typically characterized by soft or loose

foundation soils, which, under load, would cause immediate or long-term

deformation, or cause construction or post-construction instability.

• Project constraints which require retaining walls or steep slopes.

• Pavement foundations which require improvement.

• Construction equipment mobility issues, which require a working platform.

8.2.2 Step 2: Identify or Establish Performance Requirements

Performance requirements include deformation limits (horizontal and vertical), minimum 

factors of safety for stability, improved nominal resistance, stiffness parameters for 

pavements, drainage, and the available time for construction. 

8.2.3 Step 3: Identify and Assess General Site Conditions

General site conditions consider space, constructability, and environmental constraints. Space 

constraints typically refer to accessibility for construction equipment to operate safely and for 

storage of equipment and materials onsite. Work in an urban environment might be 

congested to the point that construction operations and scheduling are affected. Some ground 

modification technologies have large equipment footprints and headroom requirements that 

must be considered for their use. Environmental constraints may include the disposal of spoil 

materials (hazardous or not hazardous), disposal of water from dewatering operations, and 

the effects of construction vibrations, noise, and/or dust. In urban environments these later 

items are of particular importance and can rule out some technologies. 

8.2.4 Step 4: Assess Subsurface Conditions 

The level of detail regarding the assessment of subsurface conditions will vary significantly 

across the wide range of transportation related projects. Regardless of the project type, the 

soils which will affect the performance requirements must be identified and the necessary 

engineering properties established to perform a preliminary design for the project. At a 

minimum, the type, depth, and extent must be considered, as well as the location of the 

ground-water table. For sites with poor ground conditions, it is further valuable to have at 
1-45 

least a preliminary assessment of the shear strength and compressibility of the identified poor 

soils. For assessing the suitability of specific ground modification techniques, knowledge of 

loads and deformation criteria of the proposed structure can help to determine the extent of 

the investigation when particular techniques are under consideration. 

8.2.5 Step 5: Develop a Short-List of Geotechnologies Applicable to Site Conditions

Development of a short-list, or preliminary selection, of potentially applicable technology(s) 

to the site conditions is generally made on a qualitative basis, taking into consideration the 

performance criteria, limitations imposed by subsurface conditions, schedule and 

environmental constraints, and the level of improvement that is required. This can be done 

based on experience with ground modification technologies, using this reference manual for 

guidance on applicable technologies, or using the Technology Selection part of 

GeoTechTools. 

8.2.6 Step 6: Consider Project Constraints

Identification of constraints that might affect selection of ground modifications technologies 

can be done considering general and geotechnical constraints. 

8.2.6.1 General

General constraints include all those items not connected with the geotechnical aspects of the 

project. Included are items such as the project schedule and time; budget and cost; project 

conditions such as right-of-way limits, geometry, scale, utilities and sequence; traffic 

flow/interruption; weather; environmental; availability of agency personnel, agency 

organization and structure, project management philosophy; and contracting processes. The 

contracting process, i.e., project delivery method, and the role it might have on use of these 

techniques for the specific project should be assessed. Identification of as many of these 

constraints affecting the project as possible will allow better selection of appropriate ground 

modification technologies. 

8.2.6.2 Geotechnical

The geotechnical constraints considered here are those specific to geotechnical aspects of the 

project and ground modification technologies that might be used. Such constraints include 

subsurface conditions, including archaeological remains, contaminated groundwater, 

obstructions, and buried utilities; proprietary products/processes; knowledge and experience 

with specific technologies; allowable movements-both vertical and horizontal; stability 

requirements; noise and vibrations due to construction equipment; time for improvement to 

be effective; spoils development and disposal; environmental impacts of the technology; 
1-46 

availability of qualified contractors, personnel, materials and specialty equipment; and 

constructability of the selected technology. Identification of the poor soil types, their depths 

and extent, along with their engineering properties, is perhaps the most important 

consideration here in terms of selecting appropriate ground modification technologies. 

8.2.7 Step 7: Consider Project Risks 

Project risks can occur from not meeting schedule and budget targets, from not identifying 

geotechnical issues correctly, and from not providing an appropriate or adequate fix to the 

poor ground situation. Schedule and budget risks can occur due to contractual issues between 

the general contractor and a specialty contractor, a late notice to proceed, weather affecting 

production, site access issues, material supply problems, equipment availability issues, and 

so on. These risks need to be continuously identified and addressed as the project proceeds. 

Identification of geotechnical risks generally follows from consideration of the possible 

occurrence of such issues as slope instability, settlement, liquefaction, contamination, 

problem soils, etc. These should be identified as part of the subsurface investigation and 

carried forward as potential risks that can affect the project schedule and performance. 

Selection of a ground modification method for the project conditions that turns out not to 

work on the site soil conditions is also a risk that should be identified. 

8.2.8 Step 8: Prepare Preliminary Designs

Based on collection, synthesis and consideration of all the above items, a preliminary list of 

suitable and appropriate ground modification technologies for the project can be developed. 

To compare the various improvement technologies, preliminary design of each candidate 

method for the particular site conditions must be completed. The design for each technology 

will be unique to fit the project performance criteria and constraints. Upon completion of 

preliminary designs for the candidate technologies, preliminary cost estimates can be 

developed and compared. 

8.2.9 Step 9: Identify Alternative Solutions (Bridge, Re-route, Deep Foundations, etc.) 

There always exist alternate solutions available, and these should be evaluated along with the 

ground modification methods. Alternative solutions include the use of a bridge system, the 

use of deep foundations to support the structure rather than ground modification, or perhaps 

rerouting the alignment to avoid the poor ground conditions. 
1-47 

8.2.10 Step 10: Evaluate Project Requirements, Constraints, and Risks Against Factors 

Affecting Geotechnology Selection

For this step, additional geotechnology selection factors are considered. These factors are in 

addition to the primary factors of application, purpose of ground modification, soils type(s), 

depth of soils to be improved, etc. considered in the Step 5 development of a short-list of 

applicable technologies. Additional selection factors may vary by application and project, but 

may include the following factors: constructability (i.e., availability of material, equipment, 

staging, access, groundwater, and etc.); speed of construction; construction disruption 

(amount to traveling public and/or adjacent structures); longevity of constructed works 

(service life and maintenance requirements); ROW requirements or restrictions; aesthetics; 

environmental aspects and concerns; local familiarity with geotechnology; degree of 

establishment of the geotechnology; design procedure (availability, complexity, codified, 

etc.); contracting means and practices; cost (initial and life cycle); additional project 

constraints; and project risks. In this step constraints and risks are explicitly evaluated for 

their effect on the project.

This listing is not project specific and is presented as a tool to help evaluate the short-listed 

geotechnologies developed in Step 5. For a given project, additional selection factors may be 

applicable. Many of items may be of little to no importance on a particular project. 

The evaluation should be performed by the party that is responsible selection of the 

geotechnology to be used. Depending on the contracting method used for the project, this 

evaluation may be performed by the owner, the consulting engineer representing the owner, 

or the contractor. 

Each geotechnology selection factor is evaluated based on its relevancy and importance to 

the project requirements and site constraints and they are assigned a rating number between 

zero (0) and three (3). This is termed the weight factor (WF) for a given geotechnology 

selection factor. Three is assigned to the most relevant or important factors, one is assigned 

to the least relevant ones and zero (or NA-not applicable) is assigned to non-relevant items. 

The evaluation results should be tabulated as shown in Table 1-8. For example, speed of 

construction, amount of disturbance, longevity of constructed works, and cost are typically 

the most important wall selection factors for permanent wall systems.
1-48 

Table 1-8. Geotechnology Selection Factors

Geotechnology Selection Factor Weight Factor (WF)

Speed of construction 0 through 3 

Minimize construction disturbance 0 through 3 

Longevity of constructed works 0 through 3 

Cost of construction 0 through 3 

Constructability 0 through 3 

ROW requirements or restrictions 0 through 3 

Aesthetics 0 through 3 

Environmental 0 through 3 

Degree of establishment 0 through 3 

Familiarity with geotechnology 0 through 3 

Design procedure 0 through 3 

Contracting 0 through 3 

Life-cycle cost 0 through 3 

Additional project constraint 0 through 3 

Additional project constraint 0 through 3 

Project risk 0 through 3 

Project risk 0 through 3 

Project risk 0 through 3 

8.2.11 Step 11: Compare Short-List of Geotechnology Alternatives with Geotechnology 

Selection Factors

The applicability of all the short-listed alternatives (from Step 5) should be evaluated against 

Geotechnology Selection Factors that are defined here using a rating between one (1) and 

four (4). This is a relative rating (RR) of how well the short-listed technologies meet the 

specific geotechnology selection factor for this project. For each factor, four is assigned to 

the most suitable factor for the geotechnology being evaluated and one is assigned to the 

least suitable factor for the geotechnology being evaluated. These ratings are selected based 

on experience with the short-listed geotechnologies in meeting the selection factors.

8.2.12 Step 12: Select a Preferred Geotechnology 

This is the final step where alternatives are compared to each other in a geotechnology 

selection matrix format and the one that has the highest score is selected for the project. The 

scoring of each alternative is obtained for each geotechnology selection factor by multiplying 

WF from Table 1-7 with the 1 through 4 relative rating for each to obtain the weighted rating 

(WR) for each geotechnology selection factor. The WR for all the selection factors are then 

added together to obtain the Total Score for each short-listed technology. The geotechnology 

which has the highest score should be developed as the base design. Other high scoring 
1-49 

geotechnologies may also be included in the Contract Documents as acceptable alternates. 

An example project selection matrix is shown below in Section 8.4. 

8.3 Additional Considerations – Detailed Subsurface Investigation, Design, and Cost 

Estimate

Once a geotechnology is selected additional information may be necessary to develop the 

final design and project cost estimate. Ground modification technologies often require 

specialized geomaterial parameters for design and construction that are not obtained during 

preliminary site investigations. Supplemental detailed subsurface investigations are 

sometimes necessary to obtain the additional information. Details of needed geomaterial 

parameters are provided in the technology summaries. 

8.4 Geotechnology Selection Example

The Geotechnology Selection Factors are factors on which the project selection will be based 

and are different for each project. An example of a project selection matrix is shown in Table 

1-9. 
1-50 

Table 1-9. Sample Project Selection Matrix

Geotechnology Selection 

Factor 

Weight 

Factor 

Geotech A

Relative 

Rating

Geotech A

Weighted 

Rating

Geotech B

Relative 

Rating

Geotech B

Weighted 

Rating

Geotech C

Relative 

Rating

Geotech C

Weighted 

Rating

Speed of construction 2 3 6 4 8 2 4 

Minimize construction disturbance 2 1 2 2 4 2 4 

Longevity of constructed works 3 4 12 4 12 4 12 

Cost of construction 2 3 6 2 4 1 2 

Constructability 3 2 3 2 3 2 6 

ROW requirements or restrictions NA 0 0 0 0 0 0 

Aesthetics NA 0 0 0 0 0 0 

Environmental 1 0 0 0 0 1 1 

Degree of establishment 3 3 9 2 6 2 6 

Familiarity with geotechnology 2 2 4 1 2 2 4 

Design procedure 2 3 6 2 4 2 4 

Contracting 1 4 4 4 4 3 3 

Life-cycle cost 1 3 3 3 3 3 3 

Constraint – construction season 3 2 6 2 6 1 3 

Risk – delay due to settlement time 3 4 12 2 6 2 6 

Risk – quality assurance 2 1 2 3 6 3 6 

TOTAL SCORE 75 69 64 
1-51 

Example factors are shown in the table but others can be added as necessary. The 

Geotechnology Selection Factor Weight Factors are determined in Step 10 and Table 1-8. 

The Geotechnology Relative Ratings are determined in Step 11. The Total Score for each 

geotechnology is the sum of the Weighted Ratings as summarized at the bottom of the table. 

In this example, Geotechnology A would be advanced to final design and cost analysis, and 

Geotechnology B may be accepted as an alternate.

8.5 Combination of Geotechnologies 

Combinations of technologies should be considered in the geotechnology selection process. 

Typical combinations are noted within the Complementary Technology Sections, within each 

technology summary. Complementary technologies are also listed on the Technology Fact 

tool on the http://www.geotechtools.org website.

There are many possible combinations of technologies, as listed in Table 1-10. 

However, there are a number of combinations that are more likely to be successful. For 

example, blast densification does not densify the soil in the upper 5 feet of soil below the 

ground surface. This soil can be compacted by a shallow compaction method such as rapid 

impact compaction. Prefabricated vertical drains and fill preloading are often used together to 

accelerate consolidation and the construction schedule. Lightweight fill material is used to 

reduce the load on column supported embankments. In some cases, the subsurface conditions 

across a site or project vary significantly enough to warrant the use of different ground 

modification technologies. Numerous successful combinations that were found during a 

literature review on embankments over unstable ground (Barngrover et al. 2013) are noted in 

Table 1-10. 
1-52 

Table 1-10. Technology Combinations Found in Literature Review

Technology

PVDs and Fill Preloading Vacuum Preloading with and without PVDsVibro-CompactionDeep Dynamic Compaction Blast DensificationAggregate Columns Geotextile Encased ColumnsDeep Mixing Methods Jet Grouting Soil NailingMechanically Stabilized Earth WallsReinforced Soil Slopes Lightweight FillsColumn Supported Embankments Micropiles

PVDs and Fill Preloading X X X X X 

Vacuum Preloading with and without PVDs X 

Vibro-Compaction X X X 

Deep Dynamic Compaction X X X 

Blast Densification X X 

Aggregate Columns X X X X 

Geotextile Encased Columns X X 

Deep Mixing Methods X X X 

Jet Grouting X X 

Soil Nailing X X 

Mechanically Stabilized Earth Walls X X X X 

Reinforced Soil Slopes X 

Lightweight Fills X X X 

Column Supported Embankments X 

Micropiles X X 

Source: Barngrover et al. 2013

Blank cells=No case histories located for that combination
1-53 

9.0 REFERENCES

AASHTO. (2006). Standard Recommended Practice for Definition of Terms Related to 

Quality and Statistics As Used in Highway Construction. American Association of 

State Highway and Transportation Officials, Washington, D.C. 

ASCE. (1978). Soil Improvement-History, Capabilities, and Outlook. J.K Mitchell, Editor, 

ASCE, New York, NY, 182p. 

ASCE. (2013). Report Card for America’s Infrastructure. American Society of Civil 

Engineers, Washington, D.C. 

Barngrover, A.L., Berg, R.R., Mitchell, J.K., and Collin, J.G. (2013). White Paper on 

Integrated Technologies for Embankments on Unstable Ground. Virginia Tech Center 

for Geotechnical Practice and Research, Report CGPR #74, Blacksburg, VA, 85p. 

Charles, J.A. (2002). Ground Improvement: the Interaction of Engineering Science and 

Experience-based Technology. Géotechnique, 52(7): pp. 527-532. 

Chu, J., Varaksin, S., Klotz, U., and Menge, P. (2009). Construction Process. Proc. 17th

International Conference on Soil Mechanics and Geotechnical Engineering, 

Alexandria, Egypt, pp. 3006-3135. 

Construction Inspection and Approval, Code of Federal Regulations (4-1-11 Edition), Title 

23, Part 637, U.S. Government Printing Office, Washington, D.C. 

https://www.gpo.gov/fdsys/granule/CFR-2011-title23-vol1/CFR-2011-title23-vol1-

part637. 

DiMillio, A.F. (1999). A Quarter Century of Geotechnical Research. FHWA-RD-98-139, 

Federal Highway Administration Research, Development and Technology, U.S. 

Department of Transportation, 151p. 

Douglas, S.C., Schaefer, V.R., and Berg, R.R. (2012). Selection Assistance for the 

Evaluation of Geoconstruction Technologies for Transportation Applications. Journal 

of Geotechnical and Geological Engineering, 30(5): pp. 1231-1247. 

Douglas, S.C., Schaefer, V.R., and Berg, R.R. (2014). Development of the Geoconstruction 

Information and Technology Selection Guidance System. SHRP2 Report S2-R-2-2, 

Transportation Research Board, Washington, D.C.

Dunnicliff, J. (1988). Geotechnical Instrumentation for Monitoring Field Performance. John 

Wiley & Sons, New York, NY, 577p. 
1-54 

Elias, V., Welsh, J., Warren, J., and Lukas, R. (1999). Ground Improvement Technical 

Summaries. DTFH61-93-R-00145, Federal Highway Administration, U.S. DOT, 

Office of Technology Application. 

FHWA. (2008). Transportation Construction Quality Assurance Reference Manual, Report 

No. FHWA-NHI-08-067, Arlington, VA. 

FHWA. (2010). Technical Advisory – Development and Review of Specifications. Federal 

Highway Administration, U.S. DOT, March 24, 2010. 

http://www.fhwa.dot.gov/construction/specreview.pdf. Accessed 13 May 2015. 

FHWA. (2011). Project Development and Design Manual. Federal Lands Highway, Federal 

Highway Administration, U.S. DOT. 

Golder Associates Inc. (2011). SHRP 2 Report S2-R09-RW-2: Guide for the Process of 

Managing Risk on Rapid Renewal Projects. Second Strategic Highway Research 

Program, Transportation Research Board, National Academies, Washington, D.C. 

Gransberg, D.D. and Loulakis, M.C. (2012). Geotechnical Information Practices in DesignBuild Projects, Synthesis 429, Transportation Research Board, National Academies, 

Washington, D.C. 

Marr, W.A. (2007a). Why Monitor Performance? FMGM 2007: Seventh International 

Symposium on Field Measurements in Geomechanics, Geotechnical Special 

Publication No. 175, American Society of Civil Engineers, Reston, VA. 

Marr, W.A. (2007b). GeoEngineering in the New Millennium – Testing and Performance 

Monitoring. Keynote Paper, Proc. Civil Engineering For the New Millennium, 150th

Anniversary of Bengal Engineering College, Shibpur, India. 

Marr, W.A. (2013). Instrumentation and Monitoring of Slope Stability. Geotechnical Special 

Publication No. 231, GeoCongress 2013: Stability and Performance of Slopes & 

Embankments III, San Diego, CA, pp. 2231-2252. 

Mitchell J.K. (1981). Soil Improvement: State-of-the-Art. Proc. Tenth International Conf. on 

Soil Mechanics & Foundation Engineering, Stockholm, Sweden, June, Vol. 4, pp. 

509-565. 

Munfakh G.A. and Wyllie, D.C. (2000). Ground Improvement Engineering—Issues and 

Selection. GeoEng 2000, 19-24 Nov. 2000, Melbourne, Australia, Vol. 1: Invited 

Papers, Technomic, Lancaster, PA, pp. 333-359. 
1-55 

Schaefer, V.R., Mitchell, J.M., Berg, R.R., Filz, G.M., and Douglas, S.C. (2012). Ground 

Improvement in the 21st Century: A Comprehensive Web-Based Information System. 

Geotechnical Engineering State of the Art and Practice, Keynote Lectures from 

GeoCongress 2012, Editors: K. Rollins and D. Zekkos, Geotechnical Special 

Publication No. 226, Geo-Institute of ASCE, Reston, VA, pp. 292-293. 

Scott, S., Konrath, L., and Ferragut, T. (2014). Framework for Performance Specifications: 

Guide for Specification Writers. SHRP2 Report S2-R07-RR-3, The Second Strategic 

Highway Research Program, Transportation Research Board, Washington, D.C. 

USACE. (1999). Guidelines on Ground Improvement for Structures and Facilities. Technical 

Letter No. 1110-1-185, Department of the Army. U.S. Army Corps of Engineers, 

Washington, D.C., 42p. 

Chapter 2 

VERTICAL DRAINS 

AND 

ACCELERATED 

CONSOLIDATION

2-i 

CONTENTS

1.0 DESCRIPTION AND HISTORY ........................................................................... 2-1

1.1 Description ............................................................................................................ 2-1

1.2 Historical Overview ............................................................................................. 2-3

1.3 Focus and Scope ................................................................................................... 2-6

1.4 Glossary ................................................................................................................ 2-6

1.5 Primary References ............................................................................................. 2-7

2.0 FEASIBILITY CONSIDERATIONS .................................................................... 2-9

2.1 Applications .......................................................................................................... 2-9

2.1.1 Preloading of Soft Soils ..................................................................................... 2-9

2.1.2 Special Application, Liquefaction Reduction .................................................... 2-9

2.2 Advantages and Potential Disadvantages ........................................................ 2-10

2.2.1 Advantages ....................................................................................................... 2-10

2.2.2 Potential Disadvantages ................................................................................... 2-11

2.3 Feasibility Evaluations....................................................................................... 2-12

2.3.1 Geotechnical Criteria ....................................................................................... 2-12

2.3.2 Environmental Considerations ......................................................................... 2-12

2.3.3 Site Conditions ................................................................................................. 2-13

2.4 Limitations .......................................................................................................... 2-15

2.5 Alternate Solutions............................................................................................. 2-16

3.0 CONSTRUCTION AND MATERIALS .............................................................. 2-18

3.1 Construction ....................................................................................................... 2-18

3.1.1 Prefabricated Vertical Drains ........................................................................... 2-18

3.1.2 Staged Construction ......................................................................................... 2-24

3.2 Materials of Construction ................................................................................. 2-25

3.2.1 Prefabricated Vertical Drains ........................................................................... 2-25

3.2.2 Drainage Layer................................................................................................. 2-26
2-ii

4.0 DESIGN .................................................................................................................. 2-29

4.1 Design Considerations ....................................................................................... 2-29

4.2 Design Procedure ............................................................................................... 2-31

4.3 Preliminary Spacing Design .............................................................................. 2-36

4.4 Design Example .................................................................................................. 2-38

5.0 CONSTRUCTION SPECIFICATIONS AND CONSTRUCTION

CONTROL ......................................................................................................................... 2-39

5.1 Introduction ........................................................................................................ 2-39

5.2 Specification Development ................................................................................ 2-39

5.3 Quality Assurance .............................................................................................. 2-40

5.3.1 Site Preparation ................................................................................................ 2-40

5.3.2 Prefabricated Vertical Drain Material and Installation Equipment ................. 2-41

5.4 Instrumentation Monitoring and Construction Control ................................ 2-42

6.0 COST DATA .......................................................................................................... 2-49

6.1 Cost Components ............................................................................................... 2-49

6.2 Summary of Typical Projects ........................................................................... 2-51

7.0 CASE HISTORIES ................................................................................................ 2-52

7.1 Council Bluffs (IA) Interstate System Improvement Program ..................... 2-52

7.1.1 Project Description........................................................................................... 2-52

7.1.2 Subsurface Conditions ..................................................................................... 2-52

7.1.3 Design Concerns .............................................................................................. 2-52

7.1.4 Design .............................................................................................................. 2-53

7.1.5 Project Results ................................................................................................. 2-53

7.1.6 Project Cost ...................................................................................................... 2-53

7.2 Approach Ramps for Bridge Replacement...................................................... 2-54

7.2.1 Project Description........................................................................................... 2-54

7.2.2 Subsurface Conditions ..................................................................................... 2-55

7.2.3 Design Concerns .............................................................................................. 2-55

7.2.4 Design .............................................................................................................. 2-56

7.2.5 Project Construction Procedure ....................................................................... 2-56
2-iii

7.2.6 Project Results ................................................................................................. 2-57

7.2.7 Project Costs .................................................................................................... 2-57

7.3 Jones Creek Substation Site Fill ....................................................................... 2-60

7.3.1 Project Description........................................................................................... 2-60

7.3.2 Subsurface Conditions ..................................................................................... 2-60

7.3.3 Design Concerns .............................................................................................. 2-60

7.3.4 Design .............................................................................................................. 2-62

7.3.5 Project Results ................................................................................................. 2-62

7.3.6 Project Cost ...................................................................................................... 2-62

8.0 REFERENCES ....................................................................................................... 2-64
2-iv

LIST OF FIGURES

Figure 2-1. Prefabricated vertical drain installation for a highway embankment. ................. 2-2

Figure 2-2. Typical prefabricated vertical drain products: PVDs with corrugated inner 

core (top), PVD with studded inner core (middle), vertical earthquake drain 

(bottom)............................................................................................................................ 2-5

Figure 2-3. Typical prefabricated vertical drain installation equipment: installation rig 

(top), box mandrel (bottom). .......................................................................................... 2-19

Figure 2-4. Typical prefabricated vertical drain installation procedure: placing the 

anchor on the drain (top), inserting the mandrel into the ground (middle), cutting 

the drain after withdrawing the mandrel (bottom). ........................................................ 2-22

Figure 2-5. Typical prefabricated vertical drain splicing procedure: inserting the drain 

core within the jacket to maintain continuity (top), stapling the drain splice 

(bottom).......................................................................................................................... 2-23

Figure 2-6. Strip drains. ....................................................................................................... 2-27

Figure 2-7. Drain patterns: triangular pattern (top left), square pattern (top right), 

equivalent cylinder (bottom). ......................................................................................... 2-31

Figure 2-8. Asaoka (1978) method for determining ch and end of primary 

consolidation. ................................................................................................................. 2-34

Figure 2-9. Typical cross-section of a band-shaped drain and mandrel. ............................. 2-35

Figure 2-10. Coefficient of consolidation versus liquid limit. ............................................. 2-37

Figure 2-11. Displacement ratio versus time and stage filling. ........................................... 2-44

Figure 2-12. Cumulative displacement versus depth. .......................................................... 2-46

Figure 2-13. Incremental displacement versus depth. ......................................................... 2-47

Figure 2-14. PVD Installation showing typical instrumentation and monitoring. ............... 2-48

Figure 2-15. Settlement versus fill height for test pad. ........................................................ 2-54

Figure 2-16. Tifft Street west approach embankment field data. ........................................ 2-58

Figure 2-17. Tifft Street east approach embankment field data........................................... 2-59

Figure 2-18. Final grading plan for Jones Creek Substation fill project. ............................. 2-61

Figure 2-19. Settlement plate data and fill height versus project duration. ......................... 2-63
2-v 

LIST OF TABLES

Table 2-1. Common Uses of PVDs for Transportation Applications .................................... 2-9

Table 2-2. Methods of Prefabricated Vertical Drain Installation ........................................ 2-18

Table 2-3. Typical Unit Price Ranges for PVDs .................................................................. 2-51

Table 2-4. Estimated and Installed PVDs ............................................................................ 2-53

Table 2-5. Estimated and Installed PVDs ............................................................................ 2-62

2-1 

1.0 DESCRIPTION AND HISTORY

1.1 Description

Consolidation of soft ground using vertical drains and preloading is a technique used since 

the 1920s. The vertical drains provide a shortened pathway for water to exit the soils while 

fill preloading surcharges the foundation soils. Initially sand drains were used, then 

cardboard drains, followed by geotextile-encased drains (prefabricated vertical drains). The 

most common vertical drain at present is the use of prefabricated vertical drains (PVDs). 

PVDs are often called wick drains in the United States, but the term PVD is more appropriate 

since drainage is via pressure, and not by wicking. PVDs are used to accelerate consolidation 

of soft saturated compressible soils under load. 

PVDs are band shaped (rectangular cross-section) products consisting of a geotextile jacket 

surrounding a plastic core with drainage channels. This configuration permits pore water in 

the soil to seep into the drain for collection and transmittal up and down the length of the 

core. While there are some variations, the size of a PVD is typically 4 inches wide by 1⁄8 to 3⁄8

inches thick. Basic evolution of the term wick drains comes from the idea that they look like 

a wick, and the first available product was known as a “cardboard wick.” Other common 

terms for these drains are drainage wicks, band drains, and strip drains. 

PVDs are only one general type of vertical drain system. Aggregate columns (see Chapter 5) 

and geotextile encased columns (see http://www.GeoTechTools.org) also provide vertical 

drainage. The general principles that govern all vertical drain installations are similar for all 

types of drains. However, the cost advantages of PVDs over other vertical drain systems have 

resulted in their almost exclusive use except in unusual circumstances as outlined under 

Section 2.4 Limitations. 

The most common transportation use of PVDs is to accelerate consolidation for approach 

embankments at bridges or other embankment construction over soft soils, where the total 

post construction settlement would otherwise be unacceptable. A typical cross-section of a 

PVD installation for embankments is shown in Figure 2-1. 
2-2 

Figure 2-1. Prefabricated vertical drain installation for a highway embankment. 

A preload and a surcharge are shown in Figure 2-1. Preload is defined as the application of 

load to induce settlement and consolidation of the foundation soils. Often this is the amount 

of fill material necessary to bring the site to the final elevation, including the amount of 

additional embankment necessary to accommodate the final amount of settlement. Surcharge 

is the extra load or the added fill above final elevations that is used to accelerate settlement 

and/or minimize secondary consolidation; it is removed after the desired consolidation is 

achieved. 

Consolidation of compressible soils by vacuum preloading was conceptually introduced in 

the 1950s and has recently evolved as a reasonably reliable technology. The basic premise 

for vacuum consolidation consists of removing atmospheric pressure from a confined and

sealed soil to be consolidated and maintaining the vacuum for a predetermined period of 

time. When used in combination with PVDs, the soil is loaded approximately uniformly 

throughout its depth by the equivalent of one atmosphere of vacuum. Although used 

extensively outside the United States, vacuum preloading has seen little use in the United 

States and will not be discussed further herein. Information on vacuum preloading can be 

found on http://www.GeoTechTools.org under the Vacuum Preloading With and Without 

PVDs technology and the references therein. 
2-3 

1.2 Historical Overview

The development of PVDs almost parallels the development of the vertical sand drain 

concept. A United States patent for a sand drain system was granted in 1926 to D. J. Moran. 

The California Division of Highways, Materials and Research Department, conducted 

laboratory and field tests on vertical sand drain performance as early as 1933. From that time 

until the early 1980s, sand drains were used almost exclusively for highway projects across 

the United States that required a vertical drain solution.

In Europe, the lack of available sand with suitable drainage characteristics and the presence 

of certain environmental drawbacks led to the development of the first PVD in the late 1930s 

by Walter Kjellman, then Director of The Swedish Geotechnical Institute. This drain utilized 

three layers of cardboard, with the two outer layers serving as a filter and the middle layer as 

a 10-channel separator. The drain measured 4 inches wide by 1⁄8 inches thick and became 

known as the “Cardboard Wick.” This drain was used extensively in Sweden after 1939, 

when a machine for installation was developed. Later these drains found their way into other 

countries. Sweden subsequently, however, reverted to the use of sand drains, mainly because 

of the low permeability, poor resistance to pressure, and the low water transportation 

capability of the cardboard material. In spite of this, Oleg Wager, a colleague of Kjellman at 

The Swedish Geotechnical Institute, believed so strongly in the potential of the PVD that he 

worked on and developed an improved drain, and in 1972 patented the Geodrain. This drain 

had essentially the same dimensions as the “Cardboard Wick” but was constructed 

differently. The Geodrain consisted of a plastic core containing 27 grooved channels for 

water transportation, and was surrounded by a filter material made of paper. PVDs, as we 

know them today, had finally arrived. Almost simultaneously the Castleboard Drain was 

developed in Japan. It was very similar in appearance to the Geodrain, but with a geotextile

material, instead of paper, used as a filter.

PVDs did not come into use in the United States until the mid-to-late 1970s. Prior to this, the 

use of jetted and augered nondisplacement sand drains had proved to be an effective vertical 

drain system. Another contributing factor to this timing was that all of the practical PVD 

experience had been attained mostly in Europe and Japan.

With environmental and economic factors slowly eliminating the use of jetted and augured 

sand drains, a newly designed PVD material known as Alidrain was introduced in the 

United States. Acceptance, while initially slow due to the lack of prior experience and 

design procedures, grew quite rapidly. Today there are hundreds of projects completed 

yearly. In fact, PVDs are now used almost exclusively where a vertical drain solution is 

required. 
2-4 

Since the development of the first cardboard wick, there have been over 50 types of PVDs 

used worldwide and at least 10 in the United States. Historically, most of those installed in 

the United States are one of two basic designs. Both have an independent outside geotextile 

filter, but the core of one type is corrugated, and the other is studded. The corrugated design 

results in straight channels for flow. This type of PVD design currently accounts for nearly 

100 percent of the United States market. Commonly used brand names include 

AMERDRAIN 407®, MEBRA-DRAIN 7407®, MEBRA-DRAIN MD 88®, 

COLBONDDRAIN CX-1000® and ALIWICK®. A typical PVD of this type is shown in 

Figure 2-2 (top). 

The studded core-type design results in a more turbulent flow pattern, thereby necessitating a

slightly thicker drain in order to obtain the same flow rates. Common studded core brand 

names are ALIDRAIN® and AMERDRAIN 417®. These devices account for a small 

percentage of the United States market because of their slightly higher cost. A typical PVD 

of this type is shown in Figure 2-2 (middle). 

There are no available data from completed projects that would indicate any advantages or 

disadvantages for the two types of cores. Much of the early usage was with the studded 

design, but currently the grooved design is used most often, based on its lower material cost. 

Recently, high capacity vertical composite drains have been introduced to mitigate 

liquefaction (Rollins et al. 2004 and Rollins and Strand 2007). Termed earthquake drains 

(EQ drains), these drains consist of 3 to 6 inch diameter corrugated, perforated drain pipe 

wrapped with a geotextile filter fabric. A typical EQ drain is shown in Figure 2-2 (bottom). 
2-5 

MebraDrain MD4707® (top left), Ameridrain 407® (top right), TenCate Alidrain® (middle), 

Nilex Construction®, after Utah DOT (bottom) 

Figure 2-2. Typical prefabricated vertical drain products: PVDs with corrugated inner 

core (top), PVD with studded inner core (middle), vertical earthquake drain (bottom). 
2-6 

1.3 Focus and Scope

The principal purpose of this chapter is to acquaint the reader with the use of vertical drain 

methods. Many of the concepts and ideas stated herein are taken from practical experience 

and the basic references, listed in Section 1.5 Primary References and in the references at the 

end of this chapter. 

The content of this technical summary includes the following: 

• Applications

• Feasibility

• Construction procedures and monitoring

• Design concepts

• Specifications

• Case histories

The major use of vertical drains is for consolidation of soft soils by preloading and/or 

surcharging. Other PVD applications include: pressure relief wells to reduce pore pressures 

due to seepage; lowering perched water table conditions; and reducing liquefaction potential 

in soils.

When used in conjunction with surcharging or preloading, PVDs have the following 

principal benefits: 

• To decrease the settlement time required such that final construction can be

completed in a reasonable time, with minimal post-construction settlement

• To decrease the amount of surcharge or preload material required to achieve a

settlement in the given time

• To increase the rate of strength gain due to consolidation of soft soils when stability is

of concern

Any one of these benefits may be the sole reason for use on a particular project, or any 

combination of benefits may be the desired result. 

1.4 Glossary 

Consolidation – a time-dependent settlement process that occurs in saturated fine-grained 

soil that have a low coefficient of permeability.
2-7 

Earthquake drains (EQ drains) – high flow capacity synthetic vertical drains installed with 

a vibrating mandrel consisting of a corrugated plastic pipe three to six inches in diameter 

wrapped with a geotextile filter.

Geosynthetic – a planar product manufactured from a polymeric material used with soil, 

rock, earth, or other geotechnical–related material as an integral part of a civil engineering 

project, structure, or system. 

Geotextile (GT) – a permeable geosynthetic made of textile materials, used as a separator 

between base, subbase and subgrade layers, used as filters in drainage features, and used in 

stabilization of soft subgrade layers. 

Preconsolidation stress – the maximum past effective overburden stress to which a soil has 

been subjected.

Prefabricated vertical drain (PVD or PV drains) – band shaped (rectangular crosssection) products consisting of a geotextile jacket surrounding a plastic core with drainage 

channels. Water flows from soil through the filter into the core of the drain and from there 

upward and/or downward to the nearest free draining layer. 

Preloading – application of a load to site soils to induce settlement or consolidation of the 

foundation soils. The load may be part of the permanent embankment or be temporarily 

applied. 

Primary consolidation – The compression of the soil under load that occurs while excess 

pore pressures dissipate with time. 

Secondary compression – time dependent settlement occurring at constant effective stress 

with no subsequent changes in pore water pressure. 

Surcharge – an extra load applied above the preload to accelerate consolidation or minimize 

secondary compression. 

Wick drain – another name for prefabricated vertical drain (PVD). 

1.5 Primary References 

The primary references for Vertical Drainage and Accelerated Consolidation are the 

following: 
2-8 

• Chu, J., Bo, M.W., and Choa, V. (2004). Practical Considerations for Using Vertical

Drains in Soil Improvements Projects. Geotextiles and Geomembranes, 22: pp. 101-

117. 

• Holtz, R.D., Jamiolkowski, M.B., Lancellotta, R., and Pedroni, R. (1991).

Prefabricated Vertical Drains: Design and Performance. Construction Industry

Research and Information Association, Butterworth-Heinemann, Oxford, UK, 131p.

• Rixner, J.J., Kraemer, S.R., and Smith A.D. (1986a). Prefabricated Vertical Drains.

Report No. FHWA/RD-86/168, Vol. I: Engineering Guidelines, Federal Highway

Administration, U.S. DOT, Washington, D.C. Available at:

http://www.fhwa.dot.gov/bridge/geopub.htm.

• Rixner, J.J., Kraemer, S.R., and Smith A.D. (1986b). Shared Experience in

Geotechnical Engineering, Wick Drains Transportation Research Circular, Number

309. 
2-9 

2.0 FEASIBILITY CONSIDERATIONS

2.1 Applications

2.1.1 Preloading of Soft Soils 

In practice, PVDs are most commonly used in consolidation situations where the soil to be 

treated is a moderately to highly compressible soil with low permeability and is fully 

saturated in its natural state. Such soils are typically described as silts, clays, organic silts, 

organic clays, muck, peat, swamps, muskeg, or sludge. 

In general, soils for which PVDs are being used must be saturated and normally to slightly 

over-consolidated prior to loading. In addition, PVDs are used in under-consolidated soils in 

land reclamation applications. The loading should exceed the preconsolidation stress (the 

maximum past effective overburden stress) of the soils for the PVDs to be totally effective.

Ordinarily, PVDs are not used in highly organic materials, or where secondary consolidation 

will result in significant post-construction settlement. However, additional surcharging may 

be used with this solution to minimize the effect of secondary consolidation. 

Many of the common uses of PVDs are illustrated in Table 2-1, but are not to be considered 

all-inclusive. 

Table 2-1. Common Uses of PVDs for Transportation Applications

Field of Application

Purpose is to 

Increase

Stability

Purpose is to 

Accelerate

Settlements

Highways Roadway Embankments Yes Yes

Highway Structure Approach Fills Yes Yes

Airfield Runways and Taxiways Yes Yes

Earth Embankment Dams Yes Yes

Storage Tanks Yes Yes

Pile Foundations to Reduce “Negative” Skin Friction No Yes

Liquefaction Mitigation Yes Yes

Land Reclamation Yes Yes

2.1.2 Special Application, Liquefaction Reduction

One of the potential uses of PVDs that is currently emerging, is the installation of drains to 

reduce the potential for liquefaction of saturated granular soils. The PVDs provide a pathway 

for excess pore water pressures developed by seismic excitation to dissipate rapidly through 
2-10 

the drainage pathway provided by the PVDs. In this application, the PVDs simply replace the 

gravel drains that have been in use for liquefaction mitigation for the past 30 years. The 

PVDs for these applications are the EQ Drains, which are vertical, slotted plastic drain pipes 

of 3 to 6 inches in diameter and have a high flow capacity. The high flow capacity provides 

rapid dissipation of excess pore water pressures that, if not dissipated, would lead to 

liquefaction of the soils. 

Limited research has been conducted to determine the extent of potential liquefaction 

reduction due to EQ Drains. However, it appears that some degree of improvement will be 

obtained. Additional information can be found in Rollins et al. (2004) and Rollins and Strand 

(2007). 

2.2 Advantages and Potential Disadvantages 

2.2.1 Advantages 

2.2.1.1 Economy 

For typical projects, the cost for PVDs is much less than available alternatives such as 

aggregate drains. 

2.2.1.2 Installation

Generally, the production rate for PVDs will average between 15,000 to 20,000 lineal feet 

per day per rig. Production rates as high as two to three times these figures have been 

achieved, but for planning purposes, 15,000 feet per day is generally used. 

2.2.1.3 Continuity of Drain 

PVDs provide an assurance of a permanent drainage path, even with considerable lateral 

displacement or buckling under vertical or horizontal soil movements. 

2.2.1.4 Minimal Displacement

The typical size of mandrel used for PVDs is small enough to create minimal displacement 

during installation. The displacement effects on permeability of the disturbed soil can be 

taken into account in the PVD design process that leads to an increase of time needed for 

consolidation. There has been no evidence of a significant reduction of shear strength of in 

situ soils due to remolding. 
2-11 

2.2.1.5 Improved Quality Control 

The quality control and assurance for PVD construction is quite simple. Because continuity 

of the drain is assured during installation, the major duties of inspection are to ensure proper 

drain anchorage and proper depth attainment.

2.2.1.6 Equipment Flexibility 

There are many types and sizes of PVD installation equipment that can be easily adapted to 

field conditions. PVDs are generally installed with a static or vibratory installation force, and 

normal equipment can be adapted for a minimal amount of jetting, where necessary. Very 

lightweight equipment can also be used in unstable ground conditions. 

2.2.1.7 Low Material Storage

PVDs come in reels usually containing 450 to 1,000 feet of material. Each roll is 

approximately 3.3 feet in diameter, 4 inches thick, and can be easily stored. 

2.2.1.8 No Spoil Removal 

With the exception of situations where pre-augering or drilling through stiff layers is 

necessary to reach the soft compressible soil layer, there is no significant excess spoil 

material to be removed from the site. 

2.2.1.9 No Water Required

Except in unusual cases, PVDs are installed without jetting. Even if a minimum amount of 

jetting is required, the resulting surface runoff is minimal.

2.2.2 Potential Disadvantages 

2.2.2.1 Headroom Limitations

PVD installation equipment must be 5 to10 feet taller than the depth of installation which can 

limit their use for some sites. 

2.2.2.2 Materials Must be Stored Properly 

PVD material can degrade in sunlight and, therefore, must be stored properly. While most 

specifications require the material to be covered during storage, the effect of sunlight on the 

geotextile will not be significant unless the material is on site for more than a month. 
2-12 

2.3 Feasibility Evaluations

When PVDs are used to accelerate settlement during loading, the subsoil must meet the 

following criteria:

• Moderate to high compressibility

• Low permeability

• Full saturation

• Final embankment loads must exceed preconsolidation stress

• Secondary consolidation must not be a major concern

• Low-to-moderate shear strength

Providing the soils meet the above geotechnical criteria, the project still must be evaluated 

for the possible effects of environmental and other site conditions. 

2.3.1 Geotechnical Criteria

To ascertain the geotechnical criteria to ensure a successful design, the subsurface 

investigation must be sufficiently extensive to determine the extent and depth of the 

compressible soils and to secure high-quality undisturbed samples for consolidation and 

strength testing. In the preliminary phase, laboratory testing consisting of Atterberg Limits, 

moisture content and organic content within short depth intervals is extremely useful in 

determining strata changes and providing insight as to the depth in which undisturbed 

samples should be recovered. Sufficient consolidation testing must be performed to 

determine the over-consolidation ratios (OCR) with depth as well as shear strength testing to 

determine both undrained and drained shear strengths. The use of piezocone (CPT) 

soundings is valuable for identifying thin interbedded granular layers, which affect 

settlement-time extrapolations. Piezocone pore pressure dissipation tests could be used to 

determine field coefficients of consolidation, although the literature indicates that ch obtained 

from piezocone methods typically varies by half an order of magnitude from actual values 

(Bartlett et al. 2001).

2.3.2 Environmental Considerations 

If the in situ soils are contaminated with any kind of hazardous waste or material, then it is 

possible that the excess pore water draining through the PVD will need to be collected and 

treated. In such situations, care must be exercised to prevent the PVD from penetrating into a 

highly permeable layer, should one exist below the compressible stratum. 
2-13 

2.3.3 Site Conditions

Site topography and in situ soil conditions can have a considerable effect on the economics 

of a PVD solution. Some of the specific site and soil conditions that affect the economics or 

feasibility are listed below and subsequently discussed: 

• Uneven working surface

• Limited headroom

• Obstructions above the compressible layer

• Unstable working surface

• Depth of PVDs in excess of 100 feet

• Stiff to very stiff compressible layers

• Extremely soft layer for anchoring

• Poor site accessibility

• Overhead or subsurface utility interference

Uneven working surface. PVDs cannot be installed economically on steep slopes. Therefore, 

the area will have to be benched with widths sufficient to allow for the equipment. Generally, 

a minimum bench width of 25 feet is required. 

For shallow depths, PVDs can be installed on slopes as steep as one on five. Deeper drains 

will require a level working surface to ensure uniform drain spacing at all depths. A constant 

minor slope is preferable to an undulating surface, and it also facilitates the construction of a 

more effective drainage blanket.

Limited headroom. A rule of thumb is that the depth of a PVD needs to be 10 feet shorter 

than the available headroom in order to be economically installed. Limited headroom 

situations occur most often when installing under an existing bridge. PVDs can be installed 

vertically in segments with limited headroom, but the cost would most likely be as high as 

five times the normal unit installation price.

Obstructions above the compressible layer. Where obstructions must be penetrated above the 

compressible layer, considerable extra costs could be involved. A stiff or dense upper layer 

that can be penetrated without pre-drilling will add minimally to the cost. Obstructions such 

as concrete, rock, rubble, slag, brick, wood, riprap, stone, debris, rubbish, or trash usually 

require pre-drilling. Dense layers with SPT blow counts greater than 10 to 15 may require 

pre-drilling.
2-14 

Unstable working surface. In general, most unstable working surfaces can be made stable 

prior to the installation of PVDs with the use of geosynthetics and granular soil for the 

drainage layer (2 to 3 feet thick). The installation equipment will usually penetrate these

materials without difficulty. Where the ground cannot be stabilized, movable timber mats 

may be used for support, or specialized lightweight equipment is available at a substantial 

increase in the unit cost.

Extreme depth. PVDs have been installed to depths of 200 feet with the use of specialized 

equipment. A rule of thumb is that PVDs over 100 feet in depth will require a crane for 

installation. Depths over 120 feet require a very large crane and specialized installation 

masts.

Stiff-to-very-stiff compressible layers. If the layer that is considered compressible is quite 

stiff, the entire length of PVDs may need to be pre-augered or predrilled. In such cases, it is 

not normally advisable to use PVDs. The pre-augering or drilling will create a large void area 

around the drain and the subsequent collapse will result in excessive soil disturbance. If the 

void could be filled with sand, a PVD is not necessary to begin with. 

Extremely soft layer for anchoring. In some cases, designers have selected a depth that does 

not fully penetrate the compressible layer. In such soil with a very low shear strength, it may 

become very difficult to anchor the drain at that depth, and either additional depth will be 

necessary or special equipment procedures will be required. This situation slows production 

and adversely impacts cost.

Site accessibility. While the equipment for PVD installation is relatively easy to transport, 

there are some situations where site accessibility may add to the cost. These include multiple 

overhead obstructions on a single project or steep access roads or site grades. In other cases, 

costly access roads may be necessary to transport the equipment down steep slopes or across 

unstable areas.

Overhead or subsurface utility obstructions. Usually underground utilities can be located 

prior to PVD installation, and drains can be installed around them to avoid any problems. 

However, large sewer pipes intermixed with several other utilities may create a situation 

where drains cannot be installed for a significant width. Overhead wires can possibly present 

more of a logistic problem. If the wires cannot be de-energized, significant widths of 

treatment might need to be eliminated, or the use of angled drains specified. 

Should any of the above site conditions be encountered, it would be advisable to contact 

specialty contractors experienced in the installation of PVDs in order to determine the 
2-15 

magnitude of difficulty. All of the above cases have been encountered and overcome to 

install drains successfully. However, the additional costs can be very significant.

2.4 Limitations

It is important to remember that a PVD serves no structural function (except perhaps in 

liquefaction reduction). By providing a shorter drainage path, PVDs provide a faster release 

of excess pore water, thereby resulting in faster settlement and quicker strength gain by 

consolidation. For sites with a stability problem, the soil will initially have the same strength 

with or without PVDs installed. Therefore, in situations where stability is of concern, the rate 

of increase of load must be carefully controlled (staged construction), monitored, or if the 

final or stage fill height is less than 12 feet, vacuum consolidation may be considered. 

PVDs used without a surcharge load can accelerate only primary consolidation, i.e., water 

being squeezed out of the soil. Therefore, it is important to estimate the magnitude and time 

rate of secondary consolidation. Secondary consolidation is caused by the soil particles 

reorienting or deforming under constant load, and is not dependent on water being squeezed 

out of the soil (FHWA 2006). Secondary settlement can be minimized by placement of 

excess surcharge and/or extension of waiting periods prior to final construction. Soils with 

high organic content are prone to significant secondary consolidation and should be carefully 

evaluated for the potential magnitude. 

Other limitations on the use of PVDs should be considered. PVDs have been installed at 

depths of up to 200 feet. However, 100 to 120 feet is a typical depth limit with specialized 

equipment required for depths greater than 120 feet. 

In most situations, the flow properties of a good quality PVD will not inhibit consolidation 

times. However, for extremely deep PVDs, combined with heavy loading and relatively high 

in situ soil permeability, the flow capacity of the system could be a limiting design 

consideration. In these rare cases, well resistance in the drain occurs, and consolidation time 

will be determined more by discharge capacity of the drain than by the horizontal 

permeability of the in situ soils. Rixner et al. (1986a) and Koerner (2012) contain specific 

technical guidance for this condition. 

PVDs are not recommended where the entire length or lower length requires pre-drilling. In 

these cases, sand drains or other methods of ground modification may be considered.
2-16 

2.5 Alternate Solutions

Alternate solutions can be functionally divided into four categories, as follows: 

1. Accept time constraints without the use of any vertical drain system.

2. By-pass the compressible soil, using deep foundation elements.

3. Reduce the compressibility of the in situ soil, using a different ground modification

method.

4. Reduce the loading on the compressible soil, using lightweight fill.

When there is sufficient time for settlement to occur under the final load conditions, PVDs 

obviously are not needed. In some cases, a preload without the use of any PVDs may be all 

that is necessary to obtain consolidation within the allowable time constraints. The cost of 

this preload should be compared against the cost of using PVDs with preload. 

An alternate method would be to design for excessive post-construction settlement and 

accept the anticipated cost of repairs or corrections to the ground or structure as a long-term 

maintenance responsibility.

Use of a deep foundation is an effective and expensive means of bypassing compressible 

soils. Deep foundations may be used to support a bridge or an embankment, using a column 

supported embankment (see Chapter 6). Such solutions are usually much more expensive and 

may also limit the flexibility of future uses of the site.

Reducing the compressibility of in situ soils offers the greatest variety of solutions. While 

usually more expensive than PVDs, the following are alternative methods of improving the 

soil:

• Stone columns. This method is used in soft subsurface soils to accelerate settlement,

reduce magnitude of settlement, and provide sufficient strength increase to preclude

deep-seated global failure. A variant using geotextile encased columns (GEC) may

also be considered for very soft soil profiles. (See Chapter 5).

• Deep soil mixing. This method is used to change the in situ compression and strength

characteristics of soils. (See Chapter 7).

• Excavation and Replacement. This is a method of removing the compressible soils

and replacing with a compacted, good quality fill material.

• Lightweight fills. This method is used to reduce the embankment load. (See Chapter

3). 
2-17 

Stone columns and deep soil mixing will strengthen the soil and reduce its compressibility. 

Where stability is a significant problem, combined solutions may be warranted in order to 

achieve the desired result. For example, the unstable areas can be treated with stone columns 

or deep soil mixing, and the remaining areas treated with PVDs, or excavation of weak in situ

soils and replacement with granular materials.

The use of lightweight fills to minimize the total amount of settlement has been used more 

frequently in recent years. While lightweight fills do not result in soil improvement, they 

reduce settlement and stability problems by reducing imposed loads. 
2-18 

3.0 CONSTRUCTION AND MATERIALS

3.1 Construction

3.1.1 Prefabricated Vertical Drains 

The various methods of prefabricated vertical drain installation, which employ mostly the 

same principles, are listed in Table 2-2. 

Table 2-2. Methods of Prefabricated Vertical Drain Installation 

Installation Method Method Options

Static

1. Chain Driven* 

2. Sprocket Driven* 

3. Cable Pull-down* 

4. Heavy Weight

5. Hydraulic Piston Push 

Vibratory 

1. Offset Hammer - full supported mandrel* 

2. Direct Hammer - offset mandrel*

3. Inside Mandrel with Enlarged Shoe 

Jetting 1. Covered Mandrel with Outside Jets 

2. Jet Probe with No Covering Mandrel 

System Combination

1. Static with Vibratory*

2. Static with Jetting*

3. Vibratory with Jetting

* Currently used in the United States.

With few exceptions, all methods in Table 2-2 employ a steel covering mandrel that protects 

the PVD material as it is installed. All methods employ some form of anchoring system to 

hold the drain in place while the mandrel is withdrawn. Another common feature is that the 

PVD material comes in rolls and is threaded through the mandrel in a variety of ways. The 

major difference between the listed methods is in the technique used to insert the mandrel 

into the ground. 

Commonly used methods employ an installation mast that contains the material reels, 

mandrel, and method of installation force. Added to this is a carrier, which is a crawler 

excavator or crawler crane, depending somewhat on the depth of installation. Specifications 

usually describe the acceptable method of installation(s) as static, vibratory, and/or staticvibratory. 

Typical installation equipment and carrier pieces are illustrated in Figure 2-3. 
2-19 

Apevibro.com (top), U.S. Patent Office (bottom)

Figure 2-3. Typical prefabricated vertical drain installation equipment: installation rig 

(top), box mandrel (bottom). 

The typical mandrel used in the United States has either a diamond shape of approximately 

5.5 inches by 2.4 inches or a rectangular shape with dimensions approximately 5 inches by 2

inches. In some cases, especially with vibratory applications, a stabilizing fin will be 

necessary. In most installation masts, the PVD material is fed through the mandrel from a 
2-20 

storage reel mounted near the base of the mast. The material travels up the inside of the mast, 

over a sheave at the top, and down through the steel mandrel. 

The sequence of installation operations starts when the steel mandrel has been threaded with 

PVD material and attached to an anchor at the bottom of the mandrel. The anchor typically 

consists of a 0.5-inch diameter piece of cable or rebar or a special anchor plate made of sheet 

metal. The rebar and cable anchors will be approximately 8 inches long, and the sheet metal 

anchor plates will be made from a thickness sufficiently thin enough to allow the excess area 

to fold up around the steel mandrel, minimizing its cross-sectional area to slightly in excess 

of the cross-sectional area of the steel mandrel. A small “handle” is spot welded to the sheet 

metal anchor for insertion of the PVD material. When the sheet metal anchor is used, the 

PVD material is inserted through the handle and reinserted in the bottom of the mandrel. 

Once the anchor is in place, the mast is positioned over the location of the drain, and the 

mandrel, along with the PVD material, is inserted into the soil. During installation, the drain 

is completely protected in the steel mandrel from damage due to obstructions in the soil. 

When the PVD has reached its proper depth, the mandrel is withdrawn, leaving the PVD 

material in the ground. The mandrel progresses upward until the bottom of the mandrel is 

above the ground level. At this stage, excess PVD material is pulled through the mandrel to 

allow for sufficient “cut-off” or “stickup” above the working surface, and the PVD material 

is cut using hedge shears or other devices, and attached to another anchor. This procedure is 

then repeated at succeeding drain locations. A typical description of this procedure suggests a 

giant sewing machine. In fact, some of the installation masts are actually called drain 

stitches. 

Except for unusual installation methods, where no mandrel is used at all, or where it is 

installed in sections, the length of the drain material must be slightly longer than the desired 

installation depth. Therefore, extremely deep drains (150 feet in depth) may require very 

large or specialized equipment. PVDs have also been installed from large barges on water 

and with amphibious marsh buggies. 

Once the PVD material on the reel has been depleted, another is attached by splicing to the 

previous roll. The common method of splicing is to insert the core at end of the old roll into 

the new roll a minimum length of 6 inches. Both the cores and geotextile must overlap to 

maintain continuity of flow channel and filtration, respectively. At this point, at least 10 

staples (4 on each side and 2 in the middle) will be used to hold the ends together. The splice 

is formed so that the bottom side of a vertical drain can be inserted into the upper end to 

ensure continuous flow. 
2-21 

Where obstructions are located above the compressible stratum and cannot be penetrated 

using normal procedures, pre-drilling or offsetting the drains will be necessary. Obstructions 

within the compressible stratum that cannot be penetrated using normal installation 

procedures can only be offset.

Normally, for PVD installation, a one-to-two-person labor crew is required to handle the 

drain preparation, cutoffs, changing of drain rolls, and anchor attachment. The only other 

labor required will be that to run the carrier piece and to perform occasional repairs to the 

equipment. The construction procedures are illustrated in Figures 2-4 and 2-5. 
2-22 

Courtesy Menard Group USA

Figure 2-4. Typical prefabricated vertical drain installation procedure: placing the 

anchor on the drain (top), inserting the mandrel into the ground (middle), cutting the 

drain after withdrawing the mandrel (bottom). 
2-23 

Courtesy Menard Group USA

Figure 2-5. Typical prefabricated vertical drain splicing procedure: inserting the drain 

core within the jacket to maintain continuity (top), stapling the drain splice (bottom). 
2-24 

3.1.1.1 Specialized Equipment 

While not commonly used, there are situations that will require very specialized equipment to 

install the PVDs. Situations where this might occur include the following:

• Unstable working surfaces

• Sloped surfaces

• Subsoils that are very difficult to penetrate

• Extremely deep drains (150 feet or more in depth)

Some examples of unusual equipment were previously discussed. These include the need for 

special carrier pieces for extremely deep drains and the use of marsh buggies on unstable 

working surface situations. Other examples include special external jetting techniques 

mounted on lightweight skid platforms for use on steep slopes and the use of test boring 

equipment to install pipe drains with PVD materials for construction of relief wells in 

existing dams.

