U.S. Department of Transportation Publication No. FHWA NHI-06-088
 

Federal Highway Administration December 2006 

NHI Course No. 132012_______________________________ 

SOILS AND FOUNDATIONS 

Reference Manual – Volume I 

TTeessttiinngg

EExxppeerriieennccee

TThheeoorryy

National Highway Institute
 
NOTICE 

The contents of this report reflect the views of the authors, who are responsible for the 

facts and the accuracy of the data presented herein. The contents do not necessarily 

reflect policy of the Department of Transportation. This report does not constitute a 

standard, specification, or regulation. The United States Government does not endorse 

products or manufacturers. Trade or manufacturer's names appear herein only because 

they are considered essential to the objective of this document. 
 

 

 

 

 

Technical Report Documentation Page 

1. Report No. 2. Government Accession No. 3. Recipient’s Catalog No. 

FHWA-NHI–06-088 

4. Title and Subtitle 5. Report Date

 December 2006 

SOILS AND FOUNDATIONS 6. Performing Organization Code 

REFERENCE MANUAL – Volume I 

7. Author(s) 8. Performing Organization Report No. 

Naresh C. Samtani*, PE, PhD and Edward A. Nowatzki*, PE, PhD 

9. Performing Organization Name and Address 10. Work Unit No. (TRAIS) 

Ryan R. Berg and Associates, Inc. 

2190 Leyland Alcove, Woodbury, MN 55125 

11. Contract or Grant No. * NCS GeoResources, LLC 

640 W Paseo Rio Grande, Tucson, AZ 85737 DTFH-61-02-T-63016 

12. Sponsoring Agency Name and Address 13. Type of Report and Period Covered 

National Highway Institute 

U.S. Department of Transportation 14. Sponsoring Agency Code 

Federal Highway Administration, Washington, D.C. 20590 

15. Supplementary Notes 

FHWA COTR – Larry Jones 

FHWA Technical Review – Jerry A. DiMaggio, PE; Silas Nichols, PE; Richard Cheney, PE; 

Benjamin Rivers, PE; Justin Henwood, PE. 

Contractor Technical Review – Ryan R. Berg, PE; Robert C. Bachus, PhD, PE; 

Barry R. Christopher, PhD, PE 

This manual is an update of the 3rd Edition prepared by Parsons Brinckerhoff Quade & Douglas, Inc, in 2000. 

Author: Richard Cheney, PE. The authors of the 1st and 2nd editions prepared by the FHWA in 1982 and 1993, 

respectively, were Richard Cheney, PE and Ronald Chassie, PE. 

16. Abstract 

The Reference Manual for Soils and Foundations course is intended for design and construction professionals involved 

with the selection, design and construction of geotechnical features for surface transportation facilities. The manual is 

geared towards practitioners who routinely deal with soils and foundations issues but who may have little theoretical 

background in soil mechanics or foundation engineering. The manual’s content follows a project-oriented approach 

where the geotechnical aspects of a project are traced from preparation of the boring request through design 

computation of settlement, allowable footing pressure, etc., to the construction of approach embankments and 

foundations. Appendix A includes an example bridge project where such an approach is demonstrated. 

Recommendations are presented on how to layout borings efficiently, how to minimize approach embankment 

settlement, how to design the most cost-effective pier and abutment foundations, and how to transmit design 

information properly through plans, specifications, and/or contact with the project engineer so that the project can be 

constructed efficiently. 

The objective of this manual is to present recommended methods for the safe, cost-effective design and construction of 

geotechnical features. Coordination between geotechnical specialists and project team members at all phases of a 

project is stressed. Readers are encouraged to develop an appreciation of geotechnical activities in all project phases 

that influence or are influenced by their work. 

17. Key Words 18. Distribution Statement 

Subsurface exploration, testing, slope stability, embankments, cut slopes, shallow No restrictions. 

foundations, driven piles, drilled shafts, earth retaining structures, construction. 

19. Security Classif. (of this report) 20. Security Classif. (of this page) 21. No. of Pages 22. Price 

UNCLASSIFIED UNCLASSIFIED 462 

Form DOT F 1700.7(8-72) Reproduction of completed page authorized 
[THIS PAGE INTENTIONALLY BLANK] 
PREFACE 

This update to the Reference Manual for the Soils and Foundations course was developed to 

incorporate the guidance available from the FHWA in various recent manuals and Geotechnical 

Engineering Circulars (GECs). The update has evolved from its first two versions prepared by 

Richard Cheney and Ronald Chassie in 1982 and 1993, and the third version prepared by 

Richard Cheney in 2000. 

The updated edition of the FHWA Soils and Foundations manual contains an enormous amount 

of information ranging from methods for theoretically based analyses to “rules of thumb” 

solutions for a wide range of geotechnical and foundation design and construction issues. It is 

likely that this manual will be used nationwide for years to come by civil engineering 

generalists, geotechnical and foundation specialists, and others involved in transportation 

facilities. That being the case, the authors wish to caution against indiscriminate use of the 

manual’s guidance and recommendations. The manual should be considered to represent the 

minimum standard of practice. The user must realize that there is no possible way to cover all 

the intricate aspects of any given project. Even though the material presented is theoretically 

correct and represents the current state-of-the-practice, engineering judgment based on local 

conditions and knowledge must be applied. This is true of most engineering disciplines, but it is 

especially true in the area of soils and foundation engineering and construction. For example, 

the theoretical and empirical concepts in the manual relating to the analysis and design of deep 

foundations apply to piles installed in the glacial tills of the northeast as well as to drilled shafts 

installed in the cemented soils of the southwest. The most important thing in both applications is 

that the values for the parameters to be used in the analysis and design be selected by a 

geotechnical specialist who is intimately familiar with the type of soil in that region and 

intimately knowledgeable about the regional construction procedures that are required for the 

proper installation of such foundations in local soils. 

General conventions used in the manual 

This manual addresses topics ranging from fundamental concepts in soil mechanics to the 

practical design of various geotechnical features ranging from earthworks (e.g., slopes) to 

foundations (e.g., spread footings, driven piles, drilled shafts and earth retaining structures). In 

the literature each of these topics has developed its own identity in terms of the terminology and 

symbols. Since most of the information presented in this manual appears in other FHWA 

publications, textbooks and publications, the authors faced a dilemma on the regarding 

terminology and symbols as well as other issues. Following is a brief discussion on such issues. 

FHWA NHI-06-088 Preface 

Soils and Foundations – Volume I P - 1 December 2006 
 

• Pressure versus Stress 

The terms “pressure” and “stress” both have units of force per unit area (e.g., pounds per 

square foot). In soil mechanics “pressure” generally refers to an applied load distributed 

over an area or to the pressure due to the self-weight of the soil mass. “Stress,” on the other 

hand, generally refers to the condition induced at a point within the soil mass by the 

application of an external load or pressure. For example, “overburden pressure,” which is 

due to the self weight of the soil, induces “geostatic stresses” within the soil mass. Induced 

stresses cause strains which ultimately result in measurable deformations that may affect the 

behavior of the structural element that is applying the load or pressure. For example, in the 

case of a shallow foundation, depending upon the magnitude and direction of the applied 

loading and the geometry of the footing, the pressure distribution at the base of the footing 

can be uniform, linearly varying, or non-linearly varying. In order to avoid confusion, the 

terms “pressure” and “stress” will be used interchangeably in this manual. In cases where 

the distinction is important, clarification will be provided by use of the terms “applied” or 

“induced.” 

• Symbols 

Some symbols represent more than one geotechnical parameter. For example, the symbol Cc 

is commonly used to identify the coefficient of curvature of a grain size distribution curve as 

well as the compression index derived from consolidation test results. Alternative symbols 

may be chosen, but then there is a risk of confusion and possible mistakes. To avoid the 

potential for confusion or mistakes, the Table of Contents contains a list of symbols for each 

chapter. 

• Units 

English units are the primary units in this manual. SI units are included in parenthesis in the 

text, except for equations whose constants have values based on a specific set of units, 

English or SI. In a few cases, where measurements are conventionally reported in SI units 

(e.g., aperture sizes in rock mapping), only SI units are reported. English units are used in 

example problems. Except where the units are related to equipment sizes (e.g., drill rods), 

all unit conversions are “soft,” i.e., approximate. Thus, 10 ft is converted to 3 m rather than 

3.05 m. The soft conversion for length in feet is rounded to the nearest 0.5 m. Thus, 15 ft is 

converted to 4.5 m not 4.57 m. 

FHWA NHI-06-088 Preface 

Soils and Foundations – Volume I P - 2 December 2006 
 

•	 Theoretical Details 

Since the primary purpose of this manual is to provide a concise treatment of the 

fundamental concepts in soil mechanics and an introduction to the practical design of various 

geotechnical features related to highway construction, the details of the theory underlying 

the methods of analysis have been largely omitted in favor of discussions on the application 

of those theories to geotechnical problems. Some exceptions to this general approach were 

made. For example, the concepts of lateral earth pressure and bearing capacity rely too 

heavily on a basic understanding of the Mohr’s circle for stress for a detailed presentation of 

the Mohr’s circle theory to be omitted. However, so as not to encumber the text, the basic 

theory of the Mohr’s circle is presented in Appendix B for the reader’s convenience and as 

an aid for the deeper understanding of the concepts of earth pressure and bearing capacity. 

•	 Standard Penetration Test (SPT) N-values 

The SPT is described in Chapter 3 of this manual. The geotechnical engineering literature is 

replete with correlations based on SPT N-values. Many of the published correlations were 

developed based on SPT N-values obtained with cathead and drop hammer methods. The 

SPT N-values used in these correlations do not take in account the effect of equipment 

features that might influence the actual amount of energy imparted during the SPT. The 

cathead and drop hammer systems typically deliver energy at an estimated average 

efficiency of 60%. Today’s automatic hammers deliver energy at a significantly higher 

efficiency (up to 90%). When published correlations based on SPT N-values are presented 

in this manual, they are noted as N60-values and the measured SPT N-values should be 

corrected for energy before using the correlations. 

Some researchers developed correction factors for use with their SPT N-value correlations to 

address the effects of overburden pressure. When published correlations presented in this 

manual are based upon values corrected for overburden they are noted as N160. Guidelines 

are provided as to when the N60-values should be corrected for overburden. 

•	 Allowable Stress Design (ASD) and Load and Resistance Factor Design (LRFD) 

Methods 

The design methods to be used in the transportation industry are currently (2006) in a state of 

transition from ASD to LRFD. The FHWA recognizes this transition and has developed 

separate comprehensive training courses for this purpose. Regardless of whether the ASD or 

LRFD is used, it is important to realize that the fundamentals of soil mechanics, such as the 

FHWA NHI-06-088 Preface 

Soils and Foundations – Volume I P - 3 December 2006 
determination of the strength and deformation of geomaterials do not change. The only 

difference between the two methods is the way in which the uncertainties in loads and 

resistances are accounted for in design. Since this manual is geared towards the fundamental 

understanding of the behavior of soils and the design of foundations, ASD has been used 

because at this time most practitioners are familiar with that method of design. However, for 

those readers who are interested in the nuances of both design methods Appendix C provides 

a brief discussion on the background and application of the ASD and LRFD methods. 

FHWA NHI-06-088 Preface 

Soils and Foundations – Volume I P - 4 December 2006 
ACKNOWLEDGEMENTS 

The authors would like to acknowledge the following events and people that were instrumental 

in the development of this manual. 

•	 Permission by the FHWA to adapt the August 2000 version of the Soils and Foundations 

Workshop Manual. 

•	 Provision by the FHWA of the electronic files of the August 2000 manual as well as other 

FHWA publications. 

•	 The support of Ryan R. Berg of Ryan R. Berg and Associates, Inc. (RRBA) in facilitating the 

preparation of this manual and coordinating reviews with the key players. 

•	 The support provided by the staff of NCS Consultants, LLC, (NCS) - Wolfgang Fritz, Juan 

Lopez and Randy Post (listed in alphabetical order of last names). They prepared some 

graphics, some example problems, reviewed selected data for accuracy with respect to 

original sources of information, compiled the Table of Contents, performed library searches 

for reference materials, and checked internal consistency in the numbering of chapter 

headings, figures, equations and tables. 

•	 Discussions with Jim Scott (URS-Denver) on various topics and his willingness to share 

reference material are truly appreciated. Dov Leshchinsky of ADAMA Engineering 

provided copies of the ReSSA and FoSSA programs which were used to generate several 

figures in the manual as well as presentation slides associated with the course presentation. 

Robert Bachus of Geosyntec Consultants prepared Appendices D and E. Allen Marr of 

GeoComp Corporation provided photographs of some laboratory testing equipment. Pat 

Hannigan of GRL Engineers, Inc. reviewed the driven pile portion of Chapter 9. Shawn 

Steiner of ConeTec, Inc. and Salvatore Caronna of gINT Software prepared the Cone 

Penetration Test (CPT) and boring logs, respectively, shown in Chapter 3 and Appendix A. 

Robert (Bob) Meyers (NMDOT), Ted Buell (HDR-Tucson) and Randy Simpson (URSPhoenix) provided comments on some sections (particularly Section 8.9). 

•	 Finally, the technical reviews and recommendations provided by Jerry DiMaggio, Silas 

Nichols, Benjamin Rivers, Richard Cheney (retired) and Justin Henwood of the FHWA, 

Ryan Berg of RRBA, Robert Bachus of Geosyntec Consultants, Jim Scott of URS, and Barry 

Christopher of Christopher Consultants, Inc., are gratefully acknowledged. 

FHWA NHI-06-088 Preface 

Soils and Foundations – Volume I P - 5 December 2006 
 

SPECIAL ACKNOWLEDGEMENTS 

A special acknowledgement is due of the efforts of Richard Cheney and Ronald Chassie for their 

work in the preparation of the previous versions of this manual. It is their work that made this 

course one of the most popular FHWA courses. Their work in developing this course over the 

past 25 years is acknowledged. 

With respect to this manual, the authors wish to especially acknowledge the in-depth review 

performed by Jerry DiMaggio and time he spent in direct discussions with the authors and other 

reviewers. Such discussions led to clarification of some existing guidance in other FHWA 

manuals as well as the introduction of new guidance in some chapters of this manual. 