Layers that might require special drilling techniques are fills containing large amounts of 

rubble, concrete, old slabs or footings, buried riprap or large boulders, and any cemented 

layers. Normally, if the soil can be augered for pre-loosening, its cost should be included in 

the PVD installation. However, if it is anticipated that obstruction drilling, as described 

above, will be necessary, a special obstruction of pre-drilling pay item should be established.

3.1.2 Staged Construction 

When embankments of moderate to large heights need to be constructed on soft foundation 

soils with low strength and high compressibility, excessive settlement and instability of the 

fill can occur if the fill is placed too quickly or too high. A solution to this problem is to build 

the fill in stages, or staged construction, such that the foundation soils consolidate and gain 

strength in between additions of fill material. The rate at which fill can be placed is related to 

the consolidation characteristics of the foundation soils. Monitoring of settlement and pore 

water pressures in the foundation soils allows prediction of when additional fill can be added 

without causing excessive settlements and/or slope instability. This process is generally used 

when time exists to allow the settlement to occur during the construction process. PVDs can 

be used with staged construction to speed up the consolidation process to reduce the waiting 

times between fill placements.
2-25 

3.2 Materials of Construction 

3.2.1 Prefabricated Vertical Drains

PVDs are relatively flat and approximately 4 inch wide by 1⁄8 to 3⁄8 inch thick. Types of PVDs 

are discussed in Section 1.2. The material generally consists of a plastic core formed to make 

channels and a loose geotextile cover. These two components are equally important in the 

function of a PVD.

The purpose of a core is to create low resistance flow channels in order for water to flow 

along the length of the PVD. In addition to providing the flow path, the core maintains the 

drain configuration and shape, provides support for the filter jacket, and provides the tensile 

and compressive strength of the drain. It is important that the core have a certain amount of 

strength and flexibility. 

The function of the geotextile is to provide a surface that inhibits soil particles from 

penetrating into the core channels, while allowing passage of water into the drain. Its 

secondary purpose is to prevent closure of the interior core channels and to form the outside 

of the flow channels of the core. 

It is important that the apparent opening size (AOS) of the geotextile be such that it allows 

only a few particles to penetrate but will not allow sufficient soil particles to clog the core. 

Experience has shown that a geotextile having an AOS in the range of 0.15-0.074 mm sieve 

size (U.S. #100 to #200) will be effective. Geotextiles within this AOS range have proven to 

be very successful for all projects.

The selection of the PVD type to be used is an important part of the overall design process, 

and the following information is intended to aid in this selection process. When selecting a 

PVD, the primary parameters include jacket filter permeability and characteristics, material 

strength, flexibility, durability, discharge capacity (qw), drain resistance, and equivalent 

diameter (dw). The filter should be designed to either prevent soil particles from passing 

through, or allow formation of a natural filter cake. The filter needs to resist tearing during 

installation, and the filter characteristics need to be maintained for the duration of the 

consolidation period. The discharge capacity of the drain under lateral stresses needs to be 

considered. Guidance is provided in Rixner et al. (1986a) and Chu et al. (2004). Buckling 

and crimping of the PVD is also a concern. It is recommended to assume a conservative 

maximum discharge capacity of 3,500 ft3/year. Koerner (2012) also provides general 

guidance on geotextile filter and discharge characteristics. 

Some procedures argue that the geotextile openings should be small enough to prevent soil 

particles from passing through and causing “siltation” and reducing discharge capacity. Other 
2-26 

procedures believe that the openings should be large enough to allow for the formation of a 

natural filter. The well capacity used in design will depend on the volume of the drain 

opening, lateral earth pressure, folding and crimping of the drain, and the infiltration of fines 

into the PVD. Design criteria are presented in Holtz and Christopher (1987) for 

critical/severe conditions and for less critical/severe situations, the former having a more 

stringent design process. Projects with critical conditions include those where PVD failure 

will result in loss of life, significant structural damage, and/or the cost of repair exceeds 

initial installation cost. Projects with severe conditions include gap graded soils, high 

hydraulic gradients, dynamic flow, or reversing flow conditions. 

Soil disturbance is most dependent on the mandrel size and shape, the soil macrofabric, and 

installation procedure. Soil disturbance can slow the rate of consolidation. Therefore, it is 

critical to consider the effects of a smear zone which results from soil disturbance, which can 

be minimized through proper mandrel selection. 

3.2.2 Drainage Layer 

The second component of a PVD installation (or any vertical drain project) is the drainage 

layer or method of conveyance and discharge of water from the drains. In most cases, this is 

accomplished using a drainage blanket consisting of a granular material. Synthetic drains, 

often called strip drains, can be laid horizontally along the installation surface and connected 

to each individual PVD to provide a clear drainage path for the pore water to the atmosphere, 

without creating any head loss. Often the strip drains are outletted in gravel ditch drains at 

the edge of the embankment. 

Where sand or gravel is used, the drainage blanket can also serve as a working platform to 

help support equipment used for PVD installation. Ideally, sand should be clean and washed, 

such as concrete sand with the fraction minus the #200 sieve being less than 3 to 4 percent. 

The design of the drainage layer is based on its ability to transport the excess pore water 

without any head loss. If sand is used, it should have a minimum thickness of at least 1 foot, 

but more typically it is 1.5 foot to 3 feet. Where it is used directly over soft soils, typical 

designs require at least 1 foot of extra thickness because of potential contamination, or use of 

a geotextile separator (FHWA 2008). Where there is a possibility of mud waving, as much as 

a 3 foot thickness has been used. If gravel is used, it should also be very clean. Because of its 

permeability, gravel can be limited to a thickness of 8 inches, with geotextiles on both sides 

to keep it from being contaminated by intrusion from the lower soils or the upper 

embankment materials.

Where strip drain material is to be used in lieu of a drainage layer, there are several different 

techniques that can be utilized. Typically, the strip drain material is approximately 1 inch in 
2-27 

thickness and either 0.5 or 1 foot in width. Depending on the flow rate of the individual strip 

drain and the length and number of drains attached to each strip drain, either the smaller or 

larger size may be necessary. Typically, a strip drain is attached to every individual drain, 

which means one horizontal strip drain for every row of PVDs. Occasionally; a single strip

drain has been used for as many as three different rows, which requires that each PVD be 

connected to each strip drain. An example of a strip drain application is shown in Figure 

2-6. 

Courtesy Menard Group USA

Figure 2-6. Strip drains.
2-28 

Consideration must be given to stability of the working surface. Often the thickness of the 

granular blanket must be increased to allow for support of the PVD installation equipment. 

An alternative is to reinforce the drainage blanket with geotextiles and/or geogrids. This may 

have a twofold effect: to provide a stable working surface, and to minimize the necessary 

thickness of the drainage layer due to contamination from the soils below. A combination of 

a working platform and drainage layer is often cost effective. 

If PVDs are installed on uneven surfaces, such as on the sides of an existing embankment, 

the drainage layer effectiveness and stability must be considered. The working surface may 

have to be altered to allow for the installation of PVDs, such as a benching procedure that 

may disrupt the continuity of the drainage blanket. To ensure proper functioning, the 

drainage layer must be outletted.
2-29 

4.0 DESIGN 

4.1 Design Considerations 

The principal objective of soil preloading, with or without the use of vertical drains, is to 

achieve a desired degree of settlement in the site soils within a specified period of time. The 

design of vertical drains is primarily based on the use of fill or surcharge to preload the site 

soils and PVDs to reduce the length of the drainage path and, thereby, accelerate 

consolidation of the site soils. Vacuum consolidation (see GeoTechTools) may also be used 

as surcharge through an increase in the effective stress in the site soils. Increasingly, PVDs 

are being used for specialized applications such as liquefaction potential reduction where 

specialized knowledge is required. Such an application is beyond the scope of this technical 

summary and interested readers should see Rollins et al. (2004) and Rollins and Strand 

(2007). 

Preloading site soils to bring about consolidation requires time for the dissipation of the 

excess pore pressures developed and for settlement of the soils to occur. The use of PVDs 

shortens the drainage path and speeds up the settlement over the case of surcharging without 

vertical drains. The proper design of a PVD installation requires knowledge of the type and 

extent of the foundation soils and their pertinent engineering properties. Engineering analyses 

must include, among other items, predictions of the amount and the rate of settlement, both 

during and after construction, and the embankment stability during all phases of construction. 

For consolidation analyses, the subsurface investigation program must define the extent and 

depth of the compressible strata and secure high quality undisturbed samples to determine 

past maximum pressures, coefficients of compressibility, and coefficients of consolidation, 

both in a vertical and horizontal direction. In addition, Standard Penetration Test (SPT) blow 

counts (N60) is also significant in determining the installation costs and potential for the need 

for pre-drilling prior to placement of the PVDs. Piezocone penetration testing (PCPT or 

CPTu) is an excellent complement to SPT measurements, as it is fast, provides continuous 

profiling, and is particularly suitable for soft soils. For detailed discussions on SPT and PCPT 

the reader is referred to FHWA (2002). 

The design of vertical drain systems begins with traditional settlement analyses to determine 

the total and the time rate of settlement under final project loads without the use of drainage 

measures. Settlement analysis of cohesive soils is conducted using Terzaghi’s onedimensional consolidation theory and many textbooks detail the procedures to follow (e.g., 

see Holtz et al. 2013) as well as design manuals (see NAVFAC Design Manual 7.1). 

The design of all vertical drains systems is based upon consolidation theories. Vertical drains 

speed up the consolidation of a clay layer by introducing a radial drainage pathway into the 
2-30 

soil strata that shortens the water pathway compared to vertical consolidation only. Although 

the use of vertical drains entails two-dimensional consolidation in the vertical and horizontal 

directions, generally only the horizontal consolidation is considered in practical cases, as the 

coefficient of consolidation in the horizontal direction (ch) is usually greater than the 

coefficient of consolidation in the vertical direction (cv) and the drainage path in the 

horizontal direction is much shorter than in the vertical direction. Thus the problem reduces 

to one-dimensional radial drainage. Barron (1948) developed the first radial drainage solution 

in English for vertical sand drains, and this remains the basis for vertical drain design today. 

Barron’s work was based on the simplifying assumptions of Terzaghi’s one-dimensional 

consolidation theory that the soil is homogeneous and completely saturated, the water and 

soil grains are incompressible, the flow of water and soil compression are in one direction 

only, Darcy’s Law for water flow is valid, and the permeability and compressibility of the 

soil are constant over the increment of stress. Barron (1948) proposed the use of a triangular 

drain pattern (Figure 2-7[top left]) while Kjellman (1948) proposed a square pattern (Figure 

2-7[top right]). In either case, the boundary conditions of the problem refer to an equivalent 

soil cylinder of diameter, dc, having an impermeable outside surface and an inner cylindrical 

drain as shown in Figure 2-7(bottom) (Holtz et al. 1991). 
2-31 

Holtz et al. 1991

Figure 2-7. Drain patterns: triangular pattern (top left), square pattern (top right), 

equivalent cylinder (bottom). 

Solutions to the radial consolidation problem are available in a number of publications 

including Barron (1948), Hansbo (1979, 1981, and 2004), Rixner et al. (1986a), Holtz et al. 

(1991), and Han (2015); and reference to these is made for more complete coverage of the 

theories. The solution of Rixner et al. (1986a) is used in the next section to detail the design 

procedure. 

4.2 Design Procedure

The design of a PVD system consists of the selection of the type, spacing, and length of the 

drains to accomplish the required degree of consolidation within a specified time. The design 

process begins with the development of settlement analyses without PVDs to determine the 

total magnitude of settlement and the time rate of that settlement under the expected final 

embankment load. If the time to reach 90 to 95 percent of these projected settlements is too 
2-32 

great and beyond the allowable contract construction time, the use of PVDs is a design 

solution to reduce the time for the settlement to occur. 

The assumptions used in developing one dimensional consolidation theory have been applied 

to the development of radial drainage theory related to vertical drains, which resulted in the 

following relationship between time, drain diameter, spacing, coefficient of consolidation 

and the average degree of desired consolidation (Rixner et al. 1986a). 

[ ] 



 





− = ++

h

rs

h

c

U

FFnF

c

d

t

1

1 )( ln

8

2

[Eq. 2-1] 

where:

t = time required to achieve desired average degree of consolidation 

Uh = average degree of consolidation due to horizontal drainage

dc = diameter of the cylinder of influence of the drain (drain influence 

zone) 

ch = coefficient of consolidation for horizontal drainage 

F(n) = drain spacing factor

Fs = Soil disturbance factor (smear zone)

Fr = Well resistance factor

The drain spacing factor F(n) is determined as follows: 

F(n) ln 0.75 ( ) c

w

d simplified d

  = −    

[Eq. 2-2] 

where: 

dw = diameter of an equivalent circular drain

Note that Equation 2-1 does not consider consolidation from vertical drainage. 

The following discussion of the components of Equation 2-1 should provide an 

understanding of the factors that can affect the design of PVD systems. 
2-33 

t 

The time to achieve the required degree of consolidation ( ) for a pattern and equivalent 

spacing dc, and drain diameter dw. 

 

Normally, an average degree of consolidation of 90 to 95 percent is desired and, it is a 

function of the magnitude of post-construction settlement that the project can tolerate. It 

should be further understood that the contribution of vertical consolidation may be significant 

and should be considered for major/complex projects. Often it may approach 30 percent of 

the measured settlement. 

dc

The diameter of the cylinder of influence of each PVD. When using an equilateral triangular 

pattern, dc is 1.05 times the spacing between each drain. In a square pattern, dc is 1.13 times 

the spacing between drains. Typically, to achieve approximately 90 percent consolidation in 

3 to 4 months, designers often chose drain spacing between 3 feet to 5 feet in homogenous 

clays, 4 feet to 6 feet in silty clays, and 5 feet to 6.5 feet in coarser soils.

ch

The horizontal coefficient of consolidation (ch) of a particular layer can be obtained from 

laboratory consolidation tests. Even with proper laboratory techniques and high-quality 

undisturbed samples, the designer is fortunate to be within 50 percent of the actual 

coefficient of consolidation. Normally, only the coefficient of vertical consolidation (cv) is 

obtained from standard consolidation tests from which the horizontal coefficient is estimated. 

High quality laboratory tests have consistently shown ch to be greater than cv, the coefficient 

of consolidation in the vertical direction. A common conservative approach is to assume ch is 

directly related to cv, without direct measurement values. For design, ch is generally taken as 

1.2 to 1.5 cv, if no or only slight evidence of layering is evident upon examination of the 

partially dried clay sample. If layering of silt and sand in discontinuous lenses is evident, 

ratios of 2 to 4 are indicated. For varved clays and other deposits containing embedded and 

more or less continuous permeable layers, ratios of up to 10 can be considered. The extent of 

layering in compressible deposits is best assessed by field PCPT testing. Additional guidance 

on the determination of ch can be found in Rixner et al. (1986a). Relationships between the 

coefficient of consolidation in the vertical direction, cv, and the liquid limit, can be used for 

preliminary design purposes (as shown in Figure 2-10 later in this section). 
2-34 

The most accurate assessment of the average ch can be made from settlement data obtained 

from a test fill or from data obtained from the first stage in stage construction by using the 

Asaoka method (Asaoka 1978) shown in Figure 2-8 and discussed in Bartlett et al. (2001). 

Bartlett et al. 2001

Figure 2-8. Asaoka (1978) method for determining ch and end of primary consolidation.

Equation 2-1 does not take into consideration any consolidation from vertical drainage. The 

vertical drainage effect may be minor when the compressible layer is over 15 feet in 

thickness, and for a conservative approach, no beneficial effects may be considered for 

vertical drainage in preliminary feasibility evaluations.

When several compressible layers of varying soil properties are encountered, care must be 

taken to evaluate each layer and its effect on the total consolidation. Settlement and time rate 

of settlement computations can be performed on each layer with fixed time and spacing to 

determine if sufficient consolidation will occur within a given time constraint. For a very 

conservative approach, the lowest ch can be used to design the spacing. However, the 

magnitude of settlement should still be computed for each layer. 
2-35 

dw

Hansbo (1979) suggested that when a band-shaped drain (PVD) is used, the equivalent 

diameter, dw, should be a cylinder of having the same circumference as given by Equation 

2-3 and shown in Figure 2-9. The PVD thickness is “a” and the PVD width is “b.” 

w [ ] += bad /)(2 π [Eq. 2-3] 

Holtz et al. 1991

Figure 2-9. Typical cross-section of a band-shaped drain and mandrel. 

There have been varying methods and recommendations to determine the equivalent circular 

drain size for a PVD. Various diameters ranging from 1.5 inch to 5.5 inch have been used, 

but the most common is to use 2.5 inch. In fact, there is little difference in design if 2.0 inch 

to 4.0 inch is used. 

Fs, Fr

The above basic relationship can be modified further to consider the effects related to soil 

disturbance and well resistance. The effects of both soil disturbance and well resistance are 

often ignored for typical projects. Chu et al. (2004) provide guidance for when Fs and Fr

should be considered in the design calculations. Additional information can also be found in 

Rixner et al. (1986a) and Holtz et al. (1991). 

Considerations for Using Equation 2.1 

Sample disturbance is typically ignored except for design in soils that are highly plastic, 

sensitive, and where the coefficient of consolidation has been accurately determined. Under 

these conditions, a sample disturbance factor Fs ≈ 2 may be considered. Note that the effects 

of sample disturbance are more pronounced at drain spacing of less than 5 feet and by the 

contractor’s use of large, thick anchor plates. 
2-36 

Well resistance is rarely significant, and, therefore, ignored except for extremely deep drains 

or where a combination of high loads and very permeable soils is present. For guidance in 

these situations, the applicable reference is Hansbo (1979). 

There are three basic variables that can be manipulated in order to achieve a desired result 

from Equation 2-1. These variables are time, PVD spacing, and surcharge. If the surcharge is 

increased, yielding greater settlement, then the PVD spacing can be increased in order to 

achieve the same amount of settlement in a given time (Fellenious and Altace 1999). 

Another approach is to add surcharge in order to decrease the time required for settlement for 

the same spacing. Time can also be used as a variable, affecting the amount of surcharge or 

the PVD spacing. Using these variables, the designer can consider the cost of each, in order 

to determine the most economical solution. 

Usually, time is a constant, with the designer varying spacing or surcharge in order to 

achieve the desired results.

For consolidation of organic soils, additional surcharge must be designed to obtain the 

required settlement attributed to secondary compression within the available time, i.e., 

overconsolidate the soil for the design embankment load. 

4.3 Preliminary Spacing Design

Preliminary determination of PVD spacing using soil index data, such as liquid limits and 

project geometry, can be made. Project geometry reflects the area of loading, depth of soft 

compressible strata, and whether excess surcharge will be necessary. An approximate 

coefficient of consolidation, cv, can be estimated from liquid limit values using the graph 

shown in Figure 2-10. 
2-37 

Department of the Navy 1982

Figure 2-10. Coefficient of consolidation versus liquid limit. 

The center curve represents the case of virgin compression associated with good quality 

undisturbed samples while the lower curve is indicative of results from disturbed samples. 

The upper curve represents cv for samples above the preconsolidation stress. Additional 

guidance on the selection of cv from correlations and laboratory tests can be found in GEC 5 

– Evaluation of Soil and Rock Properties. Direct determination of the horizontal coefficient 

of consolidation, ch, is possible using piezocone dissipation test results. The procedure is 

detailed in GEC 5. 

Assuming that the desired percentage of primary consolidation has been established or is in 

the 90 to 95 percent range, the designer can then either use computer programs available for 

vertical drain design, or solve Equation 2-1 to determine a preliminary spacing consistent 

with available time. A number of computer programs have been developed that perform such 
2-38 

design procedures including FoSSA-Foundation Stress & Settlement Analysis (Adama 

Engineering 2003). Once the preliminary spacing is determined, the total project quantities 

can be computed from the project geometry.

4.4 Design Example 

A highway embankment is planned at a location where available borings indicate a profile of 

20 feet of normally consolidated clay with sand lenses over rock. Total settlement, based on 

one-dimensional drainage is estimated at 12 inches, with 90 percent occurring in 10 years. 

This estimate is based on preliminary evaluations of compression and consolidation 

coefficients obtained from classification data and liquid limit of 60.

The available construction time is on the order of 18 months; hence it is desirable to 

accelerate settlements by the use of PVDs to a point at which 90 percent of the total 

settlements would occur within about one year, without a surcharge load. To estimate 

feasibility and cost, a potential spacing must be determined. 

From Figure 2-10, using the center curve, an estimated coefficient of consolidation (cv) of 0.1 

ft2/day is obtained, and based on the existence of horizontal sand lenses in the profile, a 

coefficient of horizontal consolidation (ch) of 2 times cv can be assumed. Equation 2-1 can be 

solved for various drain spacings, using an equivalent diameter of drain of 2.5 inches to 

obtain the field time required to attain 90 percent consolidation. The solution using triangular 

spacing indicates required times on the order of 300 days at an approximate spacing of 8 feet 

or 500 days at an approximate spacing of 10 feet.

Note that this feasibility design method is only suitable for preliminary design or a quick 

check of a detailed design to determine if there are any errors. Since feasibility design is 

solely predicated on the liquid limit determined, it must be representative of the compressible 

stratum.
2-39 

5.0 CONSTRUCTION SPECIFICATIONS AND CONSTRUCTION CONTROL

5.1 Introduction

PVD construction equipment, methods and materials, are well established with a proven 

track record of many successfully completed projects in the last several decades. The design 

concepts are well understood and the accuracy of the design is almost directly related to the 

accurate determination of the horizontal coefficient of consolidation. Therefore, method 

specifications are almost always used to implement this ground modification technology.

5.2 Specification Development

The role of good specifications is to reduce the problems, risks, and cost of PVD installations 

and to ensure an installation that achieves the intended objective of the design. To 

accomplish the above, agencies usually specify the general construction method and controls 

for PVD installation. Under this contracting method, the designer details the spacing, the 

extent and length of the drains, as well as the details of the drainage blanket (usually 

specified separately under earthwork items), and the horizontal discharge drains. End result 

specifications are not common since they would require extensive soil exploration and design 

by the contractor.

The specifications should provide the PVD installer as much flexibility as possible in 

achieving the intended result. The specifications should include quality control and assurance 

procedures. Initial installation trial drains should be required to establish standard 

procedures. At this stage of construction, such items as mandrel size, depth gauges, splicing 

procedures, verticality, and materials should be carefully inspected for compliance. Visual 

observations and periodic checks can then determine any variances or concerns once 

production has begun. 

Inspection personnel should check the physical measurement of drain sizes (thickness and 

width) for compliance to specifications and variances from material submittals. If significant 

differences are noted, laboratory testing may be required to determine if PVD materials 

comply with specifications.

As part of the development of GeoTechTools, an extensive evaluation was made of 

specifications for PVDs. Eight specifications written by state DOTs, as well as the guide 

specification presented in the 2006 version of this manual, were reviewed and evaluated. 

These specifications were used to develop a guide specification entitled Standard Method 

Approach Specification for Prefabricated Vertical Drains and Fill Preloading that is intended 

to be a complete and fair specification containing commentary and instructions that is easily 

adaptable by the user for a specific project. This guide specification can be accessed on the 
2-40 

GeoTechTools website under the Prefabricated Vertical Drains and Fill Preloading 

Technology Information page. An additional specification resource is the recently published 

Wick Drain Guide Specification (Method), prepared by the Ground Improvement Committee 

of the Deep Foundations Institute (DFI 2014). This specification can be found by accessing 

the Ground Improvement Committee webpage on the DFI website (http://www.dfi.org/). 

Regardless of which PVD specification is used, it is recommended that the Hansbo (1986), 

Rixner et al. (1986a), Holtz and Christopher (1987), and Bo et al. (2005) references be 

reviewed to ensure that all PVD material and installation requirements are included. 

5.3 Quality Assurance 

One major advantage of the PVDs over other vertical drainage systems is the simplicity of 

field control. Once trial drains have been satisfactorily completed, inspection mainly consists 

of recording depths and locations of each drain, observing splices and verticality of 

equipment, taking occasional material samples for inspection and testing, and noting any 

major variances in procedure. 

To ensure proper performance, the drains must be installed in accordance with the plans and 

specifications. It is important that field inspection personnel know the procedures and 

possible ramifications of any deviations.

The construction monitoring personnel should be thoroughly familiar with the contract 

drawings and specifications and should have a good understanding of the purpose of PVDs. 

The engineer in charge should have a good understanding of the total PVD solution, 

including site preparation, fill placement, and other items that might influence the 

performance of PVDs. Inspection of the PVD installation after initial procedures are 

established can become quite repetitive and monotonous. However, when changes or 

variations occur, these should be discussed immediately with personnel familiar with the 

project design requirements. 

Other items of significance in construction, such as the drainage blanket material quality, 

embankment placement and compaction, and surcharge loading rate, must also be monitored. 

5.3.1 Site Preparation

Site preparation includes any excavation and grading to prepare the site for installation of 

PVDs. This may include site clearing, excavation and/or filling operations to bring the site to 

grade, and construction of a working platform or drainage blanket. It is also important that 

the grade be such that a drainage blanket remains continuous and surface drainage will not 

erode the blanket to the extent where it is no longer continuous. 
2-41 

During construction of a working platform or drainage blanket, the field inspection personnel 

should be monitoring for any unusual soil movements, which would be indications of mud 

waving or a potential failure. On some projects, working mats or drainage blankets were 

installed too thick, creating failures, which were very expensive to repair. In one example 

project, a contractor installed a 5 feet haul road over a 2 feet drainage blanket prior to PVD 

installation in order to have access to a piling location. This haul road caused an extensive 

failure of the entire area that was to be stabilized. 

5.3.2 Prefabricated Vertical Drain Material and Installation Equipment

Once the equipment and materials are checked for compliance with the specifications, the 

remaining field observations are rather simple. The major items to be monitored are 

verticality, depth of installation, and location. Most projects will provide an estimated depth 

or elevation across the site. It is routine to anticipate local variations from this estimate. 

However, the depth of PVDs should vary only slightly from that of adjacent drains. The 

inspector should refer to pertinent borings for a guideline to determine if proper depths are 

achieved or if there is a significant variance in soil conditions.

5.3.2.1 Material

Prior to installation, the PVD material should be visually observed to ensure that it is similar 

to that originally submitted and tested. The core and filter jacket should be continuous and 

comply with required dimensions, and the materials should not have been damaged during 

handling or storage. 

Many specifications list intricate tests to be performed on the PVD materials. Many of these 

tests are taken from manufacturer's data and quite often are far beyond what is necessary. 

However, an individual project may have specific criteria, which are dependent on some 

stringent design or performance requirement.

5.3.2.2 Equipment 

Inspection personnel should determine that the equipment complies with specification 

requirements. Some of the important items to be checked are the following: 

• Penetration method

• Mandrel shape, size, and stiffness

• Anchor type and size

• Method to measure and determine penetration depth
2-42 

• Method to measure and record installation force

• Means and procedures for pre-drilling, where necessary

Installation of trial drains to evaluate the installation equipment is recommended for all 

projects. A representative of the owner and inspection personnel should be present during the 

trial drain installation.

Variations in installation procedures, which might be necessary to penetrate to the required 

depth, should be evaluated during the trial program. Obstructions encountered are usually 

offset where possible or predrilled or removed if they encompass a significant area. 

5.3.2.3 Submittals

Most PVD specifications require several submittals for approval. Submittals include the type 

of PVD material, the material specification sheet, and the source. In addition, a method and 

sequence outlining the installation procedure is often required. The material submittal quite 

often is either related to specific specification requirements or manufacturers’ tests.

The installation submittals should address the following: 

• Size, type, weight, maximum pushing force, vibratory hammer rated energy, and

configurations of the installation rig

• Dimensions and length of mandrel

• Details of PVD anchorage

• Detailed description of proposed installation procedures

• Proposed method for splicing drains

A common additional requirement is the installer's experience. Many specifications require a 

minimum of three successful projects. This can be typically fulfilled either by the installation 

contractor's project experience or the project personnel’s past experience.

Specifications should not require the delivery of PVD materials prior to the arrival of the 

installation equipment, as this requirement would entail two mobilizations by a specialty 

contractor. Instead, material certification should be required from the manufacturer in 

advance of mobilization. 

5.4 Instrumentation Monitoring and Construction Control

Field instrumentation, such as piezometers, settlement platform and gauges, and 

inclinometers, are used to monitor performance of the PVDs and possibly control the rate of 
2-43 

construction of embankment and/or surcharge. It is important that both the designer and the 

instrumentation personnel have a full understanding of the particular piece of instrumentation 

being installed and the strata in which the instruments are being placed.

Settlement measuring devices, whether platforms, deep settlement points, or horizontal 

deflection devices, are used to measure only the rate and total amount of consolidation. 

Inclinometers are used to measure horizontal deflection with depth and to warn against 

potential stability failure. The pore pressure devices (piezometers) are used for both 

calculation of achieved consolidation rate and to quantify excessive build-up of pore pressure 

that is an indication of potential failure. One caution concerning pore pressure devices is that 

there have been a significant number of projects where the rate of settlement has not agreed 

with the rate of pore pressure dissipation. In such situations, settlement data should be given 

priority as indicators of the rate of consolidation. 

The proper selection of instrumentation devices and the frequency of monitoring during a 

project are important. For simple projects where stability is of no concern, and time is not the 

critical factor, only surface settlement platforms, which are relatively easy to install, are used. 

In situations where stability is critical, pore pressure measurements and measurements of 

horizontal deformations are also necessary. Where stability is of concern daily readings may 

be necessary both during loading and for the first few weeks after loading. 

The inclinometers should be installed at the toe of the embankment or in front of retaining 

walls, with settlement plates and piezometers beyond the crest of the embankment and/or 

near the centerline. For projects where stability is of concern, the key tool in evaluating 

stability is the measurement of the displacement ratio, DR, defined as the maximum 

cumulative horizontal displacement divided by the cumulative settlement determined from an 

adjacent settlement plate. Significant increases of displacement ratio with time or load can 

indicate when the foundation behavior is changing from a relatively stable condition 

associated with consolidation settlement to a state of excessive plastic deformation. Ladd 

suggested that when fill is placed relatively rapidly, a DR approaching 0.4 during filling may 

be measured. This appears to be the upper limit for a stable foundation in which vertical 

drains have been installed (Ladd 1991). Displacement ratios of 0.2 generally correlate with a 

slope stability FS of 1.3 or greater. 

For stable foundations DR should decrease after the completion of filling. A typical plot of 

DR vs time as developed for a multi-stage fill near the Maryland abutment of the Woodrow 

Wilson Bridge is shown in Figure 2-11. 
2-44 

Figure 2-11. Displacement ratio versus time and stage filling. 
2-45 

In addition to monitoring displacement ratios, the inclinometer data should be plotted as 

incremental displacements as a function of time and depth. The incremental displacements 

will clearly show the depth at which horizontal strain is occurring and, therefore, the location 

of potential failure plane(s). The maximum horizontal displacement can be related to soil 

strain as well, as a further check on stability. Cumulative and incremental horizontal 

displacements, at the location where the displacement ratios in Figure 2-11 have been 

computed, are shown in Figures 2-12 and 2-13. 

End of primary consolidation (EOP) can be obtained by the Asaoka method of analysis from 

settlement data. This method consists of plotting settlement on a time versus settlement plot 

of equal time steps as shown in Figure 2-8. Values of cumulative settlement at the start of a 

time step are plotted with respect to the settlement at the beginning of the next time step. 

When the settlement has not changed during a time step, primary consolidation is essentially 

complete. Details of the analysis and limitations are discussed by Bartlett et al. (2001).

It is usually best to install settlement platforms and other instrumentation after the installation 

of the horizontal drainage layer and PVDs. This allows for an unimpeded site during PVD 

installation and the ability to locate the most critical areas for instrumentation, especially for 

piezometers. If piezometers are installed prior to PVD installation, it often becomes difficult 

to locate the drain at equal distance from the piezometers’ locations, especially where 

multiple piezometers are installed at different elevations. On some projects it might be 

valuable to install some instrumentation prior to PVD installation. The resultant information 

may be useful for interpretation in critical situations. When stability is a major concern, slope 

indicators and some piezometers and settlement devices should be installed prior to PVD 

installation. A typical layout with a complete set of instrumentation is shown in Figure 2-14. 
2-46 

Figure 2-12. Cumulative displacement versus depth.
2-47 

Figure 2-13. Incremental displacement versus depth.
2-48 

Figure 2-14. PVD Installation showing typical instrumentation and monitoring.
2-49 

6.0 COST DATA

6.1 Cost Components

Often when estimating the cost of a PVD project, only the unit cost of the installed PVDs is 

considered, rather than the total cost including other ancillary items. Other factors affecting 

the total cost of the PVD solution include the following: 

• Project Size, Topography

• Obstructions, Dense Soils

• Installation Methods

• Allowable Construction and Consolidation Time

• Allowable Post-construction Settlement

• Preload and Surcharge

o Type and Material Available

o Reuse of Material

o Amount of Surcharge

• Drainage Blanket or Horizontal Drainage Path

• Design, Instrumentation and Monitoring

• Unit Cost of Installation

Note that the unit cost of installation, whether including the mobilization or not, is only one 

of many factors that affects the total cost of a PVD solution.

Some of the above factors are very difficult to quantify, and their costs may be included in 

other bid items. The following discussion will focus on all cost items since they have an 

important impact on the total cost. 

• Project Topography: The project topography can play a role in the cost of a PVD

solution. For example, it is very difficult, if not impossible, to install PVDs on

significant slopes. If there is a great variation in grade on the site, significant

earthwork may be necessary to provide a somewhat level working surface for

installation.

• Obstructions, Dense Soils. The cost of PVDs increases significantly where soils are

difficult to penetrate. Therefore, it is important to correctly identify the soils to be

penetrated and the need to achieve specified depths. Depending on methods of
2-50 

installation, the designer should be concerned with very stiff layers (generally N 

values of 15 or greater), or possible obstructions, such as boulders, rocks, previous 

foundations, known underground utilities, etc. 

• Installation Methods. The installer should be given latitude in choosing the proper 

type of equipment for installation. This will ensure the most economical solution 

when difficulties arise. A specification that allows static, static vibro, and vibro-type 

installation usually results in the most economical solution. Jetting should be allowed 

only with the approval of the engineer and where it can be shown not to have an 

environmental impact. Impact methods should not be allowed, except for pre-drilling 

through non-compressible soils. 

• Consolidation Time. Time required includes the estimated consolidation time, plus 

time for installation of the PVDs and for construction of the embankment and/or 

surcharge. Additional factors that affect time are the placement of instrumentation, 

site preparation, and drainage blanket installation. In general, the total time is the least 

variable factor because of project constraints.

• Preload and Surcharge. Preload is defined as the amount of fill material necessary to 

bring the site to the final elevation. This includes the amount of additional 

embankment necessary to accommodate the final amount of settlement. Surcharge is 

the added fill above final elevations that is used to accelerate settlement or minimize 

secondary consolidation. In many cases, the maximum amount of surcharge is 

controlled by the potential for slope instability.

Time, surcharge magnitude, and PVD spacing are the significant variables in the total 

cost, and these variables can be optimized in the design stage to determine the lowest 

total cost. If time is the critical factor, either the PVD spacing must be closer together 

or additional surcharge must be placed in order to achieve the desired settlement 

within the specified time.

• Drainage Blanket or Horizontal Drainage Path. The availability and cost of sand, or 

other granular material, can often determine whether a granular blanket will be more 

economical than a geosynthetic strip drain solution. Quite often the granular blanket 

may be necessary for a working platform, even though it would be more expensive 

than the strip drain solution. The actual cost of a granular blanket is the difference 

between the granular blanket material and the local embankment material, whereas 

the strip drain material will be a totally added cost.

• Design, Instrumentation, and Monitoring. These items depend on the complexity of 

each project, but they may be a significant cost, especially if staged construction is 

required. 
2-51 

6.2 Summary of Typical Projects 

Typical cost of PVD installations can be divided into three categories: small – up to 50,000 

lineal feet, medium – 50,000 to 300,000 lineal feet, and large – greater than 300,000 lineal 

feet.

Typical unit price ranges are listed in Table 2-3 for projects where the soils do not present 

major difficulty in penetration, do not require special equipment or are not at unusually 

difficult sites. 

Table 2-3. Typical Unit Price Ranges for PVDs 

Site Category Unit Price Range

Small $0.70 to $4.00 per lineal foot

Medium $0.50 to $1.00 per lineal foot

Large $0.30 to $0.50 per lineal foot

Usually added to these costs is a mobilization charge of $15,000 to $25,000 per rig, as well 

as the cost of the drainage blanket and instrumentation. The unit cost could be significantly 

higher in areas of severe weather, labor shortages, or difficult site conditions, and where 

there is difficulties with installation of the PVDs. 
2-52 

7.0 CASE HISTORIES

The following three case histories illustrate successful projects.

7.1 Council Bluffs (IA) Interstate System Improvement Program

7.1.1 Project Description

The Council Bluffs Interstate System Improvement Program was a comprehensive interstate 

redesign that modernized the highway system and improved mobility and safety of 

approximately 18 miles of interstate. The I-29/US 275/IA 92 (I-29/US 275) Interchange 

reconstruction was completed between 2014 and 2016. The project consisted of removal of 

the existing bridges and portions of the existing roadway embankments and the construction 

of new roadway embankments, walking trail, bridge structures and other features. Six bridges 

were reconstructed and roadway improvements included raising the highway grades and 

expanding the lanes. Construction also included the consolidation of railroad operations into 

a new common corridor. 

7.1.2 Subsurface Conditions

The site is located in the Missouri River valley floodplain. The subsoils consisted of silty 

clay and sandy silt overlying poorly graded fine to coarse grained sand of medium density. 

The upper 15 feet of the silty clay consisted of interbedded lean clay (CL) and fat clay (CH) 

with SPT blow counts (N60-values) ranging from 4-10. The lower portion of the silty clay 

was approximately 20 feet thick and consisted of fat clay (CH) with SPT blow counts 

ranging from 4 to 6. The silty clays had moisture contents of around 40 percent and void 

ratios around 1.2. Below the silty clay, silt with sand and poorly graded sand was present, 

generally at depths of about 35 feet. 

7.1.3 Design Concerns

The alluvial nature of the site soils led to design concerns about settlement and stability of 

roadway embankments. Post-construction settlement stemming from the added fill, to 

maintain the proper grade for the roadway embankments, led to settlement projections 

varying up to 30 inches, depending upon the height of fill placed. Various solutions, such as 

removal and replacement, surcharging, surcharging with PVDs, and use of stiff columns to 

transfer load were investigated. The high cost of removal and replacement eliminated this 

option. The schedule of the project and the time rate of settlement of the subsurface soils 

eliminated the option of just building the embankment and waiting for the settlement to 

occur. This led to the use of PVDs to accelerate the settlement, combined with surcharge in 

some areas, and the use of stiff columns in other areas. 
2-53 

7.1.4 Design

In areas of PVDs and surcharge, the results from laboratory testing indicated that a maximum 

total settlement of 24 inches would occur under approximately 40 feet of fill. Based on 

consolidation coefficients, it was determined that a 4-foot triangular pattern spacing of the 

PVDs would result in a settlement time period of approximately 3 to 5 months, once full 

height was attained. The PVD design depths ranged up to 35 feet. There were three areas to 

be treated using PVDs and fill loading. The estimated amount of PVDs was 18,400 in 

number with a linear estimated length of 690,750 feet, as shown in Table 2-4. 

Table 2-4. Estimated and Installed PVDs

Area

Estimated

Linear feet

Estimated 

No. of PVDs

Installed 

Linear feet

Installed

No. of PVDs

No. of Days 

to Install

3 231,000 6,600 129,494 5,850 9 

2 374,000 9,350 234,192 8,111 13 

8 85,750 2,450 74,391 2,367 4 

Total 690,750 18,400 438,077 16,328 26 

Source: Menard Group USA

A test pad was constructed as proof of concept. Typical test pad results are shown in Figure 

2-15, which indicate approximately 18 inches of settlement under 36 feet of fill. The majority 

of the settlement occurred in the first two months. 

7.1.5 Project Results

The wick drain material (PVD) used was COLBONDDRAIN CX 1000®, 4.0 inches in 

width with 1/8 inch thickness. Installation was accomplished with either an ALIMAK®

chain driven stitcher or an APE sprocket driven stitcher using a 5-1/2 inches by 3 inches 

rhombic-shaped mandrel. Installation of the PVDs in the three areas took 26 days to 

complete. As shown in Table 2-4, the installed numbers of PVDs and the total lengths were 

slightly less than the estimated values. The average installed length was about 27 feet. 

Actual settlements ranged from 4 to 20 inches and were generally attained in 2 to 3 months; 

well within the design time of 6 months. 

7.1.6 Project Cost

There were four bids for this project, ranging from $476,465 to $823,396, to include the PVD 

installation and construction of the fill embankments. Subcontractor prices for the PVDs 

were approximately $0.50 per lineal foot. Mobilization costs for multiple rigs were bid at 

$40,000. 
2-54 

Courtesy of CH2M-Hill

Figure 2-15. Settlement versus fill height for test pad.

Acknowledgments – Iowa Department of Transportation, CH2M-Hill, Menard Group USA

7.2 Approach Ramps for Bridge Replacement

7.2.1 Project Description

This project involved the replacement of the existing Tifft Street Bridge over railroads 

located between the intersection of the Tifft Street with NY Route 5 ramps and Hopkins 

Street. Replacement of the existing bridge required construction of a new structure and 

embankments, incorporating the newest design standards, including greater vertical 

clearances over railroads and increases in lane widths.

The new structure would be significantly shorter on the west end, meaning the new west 

approach embankment would extend 300 feet beyond the old abutment. The total alignment 

of the new embankment centerline and bridge would be approximately 70 feet north of the 

existing centerline. In addition, the existing bridge had to remain open until the new bridge 

could accommodate at least two lanes of traffic.
2-55 

7.2.2 Subsurface Conditions

The generalized subsurface soil profile indicated a 7 to 16 foot fill layer of cinders and slag 

overlying 2.5 feet of peat (under the west embankment only), 33 to 46 feet of soft silty clay, 

and a 10 foot layer of sand on top of limestone bedrock. The groundwater table was within 

1.5 to 3 feet of the existing surface.

7.2.3 Design Concerns

Construction of the new approach embankments, varying in height from 15 to 35 feet, would 

result in considerable settlements. The west approach embankment would result in 

settlements ranging from 24 to 36 inches without treatment. The settlement was expected to 

occur over a 6-year period after the full fill height was achieved.

On the east embankment, the proposed fill would incorporate much of the existing fills. 

Therefore, settlements of only 12 to 24 inches were expected, but the time of the settlement 

would be similar to the west approach. The designers used PVDs, on both approaches, to 

accelerate the settlements to within a 6-month time frame.

On the west approach, a significant portion of the embankment south side slope would be 

under the existing bridge. It was felt that PVDs would be necessary in this area to prevent 

significant differential settlement of the side slope in future years. Therefore, some of the 

PVDs would have to be installed to maximum depths of 55 feet with a headroom clearance of 

only 26 feet.

On the east embankment there was no need for PVDs under the existing bridge, since the 

embankment did not extend past the old abutment. However, there was still need for PVDs 

adjacent to the existing embankment. Borings indicated that some PVDs in the new east 

embankment would have to penetrate slag under the existing embankment. The depth of the 

slag was 15 feet, and it would be difficult to penetrate, as indicated by very high "N" values 

in excess of 100 in some locations. The slag material was very abrasive and would be 

difficult to drill.

Analyses indicated that side slopes of 2H:1V could be safely constructed to a height of only 

28 feet without foundation treatment. However, after the strength gain as a result of PVD 

accelerated consolidation, embankments could be constructed to the required heights of 35 

feet.
2-56 

7.2.4 Design

The west embankment was expected to settle up to a maximum 36 inches under maximum 

fill height, and the east embankment 24 inch maximum. A surcharge height of 36 inches

above final grade elevation accelerated settlement. In the area of maximum settlement and 

where stability was of concern, the PVD spacing was 4 feet in a triangular pattern. In areas of 

lesser settlement, and where stability was not a concern, the pattern spacing was 7 feet.

The total quantity of the original estimate was 229,600 lineal feet of PVDs. Added to this 

quantity was the cost of 6,500 yd3 of a granular material for the drainage blanket and 19,100 

yd2 of geotextile. New York State DOT estimated the total cost of the PVD solution, 

including auxiliary items, to be in excess of $800,000. 

7.2.5 Project Construction Procedure

PVD construction began on the east side, and it became readily apparent that special drilling 

would be necessary to penetrate the slag fill on the area near the existing embankment. 

Special air rotary drills were used to pre-drill, and the PVD installation unit had to follow 

close behind. In areas farther from the existing embankment, a combination unit of vibro and 

static force was used to install the PVDs without pre-drilling. Approximately 82,000 feet of 

PVDs were completed on the east side within 4 weeks. 

PVDs underneath the bridge on the west side were started simultaneously and completed in 

segments. A special sectional mandrel was developed such that it could be pinned together in 

a fairly rapid manner. The initial section was 20 feet long, with subsequent sections of 16 

feet. At these locations, the holes had to be predrilled through a miscellaneous fill surface 

either with an auger or a special air track machine. After several days of trial procedures, a 

maximum production rate of 25 drains per day was achieved. This compared to a maximum 

production rate of more than 300 drains per day on the locations not under the bridge. 

The PVDs on the west side that were not under the bridge were completed in 4 weeks. It was 

also necessary to angle some drains under high tension wires in order to achieve full 

coverage of the area. The total quantity of PVDs installed on the west approach was 

approximately 110,000 feet. 

To install the drains in sections underneath the bridge there had to be more than one splice 

per drain. This requirement conflicted with normal specifications, which require only one 

splice per drain.
2-57 

7.2.6 Project Results

Typical piezometers and settlement data from both the east side and west approach are shown 

in Figures 2-16 and 2-17. 

Total maximum settlement of the west approach was 28 inches; on the east approach, the 

maximum settlement was 24 inches. The settlement was achieved in both locations in 

approximately 4 to 5 months and agreed fairly well with predictions. Piezometers, slope 

indicators, and settlement plates were used to determine the time rate and amount of 

consolidation. The purpose for the slope indicators was two-fold. While stability was not 

believed to be a concern, the factor of safety was marginal. Secondarily, it was believed that 

lateral movements could be used to determine the immediate settlement component.

7.2.7 Project Costs

This project was somewhat unique in that the use of several different installation units had to 

be bid under one item. The drains were installed using standard static machines along with 

pre-drilling in some locations, static vibro machines in other locations, and a special 

installation machine underneath the bridge. 

The general contractor's bid prices were for (1) PVDs at $2.10 per foot; (2) granular drainage 

blanket at $17.00 per cubic yard.; and (3) geotextile at $1.40 per square yard. Using the 

original cost estimate, the actual prices resulted in a PVD solution less than $600,000, which 

compared quite favorably to projected costs. 

Subsequent to this project, New York State DOT added a new bid item to account for PVD 

mobilization costs. This bid item attempts to account for the need for special equipment on 

projects.
2-58 

Figure 2-16. Tifft Street west approach embankment field data.
2-59 

Figure 2-17. Tifft Street east approach embankment field data.
2-60 

7.3 Jones Creek Substation Site Fill

7.3.1 Project Description

Located in Freeport, Texas the Jones Creek Substation site fill project was a one where the 

existing ground surface was increased from elevation 5 feet MLS to elevation 23 feet to aid 

in the creation of an electrical control yard that would be above the potential flood elevation. 

The project began in July 2015 with site grading. The project scope include clearing and 

grubbing the site in preparation for grading activities, and installation of PVDs, horizontal 

strip drains, site fill and surcharge fill. The final grading plan is presented in Figure 2-18. 

7.3.2 Subsurface Conditions

The project site is located in southeastern Texas in the gulf coastal plain physiographic 

province. The area is generally comprised of fluvial and deltaic sediments. The subsoil 

consists of approximately 2-foot thick topsoil containing loose gray sandy silt (ML). The 

topsoil is underlain by 23 to 38 feet of soft to stiff gray and reddish brown to brown fat clay 

with sand (CH), which in turn is underlain by medium dense silty sand of 10 to 15 feet 

thickness. The groundwater table was generally found to be around 5 feet below ground 

surface, one day after drilling the bore holes. The fat clay was very soft to medium stiff with 

blow counts between 3 and 9. The soil has an approximate liquid limit of 69, moisture 

content of 38%, fines content between 83% and 98%, and a void ratio of 0.95. 

7.3.3 Design Concerns

The soft, compressible soils at the sites presented a large risk for long-term potential 

settlement due the 18 feet of site fill. The site grading was estimated to induce 19 inches of 

total settlement at the center of the site. Based on the engineers’ previous experience with the 

soils near the project site, they estimated that 10 to 20 percent of the total settlement would 

occur during the site grading operations (4 to 6 months) and between 25 and 50 percent of 

the consolidation would occur within the first two years of fill placement. The duration 

without PVDs and surcharge was too great for the client, due to the sensitivity of the 

electrical equipment on the planned yard. Therefore, it was deemed necessary to utilize 

PVDs to decrease the time for consolidation, and the use of surcharge to place the soil into a 

slightly over-consolidated state. Due the absence of readily available drainage blanket 

material, the engineer designed for the use of horizontal strip drains with two PVDs attached 

to each strip drain
2-61 

Figure 2-18. Final grading plan for Jones Creek Substation fill project.
2-62 

7.3.4 Design

Based on the results from three different geotechnical investigations, and subsequent 

laboratory data, it was determined that a 4-foot triangular spacing would result in 90% 

consolidation in 4 months following the placement of the fill and surcharge. The design 

depth for the PVDs was estimated to be an average of 35 feet, with allowance for 2 feet of 

cut-off allowed to attach the PVD to the horizontal strip drain. Details of the estimated and 

installed PVDs are listed in Table 2-5. 

Table 2-5. Estimated and Installed PVDs

Area

Estimated 

No. of 

PVDs

Estimated 

Length of PVDs 

incl. Cut-Offs

Estimated 

Linear Feet 

of PVDs

Installed 

Linear 

Feet

No. of 

Installed 

PVDs

No. of 

Days to 

Install

1 12,468 37 461,316 446,840 12,468 20 

7.3.5 Project Results

The wick drain material (PVD) used was MebraDrain® 7407, a two piece wick drain with a 

polypropylene core and seamed geotextile fabric measuring a combined 4.0 inches in width 

and 3/8 inches thickness. Installation was completed using a hydraulically driven wick 

stitcher with a 5 inch by 2 inch smooth rectangular mandrel to minimize soil disturbance. The

installation of the vertical drain was completed with a three person crew, two for wick drain 

installation and another for horizontal drain installation. The entire installation period was 20 

working days. The total installed quantity of wick drain was slightly less than the estimated

value, with an installed average depth of 33.8 feet as opposed to the estimated depth 35 feet. 

Total settlement for the project ranged from 8 inches to 15.6 inches, and generally occurred 

within of 2 months of placement of the fill and surcharge; and was within the design 

allowance of 4 months. Data recorded via settlement plates is presented in Figure 2-19 

against the height of fill and project duration. 

7.3.6 Project Cost

There were two wick drain bids for this project comprised of mobilization for a single 

installation unit, supply and installation of vertical wick drain, and supply and installation of 

horizontal strip drains. The prices submitted by the winning subcontractor were: $27,500 

mobilization, wick drain at approximately $0.45 per linear feet and horizontal strip drain at 

approximately $1.50 per linear foot. 

Acknowledgements – Hayward Baker Inc. 
2-63 

Courtesy of Hayward Baker Inc.

Figure 2-19. Settlement plate data and fill height versus project duration.
2-64 

8.0 REFERENCES

Adama Engineering (2003). Foundation Stress & Settlement Analysis. FoSSA 1.0, Newark, 

DE. Available at: http://www.geoprograms.com. 

Asaoka, A. (1978). Observational Procedure of Settlement Prediction. Soils and 

Foundations, Japanese Society of Soil Mechanics and Foundation Engineering, 

18(4): pp. 87–101. 

Bartlett, S.F., Monely, G., Palmer, A., and A. Soderborg, A. (2001). Instrumentation and 

Construction Performance Monitoring for the I-15 Reconstruction Project, Salt Lake 

City, Utah. Transportation Research Record No. 1772, Transportation Research 

Board, Washington, D.C., pp. 40-47. 

Barron, R.A. (1948). Consolidation of Fine-Grained Soils by Drain Wells. Transactions 

ASCE, Vol. 113, Paper No. 2346, pp. 718-742. 

Bo, M.W., Chu, J., and Choa, V. (2005). Use of Prefabricated Vertical Drains for the 

Improvement of Ultra-soft Soil. Proc. 6th International Conference on Ground 

Improvement Techniques, Coimbra, Portugal. 

Chu, J., Bo, M.W., and Choa, V. (2004). Practical Considerations for Using Vertical Drains 

in Soil Improvements Projects. Geotextiles and Geomembranes, 22: pp. 101-117. 

Chu, J., Varaksin, S., Klotz, U., and Menge, P. (2009). State of the Art Report: Construction 

Processes. Proc. 17th International Conference on Soil Mechanics and Geotechnical 

Engineering, Alexandria, Egypt, pp. 3006-3135. 

DFI. (2014). Wick Drain Guide Specification (Method). Prepared by the Ground 

Improvement Committee of Deep Foundations Institute, First Edition, 1st Printing, 

Hawthorne, N.J. http://www.dfi.org. 

Fellenious, B.H. and Altaee, A. (1999). Discussion on Design Curves for Prefabricated 

Vertical Drains. Journal of Geotechnical and Geoenvironmental Engineering, 125(4): 

pp. 338-340. 

FHWA. (2002). Subsurface Investigations. Authors: Mayne, P.W., Christopher, B.R., and 

DeJong, J., FHWA NHI-01-031, Federal Highway Administration, U.S. DOT, 

Washington, D.C., 300p. 
2-65 

FHWA. (2008). Geosynthetic Design and Construction Guidelines. Authors: Holtz, R.D., 

Christopher, B.R., and Berg, R.R., FHWA-HI-07-092, Federal Highway 

Administration, U.S. DOT, Washington, D.C., 460p. 

GEC 5. (2002). Evaluation of Soil and Rock Properties. Authors: Sabatini, P.J., Bachus, 

R.C., Mayne, P.W., Schnieder, J.A., and Zettler, T.E. FHWA IF-02-034, Federal 

Highway Administration, U.S. DOT, Washington, D.C., 385p. 

GEC 11. (2006). Soils and Foundations. Authors: Samtani, N.C. and Nowatzki, E.A., FHWA 

NHI-06-088 Vol I and NHI-06-089 Vol II, Federal Highway Administration, U.S. 

DOT, Washington, D.C., 462p. (Vol I) and 594p. (Vol II). 

Han, J. (2015). Principles and Practice of Ground Improvement. John Wiley & Sons, Inc., 

Hoboken, NJ, 418p. 

Hansbo, S. (1979). Consolidation of Clay by Band-Shaped Prefabricated Drains. Ground 

Engineering, 12(5), pp. 16-25. 

Hansbo, S. (1981). Consolidation of Fine-Grained Soils by Prefabricated Drains. Proc. 10th

International Conference on Soil Mechanics and Foundation Engineering, Vol. 3, pp. 

667-682. 

Hansbo, S. (1986). Preconsolidation of Soft Compressible Subsoil by the use of Prefabricated 

Vertical Drains. Annales des Travaux Publics de Belgique. Issue 6, 553p. 

Hansbo, S. (2004). Band Drains. Chapter 1 in Ground Improvement, 2nd Edition, M.P. 

Moseley and K. Kirsch, Editors, Spon Press, New York, NY, pp. 4-56.

Holtz, R.D. and Christopher, B.R. (1987). Characteristics of Prefabricated Drains for 

Accelerating Consolidation. Proc. Ninth European Conference on Soil Mechanics 

and Foundation Engineering, Balkema, Vol. 2, Dublin, Ireland. 

Holtz, R.D., Jamiolkowski, M.B., Lancellotta, R., and Pedroni, R. (1991). Prefabricated 

Vertical Drains: Design and Performance. Construction Industry Research and 

Information Association, Butterworth-Heinemann, Oxford, UK, 131p. 

Holtz, R.D., Kovacs, W.D., and Sheahan, T.C. (2013). An Introduction to Geotechnical 

Engineering, 2nd Edition. Prentice-Hall, Inc., Upper Saddle River, NJ, 864p. 

Ladd, C.C. (1991). Stability Evaluation During Staged Construction. Journal of Geotechnical 

Engineering, 117(4), pp. 542-615. 
2-66 

Kjellman, W. (1948). Accelerating Consolidation of Fine Grained Soils by Means of 

Cardboard Wicks. Proc. 2nd International Conference on Soil Mechanics and 

Foundation Engineering, Vol. 2, Rotterdam, The Netherlands, pp. 302-305. 

Koerner, R.M. (2012). Designing with Geosynthetics, 6th Edition. Vols. 1 and 2. Xlibris 

Corp., 914p. 

NAVFAC. (1982). Soil Mechanics. NAVFAC Design Manual 7.1, Naval Facilities 

Engineering Command, Department of the Navy, pp. 241-259. 