FHWA NHI-06-088 Preface 

Soils and Foundations – Volume I P - 6 December 2006 
 

SI CONVERSION FACTORS 

APPROXIMATE CONVERSIONS FROM SI UNITS 

Symbol When You 

Know 

Multiply By To Find Symbol 

LENGTH 

mm 

m 

m 

km 

millimeters 

meters 

meters 

kilometers 

0.039 

3.28 

1.09 

0.621 

inches 

feet 

yards 

miles 

in 

ft 

yd 

mi 

AREA 

mm2 

m2 

m2 

ha 

km2 

square millimeters 

square meters 

square meters 

hectares 

square kilometers 

0,0015 

10.758 

1.188 

2.47 

0.386 

square inches 

square feet 

square yards 

acres 

square miles 

in2 

ft2 

yd2 

ac 

mi2 

VOLUME 

ml 

l 

m3 

m3 

milliliters 

liters 

cubic meters 

cubic meters 

0.034 

0.264 

35.29 

1.295 

fluid ounces 

gallons 

cubic feet 

cubic yards 

fl oz 

gal 

ft3 

yd3 

MASS 

g 

kg 

tonnes 

grams 

kilograms 

tonnes 

0.035 

2.205 

1.103 

ounces 

pounds 

US short tons 

oz 

lb 

tons 

TEMPERATURE 

oC Celsius 1.8oC + 32 Fahrenheit oF 

WEIGHT DENSITY 

kN/m3 kilonewtons / cubic 

meter 

6.36 Pound force / cubic foot pcf 

FORCE and PRESSURE or STRESS 

N 

kN 

kPa 

kPa 

newtons 

kilonewtons 

kilopascals 

kilopascals 

0.225 

225 

0.145 

20.88 

pound force 

pound force 

pound force / square inch 

pound force / square foot 

lbf 

lbf 

psi 

psf 

PERMEABILITY (VELOCITY) 

cm/sec centimeter/second 1.9685 feet/minute ft/min 
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SOILS AND FOUNDATIONS 

VOLUME I 

TABLE OF CONTENTS 

Page 

LIST OF FIGURES .............................................................................................................. ix 

LIST OF TABLES .................................................................................................................xv 

LIST OF SYMBOLS ......................................................................................................... xviii 

1.0 INTRODUCTION.................................................................................................... 1-1 

1.1 PURPOSE AND SCOPE .............................................................................. 1-1 

1.2 SOILS AND FOUNDATIONS FOR HIGHWAY FACILITIES .................. 1-5 

1.3 ORGANIZATION OF MANUAL .............................................................. 1-10 

1.4 REFERENCES ........................................................................................... 1-12 

1.4.1 Primary FHWA References ............................................................ 1-12 

1.4.2 Other Primary References ............................................................... 1-13 

2.0 STRESS AND STRAIN IN SOILS......................................................................... 2-1 

2.1 BASIC WEIGHT-VOLUME RELATIONSHIPS ......................................... 2-3 

2.1.1 Volume Ratios .................................................................................. 2-3 

2.1.2 Weight Ratios .................................................................................... 2-4 

 2.1.3 Weight-Volume Ratios (Unit Weights) and Specific Gravity ........... 2-5 

 2.1.4 Determination and Use of Basic Weight-Volume Relations ............. 2-7 

 2.1.5 Size of Grains in the Solid Phase....................................................... 2-9 

 2.1.6 Shape of Grains in Solid Phase........................................................ 2-13 

2.1.6.1 Bulky Shape ......................................................................... 2-13 

2.1.6.2 Platy Shape .......................................................................... 2-15 

 2.1.7 Effect of Water on Physical States of Soils ..................................... 2-15 

2.2 PRINCIPLE OF EFFECITVE STRESS...................................................... 2-19 

2.3 OVERBURDEN PRESSURE...................................................................... 2-19 

2.4 VERTICAL STRESS DISTRIBUTION IN SOIL DUE TO EXTERNAL 

LOADINGS ................................................................................................. 2-22 

 2.4.1 Uniformly Loaded Continuous (Strip) and Square Footings .......... 2-24 

 2.4.2 Approximate (2:1) Stress Distribution Concept .............................. 2-26 

2.5 REPRESENTATION OF IMPOSED PRESSURES ON 

 THE po DIAGRAM ..................................................................................... 2-26 

2.6 LOAD-DEFORMATION PROCESS IN SOILS ........................................ 2-30 

 2.6.1 Time Dependent Load-Deformation (Consolidation) Process ....... 2-31 

 2.6.2 Comparison of Drainage Rates between Coarse-Grained and FineGrained Soil ..................................................................................... 2-34 

2.7 LATERAL STRESSES IN FOUNDATION SOILS ................................... 2-34 

 2.7.1 Effect of Shear Strength of Soils on Lateral Pressures.................... 2-36 

FHWA NHI-06-089 Table of Contents 

Soils and Foundations – Volume I i December 2006 
2.8 STRENGTH OF SOILS TO RESIST IMPOSED STRESSES ................... 2-38 

 2.8.1 Basic Concept of Shearing Resistance and Shearing Strength........ 2-38 

2.9 STRENGTH OF SOILS RELATED TO LATERAL 

 EARTH PRESSURES ................................................................................. 2-42 

 2.9.1 Distribution of Lateral Earth and Water Pressures .......................... 2-45 

 2.9.2 Deformations Associated with Lateral Pressures ............................ 2-47 

2.10 UNSATURATED SOIL MECHANICS...................................................... 2-47 

3.0 SUBSURFACE EXPLORATIONS........................................................................ 3-1 

3.01 Primary References ........................................................................................ 3-3 

3.1 PREPARING FOR SUBSURFACE EXPLORATION................................. 3-3 

 3.1.1 Soil Formations and Landforms......................................................... 3-6 

3.1.1.1 Residual Soils ........................................................................ 3-7 

3.1.1.2 Transported Soils ................................................................... 3-9 

 3.1.1.3 Area Concept of Explorations Based on Landforms ............. 3-9 

3.2 FIELD RECONNAISSANCE ..................................................................... 3-13 

3.3 SUBSURFACE EXPLORATION PROGRAM .......................................... 3-15 

3.4 SAMPLING TECHNIQUES AND TOOLS................................................. 3-17 

3.5 BORING METHODS .................................................................................. 3-20 

3.5.1 Auger Borings .................................................................................. 3-20 

3.5.2 Wash-type Borings........................................................................... 3-26 

3.5.3 Coring in Rocks ............................................................................... 3-26 

3.6 SAMPLING METHODS ............................................................................. 3-28 

 3.6.1 Disturbed Sampling of Soil.............................................................. 3-28 

 3.6.2 Undisturbed Sampling of Soil.......................................................... 3-28 

 3.6.3 Thin-Walled (Shelby) Tube Sampling............................................. 3-35 

 3.6.4 Undisturbed Sampling of Rock (Rock Coring) ............................... 3-38 

3.6.4.1 Core Barrels ......................................................................... 3-38 

3.6.4.2 Coring Bits ........................................................................... 3-40 

3.6.4.3 Drilling Fluid ....................................................................... 3-43 

 3.6.5 Observations During Rock Core Drilling ........................................ 3-43 

3.6.5.1 Drilling Rate/Time ............................................................... 3-43 

3.6.5.2 Core Photographs................................................................. 3-43 

3.6.5.3 Rock Classification .............................................................. 3-44 

3.6.5.4 Recovery .............................................................................. 3-44 

 3.6.5.5 Rock Quality Designation (RQD)........................................ 3-44 

3.6.5.6 Drilling Fluid Recovery ....................................................... 3-47 

 3.6.5.7 Core Handling and Labeling ................................................ 3-48 

 3.6.5.8 Care and Preservation of Rock Samples .............................. 3-50 

3.6.6 Geologic Mapping ........................................................................... 3-50 

3.7 STANDARD PENETRATION TEST (SPT) ............................................... 3-51 

 3.7.1 Energy Efficiency of Hammers ....................................................... 3-54 

 3.7.2 Effect of Overburden Stress on N-values ........................................ 3-56 

 3.7.3 Correlation of SPT N-Values with Basic Soil Characteristics........... 3-58 

 3.7.3.1 Applicability of SPTs in Gravelly Soils .............................. 3-58 

3.7.4 SPT Test Errors ................................................................................ 3-59 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I ii December 2006 
3.8 LOG OF BOREHOLE INFORMATION (“BORING LOGS”).................... 3-62 

3.8.1 Boring Log Format ........................................................................... 3-62 

3.8.2 Duties of the Logger ......................................................................... 3-62 

3.9 CONE PENETRATION TESTING (CPT).................................................. 3-67 

 3.9.1 Equipment Description and Operation ............................................ 3-68 

 3.9.2 The Standard Cone Penetration Test (CPT) .................................... 3-69 

 3.9.3 The Piezo-cone Penetration Test (CPTu) ........................................ 3-71 

 3.9.4 The Seismic Piezocone Penetration Test (SCPTu) .......................... 3-71 

3.9.5 Test Procedures ................................................................................ 3-71 

3.9.6 CPT Profiles ..................................................................................... 3-72 

3.9.7 CPT Profile Interpretation ............................................................... 3-72 

3.10 DILATOMETER TEST (DMT) .................................................................. 3-75 

3.11 PRESSUREMETER TEST (PMT).............................................................. 3-76 

3.12 VANE SHEAR TEST (VST)....................................................................... 3-77 

3.13 GROUNDWATER MEASUREMENTS..................................................... 3-78 

 3.13.1 Information on Existing Wells ......................................................... 3-78 

3.13.2 Open Borings ................................................................................... 3-79 

3.13.3 Observation Wells............................................................................ 3-79 

3.13.4 Water Level Measurements ............................................................. 3-81 

3.13.4.1 Chalked Tape ..................................................................... 3-81 

 3.13.4.2 Tape with a Float ............................................................... 3-82 

3.13.4.3 Electric Water-Level Indicator .......................................... 3-82 

3.13.4.4 Data Loggers ...................................................................... 3-82 

3.14 GUIDELINES FOR MINIMUM SUBSURFACE EXPLORATION ........... 3-83 

 3.14.1 Recommendations for Sampling Depth Intervals in Soils ............... 3-87 

 3.14.2 Recommendations for Sampling Depth Intervals in Rocks ............. 3-89 

 3.14.3 Recommendations for Water Level Monitoring in Borings ............ 3-89 

3.15 GEOPHYSICAL TESTS ............................................................................. 3-90 

 3.15.1 Types of Geophysical Tests............................................................. 3-90 

 3.15.2 Advantages and Disadvantages of Geophysical Tests..................... 3-94 

 3.15.2.1 Advantages of Geophysical Tests ...................................... 3-94 

 3.15.2.2 Disadvantages of Geophysical Tests ................................. 3-95 

 3.15.3 Examples of Uses of Geophysical Tests .......................................... 3-95 

4.0 ENGINEERING DESCRIPTION, CLASSIFICATION AND 

CHARACTERISTICS OF SOILS AND ROCKS ............................................... 4-1 

4.01 Primary References ........................................................................................ 4-3 

4.1 SOIL DESCRIPTION.................................................................................... 4-3 

 4.1.1 Consistency and Apparent Density.................................................... 4-4 

4.1.2 Water Content (Moisture) ................................................................. 4-6 

4.1.3 Color .................................................................................................. 4-6 

4.1.4 Type of Soil........................................................................................ 4-6 

 4.1.4.1 Coarse-Grained Soils (Gravel and Sand) ............................... 4-7 

4.1.4.2 Fine-Grained Soils ................................................................ 4-9 

4.1.4.3 Highly Organic Soils ........................................................... 4-10 

4.1.4.4 Minor Soil Type(s)............................................................... 4-11 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I iii December 2006 
 4.1.4.5 Inclusions ............................................................................. 4-12 

4.1.4.6 Other Descriptors ................................................................. 4-12 

4.1.4.7 Layered Soils ....................................................................... 4-12 

4.1.4.8 Geological Name ................................................................. 4-13 

4.2 SOIL CLASSIFICATION ........................................................................... 4-14 

 4.2.1 Unified Soil Classification System (USCS) .................................... 4-14 

 4.2.1.1 Classification of Coarse-Grained Soils ................................ 4-18 

 4.2.1.2 Classification of Fine-Grained Soils .................................... 4-21 

 4.2.2 AASHTO Soil Classification System .............................................. 4-26 

4.3 ENGINEERING CHARACTERISTICS OF SOILS................................... 4-31 

 4.3.1 Engineering Characteristics of Coarse-Grained Soils (Sands and 

Gravels)............................................................................................ 4-31 

 4.3.2 Engineering Characteristics of Fine-Grained Soils 

 (Inorganic Clays) ............................................................................. 4-31 

 4.3.3 Engineering Characteristics of Fine-Grained Soils 

 (Inorganic Silts) ............................................................................... 4-31 

 4.3.4 Engineering Characteristics of Organic Soils ................................. 4-32 

4.4 PRACTICAL ASPECTS OF ENGINEERING CHARACTERISTICS OF 

COARSE-GRAINED SOILS ...................................................................... 4-32 

4.5 PRACTICAL ASPECTS OF ENGINEERING CHARACTERISTICS OF 

FINE-GRAINED SOILS ............................................................................. 4-33 

4.6 DESCRIPTION OF ROCK ......................................................................... 4-34 

4.6.1 Rock Type ........................................................................................ 4-34 

4.6.2 Color ................................................................................................ 4-36 

 4.6.3 Grain Size and Shape....................................................................... 4-36 

4.6.4 Stratification/Foliation ..................................................................... 4-37 

4.6.5 Mineral Composition ....................................................................... 4-37 

 4.6.6 Weathering and Alteration ............................................................... 4-37 

4.6.7 Strength............................................................................................ 4-37 

4.6.8 Hardness........................................................................................... 4-39 

4.6.9 Rock Discontinuity .......................................................................... 4-39 

 4.6.10 Fracture Description ........................................................................ 4-41 

 4.6.11 Rock Mass Classification................................................................. 4-42 

4.7 SUBSURFACE PROFILE DEVELOPMENT ........................................... 4-45 

 4.7.1 Use of Historical Data in Development of Subsurface Profile ........ 4-48 

5.0 LABORATORY TESTING FOR GEOTECHNICAL DESIGN AND 

CONSTRUCTION .................................................................................................. 5-1 

5.01 Primary References ........................................................................................ 5-1 

5.1 QUALITY ASSURANCE FOR LABORATORY TESTING ..................... 5-2 

5.1.1 Sample Tracking ................................................................................ 5-2 

5.1.2 Sample Storage .................................................................................. 5-2 

5.1.3 Sample Handling................................................................................ 5-4 

 5.1.4 Effects of Sample Disturbance........................................................... 5-4 

5.1.5 Specimen Selection ............................................................................ 5-4 

5.2 LABORATORY TESTING FOR SOILS ..................................................... 5-5 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I iv December 2006
 5.3 LABORATORY INDEX TESTS FOR SOILS .......................................... 5-10 

5.3.1 General............................................................................................. 5-10 

5.3.2 Moisture Content ............................................................................. 5-10 

5.3.3 Unit Weight...................................................................................... 5-11 

 5.3.4 Particle Size Distribution................................................................. 5-12 

5.3.4.1 Sand Equivalent ................................................................... 5-16 

5.3.5 Atterberg Limits............................................................................... 5-16 

 5.3.5.1 Significance of the “A-line” and “U-line” on 

 Plasticity Chart..................................................................... 5-19 

5.3.6 Specific Gravity ............................................................................... 5-19 

5.3.7 Organic Content ............................................................................... 5-20 

5.3.8 Electro Chemical Classification Tests ............................................. 5-21 

5.3.9 Laboratory Classification................................................................. 5-21 

5.4 CONSOLIDATION TESTING .................................................................. 5-22 

 5.4.1 Process of Consolidation ................................................................. 5-22 

5.4.2 Consolidation Testing ...................................................................... 5-23 

5.4.3 Procedures........................................................................................ 5-24 

 5.4.4 Presentation and Understanding the Consolidation Test Results .... 5-26 

 5.4.5 Comments on the Consolidation Tests ............................................ 5-31 

 5.4.6 Useful Correlations between Consolidation Parameters and Index 

Values .............................................................................................. 5-33 

5.4.6.1 Compression Index, Cc ........................................................ 5-33 

5.4.6.2 Recompression Index, Cr ..................................................... 5-35 

 5.4.6.3 Coefficient of Vertical Consolidation, cv............................. 5-35 

 5.4.6.4 Coefficient of Secondary Compression, Cα......................... 5-37 

5.5 SHEAR STRENGTH OF SOILS ................................................................ 5-38 

 5.5.1 Concept of Frictional and Cohesive Strengths ................................ 5-38 

 5.5.1.1 Strength Due to Friction ...................................................... 5-38 

 5.5.1.2 Strength Due to Cohesion .................................................... 5-40 

 5.5.1.3 Simplified Expression for Shear Strength of Soils .............. 5-42 

 5.5.2 Strength Testing of Soils in the Laboratory..................................... 5-42 

 5.5.2.1 Unconfined Compression (UC) Tests ................................ 5-44 

5.5.2.2 Triaxial Tests ....................................................................... 5-45 

5.5.2.3 Direct Shear Tests .............................................................. 5-49 

 5.5.3 Factors Affecting Strength Testing Results ..................................... 5-51 

 5.5.4 Comparison of Laboratory and Field Strengths............................... 5-52 

 5.5.5 Selection of Design Shear Strength ................................................. 5-53 

 5.5.6 Correlations of Shear Strength Parameters with Index Parameters . 5-53 

 5.5.6.1 Undrained Shear Strength of Cohesive Soils....................... 5-53 

 5.5.6.2 Drained and Effective Shear Strength of Cohesive Soils .... 5-55 

 5.5.6.3 Shear Strength of Cohesionless Soils .................................. 5-56 

5.6 PERMEABILITY ........................................................................................ 5-58 

5.6.1 General............................................................................................. 5-58 

5.6.2 Equipment ........................................................................................ 5-58 

5.6.3 Procedures........................................................................................ 5-61 

 5.6.4 Useful Correlations of Permeability with Index Values .................. 5-62 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I v December 2006
5.7 VOLUME CHANGE PHENOMENA DUE TO LOADING AND 

MOISTURE ................................................................................................ 5-65 

 5.7.1 Swell Potential of Clays .................................................................. 5-65 

 5.7.1.1 Evaluation of Expansion (Swell) Potential .......................... 5-67 

 5.7.2 Collapse Potential of Soils .............................................................. 5-68 

 5.7.3 Expansion of Soils due to Frost Action .......................................... 5-70 

5.8 COMPACTION CHARACTERISTICS OF SOIL ..................................... 5-72 

5.8.1 Concept of Compaction .................................................................. 5-72 

5.8.2 Test Procedures ............................................................................... 5-73 

 5.8.3 Implication of Laboratory Tests on Field 

 Compaction Specifications ............................................................. 5-76 

 5.8.4 Engineering Characteristics of Compacted Soils............................. 5-81 

 5.8.4.1 Effect of Increase in Moisture Content on Shear Strength 

 of Compacted Soils .............................................................. 5-83 

5.9 ELASTIC PROPERTIES OF SOILS .......................................................... 5-83 

5.10 COMMON SENSE GUIDELINES FOR LABORATORY TESTING OF 

SOILS .......................................................................................................... 5-84 

5.11 LABORATORY TESTS FOR ROCK......................................................... 5-87 

5.11.1 Introduction ..................................................................................... 5-87 

5.11.2 Point-Load Strength Test ................................................................. 5-89 

 5.11.3 Unconfined Compressive Strength of Intact Rock Core ................ 5-91 

 5.11.4 Elastic Modulus of Intact Rock Core .............................................. 5-91 

 5.11.5 Laboratory Direct Shear Test .......................................................... 5-92 

5.12 ELASTIC PROPERTIES OF ROCKS ........................................................ 5-94 

 5.12.1 Elastic Modulus of Rock Mass ....................................................... 5-95 

5.13 COMMON SENSE GUIDELINES FOR LABORATORY TESTING OF 

ROCKS ........................................................................................................ 5-96 

5.14 PRACTICAL ASPECTS FOR LABORATORY TESTING....................... 5-98 

5.15 VARIABILITY OF MEASURED PROPERTIES ...................................... 5-99 

6.0 SLOPE STABILITY................................................................................................ 6-1 

6.01 Primary Reference ......................................................................................... 6-1 

6.1 EFFECTS OF WATER ON SLOPE STABILITY ........................................ 6-3 

6.2 DESIGN FACTOR OF SAFETY.................................................................. 6-5 

6.3 INFINITE SLOPE ANALYSIS..................................................................... 6-5 

 6.3.1 Infinite Slopes in Dry Cohesionless Soils.......................................... 6-5 

 6.3.2 Infinite Slopes in c-φ Soils with Parallel Seepage............................. 6-6 

6.4 CIRCULAR ARC FAILURE ........................................................................ 6-9 

 6.4.1 Simple Rule of Thumb for Factor of Safety .................................... 6-10 

 6.4.2 Stability Analysis Methods (General).............................................. 6-11 

 6.4.3 Ordinary Method of Slices – Step-By-Step 

 Computation Procedure ................................................................... 6-13 

6.4.4 Recommended Stability Methods .................................................... 6-20 

 6.4.5 Remarks on Safety Factor ................................................................ 6-22 

6.5 CRITICAL FAILURE SURFACE ............................................................. 6-22 

6.6 DESIGN (STABILITY) CHARTS.............................................................. 6-24 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I vi December 2006 
 6.6.1 Historical Background ..................................................................... 6-25 

6.6.2 Taylor’s Stability Charts .................................................................. 6-25 

 6.6.2.1 Determination of the Factor of Safety for a Slope ............... 6-30 

6.6.3 Janbu’s Stability Charts ................................................................... 6-33 

6.7 SLIDING BLOCK FAILURE ..................................................................... 6-43 

 6.7.1 Sliding Block – Hand Method of Analysis...................................... 6-44 

 6.7.1.1 Computation of Forces - Simple Sliding Block Analysis .... 6-45 

 6.7.2 Computation of Forces - Complicated Sliding Block Analysis....... 6-47 

6.8 SLOPE STABLITY ANALYSIS USING COMPUTER PROGRAMS ..... 6-48 

6.9 IMPROVING THE STABILITY OF EMBANKMENTS........................... 6-50 

 6.9.1 Embankment Stability Design Solutions ............................................ 6-50 

 6.9.2 Design Approach for Reinforced Soil Slopes ..................................... 6-50 

 6.9.2.1 Preliminary Feasibility Design of RSS ................................ 6-53 

6.10 IMPROVING THE STABILITY OF CUT SLOPES ................................. 6-57 

 6.10.1 Deep Seated Failure .......................................................................... 6-57 

 6.10.2 Shallow Surface Failures .................................................................. 6-58 

 6.10.3 Factor of Safety - Cut Slopes ............................................................ 6-59 

7.0 APPROACH ROADWAY DEFORMATIONS .................................................... 7-1 

7.1 TYPICAL APPROACH ROADWAY DEFORMATION PROBLEMS ...... 7-1 

7.2 INTERNAL DERFROMTION WITHIN EMBANKMENTS ...................... 7-6 

 7.2.1 General Considerations for Select Structural Backfill....................... 7-6 

 7.2.2 General Considerations for Drainage Aggregate............................... 7-6 

 7.2.3 Use of Geosynthetics to Control Internal Deformations ................. 7-10 

7.3 EXTERNAL DEFORMATION IN FOUNDATION SOILS BELOW
 

EMBANKMENTS ...................................................................................... 7-10 

 7.3.1 Procedure for Estimating Stress Distribution in Foundation Soils 

 under Fills ........................................................................................ 7-11 

7.4 COMPUTATION OF IMMEDIATE SETTLEMENT................................ 7-15 

 7.4.1 Modified Hough’s Method for Estimating Immediate Settlements. 7-16 

 7.4.1.1 Comments on the Computed Settlement of Embankments . 7-19 

7.5 COMPUTATION OF CONSOLIDATION (LONG-TERM) 

SETTLEMENTS.......................................................................................... 7-19 

 7.5.1 Correction of Laboratory One-Dimensional Consolidation Curves 7-21 

 7.5.2 Computation of Primary Consolidation Settlements ....................... 7-24 

7.5.2.1 Normally Consolidated Soils ............................................... 7-24 

7.5.2.2 Overconsolidated (Preconsolidated) Soils ........................... 7-26 

7.5.2.3 Underconsolidated Soils ...................................................... 7-27 

 7.5.3 Consolidation Rates (Time Rate of Consolidation Settlement) ....... 7-28 

7.5.3.1 Percent Consolidation .......................................................... 7-30 

 7.5.3.2 Step-by-step Procedure to Determine Amount and Time 

 for Consolidation ................................................................. 7-33 

 7.5.4 Secondary Compression of Cohesive Soils ..................................... 7-38 

7.6 LATERAL SQUEEZE OF FOUNDATION SOILS ................................... 7-40 

 7.6.1 Threshold Condition for Lateral Squeeze........................................ 7-41 

 7.6.2 Calculation of the Safety Factor against Lateral Squeeze ............... 7-42 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I vii December 2006
7.6.3 Estimation of Horizontal Movement of Abutments......................... 7-43 

7.7 DESIGN SOLUTIONS - DEFORMATION PROBLEMS ........................ 7-43 

7.7.1 Reducing the Amount of Settlement ................................................ 7-44 

 7.7.1.1 Category 1 - Increasing the Resistance ................................ 7-44 

 7.7.1.2 Category 2 - Reducing the Load .......................................... 7-44 

7.7.2 Reducing Settlement Time............................................................... 7-44 

7.7.2.1 Surcharge Treatment ............................................................ 7-45 

7.7.2.2 Vertical Drains ..................................................................... 7-46 

7.7.3 Design Solutions to Prevent Abutment Tilting ................................ 7-47 

7.8 PRACTICAL ASPECTS OF EMBANKMENT SETTLEMENT .............. 7-48 

7.9 CONSTRUCTION MONITORING AND QUALITY ASSURANCE ...... 7-48 

7.9.1 Embankment Construction Monitoring by Instrumentation ............ 7-49 

7.9.1.1 Inspector’s Visual Observation............................................ 7-49 

7.9.1.2 Types of Instrumentation ..................................................... 7-50 

 7.9.1.3 Typical Locations for Instruments ....................................... 7-51 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I viii December 2006 
LIST OF FIGURES 

Figure Caption Page 

1-1 Aerial view of a pair of 3-span Interstate 10 (I-10) bridges over local 

 roadway in Tucson, Arizona................................................................................ 1-1 

1-2 Example of 3-span roadway bridges over another roadway ................................ 1-2 

1-3 The “BIG I” stack interchange at the intersection of I-40 and I-25 in 

 Albuquerque, New Mexico (Photo: Courtesy of Bob Meyers, NMDOT) 

 (Note: A stack interchange is a free-flowing junction between two or more 

roadways that allows turning in all directions) .................................................... 1-2 

1-4 A major multi-span bridge structure over water (George P. Coleman Bridge 

 over the York River in Yorktown, Virginia) ....................................................... 1-3 

1-5 Example of a roadway bounded by cut slopes and wet lands in an 

 environmentally sensitive area............................................................................. 1-3 

1-6 Combinations of sources of information required to solve geotechnical 

engineering issues ................................................................................................ 1-5 

1-7 Geotechnical activity flow chart for a typical project using design-bid-build 

procurement process ............................................................................................ 1-8 

2-1 A unit of soil mass and its idealization ................................................................ 2-2 

2-2 Example of laboratory sieves for mechanical analysis for grain size 

 distributions ....................................................................................................... 2-11 

2-3 Sample grain size distribution curves ................................................................ 2-12 

2-4 Terminology used to describe shape of coarse-grained soils (Mitchell, 1976) . 2-14 

2-5 Conceptual changes in soil phases as a function of water content .................... 2-16 

2-6 Plasticity chart and significance of Atterberg Limits ........................................ 2-18 

2-7 Example calculation of a po-diagram................................................................. 2-21 

2-8 Schematic of vertical stress distribution under embankment loading. Graphic 

generated by FoSSA (2003) program. ............................................................... 2-22 

2-9 Vertical stress contours (isobars) by Boussinesq’s theory for continuous and 

square footings (modified after Sowers, 1979; AASHTO, 2002) ..................... 2-25 

2-10 Distribution of vertical stress by the 2:1 method 

 (after Perloff and Baron, 1976).......................................................................... 2-27 

2-11 Combined plot of overburden pressures (total and effective) and pressure due 

 to imposed loads ................................................................................................ 2-28 

2-12 Example calculation of pf with stress increments from strip load 

 on p0-diagram..................................................................................................... 2-29 

2-13 Spring-piston analogy for the consolidation process of fine-grained soils........ 2-32 

2-14 Comparison of excess pore water pressure dissipation in coarse-grained and 

 fine-grained soils................................................................................................ 2-34 

2-15 Schematic of effect of lateral stresses ................................................................ 2-35 

2-16 Schematic of lateral stress increases beneath an embankment. Graphic 

 generated by FoSSA (2003) .............................................................................. 2-37 

2-17 Basic concept of shearing resistance and strength (after Murthy, 1989)........... 2-39 

2-18 Graphical representation of shearing strength ................................................... 2-41 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I ix December 2006 
2-19 Stress states on a soil element subjected only to body stresses: (a) In-situ 

 geostatic effective vertical and horizontal stresses, (b) Insertion of 

 hypothetical infinitely rigid, infinitely thin frictionless wall and removal of 

 soil to left of wall, (c) Active condition of wall movement away from 

 retained soil, (d) Passive condition of wall movement into retained soil .......... 2-43 

2-20 Development of Rankine active and passive failure zones for a smooth 

 retaining wall ..................................................................................................... 2-45 

2-21 Failure surface, pressure distribution and forces 

 (a) active case, (b) passive case ......................................................................... 2-46 

2-22 General distribution of combined active earth pressure and water pressure ..... 2-46 

3-1 Recommended process for developing subsurface model for engineering 

 design (FHWA, 2002a) ........................................................................................ 3-2 

3-2 Typical weathering profile for metamorphic and igneous rocks (Deere and 

 Patton, 1971) ........................................................................................................ 3-8 

3-3 Typical field reconnaissance form ..................................................................... 3-14 

3-4 Common in-situ tests for geotechnical site characterization of soils (FHWA, 

2002b) ................................................................................................................ 3-18 

3-5 (a) Large diameter auger, (b) Small diameter continuous flight auger ............. 3-21 

3-6 (a) Solid stem auger and hollow stem auger (FHWA, 2002b), (b) Hollow 

stem auger components (ASTM D 4700), (c) Sizes of hollow stem auger 

flights (FHWA, 2002b), (d) Outer and inner assembly of hollow stem auger 

 (FHWA, 2002b) ................................................................................................. 3-25 

3-7 (a) Schematic of drilling rig for rotary wash methods (after Hvorslev, 1948), 

(b) Typical drilling configuration, (c) Settling basin (mud tank) ...................... 3-27 

3-8 Split barrel sampler ............................................................................................ 3-28 

3-9 Schematic of thin-walled (Shelby) tube (after ASTM D 4700) and photo of 

tube with end caps (FHWA, 2002b) ................................................................. 3-31 

3-10 Stationary piston sampler schematic (after ASTM D 4700) and photo 

 (FHWA 2002b) .................................................................................................. 3-33 

3-11 Denison sampler (FHWA, 1997) ....................................................................... 3-33 

3-12 Pitcher sampler (FHWA, 1997, 2002b) ............................................................. 3-34 

3-13 Shelby tube sealing methods: (a) Microcrystalline Wax, (b) O-Ring packer. 