Rixner, J.J., Kraemer, S.R., and Smith A.D. (1986a). Prefabricated Vertical Drains. Report 

No. FHWA/RD-86/168, Vol. I: Engineering Guidelines, Federal Highway 

Administration, U.S. DOT, Washington, D.C. Available at: 

http://www.fhwa.dot.gov/bridge/geopub.htm. 

Rixner, J.J., Kraemer, S.R., and Smith A.D. (1986b). Shared Experience in Geotechnical 

Engineering, Wick Drains Transportation Research Circular, Number 309. 

Rollins, K.M., Anderson, J.K.S., Goughnour, R.R., and McCain, A.K. (2004). Liquefaction 

Hazard Mitigation Using Vertical Composite Drains. Proc. 13th World Conference on 

Earthquake Engineering, Vancouver, B.C., Canada, Paper No. 2880. 

Rollins, K.M., and Strand, S.R. (2007). Liquefaction Mitigation Using Vertical Composite 

Drains: Full-Scale Testing for Pile Applications. Final Report for NCHRP-IDEA 

Project 103, Transportation, Research Board, The National Academies, Washington, 

D.C.
Chapter 3 

LIGHTWEIGHT FILLS 

3-i 

CONTENTS

1.0 DESCRIPTION AND HISTORY ........................................................................... 3-1

1.1 Description ............................................................................................................ 3-1

1.2 Historical Overview ............................................................................................. 3-1

1.3 Focus and Scope ................................................................................................... 3-3

1.4 Terminology and Acronyms................................................................................ 3-4

1.5 Primary References ............................................................................................. 3-5

2.0 FEASIBILITY CONSIDERATIONS .................................................................... 3-7

2.1 Functions and Design Considerations ................................................................ 3-7

2.1.1 Load Reduction .................................................................................................. 3-7

2.1.2 Shear Strength .................................................................................................... 3-8

2.1.3 Compressibility .................................................................................................. 3-8

2.1.4 Lateral Pressures ................................................................................................ 3-8

2.1.5 Drainage Characteristics .................................................................................... 3-9

2.1.6 Construction in Adverse Weather ...................................................................... 3-9

2.1.7 Seismic Considerations ...................................................................................... 3-9

2.2 Advantages and Potential Disadvantages .......................................................... 3-9

2.2.1 Advantages ......................................................................................................... 3-9

2.2.2 Potential Disadvantages ................................................................................... 3-10

2.3 Alternative Solutions ......................................................................................... 3-11

3.0 LIGHTWEIGHT FILL MATERIALS ................................................................ 3-12

3.1 Introduction ........................................................................................................ 3-12

3.2 Lightweight Fill Materials with Compressive Strength ................................. 3-12

3.2.1 Geofoam ........................................................................................................... 3-12

3.2.2 Lightweight Cellular Concrete ......................................................................... 3-13

3.3 Granular Lightweight Fills ............................................................................... 3-15

3.3.1 Tire Shreds ....................................................................................................... 3-15

3.3.2 Wood Fiber ...................................................................................................... 3-17

3.3.3 Expanded Shale, Clay, and Slate (ESCS) ........................................................ 3-18

3.3.4 Fly Ash ............................................................................................................. 3-19
3-ii

3.3.5 Blast Furnace Slag ........................................................................................... 3-20

3.3.6 Boiler Slag ....................................................................................................... 3-21

4.0 DESIGN CONCEPTS ........................................................................................... 3-22

4.1 Design Considerations ....................................................................................... 3-22

4.2 Design Procedures .............................................................................................. 3-22

4.2.1 Fills with Compressive Strength ...................................................................... 3-23

4.2.2 Granular Lightweight Fills ............................................................................... 3-30

5.0 CONSTRUCTION SPECIFICATIONS AND QUALITY ASSURANCE ....... 3-37

5.1 Specification Development ................................................................................ 3-37

5.2 Quality Assurance and Construction Control ................................................. 3-38

6.0 COST DATA .......................................................................................................... 3-40

6.1 Cost Components ............................................................................................... 3-40

7.0 CASE HISTORIES ................................................................................................ 3-43

7.1 GeoFoam ............................................................................................................. 3-43

7.2 Foamed Concrete ............................................................................................... 3-46

7.3 Tire Shreds ......................................................................................................... 3-47

7.4 Wood Fiber Embankment ................................................................................. 3-52

8.0 REFERENCES ....................................................................................................... 3-56
3-iii

LIST OF FIGURES

Figure 3-1. Embankment construction with EPS-block geofoam. ...................................... 3-13

Figure 3-2. Foamed concrete MSE wall backfill for light rail project in California. .......... 3-14

Figure 3-3. Embankment fill construction with tire shreds in Blue Earth County, 

Minnesota. ...................................................................................................................... 3-15

Figure 3-4. Embankment construction with wood fiber. ..................................................... 3-18

Figure 3-5. Placing expanded clay fill. ................................................................................ 3-19

Figure 3-6. Typical EPS-block geofoam applications involving standalone 

embankments. ................................................................................................................ 3-24

Figure 3-7. Major components of an EPS-block geofoam slope system. ............................ 3-25

Figure 3-8. Multiple potential failure surfaces to consider with EPS-block geofoam 

slope systems. ................................................................................................................ 3-25

Figure 3-9. Isometric view of typical EPS block layout. ..................................................... 3-28

Figure 3-10. Whittier Bridge typical embankment widening section. ................................. 3-44

Figure 3-11. Whittier Bridge typical ramp cross section. .................................................... 3-45

Figure 3-12. Whittier Bridge EPS block placement on embankment widening. ................. 3-46

Figure 3-13. Foamed concrete fill behind existing abutment, New York project. .............. 3-47

Figure 3-14. Marina Drive slope cross section. ................................................................... 3-49

Figure 3-15. Marina Drive longitudinal section of fill replacement. ................................... 3-50

Figure 3-16. Placement and compaction of TDA within geotextile wrap. .......................... 3-51

Figure 3-17. Completed Marina Drive repair. ..................................................................... 3-52

Figure 3-18. Soil cross-section at wood fiber fill location, Washington project. ................ 3-53
3-iv

LIST OF TABLES

Table 3-1. Densities and Specific Gravity for Various Lightweight Fill Materials............... 3-1

Table 3-2. Equivalent Soil Subgrade Values of EPS-Block Geofoam for Pavement 

Design ............................................................................................................................ 3-26

Table 3-3. Expanded Polystyrene (EPS) Design and Construction Guidelines ................... 3-27

Table 3-4. Cellular Lightweight Concrete Design and Construction Considerations ......... 3-30

Table 3-5. Tire Shreds Design and Construction Guidelines .............................................. 3-31

Table 3-6. Wood Fiber Design and Construction Guidelines .............................................. 3-32

Table 3-7. Expanded Shale, Clay, and Slate (ESCS) Design and Construction 

Guidelines ...................................................................................................................... 3-33

Table 3-8. Fly Ash Design and Construction Guidelines .................................................... 3-34

Table 3-9. Air-Cooled Blast Furnace Slag Design and Construction Guidelines ................ 3-35

Table 3-10. Boiler Slag Design and Construction Guidelines ............................................. 3-36

Table 3-11. Typical Cost Ranges for EPS-block Geofoam at Source, Delivered, and InPlace ............................................................................................................................... 3-41

Table 3-12. Typical Cost Ranges for Lightweight Fills ...................................................... 3-41

Table 3-13. Example Lightweight Fill Project Cost Comparison........................................ 3-42
3-1 

1.0 DESCRIPTION AND HISTORY

1.1 Description

The unit density of most compacted soil fills, consisting of sands, silts, or clays, generally 

ranges from about 115 to 140 pounds force per cubic foot (pcf). The use of such conventional 

earth fill material can result in significant settlement or decreased stability on some projects. 

On some projects, it is desirable to use an alternate fill material with a lower unit density to 

reduce the magnitude of applied earth loads. 

In such situations, the use of a lightweight fill material can result in reduced settlement 

and/or increased stability. The large variety of lightweight fill materials provides a large 

range in densities, ranging from less than 1 to about 90 pcf. Lightweight fills are most often 

used to reduce the applied loads to (after Caltrans 2014): 

• Eliminate or significantly reduce magnitude of embankment settlement.

• Eliminate or significantly reduce time required to achieve embankment settlement.

• Reduce lateral pressure behind retaining walls, abutments, and other structures.

• Reduce driving force in landslide repair.

• Increase embankment resistance to seismic loads (low unit weight (density) results in 

lower seismic inertial forces).

1.2 Historical Overview

Many types of lightweight fill materials have been used for roadway embankment 

construction. Some of the more common lightweight fills are listed in Table 3-1. 

Table 3-1. Densities and Specific Gravity for Various Lightweight Fill Materials

Fill Type

Range in Density

pcf

Range in 

Specific Gravity

Geofoam (RCPS) 0.70 to 3.00 0.01 to 0.05 

Cellular Concrete 20 to 80 0.4 to 1.3 

Wood Fiber 35 to 55 0.6 to 0.9 

Tire Shreds 37 to 73 0.6 to 1.2 

Expanded Shale, Clay, and Slate (ESCS) 37 to 65 0.6 to 1.0 

Fly Ash 70 to 90 1.1 to 1.4 

Boiler Slag 60 to 90 1.0 to 1.4 

Expanded Air-Cooled Slag 69 to 94 1.1 to 1.5 
3-2 

There is a wide range in the unit weight of these lightweight fill materials, but all have a 

density less than conventional soils. The composition and sources of the lightweight fill 

materials listed in Table 3-1 are discussed in Section 3. The availability of some of these 

materials and, therefore cost, will vary by geographic region. 

Lightweight fill materials have been used for decades. The worldwide interest and use of 

lightweight fill materials led to the publication of a 1997 authoritative reference, Lightweight 

Filling Materials, by the Permanent International Association of Road Congresses (PIARC 

1997). 

Geofoam is a generic term used to describe expanded polystyrene (EPS) and extruded 

polystyrene (XPS) material used in geotechnical applications. Geofoam was initially 

developed for insulation material to prevent frost from penetrating soils. The initial use of 

geofoam was in Scandinavia and North America in the early 1960s. In 1972, EPS-block 

geofoam was used as a lightweight fill for a project in Norway (PIARC 1997; Horvath 1995). 

Today, EPS-block geofoam is widely used by state transportation agencies as lightweight fill 

for embankment construction; and standards for its use are well established. 

The technique of using pumping equipment to inject foaming agents into concrete was 

developed in the late 1930s. Little is known about the early uses of this product. However, 

the United States Army Corps of Engineers used foamed concrete as a tunnel lining and 

annular fill. This product is generally job-produced as cement/water slurry, with preformed 

foam blended for accurate control and immediate placement. Today, lightweight cellular 

concrete (a.k.a. foamed concrete) is widely used by state transportation agencies for 

geotechnical fills. 

Wood fiber has been used for many years by timber companies for roadways crossing peat 

bogs and low-lying land, as well as for landslide repairs (PIARC 1997). The steel-making 

companies have produced slag by-product since the start of the iron- and steel-making 

industry. Initially, the slag was stockpiled as waste materials, but beginning around 1950, the 

slag were crushed, graded, and sold for fill materials (Lewis 1982; National Slag Association 

1968; National Slag Association 1988). 

Tire shreds are a relatively more recent source of lightweight fill materials. Tire shreds have 

been used for lightweight fill in the United States and in other countries since the mid 1980s. 

The availability of this material is increasing each year, and its use as a lightweight fill is 

further promoted by the need to dispose of tires. Tires are shredded to create tire-derived 

aggregate (TDA), which can be used as a substitute for gravel, sand and other lightweight fill 

materials (FHWA 2010). 
3-3 

It is estimated that as of 2013, about 96 percent of scrap tires generated in the United States 

are used in tire-derived fuel, ground rubber, civil engineering, and other scrap tire markets. 

Civil engineering applications consist of tire shreds used in roadway construction, landfill 

construction, septic tank leach fields, and other construction. The civil engineering 

applications use about 172,000 tons of scrap tires, which is approximately 5 percent of scrap 

tire use. (https://rma.org/) 

Expanded shale, clay, and slate (ESCS) lightweight aggregate has been used for decades to 

produce aggregate for concrete and masonry units. Beginning in about 1980, lightweight 

aggregates have also been used for geotechnical purposes. Completed projects include use 

behind a bulkhead to reduce the lateral pressures on the steel sheeting, for construction of 

roadways over soft ground, and of the lightweight fill to replace soil in a slope to reduce the 

gravitational driving force and improves stability and safety of the slope (ESCSI 2007; 

PIARC 1997; Holm and Valsangkar 1993; Stoll and Holm 1985; Valsankar and Holm 1990). 

Waste products from coal burning include fly ash and boiler slag. Both of these materials 

have been used in roadway construction (PIARC 1997). One of the first documented uses of 

fly ash in an engineered highway embankment occurred in England in 1950. Trial 

embankments led to the acceptance of fly ash fills, and other roadway projects were 

constructed in other European countries. In 1965, a fly ash roadway embankment was 

constructed in Illinois. Today, the American Coal Ash Association (ACAA) reports that 

about 2.8 million tons of fly ash is used for embankment and structural fill in the United 

States per year (ACAA 2014). Boiler slag has been used for backfill since the early 1970s. 

Approximately 50,000 tons of boiler slag is used for embankment and structural fill in the 

United States per year (ACAA 2014). 

Another byproduct used as lightweight fill is expanded or foamed air-cooled blast furnace 

slag (ACBFS), from iron and steel manufacturing. A 2009 survey of transportation agencies 

found little use of lightweight expanded ACBFS in embankment fill applications (StroupGardiner and Wattenberg-Komas 2013c). 

1.3 Focus and Scope

The purpose of this technical summary is to present an overview of the more common 

lightweight fill materials that have been used for geotechnical applications in highway 

construction. Typical geotechnical engineering parameters for the various lightweight fills 

that are important for design are provided. In addition, design and construction 

considerations unique to each of these lightweight fill materials are presented. This 

information can be used for preliminary planning purposes. 
3-4 

The technical summary discusses preparation of specifications and construction control 

procedures, and references where detailed information is available at. Approximate costs for 

the various lightweight fill materials are also presented. Four case histories are presented to 

demonstrate the effectiveness of lightweight fills for specific situations. 

This chapter should not be considered to be a design manual on use of lightweight fill, since 

the information contained herein is not presented in detail. Furthermore, the engineering 

properties of some materials can vary significantly depending upon the source of the material 

and the manufacturing process. The engineering parameters presented in this document are 

typical values, and a detailed design should include additional testing and evaluation of the 

contemplated lightweight fill material. References are given so that the reader may obtain 

more specific information regarding the lightweight fill materials, including properties, 

performance records, and applications. Key references are listed in Section 1.5. 

1.4 Terminology and Acronyms

EPS-block geofoam – generic block‐molded product form of expanded polystyrene used in 

small-strain geofoam applications

Geofoam – block or planar rigid cellular foam polymeric material used in geotechnical 

engineering applications. (ASTM 2015) 

Tire bales are produced by mechanically compressing and tying 100 (automobile) whole 

tires to form a bale, approximately 2 cubic yards in volume (Zornberg et al. 2005). 

Tire chips – pieces of scrap tires that have a basic geometrical shape and are generally 

between 12 and 50 mm [1⁄2 and 2 inches] in size and have most of the wire removed. (ASTM 

2012) 

Tire derived aggregate (TDA) – pieces of scrap tires that have a basic geometrical shape 

and are generally between 12 and 305 mm [1⁄2 to 12 inches] in size and are intended for use in 

civil engineering applications. (ASTM 2012) 

Tire shreds – pieces of scrap tires that have a basic geometrical shape and are generally 

between 50 and 305 mm [1⁄2 to 12 inches] in size. (ASTM 2012) 

Tire shreds are defined by New York State DOT as pieces of scrap tire between 2 inches 

and 12 inches in size (NYSDOT 2015a). 

A number of abbreviations are used throughout this chapter: 

• ACBFS air-cooled blast furnace slag
3-5 

• BFS blast furnace slag

• EPS expanded polystyrene geofoam

• RCPS letter designation for rigid cellular polystyrene geofoam covered by 

ASTM D6817 (2015), includes both EPS and XPS

• TDA tire derived aggregate 

• XPS extruded polystyrene geofoam 

1.5 Primary References

• ASTM (2015). Standard Specification for Rigid Cellular Polystyrene Geofoam, 

ASTM D6817-15, ASTM International, West Conshohocken, PA, 4p. 

• ASTM (2013). Standard Guide for Use of Expanded Polystyrene (EPS) Geofoam in 

Geotechnical Projects, ASTM D7180, ASTM International, West Conshohocken, PA, 

3p. 

• ASTM (2012). Standard Practice for Use of Scrap Tires in Civil Engineering 

Applications, ASTM D6270-08, ASTM International, West Conshohocken, PA, 22p. 

• Arellano, D., Stark, T.D., Horvath, J.S., and Leshchinsky, D. (2011). Guidelines for 

Geofoam Applications in Slope Stability Projects. NCHRP Project No. 24-11(02), 

National Cooperative Highway Research Program, Transportation Research Board, 

Washington, D.C., 602p. 

• Allen, T.M. and Kilian, A.P. (1993). Use of Wood Fiber and Geotextile 

Reinforcement to Build Embankment Across Soft Ground. Transportation Research 

Record No. 1422, Transportation Research Board, Washington, D.C., pp. 46-54. 

• Cheng, D. (2016). Usage Guide, Tire-Derived Aggregate (TDA), Publication # 

DRRR 2016-01545, California Department of Resources Recycling and Recovery, 

Sacramento, CA, 56p. http://www.calrecycle.ca.gov/Publications/Default.aspx. 

• ECSI (2007). Reference Manual for the Properties and Applications of Expanded 

Shale, Clay and Slate Lightweight Aggregate. Expanded Shale, Clay and Slate 

Institute, Salt Lake City, UT. 

• Kilian, A.P. and Ferry, C.D. (1993). Long Term Performance of Wood Fiber Fills. 

Transportation Research Record No. 1422, Transportation Research Board, 

Washington, D.C., pp.55-60. 

• NYSDOT (2015). Guidelines for Project Selection, Design, and Construction of Tire 

Shreds in Embankments. Geotechnical Engineering Manual, GEM-20 Revision #3, 

New York State Department of Transportation. 
3-6 

• Stark, T.D., Arellano, D., Horvath, J.S. and Leshchinsky, D. (2004a). Guideline and 

Recommended Standard for Geofoam Applications in Highway Embankments, 

NCHRP Report 529, National Cooperative Highway Research Program, 

Transportation Research Board, Washington, D.C., 58p. 

• Stark, T.D., Arellano, D., Horvath, J.S. and Leshchinsky, D. (2004b). Geofoam 

Applications in the Design and Construction of Highway Embankments, NCHRP 

Web Document 65, http://www.trb.org/main/blurbs/153437.aspx. 

• Stroup-Gardiner, M. and Wattenberg-Komas, T. (2013a). Recycled Materials and 

Byproducts in Highway Applications, Volume 1: Summary Report, NCHRP 

Synthesis 435, Transportation Research Board, Washington, D.C., 93p. 

• Stroup-Gardiner, M. and Wattenberg-Komas, T. (2013b). Recycled Materials and 

Byproducts in Highway Applications, Volume 7: Scrap Tire Byproducts, NCHRP 

Synthesis 435, Transportation Research Board, Washington, D.C., 47p. 

• Stroup-Gardiner, M. and Wattenberg-Komas, T. (2013c). Recycled Materials and 

Byproducts in Highway Applications, Volume 5: Slag Byproducts, NCHRP Synthesis 

435, Transportation Research Board, Washington, D.C., 35p. 

• Stroup-Gardiner, M. and Wattenberg-Komas, T. (2013d). Recycled Materials and 

Byproducts in Highway Applications, Volume 2: Coal Combustion Byproducts, 

NCHRP Synthesis 435, Transportation Research Board, Washington, D.C., 51p. 
3-7 

2.0 FEASIBILITY CONSIDERATIONS

2.1 Functions and Design Considerations

Lightweight fill materials have many potential highway applications, principally where the 

capacity of the underlying soft soil is too low to carry the design load and/or the estimated 

settlement of the embankment is too large. Under these conditions, the foundation soils must 

be improved or bypassed, or the embankment load reduced. Lightweight fills fulfill the latter 

function. A brief overview of lightweight fill use or function is summarized as follows. 

2.1.1 Load Reduction 

When a fill is placed on soft ground, the main driving force is from the weight of the 

embankment itself. Conventional methods of improving the foundation stability have 

included the following (Holtz 1989): 

• Removing the soft soil and replacing it with compacted select fill

• Improving the strength of the weak foundation soils through incremental surcharging

either with or without vertical drains, to speed consolidation and the development of

the shear strength

• Use of deep foundations to transfer some of the load through the weak foundation

deposits

• Use of stabilizing berms adjacent to the embankment or flatter side slopes

• Use of some form of ground modification, such as stone columns or soil mixing, to

reinforce the weak foundation soils

All of the above solutions require extra time, cost, or the acquisition of additional right of 

way to allow for construction of the fill. Some of these solutions also require specialized 

equipment and experienced labor. In these situations, it may be more advantageous to use a 

lightweight fill material so as to reduce the driving forces, thereby increasing the overall 

global stability of the fill. The reduction in driving force will depend upon the type of 

lightweight fill material used. The geotechnical properties of various types of lightweight fill 

materials are discussed in more detail in Section 4. 

A secondary benefit of the use of lightweight fill material is the reduction in settlement under 

loading. The amount of settlement will be reduced proportionately to the reduction in load. 
3-8 

2.1.2 Shear Strength

Granular lightweight fills have an angle of shearing resistance similar to natural soils, while 

cellular lightweight fills are characterized by a compressive strength. These properties result 

in internal stability within the lightweight fills. In the case of an embankment over a weak 

foundation, the shearing surface will penetrate through the lightweight fill, and the shear 

strength developed within the lightweight fill deposits will tend to increase the overall global 

stability.

2.1.3 Compressibility 

Many lightweight fill materials, such as cellular concrete, ESCS, fly ash, boiler slag, and aircooled slag have a compressibility and elasticity similar to natural soils or rock. Under static 

loading, the amount of internal compression within the fill will often be similar to that for 

conventional earth fill materials. Under dynamic loading, the resiliency of the lightweight 

materials will often be similar to the natural soils.

For certain lightweight fill materials, such as shredded tires or wood fibers, the 

compressibility of these materials under load is significantly larger than compacted soils, and 

this must be taken into account in the design. Furthermore, the resiliency of these materials is 

much larger than conventional soils. This property requires a cover of at least 3 feet of soil to 

be placed over the top of these deposits so as to reduce the resiliency, thereby enabling 

conventional pavements to be built on these deposits. 

EPS-block geofoam compressibility or stress strain behavior is generally linear to stress 

levels of about 0.5 percent. Beyond that, yielding occurs and the material is subject to timedependent creep. 

2.1.4 Lateral Pressures 

The lateral earth pressure at any depth is a function of the vertical overburden pressure 

multiplied by the coefficient of earth pressure and then reduced by the cohesion of the 

deposit. In the case of lightweight fills such as cellular concrete or EPS-block geofoam, the 

cohesion of the material is high and the densities are relatively low. Both of these factors 

tend to greatly reduce the amount of lateral earth pressure that is transmitted to adjacent 

structures such, as retaining walls, tunnels, or pile foundations below bridge abutments. 
3-9 

2.1.5 Drainage Characteristics 

Many of the granular lightweight fill materials have excellent drainage characteristics. This is 

favorable for pavement design because the presence of free water in the subgrade results in

the following: 

• High pore water pressure under traffic loading, which tends to weaken the soil

• Pumping at joints in reinforced concrete pavements

• Plastic deformation of the subgrade soils, thereby leading to rutting

• Heaving of the pavement in areas where frost penetrates into the subgrade

Good drainage is also beneficial behind a retaining wall to eliminate hydrostatic pressures.

2.1.6 Construction in Adverse Weather

It is difficult, if not impossible, to place and compact conventional soils during extremely 

cold or wet weather. However, some lightweight fills, such as EPS-block geofoam, ESCS, 

and low-density cellular concrete, have been successfully installed in poor weather. 

2.1.7 Seismic Considerations

In Japan, there have been case histories where a highway embankment constructed of 

geofoam did not fail in a severe earthquake, even though adjacent sections of a soil 

embankment did (Expanded Polystyrene Development Organization 1994). The lower unit 

weight of the material results in lower inertial forces under seismic loading. 

2.2 Advantages and Potential Disadvantages 

All lightweight fill materials have limits on their applicability. The advantages and potential 

disadvantages of all lightweight fill materials, and of individual materials as applicable, are 

listed and discussed below. 

2.2.1 Advantages 

Lightweight fills may be used in several different applications and to perform a variety of 

functions, as discussed above. Thus, the primary advantages of using lightweight fills may 

include one or more of the following: 

• Accelerate construction

• Reduce structural resistance requirements for lateral loads
3-10 

• Reduce structural resistance requirements for vertical loads

• Reduce magnitude of and time for consolidation of soft soils and settlement of

embankments and structures

• Eliminate or reduce stability problems of embankments and structures

2.2.2 Potential Disadvantages 

There are certain limitations to the use of lightweight fill materials. These limitations can be 

overcome with proper evaluation, design, and construction techniques. The limitations 

include the following: 

• Availability of the materials. Certain geographic areas may have an abundance of one

type of lightweight fill material, but not of another. As an example, wood fiber fill

would be available in lumber producing areas, fly ash and slag in heavily

industrialized areas, and ESCS in areas where production plants are present. Unless

the lightweight fill material is available on a local basis, the transportation costs could

raise the price considerably, and make these materials non-competitive.

• Construction methods. In general, all lightweight fill materials involve some special

procedures with regard to handling, transportation, placement and compaction. Some

lightweight fill materials could be difficult to place and handle. Fly ash behaves as

silt. When wet, it is very spongy, and when dry it will require dust control. Tire

shreds are very resilient when placed, thereby requiring somewhat unconventional

compaction procedures. Lightweight cellular concrete requires the use of specialized

equipment at the site to introduce air and other additives into the mixture before

placement.

• Durability of the fill deposits. Some lightweight fill materials (e.g., EPS-block

geofoam) must be protected to ensure longevity. Because EPS-block geofoam is

subject to deterioration from hydrocarbon spills, a concrete slab or a geomembrane is

generally placed over the surface of the blocks. Wood fibers can decay over a long

period of time, although recent studies indicate that the deterioration is limited to the

outer surface of an embankment. Fly ash deposits need to be protected with a soil

surface to minimize or prevent erosion of the side slopes.

• Wood fiber fills will undergo creep settlement for several years (WSDOT 2015), and

distress in and of maintenance of overlying pavement should be anticipated.

• Environmental concerns. Some of the lightweight fill materials may generate leachate

as water passes through these deposits. Fortunately, design methods and limitations

on applicability have been developed to minimize the amount of leachate, and, to
3-11 

date, these measures have worked satisfactorily. However, the additional cost of these 

measures should be considered during design. 

• Geothermal properties. Most lightweight fill materials possess geothermal properties 

that are different than soil. This can lead to accelerated deterioration of flexible 

pavements and/or problems with differential icing of pavement surfaces due to an 

alteration of the heat balance at the earth's surface. Essentially, most lightweight fill 

materials act as thermal insulation, even though this is not an intended or desirable 

function. However, this can be effectively controlled by placing a suitable thickness 

(20 to 36 inches minimum) of soil and/or paving materials over the surface of the 

lightweight fill material.

2.3 Alternative Solutions 

The feasibility of using a particular lightweight fill for a project need depends upon the 

function(s) of the modification and the method(s) selected to carry out the function. 

Feasibility evaluation includes the identification and evaluation of: technical issues, project 

development/delivery methods, performance criteria and quality assurance procedures, and 

non-technical issues that affect the utilization of ground improvement and geoconstruction 

technologies. A generalized summary of the evaluation process for use of ground 

modification technologies and materials was presented in Chapter 1.
3-12 

3.0 LIGHTWEIGHT FILL MATERIALS 

3.1 Introduction

Lightweight fill materials include both manufactured and recycled materials. Manufactured 

materials consist of products such as EPS-block geofoam that are specifically developed or 

manufactured for special design situations. These materials generally cost more than the 

recycled materials, but have unique properties that satisfy certain needs.

Recycled materials are generally the product left over from some industrial or commercial 

process. In the past, some of these materials were considered waste products and were 

landfilled or stockpiled. It has been found through experimentation that many of these 

materials have properties, such as low density, low unit cost, or relatively high permeability, 

that are desirable for fill construction. The use of these materials in roadway embankments 

has increased during the past 30 years. 

For design and construction purposes, lightweight fill materials can be grouped into two 

broad categories: i.e., materials that have an inherent compressive strength and behave 

similarly to cohesive soils in undrained loading, and materials that behave and have 

properties similar to granular soils.

3.2 Lightweight Fill Materials with Compressive Strength

3.2.1 Geofoam 

Geofoam is the generic name for any cellular material used in geotechnical applications. The 

most common geofoam materials used as a lightweight fill, are block molded expanded 

polystyrene (EPS) and extruded polystyrene (XPS), both plant manufactured. EPS-block 

geofoam is widely used by state transportation agencies. 

The manufacturing process for EPS consists of expanding small beads of polystyrene into 

spheres that contain numerous closed cells. The expanded spheres are then fused into blocks 

in a heated vacuum chamber. The heat welds the expanded beads together to form a very 

light material with a high void content. The final product has an extremely low density, but a 

relatively high strength and stiffness. The EPS blocks are similar to a large brick, but with 

typical dimensional ranges of 4 to 16 feet long, 12 to 48 inches wide and 24 to 48 inches 

thick. Different block size can be made depending upon the size of the manufacture’s mold. 

Blocks can be factory- or field-cut to any size or shape.

Detailed design guidelines and recommended standards for the use of EPS-block geofoam in 

embankments and bridge approach fills on soft ground are presented in NCHRP Report 529 
3-13 

(Stark et al. 2004a). The companion document, NCHRP Web Document 65 (Stark et al. 

2004b) provides a comprehensive treatment on the use EPS-block geofoam in embankments 

and includes design examples, construction practices, design details, case histories, and 

economic analysis. A summary of this NCHRP work on EPS-block geofoam was published 

by the Transportation Research Board (TRB 2013). 

Construction of an EPS-block geofoam embankment is shown in Figure 3-1. 

Figure 3-1. Embankment construction with EPS-block geofoam. 

3.2.2 Lightweight Cellular Concrete 

Lightweight cellular concrete is generally produced by introducing preformed foam (similar 

in appearance to shaving cream) into cement water slurry. The preformed foam is especially 

designed for concrete, and creates a network of discrete air cells within the cement matrix. 

Fly ash may be used as a partial cement replacement in the mixture. The cement water slurry 

is job-site produced, or thoroughly mixed at a plant and transferred in a ready-mix truck. The 

mixture may include some sand. 

After these materials, (cement, water, preformed foam), are blended to the specified density 

and thoroughly mixed, the resulting slurry is pumped into place. The lightweight cellular 
3-14 

concrete is generally cast in lifts ranging from 1 to 4 feet, or more, depending upon the 

specific application. Since it is cementitious in nature, hydration solidifies it, and vibratory 

compaction is not required. Subsequent lifts generally can be placed after a minimum of 12 

hours of curing. 

The quality of low-density cellular concrete is monitored through its cast density, starting 

with the wet cast density. The compressive strength is directly related to the cast density of 

the mixture.

Typically, specialized firms that supply their proprietary foaming agents will blend all 

materials on the job site and pump it to the required location. The foamed concrete can be 

produced with only one experienced person and the mixing equipment at the site, using a 

cement and water mixture delivered by a concrete supply company.

Lightweight cellular concrete backfill of an MSE wall (Chapter 10) is shown in Figure 3-2. 

Photo courtesy of MixOnSite USA, Inc.

Figure 3-2. Foamed concrete MSE wall backfill for light rail project in California.
3-15 

3.3 Granular Lightweight Fills

3.3.1 Tire Shreds 

Tire shreds (a.k.a. shredded tires and tire derived aggregate (TDA)) are produced by 

mechanically cutting tires into chips that are generally in the size range of 4 to 8 inches. 

These chips are durable, coarse grained, free draining, and have a low compacted density. 

Each cubic yard of tire shred fill contains about 75 passenger car size waste tires, so there is a 

potential for using a large number of tires in highway construction. Both steel- and glassbelted tires are used to produce tire shred aggregate. 

An application of tire shred embankment fill is illustrated in Figure 3-3. 

Photo courtesy of Gale-Tec Engineering

Figure 3-3. Embankment fill construction with tire shreds in Blue Earth County, 

Minnesota.

Here, a soil fill embankment that was under construction started moving, and threatening the 

rail line below it, due to underlying soft clay and peat deposits. Lightweight tire shred fill, 

wrapped in a geotextile, was used to replace soil fill and achieve a stable embankment (Gale 

et al. 2013). 
3-16 

Historically, the research and development of engineering criteria for use of tire shred 

lightweight fill was led by Professor Dana Humphrey and colleagues at the University of 

Maine in the 1990s. In an early FHWA (1997) User Guideline, it was noted that at least 15 

states have used tire shreds as lightweight fill and that more than 70 successful projects had 

been completed on state local or private roads (Humphrey 1996). Some projects used just the 

tire shreds as embankment fill and some used tire chips blended with soil. The 1997 

guideline also noted unresolved issues of: (i) causation of exothermic reactions that resulted 

in three scrap tire embankment fires in 1995; (ii) optimum particle size and shape of tire 

shreds; (iii) further investigation into the engineering properties of blended soil and tire shred 

fills; and (iv) field quality control methods for tire shred embankment construction. A recent 

survey of states (Stroup-Gardiner and Wattenberg-Komas 2013a), with 45 agencies 

reporting, reports that 14 states are currently using tire shreds for embankment construction. 

The engineering properties of tire shreds have been reported in technical publications and 

have been summarized by Humphrey (1998). The use of tire shreds for transportation 

embankment construction has been evaluated by Washington DOT (Baker et al. 2003) and by 

Texas Tech University (Sonti et al. 2003) for the Texas DOT. These evaluations included 

lesson learned from previous embankment construction, including some instances of 

spontaneous combustion of tire shred fills. The NCHRP 435 synthesis (Stroup-Gardiner and 

Wattenberg-Komas, 2013d), summarizes physical, chemical, environmental, and engineering 

properties of scrap tire byproducts; and various applications including embankment fill. A 

recent CalRecycle publication (Cheng 2016) provides application summaries and 

construction guidelines for TDA fills.

The state of practice for embankment construction with tire shred aggregate is presented in 

ASTM D6270 (2012) Standard Practice for Use of Scrap Tires in Civil Engineering 

Applications. Today, tire shred embankment fills generally use ASTM D6270 Type A 

gradation and are: 

• limited to a 10-foot maximum thickness, 

• encased in a geotextile, 

• covered with soil, and 

• used above the groundwater table. 

Additional design and construction guidance is provided in the New York State DOT GEM20 manual (2015a). The issues noted in the 1997 FHWA guideline are addressed within the 

ASTM standard and in the GEM-20 manual, and in the recent reports noted above. 

Additional discussion on the use of tire shred fills beneath the groundwater table are 
3-17 

presented in ASTM D6270, NCHRP 435 (Stroup-Gardiner and Wattenberg-Komas 2013d), 

and in a MN DOT reports (Oman et al. 2013; Edstrom et al. 2008). 

Since the tires are a waste product, the primary costs in some locations are associated with 

the shredding process and the transportation to the project site. In many locations, the product 

is used as a fuel source. In some states, there is rebate incentive for using scrap tires. Thus, 

the economics will depend upon the location. 

Scrap tires in two other configurations are also used for embankment construction, though 

not as frequently as tire shred fills. Tire shreds can be mixed with soil to create a composite 

fill for embankment construction. Information on soil-tire chip blended fills can be found in 

the Texas Tech University (Sonti et al. 2003) and the NCHRP 435 synthesis (StroupGardiner and Wattenberg-Komas 2013d). 

Scrap tires can also be bound together to create tire bales. A portable baling machine can 

compresses approximately 100 passenger car sized tires into a bale, bound with galvanized or 

stainless steel wire, which is about 2 cubic yards in volume and weighing about 1 ton. Thus, 

the bale has a unit weight of about 40 pcf. See the Colorado DOT Research report (Zornberg 

et al. 2005) for a summary on use of tire bales, including design, specification, cost, and 

performance monitoring discussions. 

3.3.2 Wood Fiber 

Wood fiber includes any type of wood waste generated from the handling of logs at a 

sawmill. This includes hog fuel, sawdust, and planer chips. Hog fuel is primarily bark and 

some pulverized wood that has been stripped from logs at a pulp mill. Sawdust is the small 

size pieces of wood that result from sawing of lumber. The excess material removed when 

lumber is sawed to prescribed sizes is called planer chips.

The availability of wood fiber for fill construction is geographically limited to locations 

where a significant timber industry exists. A description of the type of wood fiber used as 

lightweight fill can be found in reports by Washington DOT (WSDOT) (Kilian and 

Ferry1993; Allen and Kilian 1993). Placement of wood fiber in embankment construction is 

illustrated in Figure 3-4. 
3-18 

Figure 3-4. Embankment construction with wood fiber.

A summary of WSDOT use of wood fiber fill since the first construction in 1972 is presented 

in Baker et al. (2003). Over 20 projects were identified. The majority of applications for the 

lightweight wood fiber fill were landslide corrections, and six applications was embankment 

construction over soft soils. The average age of the projects reviewed is about 26 years. 

WSDOT notes it will continue using wood fiber as a primary source of geotechnical 

lightweight fill.

3.3.3 Expanded Shale, Clay, and Slate (ESCS) 

Synthetic aggregates can be produced from shale, clays or slate by heating these materials in 

a rotary kiln to temperatures in the range of 1,000 to 1,200 degrees C. In this process, the 

clay minerals of montmorillonite, illite, and kaolinite become completely dehydrated and will 

not re-hydrate under atmospheric conditions. The expanded vitrified mass is then screened to 

produce the desired gradation for a particular application. The particle shape may be 

rounded, cubical, or sub-angular in shape. This aggregate may have a hard ceramic outer 

shell with a system of pores that vary in size and are largely non-connected. The particles are 

durable, chemically inert, and relatively insensitive to moisture (will absorb and retain some 

water).
3-19 

The cost for producing ESCS materials is relatively high (though comparable in cost to other 

manufactured lightweight fill materials). For this reason, these products have generally been 

used as lightweight aggregates for structural concrete, but are used for some embankment 

construction. However, in areas where high-quality, naturally occurring aggregates are no 

longer present, the ESCS have been used to produce synthetic aggregates for normal roadway 

construction. 

The placement of expanded clay lightweight fill during the early stages of embankment 

construction is shown in Figure 3-5. 

Figure 3-5. Placing expanded clay fill. 

3.3.4 Fly Ash 

When coal is burned, the finer portion of the residue is airborne and carried to the smoke 

stack where it is collected via a precipitator. This portion of the waste product is classified as 

fly ash. Fly ash is defined as the finely divided residue that results from the combustion of 

ground or powdered coal and is transported from the combustion chamber by exhaust gases. 

Fly ash consists of very fine particles that are predominantly in the silt-sized gradation range. 

The particles are rounded and consist of a siliceous material.
3-20 

The chemical properties of fly ash depend upon the type of coal that is burned. ASTM 

standard C618 defines two categories of fly ash. Class C fly ash is produced from lignite or 

sub-bituminous coal, and has pozzolanic and some cementitious properties. Some Class C fly 

ashes may contain lime contents higher than 10 percent. 

Class F fly ash is produced from burning anthracite or some bituminous coals that meet the 

applicable requirements given in ASTM C618. This class of fly ash has pozzolanic 

properties. 

In recent years, the primary use for Class C and Class F fly ashes has been as an ingredient in 

concrete to reduce the cement content. This demand has raised the cost of the fly ash to the 

level where it is less competitive with other lightweight fill materials. Fly ash is sometimes a

partial cement replacement in lightweight cellular concrete. 

3.3.5 Blast Furnace Slag 

Blast furnace slag is generated from the production of iron. In the process, the blast furnace is 

charged with iron ore, limestone and/or dolomite flux, and coke for fuel. Two products are 

obtained from the furnace: molten iron and slag. The slag consists primarily of the silica and 

alumina from the original iron ore combined with the calcium and magnesium oxides from 

the flux stone. The slag leaves the furnace in a form of liquid resembling lava. Depending 

upon the manner in which the molten slag is cooled and solidified, three distinct types of 

blast furnace slag can be produced: 

• Air-cooled slag. The molten slag is permitted to run into a pit. Solidification takes 

place under the prevailing atmospheric conditions, after which cooling may be 

accelerated by water spray. After the pit has been filled and cooled, the slag is dug, 

crushed, and screened to the desired aggregate sizes. The resulting slag aggregate is 

angular and vesicular. The uniformly graded slag has the lowest density, which is 

desirable for use as a lightweight fill material.

• Expanded slag. Treatment of the molten slag with controlled quantities of water 

accelerates the solidification and increases the cellular nature of the slag, producing a 

lightweight product. Either a machine or pit process may be used to mix the water and 

molten slag.

• Granulated slag. Molten slag that is quickly chilled will form a glassy granular 

product called granulated slag. This process is the most rapid of the cooling methods, 

and little or no crystallization occurs. The granulated slag may be crushed and 

screened or pulverized for various applications. 
3-21 

Air-cooled blast furnace slag is the predominant form of slag processed in the United States, 

accounting for 90 percent of the blast furnace slag sales in 1992 (Solomon, 1993). The 

expanded slag and granulated slag are of lower unit weight and are more expensive to 

produce. Their use is generally limited to aggregates for lightweight concrete or concrete 

block. 

The principal constituents of blast furnace slag are the oxides of silica, alumina, lime, and 

magnesia. These oxides comprise 95 percent or more of the total. The remaining portion of 

the slag contains manganese, iron, and sulfur compounds. Certain slag containing high levels 

of soluble sulfur can be very acidic and, therefore, corrosive. 

Steel slag is another slag produced from the steel-making operations. Steel slag is undesirable 

for mass filling, such as embankments, because it contains the oxides of calcium and 

magnesium, which are expansive. This expansive nature has resulted in heave of concrete 

slabs. In the past, many slag piles consisted of a mixture of both blast furnace and steel slag, 

but after the undesirable effects of expansion from the steel slag became known, the industry 

began to separate slag piles so that the blast furnace slag could be processed for roadway 

construction. Steel slag is used in asphaltic concrete road surfacing mixes. 

3.3.6 Boiler Slag 

Boiler slag is a coal combustion byproduct from power plants. It is obtained from molten ash 

collected in wet, water cooled bottom boilers. It has been estimated that approximately 25 

percent of all power plant ash produced in the United States is boiler slag, 65 percent is fly 

ash, and 10 percent is dry bottom ash and cinders. Dry bottom ash is highly absorptive and 

not suitable for most embankment fills. However, boiler slag is a durable aggregate. It is 

formed when the slag flows from the furnace in a hot molten condition and is discharged in 

cold water where it crystallizes, solidifies, and forms angular black glassy particles usually 

less than 1⁄8-inch to 3⁄8-inch in size. It is generally composed of silica and ferric oxide particles 

of angular and irregular shape. 
3-22 

4.0 DESIGN CONCEPTS

4.1 Design Considerations 

The function of lightweight fills is to reduce the load on the foundation. A number of factors 

must be evaluated in order to determine appropriate candidate(s) fills. These include:

• Availability of lightweight fill materials in the geographic area of the project.

• The geotechnical engineering properties of the lightweight fill material for use in the 

design evaluation. The same, or similar, engineering analyses used for conventional 

fills would be used to evaluate the suitability of the design using granular lightweight 

fill materials. For example, if slope stability is a concern, the reduced density of the 

lightweight fill plus the angle of shearing resistance or cohesion of the lightweight fill 

should be included into the slope stability calculations (Holtz 1989; Stark et al. 

2004a). If settlement is a concern, the reduced density of the fill should be used in the 

settlement calculations. For fills characterized by compressive strength (EPS-block 

geofoam, lightweight cellular concrete), some unique design criteria must be 

evaluated (Stark et al. 2004a). 

• Drainage requirements and drainage characteristics of lightweight fill materials. 

• An evaluation of the durability, compressibility, water absorption potential, 

degradation potential, and other unique characteristics of the lightweight fill 

materials.

• Design and construction considerations. It may be necessary to incorporate certain 

features into the design to compensate for potential environmental problems or to 

reduce erosion potential. Construction with lightweight fill materials may require that 

certain field procedures be followed.

• An evaluation of the costs for using lightweight fill versus conventional construction. 

Costs for various types of lightweight fill projects are discussed in Section 7. 

However, the cost will vary for each particular site depending upon the availability of 

the lightweight fill materials at the project site.

• An evaluation of the costs for the various lightweight fill materials. Costs generally 

increase as unit weight of materials decrease, so the design needs should be balanced 

with the design requirements. 

4.2 Design Procedures

As a guide for preliminary planning and design, Tables 3-2 through 3-10 summarize some of 

the important geotechnical engineering parameters for lightweight fill materials. 
3-23 

Environmental, construction, and design considerations are also listed in these tables. Each of 

these factors should be considered for designs using lightweight fill materials. 

The design considerations are guidelines that experience has proven to be appropriate when 

working with a particular lightweight fill material. Construction guidelines refer to 

techniques that have been developed in the field to achieve proper densification and to 

minimize construction problems. 

The information presented in these tables should be used as guidelines only. For each 

specific project site, additional testing should be performed to determine the material design 

properties of the lightweight fill material that will be used. This is especially true for the 

recycled materials, since the chemical composition of these materials, as well as the 

gradation or shape after recycling, could vary from one source to another. 

In the case of manufactured lightweight fill materials, the minimum compressive strength or 

the maximum density of the product can be specified in advance, and the final product is 

relatively uniform in consistency. In this case, testing would confirm that the desired strength 

or density has been achieved. 

4.2.1 Fills with Compressive Strength

4.2.1.1 EPS-block Geofoam

The most comprehensive design, material, and construction guidelines on the use of EPSblock geofoam for highway construction are presented in NCHRP Report 529 (Stark et al., 

2004a), with particular application to highway embankments. Comprehensive guidelines 

including design examples and design detailing are contained in the report companion 

document NCHRP Web Document 65 (Stark et al. 2004b). An earlier monogram, by Horvath 

(1995), provided (then current) comprehensive information on the use of EPS-block geofoam 

in lightweight fill applications. A subsequent NCHRP study provides guidelines for EPSblock geofoam applications specific to slope stability projects (Arellano et al. (2011). 

Design of EPS-block geofoam embankments must address both strength and serviceability. 

The NCHRP Report 529 (Stark et al., 2004a) presents detailed design procedure flowcharts 

for EPS-block geofoam embankments and design charts for various embankment 

configurations and design loadings. Typical EPS-block geofoam applications of stand-alone 

embankments are shown in Figure 3-6. 
3-24 

Horvath 1995, from Stark et al. 2004a

Figure 3-6. Typical EPS-block geofoam applications involving standalone 

embankments.

The overall design process when using EPS-block geofoam is divided into three phases in 

order to consider the interaction between the three major components in the embankment 

(Stark et al. 2004a).

1. Design to preclude external (global) stability of the embankment. This should include 

considerations for settlement, bearing capacity, and slope stability under the projected 

loading conditions. 

2. Design for internal stability within the embankment mass. The design must ensure 

that the EPS-block geofoam mass can support the overlying pavement system without 

excessive immediate and creep compression.

3. Design of an appropriate pavement system for the subgrade provided by the 

underlying geofoam blocks. 

The major components of an EPS-block geofoam slope system are shown in Figure 3-7. 
3-25 

Arellano et al. 2011. Copyright National Academy of Sciences. All rights reserved.

Figure 3-7. Major components of an EPS-block geofoam slope system. 

The design of such a system must address the same three failure modes as an EPS-block 

geofoam embankment system, namely external stability, internal stability and the pavement 

support system. Detailed design procedure flowcharts for EPS-block geofoam slope systems 

are provided in Arellano et al. (2011) and NCHRP RRD 380 (2013). A complication in the 

evaluation of slope systems is that multiple potential slip surfaces must be analyzed including 

potential slip surfaces above and below the EPS-block geofoam and global stability failure 

that encompasses the EPS-block geofoam (see Figure 3-8). 

Arellano et al. 2011. Copyright National Academy of Sciences. All rights reserved.

Figure 3-8. Multiple potential failure surfaces to consider with EPS-block geofoam 

slope systems. 

Stability analyses of embankment and slope systems require the modeling and quantifying of 

both the internal shear strength of the EPS-block geofoam and of the geofoam interfaces. The 

internal shear strength of EPS-block geofoam correlates to its compressive strength. The 

interfaces typically include geofoam to geofoam, geofoam to soils, and geofoam to 
3-26 

geosynthetic material. Interface friction is an important stability design consideration, 

particularly under horizontal wind, water and/or seismic loading conditions. 

EPS-block geofoam embankments and slope systems often support an overlying roadway 

pavement. The objective in the design of an appropriate pavement system is to select the 

most economical arrangement and thickness of pavement materials for the subgrade provided 

by the supporting EPS blocks. Equivalent soil subgrade strengths for the EPS blocks can be 

used with traditional pavement design procedures. Subgrade properties as a function of EPS 

block density are listed in Table 3-2 (Stark et al. 2004a).

Table 3-2. Equivalent Soil Subgrade Values of EPS-Block Geofoam for Pavement 

Design

EPS Block 

Density

(pcf)

CBR

(%)

Young’s 

Modulus

(psi)

Resilient 

Modulus

(psi)

1.25 2 725 725 

1.5 3 1015 1015 

2.0 4 1450 1450 

The use of EPS-block geofoam requires the use of materials standards to define material 

properties. There are two materials standards are currently in use by state departments of 

transportation from ASTM International and from NCHRP reports. 

ASTM International has three standards for the use of RCPS geofoam. ASTM D6817 covers 

physical properties and dimensions, including densities and compressive resistance at 1% 

strain. There are seven grades of EPS listed in ASTM D6817 (ASTM 2015) that range in 

density from 0.70 to 2.85 pcf, with compressive resistance values of 2.2 to 18.6 psi, 

respectively. Six grades of XPS are listed with densities and compressive strengths ranging 

from 1.20 to 3.00 pcf and 2.9 to 40.6 psi, respectively. ASTM D7180 (ASTM 2013a) covers 

design considerations for the use of EPS-block geofoam in geotechnical projects. ASTM 

D7557 (ASTM 2013b) covers quality assurance sampling issues for EPS-block geofoam. 

NCHRP Report 529 (Stark et al. 2004a) provided a recommended combined material, 

product, and construction standard covering EPS-block geofoam for use as lightweight fill in 

stand-alone road embankments and related bridge approach fills on soft ground. This 

standard contained recommendations on manufacturing quality assurance, product shipment 

and storage, construction quality assurance, site preparation, placement of EPS-block 

geofoam, and pavement construction. The recommended standard was intended to be used to 

create a project-specific specification. Nichols (2008) noted that the standard had been 

included in FHWA publications as a recommended design and construction standard. In the 
3-27 

follow-on study involving EPS-block geofoam slope systems (Arellano et al. 2011), this 

standard was updated based on experience in the use of EPS-block geofoam in embankments 

and extended to include EPS-block geofoam slope systems. Key design and construction 

considerations for EPS are summarized in Table 3-3. 

Table 3-3. Expanded Polystyrene (EPS) Design and Construction Guidelines

Item Guideline

Design 

Parameters

Density, Dry: 0.75 to 2.0 pcf 

Compressive and Flexural Strength: Varies with density, 6 to14 psi 

Modulus of Elasticity: 580 to 1450 psi 

California Bearing Ratio (CBR): 2 to 4

Coefficient of Lateral Earth Pressure: Lateral pressures from adjacent soil 

mass may be reduced to a ratio of 0.1 of horizontal to vertical pressure (PIARC 

1997). 

Environmental 

Considerations

There are no known environmental concerns. No decay of the material occurs 

when placed in the ground.

Design 

Considerations

EPS blocks will absorb water when placed in the ground. Blocks placed below 

water have resulted in densities of 4.8 to 6.4 pcf after 10 years. Blocks above the 

water had densities of 1.9 to 3.2 pcf after 10 years. For settlement and stability 

analyses, use the highest densities to account for water absorption. 

Buoyancy forces must be considered for blocks situated below the water table. 

Adequate cover should be provided to result in a minimum safety factor of 1.3 

against uplift.

Because petroleum products will dissolve geofoam, a geomembrane or a 

reinforced concrete slab is used to cover the blocks in roadways in case of 

accidental spills.

Differential icing potential of pavement, due to a cooler pavement surface above 

the EPS versus pavement above a soil only subgrade. Differential icing can be 

minimized by providing a sufficient thickness of soil between the EPS and top 

of pavement surface.

Use side slopes less than or equal to 2H:1V and a minimum cover thickness of 

0.8 feet. If a vertical face is needed, cover exposed face of blocks with shotcrete 

or other material to provide long-term UV protection. 

Construction 

Considerations

The subsoil should be leveled before placement of geofoam blocks. A layer of 

sand/gravel is frequently placed as a leveling course.

When multiple layers of geofoam blocks are placed, the blocks should be placed 

at right angles to avoid continuous vertical joints and to promote interlocking. 

See Figure 3-9. A minimum of two layers of blocks must be used.

Provide a mechanical connection between blocks using a galvanized barked 

plate for shear transfer.

Place cover material over geofoam blocks as soon as possible to prevent 

displacement from wind or buoyancy. Avoid prolonged exposure to sunlight, 

which embrittles EPS.

The NCHRP combined material, product, and construction standards and the ASTM material 

standards have been beneficially applied to geofoam projects in roadway fill, slope systems, 
3-28 

and other applications. However, the presence of two standards has caused some confusion to 

users (Nichols 2008). In addition to differences in material designations, the primary 

difference between the two material standards is the “unit weight to compressive strength” 

relationships for block designations. There is long-standing industrial acknowledgment that 

EPS unit weight could be reliably indexed to the elastic limit stress – measured as the 

compressive stress at 1% strain – of the material. However, there is currently disagreement 

within the industry as to what elastic limit stress can be practically achieved for a given block 

unit weight, and what that relationship means with respect to the performance of the blocks 

in a specific fill application (Nichols 2008). Additional discussion of the use of the two 

standards in highway applications can be found in Nichols (2008), Arellano et al. (2011), and 

Horvath (2012). 

After Stark et al. 2004b

Figure 3-9. Isometric view of typical EPS block layout. 

4.2.1.2 Cellular Lightweight Concrete

Cellular lightweight concrete (a.k.a. foamed concrete) is placed as a liquid product that is 

practically self-leveling, and can be pumped over a distance as great as 3,300 feet. The 

cellular lightweight concrete will begin to harden between 2 to 6 hours after placement, and 

generally solidifies in 24 hours. Design with this product is analogous to design with 

conventional concrete. The maximum cast unit weight and related minimum compressive 

strength should be specified as dictated by design and with consideration of local suppliers of 

cellular lightweight concrete. The range of wet cast density and compressive strength that can 

be specified generally can range from 20 to 80 pcf and 10 to 300 psi, respectively. 
3-29 

Some additional design considerations for pavement support applications include the 

following (CROW 2003): 

• The heat conductivity of foamed concrete is limited. Therefore, it may not be 

advisable to construct flexible pavements directly over the product. 

• The dynamic Modulus is nearly equal to the static modulus. To account for fatigue 

under repeated loads, the tensile strength should consider a factor of safety of about 2. 

Additional tensile capacity can be obtained by the inclusion of steel mesh 

reinforcement.

• Poisson’s ratio for foamed concrete is about 0.20. 

• The cost of foamed concrete increases with cast density. 

Key design and construction considerations are summarized in Table 3-4. 
3-30 

Table 3-4. Cellular Lightweight Concrete Design and Construction Considerations

Item Guideline

Design 

Parameters

Wet Density Range: 20 to 80 pcf

Compressive Strength Range: 10 to 300 psi, depending on density 

Water Absorption:1.4 to 15 psf, depending on density 

Freeze-thaw Resistance, 100 Cycles: 92 to 98%, depending on density 

Coefficient of Lateral Earth Pressure: Negligible for vertical loads applied 

directly over the foamed concrete. Lateral pressures from adjacent soil mass 

may be transmitted undiminished. 

Environmental 

Considerations There are no known environmental concerns. 

Design 

Considerations

Dry density values will be lower than wet density values.

Buoyancy could be a problem if foamed concrete is placed below the water table 

and there is not sufficient vertical confinement. 

The lower compressive strength mixes are affected by freeze-thaw cycles. The 

product should be used below the zone of freezing or a higher compressive 

strength used. Densities greater than 37 pcf have reported excellent freeze-thaw 

resistance.

There is some absorption of water into the voids, which could affect the density 

and compressive strength. Saturation by water should be prevented by 

construction of a drainage blanket and drains. 

Construction 

Considerations

A staging area is required for batching, mixing, and placing on site. 

The foamed concrete is very fluid; formwork should be tight to avoid flow of 

the mix through joints or gaps in the forms. Polyethylene film may be used to 

prevent leakage.

If the foamed concrete is placed in a confined area, forms are not necessary, as 

the fluid mix will flow to completely fill the void.

The lift thickness should not exceed 4 feet, as the heat of hydration would have 

an adverse effect on the foam. Allow a minimum 12-hour waiting period 

between lifts. No special provisions for cold joints are necessary, although each 

lift surface should be scarified and clean. If shaping is required, the lift thickness 

should be limited to 2 feet to allow workers to shape the surface while wading in 

the fluid mix. 