 (FHWA, 2002b) ................................................................................................. 3-38 

3-14 (a) Single, (b) and (c) Double tube rock core barrels (FHWA, 1997) ............... 3-39 

3-15 Double tube core barrel. (a) Outer barrel assembly (b) Inner barrel assembly 

(FHWA, 2002b) ................................................................................................. 3-40 

3-16 Coring bits: Diamond (top left), Carbide (top right), and Sawtooth 

 (bottom center) (FHWA, 2001) ......................................................................... 3-41 

3-17 Modified core recovery as an index of rock mass quality (FHWA, 1997) ........ 3-45 

3-18 Length measurements for core RQD determination (FHWA, 1997)................. 3-46 

3-19 Core box for storage of recovered rock and labeling......................................... 3-49 

3-20 Sequence of driving split-barrel sampler during the Standard Penetration 

 Test (modified after FHWA, 2002b) ................................................................. 3-52 

3-21 (a) Stainless steel and brass liners, (b) Sample catchers (FHWA, 2002b) ........ 3-53 

3-22 SPT hammer types, (a) Donut, (b) Safety, and (c) Automatic (FHWA, 2006) . 3-54 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I x December 2006 
3-23 SPT N-values from (a) Uncorrected data, and (b) Corrected to 60% 

 efficiency (Data modified after Robertson, et al. 1983) .................................... 3-56 

3-24 Variation of overburden correction factor, CN, as a function of vertical 

 effective stress.................................................................................................... 3-57 

3-25a Example subsurface exploration log (0 - 35 ft depth) ......................................... 3-63 

3-25b Example subsurface exploration log (35 - 60 ft depth) ....................................... 3-64 

3-26 Procedures and components of the Cone Penetration Test (FHWA, 2002b) .... 3-68 

3-27 Cone penetrometer terminology (from Robertson and Campanella, 1989)....... 3-69 

3-28 Cone and piezocone penetrometers (Note the Quarter for Scale) 

 (FHWA, 2002b) ................................................................................................ 3-70 

3-29 Cone penetration testing from cone truck .......................................................... 3-71 

3-30 Piezo-cone results for Apple Freeway Bridge ..................................................... 3-73 

3-31 A commonly used simplified soil classification chart for standard electronic 

 friction cone (after Robertson, et al., 1986) ........................................................ 3-74 

3-32 Dilatometer test equipment and procedure (FHWA, 2002b) ............................. 3-75 

3-33 Pressuremeter test schematic (FHWA, 2002b) .................................................. 3-76 

3-34 Vane shear test equipment and procedure (after FHWA, 2002b)...................... 3-77 

3-35 Representative details of observation well installations. (a) Drilled-in place 

 stand-pipe piezometer, (b) Driven well point (FHWA, 1997)........................... 3-80 

4-1 Flow chart to determine the group symbol and group name for coarse- 

 grained soils (ASTM D 2488) ........................................................................... 4-17 

4-2 Evaluation of type of gradation for coarse-grained soils................................... 4-20 

4-3 Plasticity chart for Unified Soil Classification System (ASTM D 2488) .......... 4-22 

4-4a Flow chart to determine the group symbol and group name for fine-grained 

 soils (ASTM D 2488) ........................................................................................ 4-23 

4-4b Flow chart to determine the group symbol and group name for organic soils 

 (ASTM D 2488)................................................................................................. 4-24 

4-5 Range of liquid limit and plasticity index for soils in groups A-2, A-4, A-5, 

A-6 and A-7 per AASHTO M 145 (or ASTM D 3282) .................................... 4-28 

4-6 Comparison of the USCS with the AASHTO Soil Classification System 
 

 (after Utah DOT – Pavement Design and Management Manual, 2005)............ 4-29 

4-7 Comparison of soil groups in the USCS with the AASHTO Soil 

 Classification Systems (Holtz and Kovacs, 1981) ............................................. 4-30 

4-8 Example boring location plan (FHWA, 2002a) ................................................. 4-47 

4-9 Example interpreted subsurface profile (FHWA, 2002a) .................................. 4-48 

5-1 Sample laboratory test request form (Note: Only some tests are included in 

 this sample form) ................................................................................................. 5-3 

5-2 Example grain size distribution based on sieve analysis (Jumikis, 1962) ......... 5-14 

5-3 Grain size distribution curve based on data in Figure 5-2 ................................. 5-14 

5-4 Some of the equipment used for Atterberg Limits testing of soil ...................... 5-17 

5-5 Location of clay minerals on the Casagrande Plasticity Chart and Activity 

Index values (after Skempton, 1953, Mitchell, 1976, Holtz and 

 Kovacs, 1981) .................................................................................................... 5-18 

5-6 Schematic of a consolidation test....................................................................... 5-24 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xi December 2006 
5-7 (a) Components of consolidation test equipment, (b) Weighted lever arm – 

incremental load consolidation apparatus, (c) Automatic load frame and 

computerized consolidation apparatus 

(Photo courtesy of GeoComp Corporation) ....................................................... 5-25 

5-8 Consolidation test relationships (after NAVFAC, 1986a) ................................. 5-27 

5-9 Empirical correlation between Compression Index and natural (in-situ) water 

content (from Terzaghi, et al.,1996) .................................................................. 5-34 

5-10 Approximate correlations between cv and LL (NAVFAC, 1986a) ................... 5-36 

5-11 Correlation between Cαe and natural water content (NAVFAC, 1986a) ........... 5-37 

5-12 Potential contributions of various bonding mechanisms to cohesive strength 

(Ingles, 1962) ..................................................................................................... 5-41 

5-13 Schematic of an unconfined compression test ................................................... 5-44 

5-14 Schematic of a triaxial compression test (Lambe and Whitman, 1979) ............ 5-45 

5-15 (a) Failure of a loose sand specimen in a triaxial cell; and (b) Load frame, 

pressure panel, and computerized data acquisition system (Photograph 

 courtesy of GeoComp Corporation)................................................................... 5-46 

5-16 Typical stress-strain curves from CU test .......................................................... 5-48 

5-17 Schematic of the Direct Shear Test.................................................................... 5-49 

5-18 Direct shear testing box (Photograph courtesy of GeoComp Corporation)....... 5-50 

5-19 Soil sample mounted in direct shear testing apparatus (Photograph courtesy 

of GeoComp Corporation) ................................................................................. 5-50 

5-20 Relationship between the ratio of undrained shear strength to effective 

overburden pressure and plasticity index for normally-consolidated and 

overconsolidated clays (after Holtz and Kovacs, 1981) .................................... 5-54 

5-21 Relationships between φ and PI (after Terzaghi, et al., 1996)........................... 5-56 

5-22 Correlation between relative density, material classification and angle of 

internal friction for coarse-grained soils (NAVFAC, 1986a)............................ 5-57 

5-23 Correlation between relative density and SPT resistance 

 (NAVFAC, 1986a)............................................................................................. 5-57 

5-24 Rigid wall permeameter (Photograph courtesy of GeoComp Corporation) ...... 5-59 

5-25 Flexible wall permeameter (Photograph courtesy of GeoComp Corporation) .. 5-60 

5-26 The permeability of sands and gravels (after NAVFAC, 1986a) ...................... 5-64 

5-27 Classification of swell potential for soils (after Seed, et al., 1962)................... 5-66 

5-28 Chart for evaluation of collapsible soils (after Holtz and Hilf, 1961) ............... 5-68 

5-29 Approximate frost depth map for United States (Bowels, 1996) ...................... 5-71 

5-30 Hammer and mold for laboratory compaction test (tape measure is for scale 

purpose only....................................................................................................... 5-73 

5-31 Compaction curves (after Holtz and Kovacs, 1981) .......................................... 5-74 

5-32 Compactors recommended for various types of soil and rock 

 (Schroeder, 1980)............................................................................................... 5-77 

5-33 Relative density, relative compaction and void ratio concepts.......................... 5-78 

5-34 Example evaluation of economical field compaction conditions 

(after Bowles, 1979) .......................................................................................... 5-81 

5-35 Point load strength test equipment (Wyllie, 1999) ............................................ 5-89 

5-36 (a) Unconfined compression test on intact rock core, (b) Use of strain gage 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xii December 2006 
on intact rock core sample for measurement of stress-strain characteristics 

(Geomechanical Laboratory, University of Arizona) ........................................ 5-92 

5-37 Laboratory direct shear testing equipment for rock (Wyllie, 1999) .................. 5-93 

6-1 Embankment failures: (a) Infinite slope failure in embankment fill, 

(b) Circular arc failure in embankment fill and foundation soil, 

(c) Sliding block failure in embankment fill and foundation soil, and 

(d) Lateral squeeze of foundation soil ................................................................. 6-2 

6-2 Effect of water content on cohesive strength of clay ........................................... 6-4 

6-3 Infinite slope failure in dry sand .......................................................................... 6-6 

6-4 Infinite slope failure in a c-φ soil with parallel seepage ...................................... 6-7 

6-5 Typical circular arc failure mechanism ............................................................... 6-9 

6-6 Example proposed embankment ........................................................................ 6-10 

6-7 Geometry of Ordinary Method of Slices ........................................................... 6-12 

6-8 Example of dividing the failure mass in slices .................................................. 6-13 

6-9 Forces on a slice without water effect ............................................................... 6-15 

6-10 Forces on a slice with water effect..................................................................... 6-15 

6-11a Tabular form for computing weights of slices ................................................... 6-18 

6-11b Tabular form for calculating factor of safety by Ordinary Method of Slices .... 6-19 

6-12 Typical static forces acting on a slice of sliding mass without seepage ............ 6-20 

6-13 Location of critical circle by plotting contours of minimum safety factors 

for various trial circles ....................................................................................... 6-23 

6-14 Taylor’s chart for soils with friction angle (after Taylor, 1948)........................ 6-27 

6-15 Taylor's chart for φ'=0 conditions for slope angles (β) less than 54° (after 

 Taylor, 1948) ..................................................................................................... 6-28 

6-16 Stability charts for φ = 0 soils (Janbu, 1968)..................................................... 6-34 

6-17 Reduction factors to account for tension cracks to be used with stability 

 charts for φ=0 and φ > 0 soils (Janbu, 1968) ..................................................... 6-36 

6-18 Reduction factors to account for surcharge (upper) and submergence and/or 

seepage (lower) to be used with stability charts for φ=0 and φ > 0 soils 

 (Janbu, 1968) ..................................................................................................... 6-37 

6-19 Stability charts for φ > 0 (Janbu, 1968) ............................................................. 6-39 

6-20 Data for Example 6-2 ......................................................................................... 6-41 

6-21 Sliding block failure mechanism ....................................................................... 6-43 

6-22 Geometry and force components for sliding block analysis .............................. 6-44 

6-23 Example Simple Sliding Block Method using Rankine Pressure Coefficients . 6-46 

6-24 Reduction of grade line to improve slope stability ............................................ 6-52 

6-25 Use of counterweight berm to improve slope stability ...................................... 6-52 

6-26 Use of shear key to improve slope stability ....................................................... 6-52 

6-27 Failure modes for Reinforced Soil Slopes ......................................................... 6-53 

6-28 Chart solution for determining the reinforcement strength requirements 

 (Schmertmann, et al., 1987). ©The Tensar Corporation. ................................. 6-55 

6-29 Deep seated slope failure (left) and bench slope design (right) to prevent 

 slope failure........................................................................................................ 6-57 

6-30 Typical cut slope failure mechanism in clay soils ............................................. 6-58 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xiii December 2006 
7-1 (a) Elements of a bridge approach system, (b) Plan view of an approach 

system (modified after NCHRP, 1997)................................................................ 7-2 

7-2 Types of abutments (modified after NCHRP, 1990) ........................................... 7-3 

7-3 Problems leading to the existence of a bump (modified after NCHRP, 1997) .... 7-4 

7-4 Suggested approach embankment details ............................................................ 7-7 

7-5 Structural backfill placement limits for porous drainage aggregate .................... 7-8 

7-6 Pressure coefficients beneath the end of a fill (NYSDOT, 1970) ..................... 7-12 

7-7 Bearing capacity index (C') values used in Modified Hough method for 

computing immediate settlements of embankments (AASHTO, 2004 with 

2006 Interims; modified after Hough, 1959) ..................................................... 7-17 

7-8 Effect of sample disturbance on the shape of the one-dimensional 

consolidation curve (Reese, et al., 2006)........................................................... 7-21 

7-9 Construction of field virgin consolidation relationships (adapted from 

 USACE, 1994) ................................................................................................... 7-22 

7-10 Typical consolidation curve for normally consolidated soil – (a) Void ratio 

versus vertical effective stress and (b) Vertical strain versus vertical 

 effective stress.................................................................................................... 7-25 

7-11 Typical consolidation curve for overconsolidated soil – (a) Void ratio versus 

vertical effective stress and (b) Vertical strain versus vertical effective stress . 7-26 

7-12 Typical consolidation curve for under-consolidated soil – (a) Void ratio 

versus vertical effective stress and (b) Vertical strain versus vertical 

 effective stress.................................................................................................... 7-27 

7-13 Diagram illustrating consolidation of a layer of clay between two pervious 

layers (modified after Terzaghi, et al., 1996) ................................................... 7-28 

7-14 Typical settlement-time curve for clay under an embankment loading............. 7-32 

7-15 Logarithm-of-time method for determination of cv ........................................... 7-34 

7-16 Square-root-of-time method for determination of cv ......................................... 7-35 

7-17 Example time plot from one-dimensional consolidometer test for 

determination of secondary compression (USACE, 1994) ................................ 7-39 

7-18 Schematic of lateral squeeze phenomenon ........................................................ 7-40 

7-19 Examples of abutment tilting due to lateral squeeze (FHWA, 2006a) .............. 7-41 

7-20 Definitions for calculating safety factor against later squeeze (after Silvestri, 

1983) .................................................................................................................. 7-42 

7-21 Determination of surcharge time required to achieve desired settlement.......... 7-45 

7-22 Use of vertical drains to accelerate settlement................................................... 7-46 

7-23 Typical locations for various types of monitoring instruments for an 

embankment constructed over soft ground ........................................................ 7-51 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xiv December 2006 
LIST OF TABLES 

No. Caption Page 

1-1 Geotechnical involvement in project phases ....................................................... 1-9 

2-1 Summary of index properties and their application ............................................. 2-8 

2-2 Weight-volume relations (after Das, 1990) ......................................................... 2-8 

2-3 U.S. standard sieve sizes and corresponding opening dimension...................... 2-10 

2-4 Concept of soil phase, soil strength and soil deformation based on Liquidity 

 Index .................................................................................................................. 2-17 

3-1 Sources of historical site data (after FHWA, 2002a) ........................................... 3-4 

3-2 Common landforms of transported soils and their engineering significance .... 3-10 

3-3 Reference publications on in-situ testing (FHWA, 2002b) ............................... 3-19 

3-4a Soil and soft rock boring methods (FHWA, 2002a) .......................................... 3-22 

3-4b Rock core drilling methods (FHWA, 2002a)..................................................... 3-23 

3-4c Other exploratory techniques (FHWA, 2002a).................................................. 3-24 

3-5a Common samplers to retrieve disturbed soil samples 

 (modified after NAVFAC, 1986a) ..................................................................... 3-29 

3-5b Common samplers to retrieve disturbed soil samples 

 (modified after NAVFAC, 1986a) ..................................................................... 3-30 

3-6 Nominally undisturbed soil samplers (modified after NAVFAC, 1986a) ......... 3-32 

3-7 Dimensions of core sizes (FHWA, 1997) .......................................................... 3-41 

3-8 Advantages and disadvantages of the Standard Penetration Test (SPT) ........... 3-51 

3-9 Soil properties correlated with Standard Penetration Test values 

 (after Peck, et al., 1974)..................................................................................... 3-58 

3-10 Factors affecting the SPT and SPT results (after Kulhawy and Mayne, 1990) . 3-61 

3-11 Advantages and disadvantages of the Cone Penetration Test (CPT) 

 (FHWA, 2002b) ................................................................................................. 3-67 

3-12 Classification of Sensitivity Values (Mitchell, 1976)........................................ 3-78 

3-13 Guidelines for minimum number of exploration points and depth of 

 exploration (modified after FHWA, 2002a) ...................................................... 3-84 

3-14 Geophysical testing techniques (modified after FHWA, 2002a)....................... 3-91 

4-1 Evaluation of the apparent density of coarse-grained soils 

 (after Peck, et al., 1974)....................................................................................... 4-5 

4-2 Evaluation of the consistency of fine-grained soils (after Peck, et al., 1974) ..... 4-5 

4-3 Adjectives to describe water content of soils (ASTM D 2488)........................... 4-6 

4-4 Particle size definition for gravels and sands (after ASTM D 2488)................... 4-7 

4-5 Adjectives for describing size distribution for sands and gravels (after ASTM 

D 2488) ................................................................................................................ 4-8 

4-6 Field methods to describe plasticity (FHWA, 2002b) ....................................... 4-10 

4-7 Descriptive terms for layered soils (NAVFAC, 1986a) .................................... 4-13 

4-8 Basic USCS soil designations based on percents passing No. 200 (0.075mm)

 Sieve (after ASTM D 2487; Holtz and Kovacs, 1981) ............................................ 4-14 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xv December 2006 
4-9 Soil classification chart (laboratory method) (after ASTM D 2487) ................. 4-15 

4-10 Gradation based on Cu and Cc parameters ......................................................... 4-19 

4-11 Soil plasticity descriptors (based on Figures 4-3, 4-4a and 4-4b) ..................... 4-25 

4-12 Examples of description of fine-grained soils (based on Figures 4-3, 4-4a and 

4-4b)................................................................................................................... 4-25 

4-13 AASHTO soil classification system based on AASHTO M 145 (or 

ASTM D 3282) .................................................................................................. 4-27 

4-14 Example gradation limits of well-graded granular material (see Curve A in 

 Figure 4-2) ......................................................................................................... 4-33 

4-15 Example gradation limits of drainage materials (see Curve C in Figure 4-2) ... 4-33 

4-16 Rock groups and types (FHWA, 1997).............................................................. 4-35 

4-17 Terms to describe grain size (typically for sedimentary rocks) ......................... 4-36 

4-18 Terms to describe grain shape (for sedimentary rocks) ..................................... 4-36 

4-19 Terms to describe stratum thickness .................................................................. 4-37 

4-20 Terms to describe rock weathering and alteration (ISRM, 1981)...................... 4-38 

4-21 Terms to describe rock strength (ISRM, 1981) ................................................. 4-38 

4-22 Terms to describe rock hardness (FHWA, 2002b) ............................................ 4-39 

4-23 Terms to describe discontinuities (after ISRM, 1981)....................................... 4-40 

4-24 Terms to classify discontinuities based on aperture size (ISRM, 1981)............ 4-40 

4-25 Geomechanics classification of rock masses (AASHTO 2004 

with 2006 Interims)............................................................................................ 4-43 

4-26 Geomechanics rating adjustment for joint orientations (after AASHTO 2004 

with 2006 Interims)............................................................................................ 4-44 

4-27 Geomechanics rock mass classes determined from total ratings (AASHTO 

2004 with 2006 Interims)................................................................................... 4-44 

5-1 Commonly performed laboratory tests on soils (after FHWA, 2002a) ............... 5-6 

5-2 Methods for index testing of soils (after FHWA, 2002a) .................................... 5-7 

5-3 Methods for performance testing of soils (after FHWA, 2002a) ......................... 5-8 

5-4 Typical particle sizes, uniformity coefficients, void ratios and unit weights 

(from Kulhawy and Mayne, 1990)..................................................................... 5-13 

5-5 Correlations for Cc (modified after Holtz and Kovacs, 1981) ........................... 5-33 

5-6 Typical values of coefficient of vertical consolidation, cv (after Carter and 

Bentley, 1991).................................................................................................... 5-36 

5-7 Typical values of Cαe/Cc (Terzaghi, et al., 1996)............................................... 5-37 

5-8 Summary of issues relevant to shear strength evaluation in support of the 

design of typical geotechnical features (after FHWA, 2002a) .......................... 5-39 

5-9 Recommended maximum hydraulic gradient for permeability testing.............. 5-61 

5-10 Typical permeability values in soils (after Carter and Bentley, 1991) .............. 5-63 

5-11 Typical permeability values for highway materials (after Krebs and 

 Walker, 1971) .................................................................................................... 5-63 

5-12 Qualitative assessment of collapse potential (after ASTM D 5333) ................. 5-69 

5-13 Characteristics of laboratory compaction tests .................................................. 5-73 

5-14 Some values of Dr as a function of RC based on Modified and Standard 

Proctor Compaction Test ................................................................................... 5-79 

5-15 Average engineering properties of compacted inorganic soils 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xvi December 2006 
(after USBR, 1960) ............................................................................................ 5-82 

5-16 Elastic constants of various soils (after AASHTO 2004 with 2006 Interims)... 5-85 

5-17 Rate of increase of soil modulus with depth nh (tsf/ft) for sand 

(AASHTO 2004 with 2006 Interims) ................................................................ 5-85 

5-18 Common sense guidelines for laboratory testing of soils ................................. 5-86 

5-19 Common rock tests performed in the laboratory ............................................... 5-87 

5-20 Summary information on rock laboratory test methods (FHWA, 2002a) ......... 5-88 

5-21 Summary of elastic moduli for intact rock (AASHTO 2004 with 

 2006 Interims) ................................................................................................... 5-94 

5-22 Summary of Poisson's Ratio for intact rock (AASHTO 2004 with 

 2006 Interims) ................................................................................................... 5-95 

5-23 Estimation of Em based on RQD (AASHTO 2004 with 2006 Interims)............ 5-96 

5-24 Common sense guidelines for laboratory testing of rocks................................. 5-97 

5-25 Values of coefficient of variation, V, for geotechnical properties and in-situ 

tests (after Duncan, 2000)................................................................................ 5-100 

6-1 Slope stability guidelines for design.................................................................. 6-21 

6-2 Practical design solutions to mitigate embankment stability problems ............. 6-51 

6-3 Typical design solutions to mitigate cut slope stability problems ..................... 6-57 

7-1 General considerations for specification of select structural backfill.................. 7-9 

7-2 Suggested gradation for drainage aggregate ........................................................ 7-9 

7-3 Reconstruction of virgin field consolidation (modified from USACE, 1994) .. 7-23 

7-4 Average degree of consolidation, U, versus time factor, Tv, for uniform 
 

initial increase in pore water pressure ............................................................... 7-32 

7-5 Summary of abutment movements (Nicu, et al., 1971)..................................... 7-43 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xvii December 2006 
LIST OF SYMBOLS 

Chapter 1 

AASHTO American Association of State Highway and Transportation Officials 

AC Asphaltic concrete 

CMAR Construction manager at risk 

D-B Design-build 

FHWA Federal Highway Administration 

MSE Mechanically stabilized earth 

NHI National Highway Institute 

NMDOT New Mexico Department of Transportation 

PCC Portland cement concrete 

RCC Reinforced cement concrete 

RSS Reinforced soil slope(s) 

USDA United States Department of Agriculture 

USGS Unites States Geological Survey 

Chapter 2 

A Overall contact area 

AASHTO American Association of State Highway and Transportation Officials 

B Width 

c Cohesion 

c' Effective cohesion 

Dr Relative density 

Ds, DOSI Depth of significant influence 

e Void ratio 

emax Maximum void ratio 

emin Minimum void ratio 

Fa Tangential force 

Fr Shearing resistance 

G Specific gravity 

Gs Specific gravity of the solid phase 

h Embankment height 

hw Depth to water table 

K Value of proportionality constant; coefficient of lateral earth pressure 

Ka Coefficient of active earth pressure 

Ko Coefficient of lateral earth pressure “at rest” 

Kp Coefficient of passive earth pressure 

Kw Coefficient of lateral earth pressure for water 

L Length 

LI Liquidity index 

LL Liquid limit 

n Porosity 

P Load applied 

pa Active lateral pressure 

FHWA NHI-06-089 Table of Contents 

Soils and Foundations – Volume I xviii December 2006 
pf	 

 Final stress = effective overburden pressure + pressure increment due to 

external loads 

ph	  Lateral stress 

PI Plasticity index 

PL Plastic limit 

Pn	  Normal force 

po	 

 Effective overburden pressure 

pp	  Passive lateral pressure 

pt	  Total overburden pressure 

Q Load 

q, qo	 

 Unit load of embankment 

S Degree of saturation 

SI Shrinking index 

SL Shrinkage limit 

t time 

u Porewater pressure 

USCS Unified Soil Classification System 

V Volume of the total soils mass 

Va	 Volume of air phase 

Vs	 

 Volume of solid phase 

Vv	  Volume of total voids 

Vw	 Volume of water phase 

w Gravimetric water or moisture content 

W	 Weight of the total soil mass 

Wa	 Weight of air phase 

Ws	  Weight of solid phase 

Wv	 

 Weight of total voids 

Ww	 Weight of water phase 

z Depth 

zw	 Depth below water table 

δ	 Angle of interface friction 

δa 	 active translation 

δb	 

 passive translation 

∆p Pressure due to external loads 

∆u Excess pore water pressure 

γ 

	 Total unit weight 

γ'	 Effective unit weight 

γb Buoyant unit weight (same as effective unit weight) 

γd Dry unit weight 

γs Unit weight of the solid phase 

γsat Saturated unit weight 

γT or γt Total unit weight 

γw Unit weight of water 

φ	 Angle of friction 

φ'	 Effective angle of internal friction 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xix December 2006 
ν Poisson’s ratio 

θ Angle of obliquity 

θm Maximum angle of obliquity 

σ Total stress 

σ' Effective stress 

σn Normal stress 

σn ' Effective normal stress 

τ Shearing strength 

τ' Effective shear stress (strength) 

Chapter 3 

AASHTO American Association of State Highway and Transportation Officials 

AR Area ratio 

ASTM American Society for Testing and Materials 

BPT Becker (Hammer) penetration test 

CN Overburden correction factor or stress normalization parameter 

CPT Cone penetration test 

CPTu, PCPT Piezocone penetration test 

d Displacement 

DC Direct current 

De Diameter of sampler cutting tip 

Di Inside diameter of the sampling tube 

DMT Flat plate dilatometer 

Do Outside diameter of the sampling tube 

DOT Department of Transportation 

Ef Energy efficiency 

ER Energy ratio 

F Force 

FEMA Federal Emergency Management Agency 

FHWA Federal Highway Administration 

fs Sleeve friction 

g Gravitational constant 

GPR Ground penetrating radar 

h Drop height 

i Inclination 

ICR Inside clearance ratio 

ID Inside Diameter 

ISRM International Society of Rock Mechanics 

KE Kinetic energy 

LPT Large penetration test 

m Mass 

N SPT blows per foot 

N160 Overburden-normalized energy-corrected blowcount 

N60 Energy-corrected SPT-N value adjusted to 60% efficiency 

NAVFAC Naval Facilities Engineering Command 

NCR No core recovery 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xx December 2006 
Nmeas N-Value measured in the field 

OD Outside diameter 

PE Potential energy 

PMT Pressuremeter test 

po Vertical effective pressure at the depth where the SPT is performed 

PVC Polyvinyl chloride 

qc Cone tip resistance 

qt Tip resistance 

Rf Friction ratio 

RQD Rock quality designation 

St Sensitivity 

SASW Spectral analysis of surface waves 

SBT Soil behavioral type 

SCPTu Seismic cone piezocone penetration test 

SPT Standard Penetration Test 

SWPPP Storm Water Pollution Prevention Plan 

ts Shear wave time 

um Pore water pressure 

USCS Unified Soil Classification System 

USEPA United States Environmental Protection Agency 

USGS Unites States Geological Survey 

v Impact velocity 

VST Vane shear test 

W Work 

Chapter 4 

AASHTO American Association of State Highway and Transportation Officials 

ASTM American Society for Testing and Materials 

Cc Coefficient of curvature 

Cu Coefficient of uniformity 

D10 Diameter of soil particles of which 10% of the soil is finer 

D30 Diameter of soil particles of which 30% of the soil is finer 

D60 Diameter of soil particles of which 60% of the soil is finer 

DOT Department of Transportation 

EGS Effective grain size 

F Percent passing No. 200 sieve 

FHWA Federal Highway Administration 

GI Group index 

GIS Geographic information system 

GSD Grain size distribution 

IGM Intermediate geomaterial 

Is Point load index 

ISRM International Society of Rock Mechanics 

LI Liquid index 

LL Liquid limit 

N SPT blows per foot 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xxi December 2006 
N60 Energy-corrected SPT-N value adjusted to 60% efficiency 