4.2.2 Granular Lightweight Fills 

The important design and construction criteria associated with granular lightweight fills are 

summarized in Tables 3-5 through 3-10. Design proceeds using the indicated variables and 

conventional geotechnical methods associated with granular soils.
3-31 

Table 3-5. Tire Shreds Design and Construction Guidelines

Item Guideline

Design 

Parameters

Dry Density: 21 to 53 pcf loose and 30-73 pcf compacted, with various of 

gradations and reporting sources 

Angle of Shearing Resistance: 19° to 30°

Cohesion Intercept: 100 to 230 psf, use 0 for design 

Compressibility: 5 to 40 percent vertical strain over a range of 200 to 4,200 psf 

vertical stress

Permeability: 0.5 to 60 cm/sec

Type A Gradation (ASTM D6270): 8-inch maximum dimension; 100% 

passing 4-inch, a minimum of 95% passing 3-inch, a maximum of 50% passing 

the 1.5-inch, and a maximum of 5% passing the 0.2-inch sieve

Coefficient of Lateral Earth Pressure: 0.25 to 0.47

Environmental 

Considerations

The design considerations listed below address minimizing leachate generation 

and transport from tire shred fills. See NCHRP 435 (Stroup-Gardiner and 

Wattenberg-Komas 2013d), ASTM (2012), Minnesota DOT (Edstrom et al. 

2008), Washington DOT (Baker et al. 2003) for additional information and 

discussion on environmental considerations. 

Combustion 

Potential

The tire shred gradation and design considerations were developed, in part, to 

prevent combustion of tire shred fills. These design details prevent or minimize 

the amount of infiltration of water and air into tire shred fill. 

Design 

Considerations

Limit layers to 10 feet in thickness.

Keep the tire shred fill above the water table.

Provide good surface drainage of roadway surface to avoid water seepage 

through the shredded-tire fill. 

Tire shreds should be separated from the surrounding soil by completely 

wrapping with a geotextile.

Metal fragments must be firmly attached to the chips, with 98 percent embedded 

in the rubber to prevent exposed wire strands from puncturing tires or 

construction equipment. 

Place a minimum 3-foot thick soil cap on the top and side slopes of the tire chip 

fill to minimize pavement deflections and provide confinement. 

Place 2-foot soil surcharge for 60 days to minimize post construction settlement 

due to compressibility of tire shreds. 

Top of tire shred embankment should be a minimum of 5 feet below the top of 

subgrade elevation.

Multiple 10-foot tire shred layers should be separated by 3 feet of soil fill.

Drainage pipes beneath the fill should be located at least 3 feet below the bottom 

of the of the tire shred layer.

Drainage features that could provide free access to air should be avoided at the 

bottom of the fill. 

Construction 

Considerations

Spread using a track-mounted dozer in a lift thickness of 3 foot or less.

Compact using sheep’s foot rollers, smooth drum rollers, or by repeated passes 

with a D-8 dozer. 

Use multiple passes of compaction equipment, since compressibility decreases 

after 5 to 8 cycles of loading.

Anticipate 35 percent volume reduction during compaction, plus 10 percent 

shrinkage under loading of soil cover and pavement base course. 
3-32 

Table 3-6. Wood Fiber Design and Construction Guidelines

Item Guideline

Design 

Parameters

Moist Density: 45 to 60 pcf 

Angle of Shearing Resistance:

Sawdust – 25° to 27°

Hogfuel – 31°

Wood Chips – 30° to 49°

Permeability: 1 x 10-5m/s

Compressibility: Loose volume reduces 40 percent on compaction.

Vertical subgrade reaction coefficient: 1300 to 1450 psi in top 2 feet, roughly 

corresponding to a CBR of 1 

Environmental 

Considerations

Potential environmental effects of the leachate include:

depletion of available dissolved oxygen in groundwater. 

lowering of groundwater pH because of acidic nature of leachate, which has pH 

of 4 to 6. 

potential contamination of water with toxins. 

Methods to reduce contamination include: 

reducing water infiltration into wood fiber by drains and capping. 

treatment of leachate.

barriers between wood fiber fill and adjacent bodies of water. 

Design 

Considerations

Restrict particle size to 6 inches maximum to prevent development of large 

voids. Less than 30 percent should be finer than 0.5 inches to minimize the use 

of fine uniform sawdust. 

Use fresh wood fiber to prolong the life of the fill. 

Use side slopes of 1.5H:1V or flatter. 

Employ surface treatment with cover material of thickness 2 feet or more to 

protect slope from erosion and minimize deterioration of wood fibers.

Restrict height of fill to about 16 feet and reduce air penetration into wood to 

minimize the possibility of spontaneous combustion. 

Construction 

Considerations

Truck-mounted equipment is used to spread fiber in 12 to 20-inch lifts.

Two passes with a fully loaded hauling truck weighing 33 kips or more is 

usually sufficient to properly compact wood fiber. 
3-33 

Table 3-7. Expanded Shale, Clay, and Slate (ESCS) Design and Construction 

Guidelines

Item Guideline

Design 

Parameters

Dry Density, Compacted: 50 to 65 pcf 

Dry Density, Loose: 40 to 54 pcf 

Angle of Shearing Resistance: 35° loose, 37° to 44° compacted

Grain Size Gradation: 3/16 to 1 inch 

Permeability: High

Coefficient of Subgrade Reaction: 33 to 37 pci loose, 140 to 155 pci 

compacted

Environmental 

Considerations There are no known environmental concerns. 

Design 

Considerations

The material will absorb some water after placement, when continually 

submerged. Samples compacted at a water content of 8.5 percent have been 

found after 1 year to have a water content of 28 percent. Over a longer period of 

time, the estimated long-term water content would be about 34 percent. 

Side slopes of embankments should be covered with a minimum of 2.5 feet of 

soil cover.

Use side slopes of 1.5H to 1V or flatter to confine the material and provide 

internal stability.

For calculating lateral earth pressures, use an angle of shearing resistance of 35°. 

Construction 

Considerations

Particle degradation can occur from steel-tracked construction equipment. Use 2 

to 4 passes with rubber-tired rollers and lift thickness of 3 feet or less.

Fill should be unloaded at side of fill area, then distributed with lightweight 

equipment with a contact pressure of 4.5 psi or less.

Field density may be approximated in the laboratory by conducting a modified 

one-point AASHTO T 272 density test.
3-34 

Table 3-8. Fly Ash Design and Construction Guidelines

Item Guideline

Design 

Parameters

Density Range, Compacted: 70 to 90 pcf 

Shear Strength: 33° to 40°, c = 0, for Type F; Class C is self-hardening, so the 

shear strength will vary as it cures

Permeability: Range of 1 x 10-6 to 1 x 10-9 m/s

Compressibility: Cc = 0.05 to 0.37, Ccr = 0.006 to 0.04 

Grain Size Range: 0.005 to 0.074 mm 

Specific Gravity: 1.9 to 2.5

Atterberg Limits: Non-plastic

Environmental 

Considerations

The leachate is alkaline, with pH of 6.2 to 11.5. Calcium, sulfate, and boron are 

soluble constituents, which can leach and migrate.

The EPA (Rittenhouse 1993) has stated fly ash as non-hazardous. 

Design 

Considerations

Where the groundwater table is high, a drainage blanket should be provided 

below the fly ash fill to promote a capillary cutoff and prevent frost heave and 

resiliency of the subgrade. Runoff from paved surfaces should be discharged 

into a drainage system. Surface waters from peripheral areas should be diverted 

away from the embankment to minimize infiltration into the fly ash. The side 

slope of embankments should be covered with at least 2 feet of soil to prevent 

erosion. 

If concrete is to be formed directly on fly ash, a polyethylene barrier should be 

placed on the fly ash to prevent moisture absorption from the fresh concrete and 

to serve as a moisture barrier. Use fly ash in the concrete to reduce sulfate 

attack.

Construction 

Considerations

Fly ash behaves like silt, thus, dusting will occur when dry, and compaction is 

difficult when wet. 

Some means for adding water should be available on site to keep the water 

content near optimum for compaction. 

Surface protection to minimize erosion may be required. 

Compaction is obtained with smooth drum vibratory rollers or self-propelled, 

pneumatic-tired rollers.

Use 10-inch lifts and compact the fly ash immediately after spreading.

The use of test strips to develop the most efficient compaction procedures is 

advisable.
3-35 

Table 3-9. Air-Cooled Blast Furnace Slag Design and Construction Guidelines

Item Guideline

Design 

Parameters

Compacted Moist Density: 70 to 94 pcf, varies with size and gradation 

Gradation: Can be graded to any specified size from 4 inches down. 

Angle of Shearing Resistance: 35° to 40°

Permeability and Compressibility: Depends on final specified gradation. 

Generally similar to gravel and sand.

Environmental 

Considerations

Slag contains small amounts of sulfur in combined alkaline compounds. The pH 

of water in contact with slag is generally in the range of 8 to 12, which tends to 

inhibit corrosion. 

Some washing of the aggregate may be required to control the pH to 11 or less 

to meet AASHTO specifications. There are no known environmental concerns. 

Design 

Considerations

The slag behavior is similar to natural angular gravel and sand deposits. 

The highest internal stability occurs for aggregate that is well graded with a 

maximum particle size of 16 inches. The amount passing #200 sieve should be 

limited to 5 to 7 percent. However, the density increases for well-graded 

materials. If lightweight fill is desirable, uniformly graded materials should be 

specified.

Absorption in slag is usually in the range of 1 to 6 percent by weight. 

Slag is highly resistant to weathering and abrasion, and can be placed below the 

water table and next to lakes and rivers.

Construction 

Considerations

Slag can be placed and compacted in the same manner as natural gravel and 

sand. 
3-36 

Table 3-10. Boiler Slag Design and Construction Guidelines

Item Guideline

Design 

Parameters

Dry Density, Loose: 60 to 78 pcf 

Dry Density, Compacted: 82 to 102 pcf 

Optimum Moisture: 8 to 20% (Stroup-Gardiner and Wattenberg-Komas 

2013b) 

Angle of Shearing Resistance: 38° to 42°

Coefficient of Permeability: 0.3 to 0.9 mm/s 

Grain Size Range (Percent Passing): 90 to 99% on #4, 62 to 89% on #8, 16 to 

46% on #16, 4 to 23% on #30, 2 to 12% on #50, 1 to 7% on #100, and 0 to 5% 

on #200 (Stroup-Gardiner and Wattenberg-Komas 2013b) 

Atterberg Limits: Non-plastic

Compressibility: Comparable to sand, at same relative density 

Environmental 

Considerations

After 4 days of soaking, the pH of the water solution is generally in the range of 

6.7 to 7.0. 

Barium has been detected by toxicity tests, but at levels well below the EPA 

specified standard.

There are no known environmental concerns with the use of this material. 

Design 

Considerations

The aggregate is durable and satisfies acceptable limits for soundness tests.

The aggregate works well as an underdrain filter material, provided the

gradation requirements are met.

Side slopes should be covered with a minimum of 2 feet of cover material since 

exposed material has low stability.

Specify standard proctor compaction, AASHTO T 99, since some degradation 

occurs during laboratory compaction in accordance with AASHTO T 180.

Construction 

Considerations

Compact with several passes of a pneumatic roller or a smooth-drum, vibratory 

roller. Keep water content at or above optimum water content, as determined by 

AASHTO T 99. 6 to 10 passes are usually sufficient. 

Material must be kept wet since there could be a loss in stability when material 

dries.
3-37 

5.0 CONSTRUCTION SPECIFICATIONS AND QUALITY ASSURANCE

5.1 Specification Development

A discussion on specification items and guide and/or example specifications for lightweight 

fill materials are contained on http://www.GeoTechTools.org website. Additionally, guidance 

on quality assurance methods and means for lightweight fills are contained on this website. 

For lightweight fill materials, such as EPS-block geofoam and lightweight concrete, that have 

inherent strength, the compressive strength is dependent on the plant manufacturing process 

for geofoam, while that of cellular concrete is usually density related. For these lightweight 

fill materials, the specifications should include: 

• the type of lightweight fill material to be used, including the maximum and minimum

densities that are tolerable;

• the minimum compressive and/or flexural strength; and

• the lift thickness for the cellular concrete, or the block dimensions for the EPS-block

geofoam.

For lightweight fill materials that behave as a granular material, the specifications should 

include the following. 

• The type of lightweight material that is to be used, including the minimum and

maximum density as delivered to the job site.

• The gradation of the fill material.

• Other tests, depending on the type of lightweight fill material used. For example, it

might be appropriate to include tests such as durability, as measured by the Los

Angeles abrasion test, or the percent absorption of water.

• The lift thickness, the placement, and compaction procedure. For lightweight fill

materials that resemble normal soil sizes, such as gravel, sand, and silt, the

compaction can be specified in terms of a percent of AASHTO maximum density

standards. Where recycled fill materials, such as shredded tires or wood fibers are

used, the specification should describe the method of compaction. This would include

the type of equipment and the number of passes to be used to induce compaction. If

there is any doubt as to lift thickness or percent compaction, consider test strips at the

beginning of the field operations.

• The method of payment. A suggested procedure would be to base the payment on the

amount of fill to be supplied in cubic meters on the basis of the planned final cross
3-38 

sections and the existing ground surface. Under this method, shrinkage would be the 

responsibility of the contractor. 

The lightweight fill materials that fall into this granular category include wood fiber, blast 

furnace slag, fly ash and boiler slag, ESCS, and tire shreds. 

5.2 Quality Assurance and Construction Control

For the cohesive lightweight fill materials, field monitoring should include measurements of 

the density and compressive strength of the materials supplied. For EPS-block geofoam, the 

density and compressive strength will be a function of the grade delivered with appropriate 

manufacturer quality control documentation. Some manufacturers use a coloring scheme to 

differentiate between grades of EPS geofoam blocks. Samples should be obtained for quality 

assurance testing. Observations of the placements of the blocks should also be made to 

confirm that the blocks are placed without continuous joints and that shear transfer plates are 

installed between successive lifts of the blocks. The geomembrane, or concrete, covering the 

blocks should also be measured to confirm thickness and complete enclosure of the blocks. 

The seams within the geomembrane should be sealed properly, and quality assured. See 

NCHRP 529 report (Stark et al. 2004a) and the accompanying NCHRP Web Document 65 

(Stark et al. 2004b) for detailed discussions and guidance on construction, construction 

control, and quality assurance with EPS-block geofoam. 

For foamed concrete, the ingredients are mixed directly at the job site and then pumped to the 

location for use. Samples of the freshly mixed fill should be obtained at the point of 

placement in a manner similar to concrete testing for performance of density and 

compressive strength. The lift thickness of each pour should be measured to ensure that it 

does not exceed the maximum specifications and to confirm there is sufficient hardening of 

one lift before the next lift is placed. The material must support foot traffic prior to casting 

subsequent lifts. 

For the granular lightweight fill materials, monitoring and construction control generally 

follows the same procedures used for conventional soil placement and compaction. This 

would include monitoring of the lift thickness, number of passes, and moisture content at the 

time of placement, and degree of compaction. A check should also be made on the gradation 

of the material being supplied, as well as any contamination with undesirable materials.

In the case of tire shreds or wood fibers where compaction tests are not appropriate, field 

monitoring would include measurement of the lift thickness and the number of passes with 

the compaction equipment. Visual observations or measurements of the resiliency of the 

deposit during the multiple passes will also aid in determining whether additional passes of 

the compaction equipment are necessary. Test sections at the start of construction are 
3-39 

desirable to either confirm the specified number of passes to make modifications at an early 

stage in the project. See NYSDOT GEM-20 (2015a), NYSDOT GCP-19 (2015b), and 

CalRecyle Usage Guide (Cheng 2016) for detailed discussions and guidance on construction, 

construction control, and quality assurance with tire shred fill.

For projects where the tire shreds will be used, field monitoring should also be undertaken to 

confirm that material specifications are met, such as size and that there is not excessive steel 

wire with the tire shreds or that materials other than rubber are being supplied. For projects 

where wood fibers are used, the gradation of the wood fiber should be checked on a daily 

basis to confirm that there is a blend of coarse sizes to the sawdust sizes. Only fresh wood 

fibers should be used to build the fills, so as to prolong the life of the fill. The Washington 

State Department of Transportation (Kilian and Ferry 1993) has provided a classification for 

visually identifying different degrees of decomposition in the wood fiber. Project personnel 

should be provided with this classification system so as to exclude the partially decomposed 

wood fibers from the new fill. 
3-40 

6.0 COST DATA 

6.1 Cost Components

Costs for lightweight fill materials are highly variable and will depend upon a number of 

factors:

• The basic cost of the material. If the material is a waste product that can be used 

without additional recycling, such as wood fiber, the cost will be relatively low. 

Recycled materials are also relatively cheap, but crushing or shredding and sieving 

will increase the cost slightly.

• Transportation costs. If the project is located relatively far from the supply source, the 

transportation costs could be significant. Mode of transportation will also affect the 

price. For example, a lightweight slag may not be produced locally might be cost 

effective it can be barged, in lieu of long-haul trucking, close to a project site 

• Quantity of material. The amount of material to be used on the project, and the 

staging of use, can affect the average transportation cost and may affect the basic 

material cost, particularly for smaller quantities.

• Availability of materials. Some materials are not readily available in some regions, 

e.g., blast furnace slag. If the materials are produced in very low quantities, there may 

not be sufficient materials available, unless multiple sources are used for the product. 

• Placement and/or compaction costs could be higher than for soil fill. This would 

include moisture control for fly ash and boiler slag, geomembrane or concrete covers 

for geofoam, or greater numbers of compaction passes for shredded tires. 

When calculating the costs for most lightweight fills, a common denominator should be used. 

Generally, prices for granular materials are quoted by suppliers as a cost per ton of material. 

However, the density of lightweight materials varies considerably and the loose versus 

compacted density of some lightweight materials varies. For example, the density of TDA as 

stockpiled and shipped ranges from 25 to 35 pcf, while the in-place compacted density 

ranges from 45 to 50 pcf (Cheng 2016). A ton of lightweight fill material will provide a much 

greater volume than conventional soils with a higher density. For this reason, the cost 

comparison should be made on the price per in-place cubic yard. The conversion from dollars 

per ton to dollars per in-place cubic yard can be made with an estimated in-place density and 

a delivered cost per ton price. Note that the dollars per ton is frequently given as the price 

FOB at the plant or processing facility and the transportation costs must be added to this 

number. This variation of cost ranges; at source, delivered, and in-place; is illustrated in 

Table 3-11 for EPS-block geofoam. 
3-41 

Table 3-11. Typical Cost Ranges for EPS-block Geofoam at Source, Delivered, and InPlace

Material

Material Cost/yd3

FOB at source 

Delivered Material 

Cost/yd3 FOB at Project

In-Place 

Cost/yd3

EPS-block geofoam $40 to $80 $40 to $80 $40 to $100 

The benefit of one type, or density, of lightweight fill versus other types is another cost and 

design consideration. For example, use of various types and densities of lightweight fill for 

stabilization of a landslide can result in different computed stability safety factors, and level 

of risks. Or, the volume of lightweight fill could be different for the various material 

densities to achieve similar stability safety factors. 

Typical unit prices that are reported in the literature for various types of lightweight fill 

materials are summarized in Table 3-12. 

Table 3-12. Typical Cost Ranges for Lightweight Fills 

Lightweight Fill

Material 

Cost/yd3

FOB at source 

Delivered Material 

Cost/yd3 FOB 

at Project

In-Place

Cost/yd3

EPS-block geofoam $40 to $60 $60 to $125 $40 to $100 

Cellular concrete n/d $70 to $150 $250 to $340 

Wood Fiber n/d $6 to $26 $8 to $30 

Air-cooled blast furnace slag $6 to $8 n/d n/d 

Expanded blast furnace slag $11 to $15 n/d n/d 

Boiler slag $2 to $3 n/d n/d 

Fly ash $12 to $16 n/d n/d 

Expanded shale, clay, and 

slate (ESCS) $30 to $45 n/d n/d 

Tire shreds n/d $15 to $30 n/d 

n/d = no data

In some cases, available cost information is limited, and the prices could vary significantly 

from that shown in the table. These prices were generated from recent literature, suppliers, 

and user agencies, and will vary due to the factors listed above. The prices vary considerably, 

due to factors previously discussed. Note that in-place tire shred fills as low at $2/yd3 have 

been reported, and that tire shred costs may be significantly offset with state recycling 

incentives.

Engineers should contact local suppliers to obtain specific prices, or price ranges, applicable 

to their region served by their agency. Again, transportation costs should be added to 

estimate a delivered unit cost of the lightweight fill on a specific project. An example cost 
3-42 

comparison of various fills for an embankment fill project in Milpitas, California (Cheng 

2016) is presented in Table 3-13. 

Table 3-13. Example Lightweight Fill Project Cost Comparison 

Material Total Cost* 

Traditional Soil Fill $563K 

Pumice Rock $633K 

EPS-block Geofoam $1,145K 

Expanded Shale Clay $490K 

Wood Chips $545K 

Tire-derived Aggregate $334K 

* Total costs based on material, transportation 

and longevity costs; does not include installation 

costs or contractor’s overhead and profit.

Source: Cheng 2016

Additional cost considerations would include compaction methods, loose versus compacted 

volume adjustment, and any – a specialty items, such as the need for geotextiles, 

geomembranes, drainage blankets, or soil cover. 
3-43 

7.0 CASE HISTORIES 

Four case histories are presented to illustrate the various types of applications for which 

lightweight fill materials can be used in highway construction. Each project description 

provides information on the setting, the design considerations, the construction procedures, 

and the performance data. 

These case histories provide a brief synopsis, and the reader is referred to the technical 

literature for detailed discussions of these projects. References are listed for each case 

history. 

7.1 GeoFoam

The Massachusetts Department of Transportation (MassDOT) used EPS-block geofoam on 

their Whittier Bridge/ I-95 Improvement project. The existing through-truss Whittier Bridge 

constructed in 1951 is being replaced with a network tied arch structure. Each of the two 

parallel structures will carry four travel lanes, a high speed shoulder, and a breakdown lane. 

A shared-use path will be constructed on the northbound side to facilitate connections to 

walking and cycling trails. This new bridge is a signature project of the MassDOT 

Accelerated Bridge Program.

MassDOT employed several innovative construction methods to expedite project schedule, 

ensure high quality results and long lasting infrastructure, and reduce maintenance costs. One 

such technique for the Whittier Bridge/I-95 project is the use of EPS-block geofoam 

lightweight fill material. The blocks are being used for the base of the northbound breakdown 

lane, the right northbound travel lane, and the shared-use path. These areas are underlain with 

soft organic and clay soils. If the EPS blocks were not used, the contractor would have to 

excavate down 30 feet to replace the unsuitable soils, and install a drainage system as well, 

before widening the highway. The use of EPS results in significant construction advantages 

in both time and cost savings. 

Typical EPS block fill cross section on the embankment widening and on the ramp are 

illustrated in Figures 3-10 and 3-11, respectively. 
3-44 

Massachusetts Department of Transportation

Figure 3-10. Whittier Bridge typical embankment widening section.
3-45 

Massachusetts Department of Transportation

Figure 3-11. Whittier Bridge typical ramp cross section.
3-46 

EPS block placement on the embankment widening, and going around utilities, is shown in 

Figure 3-12. 

Massachusetts Department of Transportation

Figure 3-12. Whittier Bridge EPS block placement on embankment widening.

7.2 Foamed Concrete

The New York State Department of Transportation (NYSDOT) used foamed concrete fill on 

12 separate projects. The typical application was for balanced excavations (Harbuck 1993). 

In these procedures, the existing soils were excavated to a depth required to balance the 

weight of the lightweight foamed concrete fill, which would be placed to a grade higher than 

existing. Thus, no additional load was applied to the foundation soils. Generally, 3 feet of 

foamed concrete fill at a density of 30 pcf will have the same weight as 9-inch of existing fill 

with a density of 120 pcf. 

A typical example of weight balancing was for a two-span structure carrying Route 150 over 

Amtrak in Rensselaer County that was replaced with a single-span structure. The abutments 

were structurally sound and were used to support the new superstructure. It was necessary to 

raise the grade of the approach embankments by 3 feet. This is not a large grade increase, but 

the embankments were underlain by 30 to 33 feet of very soft-to-soft clay that in turn was 
3-47 

underlain by loose silt, eventually grading to a very compact silt. An analysis of the bearing 

capacity and slope stability of these soils indicated that the raising of the grade could 

jeopardize the factor of safety.

To reduce the loading on the soft soils, the upper 6 feet below final grade was constructed of 

foamed concrete with a maximum unit weight of 30 pcf. A drain was placed behind the 

existing abutment so as to reduce water pressures on the abutment. The final design is shown 

in Figure 3-13. 

Harbuck 1993, TRB

Figure 3-13. Foamed concrete fill behind existing abutment, New York project.

7.3 Tire Shreds 

The CalRecycle Usage Guide, Tire-Derived Aggregate (Cheng 2016) provides several case 

histories of TDA use in California. Applications include embankment construction over soft 

soils, roadway landslide stabilization, retaining wall backfill, light rail vibration damping, 

landfill drainage, and leach fields. The TDA application in the following project summary 

(http://www.calrecycle.ca.gov; Cheng 2016; Kennec, Inc. 2007) is the use as lightweight fill 

to stabilize an existing embankment.
3-48 

A 160-foot long section of Marina Drive in Calpella, California had been gradually settling 

and moving since the 1960s. The settlements were repaired over the years by filling with 

base course material and asphalt. The repeated repairs led to a total thickness of base course 

and asphalt of about 7 feet. This accumulating weight increased the destabilizing loads of the 

slope. 

A design to stabilize the slope with TDA lightweight fill replacement and soil reinforcement 

was engineered (Kennec, Inc. 2007). A slope cross section illustrating this repair plan is 

shown in Figure 3-14. The longitudinal, along the roadway section is shown in Figure 3-15. 
3-49 

Kennec, Inc. 2007

Figure 3-14. Marina Drive slope cross section. 
3-50 

Kennec, Inc. 2007

Figure 3-15. Marina Drive longitudinal section of fill replacement.
3-51 

The fill replacement extended to the underlying Colluvium I. Fill consisted of compacted, 

reinforced backfill; low permeability cover soils; and two zones of TDA fill. The lower zone 

of TDA was 10-foot thick, which is the maximum recommended fill height (see Table 3-5). 

The upper and lower TDA fill zones were separated by a 3-foot thick layer of low 

permeability soil. The TDA fills were wrapped in a geotextile, that functions as a permeable 

separator between the finer graded soils and more open graded TDA fill. 

A construction photograph is presented in Figure 3-16, and a view of the completed repair is 

presented in Figure 3-17. 

http://www.calrecycle.ca.gov, ©California Department of Resources Recycling and Recovery (CalRecycle). 

All rights reserved.

Figure 3-16. Placement and compaction of TDA within geotextile wrap. 
3-52 

http://www.calrecycle.ca.gov, ©California Department of Resources Recycling and Recovery (CalRecycle). 

All rights reserved.

Figure 3-17. Completed Marina Drive repair.

Design included a surface water management plan and controls. Approximately 133,000 

waste tires were used in the two lightweight fill embankment zones. This constructed repair 

option provided an overall cost savings of $740,000 over the next option. 

7.4 Wood Fiber Embankment

Approximately 600 feet of a two-lane roadway was constructed over swampy terrain in 

Washington State in 1987-88 (Allen and Kilian 1993). A generalized soil profile at the site, 

including the existing and final roadway grades is shown in Figure 3-18. 
3-53 

Allen and Kilian 1993, TRB

Figure 3-18. Soil cross-section at wood fiber fill location, Washington project. 

The standard penetration resistance values in the organic sandy silt were generally less than 

1. In place water contents ranged from 94 to 364 percent, with an average of 172 percent. 

The liquid limit ranged from 61 to 90 percent, and the plastic limit from 53 to 65 percent. 

The average unconsolidated, undrained strength determined from triaxial shear testing was 

150 psf. 

The new embankment vertical grades required up to 44 feet of fill. Side slopes were limited 

to 2H:1V to limit the right of way and the fill volume required and to minimize the amount of 

wetlands removed by construction of the embankment. Slope stability calculations indicated 

that construction of the embankment using granular soil was not practical. By assuming no 

strength gain in the foundation soil, the maximum height of granular fill that could be safely 

constructed was determined to be 12 feet. If the embankment were constructed slowly, 

allowing pore pressure dissipation and strength gain, it was estimated that 20 months would 

be required for construction of the embankment. In addition, settlement under 44 feet of 

granular fill was expected to be 8 feet, which was unacceptable because of the necessity of 

installing a culvert at the base of the fill.
3-54 

In order to obtain a minimum factor of safety of 1.25 for slope stability and 1.5 for bearing 

capacity, it was necessary to use a lightweight fill material for the embankment. Wood fiber 

was available in this area and was selected for the embankment fill. The initial 5 feet of fill 

consisted of a silty, gravelly sand to raise the base of the embankment above the prevailing 

groundwater table. The upper 4 feet of the fill was also a granular deposit to provide 

adequate subgrade support for the roadway. The remaining 35 feet of the embankment height 

consisted of the wood fiber-fill. For the stability calculations, the density of the wood fiberfill was used as 38 pcf and an angle of 40°. The granular soils were assumed to have a unit 

weight of 125 pcf and an angle of internal friction of 37°. Five layers of geotextile 

reinforcement were embedded within the lower granular fill, as well as within the wood fiber, 

so as to meet the minimum factor of safety requirements for slope stability and lateral 

spreading considerations. The geotextile reinforcement was needed only until the soil gained 

sufficient strength to support the fill without reinforcement. It was estimated that this would 

require about 8 months, and the geotextiles were designed allowing a relatively high creep 

limit of 60 percent of ultimate tensile strength.

Embankment settlement using the wood fiber fill was estimated to be 5 feet, plus an 

additional 0.4 feet of secondary consolidation over a 20-year period. An accurate settlement 

estimate was necessary to determine the amount of additional wood fiber fill necessary to 

accommodate the anticipated settlement. Re-leveling with granular soils would reduce the 

slope stability factor of safety.

The as compacted density of the wood fiber fill was determined to be 38 pcf. Compaction 

was obtained by routing hauling equipment over the entire lift thickness of 1-foot. The 

minimum mass of the hauling equipment was 33,000 lbs. 

Piezometers were installed in the organic silts and monitored as the embankment was placed 

in height. The maximum allowable pore pressure ratio was equaled or exceeded twice during 

construction. This occurred when the fill height reached 22 feet and 31 feet. In the first case, 

fill construction was stopped for 52 days; in the second case, for 130 days. 

The total time required to construct the embankment was just under 11 months. The subgrade 

was reached in September 1987, and paving began in October 1988. The measured 

settlement of the embankment immediately following construction to subgrade level was 3.3 

feet. Just prior to paving, this settlement had increased to 4 feet. In September 1992, 

settlement increased to 4.6 feet. This compares favorably with the estimated primary 

settlement of 5 feet.
3-55 

The cost of the embankment construction, including the geotextile reinforcement, was 

$972,000. If ground improvement had been undertaken with stone columns, the estimated 

cost was $1,500,000. The cost for constructing a bridge over the weak ground area was 

estimated to be $1,700,000. 
3-56 

8.0 REFERENCES

ACCA (2014). 2014 Coal Combustion Product (CCP) Production & Use Survey Report. 

American Coal Ash Association, Farmington Hills, MI, 1p. https://www.acaausa.org/Portals/9/Files/PDFs/2014ReportFinal.pdf. 

Allen, T.M. and Kilian, A.P. (1993). Use of Wood Fiber and Geotextile Reinforcement to 

Build Embankment Across Soft Ground. Transportation Research Record No. 1422, 

Transportation Research Board, Washington, D.C., pp. 46-54. 

ASTM (2015). Standard Specification for Rigid Cellular Polystyrene Geofoam, ASTM 

D6817-15, ASTM International, West Conshohocken, PA, 4p. 

ASTM (2013a). Standard Guide for Use of Expanded Polystyrene (EPS) Geofoam in 

Geotechnical Projects, ASTM D7180, ASTM International, West Conshohocken, PA, 

3p. 

ASTM (2013b). Standard Practice for Sampling of Expanded Polystyrene Geofoam 

Specimens, ASTM D7557, ASTM International, West Conshohocken, PA, 2p. 

ASTM (2012). Standard Practice for Use of Scrap Tires in Civil Engineering Applications, 

ASTM D6270, ASTM International, West Conshohocken, PA, 22p. 

Arellano, D., Stark, T.D., Horvath, J.S., and Leshchinsky, D. (2011). Guidelines for Geofoam 

Applications in Slope Stability Projects. NCHRP Project No. 24-11(02), National 

Cooperative Highway Research Program, Transportation Research Board, 

Washington, D.C., 602p. 

Baker, T.E., Allen, T.M., Jenkins, D.V., Mooney, D.T., Pierce, L.M., Christie, R.A., and 

Weston, J.T. (2003). Evaluation of the Use of Scrap Tires in Transportation Related 

Applications in the State of Washington, Washington State Department of 

Transportation, Olympia, WA, 268p. 

Caltrans (2014). Ground Improvement Section, Geotechnical Manual, California Department 

of Transportation, Sacramento, CA, 21p. 

http://www.dot.ca.gov/hq/esc/geotech/geo_manual/manual.html. 

Cheng, D. (2016). Usage Guide, Tire-Derived Aggregate (TDA), Publication # DRRR 2016-

01545, California Department of Resources Recycling and Recovery, Sacramento, 

CA, 56p. http://www.calrecycle.ca.gov/Publications/Default.aspx. 
3-57 

CROW (2003). Roads and Car Parks on Foam Concrete. CROW publication 101. Ede, The 

Netherlands. 

ECSI (2007). Reference Manual for the Properties and Applications of Expanded Shale, Clay 

and Slate Lightweight Aggregate. Expanded Shale, Clay and Slate Institute, Salt Lake 

City, UT. 

Edstrom, R.D., Jordahl-Larson, M., and Sampson, J. (2008). Oak Grove Tire Shreds Project: 

Tire Shreds below the Seasonal Groundwater Table Years 2006-2008. MN/RC 2009-

02, Minnesota Department of Transportation, St. Paul, MN, 244p. 

Expanded Polystyrene Development Organization (1994). Slope Stabilization by the EPS 

Method and Its Applications. International Geotechnical Symposium on Polystyrene 

Foam in Below Grade Applications, Honolulu, HI. 

FHWA. (2010). Recycled Materials in Roadway Construction: The Many Ways of Going 

Green, FOCUS, FHWA-HRT-10-012, Federal Highway Administration, U.S. DOT, 

Washington D.C. 

FHWA. (1997). Interim Guidelines for Shredded Tire Embankments. Director Office of 

Engineering, HNG-23, Federal Highway Administration, U.S. DOT, Washington 

D.C.

Gale,S., Lichty, N., and Forsberg, A. (2013). Bridge Approach Embankment Slope Distress: 

Analysis, Monitoring, Design & Remediation. Proc. Geo-Congress 2013, ASCE, 

Reston, VA, pp. 1384-1399. 

Harbuck, D.I. (1993). Lightweight Foamed Concrete Fill. Transportation Research Record 

No. 1422, Transportation Research Board, Washington, D.C., pp. 21-35. 

Holm, T. and Valsangkar, A. (1993). Lightweight Aggregate Soil Mechanics: Properties and 

Applications. Transportation Research Record No. 1422, Transportation Research 

Board, Washington, D.C., pp. 7-13. 

Horvath, J.S. (1995). Geofoam Geosynthetic. Horvath Engineering, Scarsdale, NY. 

Horvath, J.S. (2012). The Evolution of Generic Material Standards for Block‐Molded 

Expanded Polystyrene (EPS) Used for Small‐Strain Geofoam Applications in the 

U.S.A. Research Report No. CEEN/GE‐2012‐1, Manhattan College, School of 

Engineering, Bronx, NY, 104p. 
3-58 

Holtz, R. (1989). Treatment of Problem Foundations for Highway Embankments. NCHRP 

Synthesis of Highway Practice 147, Transportation Research Board, Washington, 

D.C.

Humphrey, D.H. (1996). Investigation of Exothermic Reaction in TDA Fill Located on SR 

100 in Ilwaco, Washington. Federal Highway Administration, U.S. DOT, 

Washington, D.C. 

Humphrey, D.H. (1998). Highway Application of Tire Shreds. New England Transportation 

Consortium, West Greenwich, RI.

Kennec, Inc. (2007). Marina Drive Slide Repair, CR 266 at M.P. 0.63. Mendocino County, 

California Construction Plans; Kennec, Inc., Long Beach, CA, 6p. 

Kilian, A.P. and Ferry, C.D. (1993). Long Term Performance of Wood Fiber Fills. 

Transportation Research Record No. 1422, Transportation Research Board, 

Washington, D.C., pp.55-60. 

Lewis, D. (1982). Properties and Uses of Iron and Steel Slag. Symposium on Slag, National 

Institute for Transport and Road Research, South Africa. Available, with statistical 

updates to 1992 as MF 182-6 at http://www.nationalslag.org/. 

National Slag Association (1968). Building Safer Highways with Slag. Technical Bulletin 

No. 264, American Road Builders Association. 

National Slag Association (1988). Processed Blast Furnace Slag, The All-Purpose 

Construction Aggregate. NSA Bulletin 188-1. 

NYSDOT (2015a). Guidelines for Project Selection, Design, and Construction of Tire Shreds 

in Embankments. Geotechnical Engineering Manual, GEM-20 Revision #3, New 

York State Department of Transportation, Albany, NY. 

NYSDOT (2015b). Sampling and Testing of Tire Shreds. Geotechnical Control Procedures, 

GCP-19 Revision #7, New York State Department of Transportation, Albany, NY. 

Oman, M.S., Gebhard, J., and Hoppe, M.J.L. (2013). Use of Tire Derived Products (TDP) in 

Roadway Construction. Report MN/RC 2013-20, Minnesota Department of 

Transportation, St. Paul, MN, 58p. 

PIARC (1997). Lightweight Filling Materials. Technical Committee 12, Permanent 

International Association of Road Congresses, AIPRC, La Defense Cedex 04, France. 
3-59 

Rittenhouse, R.C. (1993). EPA, Miracle Concrete, Non-Hazardous Wastes. Power 

Engineering Magazine, 14p. 

Solomon, C. (1993). Slag-Iron and Steel. U.S. Department of Interior, Bureau of Mines 

Annual Report. 

Sonti, K., Senadheera, S., Jayawickrama, P.W., Nash, P.T., and Gransberg, D.D. (2003). 

Evaluate the Uses for Scrap Tires in Transportation Facilities. Texas Department of 

Transportation, Austin, TX, 85p. 

Stark, T.D., Arellano, D., Horvath, J.S., and Leshchinsky, D. (2004a). Guideline and 

Recommended Standard for Geofoam Applications in Highway Embankments. 

NCHRP Report 529, National Cooperative Highway Research Program, 

Transportation Research Board, Washington, D.C., 58p. 

Stark, T.D., Arellano, D., Horvath, J.S., and Leshchinsky, D. (2004b). Geofoam Applications 

in the Design and Construction of Highway Embankments. NCHRP Web Document 

65, http://www.trb.org/main/blurbs/153437.aspx. 

Stoll, R. and Holm, R. (1985). Expanded Shale Lightweight Fill: Geotechnical Properties. 

Journal of Geotechnical Engineering, ASCE, 111(8): pp. 1023-1027. 

Stroup-Gardiner, M. and Wattenberg-Komas, T. (2013a). Recycled Materials and Byproducts 

in Highway Applications, Volume 1: Summary Report, NCHRP Synthesis 435, 

Transportation Research Board, Washington, D.C., 93p. 

Stroup-Gardiner, M. and Wattenberg-Komas, T. (2013b). Recycled Materials and 

Byproducts in Highway Applications, Volume 2: Coal Combustion Byproducts, 

NCHRP Synthesis 435, Transportation Research Board, Washington, D.C., 51p. 

Stroup-Gardiner, M. and Wattenberg-Komas, T. (2013c). Recycled Materials and Byproducts 

in Highway Applications, Volume 5: Slag Byproducts, NCHRP Synthesis 435, 

Transportation Research Board, Washington, D.C., 35p. 

Stroup-Gardiner, M. and Wattenberg-Komas, T. (2013d). Recycled Materials and 

Byproducts in Highway Applications, Volume 7: Scrap Tire Byproducts, NCHRP 

Synthesis 435, Transportation Research Board, Washington, D.C., 47p. 

TRB (2013). Guidelines for Geofoam Applications in Slope Stability Projects, Research 

Results Digest 380, NCHRP, Transportation Research Board, Washington, D.C., 26p. 
3-60 

Valsangkar, A.J. and Holm, T.A. (1990). Geotechnical Properties of Expanded Shale 

Lightweight Aggregate. Geotechnical Testing Journal, ASTM 13(X): pp.10-15. 

WSDOT (2015). Geotechnical Design Manual, M 46-03.11. Washington Department of 

Transportation, Tumwater, WA, 868p. 

Zornberg, J.G., Christopher, B.R., and Oosterbaan, M.D. (2005). Tire Bales in Highway 

Applications: Feasibility and Properties Evaluation. CDOT-DTD-R-2005-2, Colorado 

Department of Transportation Research Branch, 102p. 
Chapter 4 

DEEP COMPACTION

4-i 

CONTENTS

1.0 DESCRIPTION AND HISTORY ........................................................................... 4-1

1.1 Description ............................................................................................................ 4-1

1.2 Historical Overview ............................................................................................. 4-4

1.2.1 Deep Dynamic Compaction ............................................................................... 4-4

1.2.2 Vibro-compaction .............................................................................................. 4-5

1.3 Focus and Scope ................................................................................................... 4-6

1.4 Alternative Technologies ..................................................................................... 4-6

1.4.1 Frankipaction ..................................................................................................... 4-6

1.4.2 Blast Densification ............................................................................................. 4-6

1.4.3 Rapid Impact Compaction ................................................................................. 4-7

1.4.4 Compaction Grouting......................................................................................... 4-7

1.4.5 Vibratory Hammer Probe ................................................................................... 4-7

1.5 Alternative Solutions ........................................................................................... 4-7

1.6 Liquefaction Potential and Mitigation Assessment .......................................... 4-8

1.7 Glossary .............................................................................................................. 4-11

1.7.1 Deep Dynamic Compaction ............................................................................. 4-11

1.7.2 Vibro-compaction ............................................................................................ 4-12

1.8 Primary References ........................................................................................... 4-12

1.8.1 Deep Dynamic Compaction ............................................................................. 4-12

1.8.2 Vibro-Compaction ........................................................................................... 4-12

2.0 DEEP DYNAMIC COMPACTION ..................................................................... 4-14

2.1 Feasibility Considerations ................................................................................. 4-14

2.1.1 Applications ..................................................................................................... 4-14

2.1.2 Advantages and Potential Disadvantages ........................................................ 4-17

2.1.3 Feasibility Evaluations ..................................................................................... 4-19

2.1.4 Limitations ....................................................................................................... 4-24

2.2 Construction and Materials .............................................................................. 4-28

2.2.1 Overview .......................................................................................................... 4-28

2.2.2 Tampers............................................................................................................ 4-29

2.2.3 Lifting Equipment ............................................................................................ 4-32
4-ii 

2.2.4 Surface Stabilizing Layer ................................................................................. 4-36

2.2.5 Construction Sequencing ................................................................................. 4-37

2.3 Design .................................................................................................................. 4-38

2.3.1 Design Considerations ..................................................................................... 4-38

2.3.2 Performance Requirements .............................................................................. 4-39

2.3.3 Design Procedure ............................................................................................. 4-40

2.3.4 Example ........................................................................................................... 4-46

2.4 Construction Specifications and Quality Assurance ...................................... 4-47

2.4.1 Contracting Procedures .................................................................................... 4-47

2.4.2 Instrumentation Monitoring and Construction Control ................................... 4-50

2.5 Cost Information ................................................................................................ 4-52

2.5.1 Cost Components ............................................................................................. 4-52

2.5.2 Cost Data .......................................................................................................... 4-52

2.6 Case Histories ..................................................................................................... 4-53

2.6.1 Highway Embankment Constructed in Mine Spoil ......................................... 4-53

2.6.2 Highway Embankment on Landfill Debris ...................................................... 4-59

3.0 VIBRO-COMPACTION ....................................................................................... 4-64

3.1 Feasibility Considerations ................................................................................. 4-64

3.1.1 Applications ..................................................................................................... 4-64

3.1.2 Advantages and Disadvantages ........................................................................ 4-64

3.1.3 Feasibility Evaluations ..................................................................................... 4-65

3.2 Construction and Materials .............................................................................. 4-69

3.2.1 Construction ..................................................................................................... 4-69

3.2.2 Materials .......................................................................................................... 4-72

3.3 Design .................................................................................................................. 4-73

3.3.1 Design Considerations ..................................................................................... 4-73

3.3.2 Design Procedure ............................................................................................. 4-74

3.4 Construction Specifications and Quality Assurance ...................................... 4-79

3.4.1 Contracting Procedures .................................................................................... 4-79

3.4.2 Quality Assurance and Monitoring .................................................................. 4-80
4-iii

3.5 Cost Information ................................................................................................ 4-83

3.5.1 Cost Components ............................................................................................. 4-83

3.5.2 Cost Data .......................................................................................................... 4-86

3.6 Case Histories ..................................................................................................... 4-86

3.6.1 I-90 Mt. Baker Ridge, Seattle, WA (Hayward Baker 1989) ............................ 4-86

3.6.2 Wando Terminal, Charleston, SC (Hussin and Foshee 1994) ......................... 4-88

3.6.3 Manchester Airport, New Hampshire (Sobel et al. 1993) ............................... 4-94

4.0 REFERENCES ....................................................................................................... 4-96
4-iv

LIST OF FIGURES

Figure 4-1. Schematic illustration of deep dynamic compaction. ......................................... 4-2

Figure 4-2. The vibro-compaction process. ........................................................................... 4-2

Figure 4-3. SPT liquefaction chart for magnitude 7.5 earthquakes. ...................................... 4-9

Figure 4-4. Phase diagram model of soil behavior during deep dynamic compaction. ....... 4-15

Figure 4-5. Grouping of soils for deep dynamic compaction. ............................................. 4-21

Figure 4-6. U.S. Bureau of Mines safe level of blasting vibrations for houses and 

Office of Surface Mining regulations. ........................................................................... 4-24

Figure 4-7. Scaled energy factor versus particle velocity. ................................................... 4-25

Figure 4-8. Lateral movements at 3 m (10 feet) from drop point. ....................................... 4-27

Figure 4-9. Lateral movements at 6 m (20 feet) from drop point. ....................................... 4-27

Figure 4-10. 12 kip tamper................................................................................................... 4-30

Figure 4-11. 30 kip tamper................................................................................................... 4-30

Figure 4-12. 31 kip tamper................................................................................................... 4-31

Figure 4-13. 70 kip tamper................................................................................................... 4-31

Figure 4-14. Tamper with low contact pressure for ironing pass. ....................................... 4-32

Figure 4-15. Lifting crane of 70 kips rated capacity used with 12 kip tamper. ................... 4-33

Figure 4-16. Dragline-type lifting crane with 11⁄4-inch cable for use with 40 kip tamper. .. 4-33

Figure 4-17. Manitowoc 4000 crane with special lifting drum attached to rear of crane 

for use with 64 kip tamper. ............................................................................................ 4-34

Figure 4-18. Lifting crane built only for deep dynamic compaction with 5.0-foot 

diameter lifting hoist and single line-rated capacity of 50 ton at line speeds of 85 

feet/min used with 32 ton tampers. ................................................................................ 4-35

Figure 4-19. Multiple part line used to lift 64 kip tamper (below grade in photo). ............. 4-36

Figure 4-20. Deep dynamic compaction over a 3-foot layer of crushed rock placed on a 

landfill. ........................................................................................................................... 4-37

Figure 4-21. Trend between apparent maximum depth of influence and energy per 

drop. ............................................................................................................................... 4-41

Figure 4-22. Relationship between tamper mass and drop height. ...................................... 4-42

Figure 4-23. Variations in improvements with depth during deep dynamic compaction. ... 4-43
4-v 

Figure 4-24. Increase in SPT values in a mine spoil after deep dynamic compaction. ....... 4-54

Figure 4-25. Induced settlement following deep dynamic compaction at the mine spoil 

project in Alabama. ........................................................................................................ 4-57

Figure 4-26. Statistical variation in crater depths at the mine spoil site in Alabama. ......... 4-58

Figure 4-27. Range of soil types treated by vibro-compaction. ........................................... 4-66

Figure 4-28. Soil compactibility based on cone penetration resistance and friction ratio. .. 4-67

Figure 4-29. High-powered, probe-type vibrator utilized in vibro-compaction. ................. 4-69

Figure 4-30. Cross-section of a typical vibrator. ................................................................. 4-70

Figure 4-31. Vibrator suspended from a conventional crane. .............................................. 4-71

Figure 4-32. Vibrator suspended from a barge-mounted crane. .......................................... 4-72

Figure 4-33. Approximate variation of relative density with tributary area. ....................... 4-76

Figure 4-34. Relative density versus probe spacing for silty sands. .................................... 4-77

Figure 4-35. Typical compaction point spacing for area layouts. ........................................ 4-78

Figure 4-36. Typical compaction point layouts for column footings. ................................. 4-78

Figure 4-37. Typical data logger results: amps versus depth (left) and time versus depth 

(right). ............................................................................................................................ 4-82

Figure 4-38. Vibro-compaction on Mt. Baker Ridge’s Interstate 90. .................................. 4-87

Figure 4-39. State Pier 41, Wando Terminal. ...................................................................... 4-89

Figure 4-40. Vibro-compaction at Wando Terminal. .......................................................... 4-90

Figure 4-41. Generalized subsurface profile of Area C. ...................................................... 4-91

Figure 4-42. Densification of loose sand backfill during vibro-compaction at Wando 

Terminal. ........................................................................................................................ 4-92

Figure 4-43. Sample cone penetration results. ..................................................................... 4-93
4-vi 

LIST OF TABLES

Table 4-1. Equipment Requirements for Different Sized Tampers ..................................... 4-34

Table 4-2. Upper Bound Test Values after Deep Dynamic Compaction ............................ 4-44

Table 4-3. Applied Energy Guidelines ................................................................................ 4-46

Table 4-4. Deep Dynamic Compaction Costs...................................................................... 4-52

Table 4-5. Suitability Assessment of Granular Soils for Vibro-compaction ....................... 4-66

Table 4-6. Specifications of Several Vibrators .................................................................... 4-70

Table 4-7. Backfill Evaluation Criteria ................................................................................ 4-73

Table 4-8. Penetration Resistance and Sand Properties ....................................................... 4-74

Table 4-9. Factors Affecting Price of Vibro-compaction Projects ...................................... 4-85

Table 4-10. Cost Information Summary .............................................................................. 4-86
4-1 

1.0 DESCRIPTION AND HISTORY

1.1 Description

Deep compaction is a category description of technologies that rely on dynamic methods to 

impart high levels of energy to the ground resulting in improvement of soil properties. Kirsch 

and Kirsch (2010) divide such methods into compaction by vibration using depth vibrators or 

vibratory hammers and compaction by impact using drop weights or explosion. These 

methods are most applicable to loose cohesionless soils having little to no fines content. The 

dynamic forces imparted to the soils densify the cohesionless soils resulting in increased 

bearing capacity, increased shear strength, reduced settlement, and increased liquefaction 

resistance. In this chapter two technologies (methods) are discussed in detail: Deep Dynamic 

Compaction (DDC) and Vibro-compaction (VC). Deep dynamic compaction densifies 

materials by drop weights imparting high levels of impact energy at the surface. Vibrocompaction uses specialized vibrators lowered into the loose soils which send out horizontal 

vibrations to densify nearby materials. 

Closely related technologies include blast densification, an alternate method that densifies 

cohesionless soils through high levels of impact dynamic energy; rapid impact compaction 

(RIC) that uses an excavator-mounted drop weight to compact soils to depths of 10 to 20 

feet; and a group of technologies similar to vibro-compaction termed vibro-replacement, 

whereby additional material is added to the ground as the dynamic vibrations are imparted. 

Vibro-replacement methods include vibro-stone columns, vibro-concrete columns and sand 

compaction piles. These are further detailed in Chapter 5 Aggregate Columns. 

Deep dynamic compaction is a method of ground modification that results from the 

application of high levels of energy at the ground surface. The energy is applied by 

repeatedly raising and dropping a tamper with a mass ranging from (10 to 40 kips) at heights 

ranging from 30 to 100 feet. The tamper is lifted and dropped by a conventional crane with a 

single cable plus a winch that has a free spool attachment that allows the single cable to 

unwind with minimum friction. The tamper’s energy of impact at the ground surface results 

in densification of the deposit to depths that are proportional to the energy applied. The depth 

of improvement generally ranges from 10 to 35 feet for light- to heavy-energy applications, 

respectively. Following the high energy level application, the surface of the deposit is in a 

loose condition to a depth equal to the depth of the craters. This surface is then compacted on 

a tight grid basis, with a low-level energy application called an ironing pass. 

The deep dynamic process is illustrated schematically in Figure 4-1. The arrows represent

energy that is transmitted into the soil mass following impact of the tamper. The predicted 

depth of improvement is shown as a function of the energy of a single drop. 
4-2 

Figure 4-1. Schematic illustration of deep dynamic compaction.

If ground improvement is needed to provide a suitable bearing stratum for an embankment or 

a structure, deep dynamic compaction may be a viable solution. Deep dynamic compaction is 

generally the most economical form of site improvement. Conventional cranes are used to lift 

and drop the tampers, which allow local contractors to compete with specialty contractors to 

perform this work. Deep dynamic compaction can be used on a wide variety of deposits, 

ranging from silty soils to boulder-sized granular deposits. This procedure can improve both 

fill deposits and natural soil deposits. 

Vibro-compaction is a ground modification technique that uses specially designed probetype, depth vibrators for in situ densification of loose sands and gravels, as shown in Figure 

4-2. 

Figure 4-2. The vibro-compaction process. 
4-3 

Originally called vibroflotation, vibro-compaction was accomplished with water jetting, 

hence the name (Vibroflotation Foundation Co. 1980). Subsequent equipment development 

featuring higher horsepower and amperage made a dry operation possible. The majority of 

vibro-compaction projects, however, are accomplished by the jetting water (wet) method, and 

this chapter reflects this tendency.

The mechanism of densifying granular, cohesionless soils with vibrators can be briefly 

described as follows: Mechanical vibrations and simultaneous application of water nullify the 

effective stresses between the soil grains, which are rearranged, unconstrained, and 

unstressed under the action of gravity to the densest possible state, thus providing permanent 

compaction. In the immediate vicinity of the vibrator, the soil is saturated, liquefies locally 

and temporarily under the influence of the vibrations (Moseley and Priebe 1993). 

There are numerous natural and man-made deposits where vibro-compaction, can be applied, 

including densification of granular hydraulic fills, coastal plain sediments, glacial deposits, 

alluvial soils, and miscellaneous granular fills and/or deposits to permit construction of 

shallow foundations. Also, liquefaction potential can be reduced by vibro-compacting loose, 

granular soil to a density beyond the threshold density triggering liquefaction. In earth 

retaining problems, the process can be performed prior to wall construction to decrease active 

earth pressure and increase passive resistance as the density is improved. Generally, vibrocompaction can be used to achieve the following results: 

• Increased soil bearing resistance, permitting shallow foundation construction

• Reduced foundation settlements

• Increased resistance to liquefaction

• Increased shear strength

• Reduced permeability

• Filling of voids in treated areas

As with any ground modification technique, vibro-compaction has its limitations. The 

improved in situ soil characteristics depend on the in situ soil type and its gradation, the 

spacing of the compaction points, the characteristics of the equipment utilized, and the 

compaction duration. All of these factors affect the outcome of the project.
4-4 

1.2 Historical Overview

1.2.1 Deep Dynamic Compaction

Impacting soil deposits with tampers dropped from varying heights has been used for 

centuries. The Romans reportedly used this process to densify loose soils (Kerisel 1985). 

Cohesionless soils in Germany were compacted with a 4 kip tamper and a 5-foot drop from a 

steam shovel in 1933 (Loos 1963). The Corps of Engineers experimented with heavy 

tamping at the Franklin Falls Dam construction site in 1936 (USACE 1938). In 1955, 

dynamic compaction was used in South Africa to densify loose soils to support a 250 feet 

diameter crude oil tank (Hobbs 1976). In Russia, heavy tampers were used to compact 

loessial silty and sandy soils, beginning about 1960 (Bobylev 1963). 

Use of deep dynamic compaction began on a regular and continuing basis in Europe in 1969 

and in the United States in 1971. In Europe, tampers of 18 to 22 kips were dropped from 

heights of 25 to 40 feet to densify fill deposits. This process was called heavy tamping, and 

was generally used in good quality fill deposits, such as rock waste, rubble, and sand. After a 

few years, the process was expanded to include fine grain soil deposits, and the name was 

changed to dynamic consolidation (Menard and Broise 1975). 

In the United States, densification was initially achieved using tampers in the range of 4 to 12 

kips with drops of 20 to 35 feet to densify loose rubble fill and granular deposits to support 

lightly loaded structures. Later, tampers up to 33 kips were used to densify former landfills. 

This technique was initially called “pounding,” but eventually became known as dynamic 

compaction (FHWA 1986). Herein, the term deep dynamic compaction is used to 

differentiate this method from shallow dynamic compaction methods such as high energy 

impact rollers.