NAVFAC Naval Facilities Engineering Command 

PI Plasticity index 

PL Plastic limit 

RMR Rock mass rating 

RQD Rock quality designation 

SPT Standard penetration test 

U.S. United States 

USCS Unified Soil Classification System 

USDA United States Department of Agriculture 

Chapter 5 

A Activity index 

A Area 

AASHTO American Association of State Highway and Transportation Officials 

ASTM American Society for Testing and Materials 

B Bulk modulus 

c Cohesion 

C Permeability coefficient 

Cα Coefficient of secondary compression 

c' Effective stress cohesion intercept 

c'cu Effective stress cohesion from CU test 

CBR California Bearing Ratio 

Cc Compression index 

Ccε Modified compression index 

CD Consolidated drained triaxial test 

CF Clay fraction 

ch Coefficient of horizontal consolidation 

co Compacted cohesion 

CP Collapse potential 

Cr Recompression index 

Crε Modified recompression index 

csat Saturated cohesion 

CU Consolidated undrained triaxial test 

cu Apparent cohesion 

cv Coefficient of consolidation 

cv Coefficient of vertical consolidation 

Cα Secondary compression index 

Cαε Modified secondary compression index 

D Distance between contact points of platens 

D Vane diameter 

D10 Diameter of soil particles of which 10% of the soil is finer 

D60 Diameter of soil particles of which 60% of the soil is finer 

De Equivalent core diameter 

Dmax Maximum diameter of soil particle 

Dmin Minimum diameter soil particles 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xxii December 2006 
Dr Relative density 

ds Equivalent diameter 

e Void ratio 

emax Maximum void ratio 

emin Minimum void ratio 

E Young modulus 

Ei Elastic modulus of intact rock 

Em Elastic modulus of rock mass 

emax Maximum void ratio 

emin Minimum void ratio 

eo Initial void ratio 

E or Es Elastic modulus 

FHWA Federal Highway Administration 

g Gravitational constant 

G Shear modulus 

GS Specific gravity 

H Height of vane 

H Soil layer thickness 

H/D Height to diameter ratio 

Ho Initial height of specimen 

iB Angle of taper at the bottom of the vane 

IL Incremental load 

Is Point load strength index 

Is(50) Size-corrected point load strength index 

iT Angle of taper at the top of the vane 

k Hydraulic conductivity 

kPLT Size correction factor 

LI Liquid index 

LIR Load increment ratio 

LL Liquid limit 

LVDT Linear variable differential transducer 

md Man-days 

MPC Modified Proctor compaction 

Ms Mass of solid component of sample 

Mt Total mass 

N Normal stress 

N SPT blows per foot 

N160 Overburden-normalized energy-corrected blowcount 

N60 Energy-corrected SPT-N value adjusted to 60% efficiency 

NAVFAC Naval Facilities Engineering Command 

NC Normally consolidated 

nh Rate of increase of soil modulus with depth 

OC Over consolidated 

OCR Overconsolidation ratio 

OMC Optimum moisture content 

P Breaking load 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xxiii December 2006 
pc Maximum past effective stress 

pc Preconsolidation pressure 

PI Plasticity index 

PL Plastic limit 

po Effective overburden pressure 

pt Total vertical stress 

qc Cone tip resistance 

qu Unconfined compression stress 

RC Relative compaction 

RMR Rock mass rating 

RQD Rock quality designation 

S Degree of saturation 

S, St Sensitivity 

scollapse Collapse settlement 

SL Shrinkage limit 

SPC Standard Proctor compaction 

SPT Standard penetration test 

sr, VST Remolded undrained shear strength (obtained by using VST data) 

st, VST Sensitivity (obtained by using VST data) 

su Undrained shear strength 

su, VST Undrained shear strength (obtained by using VST data) 

su/po Undrained strength ratio 

T Tangential (shear) force 

T Torque (related to VST) 

t Vane edge thickness 

t100 Time corresponding to 100% of primary consolidation 

Tmax Maximum torque (related to VST) 

Tnet Difference between Tmax and Trod 

Trod Rod friction (related to VST) 

u Pore water pressure 

UC Unconfined compression test 

U.S. United States 

USBR United States Bureau of Reclamation 

USCS Unified Soil Classification System 

UU Unconsolidated undrained triaxial test 

V Coefficient of variation 

Vs Volume of soil solids 

VST Vane shear test 

Vt Total volume 

W Specimen width 

w Water content 

wn Natural moisture content 

wopt Optimum moisture content 

Ws Weight of solid component of soil 

Wt Total weight 

Z Depth below ground surface 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xxiv December 2006 
 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

∆e Change in void ratio 

∆Hc Change in height upon wetting 

∆σ Incremental stress 

ε Strain 

γ Unit weight 

γ' Effective unit weight 

γb Buoyant unit weight (same as effective unit weight) 

γd field Field dry unit weight 

γd or γdry Dry unit weight 

γd-max Maximum dry unit weight 

γs Unit weight of solid particles in the soil mass 

γsat Saturated unit weight 

γt or γtot Total unit weight 

γt Moist unit weight of compacted soil 

γw Unit weight of water 

φ Angle of internal friction 

φ' Effective friction angle 

φ Friction 

φ' Peak effective stress friction angle 

φ' cu Effective friction angle from CU test 

φ' r Residual effective stress friction angle 

μ Coefficient of friction 

ν Poisson ratio 

ρ Density 

ρd or ρdry Dry mass density 

ρt or ρtot Total mass density 

ρt Moist (total) mass density 

σ' Effective normal stress 

σc Uniaxial compressive strength 

σn Normal stress 

σ' p Preconsolidation stress 

σvo Total vertical stress 

τ Shear stress 

%C Percent collapse 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xxv December 2006 
Chapter 6 

AASHTO American Association of State Highway and Transportation Officials 

b Unit width 

b Width of slice 

c Cohesion 

c Cohesion component of shear strength 

c Unit cohesion 

c' Effective cohesion 

CD Consolidated drained triaxial test 

cd Developed cohesion 

CU Consolidated undrained triaxial test 

d Depth factor 

D Depth ratio 

Fc Average factor of safety with respect to cohesion 

FHWA Federal Highway Administration 

FS or FOS Factor of safety 

Fφ Average factor of safety with respect to friction angle 

h Depth less than or equal to the depth of saturation 

H Height 

H Height of soil layer in active wedge 

h Slope depth 

H Slope height 

H'w Height of water within the slope 

HFill Fill height 

hi Height of layer at center of slice 

Ht Tension crack height 

hw Depth from groundwater surface to the centroid point on the circle 

Hw Depth of water outside the slope 

Hzone Height of zone 

IN Interslice normal (horizontal) force 

IS Interslice shear (vertical) force 

Ka Coefficient of active earth pressure 

Kp Coefficient of passive earth pressure 

l Arc length of slice base 

Ls Radius of circle 

Lw Level arm distance to the center of rotation 

N Normal force component or total normal force 

N Number of reinforcement layers 

N' Effective normal force component 

Ncf Critical stability number 

No Stability number 

Ns Stability number 

Pa Active force (driving) 

po In-situ vertical effective overburden pressure 

Pp Passive force (resisting) 

q Surcharge load 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xxvi December 2006 
R Moment arm 

Rc	 Coverage ratio of the reinforcement 

RSS Reinforced soil slope 

S Frictional force along failure plane 

S Shear strength along failure plane 

SPT Standard penetration test 

Sv	  Vertical spacing of reinforcement 

T Tangential force component 

Ta Sum of available tensile force per width of reinforcement for all reinforcement 

layers 

tan φ	 Coefficient of friction along failure surface 

TMAX	 

 Maximum design tension 

TS-MAX	  Maximum tensile force 

Tzone	 Maximum reinforced tension required for each zone 

U Pore water force 

u Water pressure on slice base 

u Water uplift pressure against failure surface 

UU Unconsolidated undrained triaxial test 

W	 Weight of slice 

Wi	 

 Partial weight 

WT	 Total slice weight 

α	 Angle between vertical and line drawn from circle center to midpoint of slice 

base 

αw	 Slope of water table from horizontal 

γFill Fill soil unit weight 

μ'w	 Seepage correction factor 

μq	 Surcharge correction factor 

μt	 Tension crack correction factor 

μw	 Submergence correction factor 

σ	 The total normal stress against the failure surface slice base due to the weight of 

soil and water above the failure surface 

ΣWi	  Total weight of slice 

β	 

 Angle of slope 

β	 Inclination of the slope 

φ	 Angle of internal friction 

φ'	 Effective angle of internal friction 

φd	 Developed angle of internal friction 

γ	 Unit weight of soil 

γ	 Unit weight of soil in the active wedge 

γi	  Unit weight of layer i 

γ 

	 Effective unit weight 

γm Moist unit weight 

γsat Saturated unit weight 

γt Total soil unit weight 

γw Unit weight of water 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xxvii December 2006 
σ' n Effective stress between soil grains 

τ Frictional shearing resistance 

τ Total shear strength 

τd Developed shear strength 

Chapter 7 

AASHTO American Association of State Highway and Transportation Officials 

C′ Bearing capacity index 

Cc Compression index 

Ccε Modified compression index 

Cr Mean slope of the rebound laboratory curve 

Crε Modified recompression index 

cv Coefficient of consolidation 

Cα Coefficient of secondary consolidation (determined from lab consolidation test) 

Cαε Modified secondary compression index 

DS Depth of soft soil beneath the toe of the end slope of the embankment 

e Void ratio 

eo Initial void ratio at po 

FHWA Federal Highway Administration 

FSSQ Safety factor against failure by squeezing 

H Height of the fill 

H Thickness of soil layer considered 

Hd Distance to the drainage boundary 

hf Fill height 

Ho Layer thickness 

ID Inner Diameter 

N160 Number of blows per foot corrected for overburden and hammer efficiency 

NCHRP National Cooperative of Highway Research Program 

OCR Over consolidation ratio 

pc Maximum past effective stress 

pc Maximum past vertical pressure (preconsolidation) 

pf Final effective vertical stress at the center of layer n 

pf Final pressure applied to the foundation subsoil 

pf Final stress 

pf Total embanklment pressure 

PI Plasticity index 

po Effective overburden pressure 

po Existing effective overburden pressure 

po Initial effective vertical stress at the center of layer n 

RSS Reinforced soil slope 

S Degree of saturation 

S Settlement 

Sc Settlement due to primary consolidation 

SPT N Number of blows per foot (blow/0.3m) 

SPT Standard penetration test 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xxviii December 2006 
 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

Ss Settlement due to secondary compression 

St Settlement at time t 

su Undrained shear strength of soft soil beneath embankment 

Sultimate Settlement at end of primary consolidation 

t Time 

t1 lab Time when secondary compression begins 

t1 Time when approximately 90% of primary compression has occurred 

t100 Time for 100%of primary consolidation 

t2 lab Arbitrary time on the curve 

t2 The service life of the structure or any time of interest 

t90 Time for 90%of primary consolidation 

Tv Time factor 

U Average degree of consolidation 

us Hydrostatic pore water pressure at any depth 

us Initial hydrostatic pore water pressure 

USACE United States Army Corps of Engineers 

usb Hydrostatic pore water pressure at bottom of layer 

ust Hydrostatic pore water pressure at top of layer 

ut Total pore water pressure at any depth after time t 

ZI Zone of influence 

∆e Change in void ratio 

∆H Settlement 

∆p Distributed embankment pressure 

∆p Load increment 

∆p Stress increase 

∆po Effective vertical stress increment 

∆pt Applied vertical stress increment 

∆u Excess pore water pressure at any depth after time t 

∆ui Initial excess pore water pressure 

εv Vertical strain 

γ Unit weight of fill 

γ' Effective unit weight 

γb Buoyant unit weight (same as effective unit weight) 

γf Fill unit weight 

θ Angle of slope 

FHWA NHI-06-088 Table of Contents 

Soils and Foundations – Volume I xxix December 2006 
[THIS PAGE INTENTIONALLY BLANK]
 
 

CHAPTER 1.0 

INTRODUCTION 

1.1 PURPOSE AND SCOPE 

The Soils and Foundations course is sponsored by the National Highway Institute (NHI) to 

provide practical knowledge in geotechnical and foundation engineering for both civil 

engineering generalists and geotechnical and foundation specialists. The course is developed 

around the design and construction aspects of a highway project that includes bridges, 

earthworks and earth retaining structures. Bridges can range from single span bridges to 

multi-span bridges as part of a stack interchange. Bridges may be constructed over land, in 

which case they are known as viaducts, or over water. Examples of transportation facilities 

that include bridge structures are shown in Figures 1-1 to 1-4. Not all highway projects 

include bridge structures. Figure 1-5 shows an example of a highway corridor without 

bridges and in an environmentally sensitive area. 

Figure 1-1. Aerial view of a pair of 3-span Interstate 10 (I-10) bridges over a local 

roadway in Tucson, Arizona. 

FHWA NHI-06-088 1 – Introduction 

Soils and Foundations – Volume I 1 - 1 December 2006 
 

Figure 1-2. Example of 3-span roadway bridges over another roadway. 

Figure 1-3. The “BIG I” stack interchange at the intersection of I-40 and I-25 in 

Albuquerque, New Mexico (Photo: Courtesy of Bob Meyers, NMDOT) (Note: A stack 

interchange is a free-flowing junction between two or more roadways that allows 

turning in all directions). 

FHWA NHI-06-088 1 – Introduction 

Soils and Foundations – Volume I 1 - 2 December 2006 
 

Figure 1-4. A major multi-span bridge structure over water (George P. Coleman Bridge 

over the York River in Yorktown, Virginia). 

FHWA NHI-06-088 1 – Introduction 

Soils and Foundations – Volume I 1 - 3 December 2006 

 

 

 

 

 

 

Cut Slopes 

Wetlands 

Figure 1-5. Example of a roadway bounded by cut slopes and wetlands in an 

environmentally sensitive area. 
 