Before 1975, European and American practices developed independently and were somewhat 

experimental. In 1975, a technical article by Menard and Broise (1975) was published that 

dealt with the theoretical and practical aspects of deep dynamic compaction. A formula was 

presented for predicting the depth of improvement as a function of the applied energy. This 

article presented sufficient information to place the deep dynamic compaction process on a 

sound technical basis. Subsequent articles published in Europe and the United States used 

this paper as the starting point to expand and exchange the knowledge base of deep dynamic 

compaction. In the United States, the state of practice was summarized by Lukas in

”Dynamic Compaction for Highway Construction” (FHWA 1986) which presented 

guidelines for implementation. Guidelines for implementation have been kept current in 

FHWA publications (FHWA 1986, GEC 1 1995) and this manual. 
4-5 

1.2.2 Vibro-compaction 

The fundamental concept of vibro-compaction ground treatment was developed in Germany 

during the early 1930s for compaction of variable and loose naturally occurring sand deposits 

to depths of 65 feet. It was used in the densification of underwater sands for the seaport 

developments of northern Germany. Previous methods of compacting sand deposits consisted 

essentially of surface vibration or rolling. To overcome the limitations of these methods, a 

technique was developed whereby a metal tube or probe, which had within its lower end an 

electric motor driving an eccentric weight, was inserted into the ground. Vibrations were 

imparted into the ground as the tube was inserted to a required depth. To assist penetration of 

the vibrator, pressurized water was jetted through the tip of the probe. This original process 

was patented under the name Vibroflotation. 

In 1948, the first U.S. vibro-compaction project was performed in Cape May, NJ. By the 

early 1970s, vibro-compaction was generally accepted as the premier method of densifying 

deep deposits of sand. Since that time, vibro-compaction techniques have been improved and 

utilized to solve various types of geotechnical problems involving loose, granular soils, 

ranging from foundation settlement to poor bearing capacity. 

There are several systems that have been historically identified as vibro-compaction. These 

include vibroflotation, Terra-Probe®, and the vibratory beam. Vibroflotation refers to 

compaction by means of a vibroprobe, sometimes called a vibroflot that densifies with 

predominantly horizontal vibrations, while other techniques such as Terra-Probe®, VibroWing®, and Tri-Star® or Y-Probe®, utilize a top pile vibrator that densifies with 

predominantly vertical vibrations, which normally require closer spacing and are less 

efficient (Welsh 1986 and Schaefer 1997). 

The Great Alaska and Niigata earthquakes of 1964 brought the issue of liquefaction of soils 

to the forefront. As a consequence, procedures to estimate the seismic response of level 

ground, embankments, and slopes have been improved. More recently, the extensive 

geotechnical failures during the Loma Prieta, Northridge, and Kobe earthquakes have served 

to highlight the importance of ground improvement for seismically unstable sites. At Kobe, 

observations of sites where a vibro-densification process had been used indicated that while 

areas outside of the improved sites showed significant evidence of liquefaction in the form of 

settlements and sand boils, the improved ground either precluded liquefaction or limited 

deformations to a minimum.
4-6 

1.3 Focus and Scope

Deep dynamic compaction and vibro-compaction are mature, well-established methods of 

ground modification. The purpose of this chapter is to provide an overview of these methods 

and to present the state of the practice in their utilizing for improvement of site soils. This 

chapter addresses feasibility considerations, including applications, advantages, limitations, 

and potential disadvantages; construction equipment and required materials; preliminary 

design concepts; specification and quality assurance development; cost information; and case 

histories.

1.4 Alternative Technologies

A number of alternative methods of compacting ground in situ exist. Many of these 

techniques also densify cohesionless soils with some adding structural support to the ground. 

These techniques are as follows: 

1.4.1 Frankipaction 

The Frankipaction system is essentially an adaptation of the Franki pile to achieve 

compaction. The Franki pile is a driven, enlarged-base pile. The pile casing is positioned 

on the ground and a charge of dry concrete is placed within the bottom of the casing. A 

drop hammer is used to drive the concrete, which then forms a dense plug that penetrates 

the ground, dragging the casing down with it. When the pile has reached the desired depth, 

the concrete end-plug within the casing is driven down and outward to form the pile base. 

Successive charges of concrete are then placed and compacted as the casing is withdrawn,

forming the pile shaft. In Frankipaction, sand or gravel is used instead of concrete, and a 

compaction pile is created. The major advantage of this system is the ability to put large

volumes of stone in layers of soft or compressible soils to minimize settlements or potential 

shear failures of these layers, while still achieving densification in the granular layers.

1.4.2 Blast Densification

Blast densification has been used since the mid-1950s. While it is potentially economical 

compared with other systems, it is generally only useful where densification is necessary 

over large areas and at great depths (>100 feet). The technique has an inherent potential to 

have a negative effect on both the natural and man-made environment. Blast densification is 

more vulnerable to litigation than most other densification techniques and therefore requires 

a high degree of expertise and experience in its execution. Despite these drawbacks, the 

technique should be considered useful under the right circumstances. Washington State DOT 

has reported a blast densification project in which potentially liquefiable, loose debris 
4-7 

avalanche deposits from the 1980 eruption of Mt. St. Helens was densified by blasting 

techniques (Kimmerling 1994). 

1.4.3 Rapid Impact Compaction 

Rapid impact compaction uses a 5 to 9 ton excavator-mounted, weight that is dropped about 

4 feet on to a 5-foot diameter tamper capable of imparting 40 to 60 blows per minute. The 

resulting force of this RIC process densifies soils to depths of up to 10 to 20 feet. The depth 

of compaction is dependent on the compaction energy level, soil properties, and groundwater 

conditions. 

1.4.4 Compaction Grouting

Compaction grouting is a method in which cohesionless or weak soil, soil with fractures and 

air pockets, or soil that has settled, is densified using a thick, low-slump grout. The grout 

forms a bulb at the tip of the grout pipe, displacing the soil. Soil between the grout bulbs is 

thus compacted and strengthened. The common applications of this technique are described 

in the Chapter 8, Grouting. The use of compaction grouting as a densification tool is 

relatively new. Although vibro-compaction and deep dynamic compaction are significantly 

more economical in granular soils, compaction grouting might prove economical in finer 

grained soils or layered soils where strength gain is necessary. Compaction grouting also has 

economic advantages where only localized layers at depth need treatment. 

1.4.5 Vibratory Hammer Probe

The vibratory hammer probe method differs from the vibro-compaction method in that the 

vibrations are transmitted vertically down the attached pipe of a typical diameter of 2.6 feet. 

The vibro-compaction method transmits a horizontal vibration over a distance of up to 13 

feet. The frequency of vibration of the vibratory hammer probe and the location of the 

vibrator result in a less-effective densification process and therefore must be used on 

significantly closer pattern spacing. Usually this results in higher overall cost for 

densification. 

1.5 Alternative Solutions 

Prior to deciding upon ground improvement by densification, alternate foundation solutions 

should be investigated to compare construction time and costs. Aside from relocating the 

project to a more favorable location, common foundation alternatives include the following: 

• Foundations of piles or drilled shafts to transfer the loads to deeper levels. A

structural slab or a geosynthetic-reinforced thick, crushed-stone mat may be required
4-8 

at the ground surface to transfer embankment loading into the deep foundations (see 

Chapter 6). 

• Excavation and replacement of the weak ground. Sometimes this solution may also

entail use of sheeting and bracing in confined working areas to retain the adjacent soil

mass or dewatering when the water table is high. Contaminated soil deposits could

result in a very high excavation cost if the spoil must be taken to a special waste site.

• Preloading of the weak deposits to increase strength and decrease compressibility.

This solution frequently can result in time delays of 6 months to one year to allow the

preloading to be effective. Wick drains could be used to accelerate the consolidation

process. If materials for surcharging are not available on site, they must be imported

and then excavated and removed after the preloading is completed (see Chapter 2).

• Vibro-replacement could be considered to stiffen clayey soil deposits (see Chapter 5).

• Deep soil stabilization techniques, such as jet grouting, deep soil mixing, or

compaction grouting, can be considered (see Chapters 7 and 8).

1.6 Liquefaction Potential and Mitigation Assessment 

The identification of liquefiable soils is beyond the scope of this manual, but typically they 

are loose sand deposits either natural or man-made. The chief strategy to improve loose sand 

deposits to resist liquefaction is to increase their density and/or contain the liquefiable 

deposit. In developing design guidelines for highway structures and embankments, Cooke 

and Mitchell (1999) provide the following recommendations:

• Improvement of liquefiable soils should extend to the bottom of the liquefiable

material and extend laterally to a distance equal to the depth of treatment.

• Improvement for reducing lateral deformations of embankments is more effective

when the foundation is treated in a zone between the crest and the toe of the

embankment.

• Field performance suggests that the effect on structures will be minor when the

supporting ground is improved to the “no liquefaction” side of liquefaction potential

curves, recommended in the simplified method for liquefaction potential assessment.

The deep compaction techniques of DDC and VC may be used at sites with soils that may be 

susceptible to liquefaction during earthquakes. Saturated sands, silty sands, sandy silts, and 

silts are likely to be in this category. When DDC or VC are used to densify soils for the 

support of embankments and structures, it is also necessary to confirm that there will not be a 

risk of liquefaction or other ground disturbance that could lead to loss of support and lateral 

spreading. The initial assessment of whether the soil at a site will liquefy in an earthquake is 
4-9 

made in terms of whether the in situ shear strength under cyclic loading, represented as a 

Cyclic Resistance Ratio (CRR), is less than the cyclic shear stress that will cause 

liquefaction, termed the Cyclic Stress Ratio (CSR).

Combinations of CSR and strength of the soil layer, usually determined in situ by means of 

penetration tests, have been found that define the boundary between liquefaction and no 

liquefaction over a range of peak ground motion accelerations. This boundary has been 

determined through extensive analyses of case history data from many earthquakes. Standard 

Penetration Tests (SPT), Cone Penetration Tests (CPT), and Becker Penetration Tests for 

soils containing gravel and cobbles (BPT) are used to determine the CRR. Values of CRR are 

defined by the points on the boundary curve that separates liquefaction and no liquefaction 

zones on a plot of CSR vs. penetration resistance or shear wave velocity corresponding to the 

measured and corrected in situ property. An example of such a plot for liquefaction analysis 

using the SPT is shown in Figure 4-3. 

Youd et al. 2001

Figure 4-3. SPT liquefaction chart for magnitude 7.5 earthquakes. 
4-10

Thus, if a site underlain by saturated clean sand has a corrected blow count (N1)60 of 10 

blows per foot and the anticipated cyclic stress ratio under the design earthquake is 0.25, the 

soil will liquefy unless the normalized penetration resistance (N1)60 is increased to greater 

than 22 blows per foot by densification, or the cyclic stress ratio is reduced by transferring 

some or all of the dynamic shear stress to reinforcing elements. Similar plots are available in 

terms of normalized CPT tip resistance qc1N. In each case the penetration resistance is 

normalized to an effective overburden stress of 1 atmosphere. 

Although straightforward in concept, the liquefaction potential analysis is complex in 

application, because (1) the CSR depends on the input motions within the soil layer which, in 

turn, depend on such factors as earthquake magnitude and intensity, distance from the 

epicenter, geologic setting, rock conditions, and soil profile characteristics, (2) the CRR 

depends on such factors as overburden stress, fines content of the soil, and static shear stress, 

and (3) determination of normalized values of the penetration resistance involves several 

corrections to the measured values, especially in the case of the SPT.

Information about input ground motions can be obtained from local experience and recorded 

ground motions near the site, if available, or from seismicity information obtainable from the 

following sources: 

• United States Geological Survey Ground Motion Calculator

(https://earthquake.usgs.gov/hazards/designmaps/), which can be used to obtain peak

rock accelerations for the site

• USGS Earthquakes Hazards Program website (https://earthquake.usgs.gov/hazards/),

which provides design ground motions for buildings and bridges; interactive fault

maps; scenarios of ground motions and effects of specific hypothetical large

earthquakes; and seismic hazard maps and site-specific data which includes a Beta

version of an unified hazard tool that enables determination of site-specific ground

motion parameters.

Widely used liquefaction correlation diagrams for SPT and CPT, along with discussions of 

how to make the necessary computations to obtain the CSR, (N1)60, qc1N, and the CRR are 

given in Youd, et al. (2001) and Idriss and Boulanger (2008). 

The usual design procedure for ground improvement to mitigate liquefaction using DDC and 

VC is to require that the soil be densified sufficiently to attain a factor of safety against 

liquefaction triggering, defined by CRR/CSR, greater than 1.5, with a minimum of 1.3, 

although no single value may be suitable for all conditions owing to the many factors that 

influence each specific site and problem. Each case needs to be judged on its own merit in 

the event there are a few points where the safety factor criteria are not met. A few scattered 
4-11 

locations where the safety factor is below the minimum is quite different from several low 

values that are grouped closely together. If a value fails by a large amount it is more 

significant than if it fails to meet the minimum by a small amount. Settlement attributed to 

earthquake shaking can be estimated using methods outlined in FHWA GEC 3 (1997). 

1.7 Glossary 

1.7.1 Deep Dynamic Compaction

This glossary describes terminology unique to dynamic compaction. 

Applied energy – Average energy applied at ground surface, which is calculated on the basis 

of the sum of all the energy applied by dynamic compaction divided, by the surface area of 

the densified soil. 

Contact pressure – The weight divided by the base area of the tamper.

Crater – Depression in the ground at the drop point location that results from energy 

application. 

Depth of improvement – Maximum depth to which measurable improvement is attained. 

Drop energy – Energy per blow, which is calculated on the basis of the tamper mass, 

multiplied by the drop height.

High-level energy – Energy applied to cause densification to the depth of improvement. 

Induced settlement – Average ground settlement following densification, which is 

determined by elevation readings taken before and after dynamic compaction. 

Low-level energy – Energy applied to compact the surface deposits to the depth of crater 

penetration following high-level energy application. Low-level energy application frequently 

is called the ironing pass. 

Pass – The application of a portion of the planned energy at a single drop point location. 

Multiple drops are required to deliver the energy at each drop point. If all the drops cannot be 

applied at one time because of deep craters or excess pore water pressures, another pass or 

passes will be required after excess pore water pressures dissipate, or the craters are filled 

with granular fill. There is generally a waiting period of at least a few days between passes. 

Phase – Describes the pattern in which the energy will be applied. For example, every other 

drop point of the grid pattern could be selected to be densified as Phase 1. After completion 
4-12 

of Phase 1, the intermediate drop points could be densified as Phase 2. Some projects use 

only one phase; others have been undertaken with as many as five phases.

1.7.2 Vibro-compaction

This glossary describes terminology unique to vibro-compaction. 

Vibro-compaction – A ground modification technique that uses specially designed probetype, depth vibrators for in situ densification of loose sands and gravels. 

Vibro-diplacement – Refers to the dry, top or bottom feed process; almost no in situ soil 

appears at the surface, but is displaced by the backfill material.

Vibroflotation – Refers to compaction by means of a vibroprobe, sometimes called a 

vibroflot that densifies with predominantly horizontal vibrations. Vibroflotation is the 

original name for the process and is used synonymously with vibro-compaction. 

Vibro-replacement – Refers to the wet, top-feed process in which jetting water is used to aid 

the penetration of the ground by the vibrator. Due to the jetting action, part of the in situ soil 

is washed to the surface. This soil is then replaced by backfill material. 

1.8 Primary References

1.8.1 Deep Dynamic Compaction

• FHWA. (1986). Dynamic Compaction for Highway Construction, Author: Lukas, R., 

FHWA/RD-86/133, Federal Highway Administration, U.S. DOT, Washington, D.C. 

• GEC 1. (1995). Dynamic Compaction. Author: Lukas, R., FHWA SA-95-037, 

Federal Highway Administration, U.S. DOT, Washington, D.C., 97p. 

• Slocombe, B. (2013). Dynamic Compaction. Chapter 3 in Ground Improvement, 

Third Edition, K. Kirsch and A. Bell, Editors, CRC Press, Taylor & Francis Group, 

Boca Raton, FL, pp. 57-85. 

1.8.2 Vibro-Compaction 

• Kirsch, K. and Kirsch, F. (2010). Ground Improvement by Deep Vibratory Methods. 

Spon Press, 189p. 

• Massarsch, K.R. and Fellenius, B.H. (2002). Vibratory Compaction of CoarseGrained Soils. Canadian Geotechnical Journal, 39(3): pp. 695-709. 
4-13 

• Massarsch, K.R. and Fellenius, B.H. (2005). Deep Vibratory Compaction of Granular 

Soils. Chapter 19 in Ground Improvement – Case Histories, B. Indranatna and J. Chu, 

Editors, Elsevier Publishers, pp. 633-658. 
4-14 

2.0 DEEP DYNAMIC COMPACTION 

A number of fundamental questions must be addressed before proceeding with an in-depth 

evaluation of deep dynamic compaction at a specific project site. These include the 

following: 

• What are the typical applications?

• What are the advantages and disadvantages?

• What types of deposits can be improved? 

• Are there limitations or environmental considerations? 

• Are there alternate site improvement methods?

• Is it cost-effective compared to other alternatives?

The following sections provide answers and insight to these questions. 

2.1 Feasibility Considerations

2.1.1 Applications

The primary purpose of deep dynamic compaction is to densify natural and fill deposits to 

increase bearing resistance, reduce settlement, minimize collapse potential of large voids or 

collapse-susceptible soils, and mitigate liquefaction potential so that engineered structures 

such as bridges, abutments, retaining structures, and embankments can be constructed safely 

and economically. Discussion of the uses of deep dynamic compaction is best split into the 

following: 

• Densification of loose deposits 

• Collapse of large voids and collapse-susceptible soils

• Related applications

2.1.1.1 Densification of Loose Deposits

The primary use of deep dynamic compaction is to densify loose deposits so as to reduce the 

settlement that would otherwise occur under load application. Such deposits include natural 

soils and fill deposits of soils, manmade debris, byproducts, or any combination of these. 

However, deep dynamic compaction works best on dry granular materials including sands, 

gravels, ashes, mine spoils, and saturated, free-draining soils. The densification results from 

rearrangement of the material particles into a more compact state, resulting in a lower void 

ratio. In partially saturated soils, densification is similar to laboratory impact compaction by 
4-15 

the proctor method. In saturated or nearly saturated fine sands and silts, excess pore water 

pressures develop on impact, which dislodges some of the point-to-point contacts between 

soil particles. Following dissipation of the pore water pressures, the soil grains rearrange into 

a denser state of packing at a lower water content.

In areas subject to seismic activity, saturated sand or silty sand deposits that are stable under 

static conditions may liquefy during a seismic event. Deep dynamic compaction induces high 

pore water pressures into these soils which liquefies the soils resulting in a denser state after 

dissipation of the pore water pressure, which in turn, makes them non-susceptible or lesssusceptible to liquefaction from earthquakes. 

Deep dynamic compaction has been used to improve fine-grained, cohesive soils, although 

such soils are more difficult to improve with DDC. In fine-grained soils with low saturation, 

DDC can be effective in collapsing the air void spaces. DDC works best in soils in which the 

water content is less than the plastic limit (PL). Generally, DDC is not recommended with 

soils in which the plasticity index (PI) of the soil is greater than 8 and which possess high 

degrees of saturation. In such soils the applied load is transferred to the pore water and the 

low permeability does not allow timely dissipation of the pore water pressures, limiting the 

settlement induced in such soils. Heave of fine-grained soils often occurs when subjected to 

DDC.

The soil behavior during deep dynamic compaction is illustrated in Figure 4-4. 

Courtesy of Soletanche Bachy

Figure 4-4. Phase diagram model of soil behavior during deep dynamic compaction. 
4-16 

The soil deposit is characterized as a phase diagram consisting of three portions, i.e., solid 

materials, liquids, and gas. On the left side of Figure 4-4, the phase diagram represents the 

original state of the soil deposit. During the first phase of energy application, excess pore 

water pressure develops in the liquid portion and then dissipates with time. The time required 

for pore pressure dissipation is a function of the permeability of the soil mass and the 

distance to a drainage path. There is some induced ground settlement, which is represented 

by a lowering of the top of the phase diagram. This induced settlement results from expulsion 

of some of the gas and liquid for the voids, which causes a lower void ratio. The solid portion 

remains the same.

During the second phase of energy application, the pore water pressures temporarily increase 

during tamping, but this time, the magnitude is slightly less. At the end of the second phase, 

the excess pore water pressures have once again dissipated back to the original condition. 

The thickness of the liquid portion decreases slightly less than during Phase 1, and the gas 

portion is also slightly less than in Phase 1. The induced settlement increases beyond what 

occurred during Phase 1. The void ratio of the soil also has been lowered. 

During Phase 3, excess pore water pressures are again generated, but to a lesser extent than 

during previous phases. Once again, some water is expelled as the pore water pressures 

dissipate and the induced settlement is slightly increased.

The final state of the soil mass is shown on the right side of Figure 4-4. The volume of the 

solids remains the same as the initial state, but the volume of the liquids and the gas 

decreases. The reduced void ratio results in reduced compressibility and increased strength of 

the deposit. 

2.1.1.2 Collapse of Large Voids and Collapse-Susceptible Soils

Large voids may exist within the soil mass in either natural or fill deposits. Deep dynamic 

compaction is used to collapse the voids and provide a more uniform foundation that reduces 

differential settlement.

In karst formations, voids often develop within the soil deposit as a result of erosion into an 

underground karstic limestone cavern created by the dissolution of limestone or dolomite. 

The risk of unforeseen large settlement or soil collapse into the void can be reduced by deep 

dynamic compaction. For DDC to be effective, the void should be located within the 

effective depth of treatment as shown in Figure 4-1. 

Man-made fill deposits from construction debris, solid waste, mine spoil, or mineral 

processing may have large voids created by the filling process. Voids can also be caused by 
4-17 

buried vessels, such as drums or pipes. These voids could be collapsed by DDC, provided 

they are within the zone of treatment.

In mine spoil deposits, voids are frequently present in zones where slabs of rock or nested 

clusters of large boulders exist. During the placement of these materials, no effort is 

generally made to compact the deposits or to isolate the larger chunks within the soil matrix 

of the mine spoil. In time, these voids eventually fill in with erosion of the finer materials 

into the larger voids, which in turn results in ground subsidence. 

Windblown soils such as loess and partially saturated soils deposited in arid environments 

often have a loose soil structure in which the soil grains are weakly cemented together. These 

soils are susceptible to collapse upon wetting as the cementitious bonds between the particles 

are dissolved and the soils collapse into a denser state of packing. Deep dynamic compaction 

can be used to overcome the cementitious bonds and move the particles into a more dense 

condition to reduce or eliminate settlement upon wetting.

2.1.1.3 Related Applications

In weak saturated soils relatively deep craters greater than 5 feet can develop. If these craters 

are filled with coarse granular material and supplemental energy applied, the granular 

material will be driven into the weak deposit. This type of improvement is strictly speaking 

not deep dynamic compaction and is called dynamic replacement. The dynamic compaction 

equipment is used to produce the improvement, so this procedure is a related form of ground 

improvement. The depth of improvement is generally less than about 10 to 13 feet.

2.1.2 Advantages and Potential Disadvantages 

2.1.2.1 Advantages 

The advantages of deep dynamic compaction include the following:

• Impacting the tamper into the soil serves as both a probing and a correcting tool. If 

there are weak ground conditions or large voids in local areas, the tamper will 

penetrate further into the ground than in adjacent areas, thereby causing large crater 

depths. This provides the field engineer with immediate feedback on ground 

response. A decision can then be made regarding further energy application in this 

area to correct the poor ground condition or, if the deposit will not compact upon 

energy application, to undercut and remove the poor ground. This probing aspect of 

deep dynamic compaction is important in heterogeneous deposits, such as old 

landfills, mine spoils, or in karst deposits where voids are present in local areas. 
4-18 

• Densification of the deposits can be observed as the work proceeds. If multiple passes 

are made over an area, each succeeding pass generally will result in an average crater 

depth less than that of the prior pass. This is an indication of the improvement from 

the resistance of the ground. Ground settlement readings are usually taken before and 

after each application of energy, and the amount of ground compression is an 

indicator of the degree of improvement achieved. Normally, ground compressions of 

5 to 10 percent of the thickness corresponding to the depth of treatment, as predicted 

by Equation 4-1, occurs during densification. In extremely loose fills, such as recent 

landfills, the ground compression can be 20 to 25 percent. 

• Deep dynamic compaction can be used at sites with a very heterogeneous mixture of 

deposits and at sites with gradation ranges from large boulders and broken concrete to 

silty soil particles. Deep dynamic compaction is effective in densifying all of these 

deposits with the same equipment. Furthermore, deposits that were formerly thought 

uncompactable, i.e., building rubble debris or decomposed sanitary landfills, can be 

compacted by this method. 

• Densification usually results in a bearing stratum having a more uniform 

compressibility, which minimizes differential settlements. Weaker zones within the 

deposit undergo the most improvement, which eliminates zones of potentially high 

compressibility.

• Densification can be achieved below the water table in pervious and semi-pervious 

deposits, which eliminates costly dewatering and/or lateral bracing systems required 

for conventional excavation and replacement techniques. 

• Except for the very heavy tampers and the high drop heights, non-specialty 

contractors can perform deep dynamic compaction on a local basis, making the cost 

for deep dynamic compaction very competitive. For the larger tampers and the higher 

drop heights, the equipment must be modified because cable and drum wear is higher 

than normal. In this case, specialty contractors are required to perform the work. 

• Deep dynamic compaction can proceed during inclement weather conditions, 

including freezing weather or rain, provided precautions are taken to minimize water 

accumulation and frost penetration. Excess surface water should be removed by 

sloping the ground to shed water or by pumping. Deep frost penetration in fine grain 

deposits should be prevented by covering the surface with soil or straw removed 

immediately prior to deep dynamic compaction.

2.1.2.2 Potential Disadvantages

Potential disadvantages of deep dynamic compaction include the following: 
4-19 

• Deep dynamic compaction produces ground vibrations that can travel significant 

distances from the point of impact. In urban areas, this may require the use of 

lightweight tampers and low drop heights, as well as limiting deep dynamic 

compaction to areas well within the property lines. At some sites, shallow isolation 

trenches have been cut through the upper portion of the soil mass to reduce the 

transmission of energy off site. 

• To prevent surface softening of the soil mass, as well as sticking of the tamper into 

the soil, the ground water table should be located more than 6 feet below ground 

surface. If the water table is higher than 6 feet, it may be necessary either to lower the 

water table by pumping or raise the grade by soil placement.

• At sites consisting of very loose deposits, such as recent landfills, it is frequently 

necessary to place a layer of granular material, such as gravel or crushed stone, at the 

surface to provide a working platform for equipment operation and to limit tamper 

penetration at impact. The surface layer also provides confinement for the underlying 

weak deposits. The cost of the granular fill can significantly add to the cost of the 

deep dynamic compaction operation. 

• Lateral ground displacements of 1 to 3 inches have been measured at distances of 

about 20 feet from the point of impact of 30 to 60 kip tampers. Utilities or buried 

vessels within the zone of influence could be displaced or damaged. 

2.1.3 Feasibility Evaluations

2.1.3.1 Geotechnical

The suitability of deep dynamic compaction is evaluated with the following soil parameters:

• Permeability of the soil mass, which can be inferred from soil classification

• Degree of saturation, which is related to the position of the water table 

• Length of drainage path 

• Soil stratigraphy, such as buried hard or weak layers 

Deep dynamic compaction works best on deposits where the degree of saturation is low, the 

permeability of the soil mass is high, and the drainage is good. Deposits considered most 

appropriate for deep dynamic compaction are pervious granular soils. If these deposits are 

situated above the water table, densification is immediate, as the soil particles are compacted 

into a denser state. If these deposits are situated below the water table, the permeability of the 

soils is usually high, and excess pore water pressures generated by the impact of the tamper 

dissipate almost immediately and improvement is correspondingly immediate. Pervious 
4-20 

granular deposits include natural sands and gravels and fill deposits consisting of building 

rubble, granular mine spoil deposits, industrial waste fills, such as certain types of slag, and 

decomposed refuse. 

Deposits for which deep dynamic compaction is not appropriate include saturated clayey 

soils (either natural or fill). In saturated deposits, improvements cannot occur unless the 

water can be expelled from the voids. In clayey soils where the permeability is low, the 

excess pore water pressures generated during deep dynamic compaction require a lengthy 

period of time to dissipate, which renders deep dynamic compaction impractical for these 

deposits. Furthermore, the degree of improvement is generally minor in saturated clayey 

deposits. The upper portions of these deposits can be improved by dynamic replacement (see 

section 2.1.1.3).

Intermediate between the two extremes of granular pervious soils and saturated clay deposits 

is a third category of semi-pervious soils. Silts, clayey silts, and sandy silts are in this 

category. Deep dynamic compaction will work in these deposits, but because of the lower 

than desired permeability, the energy must be applied using multiple phases that allow excess 

pore water pressures to dissipate between energy applications. Sometimes the excess pore 

water pressure dissipation occurs over periods of days to weeks. At some project sites, wick 

drains have been used to shorten the drainage path, and thereby dissipate pore pressure. 

Unfortunately, not all the influential soil parameters are determined as part of the 

preconstruction field exploration process, nor can they be measured with accuracy. An 

estimate of the field permeability can be obtained through slug tests performed in boreholes, 

but this is not undertaken for most projects. Instead, the permeability is generally inferred 

from soil classification or soil index tests. Some idea of the length of the drainage path can be 

obtained from examination of the soil boring logs, although thin sand seams within a finegrained soil deposit could escape detection. Thus, drainage can only be estimated 

subjectively. In fine-grained soils, the degree of saturation can be estimated from laboratory 

tests such as unit weight, water content, and specific gravity. In coarse-grained soils, it can be 

assumed that the soils are partially saturated above the water table and fully saturated below. 

A guide for estimating the suitability of deep dynamic compaction for various soil deposits is 

provided in Figure 4-5, based upon conventional index tests. 
4-21 

GEC 1 1995, After FHWA 1986

Figure 4-5. Grouping of soils for deep dynamic compaction.

Three categories of deposits shown in Figure 4-5 are summarized below.

• Zone 1 represents pervious soil deposits where deep dynamic compaction is best 

suited. The permeability of these deposits is generally greater that 10-5 m/s. 

Improvements are obtained regardless of whether these deposits are partially or fully 

saturated.

• Zone 3 represents impervious clayey deposits. The permeability of these soils is 

generally less than 10-8 m/s. Deep dynamic compaction is not recommended if these 

deposits are saturated or nearly saturated. Some improvement has been achieved in 

clayey fill deposits that are only partially saturated, provided the water content at the 

time of densification is at or below the plastic limit.

• Zone 2 represents intermediate soil deposit where improvements can be achieved, 

provided the energy is applied in multiple phases to allow complete pore pressure 
4-22 

dissipation between phases. The permeability of these deposits is generally between 

10-5 to 10-8 m/s.

The suggested procedure for evaluating whether the soil deposits at any site are suitable for 

deep dynamic compaction is to first determine the grain size distribution of representative 

samples to determine where the gradation plots on Figure 4-5. If the gradation plots in Zone 

1, proceed with the design and construction sequence for deep dynamic compaction. If the 

soils are in the range of Zones 2 or 3, additional testing, including water content and specific 

gravity, will help determine the degree of saturation. It would also be necessary to determine 

the position of the water table and to establish whether any pervious seams are present within 

the soil deposit. Additional field permeability testing should be considered since field 

permeability, not grain size gradation, dictates the suitability of the soil. If the testing 

confirms that the deposit is in Zone 3, consider alternate methods of ground improvement. If 

the deposits are classified as Zone 2, deep dynamic compaction will work, but the energy 

should be applied in phases. 

On some projects, test sections are used to evaluate the effectiveness of deep dynamic 

compaction before construction. This is especially important when the soil stratigraphy is 

variable. Strata of hard or weak deposits within an otherwise uniform deposit can affect the 

depth and degree of improvement. Soil borings with Standard Penetration Tests (SPT), Cone 

Penetrometer Tests (CPT), or Pressuremeter Tests (PMT) are generally completed before and 

after energy application to measure soil improvement. Different levels of energy can be 

applied and evaluated. In fine-grained soils, piezometers can be installed to measure the 

magnitude of pore water pressures generated by deep dynamic compaction and the time for 

dissipation. Information gathered from a properly instrumented and monitored test section is 

very helpful in preparing more meaningful specifications that could lower construction costs 

and prevent construction delays.

2.1.3.2 Environmental Considerations 

The key environmental considerations are noise, vibrations, and lateral movements, which 

can affect neighboring structures and people. Typical noise levels on an open site can reach 

100 to 120 dB at 40 feet decreasing to less than 80 db at 80 feet (Dumas and Beaton 1992). 

Normally, the noise issue is only present during short periods during the lifting and dropping 

cycle. Ground vibrations can travel significant distances and may travel offsite affecting 

nearby property and structures. Detailed information regarding vibration analysis is included 

in this document in Section 2.1.4. Lateral ground displacement adjacent to craters may 

damage existing utilities and structures. During construction, airborne debris can cause 

damage or injury unless proper precautions are taken. 
4-23 

2.1.3.3 Site Conditions 

• The use of deep dynamic compaction requires sufficient headroom for the cranes

(equipment).

• Adjacent buildings, structures, utilities and property must be monitored for potential

vibrations when using DDC.

• Vibration, noise and lateral movements.

Site conditions should always be considered when selecting a ground modification 

technology. Site topography and in situ soil conditions can have a considerable effect on the 

economics of a DDC solution. The site investigation should establish the site soils, whether 

clay, silt, sand or gravel; soil densities and water contents; the groundwater level; and the 

presence of overhead wires, buried utilities, or nearby structures. Some of the specific site 

and soil conditions that affect the economics or feasibility are listed below:

• Uneven and unstable working surface. The cranes for DDC need a level and stable

working surface. If the ground surface is uneven, a working layer can be placed on

which the cranes can operate. This layer is best comprised of free-draining materials.

The weight of the cranes must be accommodated by the working platform.

• Headroom. To the large sized cranes used for DDC, sufficient headroom must be

available for the operation to proceed. Overhead wires can possibly present a logistic

problem. Overhead lines must be de-energized to avoid possibilities of accidental

impact or arcing of current from the lines to the mast.

• Obstructions above the compressible layer. The key issue with obstructions such as

concrete, rock, rubble, slag, brick, wood, riprap, stone, debris, rubbish, or trash near

the surface is the reduction in energy reaching the compressible layer. If these

materials are present in the upper reaches, DDC will help to densify such materials,

but the energy transmission to the materials below will be reduced and additional

drops may be necessary to achieve the desired compaction.

• Depth. The soil is improved to depths between 10 to 35 feet. If deeper improvement

is necessary, deep dynamic compaction in combination with other systems, such as

grouting and stone columns, should be considered.

Should any of the above site conditions be encountered, it would be advisable to contact 

specialty contractors experienced in DDC in order to determine the magnitude of difficulty. 
4-24 

2.1.4 Limitations

Whenever a tamper impacts the ground, vibrations are transmitted through the subsurface 

with diminishing intensity, as the distance from the point of impact increases. If deep 

dynamic compaction is undertaken close to property lines, ground vibrations transmitted off 

site should be considered. 

There are no general regulations of construction and industrial vibrations. Based on research 

related to blasting, the U.S. Bureau of Mines found that building damage is related to particle 

velocity (Siskind et al. 1980). The Bureau of Mines developed Figure 4-6 based on 

experiences with damage measurements made in residential construction from blast-induced 

vibrations. 

After Svinkin 2015

Figure 4-6. U.S. Bureau of Mines safe level of blasting vibrations for houses and Office 

of Surface Mining regulations. 
4-25 

The limiting particle velocity depends upon the frequency of the wave form. Normally, deep 

dynamic compaction results in frequencies of 5 to 12 Hz. Using Figure 4-6 as a guide, this 

would limit peak particle velocities to values of 1⁄2 inch/sec for older residences with plaster 

walls and 3⁄4 inch/sec for more modern constructions with drywall. Peak particle velocities 

that exceed the values given in Figure 4-6 do not mean that damage will occur. Rather, these 

values are the lower threshold beyond which cracking of the plaster or drywall may occur. 

Data generated by the U.S. Bureau of Mines indicate that minor damage occurs when the 

particle velocity exceeds 2 inch/sec, and major damage occurs when the particle velocity 

exceeds about 7.6 inch/sec. Thus, keeping the particle velocity less than about 1⁄2 to 3⁄4 

inch/sec should be a reasonably conservative value to minimize damage.

Normally, the ground vibrations are measured with a seismograph at the time of construction. 

The readings are taken on the ground adjacent to nearby structures. However, before starting 

deep dynamic compaction operations, it is necessary to predict the particle velocity of ground 

vibrations, because this may affect the level of energy application in close proximity to 

existing facilities. For planning purposes, Figure 4-7 can be used. 

FHWA 1986

Figure 4-7. Scaled energy factor versus particle velocity.
4-26 

The square root of the energy of a single drop (drop height times the mass of tamper) divided 

by the distance from the point of impact is used to calculate the scaled energy factor. This 

value, and the type of soil deposit that most closely resembles the soil being densified, can 

then be used to predict the particle velocity. The basis of Figure 4-7 is actual readings 

obtained at specific sites.

If the predicted particle velocity is higher than desired, it will be necessary either to reduce 

the energy or increase the distance between the point of impact and the adjacent facility. 

Either would reduce the scaled energy factor. At some sites, trenches have been dug along 

the property line to reduce the particle velocity. This was found to be partially helpful in 

reducing the surface waves that travel off site. The effectiveness of the trenches can be 

established at the time of construction from vibration readings taken on the near and far side 

of the trench following impact of the tamper. 

Even though damage may not occur, ground vibrations will still be felt by humans; this can 

be annoying and lead to complaints. The relative response by humans is shown on the right 

side of Figure 4-7. Ground vibrations would be disturbing to people in the range of 1⁄2 to 3⁄4 

inch/sec (13to 19 mm/sec). Some education on the part of the owner or contractor will be 

necessary to reduce the fears of the adjacent property owners if the ground vibrations are in 

this range. 

Buried utilities tolerate higher vibration levels without damage than do buildings because the 

utilities are surrounded in a soil mass. The utility moves with the soil mass and remains 

confined. Water mains and water pipes have sustained particle velocities of 3 inch/sec

without damage. 

Deep dynamic compaction also causes lateral ground shifting. Measurements taken for three 

different soil types, with inclinometers located at 3 m and 6 m (10 to 20 feet) from the point 

of impact of tampers in the range of 15-29 Mg (33-64 kips) are presented in Figures 4-8 and 

4-9. 
4-27 

GEC 1 1995, After FHWA 1986

Figure 4-8. Lateral movements at 3 m (10 feet) from drop point. 

GEC 1 1995, After FHWA 1986

Figure 4-9. Lateral movements at 6 m (20 feet) from drop point. 
4-28 

At a distance of 3.0 m (10 feet) from the point of impact, lateral deformations as much as 254 

to 305 mm (10 to 12 inches) were observed, while at 6 m (20 feet), lateral deformations 

ranging from 13 to 76 mm (1⁄2 to 3 inches) were reported. These are permanent deformations 

that occur as the ground is displaced laterally following impact.

Based upon these measurements and observations from other sites, it appears that deep

dynamic compaction operations with tampers ranging from 15-30 Mg (33-66 kips) should 

not be undertaken within about 6 to 7 m (23 to 26 feet) of any buried structures, if movement 

could cause damage to these structures. This could include a utility or a shallow structure 

foundation. 

Occasionally, debris may become airborne following the impact as the tamper strikes the 

ground. This is likely to occur in dry soil deposits and those that contain larger objects, such 

as cobbles, boulders, or even landfill sites containing bricks and bottles. To avoid being hit 

with flying debris, a safe working distance from the point of impact should be delineated. 

Where deep dynamic compaction is to be undertaken immediately adjacent to a street or 

another facility, it may be necessary to erect a shield or barrier to deflect flying particles.

Safety and cost issues arising from rapid cable wear must be addressed by the contractor. The 

solution may include the use of swivels, the reduction of cable lengths, use of a tag line, use 

of non-rotating surface cables, placement of rubber tires over the tamper, or other specialized 

equipment modifications or features (FHWA 1986). 

There is a depth limitation on improvement. The heaviest tamper that can be lifted with 

conventional equipment is about 40 kips with a drop height of 75 to 90 feet. This will result 

in a maximum improvement depth of about 36 feet. If deeper improvement is necessary, deep 

dynamic compaction in combination with other systems, such as grouting and stone columns, 

should be considered.

2.2 Construction and Materials

2.2.1 Overview

Conventional lifting cranes are used for deep dynamic compaction projects where the tamper 

size is less than about 40 kips. Tampers are sometimes built especially for deep dynamic 

compaction, while other tampers have been fashioned from steel ingots or other sources, such 

as a bank vault door. 
4-29 

Associated pieces of equipment include a front-end dozer for leveling the ground after the 

craters are formed. Imported granular material may also be required to provide a firm 

working mat across the site.

Because of the conventional and readily available equipment required for most deep dynamic 

compaction projects, many general, demolition, or excavation contractors will perform deep 

dynamic compaction when properly guided and monitored.

2.2.2 Tampers

Deep dynamic compaction generally is performed with tampers ranging from 11 to 60 kips. 

The lighter tampers are used where the thickness of the deposit is relatively thin, such as 10 

feet, and the heaviest tampers are used where the deposit is about 30 to 40 feet thick.

The tampers must be very rugged because high stresses are induced in the tamper when it 

strikes the ground. Most tampers are constructed of solid steel, but some have a steel base 

plate and steel sidewalls, with the interior filled with concrete. Tampers constructed solely of 

concrete have a relatively short life.

The tampers should have a flat base and can be either square, round, or hexagonal. Generally, 

the tamper rotates as it is lifted because the cable unwinds due to the heavy load. Therefore, a 

round tamper will always hit on the same imprint. Square tampers have also been used, but a 

rounded crater pattern develops. Sometimes guy wires are used to keep the tamper from 

rotating as it is lifted and dropped. 

The contact pressure at the base of the tamper should be 5 to 10 psi. This pressure is obtained 

by dividing the weight of the tamper by the area of the base. A significantly higher tamperpressure could cause penetration into the ground without densification. Pressures less than 

about (5 psi will distribute the energy over too wide an area to cause deep densification. 

Photographs of 12, 30, 31, and 70 kip tampers are shown in Figures 4-10 through 4-13, 

respectively. 
4-30 

Figure 4-10. 12 kip tamper. 

Figure 4-11. 30 kip tamper. 
4-31 

Figure 4-12. 31 kip tamper.

Figure 4-13. 70 kip tamper. 
4-32 

The 12 kip tamper has a round 6 inch thick steel base plate and a 1 inch thick steel cylinder 

that is filled with concrete. The 30 kip tamper also has a large base plate with steel plates 

stacked on top to make up the remaining portion of the tamper. The plates are welded and 

bolted together. The 70 kip tamper is solid steel.

Tampers with a low contact pressure are sometimes used to provide surface compaction after 

the deep densification is finished. The contact pressure at the base of these tampers is 

typically 1.5 to 5 psi. A tamper used for the ironing pass is shown in Figure 4-14. On some 

projects, surface densification is achieved with either conventional compaction equipment or 

by proofrolling with a fully loaded dump truck. 

Figure 4-14. Tamper with low contact pressure for ironing pass. 

2.2.3 Lifting Equipment

To deliver the maximum amount of energy to the ground, the tampers are lifted and dropped, 

using a single cable and a drum on the lifting equipment, which has a free spool. The only 

losses that occur from this type of equipment are friction in the free-spool drum, friction of 

the cable over the upper sheave of the boom, and some air resistance at the base of the 

tampers. Typical pieces of equipment used to raise and drop the tamper are shown in Figures 

4-15 through 4-17. 
4-33 

Figure 4-15. Lifting crane of 70 kips rated capacity used with 12 kip tamper. 

Figure 4-16. Dragline-type lifting crane with 11⁄4-inch cable for use with 40 kip tamper. 
4-34 

Figure 4-17. Crane with special lifting drum attached to rear of crane for use with 64 

kip tamper. 

There are also specialized pieces of equipment that will raise the tamper with multiple-part 

lines. Then a release mechanism allows the tamper to free fall. This also is an acceptable 

method for applying the energy. 

Conventional lifting cranes generally are used for tampers in the size range of 10 to 40 kips. 

The type of crane and cable size required to repeatedly raise and drop tampers of different 

sizes are listed in Table 4-1. 

Table 4-1. Equipment Requirements for Different Sized Tampers

Tamper Weight Crawler Crane Size Cable Size

8 to 16 kips 80 to 100 kips 3⁄4 to 7⁄8 inch

16 to 28 kips 100 to 200 kips 7⁄8 to1 inch

30 to 36 kips 200 to 250 kips 1 to 11⁄8 inch

36 to 50 kips 300 to 350 kips 11⁄4 to 11⁄2 inch

For tamper weights in the range of 35 to 50 kips, the conventional lifting equipment must be 

modified to prevent breakdowns. When the tamper is released, a rocking motion causes a 

stress in the shaft connecting the cab to the tracks. To reduce this stress, spuds can be placed 

on the rigs to reduce rocking. In addition, the lifting drums are enlarged and made thicker to 

withstand the forces. 
4-35 

Specialized lifting equipment has been devised for tampers 50-60 kips. A crane developed 

for large-sized tampers is shown in Figure 4-18. 

Figure 4-18. Lifting crane built only for deep dynamic compaction with 5.0-foot 

diameter lifting hoist and single line-rated capacity of 50 ton at line speeds of 85 

feet/min used with 32 ton tampers. 

The lifting drum is approximately 5 feet in diameter. The upper carriage of the crane is fixed 

to the tracks, so the crane cannot rotate in a horizontal direction. This reduces damage to the 

connection between the cab and the tracks but makes maneuvering around the site more 

difficult.

A crane using multiple part lines to lift a 64 kip tamper, that is then allowed to free fall and 

impact the ground, is shown in Figure 4-19. 
4-36 

Figure 4-19. Multiple part line used to lift 64 kip tamper (below grade in photo). 

2.2.4 Surface Stabilizing Layer

Where soft ground conditions prevail, it may be necessary to place a surface stabilizing layer 

to allow travel of the deep dynamic compaction equipment across the site, as well as to 

reduce penetration of the tamper into the ground. Soft deposits would include fairly recent 

landfills with a thin cover, or a mine spoil deposit that has weathered to a softer clay 

consistency at the surface.

The stabilizing layer usually consists of a granular material with a typical gradation range 

from 6 inch maximum size, down to sand size. The thickness of these layers depends upon 

the stability of the surface deposits, but thicknesses ranging from 1 to 3 feet have been used 

successfully. Deep dynamic compaction equipment is shown in Figure 4-20 working on a site 

where crushed rock was placed to a depth of 3 feet over a landfill.
4-37 

Photo courtesy Bob Lukas, Ground Engineering Consultants, Inc.

Figure 4-20. Deep dynamic compaction over a 3-foot layer of crushed rock placed on a 

landfill. 

At sites where the deposits are more stable (i.e., building rubble deposits, loose granular 

outwash deposits, or old decomposed landfills that are elevated above the water table), 

surface stabilizing layers are not needed. Because the stabilizing layer can cost as much as 

the deep dynamic compaction, it is used only where absolutely necessary. 

2.2.5 Construction Sequencing

Deep dynamic compaction is generally undertaken on a grid pattern throughout the entire 

project area and extends beyond the limits of the project for a distance equal to the thickness 
4-38 

of the weak deposit being densified. Energy can be applied using single or multiple phases, 

as single or multiple passes.

A phase is the application of the energy in a specific pattern. For instance, initial energy 

applications undertaken on 25 foot centers across the area could be labeled Phase 1. Phase 2 

could be the application of energy midway between initial grid points. A single phase of 

energy application is most often used on projects where pore pressures dissipate rapidly and 

the energy can be applied on grid point locations immediately adjacent to a grid point that 

has just been densified. Multiple phases are used on projects where pore pressures rise and 

take some time to dissipate; therefore, it is more effective to apply energy on an intermittent 

grid pattern.

A pass is the application of energy in increments at each specific drop point location. For 

instance, if the plan is to impart 12 drops at a specific grid point location, but only 3 drops 

can be applied before the crater depths become excessive or ground heaving occurs, the first 

3 drops would be called the first pass. After the first pass is completed, pore water pressures 

are allowed to dissipate, and the craters are filled. As additional drops are applied, they 

would be called the second pass. In fine-grained deposits, it is sometimes necessary to use 3 

or 4 passes, whereas in the more pervious deposits, only 1 pass is needed. 

While most projects are dynamically compacted on a grid pattern, some projects require 

additional energy application at specific locations. At the footing locations, for instance, 

additional energy can be applied. On projects where there are karst formations, the induced 

settlement may be larger in some areas, which indicates the presence of voids; additional 

energy can be applied at these locations.

2.3 Design

2.3.1 Design Considerations 

If the preliminary design considerations discussed in Section 2 indicate that deep dynamic 

compaction will be appropriate on a project, the next step is to prepare a more specific or 

detailed plan for the deep dynamic compaction procedures to be used. This plan would 

include the following: 

• Determine the project performance requirements for the completed structure

• Select the tamper mass and drop height to correspond to the required depth of

improvement

• Estimate the degree of improvement that will result from deep dynamic compaction
4-39 

• Determine the applied energy to be used over the project site to produce the 

improvement 

The information presented in this chapter is a brief summary of the design concepts. The 

FHWA GEC 1 (1995) provides additional details. 

2.3.2 Performance Requirements

Deep dynamic compaction densifies the soil mass and this, in turn, improves soil shear 

strength and reduces compressibility. The minimum property values required for adequate 

performance of the new facility should be determined using conventional analysis, which is 

usually based upon Standard Penetration Test (SPT), Cone Penetration Test (CPT), or 

Pressure Meter Test (PMT) results. The required property values can then be compared with 

the estimated property improvements, following densification by methods outlined in this 

chapter. On this basis it can be determined whether deep dynamic compaction is capable of 

producing the desired effect.

As an example, if a roadway embankment is to be constructed over weak ground, one 

concern is settlement of the embankment. Most embankments can withstand up to 6 inches of 

settlement without detrimental performance of the pavement system, provided the settlement 

is reasonably uniform, doesn't occur next to a pile-supported structure, and occurs very 

slowly (Holtz 1975). Based upon the estimated properties of the soil following densification, 

a settlement prediction can be made for the height of embankment to be constructed, to 

determine if the settlement will be less than 6 inches. If so, deep dynamic compaction would 

satisfy the project requirements. Conversely, if the predicted settlement using conventional 

analysis would exceed what the embankment can tolerate, other methods of supporting the 

embankment should be considered. 

Where bearing capacity of the foundation soil is a concern, a similar procedure could be 

followed. Estimated properties based on procedures given in this chapter could be used in a 

conventional analysis to determine if the bearing capacity has increased sufficiently to 

prevent failure.

If the purpose of densification is to prevent liquefaction, conventional analyses are generally 

based upon SPT tests or CPT tests. The minimum SPT or CPT value to prevent liquefaction 

is determined by this analysis. The estimated SPT or CPT value following densification can 

be determined from the procedures in this chapter. If the minimum SPT and CPT values are 

attainable, then deep dynamic compaction would be one suitable method of ground 

improvement. 
4-40 

The procedures in this chapter for estimating improvement are conducted before the project 

begins, but verification of the improvement by additional field testing is required.

In uncontrolled fill deposits, there is always the possibility of an extremely loose pocket of 

soil in an otherwise medium dense fill deposit. This loose pocket of soil may not have been 

encountered in any of the initial borings, but if the engineer is aware that the site consists of 

uncontrolled fill, deep dynamic compaction could reduce the risk of differential settlement 

from these unforeseen pockets of loose ground. Since deep dynamic compaction is 

undertaken on a grid pattern throughout a site, the presence of the loose pockets will show up 

during the compaction process. Additional energy can be applied to densify these loose fills, 

or some of the loose deposits can be undercut and replaced with a granular material that is 

then dynamically compacted.

2.3.3 Design Procedure

2.3.3.1 Depth and Degree of Improvement 

Depth of improvement is a function of a number of variables, including the mass of the 

tamper, the drop height, the soil type, and the average energy applied. All factors are 

accounted for in the following relation: 

2/1 = WHnD )( [Eq. 4-1] 

where, 

D = depth of improvement in meters 

n = empirical coefficient that is approximately 0.5 for most soils. As 

shown in Figure 4-21, n varies between 0.3 and 0.8. A value of 0.4 is 

typically used for landfills. 

W = mass of tamper in Megagrams (metric tonnes = 1.10 US ton) 

H = drop height in meters (1 meter = 3.28 feet)
4-41 

Mayne et al. 1984

Figure 4-21. Trend between apparent maximum depth of influence and energy per 

drop. 

Other factors affecting the depth of improvement include the presence of soft energy 

absorbing layers, such as low strength clay or organic deposits, and/or hard layers at the 

surface that do not allow the energy to be transmitted to greater depths. A more thorough 

discussion of factors influencing the depth of improvement is presented in FHWA (1986) and 

Han (2015). 

Various combinations of W and H can be used in Equation 4-1, depending upon equipment 

availability. The relationship between drop height and tamper mass that has been used on 

numerous projects is summarized in Figure 4-22. This figure can be used as a starting point 

with adjustments in W and H made after a contractor is selected.
4-42 

1 meter = 3.28 feet; 1 Mg = 2.2 kips

Mayne et al. 1984

Figure 4-22. Relationship between tamper mass and drop height. 

The amount of improvement resulting from deep dynamic compaction is generally measured 

by conventional in situ testing techniques such as SPT, CPT, or PMT. Test values obtained 

after deep dynamic compaction are compared with initial values before deep dynamic 

compaction to monitor the improvement. The greatest amount of improvement is generally 

near the upper portion of the soil layer densified and then decreasing with depth. The 

variation of improvement with depth is illustrated in Figure 4-23. 
4-43 

GEC 1 1995, After FHWA 1986

Figure 4-23. Variations in improvements with depth during deep dynamic compaction.

The degree of improvement achieved is primarily a function of the average energy applied at 

the ground surface. In general, the greater the amount of energy, the greater the degree of 

improvement. However, there are limitations to the maximum SPT, CPT, or PMT values that 

can be reached.

Based upon a review of a number of project sites, upper-bound test values of SPT, CPT, and 

PMT tests are shown in Table 4-2 (FHWA 1986). These indicate maximum values of 

improvement at depths of D/3 to D/2. Above and below this depth range, the test values 
4-44 

would be less. At project sites where these test values were measured, the average energy 

applied was typically on the order of 41,300-62,000 ft-lbf/ft2. If a lesser amount of energy is 

applied, or if there is some other complicating factor, such as energy absorbing layers or a 

hard surface crust that does not allow the energy to penetrate deeper, the degree of 

improvement could be significantly less than that shown in Table 4-2. 

Table 4-2. Upper Bound Test Values after Deep Dynamic Compaction 

Soil Type 

Standard 

Penetration

Resistance

(blows/12 inches) 

Static Cone 

Tip

Resistance

(tsf) 

Pressuremeter

Limit Pressure 

(tsf) 

Pervious coarse-grained soil:

• sands and gravel 30 to 50 200 to 300 20 to 25 

Semipervious soil: 

• sandy silts

• silts and clayey silts

25 to 35 

20 to 35 

135 to 177 

104 to 135 

14.6 to 20 

10.4 to 14.6 

Partially saturated impervious 

deposits: 

• clay fill and mine spoil

20 to 40* N/A 14.6 to 20 

Landfills 15 to 40* N/A 5.2 to 10.4 

* Higher test values may occur because of large particles in the soil mass.

Source: FHWA 1986

The upper-bound test values given in Table 4-2 should be used only for estimating the 

maximum degree of improvement that is possible at a project site. The actual improvement 

may be less and should be measured by field tests, such as SPT, CPT, or PMT, after deep 

dynamic compaction. Some testing should be conducted while the dynamic compaction 

equipment is at the site so additional energy application can be applied if the degree of 

improvement has not reached the desired value. However, it has been found that there is a 

delayed improvement with time following deep dynamic compaction, after all excess pore 

water pressures have dissipated. Lukas (1997) reported on 5 projects where improvements in 

soil properties were measured years after deep dynamic compaction. The improvements in 

standard penetration or pressuremeter tests ranged from 15 to 200 percent. Schmertmann 

(1991) stated that improvements of 50 to 100 percent in many soil properties can occur. 

The reason for the strength gain with time is not fully known. This phenomenon has been 

attributed by various authors to cementation between soil grains, secondary consolidation, 

and aging. This delayed strength gain occurs mostly in fine-grained soils, but has also been 

measured in granular soils. Thus, if the tests taken at the end of deep dynamic compaction are 

only borderline acceptable, additional testing could be conducted at a later date to confirm 

the improvement that occurs in soil properties with time.
4-45 

A detailed discussion of the reasons for delayed improvements from site improvement is 

discussed in Schmertmann (1991). 

2.3.3.2 Energy Requirements

Deep dynamic compaction is generally undertaken on a grid pattern throughout the area. For 

this reason, it is convenient to express the applied energy in terms of average values. This 

average applied energy can be calculated on the basis of the following formula: 

2 ( )

))()()((

grid spacing

PNHW AE =

[Eq. 4-2] 

where, 

AE = applied energy in tm/m2 (9.76 tm/m2 = 1 TSF)

N = number of drops at each specific drop point location 

W = tamper mass in Mg (metric tonnes; 1 metric tonne = 1.10 US ton) 

H = drop height in m (1 meter = 3.28 feet) 

P = number of passes 

Grid spacing in m (1 meter = 3.28 feet)

In using Equation 2, the following unit conversions may be useful, depending on units used: 

1 metric tonne = 9.807 kN = 1 Mg 

1 U.S. ton = 0.907 metric tonne 

1 tonne-meter/m2

 = 9.807 kJ/m2 = 0.1024 TSF 

If different sized tampers and drop heights are used, the average applied energy would be the 

sum of all the levels of effort. For example, the high level of energy is generally applied with 

a heavy tamper and a high drop height. This is frequently followed with a low level of energy 

(called an ironing pass), using a smaller sized tamper and a lower drop height. The average 

applied energy would be the sum of the energy imparted to the site.

Most projects have been completed using energy levels ranging from approximately 21,000 

to 62,000 ft-lbf/ft2

 (1 to 3 MJ/m2). However, the energy can be varied for each project site 

depending upon the degree of improvement needed to satisfy the project requirements. 
4-46 

During the initial design or planning stages, an estimate of the required energy can be 

obtained using Table 4-3 in units of volume. 

Table 4-3. Applied Energy Guidelines

Type of Deposit 

Unit Applied 

Energy ft-lbf/ft3 

(kJ/m3) 

Percent

Standard

Proctor 

Energy

Pervious coarse-grained soil (Zone 1 of Figure 4-5) 4,100 to 5,200 

(200 to 250) 33 to 41 

Semipervious fine-grained soils (Zone 2 and clay 

fills above the water table and Zone 3 of Figure 4-5) 

5,200 to 7,200 

(250 to 350) 41 to 60 

Landfills 12,400 to 22,700 

(600 to 1100) 100 to 180 

Note: Standard Proctor energy equals 12,400 ft-lbf/ft3

This table incorporates some of the variables that influence the amount of energy required to 

achieve adequate improvement. This includes the thickness of the deposit, the type of soil 

being densified, and the initial relative density of the deposit.

The deposits are grouped into three broad categories of soils, with landfills requiring the 

greatest amount of energy and coarse-grained soils the least. The thickness of the deposit has 

been considered in Table 4-3 by listing the applied energy in terms of a unit volume. The 

average energy to be applied at the surface of the deposit can be obtained by multiplying the 

Unit Applied Energy by the thickness of the deposit. The initial relative density of the deposit 

is taken into account by showing a range in the amount of energy required. Within any 

particular soil type, the loosest deposits would require the higher level of suggested energy, 

and the denser deposits would require the lower level of suggested energy. 

2.3.4 Example 

An example helps illustrate the use of Table 4-3 for planning purposes. Consider a site that 

consists of a loose rubble fill with a thickness of 26 feet. The rubble fill can be classified as a 

pervious coarse-grained soil, typical of Zone 1. Because the deposit is in a loose condition, 

the upper bound of applied energy for this deposit of 5,200 ft-lbf/ft2 would be used. Since the 

deposit is 26 feet thick, 41,000 ft-lbf/ft2

 of average applied energy is required. Substituting 

this number into Equation 4-2, and using the tamper mass and drop height from Equation 4-

1, different combinations of drops, grid spacing, and passes can be evaluated. 

The relationship between the number of drops, N, and the number of passes P, and the 

calculated applied energy is developed by Equation 4-2. The required mass of tamper, W, 

and drop height, H, is calculated from Equation 4-1. The grid spacing usually is selected as 
4-47 

1.5 to 2.5 times the diameter of the tamper. Equation 4-2 is solved for N x P. It is more 

efficient for the contractor to apply all the energy in one pass, than multiple passes. Using 

P=1, the number of drops at each grid point can be determined. 

Multiple passes are used when not all the drops can be made at each grid point at one time, 

while the product of N and P remains the same whether there is a single pass or multiple 

passes.

The energy requirements shown in Table 4-3 can be related to standard proctor energy 

(ASTM D698), which is equal to 12,400 ft-lbf/ft2

. Pervious coarse-grained soils generally 

require about 33 to 41 percent of this energy to achieve adequate densification, and this is 

caused by some densification that occurs naturally in these deposits after they have been in 

place for a while. Conversely, landfill deposits require 100 to 180 percent of the standard 

proctor energy. Landfills are generally in an extremely loose condition and may even be 

underconsolidated, so a significantly higher level of energy is required to densify these 

formations. 

Close monitoring of the field operations coupled with SPT, CPT, or PMT during deep 

dynamic compaction should be used to verify whether the desired improvement is reached. 

2.4 Construction Specifications and Quality Assurance 

This chapter presents a brief summary of methods that can be used in contracting for deep 

dynamic compaction, plus an overview of the monitoring that normally occurs during the 

field work. Complete construction specifications are detailed in FHWA GEC 1 (1995). 

Additional information on specifications can also be found in GeoTechTools. 

2.4.1 Contracting Procedures

The designer will develop a set of drawings and specifications for the contracting package. 

This information is used by the construction personnel. It establishes the basis of payment, 

and details of the quality assurance program. Two contracting approaches are available in 

preparing the plans and specifications: 

• A method approach

• A performance approach

With the method approach, the designer outlines the work plan that the contractor is to 

follow. Information is provided on the deep dynamic compaction process, including the size 

of tamper, the drop height, the average energy to be applied, whether the energy must be 
4-48 

applied in single or multiple passes or phases, the need for a working mat, and all other facets 

of deep dynamic compaction that are necessary to achieve the proper improvement.

The method specification approach is more common as knowledge of deep dynamic 

compaction and the amount of energy to apply becomes better known among designers. The 

method specification also allows for general contractors to undertake some of the work. 

With the performance approach, the contractor is responsible for achieving the desired result. 

The designer in this case specifies the required depth and degree of improvement and also 

provides information regarding the site and soil conditions. The contractor is free to select 

the size of the tamper, the drop height, and other specifics of deep dynamic compaction to 

achieve the requirements set forth by the designer.

Only experienced contractors should be allowed to bid work under a performance approach. 

There are approximately five to eight specialty contractors in the United States who 

specialize in deep dynamic compaction. Since these contractors will do a significant amount 

of engineering and planning of the deep dynamic compaction operation, the cost for deep 

dynamic compaction using this approach is generally higher than for projects where method 

specifications are used.

2.4.1.1 Method Specification 

When preparing a method specification, the designer should include the following items in 

the plans and specifications:

• Tamper mass and size

• Drop height

• Grid spacing

• Applied energy

• Number of phases or passes

• Site preparation requirements

• Surface compaction after deep dynamic compaction

• Drawings of the working area

• Subsurface investigation data

• Test section design, if relevant
4-49 

The owner or designer is responsible for the following: 

• Monitoring during construction

• Borings and tests after deep dynamic compaction

The contractor is responsible for these items: 

• Providing adequate equipment to complete the work in a timely manner

• Ensuring the safety of personnel and equipment

• Providing a work plan subject to approval by designer

2.4.1.2 Performance Specification 

If a performance specification is prepared, the designer should specify the desired end 

product including these items: 

• Minimum soil property value to be achieved and the method of verification

• Maximum permissible settlement

• Other objectives of site improvement

• Minimum contract qualification requirements

The owner or designer also provides the initial subsurface data and the lateral extent of the 

project site. Some site monitoring is also required, as it is in the interest of the owner to be 

aware of the details of the field operations, including any changes in the planned scope. 

The contractor is required to meet the minimum specified final product and is responsible for 

the following: 

• Proper equipment and a work plan

• Meeting the project deadline

• Safety

• Field monitoring

• Additional subsurface exploration, as required to properly prepare a deep dynamic

compaction plan

• Verification of the end product
4-50 

2.4.2 Instrumentation Monitoring and Construction Control 

Regardless of whether deep dynamic compaction is performed under the method or the 

performance approach, it is essential to monitor the deep dynamic compaction operations. 

During deep dynamic compaction, close observation of the procedures should be undertaken, 

because it is frequently necessary to adjust the field program. Reasonable and practical 

adjustments can only be made if good monitoring information is available. Monitoring 

during deep dynamic compaction includes these procedures:

• Where deep dynamic compaction is planned adjacent to structures, vibration readings

should be taken to determine if ground vibrations pose a potential risk for the

buildings. Building damage has been correlated to peak particle velocities as

measured by seismograph, and there are charts available for determining permissible

levels of vibrations.

• Measuring crater depths and adjacent ground surface heave at occasional drop point

locations to determine the proper number of drops that can be applied at a location for

maximum efficiency. As an example, the first few drops generally result in significant

crater displacements without surrounding ground heave. After six or seven drops at

one point, the crater depths may become significantly less for each drop, and some

ground heave may occur. At some point, the additional crater volume produced by a

single drop may be equaled by ground heave adjacent to the drop point location,

which means that there is no longer any compaction occurring in the soil mass.

Because there is merely a ground displacement at this time, application of additional

energy would not produce densification.

• The average ground subsidence following deep dynamic compaction in an area

should be measured. Settlement readings should be obtained on a grid basis over an

area before and after deep dynamic compaction to determine the average induced

subsidence. Generally, the induced subsidence ranges from 5 to 10 percent, except in

landfills, where it could be higher.

• Field testing could be performed while the deep dynamic compaction is under way to

confirm that the desired improvements are being reached. This could include soil

borings with SPT, CPT, or PMT. At some sites, such as landfills, in situ soil testing is

often meaningless and load tests are frequently made with settlement readings

generally taken over a period of at least 1 week to measure performance. Field testing

is only an indirect indicator of the strength and compressibility of the deposits shortly

after deep dynamic compaction. The properties of some soils improve with time,

which must be kept in mind when evaluating the results.
4-51 

• Where there are fine-grained materials, such as silts or clayey silts, it may be helpful

to install piezometers in the deposits below the water table. The purpose of this

monitoring is to determine the magnitude and rate of dissipation in excess pore water

pressure during deep dynamic compaction. This would determine the waiting period

between drops applied at a location.

• In addition to the testing described above, qualified personnel should make general

observations of the deep dynamic compaction operations. Adjustments, in the field, of

the tamping program can be made depending upon these observations. For instance,

where ground subsidence is much greater in one area than in the remaining areas, this

could indicate an extremely loose pocket that requires additional tamping. Excessive

ground heave in other areas might indicate soils that will not properly compact, which

may require either a greater waiting period between densification or partial

undercutting and replacement with a soil that will densify under deep dynamic

compaction.

After the deep dynamic compaction is completed, conduct additional field explorations to 

confirm the degree and depth of improvement. This investigation is undertaken with soil 

borings along with SPT, CPT, or PMT. Because improvements in SPT, CPT, and PMT 

values have been observed 30 to 60 days following completion of deep dynamic 

compactions, it would be helpful if borings and field testing could be undertaken at that time. 

On some projects, settlement plates and inclinometers have been installed following deep 

dynamic compaction to monitor the movement of the subsoils during construction of the 

embankment or structure. While not necessary, this practice does provide useful information 

to evaluate the effectiveness of the deep dynamic compaction.

For projects completed under a method type of specification, the owner and designer are 

responsible for monitoring operations. This would include providing field personnel during 

the deep dynamic compaction to confirm the depth and degree of improvement. 

For projects completed under a performance type specification, the contractor is responsible 

for providing field monitoring and providing borings with SPT, CPT, or PMT after deep 

dynamic compaction to confirm the depth and degree of improvement. However, it is in the 

owner's interest to have field personnel on the site, because any changes to be made in the 

deep dynamic compaction plan would need to be mutually agreed upon between the owner 

and the contractor. The borings and field tests made after deep dynamic compaction should 

also be monitored to confirm that the improvement has been achieved. 
4-52 

2.5 Cost Information 

2.5.1 Cost Components

The cost of deep dynamic compaction will vary as a function of the depth of improvement 

required. A greater depth of improvement requires a heavier tamper and a higher drop height. 

Therefore, the equipment to undertake this work also requires a much heavier lifting crane. 

For tampers up to 40 kips, costs are relatively predictable because conventional lifting 

equipment can be used to raise and drop the tamper repeatedly. A comparison of costs for 

deep dynamic compaction versus the size of tamper are presented in Table 4-4. 

Table 4-4. Deep Dynamic Compaction Costs

Size of Tamper Required

kips

Unit Cost 

Dollars/ft2

4 to 8 1.10 to 1.50 

8 to 16 1.50 to 2.00 

16 to 20 2.00 to 3.00 

20 to 90 Negotiated for each job 

It should be noted that these costs include mobilization for large projects. On small projects, 

the mobilization costs should be added to the unit rates in the table. The unit costs do not 

include the cost of the quality assurance program or the cost of granular fill, if required, to 

fill craters or provide a surface stabilizing layer.

2.5.2 Cost Data

Costs for deep dynamic compaction vary considerably depending upon geographic location, 

type of deposit to be densified, and availability of local contractors. For lighter tampers (less 

than about 15 to 20 kips, many excavating and earth moving contractors have adequate 

equipment on hand to undertake this work. When the size of the tamper is in a range 20 to 40 

kips, heavier lifting equipment is required, so much of this work is done by specialty 

contractors.

When the tamper exceeds 35 to 40 kips, specialized equipment will be necessary. This

equipment could consist of a normal lifting crane modified with a large-diameter hoist for 

repeatedly raising and dropping the tamper. A large crane with a special hoist on the back 

for repeatedly raising and dropping a 64 kip tamper is shown in Figure 4-17. A specialized 

crane that has been developed for lifting tampers in the range of 64 to 100 kip is shown in 

Figure 4-18. The costs for deep dynamic compaction are much higher for these specialized 

pieces of equipment using the heavier tampers. 
4-53 

For planning purposes, the desired depth of improvement is selected, and then Equation 4-1 

is used to predict the proper energy. After selecting a drop height, the mass of the tamper is 

then determined. The cost may be estimated using Table 4-4. Additional information on 

estimating costs are presented in FHWA GEC 1 (1995) and GeoTechTools. 

In addition to the cost for deep dynamic compaction, there may be other factors:

• The construction of a working platform over soft ground

• Additional fill requirements to maintain the original grade

• Undercutting of weak deposits, such as zone 3 soils that don't respond to deep

dynamic compaction

• Pre-construction trial test sections, and/or pre- and post-construction monitoring and

testing

Though a granular blanket generally is not used on firm ground, on weak deposits, such as 

landfills, it may be necessary to import 1 to 3 feet of a crushed rock or granular material. This 

would add to the cost for deep dynamic compaction. If the ground subsidence is significant 

and the grade must be maintained, additional fill would be required to compensate for the 

induced settlement, which could be 5 to 10 percent of the thickness of the densified ground.

2.6 Case Histories 

Two case histories are presented to illustrate typical applications.

2.6.1 Highway Embankment Constructed in Mine Spoil

2.6.1.1 Project Description

Interstate 65 in Jefferson County, Alabama, was extended over a mine spoil area for 

approximately 2,500 feet. At either end of the mine spoil area, the final grades required 

approximately 15 feet of new fill, but in the center, there was a cut of 45 feet. From this final 

grade, there would still be 50 to 120 feet of mine spoil below the roadway. The new roadway 

was to be a four-lane highway with a width of 182 feet. 

2.6.1.2 Soil Conditions 

The mine spoil is classified as a mixture of rock fragments of shale, siltstone, and sandstone 

embedded within the soil matrix of silt and sand, or clayey silt and sand. Approximately 50 

percent of the mine spoil was larger than 2 inches. 
4-54 

The area was strip mined from 1977 to 1980. The initial subsurface investigations for the 

interstate occurred in 1983, and construction of the roadway started in 1984. Thus, the mine 

spoil was of relatively recent age at the time of construction.

Soil borings were made through the mine spoil using standard penetration tests (SPT). 

Excluding the very high SPT values where weathered chucks of rock were encountered, the 

majority of the mine spoil had SPT values ranging from 15-20 blows per 1 foot. There were 

some areas where the SPT values were as low as 5-10 blows per 1 foot and other areas as 

high as 25-35 blows per 1 foot. The variation in SPT value at depth is illustrated in Figure 4-

24. 

Figure 4-24. Increase in SPT values in a mine spoil after deep dynamic compaction.

Ground water was encountered at depths in excess of 100 feet below ground surface, and was 

not a factor on this project. 
4-55 

2.6.1.3 Design Concerns 

The Alabama Highway Department and its consultants realized that there were large 

variations in material classifications, strengths, and compressibilities in relatively short 

vertical and horizontal distances over the entire site. For this reason, site improvement by 

dynamic compaction was recommended in order to achieve the following: 

• Provide a more uniform subgrade for the new pavement 

• Crush existing large boulders and rock fragments that are subject to deterioration 

• Densify localized areas of loose/soft soil in the upper 25 to 30 feet, which could 

subside after highway construction 

Excavation of 25 feet of mine spoil followed by recompaction with conventional compactors 

was considered an alternate to deep dynamic compaction. However, deep dynamic 

compaction was found to be significantly less expensive. 

2.6.1.4 Predicted Densification Procedure

Using the guidelines presented in this chapter, the first step was to calculate the tamper mass 

and drop height for a desired depth of improvement of 25 to 30 feet (7.6 to 9 m). 

Using D = 30 feet (9 m), n = 0.5 as average, and Equation 4-1, 

• 9 m = 0.5(WH)1/2

• WH = 324 

From Figure 4-22, using W = 20 Mg (44 kip), H ranges from 15 to 31 m (49 to 103 feet); use 

H = 16.5 m (54 feet), since WH = 20 x 16.5 = 330 > 324 Mg-m.

The second step was to calculate the energy to apply. Using Table 4-3 as a guide, for Zone 2 

type deposits 

• E = 250 to 350 kJ/m3 (5,200 to 7,200 ft-lbf/ft2). 

Because the deposits are already in a medium-dense condition, use E = 250 kJ/m3 (5,200 ftlb/ft2

). For a 9 m (30 feet) thick deposit, applied energy = 9 m x 250 kJ/m3 = 2250 kJ/m2. 

The third step was to determine the grid spacing and number of drops. Assuming a 20 Mg 

(44 kip) tamper dropped 16.5 m (54 feet), a reasonable spacing would be 3 m (10 feet), as the 

tamper diameter is likely to be 2 m (6.5 feet). Using Equation 4-2, and assuming only one 

pass and an ironing pass, which will apply 250 kJ/m2

 energy, the number of drops can be 

computed. 
4-56 

2

2

)3(

)()5.16()/10()20( /)2502250(

m

Mg kN NmMg − kJ m =

For the grid spacing of 3 m x 3 m, N = 5.5; or six drops at each grid point. 

After dynamic compaction, the maximum SPT value according to Table 4-2 would be 25 to 

35 for sandy silts, and 20 to 35 for clayey silts. Estimated induced settlement is 5 percent (9 

m) = 0.45 m (1.5 feet). 

2.6.1.4 Actual Densification Procedure 

Based upon results of a test section, it was determined that densification to a depth of 7.6 to 9 

m (25 to 30 feet) could be achieved using a 20-Mg (44-kip) tamper with a drop height of 19 

m (62 feet). The high-level energy was applied using five drops at each grid point location, 

with a spacing of 3.0 m (10 feet ) between grid points. 

After the high-level energy was applied, the ground surface was leveled and an ironing pass 

completed using the same tamper with a drop height of 5.8 m (19 feet ), a grid spacing of 1.8 

m (6 feet ), and one drop per grid point location. 

This procedure resulted in an average applied unit energy of 2.1 MJ/m2 (142 ft-kip/ft2

) for the 

primary energy application and an additional 0.36 MJ/m2 (24 ft-kip/ft2

) during the ironing 

pass which is approximately the same energy previously calculated using Table 4-3. 

2.6.1.5 Ground Improvement

In Figure 4-25, soil borings with SPT values made after dynamic compaction are compared 

to the borings with SPT values before dynamic compaction. This data indicates the 

improvements were obtained to depths of approximately 10.7 m (35 feet) and that the SPT 

values increased significantly.
4-57 

Figure 4-25. Induced settlement following deep dynamic compaction at the mine spoil 

project in Alabama.

Another indication of ground improvement was the amount of induced ground settlement by 

dynamic compaction. Within the test sections, ground elevations were taken on a grid pattern 

and measured following various levels of energy application. The data are summarized in 

Figure 4-26. 
4-58 

Figure 4-26. Statistical variation in crater depths at the mine spoil site in Alabama. 

At test pad 1, the average ground subsidence following full application of the primary energy 

was approximately 0.6 m (2 feet). At test pad 2, approximately 0.6 m (2 feet) of induced 

settlement was observed after energy application, corresponding to about 90 percent of the 

prescribed energy.

During the production phase of dynamic compaction, the average induced ground settlement 

was 0.5 m (1.6 feet), although it could have been more because some fill was brought in 

during the leveling of the craters. In local areas, the average induced settlement was 

significantly higher. The variation in the crater depths that were observed in certain sections 

of the dynamically compacted area are illustrated in Figure 24. The normal average crater 

depth for the mine spoil was 1 to 1.1 m (3.3 to 3.6 feet), but crater depth as high as 1.5 to 2.7 

m (5 to 9 feet) occurred in some locations, indicating a soft or void area. Additional highenergy tamping was undertaken in the soft area after ground leveling and placement of fill to 

raise the grade.

2.6.1.6 Contracting Procedure and Cost 

A method specification was prepared for this project, and non-specialty, as well as specialty, 

contractors were allowed to bid. The project was awarded to an excavating contractor. After 

an initial 2-week trial period with some experimentation, the work proceeded on a reasonably 

good schedule. One hundred working days were required to dynamically compact 

approximately 37,200 m2 (44,500 yd2

) for an average of 372 m2 (445 yd2) per day. When 
4-59 

considering only production time, the average tamping rate was 63 impacts per hour for the 

primary phase and 67 impacts per hour for the ironing pass. 

The project was bid on a price per drop that included an overall mobilization charge for all 

facets of the embankment construction. Therefore, the portion of the mobilization attributed 

to dynamic compaction is difficult to determine. The bid price per drop was $2.90 for the 

high-energy phase and $2.65 per drop for the ironing pass. Using the prices given and an 

estimate for the mobilization, the cost for the dynamic compaction was approximately $7.20 

per m2

 ($6.00 per yd2) of area treated. This amounted to about 4 percent of the total project 

cost.

The estimated cost for excavation of the upper 6.1 m (20 feet) of soil followed by placement 

in lifts and compaction would have been approximately 2.6 times the cost of the dynamic 

compaction. 

2.6.2 Highway Embankment on Landfill Debris

2.6.2.1 Project Description

The Route 7 bypass around Manchester, Vermont, crosses two areas underlain by old refuse. 

The southern area, designated as area 1, is approximately 61 m (200 feet) long, and the 

planned embankment extended to a height of approximately 7 m (23 feet) above present 

grade. The northerly area, designated as area 2, is approximately 91 m (300 feet) long, and 

the new planned embankment extended to heights of 3 to 3.7 m (10 to 12 feet) above present 

grade.

2.6.2.2 Subsurface Conditions 

Both landfills were covered with about 0.6 m (2 feet) of gravelly glacial till that was used as 

a cover material. Below this level, old landfill material was present to depths ranging from 1 

to 3.4 m (3.3 to 11.2 feet), but averaging about 2 m (6.6 feet) in area 1. The landfill was 

described as consisting of miscellaneous materials, including metals, plastic, bags, glass, and

trash. No paper, food, or other biodegradable materials were encountered within the landfill. 

Occasional seams or layers of silty sand were encountered within the trash, but these were 

probably thin layers of daily cover. Standard penetration tests ranged from 10 to 14 blows for 

300 mm (1 foot), with some values as low as 7. 

At area 2, the thickness of the trash was approximately 1 to 6 m (3.3 to 20 feet) and averaged 

3 m (10 feet). The trash consisted of the same classification as area 1. The water table was 

determined to be at a depth of 5.5 m (18 feet) in landfill area 2. 
4-60 

Both landfills were underlain by a medium-dense to dense silty sand and gravel containing 

boulders and cobbles. 

The age of the landfills at the time of dynamic compaction was determined to be 

approximately 14-18 years after closure. Ordinarily, this would mean the landfill was in the 

middle age of decomposition. However, the absence of organic materials within the trash 

indicated that it was in an older stage of decomposition. Because no methane gas was noted 

on the boring logs, it was likely that the decomposition of the highly organic materials was 

complete.

2.6.2.3 Design Concerns 

When landfills decompose, a relatively loose structure is all that remains, creating the 

potential for significant total and differential settlement. For this reason, some method of 

ground improvement was necessary, and dynamic compaction was selected to reduce the 

potential for this predicted movement. 

2.6.2.4 Predicted Densification Procedure

Using the guidelines in this chapter, the first step was to calculate the tamper mass and drop 

height for a desired depth of improvement ranging from 3.4 m (11.2 feet) maximum in area 1 

and 6 m (20 feet) maximum in area 2. 

For area 2:

Using Equation 4-1, with n = 0.4 for a landfill, and D = 6 m (20 feet), 

• 6 m = 0.4 (WH)1/2

• WH = 225 Mgm 

From Figure 4-22, the desired energy can be obtained as the product of tampers ranging from 

about 13 to 16 Mg multiplied by drop heights ranging from 14 to 30 m. 

For a 14-Mg tamper, use:

m

Mg

mMg H 16

14

225 = =

For area 1:

Using Equation 4-1, with n = 0.4, D = 3.4 m, and W = 14 Mg, 
4-61 

m

Mg

H 2.5

14

)4.0/4.3(

2

= =

The second step was calculating how much energy to apply. Using Table 4-3 as a guide, E = 

600-1100 kJ/m3 for a landfill. Because the SPT values indicated a medium dense condition, E 

= 800 kJ/m3 was selected.

Using the average fill thickness, the required average applied energy can be calculated: 

• For area 2, E = (3 m) x (800 kJ/m3

) = 2400 kJ/m2 = 2.4 MJ/m2

• For area 1, E = (2 m) x (800 kJ/m3) = 1600 kJ/m2 = 1.6 MJ/m2

The third step was determining the grid spacing and number of drops. Assume all the energy 

can be applied in one pass. For a 14-Mg tamper, the diameter is typically 1.6 m (5.2 feet), 

suggesting a grid spacing of 2.3 m (7.5 feet). Because the highway department planned on 

using a surface compactor following dynamic compaction, the ironing pass was eliminated. 

The number of drops can now be calculated using Equation 4-2. 

For area 2, a grid spacing of 2.3 m, and 1 pass 

2

2

)3.2(

)1)()(16)(/10)(14( /2400

m

Mg kN NmMg kJ m =

where N = 5.66 drops or 6 drop per grid point. 

For area 1, a grid spacing of 2.3 m, and 1 pass 

2

2

)3.2(

)1)()(2.5)(/10)(14( /1600 Mg kN NmMg kJ m =

where N = 11.6 or 12 drops per grid point. 

To reduce the number of drops, the drop height could be increased because the equipment 

provided for area 2 will have the capacity to lift the tamper to 16 m. To have the same 

number of drops (6) as for area 2, use this equation to calculate drop height: 

2

2

)3.2(

)1)(6)()(/10)(14( /1600 Mg kN dropsHMg kJ m =

where H = 10 m 
4-62 

After dynamic compaction, the maximum SPT value according to Table 4-2 would be 

anticipated to be 15 to 40. 

The estimated induced settlement would be (20%)(3 m) = 0.6 m (1 feet) in area 2, and 

(20%)(2 m) = 0.4 m (1.3 feet) in area 1, based on experience with landfills. 

2.6.2.5 Actual Densification Procedure

Before starting dynamic compaction, the site was leveled by lowering the elevation in the 

high portion of the site, then placing some of the debris in the lower portion of the site. 

Because the debris was variable, a 0.6-m (2 feet) blanket of silty sandy gravel was placed on 

the surface as a working mat.

The specifications required a 13.6-Mg (30 kip) (minimum) tamper, an 18-m (60 feet) drop in 

the shallow fill area, and a 27-m (90 feet) drop height in the deeper fill area. The contractor 

used a 14-Mg (31 kip) tamper with 18 and 27 m (60 and 90 feet) drop heights and elected to 

apply the energy in three phases. The first phase consisted of dynamic compaction on a grid 

basis with a spacing of 4.6 m (15.4 feet) between drop points. The second phase consisted of 

the same grid pattern, offset from the first by 2.3 m (7.5 feet) so as to be situated between the 

phase 1 points. The third phase consisted of energy applied at the phase 1 drop point 

locations. Seven drops were applied at each drop point location. This resulted in an average 

energy at the ground surface of 2.5 MJ/m2

 (169 ft-kip/ft2) for area 1 and 3.75 MJ/m2 (254 ftkip/ft2

) for area 2, which is considerably more energy than required by the calculations, based 

on the presented guidelines. However, the specifications required a drop height much greater 

than that required by Equation 4-2, thereby resulting in more energy being applied. 

Crater depths were monitored during dynamic compaction. In the first phase, the crater 

depths were typically 1 m (3.3 feet). In the third phase, which took place at the same location 

as the first phase, the crater depths were 0.5 m (1.6 feet). Heave measurements were taken 

adjacent to the drop point locations, and heave was not observed. 

2.6.2.6 Ground Improvement

No soil borings were made after dynamic compaction. The initial plan was to install 

settlement plates along the completed sections of the roadway, but this was not undertaken. 

Discussions with the highway engineer indicate that the pavement sections have performed 

well in this area and there is no evidence of settlement.
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2.6.2.7 Contracting Procedure and Cost 

A method specification was prepared by the agency for this project. The specification 

included the tamper weight, drop heights ranging from 18 m (59 feet) in the shallow fill areas 

to 27 m (88.6 feet) in the deep fill areas, plus the number of phases of energy application and 

the spacing between the drop point locations. The number of drops at each location was 

specified to range from a minimum of 6 to a maximum of 10. 

A specialty contractor was awarded this project and was able to demonstrate that 7 drops per 

phase were sufficient to achieve satisfactory densification. The cost for dynamic compaction 

was $10.25 per m2 ($8.57 per yd2). This cost does not include the placement of the 0.6 m (2 

feet) gravel blanket used as a working mat.
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3.0 VIBRO-COMPACTION

3.1 Feasibility Considerations

Vibro-compaction can be used to achieve a number of design objectives, as discussed in 

Chapter 1. This section discusses applications for transportation facilities, as well as 

advantages, disadvantages, limitations of the system, and feasibility.

3.1.1 Applications

This section focuses on the use of vibro-compaction as a solution to problems related to 

transportation projects. Thousands of vibro-compaction projects have been completed in the 

United States, with about 10 percent being transportation related. 

For transportation projects, vibro-compaction can be used to treat problems related to the 

following: 

• Foundation soils beneath proposed structures

• Highway embankment fills

• Tunnels - compaction of overburden soils

• Densification of artificial tunnel islands

• Mitigation of liquefaction potential for transportation applications:

o Compaction to stabilize pile foundations driven through loose granular

materials

o Densification for abutments, piers, and approach embankment foundations

• Compaction of underwater embankment fills

• Compaction in areas of potential cavities beneath embankments to pre-settle and fill

such voids prior to construction of a structure

3.1.2 Advantages and Disadvantages 

3.1.2.1 Advantages 

As an alternative to deep foundations, vibro-compaction is usually more economical and 

often results in significant time savings. Loads can be spread from the footing elevation, thus 

minimizing problems from lower, weak layers. Densifying the soils with vibro-compaction 

can considerably reduce the risk of seismically induced liquefaction. Vibro-compaction is a 

cost-effective alternative to removal and replacement of poor load-bearing soils. The use of 
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vibro-compaction allows maximum improvement of granular soils to a depth of about 165 

feet, with generally recommended depth of about 100 feet. The vibro-compaction method is 

effective both above and below the natural groundwater level. 

3.1.2.2 Disadvantages

The primary disadvantage of vibro-compaction is that it is effective only in granular, 

cohesionless soils. The realignment of the sand grains and, therefore, proper densification 

generally cannot be achieved when the granular soil contains more than 12 to 15 percent silt 

OR more than 2 percent clay. The maximum depth of 165 feet may be considered a 

disadvantage, but there are very few construction projects that will require densification to a 

greater depth.

Like all ground modification techniques, a thorough soil investigation program is required. A 

more detailed soils analysis may be required for vibro-compaction than for a deep foundation 

project. This is because the vibro-compaction process utilizes the native soil to the full depth

of treatment to achieve the end result. A comprehensive understanding of the total soil profile 

is therefore necessary. A vibro-compaction investigation will require continuous standard 

penetration tests (SPT), and/or cone penetrometer (CPT), as well as gradation tests to verify 

that the soils are suitable for vibro-compaction.

3.1.3 Feasibility Evaluations

3.1.3.1 Geotechnical (In Situ Soil Gradations) 

Vibratory compaction of soils is most effective on granular materials having little to no fines 

or low cohesion or plasticity. A quick assessment of the suitability of granular soils for 

treatment by vibro-compaction was proposed by Degen (1997) on the basis of the Unified 

Soil Classification System (USCS) and is shown in Table 4-5. 
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Table 4-5. Suitability Assessment of Granular Soils for Vibro-compaction

Soil Type USCS Comments on Suitability for VC

Gravel, well graded GW Well suited for VC, potential penetration difficulties 

with less powerful machines 

Gravel, poorly graded GP If D60/D10 ≤ 2 compaction only marginal (trail 

compaction recommended) 

Gravel, silty or clayey GM, GC Compaction not possible if clay content >2% and silt 

content >10% 

Sand, well graded SW Ideally suited

Sand, poorly graded SP If D60/D10 ≤ 2 compaction only marginal (trail 

compaction recommended) 

Sand, silty SM Compaction inhibited if silt content >8%

Sand, clayey SC Compaction inhibited if clay content >2% 

Sources: Kirsch and Kirsch 2010 and Degen 1997

The suitability of a soil for vibro-compaction methods has generally been determined on the 

basis of grain size distribution, as shown in Figure 4-27. 

Figure 4-27. Range of soil types treated by vibro-compaction.
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Soils with grain size distribution curves lying entirely within zone B are ideally suited for 

vibro-compaction with fines content below 10 percent. Soils in zone A are well compactible 

but the increasing gravel content are resultant high permeability may affect the ability of the 

depth vibrator to penetrate to desired depths. If the grain-size distribution curve falls in zone 

C, it is advisable to backfill with gravel in lieu of sand during the compaction process. This 

will improve the contact between the vibrator and the treated soil and drastically increase 

compaction time. The soils with grain size distribution curves partly or entirely in zone D are 

not readily compactable by vibro-compaction. However, these soils can be improved by 

vibro-replacement, as described in Chapter 5 Aggregate Columns. 

An alternative way to assess suitability for vibro-compaction is the use of cone penetration 

tests. An advantage of using cone penetration test is that continuous readings of cone tip 

resistance and sleeve friction can be obtained as compared to spot samples collected for 

grain-size analyses. The suitability classification proposed by Massarsch (1991) is shown in 

Figure 4-28. 

After Massarsch 1991

Figure 4-28. Soil compactibility based on cone penetration resistance and friction ratio. 

For cohesionless soils with natural dry densities less than their maximum dry densities, the 

influence of vibrations will result in a rearrangement of their grain structures. Under the 

influence of induced vibrations, the inter-granular forces between the grains in non-cohesive 

soils are temporarily nullified. The grains are then rearranged, unconstrained, and unstressed 

under the action of gravity to a more dense state. The void ratio and compressibility of the 
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soil treated by vibratory means will be decreased, and the angle of shearing resistance 

increased. The treated, compacted soil is capable of sustaining higher bearing pressures for 

the same settlements as the untreated soil, and undergoes smaller settlement for the same 

bearing pressure, with the settlement generally being only elastic. The achievable reduction 

in void ratio depends on grain shape, soil composition, and vibration intensity (Moseley and 

Priebe 1993). 

By advancing the vibrator to the desired level and withdrawing it from the ground in a 

specific manner, the granular soils are compacted by the horizontal vibration forces. A 

compact soil cylinder is thus formed, with the diameter determined by the grain size 

distribution, the soil density, and the vibrator characteristics. By arranging compaction points 

in suitable patterns, soil masses can be compacted homogeneously. 

The increased density of the granular soils results in the downward movement of the soil 

around the vibrator and creates a cone-shaped depression at the surface. This depression must 

be continuously in-filled with granular fill material. If on-site material is used, then the 

original ground surface will be lowered. Alternatively, ground level can be maintained by 

adding imported granular fill material, which is compacted simultaneously with the natural 

soil. 

The vibro-compaction process subjects the soil mass to high accelerations during 

compaction. These levels of dynamic strain are unlikely to be repeated, even under 

earthquake loading. Provided that the design earthquake criteria are not exceeded during a 

seismic event, the treated ground can be expected to perform as designed. 

Soil compaction, as achieved in the vibro-compaction process through the rearrangement of 

soil particles, is not possible in cohesive, fine-grained soils. The cohesion between the 

particles prevents rearrangement and compaction from occurring. 

3.1.3.2 Environmental Considerations 

The dry method of vibro-compaction is only viable in clean, sandy, fully saturated soils. The 

great majority of vibro-compaction projects are therefore accomplished by the wet method. 

Although the vibro-compaction technique is used for densifying primarily granular soils, the 

jetting water effluent will nevertheless require temporary construction provisions to contain 

and dispose of any silt and clay in the effluent. With current awareness of potential 

environmental problems, geotechnical exploration programs should include not only the 

classification of the soil type and location of groundwater, but also the examination and 

classification of any potential contaminants in the soil and groundwater. If contaminants are 

uncovered in the original exploration program, a determination should be made as to whether 
4-69 

they can be treated at the site during the vibro-compaction program. If this is not the case, 

then an alternative densification program, such as the dry bottom feed stone column 

technique (which does not produce jetting water effluent), or other solutions, should be 

considered. 

3.2 Construction and Materials

The vibro-compaction process uses crane-mounted depth vibrators and appropriate backfill 

material. This section discusses construction equipment and the suitability of backfill 

material.

3.2.1 Construction 

The equipment used to achieve the necessary densification are high-powered, probe-type 

vibrators ranging from 12 to 16 inches in diameter and 10 to 15 feet in length, as shown in 

Figure 4-29 and Table 4-6. 

Courtesy of Hayward Baker

Figure 4-29. High-powered, probe-type vibrator utilized in vibro-compaction. 
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Table 4-6. Specifications of Several Vibrators

Vibrator 

Length 

m

Dia. 

mm

Weight 

kg

Motor 

kW

Speed 

rpm

Ampl. 

mm

Dynamic 

Force 

kN

Bauer TR13 3.1 300 1000 105 3250 6 150 

Bauer TR85 4.2 420 2090 210 1800 22 330 

Keller M 3.3 290 1600 50 3000 7.2 150 

Keller S 3.0 400 2450 120 1800 18 280 

Keller A 4.3 290 1900 50 2000 13.8 160 

Keller L 3.1 320 1815 100 3600 5.3 201 

Vibro V23 3.6 350 2200 130 1800 23 300 

Vibro V32 3.6 350 2200 130 1800 32 450 

Source: Layne Christiansen Company

Note: See table at front of manual for SI conversions.

A set of rotating eccentric weights housed inside the probe is mounted on a vertical shaft. 

Vibrations (induced by rotating these weights) are produced close to the bottom of the unit. A 

motor located within the casing, as shown in Figure 4-30, drives the rotating shaft. 

Figure 4-30. Cross-section of a typical vibrator.

To drive the assembly shown in Figure 4-30, an electrically driven motor is usually 

employed, driven by motors typically in the 100-130 kW range. The vibrations produced by 

these units are generated at the nose of the unit and, as a result of the rotation of the weights, 
4-71 

emanate radially in the horizontal plane away from the unit. The units now in general use 

generate dynamic forces from 33,750 to 100,000 lbs (150 to 450 kN) at frequencies ranging 

from 1,800-3,200 rpm. However, for vibro-compaction, vibrators operating at lower 

frequencies will usually produce better densification results than those operating at higher 

frequencies. This is because low frequency vibrators usually have a higher amplitude, which 

translates into a greater compactive effort. Additionally, the natural frequency of most 

densifiable soils is closer to 1500 rpm than to 3000 rpm. Selected available vibrators and 

some of their operating characteristics are described in Table 4-6. 

Follower tubes of a similar or lesser diameter are attached to the vibrating unit in order to 

extend its length to allow treatment of soils at depth. The follower tubes are attached to the 

vibratory unit by means of an isolation coupling, thus preventing the vibrations from 

traveling up the follower tubes, negating the problem of energy losses at depth. 

The complete assembly is supported from a standard crane (Figure 4-31), a specially built 

hydraulic crawler crane, or a crane that is mounted on a barge (Figure 4-32), depending upon 

the site conditions. 

Figure 4-31. Vibrator suspended from a conventional crane. 
4-72 

Figure 4-32. Vibrator suspended from a barge-mounted crane.

The vibro-compaction operation necessitates the use of water- or air-jetting to facilitate the 

penetration of the vibrator and to densify the soil. Therefore, water- or air-feed hoses, as well 

as water or air pumps, are also required.

3.2.2 Materials

In order to transmit vibrations from the vibrator into the in situ soil and achieve adequate 

compaction, it is necessary to supply sufficient backfill material to fill the void created by the 

densification process. Fine sands, coarse sands, rounded gravel, crushed stone, recycled 

aggregate, and slag have all been used as backfill material. Slag has the advantage of being 

inexpensive in some locations, but does not settle as fast as other material with comparable 

gradation. Coarse materials with little or no fines make the best backfill. However, if the 

particle size becomes too large, the gravel will arch in the annular space between the follower 

tube and the void, preventing backfill from reaching the vibrating tip. 

The suitability of the backfill appears to be a function of the backfill quantity that can 

accumulate around the vibrating tip in a fixed period of time. The backfill gradation is the 

most significant factor controlling the rate at which the backfill settles through the wash 

water and accumulates around the tip. A rating system has been developed to judge the 

suitability of backfill material for vibro-compaction, based on the settling rate of the backfill 

in water and project experience (Brown 1977). This rating is dependent on a “suitability 

number” and is a function of the grain size diameters of the backfill material. The equation 

used in this calculation is as follows:
4-73 

 

[Eq. 4-3] 

where D50, D20, and D10 are the grain size diameters, in millimeters, at 50 percent, 20 percent, 

and 10 percent passing. The qualitative categories of backfill using utilizing this rating 

system are listed in Table 4-7. 

Table 4-7. Backfill Evaluation Criteria

Suitability 

Number Rating

0 to 10 Excellent

10 to 20 Good 

20 to 30 Fair

30 to 40 Poor 

>50 Unsuitable 

Source: Brown 1977 

The quality of backfill material affects the allowable withdrawal rate of the vibrator. Within 

reasonable limits, the lower the suitability number, the faster the backfill will settle, and the 

faster the vibrator can be withdrawn and still achieve acceptable compaction. Backfill 

normally consists of material graded as sand, or sand and gravel, with less than 10 percent by 

weight passing the #200 sieve, and containing no clay. 

3.3 Design 

Similar to other ground modification methods, design of a vibro-compaction program 

requires definition of the problem, identification of all possible solutions, and the 

development of performance requirements for the improved soil. Depending on the type of 

project being designed, the prime consideration could be total or differential settlement, 

bearing capacity, or a seismic/liquefaction resistance requirement.

3.3.1 Design Considerations 

If loose granular soils are identified as the site problem, then densification by vibrocompaction will be a potential technical solution, especially if the loose deposit is deeper 

than 35 feet as measured from the surface. The relationship between penetration resistance 

from subsurface investigations and soil properties that are useful in assessing the current 

density and the feasibility of improvement, as well as in identifying the potential targets and 

performance requirements for the improved soil, are indicated in Table 4-8. 
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Table 4-8. Penetration Resistance and Sand Properties

Penetration 

Resistance

Very 

Loose Loose 

Medium 

Dense Dense

Very

Dense

SPT N-value 

(blows/foot)* < 4 4 to 10 10 to 30 30 to 50 > 50 

CPT cone resistance

(kg/cm2 or tsf) < 50 50 to 100 100 to 150 150 to 200 > 200 

Equivalent Relative

Density (%)* < 15 15 to 35 35 to 65 65 to 85 85 to 100 

Dry Unit Weight (pcf) < 90 90 to 100 100 to 115 115 to 130 > 130 

Friction Angle, degrees < 30 30 to 3235 35 to 40 40 to 45 > 45 

Cyclic Stress Ratio

Causing Liquefaction** < 0.04 > 0.04 to0.12 

> 0.12 to 

0.33 

> 0.33 to 

0.40 

Shear Wave Velocity

(ft/s)*** < 400 400 to 525 525 to 650 650 to 740 > 750 

* Normally consolidated sand

** Seed et al. 1983

*** Debats and Sims 1997

If vibro-compaction is selected as the improvement method, the following parameters must 

be determined:

• Gradation of the in situ soils, including silt and clay content

• Existing relative density, or looseness, of the in situ soils

• Required density improvement necessary to solve the project's requirements and,

once determined, whether this improvement is feasible

3.3.2 Design Procedure 

The significant engineering properties of a granular soil – compressibility, shear resistance, 

permeability, resistance to dynamic loading – are largely dependent on the state of 

compaction, typically expressed in terms of relative density for clean granular materials. The 

term “relative density,” or Dr, is defined as follows: 

 

[Eq. 4-4] 

where, 

γd = dry unit weight of the soil in its natural state

γd(min) = dry unit weight of the soil in its loosest state
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γd(max) = dry unit weight of the soil in its densest state

In this calculation, γd(min) and γd(max) should be determined in accordance with current ASTM 

procedures [ASTM D-2049]. 

High relative density corresponds to high bearing resistance with low settlement. For seismic 

loading, resistance to liquefaction in granular soil is a function of relative density. In 

earth-retaining problems, active pressure decreases and passive resistance increases, as 

relative density increases. With vibro-compaction, the angle of internal friction is increased 

on average between 5 and 10 degrees, resulting in much higher shear resistance. The stiffness 

of the improved soils is increased, and consequently settlements are greatly reduced.

In addition to soil gradation, the area influenced (the tributary area) by each compaction 

point for a specified relative density depends on the compaction method used and the specific 

characteristics of the vibrator, which may not be known in advance. As shown in Table 4-6, 

vibrator characteristics vary widely.

The approximate relationships between relative density, soil type, and treatment area for a 

specific vibrator are shown in Figure 4-33. It is unlikely, even for heavy loading, that it will 

be necessary to achieve a relative density above 85 percent.
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Figure 4-33. Approximate variation of relative density with tributary area. 

A chart useful in estimating the probable level of improvement that can be obtained by vibrocompaction is shown in Figure 4-34. It is based on the lower bound soil gradation (silty sand) 

indicated in Figure 4-33. Similar charts can be developed for coarser granular soils from 

Figure 4-33. 
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Figure 4-34. Relative density versus probe spacing for silty sands. 

The project designer is responsible for setting the requirements for the project with an 

appropriate safety factor and the best method of confirmation testing. For most vibrocompaction projects, the following performance criteria should be considered: 

• 60 percent relative density for floor slabs, flat bottom tanks, embankments 

• 70 to 75 percent relative density for column footings, bridge footings 

• 80 percent relative density for machinery and mat foundations 

3.3.2.1 Probe Spacing and Patterns

A typical vibro-compaction program is designed with various probe spacing and patterns. 

The distance between compaction points is critical, as the density generally decreases as the 

distance from the probe increases. Stronger vibroprobes allow for wider spacing under the 

same soil conditions. 

The area compaction point pattern also affects the densification. An equilateral triangular 

pattern is primarily used to compact large areas, since it is the most efficient pattern. The use 

of a square pattern instead of an equilateral triangular pattern requires 5 to 8 percent more 

points to achieve the same minimum densities in large areas.

Given the in situ soil gradation and relative density required, the spacing of compaction 

points can be determined. Typical area patterns and spacing for 80 percent relative density 

requirements are illustrated in Figures 4-35 and 4-36. The spacing of the vibro-compaction 

points would be wider for a lower relative density requirement. 
4-78 

Figure 4-35. Typical compaction point spacing for area layouts. 

Figure 4-36. Typical compaction point layouts for column footings. 

3.3.2.2 Performance Requirements

Performance requirements, such as total or differential settlement, bearing resistance or 

reduced liquefaction potential, can all be related to a desired in situ relative density. After 
4-79 

compaction is completed, the effectiveness of compaction is normally evaluated to verify 

contractual compliance and to verify that the compacted soils will perform adequately under 

the intended loads or seismic event. A number of testing methods have been used, including 

Standard Penetration Testing, cone penetrometer soundings, shear wave velocity, or load 

tests. These methods, and their advantages and disadvantages, are described in Chapter 5. 

The potential for liquefaction can be evaluated using SPT blow counts and the cyclic shear 

stress ratio (CSR) at any depth.

3.4 Construction Specifications and Quality Assurance

3.4.1 Contracting Procedures

Vibro-compaction may be performed under either a method-type specification or a 

performance-type specification. Under a method-type specification, the specifying agency 

details a specific procedure and pattern spacing to achieve the required improvement. Bids 

are invited from contractors suitably equipped to perform the work. With this type of 

specification, the specifying agency assumes the risk, and a full knowledge of the ground 

improvement technology and equipment is required. If this knowledge is not available within 

the specifying agency, a method type specification is not advised. 

Under a performance-type specification, the required end result is specified and the 

contractor assumes responsibility for achieving it. This approach does not require in-depth 

knowledge within the specifying agency. The contractor has the flexibility of selecting the 

procedure and pattern spacing to meet the design criteria.

Specifications and contracting procedures for vibro-compaction have changed significantly 

over the years. Where once the specifications stated a specific procedure and pattern spacing, 

variances in equipment and methods today favor placing the responsibility for achieving the 

required improvement on the contractor. Whereas the vibro-compaction method itself may 

still be specified, the contractor adopts the procedure and pattern spacing to achieve project 

objectives. Most vibro-compaction specifications today are performance based. 

This section discusses contracting methods and quality control and inspection procedures. A 

guide to the preparation of a typical specification is included. Since the responsibility for 

achieving the design criteria for the ground improvement usually rests with the contractor 

under a performance specification, the focus of this chapter reflects this norm. 

Most vibro-compaction projects require a certain degree of densification, which can be 

specified as follows:

• Minimum or average percent relative density
4-80 

• Minimum or average (spt) blow count

• Minimum or average cone penetration resistance

• Minimum or average size of gravel or sand column

• Minimum amount of backfill material added

• Minimum load bearing requirement

All of the above have been used in past projects and, for the most part, have been successful. 

However, the best specification is one that allows for some variance of results within 

specified limits. Also, past experience has shown Standard Penetration Test blow counts to 

be misleading in certain stratifications and that specific percentage degrees of density are 

difficult to measure. 

The technical literature has shown evidence that verification testing procedures can give 

misleadingly low results if performed immediately after densification, and that results can 

increase significantly with time (Mitchell and Solymar 1984; Debats and Sims 1997). A 

minimum wait of 5 days is recommended before performing verification testing, but a wait of 

about 10 days is preferable. The effectiveness of soil improvement with time after treatment 

should be considered in performing tests and interpreting test results. 

A guide performance specification can be found in GeoTechTools. The format of the guide 

specification is deliberately generic. The responsible party should be inserted as appropriate 

when developing specifications for a particular project. Italicized terms or descriptions allow 

for flexibility to adapt to the specific requirements of the project to be improved by vibrocompaction. Where necessary, additional explanatory notes are included. 

3.4.2 Quality Assurance and Monitoring

The quality assurance plan and inspection activities are developed well in advance of the 

vibro-compaction work. The duties of the Contractor and the Owner/Engineer with respect to 

QA/QC are dependent on the type of specification under which the work is being 

accomplished.

Under a method specification, development of the QA/QC plan, review of plans and 

specifications, and acceptance of backfill material are the responsibility of the Owner's 

Engineer. During performance of the work, the Owner's Engineer is responsible for all onsite inspection and testing. (Under a performance specification, these latter responsibilities lie 

with the Contractor). The Contractor provides the Owner's Engineer with testing results to 

verify that improvement criteria are being met. Under a performance specification, the 
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Owner is generally obligated to perform some independent verification testing based on the 

results supplied by the Contractor. 

Under either type of specification, the inspection process during the actual compaction 

process should include the following:

1. Verification that probes penetration depth is acceptable

2. Verification that probe withdrawal rate is acceptable

3. Monitoring the probe penetration rate to obtain a rough indication of the type and 

density of soil penetrated 

4. Verification that compaction points are at the proper locations 

5. Monitoring the volume of backfill added to obtain an indication of the densities 

achieved

6. Verification that backfill gradation is acceptable

7. Monitoring of ammeter or hydraulic pressure readings to verify that the build-up is 

sufficient

8. Verification that the probes are operating at appropriate speeds

9. Verification that induced vibrations are not excessive when operating close to 

existing structures

Examination of the data logger records are a valuable QC tool and can be used to monitor 

uniformity of the compactive effort with depth. The length of time spent at each stage of 

compaction (1.5 to 2.0 feet) depends on the soil reaction and is shown on the right side of 

Figure 4-37, which measures from the bottom up. Generally the finer the soil, the longer the 

time required to achieve the same degree of compaction. The electrical current drawn by the 

motor increases as the soil around the vibroprobe densifies, as shown on the left side of 

Figure 4-37. 
4-82 

Debats and Sims 1997

Figure 4-37. Typical data logger results: amps versus depth (left) and time versus depth 

(right). 

During the compaction process, the adequacy of compaction is periodically verified for 

quality control and acceptance purposes. These checks verify contractual compliance and 

compacted-soils performance under the intended loadings. A number of methods are used, 

including borings with SPTs, static CPTs, measurement of the surface subsidence, density 

measurements on undisturbed samples, and downhole nuclear densimeters. Each method has 

certain advantages and disadvantages.

The SPT is the most widely available method and the most widely used. However, it is also 

the least reliable method for estimating potential settlements, bearing capacity, and relative 

density of the compacted soils. SPT resistance N values are variable depending upon a 
4-83 

number of factors. There is also significant scatter in the correlations of SPT resistance with 

relative density and with the soil properties needed to estimate settlement and bearing 

capacity. In addition, if data are obtained before pore pressures have dissipated, the 

penetration resistance will not be representative of the actual degree of soil improvement. 

SPT data are usually taken at 5 foot intervals, which is inadequate to properly evaluate the 

vertical variability of the vibro-compaction. However, this can be overcome by specifying 

continuous sampling, as is frequently the case. 

The static CPT overcomes most of the disadvantages encountered with the SPT, and is 

considered the best available QA/QC method. The CPT is particularly advantageous since it 

is relatively inexpensive and can be used directly to estimate settlements in compacted areas. 

The cone resistance, however, will underestimate the degree of improvement if excess pore 

pressures are present.

Measurement of surface subsidence is an excellent way of monitoring the average increase in 

relative density, when the fill material is obtained from the compacted area. This method can 

also be used to check compaction of large areas if the quantity of imported fill is known. As a 

practical matter, it is difficult to accurately verify compaction achieved for footings with this 

method, and it is not possible to check for the minimum compaction achieved. 

Downhole nuclear densimeters offer an alternate method for verifying final densities, but 

have not been used enough to establish their advantages and disadvantages relative to vibrocompaction. With this method, a small diameter aluminum pipe is placed in the ground to the 

planned compaction depth prior to compaction. Before and after compaction, a site-calibrated 

nuclear probe is lowered down the casing to obtain a continuous density-moisture- content 

profile. This method indicates the density within approximately 150 mm (6 in.) of the 

aluminum pipe. 

3.5 Cost Information 

3.5.1 Cost Components

Using the criteria described in Section 3.3, vibro-compaction point spacing can be 

determined. The total area requiring improvement can then be divided by the effective area 

of each point to determine the number of vibro-compaction locations required. In estimating 

costs, it is important to include the perimeter zone outside the limits of loaded area or 

influenced by vibro-compaction in the surface area calculation so that the project

requirements are accurately matched. The depth of improvement required can then be 

multiplied by the number of points to determine budget footage of vibro-compaction. It is 

normally more economical to lower the entire site elevation by the vibratory compaction 
4-84 

effort rather than add granular backfill from the surface. A typical price per linear foot of 

vibro-compaction would be $5 when no backfill is placed around the probe and $8 when 

granular backfill is added. The specific backfill cost will vary significantly on a local basis. 

In addition, mobilization/demobilization costs should be added. Other costs that should be 

considered include the following: 

• Surface densification. With the lack of overburden restraint, the upper 3 feet of soil

will have to be densified by conventional surface compaction methods.

• Additional fill to raise the site to the required grade and, in the case of no added

backfill, to compensate for the site's depression. This cost will depend on the

looseness of the in situ soil and the specified degree of densification.

• Verification testing. Standard Penetration Tests (SPT) are normally specified and, to

ensure uniformity, some tests should be continuous. On large projects, Cone

Penetrometer Tests (CPT) commonly supplements SPTs. Both of these tests are

performed at the centroid of the vibro-compaction points, thus giving the lowest

readings. An average reading could be obtained by testing in the middle of a line

connecting two points.

For typical area densification problems, the cost range for a vibro-compaction solution will 

vary from $1 to $3/yd3

 of densified in situ soils, depending mainly on the size of the project, 

gradation of in situ soils, and degree of densification required. However, in marginal soils 

where special backfill is required, the costs could be significantly higher, yet the total 

economics may justify a vibro-compaction solution. 

The many factors that can affect the pricing of a vibro-compaction project are listed in Table 

4-9. 
4-85 

Table 4-9. Factors Affecting Price of Vibro-compaction Projects

Category Factors 

In situ Material

• Type of Material

• In situ Density

• In situ Cementation

Backfill Material • Type

• Cost

Densification

Requirements

• Load Bearing

• Degree of Densification

1. Average relative density

2. Minimum relative density

Project

Requirements

• Size

• Depth of Densification

• Overburden

• Type

1. Footing compaction

2. Area compaction

• Specifications

• Location of Project

1. Labor and union considerations

2. Support equipment availability

3. Weather - freezing weather conditions

Pricing

• Compaction Spacing

• Unit Pricing

1. Linear foot

2. Cubic yard

The following procedure may be used for estimating the cost of vibro-compaction: 

1. Determine the performance requirement. Section 3.3 lists typical requirements for

most projects.