Highway projects can involve a full range of geotechnical engineering assessments and 

alternatives depending on the complexity of the project. For example, the foundations for the 

bridge piers and abutments may be shallow foundations, or deep foundations such as driven 

piles and/or drilled shafts. The approach embankments may be unreinforced slopes or 

reinforced soil slopes (RSS). Cut slopes may be in rocks and/or soils. Retaining walls may 

be used at abutments and/or along approaches and may consist of cantilevered walls or 

mechanically stabilized earth (MSE) walls. The ground under the bridge may be soft and 

require improvement. Similarly, the transportation corridor may traverse wetlands and 

special ground improvement measures may be required. Pavements seen in Figures 1-1 to 1-

3 and 1-5 may be constructed of asphaltic concrete (AC), Portland cement concrete (PCC) or 

reinforced cement concrete (RCC) on a variety of subgrade materials1 . 

Recognizing the need for consistent guidance for practitioners involved in the planning, 

design and construction of transportation facilities that include bridges and associated 

structures, the Federal Highway Administration (FHWA) developed the first version of this 

manual in 1982. Subsequently, the manual was revised in 1993 and in 2000. The present 

reference manual, which is the fourth edition, represents a significant update and supersedes 

earlier editions of the manual. In particular, this manual has been updated to reflect the 

current standard of geotechnical practice in the planning, design and construction of 

transportation facilities. As part of this effort, this edition provides guidance consistent with 

that found in the latest FHWA manuals and courses. 

This edition of the manual, like the earlier editions, is geared towards the practicing engineer 

who routinely deals with soils and foundations problems on highway projects but who may 

not have a thorough theoretical background of soil mechanics or foundation engineering. 

The overall goals of this manual are: (i) to explain geotechnical engineering principles, and 

(ii) to provide sound methods and recommendations related to safe, cost-effective design and 

construction of geotechnical features. The reader is encouraged to develop an appreciation 

for the design and construction of geotechnical features in all phases of a project that may 

influence or could be influenced by his/her work (cost, quality, time, and performance). 

Coordination among generalists and specialists in all project phases is stressed. 

The manual contains an appendix (Appendix A) wherein the geotechnical engineering input 

to a bridge project is traced from conception (scoping) to completion (post construction) in a 

serialized illustrative problem that incorporates many of the technical concepts presented in 

the course. The bridge project used in Appendix A is based on an actual project in the State 

of New York. 

 

1

 Pavement structures are not addressed in this manual. 

FHWA NHI-06-088 1 – Introduction 

Soils and Foundations – Volume I 1 - 4 December 2006 
 

1.2 SOILS AND FOUNDATIONS FOR HIGHWAY FACILITIES 

Civilization’s earliest attempts at construction probably involved soil; however, the 

understanding of the role of soil as a foundation or building material developed by trial and 

error. Since the early 20th century, an improved understanding of soil behavior has been 

achieved by applying the principles of physics, solid mechanics, fluid mechanics, strength of 

materials, and structural engineering to define soil behavior. The body of knowledge 

developed by analyzing soil behavior on a theoretically sound basis is called "soil 

mechanics" and its application to solution of actual problems is called “geotechnical 

engineering.” Soil is a complex three-phase medium that contains various amounts of water 

and/or air surrounding the solid particles. It is not a solid mass, i.e., a continuum, as many of 

the theories of solid mechanics require. Therefore, an entirely theoretical solution of the 

most commonly encountered soil problems is not practical. The most practical solution to 

soil problems can be reached by a combination of the sources of information as illustrated in 

Figure 1-6. 

Testing 	 Theory

Experience

Figure 1-6. Combinations of sources of information required to solve geotechnical 

engineering issues. 

1.	 Experience obtained from previous projects can be developed into the empirical or 

"rule of thumb" procedures followed by some engineers/specialists today. Often 

some geotechnical designers rely almost exclusively on experience. The weakness of 

using this approach exclusively is that experience does not always recognize the 

factors that cause differences in the engineering properties of soils. What works well 

at one location may not succeed with the same type of soil at another location because 

of a change in conditions, such as water content. The current state of the practice 

FHWA NHI-06-088 1 – Introduction 

Soils and Foundations – Volume I 1 - 5 December 2006 
 

requires the geotechnical specialist to rely on testing and theory in addition to 

experience or rules of thumb. 

2.	 Testing of representative samples of soil in the field and laboratory is required to 

obtain information on the engineering properties and the characteristics of soils. The 

results of subsequent engineering analyses will be only as good as the soils data used 

as input. 

3.	 Theory based on principles from various fields of engineering and science tempered 

by assumptions to fit reality is used to explain or predict the behavior of soils under 

various conditions. 

The engineering analysis of soils is often more complex than the analysis of other 

construction materials because soil is not a continuum. Therefore, soil typically does not 

strictly meet the assumptions of the theories of solid mechanics and strength of materials. By 

contrast, steel and concrete are relatively uniform solids that have predictable properties. For 

example, the strength of steel is predictable within the elastic range of loading. Even though 

the strength of steel and concrete may be "ordered,” that strength will be essentially constant 

under a wide range of climatic conditions. Structures can then be built of these materials 

with a high degree of confidence regarding the material strength. 

The engineering properties of the soils, on the other hand, can vary widely over time and 

space so that their physical properties cannot be defined accurately at all locations for all 

conditions. Since soils are composed of a mixture of three dissimilar materials - soil solids, 

liquid fluids (usually water), and gaseous fluids (usually air) - their properties are influenced 

by the interaction of these three phases in the soil mass. Some of the factors that influence 

the behavior of soil are: 

1.	 size, shape, and distribution of soil particles, 

2.	 mineralogy, 

3.	 degree of packing of soil particles, 

4.	 amount of water in the soil, 

5.	 climatic conditions, and 

6.	 degree of confinement (i.e., depth). 

In short, engineers should understand that the engineering properties of soils can be 

significantly influenced by many factors. 

The success or failure of a geotechnical feature is often decided in the early stages of a 

project. Geotechnical engineering is a specialized field. Therefore, to assure success of a 

FHWA NHI-06-088 1 – Introduction 

Soils and Foundations – Volume I 1 - 6 December 2006 
 

 

project, the input of a qualified and experienced geotechnical specialist should begin at 

project inception and continue until completion of construction. Geotechnical designs are 

based upon soil properties that are generally defined from a subsurface exploration and 

laboratory testing of a very minute physical sampling of the soils. The volume of site soils 

excavated and exposed during construction is many orders of magnitude greater than that 

from the subsurface explorations. Thus, a great deal of geotechnical information can and 

should be gathered during the construction phase of a project to validate or revise the 

geotechnical design parameters. A geotechnical design should not be considered complete 

until construction has been successfully completed. A geotechnical specialist should also 

be involved during post-construction activities such as instrumentation monitoring, 

participating in resolution of contractor disputes and claims activities, and documenting 

lessons learnt on the project. 

Based on the above considerations, early interactions at a project’s scoping phase among the 

geotechnical specialist, other engineers/specialists, the project manager and the contractor 

will prevent the design of a project element, or even worse the construction of an element, 

such as alignment or grade, that may require costly foundation treatment later. It is 

imperative that good communication and interaction exist among the geotechnical 

specialist, structural specialist, construction specialist, project manager and contractor 

throughout the design and the construction process. Such interactions and involvement 

are required to insure a cost-effective design and to minimize change orders and contract 

disputes resulting from design deficiencies and/or misunderstandings during construction. 

The importance of communication and interaction is stressed throughout this manual 

and cannot be overemphasized. 

The flow chart in Figure 1-7 and the six phases identified in Table 1-1 describe the details of 

geotechnical involvement in a typical project using the design-bid-build (D-B-B) 

procurement process where the geotechnical specialist interacts with the owner to provide 

information to the contractor. There are several other procurement processes such as the 

design-build (D-B) process and the Construction Manager at Risk (CMAR) process. In each 

such alternative process, the geotechnical specialist is concurrently dealing with both the 

owners and the contractors, with the geotechnical specialist’s direct client being the 

contractor. Even though the geotechnical involvement is somewhat different in each of these 

types of procurement processes, it is important to realize that all the items listed in Figure 1-7 

as well as in Table 1-1 must be addressed to achieve a successful project. 

FHWA NHI-06-088 1 – Introduction 

Soils and Foundations – Volume I 1 - 7 December 2006 
 

 

Project Initiation (Scoping) 

Meet with Bridge and Highway Designer to Define Geotechnical Scope 

Review Existing Data 

Obtain Necessary Permits and Perform Site 

Exploration Program 

Prepare Project-Specific Laboratory 

Testing Program 

Perform Analysis and Design 

(Include Evaluation of Alternates and Cost) 

Prepare Geotechnical Design Report 

Prepare Project-Specific Special Provisions 

Review Final Plans, Specifications and Special Provisions to 

Ensure Compliance with Geotechnical Design Report 

Establish Construction Quality Assurance Criteria 

and Monitoring Program 

Troubleshoot Construction Problems 

Provide Post-Construction Services such as Instrumentation 

Monitoring, Resolution of Disputes and Documenting Lessons 

Figure 1-7. Geotechnical activity flow chart for a typical project using design-bid-build 

procurement process. 

FHWA NHI-06-088 1 – Introduction 

Soils and Foundations – Volume I 1 - 8 December 2006 
 

 

Table 1-1 

Geotechnical involvement in project phases
 

Phase Function 

PHASE 1 1. Study project information, scope and existing data. (a) USGS topographic 

Planning sheets. (b) USDA soil maps. (c) groundwater bulletins. (d) air photos. 

2. Conduct site inspection with project manager. (a) inspect nearby structures for 

settlement, scour, etc. (b) assess site conditions. 

3. Prepare terrain reconnaissance report for planning engineer. Include: (a) 

anticipated soil, rock and water conditions. (b) major problems or costs that will 

hinder or preclude construction of the facility. (c) right-of-way required for 

possible special geotechnical treatment. (d) beneficial shifts in alignment. 

PHASE 2 

Design 

Alternatives 

1. Assess facility locations with regard to major soil issues. 

2. Provide input for specific uses, e.g., soil/rock scour. 

3. Implement subsurface exploration and laboratory testing programs after design 

approval. 

PHASE 3 

Prepare 

Detail Plans 

1. Review and interpret subsurface information from field and laboratory work. 

2. Provide preliminary input to bridge/roadway engineer. 

3. Submit report to bridge and roadway engineer summarizing the investigations 

along with recommendations. Include: (a) coordination with roadway 

construction. (b) alternate foundation design. (c) subsurface profile. (d) special 

provisions and specifications. 

PHASE 4 

Final Design 

1. Review final plans 

2. Make appropriate adjustments to geotechnical information if necessary 

PHASE 5 

Construction 

1. Provide geotechnical support to the resident engineer during construction. 

Examples are as follows: 

(A) Driven Piles: (a) submit wave equation analysis to bridge engineer. (b) 

hammer approval. (c) stress analysis. (d) required blow count. (e) special 

effects, etc. 

(B) Drilled Shafts: (a) shaft excavation information, e.g., need for casing or 

slurry. (b) steel placement tolerances. (c) tube placement for integrity testing. 

(d) concreting requirements. (e) post-installation integrity tests, etc. 

(C) Spread footings: (a) evaluation criteria of stiffness of soils at base of footing 

excavation, etc. 

(D) Retaining Walls: (a) construction process based on whether wall is top-down 

or bottom-up construction. (b) backfill compaction requirements, etc. 

(E) Slopes/Embankments: (a) backfill compaction requirements. (b) final grading 

of a slope, etc. 

2. Attend preconstruction meeting with resident engineer and foundation inspector. 

Explain various important geotechnical issues: (a) general geologic profile. (b) 

design basis. (c) wave equation analysis for driven piles. (d) end and skin 

resistance values taking into account strain compatibility for drilled shafts. (e) 

possible geotechnical problems. 

3. Troubleshoot soils-related problems as required. 

4. Assist with structural foundation load tests as required. 

PHASE 6 

Post 

Construction 

1. Review actual pile results versus predicted. Include: (a) blow count for driven 

piles. (b) installation methods for drilled shafts. (c) length. (d) field problems. 

(e) load test capacity. 

2. Participate in contractor disputes and claims activities. 

FHWA NHI-06-088 1 – Introduction 

Soils and Foundations – Volume I 1 - 9 December 2006 
 

1.3 ORGANIZATION OF MANUAL 

The organization of this manual follows a project-oriented approach whereby a typical bridge 

project is traced from scoping stage through design computation of settlement, allowable 

footing pressure, selection of earth retaining structure, to the construction of approach 

embankments, pile driving or shaft drilling operations, etc. Recommendations are presented 

on how to layout borings efficiently, how to minimize approach embankment settlement and 

eliminate the bump at the end of-the bridge, how to design the most cost-effective deep 

foundation, and how to transmit design information properly to contractors directly through 

plans, specifications, and special provisions and/or indirectly through contact with the project 

engineer. 

The concepts presented in various chapters are concise and specifically directed at a 

particular operation in the geotechnical design and construction process. Basic example 

problems are included in several sections to illustrate how concepts are used and for handson knowledge. Continuity between chapters is achieved by sequencing the information in the 

normal progression of a geotechnical project. In addition, the manual contains an appendix 

(Appendix A) with the solution to geotechnical issues, in a serialized format, for a highway 

project involving a bridge and approach embankment over soft ground. In each phase of the 

fictitious project the geotechnical concepts are developed into specific designs or 

recommendations for that segment of the problem. 

The organization of the manual and a summary of the material presented in each chapter 

follow. 

•	 Chapter 1 – this chapter (Introduction) presents the purpose and scope of NHI’s Soils and 

Foundation course and provides introductory material about geotechnical activities 

related to the design and construction aspects of a highway project. 

•	 Chapter 2 (Stress and Strain in Soils) presents basic phase (weight-volume) relationships, 

effective stress principles, computation of overburden pressures, estimating vertical and 

horizontal stresses in soils due to external (superimposed) loads on geomaterials. 

•	 Chapter 3 (Subsurface Explorations) presents basic information on subsurface 

exploration procedures including terrain reconnaissance, subsurface investigation 

methods, standard penetration test procedures, undisturbed soil sampling, and guidelines 

for the geotechnical investigation of both roadway and structure sites. 

FHWA NHI-06-088 1 – Introduction 

Soils and Foundations – Volume I 1 - 10 December 2006 
 

•	 Chapter 4 (Engineering Description, Classification and Characteristics of Soils and 

Rocks) discusses the basic engineering characteristics of the main soil and rock groups, 

and presents procedures for describing and classifying soils and rocks, and for developing 

a subsurface profile. 

•	 Chapter 5 (Laboratory Testing for Geotechnical Design and Construction) presents 

several commonly used laboratory tests for soils and rocks including soil classification, 

basic consolidation and strength testing concepts. This chapter also includes guidelines 

for laboratory testing on a typical highway project, and a procedure for summarizing and 

choosing design values from laboratory tests. 

•	 Chapter 6 (Slope Stability) presents the general procedures for the stability analysis of 

embankments and cut slopes. Basic methods of analysis are shown and explained with 

emphasis on practical application to highway embankments. Stability charts are 

presented for a rapid preliminary evaluation of slope stability. Remedial methods are 

discussed for stability problems. 