2. Determine the number of compaction points required from the performance

requirement, resulting compaction point spacing, and total project size.

3. Determine the required depth of compaction from the subsurface investigation and

project requirements.

4. Cost of vibro-compaction = (# of compactions x depth x unit price) + mobilization.
4-86 

5. Price includes supervision, labor, equipment, tools, utilities and backfill added during

compaction. (About 1 cubic yard of backfill added for each 5 linear feet of

compaction.) Rate of production = 300 linear feet per vibrator per 8 hour day.

Add the cost of additional fill to raise the grade. 

3.5.2 Cost Data 

Cost information for transportation related vibro-compaction projects, where approximate 

unit costs are available, is summarized in Table 4-10. 

Table 4-10. Cost Information Summary

Pay Item 

Description

Quantity 

Range Unit

Low 

Unit 

Price

High 

Unit 

Price

Factors Which May 

Potentially Impact Costs 

Mobilization 1 LUMP 

SUM $20,000 $30,000 

Mobilization cost 

increases for distances 

greater than 500 miles. 

Phased projects may 

require multiple 

mobilizations.

Vibrocompaction Greater 

than 2,500 LF $5.00 $9.00 

Production rates increase 

as depth increases.

In situ density of soils 

impacts the average 

production rate. 

Granular Fill 

Material – TON $7.00 $20.00 

Material specifications and 

haul distance will impact 

unit costs. 

3.6 Case Histories

The vibro-compaction technique is used to achieve a variety of design objectives. The case 

histories selected for this chapter represent different transportation applications. 

3.6.1 I-90 Mt. Baker Ridge, Seattle, WA (Hayward Baker 1989) 

Environmental considerations played a major role in an extensive improvement and 

expansion program for the I-90 corridor through the Mt. Baker Ridge area in Seattle, WA. 

With stretches of the improved interstate designed to carry 50,000 vehicles each way daily, 

the impact on residential communities was alleviated by deep-cut construction 

accommodating covered roadways. The roadway structures would support landscaped parks, 

effectively reclaiming these construction areas.
4-87 

Massive pier footings on grade were required to support the covered roadway cross-section. 

In addition to providing vertical support for the cross-section, the footings were also designed 

to carry the lateral load of embankment soils placed behind the wall. For units 9 and 10 of the 

2,590-feet-long roadway section abutting the Mt. Baker Tunnel, the 300-foot-long by 30-

foot-wide footing was to be placed directly on soils previously placed for the existing 

highway embankment. Originally, the footing design assumed a 5,960 psf allowable bearing 

on to the fill soils. However, subsequent geotechnical investigation determined that the loose 

to medium-dense, silty, gravelly, fine-to-medium sand fill (approximately 20 foot depth) 

could not support the 5,960 psf loading without extensive settlement. Washington 

Department of Transportation engineers, in conjunction with their consultants, considered 

both deep foundations and in situ soil improvement. Based on time and cost considerations, 

vibro-compaction was chosen as the best solution. 

To meet densification criteria, stone backfill was specified for the vibro-compaction process. 

At 540 compaction points, a 120 kW vibrator densified the soils to a depth averaging 

between 15 to 20 feet, as shown in Figure 4-38. 

Hayward Baker 1989

Figure 4-38. Vibro-compaction on Mt. Baker Ridge’s Interstate 90.
4-88 

This treatment depth allowed for densification/reinforcement through the existing soils and 

through the loose upper zone of the underlying sandy glacial deposit. The area of treatment 

included a 10 to 20 foot perimeter around the entire footing. The compaction points were 

spaced on a 6 foot grid pattern with the design intent to limit settlement to a specified 

requirement of 3⁄4 inch. 

Two plate-load tests were performed at selected compaction point locations during the work. 

A 6 foot by 6 foot plate was placed directly over each of two points and loaded in increments 

to 210,000 lbf. The total load represented a uniform 5,960 psf pressure on the test plate. The 

test work indicated that average total settlement under the working design load was 

approximately (0.5 inch) with permanent plastic deformation upon unloading indicated to be 

approximately 1⁄4 -inch. 

3.6.2 Wando Terminal, Charleston, SC (Hussin and Foshee 1994) 

In South Carolina, a site improvement challenge involved the expansion of Wando Terminal, 

a State port facility in Mount Pleasant, near Charleston. The expanded terminal was designed 

to serve as a docking facility and as a (267,000 yd2 (225,000 m2

) concrete-paved container 

storage yard. The site of the expansion section was located north of the existing facility 

(Figure 4-39). 
4-89 

Hussin and Foshee 1994

Figure 4-39. State Pier 41, Wando Terminal. 

The storage yard for the expansion was divided into three areas: Area A (53,000 yd2

 (44,500 

m2

), Area B (87,000 yd2 (72,900 m2), and Area C (131,000 yd2 (109,300 m2). (Areas A and B 

were formerly marshlands that, over 10 years ago, had been filled to elevation +22.0 feet (6.7 

m) MLW or higher. The long-term surcharging of these areas had consolidated the 

underlying marsh deposit sufficiently to eliminate the need for additional ground 

improvement. However, Area C was composed of virgin marshlands. 

Much of the Charleston peninsula is composed of former marshland, filled over the last 350 

years with both earthen materials and man-made debris. Many different structures have been 

built within these areas, and numerous problems have resulted, including areal subsidence in 

the range of 2 to 4 inches (50-100 mm) per year over the life of a structure. In the early 

design stages for the new container storage yard, the owner decided that the above 

settlements could not be tolerated. Since the existing Wando terminal is viewed as the 

showcase of South Carolina State Port Authority's Charleston facilities, the expansion was 

required to be of comparable quality. Replacing some, or all, of the deep deposits of marsh 

mud in Area C with less compressible soil was determined to be the only option. 
4-90 

A generalized profile of the subsurface conditions within Area C is depicted in Figure 4-40. 

Hussin and Foshee 1994

Figure 4-40. Vibro-compaction at Wando Terminal. 

This profile essentially represented the worst-case conditions for analyzing the various 

ground improvement alternatives being considered to create the container storage yard. As 

can be inferred from the soil properties listed in Figure 4-41, the marsh mud was extremely 

soft and compressible. Although still relatively soft, the intermediate "firm" clay was more 

consistent and did not present the same design challenges with respect to compressibility and 

stability. The lower stratum (Cooper Marl), due to its high over-consolidation ratio, is 

virtually incompressible. 
4-91 

Hussin and Foshee 1994

Figure 4-41. Generalized subsurface profile of Area C.

It was determined that to achieve acceptable results, the required foundation improvement 

program would have to be a three-step process:

1. Dredge the soft clay to elevation -25 feet (-7.5 m) and replacing that material with 1.2

million cubic yards (meters) of clean sand to elevation +10 feet (+3 m).

2. Install vertical drains (wick drains) to accelerate the consolidation of the underlying

clays.

3. Transform the very loose sand backfill into dense sand using vibro-compaction

(Figure 4-42).
4-92 

Hussin and Foshee 1994

Figure 4-42. Densification of loose sand backfill during vibro-compaction at Wando 

Terminal.

The third step, involving vibro-compaction, would densify the backfill and eliminate a costly 

intermediate step of dewatering the site.

By selecting a dredge level of -25 feet (-7.5 m) MLW, only isolated pockets of the highly 

compressible marsh mud would be left in place. By backfilling with clean sand (fines # 5 

percent), and inserting wick drains to elevation -40 feet (-12.2 m) MLW on 5 foot (1.5 m) 

centers, it was estimated that the maximum post-construction settlement of the container yard 

would include 3.5 to 7 inches (90 to 180 mm) of primary consolidation over the first 2 years, 

and up to 4 inches (100 mm) of secondary compression over the next 50 years. 

Once the specifics of the program were determined, the backfilling of the 130,720 yd2

, 

(109,300 m2) excavation with underwater fill could begin. This fill was specified to be fine 

sand with less than 1 percent clay, and less than 5 percent fines (silt and clay), by weight. 

The contractor elected to fill the excavation by hydraulically pumping the sand from a central 

dumping area. 

With the backfill and wick drains in place, vibro-compaction (utilizing 4 rigs working double 

shifts, 6 days per week for 5 months) could complete the improvement program (Figure 4-

41). At the completion of this process, the sand backfill was completely densified, lowering 

the surface elevation from 35 feet to 31 feet (10.7 m to 9.5 m). 
4-93 

Pre-treatment and post-treatment cone penetrometer test results (Figure 4-43) confirmed that 

the required minimum densification had been achieved. Vibro-compaction densification 

stabilized the soils to support the weight of the containers. 

Hussin and Foshee 1994

Figure 4-43. Sample cone penetration results.

Additionally, since the Wando Terminal site is situated within one of the most prominent 

areas of seismicity along the Atlantic Seaboard, the densification served as a precaution 

designed to prevent liquefaction of soils should an earthquake occur (Charleston was struck 

by a large earthquake in 1886). 

The uniqueness of this large, multi-step project was twofold. First, its size (130,700 yd2

(109,300 m2

)) proved to be one of the largest areas treated to date. Second, the program 

achieved project goals by backfilling the excavation through the water, instead of traditional 

dewatering and filling the hole with compacted layers of sand. Vibro-compaction proved to 

be equally effective and considerably more economical than the dewatering alternative.
4-94 

3.6.3 Manchester Airport, New Hampshire (Sobel et al. 1993)

Construction of a new, 160,000 ft2 (15,000-square-meter) terminal building at Manchester 

Airport, New Hampshire, over loose, sandy, potentially liquefiable soils required that a 

ground improvement program be developed to mitigate the risk of liquefaction during a 

seismic event.

Design phase borings had revealed delta-deposited, clean, uniformly graded, saturated, fineto-medium sands from depths of 12 to 45 feet (3.7 to 13.7 m). Laboratory gradation and 

Standard Penetration Testing revealed the potential for seismically induced liquefaction. The 

design of the densification program was based on specific parameters developed from 

1. Methodology proposed to determine the factor of safety against the occurrence of

liquefaction (Seed et al. 1985).

2. Correlation of SPT values to volumetric strain (Tokimatsu and Seed 1987).

Analysis performed in accordance with the above indicated a factor of safety against 

liquefaction of less than unity under regional design criteria and a volumetric strain of 10 

percent of the layer thickness that translated into a potential for 1 foot (0.3 m) of settlement 

below the building footprint. 

Both deep foundation and ground improvement alternatives to allow shallow footing 

construction were evaluated. The deep foundation alternatives were eliminated due to cost 

considerations and the uncertainty of performance under liquefaction conditions. Of the 

ground improvement alternatives considered (including excavation/replacement, dynamic 

compaction, deep blasting, and compaction grouting), vibro-compaction was selected 

because of its cost-effectiveness and proven success record in sands. 

The vibro-compaction design was required to meet seismic criteria of a design earthquake 

magnitude of 6.0, a peak ground acceleration of 0.12g, and a minimum factor of safety 

against liquefaction of 2.0. Allowable differential settlement was determined to be 0.5 inch 

(12 mm), with a 1 inch (25-mm) allowable total settlement.

To meet these criteria, compaction points were located on a 10 foot by 10 foot (3 m by 3 m) 

grid. The necessary depths of compaction were determined to be 26 feet (8 m) and 36 feet 

(11 m). Although design borings had identified potentially liquefiable soils to 45 feet (13.7 

m), actual depth-of-treatment selection was based upon performance studies of Japanese sites 

where liquefaction had occurred. 
4-95 

The spacing and depth of treatment used resulted in the minimum specified relative density 

where coarse, clean sand was present. Where post-treatment tests indicated that loose relative 

density conditions remained, the spacing was reduced. 

The project required over 2,600 compaction points. Thirty-four post-treatment SPTs were 

conducted, typically at 1100 yd2 (900 m2

) intervals, at the centroid of the compaction point 

grid to assess the vibro program. Compliance with project specifications was generally 

achieved after initial treatment. Where SPT values were at or slightly below specified values, 

at depths ranging from 12 feet to 17 feet (3.7 m to 5.2 m), it was attributed to the presence of 

a dense crust of coarse sand temporarily arching over the loose material below. Subsequent 

testing, after a waiting period of 1 to 3 weeks, showed that, in most instances, N values had 

increased with time to meet the specified criteria.
4-96 

4.0 REFERENCES

Bobylev, L.M. (1963). Distribution of Stresses, Deformations, and Density in Soil During 

Consolidation of Made Ground by Tamping Plates. Osnovoniya, Fundamenty; 

Mekhonika Gruntov. 

Brown, R.E. (1977).Vibroflotation Compaction of Cohesionless Soils. Journal of 

Geotechnical Engineering, ASCE, 103(12): pp. 1437-1451. 

Cooke, H. G. and Mitchell, J.K. (1999). Guide to Remedial Measures for Liquefaction 

Mitigation at Existing Highway Bridge Sites. MCEER-99-0015, Multidisciplinary 

Center for Earthquake Engineering Research, University of Buffalo, Buffalo, NY. 

Debats, J.M. and Sims, M. (1997). Vibroflotation in Reclamations in Hong Kong. Ground 

Improvement, Vol. 1, pp. 127-145. 

Degen, W. (1997). 56m Deep Vibro-Compaction at German Lignite Mining Area. Proc. 3rd

International Conference on Ground Improvement Systems. London, UK. 

Dumas, J.C. and Beaton N.F. (1992). Dynamic Compaction, Suggested Guidelines for 

Evaluating Feasibility- for Specifying- for Controlling. Proc. Canadian Geotechnical 

Conference, pp. 54-1 to 54-12. 

FHWA. (1986). Dynamic Compaction for Highway Construction, Author: Lukas, R., 

FHWA/RD-86/133, Federal Highway Administration, U.S. DOT, Washington, D.C. 

GEC 1. (1995). Dynamic Compaction. Author: Lukas, R., FHWA SA-95-037, Federal 

Highway Administration, U.S. DOT, Washington, D.C., 97p. 

GEC 3. (1997). Geotechnical Earthquake Engineering for Highways, Volume I – Design 

Principles, Authors: Kavazanjian, E., Jr., Matasović, T., Hadj-Hamou, F., and 

Sabatini, P.J., FHWA SA-97-076, Federal Highway Administration, U.S. DOT, 

Washington, D.C. 

Han, J. (2015). Principles and Practice of Ground Improvement. John Wiley & Sons, Inc., 

Hoboken, NJ, 418p. 

Hayward Baker. (1989). I-90 Lid Structure - Mt. Baker Ridge, Seattle, Washington. Vibro 

Systems Case Histories, GKN Hayward Baker, 4p. 

Hobbs, N.B. (1976). Discussion of Symposium: Ground Treatment by Deep Compaction. 

Institution of Civil Engineers, London, UK, 104p. 
4-97 

Holtz, R.D. (1975). Treatment of Soft Foundations for Highway Embankments. National 

Cooperative Highway Research Program, Syntheses of Highway Practice Report 29, 

Transportation Research Board, Washington, D.C., 25p. 

Hussin, J.D. and Foshee, F. (1994). Wando Terminal Ground Improvement Program. Proc. 

Dredging ‘94, ASCE, New York, NY, pp. 1416-1425. 

Idriss, I.M. and Boulanger, R.W. (2008). Soil Liquefaction During Earthquakes. Earthquake 

Engineering Research Institute Monograph MNO-12, 235p. 

Kerisel, J. (1985). The History of Geotechnical Engineering up Until 1700. Proc. XI 

International Conference on Soil Mechanics and Foundation Engineering, San 

Francisco, CA, Golden Jubilee Volume, A. A. Balkema, pp. 3-93. 

Kimmerling, R.E. (1994). Blast Densification for Mitigation of Dynamic Settlement and 

Liquefaction. Report No. WA-RD 348.1, Washington Department of Transportation, 

Tumwater, WA.

Kirsch, K. and Kirsch, F. (2010). Ground Improvement by Deep Vibratory Methods. Spon 

Press, 189p. 

Loos, W. (1963). Comparative Studies of the Effectiveness of Different Methods for 

Compacting Cohesionless Soils. Proc. 1st International Conference on Soil 

Mechanics and Foundation Engineering, Vol. 3, pp. 174-179. 

Lukas, R. (1997). Delayed Soil Improvement after Dynamic Compaction. Ground 

Improvement, Ground Reinforcement, Ground Treatment Developments 1987-1997. 

V.R. Schaefer, Editor, Geotechnical Special Publication No. 69, Geo-Institute of 

ASCE, New York, NY, pp. 409-421. 

Massarsch, K.R. (1991). Deep Soil Compaction Using Vibratory Probes in Deep Foundation 

Improvement. Standard Technical Publication 1089, ASTM, Philadelphia, PA, pp. 

297-319. 

Massarsch, K.R. and Fellenius, B.H. (2002). Vibratory Compaction of Coarse-Grained Soils. 

Canadian Geotechnical Journal, 39(3): pp. 695-709. 

Massarsch, K.R. and Fellenius, B.H. (2005). Deep Vibratory Compaction of Granular Soils. 

Chapter 19 in Ground Improvement – Case Histories, B. Indranatna and J. Chu, 

Editors, Elsevier Publishers, pp. 633-658. 
4-98 

Mayne, P.W., Jones, J.S., and Dumas, J.C. (1984). Ground Response to Dynamic 

Compaction, Journal of Geotechnical Engineering, ASCE, 110(6): pp. 757-774. 

Menard, L. and Broise, Y. (1975). Theoretical and Practical Aspects of Dynamic 

Consolidation. Geotechnique, 25(1): pp. 3-18. 

Mitchell, J.K. and Solymar, Z.V. (1984). Time Dependent Strength Gain in Freshly 

Deposited or Densified Sand. Journal of Geotechnical Engineering, ASCE, 110(10): 

pp. 1415-1430. 

Moseley, M.P. and Preibe, H.J. (1993). Vibro Techniques. Chapter 1 in Ground 

Improvement, M.P. Moseley, Editor, Blackie Academic & Professional, Glasgow, 

Scotland, pp. 1-19. 

Schaefer, V.R. Editor (1997). Ground Improvement, Ground Reinforcement, Ground 

Treatment Developments 1987-1997. Geotechnical Special Publication No. 69, GeoInstitute of ASCE, New York, NY, 616p. 

Schmertmann, J. (1991). The Mechanical Aging of Soils. Journal of Geotechnical 

Engineering, ASCE, 117(9): pp. 1288-1330. 

Seed, H.B., Idriss, I.M., and Arango, I. (1983). Evaluation of Liquefaction Potential Using 

Field Performance Data. Journal of Geotechnical Engineering, ASCE, 109(3): pp. 

458-482. 

Seed, H.B., Tokimatsu, K., Harder, L., and Chung, R. (1985). Influence of SPT Procedures in 

Soil Liquefaction Resistance Evaluations. Journal of Geotechnical Engineering,

ASCE, 101(12): pp 1245-1445. 

Siskind, D.E., Stagg, M.S., Kopp, J.W., and Dowding, C.H. (1980). Structure Response and 

Damage Produced by Ground Vibration from Surface Mine Blasting. Bureau of 

Mines, Department of Investigation, RI 8507. 

Slocombe, B. (2013). Dynamic Compaction. Chapter 3 in Ground Improvement, Third 

Edition, K. Kirsch and A. Bell, Editors, CRC Press, Taylor & Francis Group, Boca 

Raton, FL, pp. 57-85. 

Sobel, J.M. Baez, J.I., and Swekosky, F.J. (1993). Liquefaction Risk Management - 

Manchester Airport. Proc. Third International Conference On Case Histories In 

Geotechnical Engineering, St. Louis, MO, pp. 1709-1713. 
4-99 

Svinkin M.R. (2015). Tolerable Limits of Construction Vibrations. Practice Periodical on 

Structural Design and Construction, ASCE, 20(2): 04014028-1 - 04014028-7. 

Tokimatsu, K. and Seed, H.B. (1987). Evaluation of Settlements in Sands Due to Earthquake 

Shaking. Journal of Geotechnical Engineering, ASCE, Vol. 113, No. 8. pp. 861-878. 

USACE. (1938). Compaction Tests and Critical Density Investigation of Cohesionless 

Materials for Franklin Falls Dam. U.S. Engineer Office, Boston, MA. 

Vibroflotation Foundation Co. Literature. (1980). 8p. 

Welsh, J.P. (1986). In-Situ Testing For Ground Modification Techniques. Use of In-Situ 

Tests in Geotechnical Engineering, Geotechnical Special Publication No. 6, ASCE, 

New York, NY, pp. 322-335. 

Youd, T.L., Idriss, I.M., Andrus, R.D., Arango. I., Castro, G., Christian, J.T., Dobry, R., 

Finn, W.D.L., Harder, L.F., Hynes, M.E., Ishihara, K., Koester, J.P., Liao, S.S.C., 

Marcuson, W.F., Martin, G.R., Mitchell, J.K., Moriwaki, Y., Power, M.S., Robertson, 

P.K., Seed, R.B., and Stokoe, K.H. (2001). Liquefaction Resistance of Soils: 

Summary Report from the 1996 NCEER and 1998 NCEER/NSF Workshops on 

Evaluation of Liquefaction Resistance of Soils. Journal of Geotechnical and 

Geoenvironmental Engineering, ASCE, 127(10): pp. 817-833. 

Chapter 5 

AGGREGATE COLUMNS

5-i 

CONTENTS

1.0 INTRODUCTION.................................................................................................... 5-1

1.1 Description and History ...................................................................................... 5-1

1.1.1 Description ......................................................................................................... 5-1

1.1.2 Stone Columns ................................................................................................... 5-1

1.1.3 Rammed Aggregate Piers .................................................................................. 5-2

1.2 Historical Overview ............................................................................................. 5-2

1.3 Focus and Scope ................................................................................................... 5-3

1.4 Primary References ............................................................................................. 5-4

1.5 Related Technologies ........................................................................................... 5-4

1.5.1 Vibro-Concrete Columns ................................................................................... 5-4

1.5.2 Controlled Modulus Columns ............................................................................ 5-5

1.5.3 Design and Construction Considerations of VCCs and CMCs ......................... 5-5

2.0 DESIGN CONSIDERATIONS ............................................................................... 5-6

2.1 Applications .......................................................................................................... 5-6

2.1.1 Embankments ..................................................................................................... 5-6

2.1.2 Bridge Approach Fills ........................................................................................ 5-6

2.1.3 Bridge Abutment and Foundation Support ........................................................ 5-7

2.1.4 Liquefaction ....................................................................................................... 5-7

2.2 Advantages and Potential Disadvantages of Aggregate Columns ................... 5-7

2.2.1 Advantages ......................................................................................................... 5-7

2.2.2 Disadvantages .................................................................................................... 5-8

2.3 Feasibility Evaluations......................................................................................... 5-8

2.3.1 Geotechnical ...................................................................................................... 5-8

2.3.2 Environmental Considerations ........................................................................... 5-9

2.3.3 Site Consideration ............................................................................................ 5-10

2.4 Limitations .......................................................................................................... 5-10

2.5 Alternative Modification Methods ................................................................... 5-10

2.5.1 Gravel Drains ................................................................................................... 5-10

2.5.2 Sand Compaction Piles .................................................................................... 5-10

2.5.3 Rammed Stone Columns.................................................................................. 5-10
5-ii

3.0 CONSTRUCTION MATERIALS AND EQUIPMENT .................................... 5-12

3.1 Stone Columns .................................................................................................... 5-12

3.1.1 Construction ..................................................................................................... 5-12

3.1.2 Backfill Material .............................................................................................. 5-23

3.2 Rammed Aggregate Columns ........................................................................... 5-23

3.2.1 Construction ..................................................................................................... 5-23

3.2.2 Backfill Material .............................................................................................. 5-26

4.0 DESIGN .................................................................................................................. 5-27

4.1 Stone Columns .................................................................................................... 5-27

4.1.1 Design Considerations ..................................................................................... 5-27

4.1.2 Design Procedure ............................................................................................. 5-28

4.1.3 Seismic Design................................................................................................. 5-39

4.2 Rammed Aggregate Piers .................................................................................. 5-42

4.2.1 Design Considerations ..................................................................................... 5-42

4.2.2 Design Procedures ........................................................................................... 5-43

4.2.3 Settlement Analysis ......................................................................................... 5-44

4.3 Design Examples ................................................................................................ 5-46

4.3.1 Rammed Aggregate Piers ................................................................................ 5-46

4.3.2 Stone Columns ................................................................................................. 5-47

4.4 Design Verification............................................................................................. 5-49

5.0 CONSTRUCTION SPECIFICATIONS AND QUALITY ASSURANCE ....... 5-52

5.1 Aggregate Column Performance Specification ............................................... 5-52

5.2 Field Inspection and Improvement Verification ............................................. 5-52

5.2.1 Stone Columns ................................................................................................. 5-53

5.2.2 Rammed Aggregate Columns .......................................................................... 5-55

5.2.3 Verification Testing ......................................................................................... 5-57
5-iii

6.0 COST DATA .......................................................................................................... 5-59

7.0 CASE HISTORIES ................................................................................................ 5-62

7.1 Rammed Aggregate Piers Case History ........................................................... 5-62

7.1.1 Basic Information............................................................................................. 5-62

7.1.2 Project Summary .............................................................................................. 5-62

7.2 Stone Columns Case History ............................................................................ 5-65

7.2.1 Basic Information............................................................................................. 5-65

7.2.2 Resources ......................................................................................................... 5-65

7.2.3 Project Summary .............................................................................................. 5-65

8.0 REFERENCES ....................................................................................................... 5-67
5-iv

LIST OF FIGURES

Figure 5-1. Soils applicable for stone columns. ..................................................................... 5-2

Figure 5-2. Top feed vibro-replacement. ............................................................................. 5-12

Figure 5-3. Bottom feed vibro-displacement. ...................................................................... 5-12

Figure 5-4. Suspended vibrator. ........................................................................................... 5-14

Figure 5-5. Typical vibrator cross-section. .......................................................................... 5-14

Figure 5-6. Truck mounted crane utilized for top feed vibro-replacement.......................... 5-16

Figure 5-7. Stone column dry bottom feed rig. .................................................................... 5-17

Figure 5-8. Dry bottom feed rig. .......................................................................................... 5-18

Figure 5-9. Top feed vibro rig.............................................................................................. 5-19

Figure 5-10. Marine double-lock gravel pump. ................................................................... 5-20

Figure 5-11. Quality control output over time for dry bottom feed vibro-displacement: 

depth (left), amperage (middle) and gravel (right). ....................................................... 5-21

Figure 5-12. Quality control output over depth for dry bottom feed vibro-displacement: 

amperage (left), compaction (middle), and column diameter over depth (right). .......... 5-22

Figure 5-13. Replacement method, rammed aggregate pier construction process: from 

left to right, (1) make a cavity, (2) place stone at bottom of cavity, (3) ram stone to 

form bottom bulb, (4) place and ram thin lifts to form undulated side shaft. ................ 5-24

Figure 5-14. Rammed aggregate pier tamper. ..................................................................... 5-25

Figure 5-15. Replacement method, rammed aggregate pier with predrilling. ..................... 5-25

Figure 5-16. Equilateral triangular pattern of stone columns. ............................................. 5-29

Figure 5-17. Unit cell idealization. ...................................................................................... 5-30

Figure 5-18. Failure modes of a single stone column in a homogenous soft layer. ............. 5-33

Figure 5-19. Failure modes of a single stone column in a non-homogeneous cohesive 

soil. ................................................................................................................................. 5-33

Figure 5-20. Failure modes of stone column groups. .......................................................... 5-34

Figure 5-21. Comparison of elastic theories and field observations. ................................... 5-36

Figure 5-22. Notation used in average stress method stability analysis. ............................. 5-38

Figure 5-23. Relationship between stress ratio causing liquefaction and (N1)60 values 

for silty sands for 7.5 magnitude earthquakes. .............................................................. 5-41
5-v 

Figure 5-24. Example problem 1 geometry. ........................................................................ 5-46

Figure 5-25. Aggregate column ground improvement layout. ............................................ 5-47

Figure 5-26. Example problem 2 geometry and soils. ......................................................... 5-48

Figure 5-27. Example problem 2 settlement ratio determination. ....................................... 5-49

Figure 5-28. Rammed aggregate pier installation. ............................................................... 5-63

Figure 5-29. Completed MSE wall supported on rammed aggregate piers. ........................ 5-64

Figure 5-30. Stone column installation. ............................................................................... 5-66

LIST OF TABLES

Table 5-1. Suitability for Testing Aggregate Columns ........................................................ 5-57

Table 5-2. Unit Costs ........................................................................................................... 5-60

5-1 

1.0 INTRODUCTION

1.1 Description and History

Over the past 30 years, aggregate column technology has become established in the United 

States as a viable ground modification technique. It has been applied extensively for 

remediation and new construction of transportation facilities. Construction of highway 

embankments using conventional design methods, such as preloading, dredging, and soil 

displacement techniques, can often no longer be used due to environmental restrictions and 

post-construction maintenance expenses. Aggregate columns have a proven record of 

experience and are ideally suited for use in clays, silts, loose silty sands, and uncompacted 

fills. The history and development of aggregate columns, the focus and scope of this 

technical summary and primary references are discussed in this section. 

1.1.1 Description

This technical summary on aggregate columns includes both rammed aggregate piers and 

stone columns. The similarities and differences of both types of columns will be presented in 

the following sections of the document. When discussing the attribute of both stone columns 

and rammed aggregate piers the term aggregate columns will be used. However, if an 

attribute is specific to only one of the two types of aggregate columns, the specific column 

will be identified as a stone column or rammed aggregate pier.

1.1.2 Stone Columns

Stone column construction is accomplished by down-hole vibratory methods. The technique 

of creating stone columns involves the introduction of backfill material into the soil so that 

dense and sometimes deep columns of aggregate are formed that are tightly interlocked with 

the surrounding soil. 

The stone column construction technique is known as either vibro-replacement or vibrodisplacement, as follows:

• Vibro-replacement - Generally refers to the wet, top feed process in which jetting

water is used to aid the penetration of the ground by the vibrator. Due to the jetting

action, part of the in situ soil is washed to the surface. This soil is then replaced by the

backfill material (e.g. stone). A dry feed process, with soils requiring predrilling, will

also result in in situ soil being brought to surface.

• Vibro-displacement - Generally refers to the dry, top or bottom feed process; almost

no in situ soil appears at the surface, but is displaced by the backfill material.
5-2 

The product of both the vibro-replacement and vibro-displacement construction methods is 

generically referred to as a stone column.

1.1.3 Rammed Aggregate Piers

Rammed aggregate piers are installed by drilling 18- to 36-inch diameter holes into the 

foundation soils and ramming lifts of well-graded aggregate within the holes to form stiff, 

high-density aggregate columns. The drilled holes typically extend from 7 to 33 feet below 

grade. The first lift of aggregate forms a bulb below the bottoms of the piers. Subsequent lifts 

of aggregate are typically 12 inches in thickness. Ramming takes place with a high-energy 

beveled tamper that both densifies the aggregate and forces the aggregate laterally into the 

sidewalls of the hole. This action increases the lateral stress in the surrounding soil, further 

stiffening the stabilized composite soil mass.

1.2 Historical Overview

For over 50 years, deep vibrators have been used to improve the bearing resistance and 

settlement performance of weak soils. As early as 1936, methods and equipment were 

developed that enabled the compaction of non-cohesive soils to depths of 60 feet with 

excellent results. This original process is now referred to as vibro-compaction or vibroflotation. Stone-column technology developed as a natural progression from vibrocompaction and extended vibro-system applications beyond the relatively narrow application 

of densification of clean, granular soils, as shown in Figure 5-1. 

Courtesy Hayward Baker

Figure 5-1. Soils applicable for stone columns.
5-3 

The ability to compact soil depends mainly on the grain size distribution of the soil. Soils 

with grain size distribution curves lying entirely on the coarse side of the hatched line in 

Figure 5-1 are generally readily compacted by the vibro-compaction process. If the grain size 

distribution curve falls to the right of the dotted line the soil is not readily compacted by 

vibro-compaction. It is for these types of soil and their related problems that necessitated the 

development of stone column technology. 

It is interesting to note that one of the first documented uses of stone columns was for the Taj 

Mahal in India, completed in 1653. This historic structure has been successfully supported 

for more than three centuries by hand-dug pits backfilled with stone. The concept of stone 

columns was also used in France in the 1830s to improve native soil (FHWA 1983). Modern 

techniques were first implemented during the 1960s in Europe. After extensive use in 

Europe, the stone column technique was introduced in the United States in the 1970s, but 

saw limited use in its first 12 years, with only 21 completed projects. However, by 1994, this 

number had increased to over 400. This growth is due to the better understanding of the 

design concepts and economics of stone column techniques and the fact that more projects 

were being built on sites with poor soil. Today stone columns are used extensively to 

improve the bearing resistance of soft compressible soils. 

Rammed aggregate piers were developed in the United States in 1984. The concept of short 

dug pits backfilled with aggregate to support structures (i.e., rammed aggregate piers) is not 

new and has been previously documented in the literature. Refinements and improvement to 

this basic technique have been introduced in the last 25 years under the trade name Geopier®. 

These rammed aggregate piers are a ground improvement system that is used extensively to 

improve the bearing resistance of foundation soils. Rammed aggregate piers are installed by 

drilling 2 to 3 foot diameter holes into the foundation soils and ramming lifts of either wellgraded or open graded aggregate within the holes to form very stiff, high-density aggregate 

piers. The drilled holes typically extend from 7 to 33 feet below grade. 

1.3 Focus and Scope

The focus and scope of this technical summary on aggregate columns is to provide guidance 

on: applications where the technology can be utilized, design, contracting methods, and 

quality assurance. References are cited where more detailed technical information can be 

obtained, and typical costs are given in order to make a preliminary technical and economic 

evaluation as to whether aggregate columns, and related technologies, are appropriate for a 

given site and application. It is the intent of this document to serve as a reference on 

aggregate columns and how they may be best utilized on a ground modification project by 

discussing their construction, utilization, and limitations. 
5-4 

1.4 Primary References

• FHWA. (1983). Design and Construction of Stone Columns. Authors: Barksdale,

R.D. and Bachus, R.C., FHWA/RD-83/026, Federal Highway Administration, U.S.

DOT, Vol I and Vol II.

• Collin, J.G. (2007). Evaluation of Rammed Aggregate Piers by Geopier Foundation

Company Final Report, Technical Evaluation Report prepared by the Highway

Innovative Technology Evaluation Center, ASCE, 86p.

1.5 Related Technologies 

There are a variety of column technologies that are related, and similar, to aggregate 

columns. Many of these are proprietary technologies developed by ground modification 

contractors. Some are equipment and installation variations, and may be more suited to 

specific installation conditions, such as beneath the water table or in very soft soils. Many of 

these related technologies use Portland cement binder with the aggregate and, thus, a more 

rigid (cemented aggregate) column is constructed. Another cement based column option is to 

use concrete for construction of the columns. Two common cement based columns are vibroconcrete columns (VCCs) and controlled modulus columns (CMCs); these are briefly 

discussed below. Cement based, concrete columns may be used in softer soils, without 

casing, and can be used to produce a stiffer element than aggregate columns.

1.5.1 Vibro-Concrete Columns

Vibro-Concrete Columns (VCC) are considered a sister technology to stone columns, with 

concrete replacing the stone in the column. The vibro-concrete column is a non-proprietary 

process that employs a vibro-displacement (i.e., bottom feed) depth vibrator to penetrate the 

soils to a level suitable for bearing. Concrete is pumped through the vibrator assembly during 

initial withdrawal. The vibrator then re-penetrates the concrete, displacing it into the 

surrounding soil to form an enlarged column base. The vibrator is then slowly withdrawn as 

concrete is pumped at a maintained pressure to form a continuous shaft of concrete up to 

ground level. At ground level, a slight mushrooming of the concrete column is constructed to 

assist the transfer of the applied load to the vibro-concrete column. 

The vibro-concrete column was first developed in Europe in 1976. Since stone columns 

derive their strength and settlement characteristics from the surrounding soil, their capacities 

are significantly reduced in very soft clay or peat with a thickness greater than 1 to 2 times 

the diameter of the column. Vibro-concrete columns were developed to treat these soils. 

Instead of feeding stone to the tip of the vibrator, concrete is pumped through an auxiliary 

tube to the bottom of the hole. This method can offer the ground modification advantages of 
5-5 

the vibro-systems, with the load carrying characteristics of a deep foundation. The first 

installation of vibro-concrete columns in the United States was in 1994 in Pennsylvania, and 

was used in support of an oil storage tank. They have been used to support embankments 

over soft organic soils in many states including Florida, Illinois, Maryland, New Jersey, 

Puerto Rico, and South Carolina. 

See http://www.GeoTechTools.org for detailed information and guidance on VCCs as well as 

for other cement based column technologies. 

1.5.2 Controlled Modulus Columns

Controlled modulus columns (CMCs) are produced with a proprietary process and are similar 

to VCCs in that the final product is a concrete column. CMCs are constructed using a reverse 

auger method where the auger displaces the site soil until an adequate bearing layer is 

reached. The auger has a hollow stem through which a low slump concrete is pumped as the 

auger is withdrawn. CMCs are a patented technology. 

The CMC technique was developed in the early 90’s in France by the Menard Group. 

Menard developed a series of specifically designed machines and tooling that enabled the 

technology to rapidly grow in use in Europe. DGI-Menard introduced the technology in 2003 

in the USA and Canada with the first CMC project in Vermont for the support of a home 

improvement store. 

1.5.3 Design and Construction Considerations of VCCs and CMCs

A generalized summary of the factors affecting the feasibility of stabilizing soft ground with 

VCCs and CMCs follows: 

1. The allowable design load for VCCs and CMCs is a function of the diameter of the

column, the allowable strength of the concrete, and the strength of the bearing layer.

Typical column diameters range from 18 to 24 inches for VCCs and 10 to 18 inches

for CMCs. Typical allowable design loads range from 150 to 250 kips for VCCs and

75 to 150 kips for CMCs.

2. VCCs and CMCs are typically used in very soft clay and organic soils.

3. Typical lengths vary from 20 to 75 feet.

See http://www.GeoTechTools.org for information on design, specification, quality 

assurance and construction of VCCs. This information can also provide some guidance for 

other cement based column technologies.
5-6 

2.0 DESIGN CONSIDERATIONS 

Aggregate column construction involves the partial replacement or displacement of 

unsuitable subsurface soils with a vertical column of compacted aggregate. This section

discusses applications, advantages and disadvantages, and design considerations for 

aggregate columns. 

2.1 Applications

Aggregate columns can be applied to increase bearing resistance, reduce total and differential 

settlements, accelerate the time rate of settlement, improve slope stability, and reduce the 

liquefaction potential of soil. Typical applications include foundation improvement for the 

construction of highways, embankments, warehouses, and light industrial buildings. 

2.1.1 Embankments

One typical application of aggregate column technology is the stabilization of large area 

loads such as highway embankments. The use of aggregate columns offers a practical 

alternative, where conventional embankments cannot be constructed due to stability 

considerations. Applications include moderate-to-high fills on soft soils, fills that may be 

contained by mechanically stabilized earth, and construction on slopes where stability cannot 

otherwise be obtained. An important related highway application is slope stabilization. In 

1987, the Soil Mechanics Bureau, New York State DOT, reported on the use of the dry 

bottom feed vibro-displacement method to solve a slide problem (Sung and Ramsey 1988).

A considerable amount of highway widening and reconstruction work has occurred over the 

last several decades. Some of this work involved building additional lanes immediately 

adjacent to existing highways constructed on moderate-to-high fills over soft cohesive soils, 

such as those found in wetland areas. For this application, differential settlement between the 

existing and new construction is an important consideration, in addition to embankment 

stability. Support of these new fills on aggregate columns offers a viable design alternative to 

conventional construction. 

2.1.2 Bridge Approach Fills

Aggregate columns can be used to support bridge approach fills, to provide stability, and to 

reduce the costly maintenance problem from settlement at the joint between the approach fill 

and bridge. In 1989, the Texas DOT used 13,000 lineal feet of aggregate columns to support 

mechanically stabilized earth walls for the U.S. 77 overpass situated in Brownsville. In 1990, 

the Texas DOT utilized 42,000 lineal feet of 13 to 20 foot long aggregate columns for 

Brownsville Road over U.S. 77. 
5-7 

Aggregate column supported embankments can be constructed to greater heights than 

conventional approach embankments over soft foundation soils. Therefore, the potential 

exists to reduce the length of bridge structures by extending the approach fills supported on 

aggregate columns. Embankment fills can be placed faster due to the combined effects of 

accelerated drainage and consolidation, and the increase in shear strength supplied by 

aggregate columns. 

2.1.3 Bridge Abutment and Foundation Support

Aggregate columns can be used to support bridge abutments at sites that are not capable of 

supporting abutments on conventional shallow foundations. At such sites, an important 

additional application involves the use of mechanically stabilized earth walls supported on 

stone columns. 

Another potentially cost effective alternative to pile foundations for unfavorable site 

conditions is to support single span bridges, their abutments, and their approach fills on 

aggregate columns. This technique minimizes the differential settlement between the bridge 

and approach fill. 

2.1.4 Liquefaction 

In earthquake prone areas, aggregate columns can be used to reduce the liquefaction potential 

of cohesionless soils supporting embankments, abutments, and soils beneath shallow 

foundations. Aggregate columns can also be used to reduce the liquefaction potential of 

cohesionless soils surrounding existing or proposed pile foundations. This application has 

been used quite extensively for major bridges on pile foundations through liquefiable soils in 

the Pacific Northwest.

2.2 Advantages and Potential Disadvantages of Aggregate Columns 

2.2.1 Advantages 

Aggregate columns are a technical and potentially economical alternative to deep 

foundations, capable of improving the soil sufficiently to allow less expensive, shallowfoundation construction. Aggregate columns are also more economical than the removal and 

replacement of deep, poor bearing soils, particularly on larger sites where the groundwater is 

close to the surface. Where the infrastructure precludes high-vibration techniques, such as 

conventional pile driving, dynamic compaction or deep blasting, the low-vibration aggregate 

column technique is often viable. If time is critical to project start-up, site modification by 

aggregate column installation can be achieved quicker than by pre-loading the soils. In 

seismic areas, aggregate columns can reduce dynamics settlements to acceptable levels, and 
5-8 

in some cases may densify the soils beyond the threshold of liquefaction. Aggregate columns 

also provide radial drainage and a vertical drainage path for excess pore water pressure 

dissipation when low fines content aggregate is used, as well as densifying the liquefiable 

soils.

2.2.2 Disadvantages 

Aggregate columns are not a solution for all soft soil problems. Strata of peat and other 

organic materials, and very soft clays with a thickness greater than the diameter of the 

aggregate column can be inappropriate for aggregate column construction, as they offer 

inadequate lateral support to effectively create the column or to ensure long-term 

performance. Dense overburden, boulders, cobbles, or other obstructions may require predrilling prior to installation of stone columns (rammed aggregate piers that auger a hole to 

create the column generally can overcome this disadvantage). 

Cost, when compared to other solutions, can be a disadvantage of aggregate columns. The 

need to channel and dispose of spoil water in wet feed construction and lateral ground 

displacement with a dry construction process may be major disadvantages at some locations. 

Removal of spoil from the rammed aggregate pier, or predrilled stone column, installation 

may be a major disadvantage when contaminated soils are present. Rammed aggregate piers 

are more costly to install when casing is required, and when casing is used this technology 

may not densify granular soils as effectively as stone columns. 

2.3 Feasibility Evaluations

The aggregate column technique of ground modification has been successful in: (1) 

improving stability of both embankments and natural slopes; (2) increasing bearing 

resistance; (3) reducing total and differential settlements; (4) reducing the liquefaction 

potential of cohesionless soils; and (5) increasing the time rate of settlement.

2.3.1 Geotechnical

The degree of densification resulting from the installation of vibro systems is a function of 

soil type, silt and clay content, soil plasticity, pre-densification relative densities, vibrator 

type, stone shape and durability, aggregate column area, column spacing, and energy applied. 

Experience has shown that soils with less than 15 percent passing a #200 sieve, and clay 

contents of less than 2 percent will densify due to vibrations. Clayey soils do not react 

favorably to the vibrations, and the improvement in these soils is measured by the percent of 

soil replaced and/or displaced by the aggregate column. In the case of clayey soils, the 

ground improvement is achieved by reinforcing the soil. 
5-9 

A generalized summary of the factors affecting the feasibility of stabilizing soft ground with 

aggregate columns is as follows: 

1. The allowable design loading of an aggregate column should be relatively uniform 

and is limited by the lateral support the in situ soil can develop. Typically, with good 

lateral support, a maximum of 110 kips per column is used; and typically, the 

composite factored bearing resistance is increased to 2,000 to 8,000 psf. 

2. The most significant improvement is likely to be obtained in compressible silts and 

clays ranging in shear strength from 300 to 1000 psf. 

3. Aggregate columns should not be used in highly sensitive soils. Special care must be 

taken when using aggregate columns in soils containing organics and peat lenses or 

layers with undrained shear strength of less than 300 psf. Because of the high 

compressibility and low strength of these materials, little lateral support may be 

developed and large vertical deflections of the columns may result. When the 

thickness of the organic layer is greater than one to two aggregate column diameters, 

the ability to develop consistent column diameters becomes questionable. 

4. Ground improved with stone columns reduces settlements typically by 50 to 70 

percent of the unimproved ground response and differential settlements from 5 to 15 

percent of unimproved soil response. Ground improvement with rammed aggregate 

piers can reduce settlement to less than 1 inch, in some loading and subsurface 

conditions. 

5. Due to the development of excessive resistance to penetration of the vibrator a 

practical upper limit is in the range of an undrained strength of 1000 to 2000 psf for 

stone columns. If stone columns are used in these stiff soils or through stiff lenses, the 

column hole is commonly pre-bored, which is often the case in landslide projects. 

This may result in a significant additional cost.

6. The installation of rammed aggregate piers using the typical replacement method 

(drilled method) in soils that do not stand open during drilling (i.e., loose granular 

soils, very soft cohesive soils) may require the use of temporary casing, which 

reduces the installation rate and increases the cost of the piers.

7. Typically, the maximum practical depth of stone columns and rammed aggregate 

piers is 100 feet and 35 feet respectively.

2.3.2 Environmental Considerations 

The selection of the most appropriate aggregate column installation method should consider 

the environmental effects of the installation. Soil spoils must be contained, particularly fines 
5-10 

from air or water jetting operations. The designer may select an alternate column system that 

does not replace the in situ soils.

2.3.3 Site Consideration

Site conditions should always be considered when selecting a ground modification

technology. The installation of aggregate columns requires sufficient headroom (typically 8 

to 10 feet more than the depth of penetration of the column) for the construction equipment. 

Adjacent buildings and structures must be monitored for heave when using vibrodisplacement stone columns. 

2.4 Limitations

The major limitation for aggregate columns is that they are not appropriate in very soft and 

sensitive fine-grained soils and organics. Stone may not be readily available near the project 

site, leading to potentially significant cost ramifications. Rammed aggregate piers have the 

additional limitation on the depth of the column (i.e., typically 35 feet). 

2.5 Alternative Modification Methods

The following alternative methods, which are similar in concept to aggregate piers, have 

been used. 

2.5.1 Gravel Drains

In Japan, gravel drains are installed by backfilling inside a casing and densifying the stone 

with an interior vibrator as the casing is extracted. This provides a good drain, but does little 

to densify the soil outside the casing. For soils with a high liquefaction potential, gravel 

drains alone may not be able to handle the excess pore pressures, and liquefaction may still 

occur.

2.5.2 Sand Compaction Piles

This system is also used extensively in Japan. Sand compaction piles are constructed by 

using a vibratory hammer to install a steel casing to the desired elevation. The casing is filled 

with sand as it is extracted. For more information see The Sand Compaction Pile Method

(Kitazume 2005). 

2.5.3 Rammed Stone Columns

In Belgium, rammed stone columns have been constructed by driving a casing, placing 

granular backfill and dropping a heavy weight on the stone as the casing is extracted. While
5-11 

this system can create some compaction of the surrounding soil, it is a very slow process 

(250 ft/shift/rig) and, therefore, not economically competitive. 

The following alternative methods are covered in the other chapters of this manual:

• Prefabricated vertical drains either with or without preloading – Chapter 2

• Deep and mass mixing methods – Chapter 7

• Jetted grout columns – Chapter 8
5-12 

3.0 CONSTRUCTION MATERIALS AND EQUIPMENT

3.1 Stone Columns

3.1.1 Construction

The primary methods of constructing stone columns are vibro-replacement (wet, top feed) 

and vibro-displacement (dry, top or bottom feed). Where environmental concerns are strong, 

the dry process will typically be required. However, the wet process is more economical, if 

environmental concerns are not as relevant to the project. These processes are illustrated in 

Figures 5-2 and 5-3. 

Figure 5-2. Top feed vibro-replacement.

Figure 5-3. Bottom feed vibro-displacement.
5-13 

3.1.1.1 Vibro-Replacement (Wet, Top Feed)

The original stone column installation technique, called vibro-replacement or the wet 

process, utilizes a high-pressure jet of water to open a hole that the probe follows into the 

ground. The probe is then retracted in increments, and stone is introduced into the void from 

the surface (Figure 5-2). After every increment, the probe is lowered into the new column 

material, thereby densifying and compacting the stone column and, potentially, the 

surrounding soil (depending on percent fine content). This method is best suited for sites with 

soft to firm soils with undrained shear strengths of 300 to 1000 psf and a high groundwater 

table.

3.1.1.2 Vibro-Displacement (Dry, Top and Bottom Feed) 

As the jetting water effluent from the vibro-replacement method includes the finer portion of 

the in situ soil, environmental problems encountered in containment, removal, and disposal 

of the effluent had to be addressed. To resolve these problems, the dry top and dry bottom 

feed techniques were developed. Using the oscillations of the vibrator coupled with its

deadweight, air jetting, and/or pre-augering, the vibrator is inserted into the ground without 

the use of jetting water. For shorter stone columns, the stone can still be fed into the annulus 

created by the vibrator from the surface, as shown in Figure 5-2. For deeper treatment or

where the hole may collapse, the stone is fed to the bottom of the vibrator through an 

attached tremie tube as shown in Figure 5-3. The first major use of the dry bottom feed vibrodisplacement system in the United States was for the Steel Creek Dam foundation at the 

Department of Energy's Savannah River Plant, South Carolina, in 1985 (Dobson 1987). 

3.1.1.3 Equipment 

The equipment used to form stone columns is comprised of the following: 

• Vibrator, which is suspended from extension tubes with air or water jetting systems

• Crane or base machine, which supports the vibrator and extension tubes

• Stone delivery system

• Control and verification devices

The principal piece of equipment used to achieve compaction is the vibrator, which ranges in 

diameter from 12 to 16 inches and in length from approximately 10 to 16 feet. A suspended 

vibrator is shown in Figure 5-4, and a cross section of a typical vibrator is shown in 

Figure 5-5. 
5-14 

Figure 5-4. Suspended vibrator.

Figure 5-5. Typical vibrator cross-section.
5-15 

Horizontal vibrations are produced close to the base of the vibrator and are induced by 

rotating eccentric weights mounted on a shaft and driven by a motor located in the upper part 

of the vibrator casing. Both electric and hydraulic power can be used to power the motor. 

Early units were driven by motors in the 22 to 60 kW range, but more recent machines 

develop up to 125 kW. Centrifugal forces of up to 60 kips at frequencies varying from 1200 

to 3000 rpm are currently achieved. Abrasion resistant wear plates are added to the sides of 

the vibrator, protecting it from excessive wear during raising and lowering from the ground. 

Fins located on the sides of the vibrator reduce rotation. Follower or extension tubes, 

typically of a similar or smaller diameter to the vibrator unit, are attached to it and allow 

treatment of soils at depth. An elastic coupling is used to isolate the vibrator from the 

extension tubes and to prevent vibrations from traveling up the extension tubes to the 

supporting crane or base machines. 

Water or air can be conveyed to the top of the extension tubes by flexible hoses and, 

subsequently, through the extension tubes to the vibrator. The water or air is generally fed to 

the nose of the vibrator to assist penetration into the soil. The thickness of soil to be treated 

determines the overall length of vibrator, extension tubes, and lifting equipment, which, in 

turn, determines the size of crane to be used. The vibrator is suspended from the boom of a 

crane; a 33-foot probe can be easily handled using a 40 ton crane with a 40-foot boom. 

Penetration of the probe is accomplished by vibration, jetting media (air or water), and dead 

weight. The greater the depth of soil to be treated, the larger the required crane.

The construction of stone columns requires the importation and handling of substantial 

quantities of granular material. The granular material is routinely handled with front end 

loaders, working from a stone pile and delivering stone to each stone column location. For 

the top feed method, the stone is end-dumped into the hole created by the vibrator. For the 

bottom feed system, stone is fed into a skip. The skip can then supply pipes in the vibrator 

and extension tube assembly with stone. The pipes lead to the vibrator nose and, during 

operations, stone is fed continuously to the very point of compaction. Vibrators can be 

retained in the ground during compaction work, thus maintaining the hole in an open 

condition, and enabling a high integrity stone column to be constructed. Alternate stone 

transport systems have been developed, which allows the transport of stone backfill through a 

6-inch hose instead of a skip, along a leader. Typical equipment to install stone columns is 

illustrated in Figures 5-6 through 5-9. 
5-16 

Figure 5-6. Truck mounted crane utilized for top feed vibro-replacement.
5-17 

Courtesy Hayward Baker

Figure 5-7. Stone column dry bottom feed rig. 
5-18 

Courtesy Treviicos

Figure 5-8. Dry bottom feed rig. 
5-19 

Courtesy Subsurface Constructors

Figure 5-9. Top feed vibro rig. 

A method developed almost 20 years ago, the marine double-lock gravel pump technique (a 

patented process), has been developed to deliver stone to the bottom of the vibrator in 

underwater applications. This method transports the gravel through a system of hoses using 

air pressure supplied through an air compressor. The double-lock system provides excess air 

pressure at the tip of the vibroprobe at all times. This minimizes the potential for soil 

intrusion into the discharge pipe. This method is illustrated in Figure 5-10. 
5-20 

Vibroflotation Group

Figure 5-10. Marine double-lock gravel pump. 

Instrumentation packages that provide a continuous record of construction data for each stone 

column are now common. Measurements of depth, power consumption, and stone 

consumption are recorded against time and provided on a printout at the time of construction. 

Such instrumentation is available for bottom feed vibrator systems and has been used in 

Europe since the 1980s and in the United States since 1993. Sample output is shown in 

Figures 5-11 and 5-12. 
5-21 

Vibroflotation Group

Figure 5-11. Quality control output over time for dry bottom feed vibro-displacement: 

depth (left), amperage (middle) and gravel (right). 
5-22 

Vibroflotation Group

Figure 5-12. Quality control output over depth for dry bottom feed vibro-displacement: 

amperage (left), compaction (middle), and column diameter over depth (right). 
5-23 

3.1.2 Backfill Material

The size, gradation, and shape of backfill for stone columns usually depends upon the: 

• construction technique used,

• subsoil properties that the column is being constructed in,

• application for which the columns are being designed for, and

• local availability of materials.

Backfill may vary by method or technique of placement, i.e., top feed or bottom feed, jetting 

or not, water or air jetting, etc. The method of placement is also a function of the subsoils 

characteristics. Furthermore, the application has to be considered in the selection of the 

backfill. For example, drainage characteristics are crucial in liquefaction prevention and 

shear strength is critical in slope stabilization projects. Whereas, drainage and shear strength 

properties are generally not critical for increasing bearing resistance applications and, of 

course, local availability and material costs will factor into column backfill selection.

For vibro-replacement stone columns, subangular or angular gravel of nearly uniform 

grading 1.0- to 2.5-inch in size is often used. This size backfill passes easily around the 

vibrating probe, while it is still in the hole. The larger sized in situ material suspended in the 

water usually fills the voids between the stone resulting in a rigid column. 

An important factor in the successful construction of wet stone columns is keeping the 

flushing water flowing at all times to wash out the soil fines that infiltrate the stone and to aid 

in stabilizing the hole. Keeping the probe in the hole at all times during installation increases 

the stability of the jetted hole.

For vibro-displacement, well graded backfill with a gradation from 0.4- to 3-inch or up to 4-

inch is generally used to achieve mechanical interlock and filling of voids. The finer backfill 

sizes are included to provide an intermediate particle size between the in situ clay and gravel. 

The bottom feed method is restricted to aggregates of approximately 0.4- to 1.4-inch in size 

to avoid blockage of the equipment. 

3.2 Rammed Aggregate Columns 

3.2.1 Construction

The primary method of constructing rammed aggregate piers is the replacement method, 

shown in Figure 5-13. 
5-24 

Courtesy Geopier Foundation Company

Figure 5-13. Replacement method, rammed aggregate pier construction process: from 

left to right, (1) make a cavity, (2) place stone at bottom of cavity, (3) ram stone to form 

bottom bulb, (4) place and ram thin lifts to form undulated side shaft. 

The replacement method consists of the following: 

• Excavate pier to design depth (make cavity), use casing if hole will not stay open

• Place open graded stone at bottom of cavity, in a 24-inch thick lift

• Ram stone at bottom of cavity

• Place and ram 12-inch lifts of stone until the elevation of the top of the column is

achieved.

Rammed aggregate pier construction equipment is shown in Figures 5-14 and 5-15. 
5-25 

Courtesy Geopier Foundation Company

Figure 5-14. Rammed aggregate pier tamper.

Courtesy Geopier Foundation Company

Figure 5-15. Replacement method, rammed aggregate pier with predrilling.
5-26 

The construction equipment consists of three pieces of equipment: an excavator-mounted 

drill, an excavator-mounted hammer, and a skid-steer loader. The excavator-mounted drill is 

a conventional excavator with typically 2- to 3-foot diameter drilling tools. Common 

excavators are generally used to minimize problems and costs associated with transportation 

of large construction equipment. The excavator-mounted tamper is a conventional excavator 

with a modified concrete breaker attached to the machine that is capable of delivering 245 to 

650 kip-lbf per 1 foot per lift of energy for tamping, that both densifies the aggregate and 

forces the aggregate laterally into the sidewalls of the hole. This action increases the lateral 

stress in the surrounding soil. A composite alloy shaft with an attached beveled-hammer is 

connected to the modified concrete breaker. The size of the tamper should be at least 85% of 

the plan area of the cavity. The third piece of equipment is a skid-steer loader that delivers 

the aggregate to the hole. Most skid-steer loaders are track-driven to provide stability and 

better maneuverability on muddy sites.

3.2.2 Backfill Material 

For rammed aggregate pier construction, clean 1- to 3-inch stone is commonly specified for 

the bottom bulb. The same material is used throughout the pier if radial drainage of the pier is 

included in the design solution; otherwise, a well-graded base course aggregate is used.
5-27 

4.0 DESIGN

4.1 Stone Columns

4.1.1 Design Considerations 

Although the method of introducing the backfill material, and gradation of backfill, is 

somewhat different for vibro-replacement and vibro-displacement, the design approach is 

similar for both techniques. 

The development and rationale of the various design theories for stone columns are outside 

the scope of this technical summary. Sufficient design information is presented to assess the 

feasibility of stone columns. For development of the design theories and in-depth design 

criteria, FHWA (1983). The publication The Design of Vibro Replacement by Priebe (1995) 

updates earlier work and was a popular and widely used design method. However, for the 

most current guidance on the design of stone columns, the reader is referred to 

http://www.GeoTechTools.org. 

In practice, the design of stone columns is to a large extent semi-empirical. Specific state-ofthe-practice design recommendations are given for bearing resistance, settlement, and 

stability analyses. These design recommendations give a rational basis upon which to 

evaluate stone columns. Theoretical results, of course, should always be supplemented by 

past experience and sound engineering judgment. 

The present methods used for analysis and design range from experience based semiempirical methods to finite element analyses. These methods have been typically indexed to 

full-scale field tests, laboratory and analytical models to study and predict the load carrying 

capacity, settlement behavior, shear resistance, and mode of failure of the soil stone-column 

system. 

Weak soils reinforced with stone columns act as a composite medium, exhibiting increased 

stiffness with reduced spacing, increased column cross-sectional area, and angle of friction 

for the imported stone. The columns are stiffer than the in situ soils they replace or displace, 

and rely on the lateral support of the adjacent soil to function properly. Consequently, the 

columns must have adequate lateral support to preclude a bulging failure and terminate 

typically in a denser stratum to preclude a bearing resistance failure.

Since the stone column is more rigid than the surrounding soil, it settles less than the 

adjoining soil under load. Therefore, it carries, by arching, a larger portion of the imposed 

load. As further consolidation of the in situ soil occurs, additional load transfer to the stone 

column occurs until an equilibrium condition is reached. This transfer of load to the stiffer, 
5-28 

less compressible column, results in decreased settlement for the entire stone-column 

foundation. 

4.1.2 Design Procedure

Stone columns are typically selected to increase bearing resistance, reduce settlement, 

accelerate consolidation time rate, increase shear strength, reduce liquefaction potential, or 

provide any combination of the above. 

Preliminary design methods and assumptions to achieve the desired end result are outlined in 

this section.

The generalized design process for embankment support is as follows: 

1. Perform embankment design without stone columns to determine the overall 

settlement and global stability to determine if stone columns or another form of 

ground modification are required. If yes proceed to step 2. 

2. Assume an area replacement ratio and column diameter.

3. Determine the spacing based on the assumed area replacement ratio and column 

diameter. 

4. Check the load bearing resistance of the stone column to see if it meets the project 

requirements. If not revise the column diameter and re-check.

5. Determine the total settlement of the embankment supported on the stone columns. 

6. Check the time rate of settlement. If the time for settlement is too large consider 

changing the column spacing. 

7. Check global stability.

4.1.2.1 Unit Cell Concept 

For purposes of settlement and stability analyses, it is convenient to associate the tributary 

area of soil surrounding each stone column with the column, as illustrated in Figures 5-16 

and 5-17. 
5-29 

FHWA 1983

Figure 5-16. Equilateral triangular pattern of stone columns. 
5-30 

FHWA 1983

Figure 5-17. Unit cell idealization.

Although the tributary area forms a regular hexagon about the stone column, it can be closely 

approximated as an equivalent circle having the same total area. The resulting equivalent 

cylinder of material having a diameter De enclosing the tributary soil and one stone column is 

known as the unit cell. The stone column is concentric to the exterior boundary of the unit 

cell. 

4.1.2.2 Area Replacement Ratio

The volume of soil replaced by stone columns has an important effect upon the performance 

of the improved ground. To quantify the amount of soil replacement, the Area Replacement 
5-31 

Ratio, αsc, is defined as the fraction of soil tributary to the stone column replaced by the 

stone: 

A

Asc α sc = [Eq. 5-1] 

where “Asc” is the area of the stone column after compaction and “A” is the total area within 

the unit cell. Typical ratios used are in the range of 0.10 to 0.30. The literature also describes 

the ratio asc, the area improvement ratio, which is the inverse of an area replacement ratio.

4.1.2.3 Spacing and Diameter

Stone column diameters vary between 1.5 and 4 feet, but are typically in the range of 3.0 to 

3.6 feet for the dry method, and somewhat larger for the wet method. 

Triangular, square, or rectangular grid patterns are used, generally with center-to-center 

column spacing of 5 to 12 feet. For footing support, they are installed in rows or clusters. For 

both footing and wide area support, they should extend beyond the loaded area. 

4.1.2.4 Stress Ratio

The relative stiffness of the stone column to the in situ soil, as well as the diameter and 

spacing of the columns, determines the sharing of the imposed area vertical load between the 

column and the in situ soil.

Since the deflection in the two materials is approximately the same, equilibrium 

considerations indicate the stress in the stiffer stone column must be greater than the stress in 

the surrounding soil. The assumption of equal deflection is frequently referred to as an equal 

strain assumption, which both field measurements and finite element analyses have indicated 

to be valid. 

The stress concentration or stress ratio n, defined as the stress in stone column divided by the 

in situ soil stress, is dependent upon a number of variables, including the relative stiffness 

between the two materials, length of the stone column, area ratio and the characteristics of 

the granular blanket placed over the stone column. Measured values of stress ratio have 

generally been found to be between 2.0 and 5.0, and theory indicates this concentration factor 

should increase with time. Since secondary settlement in reinforced cohesive soils is greater 

than in the stone column, the long-term stress in the stone column could be larger than at the 

end of primary settlement.
5-32 

For preliminary design, the determination of a design stress ratio is the key element in stone 

column design; and, unfortunately, it is based largely on experience, although theoretical 

solutions are available.

A high stress ratio (3 to 4) may be warranted if the in situ soil is very weak and the column 

spacing very tight. For stronger in situ soils and large column spacings, lower bound stress 

ratios (2 to 2.5) are indicated. For preliminary design, a ratio of 2.5 is often conservatively 

used for stability and bearing resistance calculations.

Once a stress ratio has been assumed or determined, the stress on the stone column, σsc, and 

on the surrounding soil, σsoil, can be calculated for each replacement ratio, αsc, and any 

average stress condition, q, that would exist over the unit cell as follows: 

soil

sc n

σ

σ

=

[Eq. 5-2] 

For equilibrium of vertical forces for a given asc 

( ) ( ) scsc soil sc q 1−+= ασασ [Eq. 5-3] 

For a given stress concentration ratio, the stress on the unimproved soil is: 

[ ] ( ) sc

soil

n

q

α

σ

−+ 11 =

[Eq. 5-4] 

and on the stone column: 

1 1 ( ) sc

sc

nq

n

σ

α =

  + −   [Eq. 5-5]

4.1.2.5 Stone Column Vertical Load Capacity

In determining the ultimate load capacity of a stone column or a stone column group, the 

possible modes of failure to be considered are illustrated in Figures 5-18 to 5-20. 
5-33 

FHWA 1983

Figure 5-18. Failure modes of a single stone column in a homogenous soft layer. 

FHWA 1983

Figure 5-19. Failure modes of a single stone column in a non-homogeneous cohesive soil.
5-34 

FHWA 1983

Figure 5-20. Failure modes of stone column groups.
5-35 

Caution should be given to avoiding local bulging failures due to very weak or organic layers 

of limited thickness (Figure 5-19). Bulging would have an effect on the time rate and 

magnitude of settlement, and may be of concern with respect to stability and stone column 

shear strength. Use of a bulging analysis for a single column to predict group behavior gives 

an approximate conservative solution. 

The rational prediction of the bearing resistance of stone column groups loaded by either a 

rigid foundation or a flexible load, such as an embankment, is still in the development stage. 

As a result, past experience and engineering judgment should be used in addition to theory 

when selecting a design stone column load. 

Frequently, the ultimate capacity of a stone column group is predicted by multiplying the 

single column capacity by the number of columns in the group. Small-scale model studies 

using a rigid footing indicate this approach is probably slightly conservative for soft cohesive 

soils. The bearing resistance of an isolated stone column or a stone column located within a 

group can be expressed in terms of nominal bearing resistance of the stone column: 

= Ncq cn [Eq. 5-6] 

where qn, c, and Nc are the nominal bearing resistance of the stone column can carry, the 

undrained shear strength of the surrounding cohesive soil, and the bearing capacity factor for 

the stone column, respectively. Bearing capacity factors between 18 and 22 have been found 

to provide good estimates. 

Cavity expansion theory indicates that the ultimate capacity and, hence, Nc is dependent upon 

the compressibility of the soil surrounding the stone column. Hence, soils with organics or 

other soft clays would be expected to have a smaller value of Nc compared to stiffer soils. 

For soils having a reasonably high initial stiffness, an Nc of 22 is recommended; for soils 

having low stiffness, an Nc of 18 is recommended. Low stiffness soils would include peats, 

organic cohesive soils, and very soft clays with plasticity indices greater than 30. High 

stiffness soils would include inorganic soft-to-stiff clays and silts. The recommended values 

of Nc are based on a back-analysis of field test results. In this analysis, the strengths of both 

the soil and stone column were included. A resistance factor of 3 is recommended for design 

if using Equation 5-6. 

Typically, single column design loads of 40 to 60 kips can be used in soft to medium stiff 

clays.
5-36 

4.1.2.6 Settlement

Reduction of settlement is one of the improvement benefits achieved by the use of stone 

columns. The reduction of settlement has been estimated by both pseudo-elastic and elastoplastic methods, considering both isolated and wide spread loading using a unit cell concept. 

The predicted improvement, often expressed as the settlement ratio "n", defined as the ratio 

of settlement without stone columns to that with stone columns, is typically related to the 

area replacement (αsc) or area improvement (1/αsc) ratio. The settlement of the non-improved 

zone is determined by conventional settlement analyses. Improvement predictions based on 

some theoretical analytical methods, as well as results from field measurements, are shown in 

Figure 5-21 (Greenwood and Kirsch 1984). 

Wallays et al. 1983

Figure 5-21. Comparison of elastic theories and field observations. 

Han (2015) presents three methods for calculating the settlement of granular column 

reinforced foundations. The three are stress reduction method, improvement factor method, 

and elastic-plastic method.
5-37 

It should be noted in Figure 5-21 that the settlement ratio “n” was determined analytically by 

various researchers as a function of the ratio of the Modulus of the stone column (Esc) to the 

in situ soil modulus (Esoil), or a measure of the strength of the stone column (N) to the shear 

strength of the in situ soil (cu).

For preliminary estimates, the Priebe curve may be used to evaluate the upper bound 

effectiveness and cost at various spacings. It should be further noted that the Equilibrium 

Method outlined in FHWA (1983) Design and Construction of Stone Columns is roughly 

equivalent to the Balaam relationships shown in Figure 5-21 and represents an average or 

lower bound estimate suitable for preliminary analyses.

4.1.2.7 Rate of Settlement

Stone columns substantially alter the time-rate of settlement as radial drainage governs. 

Therefore, time-rate of settlement computations are identical to the computations performed 

for vertical sand drains and prefabricated vertical drains (see Chapter 2 –Prefabricated 

Vertical Drains). The effect of disturbance or smear during installation, which reduces radial 

flow, can be roughly accounted for by reducing the diameter of the column by 50 to 80 

percent of its design diameter. A larger disturbance or smear zone should be anticipated with 

the dry-displacement construction method and for all installations in sensitive clays.

4.1.2.8 Shear Strength Increase

For slope stability analyses, an average shear strength of the soil/stone column composite 

material has been used in the past to estimate the stability of the embankment. How this 

approach is used is illustrated in Figure 5-22. 
5-38 

FHWA 1983

Figure 5-22. Notation used in average stress method stability analysis. 

However, recent research has shown that the average strength approach may overestimate the 

factor of safety by 10% for an undrained condition (Zhang et al. 2014; Abusharar and Han 

2011). 

The composite strength is a function of the undrained shear strength of the in situ soil, the 

frictional resistance of the column, the area replacement ratio, the stress ratio, and the loading 

condition. For significant improvement to occur, a relatively close spacing and a substantial 

overburden pressure is necessary to mobilize the frictional strength of the column.

The average strength,τ, and average unit weight, γ , parameters can be determined as 

follows:

( ) sc u sc v sc 1 c +−= tanφσαατ [Eq. 5-7] 

( ) 1 sc sc soil sc γ γα γ α = +− [Eq. 5-8] 
5-39 

where, 

τ = average weighted shear strength

cu = undrained shear strength of in situ soil 

γ = average unit weight 

γsoil, γsc = unit weight of soil and stone column 

φsc = angle of friction for stone column 

σv = stress due to embankment loading 

αsc = area replacement ratio

For design, the angle of internal friction φs of the stone column typically used varies from 40 

to 45 degrees. The lower angles should be considered for gravel mixtures, and the higher 

angles for angular crushed stone mixtures. 

Note that in landslide remediation projects, the stress ratio is 1, and consequently the strength 

parameters are essentially a weighted average.

Stability analyses may be performed using a total stress approach by assigning φ = 0 for endof-construction conditions, or, using an effective stress approach, by assigning c = 0 for longterm conditions. A target factor of safety of 1.2 to 1.3 is considerate adequate. 

Over the last decade slope stability programs have advanced to the point where it is now easy 

to model discrete stone columns and not use the average shear strength method described 

above. For a complete description of the stability analysis methods using discrete stone 

columns in the model, see http://www.GeoTechTools.org. 

4.1.3 Seismic Design

In the United States, there has been an effort to evaluate the liquefaction potential of soils 

from in situ density data and to modify and improve the properties of these soils. 

"Quantitative Evaluation of Stone Column Techniques for Earthquake Liquefaction 

Mitigation" (Baez and Martin, 1992), Soil Improvement for Earthquake Hazard Mitigation, 

(Hryciw Editor 1995), Advances in the Design of Vibro Systems for Improvement of 

Liquefaction Resistance (Baez 1993) and Review of Verification and Validation of Ground 

Improvement Techniques for Mitigation of Liquefaction (Woeste et al. 2016) provide 

recommendations on how to quantify the benefits of ground modification using stone 
5-40 

columns and how to evaluate the actual safety factor against seismic liquefaction. The 

benefits of stone columns with respect to liquefaction mitigation are that the soil around the 

column is densified, the drainage of excess pore water is facilitated, and the stiffer (i.e., the 

stone column is stiffer than the surrounding soil) stone column accepts higher seismic stress 

than the surrounding soil. The approach presented below is a simplified procedure that only 

considers the benefit of soil densification. This approach is appropriate for preliminary 

designs and more rigorous analysis may be warranted for final design. 

4.1.3.1 Soil Density

It is well understood that under cyclic loading, pore pressure generation in a dense soil occurs 

more slowly than in loose sand. Therefore, liquefaction potential can be reduced by 

increasing soil density. For loose sands, once the state of initial liquefaction is reached, large 

ground deformations may occur due to their lower initial strength. In dense sands, when peak 

pore pressures become equal to the initial confining pressure, the larger shear strains 

mobilize significant dilation of the sand structure, thereby maintaining significant residual 

stiffness and strength. 

Densification has been used frequently for reducing the potential for liquefaction. Seed et al. 

(1985) developed empirical liquefaction curves which correlate cyclic stress ratio to 

corrected penetration resistance. The cyclic stress ratio (CSREQ) is determined as follows:





 



 



 



 = ' max 65.0 v

v EQ dr

g

a CSR

σ

σ

[Eq. 5-9] 

where,

amax = maximum ground acceleration

σv = total vertical stress at any depth z

σv′ = effective vertical stress at any depth z

rd = stress reduction factor

For preliminary designs the stress reduction factor may be estimated as a function of depth 

(z) based on the following (note that the following equations were developed for metric 

units): 

rd = 1.0 - 0.00765z for z ≤ 30 ft. 
5-41 

rd = 1.174 - 0.0267z for 30 ft. < z ≤ 75 ft. 

rd = 0.744 - 0.008z for 75 ft. < z ≤ 100 ft. 

rd = 0.5 for z < 100 ft. 

The relationship between stress ratio causing liquefaction and corrected SPT “(N1)60” values 

for silty sands for a magnitude 7.5 earthquake is shown in Figure 5-23. This figure may be 

used to estimate the required improvement in soil density to prevent liquefaction.

Seed et al. 1985. 

Figure 5-23. Relationship between stress ratio causing liquefaction and (N1)60 values for 

silty sands for 7.5 magnitude earthquakes. 
5-42 

4.1.3.2 Spacing

The spacing of the stone columns may be determined using Figure 5-23 to determine the 

corrected SPT “N” value. Use Table 4-8 from the Chapter 4 Deep Compaction to correlate 

SPT “N” values to relative density. Then use Figure 4-34 (from Chapter 4 Deep Compaction) 

to estimate the required stone column spacing to improve the soil to the required penetration 

resistance at the mid-point between columns. 

4.1.3.3 Permeability

In order to avoid significant generation of pore water pressures within the stone column, it is 

recommended that the permeability of the stone column be at least two orders of magnitude 

larger than the treated soil. This recommendation can be achieved by selection of the 

gradation for the stone column, with due regard to piping considerations outlined below. 

4.1.3.4 Piping Prevention 

There is a likelihood that hydraulic gradients may exceed critical gradients (greater than one). 

This situation may initiate a movement of fines from the natural soil into the large, open pore 

structure of the stone column during seismic loading, leading to the development of cavities 

within the soil structure and potentially undesirable volume change. In reality, due to the 

short duration of the strong motion, it is unlikely that much soil material could be carried into 

the stone column. 

Based on experimental data, the following relationship is recommended for piping prevention 

under any loading condition based on the grain size distribution of the stone column and the 

surrounding soil. Adherence to these criteria will ensure maximum permeability and prevent 

piping of the soil: 

15 15 85 20 9 GS << DDD S [Eq. 5-10] 

where DS15 is the diameter of soil particle passing 15 percent, DG15 is the diameter of gravel 

(stone) passing 15 percent, and DS85 is the diameter of soil particle passing 85 percent in a 

grain size analysis test.

4.2 Rammed Aggregate Piers

4.2.1 Design Considerations 

The design concept used for rammed aggregate piers is almost identical to that used for stone 

columns. 
5-43 

For area wide ground improvement applications, the design method is identical to that 

previously detailed for stone columns. The stone within the rammed aggregate piers having 

been compacted by impact ramming typically exhibits a somewhat higher effective friction 

angle, in the range of 45 to 50 degrees, and potentially higher stiffness (modulus). 

Consequently, the ratio of the stiffness (modulus) of the rammed aggregate piers to the 

stiffness of the in situ soil should be somewhat higher than for stone columns, resulting in a 

higher design stress ratio than previously identified for stone columns. Although the Geopier 

design manual suggests stress ratios of 20 or higher, a stress ratio of 5 to 10 for area ground 

improvement applications under flexible embankment loading appears warranted until 

considerably more field data in support of a higher ratio is developed. 

For structure foundation support under rigid footings a somewhat higher stress ratio (10) may 

be considered, with anticipated settlements and pier capacity conventionally computed, based 

on the loads on each element. The load on the rammed aggregate pier and on the in situ soil 

is based on the chosen stress and replacement ratios. The design area replacement ratio is 

determined after evaluating settlement of the unimproved soil. A minimum replacement ratio 

of 0.33 is generally recommended, as noted in the HITEC Evaluation Report (Collin 2007). 

4.2.2 Design Procedures

Rammed aggregate piers are typically selected to increase bearing resistance, reduce 

settlement, increase shear strength, or provide any combination of the above. 

Preliminary design methods and assumptions to achieve the desired end result are outlined in 

this section.

The generalized design process for an embankment support is as follows: 

1. Perform embankment design without rammed aggregate piers to determine the overall 

settlement and global stability to determine if rammed aggregate piers or another 

form of ground improvement are required. If so proceed to step 2. 

2. Assume an area replacement ratio and column diameter.

3. Determine the spacing based on the assumed area replacement ratio and column 

diameter. 

4. Check the load bearing resistance of the rammed aggregate pier to see if it meets the 

project requirements. If not revise the column diameter and re-check.

5. Determine the total settlement of the embankment supported on rammed aggregate 

piers.
5-44 

6. Check the time rate of settlement. If the time for settlement is too large consider 

changing the column spacing. 

7. Check global stability.

The design process for rammed aggregate piers is similar in many respects to stone columns. 

However, the major difference in design of the two systems is with respect to settlement 

analysis that is presented in the following sections. 

4.2.3 Settlement Analysis

Rammed aggregate pier settlement control design methodology is based on a two-layer 

settlement approach as described by Lawton et al. (1994), Fox and Cowell (1998), and 

Wissmann et al. (2002). The installation of rammed aggregate piers within the aggregate 

column-reinforced zone, referred to as the upper zone, creates a stiffened, engineered zone 

with reduced compressibility that reduces settlement of embankments and transportationrelated structures. The settlement below the rammed aggregate pier-reinforced zone, referred 

to as the lower-zone, is evaluated using conventional geotechnical analysis approaches. The 

total settlement (stot) of the transportation structures is evaluated as the sum of the upper zone 

settlement (suz) and the lower zone settlement (slz):

tot uz lz += sss [Eq. 5-11] 

4.2.3.1 Settlement in the Rammed Aggregate Pier-Reinforced Zone 

Settlement in the rammed aggregate pier-reinforced zone (upper zone) is estimated by first 

calculating the top-of-pier stress (qg) using the following equation: 





 





+− =

aas 1

s

g RRn

n

qq

[Eq. 5-12] 

where, 

q = average applied bearing pressure

Ra = ratio of the cross-sectional area coverage of the rammed aggregate 

piers to the matrix soil

ns = stress concentration ratio between the rammed aggregate piers and the 

matrix soil
5-45 

Research has shown that stress concentration ratios range from 4 to 45 for rigid footings. 

Because embankments and most MSE walls are not rigid structures, stress concentration

ratios may be lower than those observed for rigid footings and should be selected with care. 

In addition, the stress concentration ratio is related to the stiffness of the matrix soil with 

larger ratios resulting at softer soil sites. Suggested stress concentration ratios ranging from 5 

to 10 may be used for settlement control of embankments. 

The settlement of the rammed aggregate pier-reinforced zone is estimated as the top-oframmed aggregate pier stress, qg, divided by the rammed aggregate pier stiffness modulus, 

kg,: 

g

g

uz k

q

s =

[Eq. 5-13] 

Design rammed aggregate pier stiffness modulus values range from 75 pci to 360 pci for 

support of rigid footings. Conservative stiffness modulus values should be used for support 

of embankments and transportation-related structures (Collin 2007). 

4.2.3.2 Settlement below the Rammed Aggregate Pier Reinforced Zone 

Settlement below the rammed aggregate pier-reinforced zone is evaluated using conventional 

geotechnical approaches, consisting of either elastic settlement analyses or consolidation 

analyses using equation 5-14 for cohesionless or overconsolidated cohesive soils and

equation 5-15 for normally-consolidated cohesive soils 

E

Hq slz

∆=

[Eq. 5-14] 















 ∆+

















+ =

o

o

o

lz c p

qp H

e

cs log 1

1

[Eq. 5-15] 

where H is the thickness of the lower zone, E is the matrix soil elastic modulus within the 

lower zone, cc is the matrix soil coefficient of compressibility, eo is the matrix soil void ratio, 

po is the vertical effective stress at the mid-point of the compressible layer, and ∆q is the 

average bearing pressure applied by the embankment. The average applied bearing pressure 

is the product of the applied pressure and the stress influence factor, Iσ. The stress influence 

factor can be determined using either Boussinesq or Westergaard’s method. Rammed 

aggregate pier reinforced soil is typically considered a layered soil and therefore 

Westergaard’s method is typically used. However, for embankments, the stress influence 
5-46 

factor within the lower zone is typically assumed to be 1.0 because of the large lateral extent 

of embankment fills. 

Typically, elastic modulus settlement approaches are used to estimate settlement in granular 

soils and heavily over-consolidated cohesive soils. Matrix soil equivalent elastic modulus 

values may be estimated using published correlations from SPT N-values, undrained shear 

strengths, CPT tip resistances, or other in situ tests. Consolidation settlement approaches are 

used to evaluate settlement in normally-consolidated or lightly over-consolidated cohesive 

soils.

4.3 Design Examples 

4.3.1 Rammed Aggregate Piers

The following design example is provided to demonstrate the method to determine settlement 

of an embankment supported on rammed aggregate piers. 

4.3.1.1 Problem 

A new embankment is to be constructed over a soft clay layer that is underlain by rock. The 

geometry of the embankment and soil stratigraphy are shown in Figure 5-24. 

Figure 5-24. Example problem 1 geometry.

Determine the total settlement that will occur after the embankment is constructed. The 

spacing of the columns is 5 feet and the diameter of the columns is 2.75 feet. 

4.3.1.2 Total Settlement Magnitude without Ground Improvement 

zP ( ) ( ft pcf pcf ) psf o sat w γγ =−= − =4324.621205.7

γ Hq == pcf ( ) ft = 500,220125 psf
5-47 

( ) ( ) ft inches

P

qP H

e

cS

o

o

o

c 2283.1

432

2500432 log15 7.011 log 25.0

1

1  ==

 



 +

















+  =



 



















 ∆+

+ =

The total expected settlement of the embankment without ground improvement is 22 inches. 

The proposed aggregate column layout for the embankment is shown in Figure 5-25. 

Figure 5-25. Aggregate column ground improvement layout. 

Determine the anticipated amount of settlement with the rammed aggregate piers.

4.3.1.3 Settlement Magnitude with Rammed Aggregate Piers

( ) 27.0

25.5

94.5

2 == = ft

ft

a

A

R g

a

sd ( )=== 25.5505.105.1 e

γ Hq == pcf ( ) ft = 500,220125 psf

( ) psf psf RRn

n

qq

aas

s

g 383,6 127.027.06

6 500,2 1 =



 





+−  =



 





+− =

( )

inches

ft

in pci

psf

k

q

s

g

g

uz 68.0

14465

383,6

2

2

=





 



 ==

4.3.2 Stone Columns

4.3.2.1 Problem 

A new embankment is to be constructed over a soft clay layer that is underlain by dense sand 

with and an average N160 = 48. The geometry of the embankment and soil stratigraphy are 

shown in Figure 5-26. 
5-48 

Figure 5-26. Example problem 2 geometry and soils. 

Determine the total settlement that will occur after the embankment is constructed. The 

spacing of the columns is 5.7 feet and the diameter of the columns is 3.0 feet. Assume that no 

settlement on the dense sand will occur. 

4.3.2.2 Total Settlement Magnitude without Ground Improvement 

zP ( ) ( ft pcf pcf ) psf o sat w γγ =−= − = 440,14.6212025

γ Hq == pcf ( ) ft = 875,115125 psf

( ) ( ) ft inches

P

qP H

e

cS

o

o

o

c 2726.2

440,1

875,1440,1 log50 6.011 log 2.0

1

1  ==

 



 +

















+  =



 



















 ∆+

+ =

4.3.2.3 Settlement Magnitude with Stone Columns

( )

( ) 6.3

0.3

7.5

2

2

== Asc

A

Using the Priebe curve from Figure 5-27, determine the settlement ratio. 
5-49 

After Wallays et al. 1983

Figure 5-27. Example problem 2 settlement ratio determination.

The settlement ratio is 2.7. Therefore, the settlement of the embankment using stone columns 

as ground improvement is 10 inches. 

4.4 Design Verification

As an important adjunct to design, a field verification program of load tests and in situ testing 

must be developed and implemented through appropriate construction specification 

requirements. A program should be specified, regardless of the contracting method. 

A combination of load tests on aggregate columns constructed before, during, and after 

production should be specified to verify the design assumptions and the performance 

specification. There are three types of load tests: (1) short-term tests, which are used to 

evaluate ultimate stone column bearing resistance, (2) long-term tests, which are used to 

measure the consolidation settlement characteristics; and (3) horizontal or composite shear 

tests, which are used to evaluate the composite aggregate-soil shear strength for use in 
5-50 

stability analyses. The most common of these tests is the short-term load test on a single 

column. 

The short-term load tests, similar to pile load tests, should be performed after all excess pore 

pressures induced during construction have been dissipated. The load increment should 

closely correspond to the actual loading. For example, if the actual foundation load will be 

applied very slowly, a load increment of approximately 10 percent of the ultimate should be 

used. A rapid loading may result in immediate settlement, as well as consolidation 

settlement. If the actual load will be applied rapidly, a load increment of 20 to 25 percent of 

ultimate should be used. The tests are generally performed to 150% of the design load, and 

the measured settlement is compared to project settlement tolerance. For example, a final 

acceptance criterion of 1 inch of settlement at 150 to 200 percent of the design load appears 

to be a reasonable criterion for columns supporting a structure. 

The long-term settlement of the stone column foundation is usually estimated from the 

results of short-term load tests on single stone columns. Mitchell (1981) reported that the 

foundation settlement due to a uniform loading of a large area was 5 to 10 times greater than 

the settlement measured in a short-term load test on a single column. However, there is very 

little field data available to confirm this behavior. Therefore, it is recommended that longterm load tests on a group of columns be conducted in conjunction with short-term load tests 

to develop an estimate of the settlement of the stone column foundation. The long-term load 

tests should be conducted on a minimum of three to four stone columns located within a 

group of 9 to 12 columns having the proposed spacing and pattern. The load should be 

applied over the tributary area of the columns and left in place until the cohesive soil reaches 

a primary degree of consolidation of 90-95 percent. The applied load could consist of column 

backfill material, native material, and/or the dead weight from the short-term load tests. The 

results of these tests will provide valuable information for estimating the ultimate settlement 

of the stone column foundation. 

During the production phase of construction, a few short-term load tests can be performed for 

quality control purposes. These tests are referred to as proof tests and are used to verify 

quality control during production. The load applied in the proof test is usually 150 to 200 

percent of the allowable/design load.

In situ testing to evaluate the effect of the stone column construction on the native cohesive 

soil can be also specified. However, the specified test method should be selected on the basis 

of its ability to measure changes in lateral pressure in cohesive soils. The cone penetrometer 

(CPT), the flat plate dilatometer (DMT), and the pressuremeter (PMT) appear to provide the 

best means for measuring the change, if any, in lateral stress due to stone column 

construction. Due to the limited amount of information that will be obtained from CPT, 
5-51 

DMT, or PMT testing after column construction, it is recommended that long-term load tests 

on groups of stone columns be conducted instead of in situ tests. However, extensive in situ

testing should be conducted during the initial subsurface investigation to reliably estimate the 

soil profile and the stone column design parameters. 

For rammed aggregate pier construction, a Modulus test and a Bottom Stabilization test have 

been developed and are used as quality assurance checks. For details, consult the HITEC 

Evaluation Report (Collin 2007). 
5-52 

5.0 CONSTRUCTION SPECIFICATIONS AND QUALITY ASSURANCE

Like other methods of specialty construction, unless the specifying agency has expertise in 

the design, construction, and inspection of aggregate columns, it is good practice to specify 

that the work be accomplished under a performance type specification. If the specifying 

agency has the necessary experience with the aggregate column technique, a method 

specification may be utilized.

5.1 Aggregate Column Performance Specification

As part of the development of GeoTechTools, an extensive evaluation was made of 

specifications for aggregate columns. Twenty-one specifications written by state DOTs and 

other agencies were reviewed and evaluated. Of the assessed specifications, two

specifications were only applicable to rammed aggregate piers, 14 specifications were only 

applicable to stone columns, and five specifications were applicable to both rammed 

aggregate piers and stone columns. These specifications were used to develop a guide 

specification entitled Guide Specification for Aggregate Columns that is intended to be a 

complete and fair specification containing commentary and instructions that are easily 

adaptable by the user for a specific project. This guide specification can be accessed at

http://www.GeoTechTools.org under the Aggregate Columns Technology Information page 

and is applicable to both stone columns and rammed aggregate piers). 

An outline of the current Guide Specification for Aggregate Columns, illustrating what items 

should be contained in such a specification, follows. 

PART 1 GENERAL

1.01 INTRODUCTION

1.02 INTENT 

1.03 STANDARDS AND REFERENCES

1.04 DEFINITIONS

1.05 SCOPE OF WORK 

1.06 SUBMITTALS

1.07 QUALIFIED CONTRACTORS

1.08 QUALITY ASSURANCE
5-53 

PART 2 EQUIPMENT AND MATERIALS

2.01 EQUIPMENT

2.02 BACKFILL MATERIALS

PART 3 EXECUTION

3.01 SITE INSPECTION

3.02 AGGREGATE COLUMN CONSTRUCTION

3.03 PERFORMANCE CRITERIA

3.04 FIELD QUALITY ASSURANCE

3.05 REJECTION OF AGGREGATE COLUMNS

3.06 EXCAVATION OF COLUMNS TOPS, AND UTILITIES 

3.07 SUBGRAD PREPARATION 

3.08 RESTRICTIONS

PART 4 PAYMENT

4.01 METHOD OF PAYMENT

5.2 Field Inspection and Improvement Verification

Verification and detailed field inspection of aggregate column construction is a very 

important, but often neglected, aspect. Thorough field surveillance by both the Engineer and 

Contractor is essential in the construction of aggregate columns. Furthermore, good 

communication should be maintained at all times between the inspection personnel, 

Contractor, Project Engineer and Designer.

5.2.1 Stone Columns

A comprehensive stone column Quality Assurance (QA) assessment program usually 

consists of several QA methods. Gradation, specific gravity, loose density, and compacted 

density tests should be run on the stone to be installed, with a frequency of one test for each 

5,000 tons of material prior to construction to ensure compliance with specifications. Stone 

column performance is dependent upon the integrity of the column. It is important that the 
5-54 

minimum column diameter and required compacted density of the stone be achieved in order 

to ensure the desired performance. During construction, stone consumption, in terms of 

buckets of a known weight or volume, should be monitored as a function of depth. Based on 

the loose and in-place, compacted density of the stone, it is possible to estimate the column 

diameter. Barksdale and Bachus (FHWA 1983) provide a method for estimating the in-place 

density of the stone based on loose and compacted density tests. Measurements should 

typically be taken at a maximum of 5-foot increments to determine the column’s crosssectional area profile versus depth. Decreased rate of stone consumption may indicate caving 

of the hole or failure to attain adequate displacement and replacement of the surrounding 

ground. For any group of 50 consecutively installed stone columns, the average diameter 

over the total length should not be less than as specified in the contract documents. No stone 

column should have a diameter less than 90% of the minimum diameter specified in the 

contract documents. Verticality of the rig should be monitored, and no stone column axis 

should be inclined from the vertical by more than 2 inches in 10 feet. During construction of 

the column, each lift should be re-penetrated until the specified amp-meter reading is 

achieved, thus indicating good input energy from the vibrator probe to the stone. In general, 

it is recommended that, as a minimum, the vibrator free-standing current reading plus at least 

40 additional amps be developed. 

For projects requiring the improvement of large areas, it is desirable to subdivide the total 

area into approval or acceptance zones on the order of 100 feet on a side. Completing the 

work with timely approval on a zone-by-zone basis means that the contractor may proceed 

without risk of having to return late in the project to correct deficiencies that developed early 

in the project. 

All construction records should be furnished to the engineer, with the following data to be 

obtained during column installation: 

• Stone column reference number 

• Measurement of rig verticality 

• Elevation of top and bottom of each stone column 

• Number of buckets of stone backfill in each stone column 

• Amperage achieved as a function of depth; the date and column identification should 

be written on each record 

• Time to penetrate and time to form each stone column 

• Details of obstructions, delays, and any unusual ground conditions 

• Digital data log of amperage, depth, and stone consumption 
5-55 

Post-construction QA is dependent on the specific application and the type of ground in 

which the stone columns are installed. For slope stabilization, structure or embankment 

support, settlement reduction, liquefaction mitigation, and prevention of lateral spreading 

applications in silty and clayey sands where densification is required, in situ testing (SPT, 

CPT, or PMT) should be conducted at central points between the columns. Penetration 

resistance should be verified against values that were used to determine column spacing. The 

same test method should be utilized both before and after the stone column installation to 

verify soil improvement. 

Stone column installation is not expected to induce densification of soft, saturated clays. If 

the columns are to support a structure or embankment in such soils, load tests are sometimes 

required to determine the short-term capacity and settlement of the column. Short-term load 

tests should be conducted in accordance with ASTM D1143, Standard Test Methods for 

Deep Foundations Under Static Axial Compressive Load, on individual columns after all 

pore pressures induced by construction have dissipated. If settlement is a primary concern, 

longer-term load tests are highly recommended, with settlement readings generally taken 

over a one-week period. The longer-term load tests should be conducted on a minimum of 

three to four stone columns located within a group of nine to 12 columns having the proposed 

spacing and pattern. The load should be applied over the tributary area of the columns and 

may consist of column backfill material, native material, and/or the dead weight from the 

short-term load tests. Concrete blocks and reaction pile systems may also be used for load 

testing of single columns. Surveying methods should be used to ensure proper column 

spacing and location. No column should be more than 4 inches from the specified center 

location unless an obstruction is encountered. In case of an obstruction, the Engineer should 

be notified to determine the maximum allowable offset. Gradation analyses on samples taken 

from installed columns may be used to confirm that the in situ gradation matches the 

specifications and that the columns have not been penetrated by excessive amounts of fines 

from the surrounding ground. Such testing may be appropriate for the owner's information in 

a method specification, but columns cannot be rejected for failing to meet a post-installation 

gradation criterion if other provisions of a method specification have been followed. 

5.2.2 Rammed Aggregate Columns 

A comprehensive rammed aggregate pier QC/QA assessment program usually consists of 

several QC/QA methods. It is the responsibility of the QC representative to coordinate with 

the General Contractor on footing layout and pier elevations, observe installation procedures, 

ensure the aggregate moisture content is within acceptable limits, perform tests on production 

piers, and implement corrective measures when necessary. The Bottom STAbilization test 

(BSTA) is used to verify piers have an adequate stabilized bottom (Collin 2007). It involves 

re-tamping the bottom of the piers to verify that displacement is within acceptable limits. A 
5-56 

pattern of successful BSTA tests is sufficient to reduce BSTA verification to spot checks 

(Fox and Cowell 1998). The Dynamic Cone Penetrometer (DCP) is used in general 

accordance with ASTM STP 399 Vane Shear and Cone Penetration Resistance Testing of In 

situ Soils to verify aggregate densification within the top few feet of the pier. If average 

penetration resistance measured consistently exceeds 15 blows, and less than 10% of tests 

fall below 15 blows per 1.75 inches, then testing may be reduced to spot checks (Fox and 

Cowell 1998). Modulus testing is used to verify stiffness modulus design assumptions and is 

based largely on ASTM D1143 Standard Test Methods for Deep Foundations Under Static 

Axial Compressive Load. Typically, one stiffness modulus test is conducted per project site 

for small projects. On larger projects, between two and four stiffness modulus tests may be 

conducted. As a general rule, one stiffness modulus test is performed per 1,000 piers (Collin 

2007). Uplift tests are conducted when necessary to verify the performance of piers in 

tension. They are largely based on ASTM D3689 Standard Test Methods for Deep 

Foundations Under Static Axial Tensile Load and generally follow the same load and holding 

criteria as the modulus test. Often, it is possible to conduct an uplift load test at the same time 

as the modulus load test, since uplift pier elements are generally used as anchor reactions for 

the modulus test load frame. All loading and test procedures are available in Collin (2007). 

Surveying methods should be used to verify pier locations. The center of each pier should be 

within four inches of the plan location. 

Included in the QC procedures should be the completion of daily reports during installation, 

which include the following information: 

• Footing and pier location 

• Pier length and drilled diameter 

• Planned and actual pier elevations at the top and bottom of the element 

• The number of lifts and time of tamping for each lift placed 

• Average lift thickness for each pier 

• Documentation of soil conditions during drilling for comparison with soil conditions 

in boring logs 

• Depth to groundwater, if encountered 

• Documentation of any unusual conditions encountered (e.g., sloughing) 

• Type and size of densification equipment used. 

QA procedures include monitoring installation of modulus and uplift load test piers, 

monitoring load tests, performing DCP testing, and monitoring daily pier installation, 

including observing subsurface conditions and soils during installation. Gradation analyses 
5-57 

on samples taken from installed columns may be used to confirm that the in situ gradation 

matches the specifications and that the columns have not been penetrated by excessive 

amounts of fines from the surrounding ground. Such testing may be appropriate for the 

owner's information in a method specification, but columns cannot be rejected for failing to 

meet a post-installation gradation criterion if other provisions of a method specification have 

been followed. 

5.2.3 Verification Testing

The testing of soils reinforced by aggregate columns should address the different response of 

the ground when testing granular soils in comparison to predominantly cohesive soils. In situ 

tests are more appropriate where densification of the in situ soil is anticipated. Load tests are 

also appropriate for these soils, as well as mixed and cohesive soil profiles. Guidance on the 

usefulness of certain commonly performed test methods (Esrig and Bachus 1991) is 

presented in Table 5-1. 

Table 5-1. Suitability for Testing Aggregate Columns

Test Granular Cohesive Comments

Dynamic Cone 2 1 Too insensitive to reveal clay lenses. Can 

locate dense layers and buried features.

Mechanical Cone 3 1 Rarely used.

Electric Cone 4 2 

Particle size important. Can be affected by 

lateral earth pressures generated by 

treatment. Best test for seismic 

liquefaction evaluation. 

Boreholes + SPT 3 2 Efficiency of test important. Recovers 

samples.

Dilatometer 3 1 Rarely used.

Pressuremeter 3 1 Rarely used.

Small Plate

Load Test 1 1 Does not adequately confine stone column. Affected by pore water pressures.

Large Plate

Load Test 2 2 Better confining action.

Zone Loading 4 4 Best test for realistic comparison with 

foundations. 

Full-Scale 5 5 Rare

Note: Suitability ranking varies from 1 as least suitable to 5 as most suitable.

Short duration tests on of 2-foot diameter (small plates in Table 5-1) metal plates are the 

most common form of testing aggregate columns in Great Britain. This is due to their speed 

and low cost. However, such tests can only stress the soils to shallow depths and have been 

susceptible to misinterpretation of actual aggregate column behavior, particularly when 
5-58 

residual porewater pressures are present in the ground. To partially get around these obstacles 

large diameter plate test where the diameter of the plate is equal to the diameter of the 

column are typically used in the United States. 

To overcome these limitations, and to provide more realistic simulation of applied loads, 

zone loading or dummy footing tests are occasionally performed. Here, loadings of up to 3 

times the design bearing pressure are applied over a group of aggregate columns, typically of 

4 to 9 in number. Significant expense is involved with these tests. As a result, these tests tend 

to be performed on larger contracts or where the soil profile is variable; in combination with 

plate tests to permit correlation between individual aggregate columns and group 

performance.

It is important that the loaded area be of sufficient dimension and magnitude to induce 

significant stress into the “critical layer.” This stratum is normally the weakest cohesive layer 

of significant thickness. This layer determines the allowable load of the aggregate column. 
5-59 

6.0 COST DATA

This section presents guidelines for preparing budget estimates in order to determine the 

economic feasibility of aggregate columns. There are many factors affecting the price of 

aggregate column construction, including labor, the price and availability of stone, weather, 

environment, etc. Therefore, it is recommended that experienced contractors with a record of 

installing aggregate columns be contacted to verify both the budget cost calculations and the 

technical feasibility of aggregate column installation.

The costs of aggregate columns on a highway project are typically captured in a contract bid 

item which is measured by the lineal foot (LF). Included in this bid item are the material, 

equipment, labor, and incidentals to construct an aggregate column. Mobilization associated 

with the installation of aggregate columns may be measured and paid for separately. 

Construction cost items that are associated with aggregate columns, along with approximate 

cost ranges, are listed in Table 5-2. 
5-60 

Table 5-2. Unit Costs

Cost ranges are based on data (i.e., review of State DOT’s bid tabs for aggregate columns) 

from 2007 through 2010. Readers should carefully examine the project characteristics and 

constraints and determine to what degree, if any, these factors may influence the actual cost 

associated with constructing aggregate columns. For many aggregate column applications, a 

working platform will be required. These costs should be included when comparing this 

technology with others. The cost of the geosynthetic for the working platform is also 

provided in this table. 

Pay Item 

Description

Quantity 

Range Unit

Low Unit 

Price

High Unit 

Price

Factors that May 

Impact Costs

Aggregate 

Columns 

Greater 

than 

1,000 

LF $20.00 $60.00 

Cost of aggregate materials 

is sensitive to material 

specification and haul 

distance.

Unit costs will decrease as 

total quantity increases.

Typical price range is $20 to 

$40 per lineal foot. 

Mobilization 1 Rig $20,000 $100,000 

Mobilization cost increases 

for distances greater than 

500 miles. 

Phased construction may 

require multiple 

mobilizations.

High price for rigs for 

moderate depth treatment is 

$40,000. 

Embankment 

Greater 

than 

5,000 

CY

Use 

agency 

data

Use 

agency 

data

Use historical costs that are 

representative of the project 

quantity, project conditions 

and project location 

Working 

Platform 

Geosynthetic 

Greater 

than 

5,000 

SY $1.00 $3.50 

Geogrids are more 

expensive than geotextiles

Heavier geotextiles cost 

more

Specified overlap widths 

impact the total quantity of 

material required.

Granular Fill 

Material 

Greater 

than 

2,500 

Tons $7.00 $20.00 Material specification and 

haul cost will impact costs
5-61 

Using the information in the preceding sections, a determination can be readily made as to 

the depth of the aggregate column installation and the spacing required to satisfy the design 

intent. The area of treatment should take into consideration the effect of the proposed loading 

on the soil being improved by aggregate columns. It is recommended that for aggregate 

column installation the loads be considered as being transmitted on a 45-degree angle around 

the specified treatment zone perimeter. This will extend the area that requires improvement.

A spreadsheet is available at http://www.GeoTechTools.org for performing preliminary 

budgets for aggregate columns. 
5-62 

7.0 CASE HISTORIES

Representative case histories of transportation-related construction projects are presented to 

illustrate the application of rammed aggregate pier and stone column technologies. 

7.1 Rammed Aggregate Piers Case History

7.1.1 Basic Information 

• Project Name: US 90 at SH 6

• Project Location: Sugarland, Texas 

• Owner: Texas Department of Transportation

• Engineers: Geotechnical Engineer – HVJ Associates | Structural Engineer: Chiang 

Patel & Yerby

• Contractor: W.W. Webber, Inc. 

• Year Constructed: 2006 

7.1.2 Project Summary

This project consisted of ground improvement for support of several MSE walls located at 

the US90 and SH6 interchange. This was the first MSE wall application that was supported 

by the rammed aggregate piers that was monitored and instrumented by FHWA’s Houston 

office. The instrumented MSE wall had a maximum height of 27 feet.

7.1.2.1 Subsurface Conditions 

The soil conditions consisted of soft to medium stiff clay to 30 feet below ground surface, 

underlain by sandy silt to silty sand from 30 to 40 feet, overlying sand to silty sand to the 

maximum explored depth of 60 feet below ground. 

7.1.2.2 Technology Used

Rammed aggregate piers provided a cost-effective solution for this MSE wall project saving 

clients 20 to 50% compared to traditional deep foundation alternatives. Using rammed 

aggregate piers to reinforce good to poor soils, this ground modification technique allows for 

visible inspection of the spoils, and the opportunity to address changing ground conditions as 

they happen. It is an effective replacement for massive over-excavation and replacement or 

deep foundations, including driven piles, drilled shafts or auger cast-in-place piles. The 

rammed aggregate pies are constructed by applying direct vertical ramming energy to 

densely compact successive lifts of high quality crushed rock to form high stiffness 
5-63 

engineered elements. The vertical ramming action also increases the lateral stress and 

improves the soils surrounding the cavity, which results in foundation settlement control and 

greater bearing pressures for design. Vertical impact ramming results in high density and 

high strength columns providing superior support capacity, increased bearing pressure up to 

10,000 psf and excellent settlement control. 

7.1.2.3 The Construction Process 

The unique installation process utilizes pre-augering and vertical impact ramming energy to 

construct rammed aggregate piers, which exhibit high strength and stiffness. The process first 

involves drilling a cavity. Drill depths normally range from about five to 30 feet, depending 

on design requirements. Pre-drilling allows you to see the soil between the borings, ensuring 

that the piers are engineered to reinforce the right soils. Layers of aggregate are then 

introduced into the drilled cavity in lifts (Figure 5-28). 

Courtesy Geopier Foundation Company

Figure 5-28. Rammed aggregate pier installation. 

A patented beveled tamper rams each layer of aggregate using vertical impact ramming 

energy, resulting in high strength and stiffness. The ramming action densifies aggregate 

vertically and forces aggregate laterally into cavity sidewalls. This results in excellent 

coupling with surrounding soils and reliable settlement control. 
5-64 

7.1.2.4 Cost Information 

The total contract value was $751,946. 

7.1.2.5 Solution 

A total of 1,411 rammed aggregate piers with spacing that ranged from four to nine feet oncenter were installed beneath wall heights of 14 feet or greater (Figure 5-29). 

Courtesy Geopier Foundation Company

Figure 5-29. Completed MSE wall supported on rammed aggregate piers.

As a result the factors of safety for bearing resistance instability and global stability were 

increased to greater than 2.0 and 1.3, respectively as well as allowing rapid pore water 

pressure dissipation by radial drainage into the columns. Horizontal displacement at the base 

of the walls was measured to be less than one and a half inches. The modulus test results 

showed a total movement of 0.69 inches at a stress of more than 22,000 psf, indicating a pier 

stiffness greater than twice the assumed design value. 
5-65 

7.2 Stone Columns Case History 

7.2.1 Basic Information 

• Project Name: Route 22 

• Project Location: Wadhams, NY 

• Year Constructed: 1987 

7.2.2 Resources

Sung, J.T. and Ramsey, I.S. (1988). Slope Stabilization by Stone at Wadhams, NY. Report by 

Soil Mechanics Bureau, New York State Department of Transportation, State Campus, 

Albany, NY. 

7.2.3 Project Summary

Stone columns were used to stabilize a 220 foot long slope along New York Route 22 near 

Wadhams, NY. 

7.2.3.1 Subsurface Conditions 

Three meter thick layer of silty clay overlying a 10- to 20-foot layer of over-consolidated, 

soft silty clay. This clay layer is underlain by a layer of silty gravel in which artesian 

groundwater conditions were encountered. The liquidity index and activity of the clay were 

1.0 and 0.5, respectively. 

7.2.3.2 Technology Used

A stabilizing berm, shear key, and stone columns were considered. Berm treatment would 

require additional right-of-way in a wetland area, and shear key would require extensive 

excavation. Stone columns installed by the dry, bottom feed methods were found to be the 

most technically feasible, environmentally acceptable, and economic solution. 

7.2.3.3 The Construction Process 

The stone columns were installed through the soft clays into the gravel layer to intercept the 

slip plane near the gravel/clay interface at a depth of 16 feet. A photograph during 

installation is shown in Figure 5-30. 
5-66 

Sung and Ramsey 1988

Figure 5-30. Stone column installation. 

7.2.3.4 Performance Monitoring

Prior to construction, slope movement was measured at approximately 1/32-inch per day. 

During installation, the total movement was 1/8-inch. Eight years after the project 

completion; little to no additional movement had been recorded. 
5-67 

8.0 REFERENCES

Abusharar, S. and Han, J. (2011). Two-dimensional Deep-seated Slope Stability Analysis of 

Embankments over Stone Columns. Engineering Geology, 120: pp. 103-110. 

Baez, J.I. and Martin, G.R. (1992). Quantitative Evaluation of Stone Column Techniques for 

Earthquake Liquefaction Mitigation. Proc. Tenth World Conference on Earthquake 

Engineering, A.A. Balkema, Brookfield, VT, pp. 1477-1483. 

Baez, J.I. (1993). Advances in the Design of Vibro Systems for the Improvement of 

Liquefaction Resistance. Proc. Symposium on Ground Improvement, Vancouver, 

British Columbia.

Collin, J.G. (2007). Evaluation of Rammed Aggregate Piers by Geopier Foundation 

Company Final Report, Technical Evaluation Report prepared by the Highway 

Innovative Technology Evaluation Center, ASCE, 86p. 

Dobson, T. (1987). Case Histories of the Vibro Systems to Minimize the Risk of 

Liquefaction. Soil Improvement – A Ten Year Update, Welsh, J.P., Editor,

Geotechnical Special Publication No. 12, ASCE, New York, NY, pp. 167-183. 

Esrig, M.I. and Bachus, R.C., Editors. (1991). Deep Foundation Improvements: Design, 

Construction, and Testing. Proc. Symposium on Design, Construction, and Testing of 

Deep Foundation Improvements: Stone Columns and Other Related Techniques. 

Special Technical Publication 1089, ASTM, Philadelphia, PA, 337p. 

FHWA. (1983). Design and Construction of Stone Columns. Authors: Barksdale, R.D. and 

Bachus, R.C., FHWA/RD-83/026, Federal Highway Administration, U.S. DOT, Vol I 

and Vol II. 

Fox, N.S. and Cowell, M.J. (1998). Geopier Foundation and Soil Reinforcement Manual, 

Geopier Foundation Company, Inc., Scottsdale, AZ.

Greenwood, D.A. and Kirsch, K. (1984). Specialist Ground Treatment by Vibratory and 

Dynamic Methods. State-of-the-Art Report, Piling and Ground Treatment, Thomas 

Telford Ltd., London, UK, pp. 17-45. 

Han, J. (2015). Principles and Practice of Ground Improvement. John Wiley & Sons, 

Hoboken, NJ, 418p. 

Hryciw, R.D., Editor. (1995). Soil Improvement for Earthquake Hazard Mitigation. ASCE 

Geotechnical Special Publication No. 49, ASCE, New York, NY, 141p. 
5-68 

Kitazume, M. (2005). The Sand Compaction Pile Method. CRC Press, Boca Raton, FL, 232p. 

Lawton, E.C., Fox, N.S., and Handy, R.L. (1994). Control of Settlement and Uplift of 

Structures Using Short Aggregate Piers. Use of In Situ Tests in Geotechnical 

Engineering, Clemence, S.P., Editor, Geotechnical Special Publication No. 6, ASCE, 

New York, NY, pp. 121-132. 

Mitchell, J.K. (1981). Soil Improvement: State-of-the-Art. Proc. 10th International 

Conference on Soil Mechanics and Foundation Engineering, Stockholm, Sweden, 

Vol. 4, pp. 509-565. 

Priebe, H.J. (1995). The Design of Vibro Replacement. Ground Engineering, 28(10): pp. 31-

37. 

Seed, H.B., Tokimatsu, K., Harder, L.F., and Chung, R.M. (1985). Influence of SPT 

Procedures in Soil Liquefaction Resistance Evaluations. Journal of Geotechnical 

Engineering, ASCE, 111(12): pp. 1425-1445. 

Sung, J.T. and Ramsey, I.S. (1988). Slope Stabilization by Stone at Wadhams, NY. Report by 

Soil Mechanics Bureau, New York State Department of Transportation, State 

Campus, Albany, NY. 

Wallays, M., Dalapierre, J., and Van Den Poel, J. (1983). Load Transfer Mechanism in Soil 

Reinforced by Stone or Sand Columns. Eighth European Conference on Soil 

Mechanics and Foundation Engineering, Helsinki, Finland, pp. 313-317. 

Wissmann, K.J., FitzPatrick, B.T., White, D.J., and Lien, B.H. (2002). Improving Global 

Stability and Controlling Settlement with Geopier® Soil Reinforcing Elements. Proc. 

4th International Conference on Ground Improvement Techniques. Kuala Lumpur, 

Malaysia.

Woeste, D., Green, R., Rodrigues-Marek, A., and Ekstrom, L. (2016). A Review of 

Verification and Validation of Ground Improvement Techniques for Mitigation of 

Liquefaction. CGPR Report No. 86, Center for Geotechnical Practice and Research, 

Virginia Polytechnic Institute and State University, Blacksburg, VA. 

Zhang, Z., Han, J., and Ye, G. (2014). Numerical Investigation on Factors for Deep-seated 

ASlope Stability of Stone Column-supported Embankments over Soft Clay. 

Engineering Geology, 168C: pp. 104-113.