•	 Chapter 7 (Approach Roadway Deformations) distinguishes between internal and 

external settlement within and below embankment fills. Recommendations are provided 

for select fill and compaction control for soils placed near abutments. Immediate (i.e., 

short-term) and consolidation (i.e., long-term) settlement, and lateral squeeze are 

discussed and methods of analysis are presented. 

•	 Chapter 8 (Shallow Foundations) presents the FHWA-recommended foundation design 

procedure for shallow foundations in soils and rocks. The analysis for both bearing 

capacity and settlement are discussed and the application of results is presented. 

Economic considerations of shallow versus deep foundations are discussed. 

•	 Chapter 9 (Deep Foundations) discusses basic concepts in the selection and design of 

both driven piles and drilled shafts in soils and rocks. Analyses for skin friction and end 

bearing are addressed for cohesive soil, cohesionless soils and rocks. Foundation 

installation effects on design are discussed as well as group effects, negative skin friction 

and deep foundation settlement. The components of pile driving equipment are 

presented. The use of driving formulae and the wave equation analysis in construction is 

introduced monitoring. Generic information is presented on the use of load testing. 

Construction considerations for drilled shafts are also presented. 

FHWA NHI-06-088 1 – Introduction 

Soils and Foundations – Volume I 1 - 11 December 2006 
 

•	 Chapter 10 (Earth Retaining Structures) presents the basic lateral earth pressure theories, 

briefly introduces various wall types, presents a wall classification system, and presents 

the external stability analysis for a typical fill wall. 

•	 Chapter 11 (Geotechnical Reports) presents outlines for various types of geotechnical 

reports, discussions on subsurface profiles, guidelines on the use of disclaimers, and 

suggestions for how to incorporate geotechnical information into contract documents. 

1.4 REFERENCES 

A detailed list of references is provided in Chapter 12. However, certain primary references 

were used to develop materials for many sections in this document. In addition, FHWA has 

either developed or is in the process of developing detailed guidance in the topic areas 

covered in this document. Most of those documents are reference manuals for geotechnical 

courses developed for the National Highway Institute. Both the FHWA and other primary 

references are listed below. The reader is directed to the web site for the FHWA National 

Geotechnical Team (NGT), http://www.fhwa.dot.gov/engineering/geotech/index.cfm, to 

obtain information on all geotechnical publications and software that have been developed by 

FHWA. The NAVFAC manuals and many other public domain manuals can be downloaded 

from http://www.geotechlinks.com. 

1.4.1 Primary FHWA References 

FHWA (1988). Checklist and Guidelines for Review of Geotechnical Reports and 

Preliminary Plans and Specifications. Report No. FHWA ED-88-053, Federal 

Highway Administration, U.S. Department of Transportation. Revised 2003. 

FHWA (1999). Drilled Shafts: Construction Procedures and Design Methods. Report No. 

FHWA-IF-99-025, Authors: O’Neill, M. W. and Reese, L. C. Federal Highway 

Administration, U.S. Department of Transportation. 

FHWA (2001a). Soil Slopes and Embankment Design Reference Manual. Report No. 

FHWA-NHI-01-026. Authors: Collin, J.G., Hung, J.C., Lee, W.S., and Munfakh, G., 

Federal Highway Administration, U.S. Department of Transportation. 

FHWA (2002a). Geotechnical Engineering Circular 5 (GEC5) - Evaluation of Soil and Rock 

Properties. Report No FHWA-IF-02-034. Authors: Sabatini, P.J, Bachus, R.C, 

FHWA NHI-06-088 1 – Introduction 

Soils and Foundations – Volume I 1 - 12 December 2006 
 

 

Mayne, P.W., Schneider, J.A., Zettler, T.E., Federal Highway Administration, U.S. 

Department of Transportation. 

FHWA (2002b). Subsurface Investigations (Geotechnical Site Characterization). Report No. 

FHWA NHI-01-031, Authors: Mayne, P. W., Christopher, B. R., and DeJong, J., 

Federal Highway Administration, U.S. Department of Transportation. 

FHWA (2002c). Geotechnical Engineering Circular 6 (GEC6), Shallow Foundations. 

Report No. FHWA-SA-02-054, Author: Kimmerling, R.E. 2002, Federal Highway 

Administration, U.S. Department of Transportation. 

FHWA (2006a). Design and Construction of Driven Pile Foundations - Vol. I and II, Report 

No. FHWA-NHI-05-042 and FHWA-NHI-05-043, Authors: Hannigan, P.J., G.G. 

Goble, G. Thendean, G.E. Likins and F. Rausche., Federal Highway Administration, 

U.S. Department of Transportation. 

Geotechnical Engineering Notebook. FHWA Geotechnical Guidelines GT1 –GT16. 

http://www.fhwa.dot.gov/engineering/geotech/index.cfm. 

1.4.2 Other Primary References 

AASHTO (1988). Manual on Foundations Investigations, Standard Specifications for 

Highway Bridges, American Association of State Highway and Transportation 

Officials, Washington, D.C. 

AASHTO (2002). Standard Specifications for Highway Bridges. 17th Edition, American 

Association of State Highway and Transportation Officials, Washington, D.C. 

AASHTO (2004 with 2006 Interims). AASHTO LRFD Bridge Design Specifications, 3rd 

Edition, American Association of State Highway and Transportation Officials, 

Washington, D.C. 

AASHTO (2006). Standard Specifications for Transportation Materials and Methods of 

Sampling and Testing, Parts I and II. American Association of State Highway and 

Transportation Officials, Washington, D.C. 

ASTM (2006). Annual Book of ASTM Standards – Sections 4.02, 4.08, 4.09 and 4.13. 

ASTM International, West Conshohocken, PA. 

FHWA NHI-06-088 1 – Introduction 

Soils and Foundations – Volume I 1 - 13 December 2006 
 

Holtz, R. D. and Kovacs, W. D. (1981). An Introduction to Geotechnical Engineering, 

Prentice Hall, Englewood Cliffs, NJ. 

Lambe, T. W., and Whitman, R. V. (1979). Soil Mechanics: SI Version, John Wiley & Sons, 

Inc., New York, NY. 

NAVFAC (1986a). Design Manual 7.01 - Soil Mechanics, Department of the Navy, Naval 

Facilities Engineering Command, Alexandria, VA. (can be downloaded from 

http://www.geotechlinks.com). 

NAVFAC (1986b). Design Manual 7.02 - Foundations & Earth Structures, Department of 

the Navy, Naval Facilities Engineering Command, Alexandria, VA. (can be 

downloaded from http://www.geotechlinks.com). 

FHWA NHI-06-088 1 – Introduction 

Soils and Foundations – Volume I 1 - 14 December 2006 
 

CHAPTER 2.0 

STRESS AND STRAIN IN SOILS
 

Soil mass is generally a three phase system that consists of solid particles, liquid and gas. 

The liquid and gas phases occupy the voids between the solid particles as shown in Figure 2­

1a. For practical purposes, the liquid may be considered to be water (although in some cases 

the water may contain some dissolved salts or pollutants) and the gas as air. Soil behavior is 

controlled by the interaction of these three phases. Due to the three phase composition of 

soils, complex states of stresses and strains may exist in a soil mass. Proper quantification of 

these states of stress, and their corresponding strains, is a key factor in the design and 

construction of transportation facilities. 

The first step in quantification of the stresses and strains in soils is to characterize the 

distribution of the three phases of the soil mass and determine their inter-relationships. The 

inter-relationships of the weights and volumes of the different phases are important since 

they not only help define the physical make-up of a soil but also help determine the in-situ 

geostatic stresses, i.e., the states of stress in the soil mass due only to the soil’s self-weight. 

The volumes and weights of the different phases of matter in a soil mass shown in Figure 2­

1a can be represented by the block diagram shown in Figure 2-1b. Such a diagram is also 

known as a phase diagram. A block of unit cross sectional area is considered. The symbols 

for the volumes and weights of the different phases are shown on the left and right sides of 

the block, respectively. The symbols for the volumes and weights of the three phases are 

defined as follows: 

Va, Wa : volume, weight of air phase. For practical purposes, Wa = 0. 

Vw, Ww: volume, weight of water phase. 

Vv, Wv : volume, weight of total voids. For practical purposes, Wv = Ww as Wa = 0. 

Vs, Ws : volume, weight of solid phase. 

V, W : volume, weight of the total soil mass . 

Although Wa = 0 so that Wv = Ww, Va is generally > 0 and must always be taken into 

account. Since the relationship between Va and Vw usually changes with groundwater 

conditions as well as under imposed loads, it is convenient to designate all the volume not 

occupied by the solid phase as void space, Vv. Thus, Vv = Va + Vw. Use of the terms 

illustrated in Figure 2-1b, allows a number of basic phase relationships to be defined and/or 

derived as discussed next. 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 1 December 2006 
Soil 

Particles 

Voids 

(filled with 

water and air)
 

Volume Weight 

(a) 

Va Air Wa ≈ 0 

Ww 

Ws 

W 

Vv 

V 

Vw Water 

Vs Solid 

1 unit 

(b) 

Figure 2-1. A unit of soil mass and its idealization. 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 2 December 2006 
 

 

2.1 BASIC WEIGHT-VOLUME RELATIONSHIPS 

Various volume change phenomena encountered in geotechnical engineering, e.g., 

compression, consolidation, collapse, compaction, expansion, etc. can be described by 

expressing the various volumes illustrated in Figure 2-1b as a function of each other. 

Similarly, the in-situ stress in a soil mass is a function of depth and the weights of the 

different soil elements within that depth. This in-situ stress, also known as overburden stress 

(see Section 2.3), can be computed by expressing the various weights illustrated in Figure 2­

1b as a function of each other. This section describes the basic inter-relationships among the 

various quantities shown in Figure 2-1b. 

2.1.1 Volume Ratios 

A parameter used to express of the volume of the voids in a given soil mass can be obtained 

from the ratio of the volume of voids, Vv, to the total volume, V. This ratio is referred to as 

porosity, n, and is expressed as a percentage as follows: 

Vv n = x100 2-1

V 

Obviously, the porosity can never be greater than 100%. As a soil mass is compressed, the 

volume of voids, Vv, and the total volume, V, decrease. Thus, the value of the porosity 

changes. Since both the numerator and denominator in Equation 2-1 change at the same 

time, it is difficult to quantify soil compression, e.g., settlement or consolidation, as a 

function of porosity. Therefore, in soil mechanics the volume of voids, Vv, is expressed in 

relation to a quantity, such as the volume of solids, Vs, that remains unchanging during 

consolidation or compression. This is done by the introduction of a quantity known as void 

ratio, e, which is expressed in decimal form as follows: 

 

V 

e = v 2-2 Vs 

Unlike the porosity, the void ratio can have values greater than 1. That would mean that the 

soil has more void volume than solids volume, which would suggest that the soil is “loose” 

or “soft.” Therefore, in general, the smaller the value of the void ratio, the denser the soil. 

As a practicality, for a given type of coarse-grained soil, such as sand, there is a minimum 

and maximum void ratio. These values can be used to evaluate the relative density, Dr (%), 

of that soil at any intermediate void ratio as follows: 

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Soils and Foundations – Volume I 2 - 3 December 2006 
(emax − e) Dr = x100 2-2a (e − e ) max min 

At e = emax the soil is as loose as it can get and the relative density equals zero. At e = emin 

the soil is as dense as it can get and the relative density equals 100%. Relative density and 

void ratio are particularly useful index properties since they are general indicators of the 

relative strength and compressibility of the soil sample, i.e., high relative densities and low 

void ratios generally indicate strong or incompressible soils; low relative densities and high 

void ratios may indicate weak or compressible soils. 

While the expressions for porosity and void ratio indicate the relative volume of voids, they 

do not indicate how much of the void space, Vv, is occupied by air or water. In the case of a 

saturated soil, all the voids (i.e., soil pore spaces) are filled with water, Vv = Vw. While this 

condition is true for many soils below the ground water table or below standing bodies of 

water such as rivers, lakes, or oceans, and for some fine-grained soils above the ground water 

table due to capillary action, the condition of most soils above the ground water table is 

better represented by consideration of all three phases where voids are occupied by both air 

and water. To express the amount of void space occupied by water as a percentage of the 

total volume of voids, the term degree of saturation, S, is used as follows: 

V S = w x100 2-3 Vv 

Obviously, the degree of saturation can never be greater than 100%. When S = 100%, all the 

void space is filled with water and the soil is considered to be saturated. When S = 0%, 

there is no water in the voids and the soil is considered to be dry. 

2.1.2 Weight Ratios 

While the expressions of the distribution of voids in terms of volumes are convenient for 

theoretical expressions, it is difficult to measure these volumes accurately on a routine basis. 

Therefore, in soil mechanics it is convenient to express the void space in gravimetric, i.e., 

weight, terms. Since, for practical purposes, the weight of air, Wa, is zero, a measure of the 

void space in a soil mass occupied by water can be obtained through an index property 

known as the gravimetric water or moisture content, w, expressed as a percentage as 

follows: 

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Soils and Foundations – Volume I 2 - 4 December 2006 
Ww w = x100 2-4 Ws 

The word “gravimetric” denotes the use of weight as the basis of the ratio to compute water 

content as opposed to volume, which is often used in hydrology and the environmental 

sciences to express water content. Since water content is understood to be a weight ratio in 

geotechnical engineering practice, the word “gravimetric” is generally omitted. Obviously, 

the water content can be greater than 100%. This occurs when the weight of the water in the 

soil mass is greater than the weight of the solids. In such cases the void ratio of the soil is 

generally greater than 1 since there must be enough void volume available for the water so 

that its weight is greater than the weight of the solids. However, even if the water content is 

greater than 100%, the degree of saturation may not be 100% because the water content is a 

weight ratio while saturation is a volume ratio. 

For a given amount of soil, the total weight of soil, W, is equal to Ws + Ww, since the weight 

of air, Wa, is practically zero. The water content, w, can be easily measured by oven-drying 

a given quantity of soil to a high enough temperature so that the amount of water evaporates 

and only the solids remain. By measuring the weight of a soil sample before and after it ahs 

been oven dried, both W and Ws, can be determined. The water content, w, can be 

determined as follows since Wa = 0: 

W − Ws Ww w = = x100 2-4a Ws Ws 

Most soil moisture is released at a temperature between 220 and 230 o

F (105 and 110oC). 

Therefore, to compare reported water contents on an equal basis between various soils and 

projects, this range of temperature is considered to be a standard range. 

2.1.3 Weight-Volume Ratios (Unit Weights) and Specific Gravity 

The simplest relationship between the weight and volume of a soil mass (refer to Figure 2­

1b) is known as the total unit weight, γt, and is expressed as follows: 

W Ww + Ws γ t = = 2-5

V V 

The total unit weight of a soil mass is a useful quantity for computations of vertical in-situ 

stresses. For a constant volume of soil, the total unit weight can vary since it does not 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 5 December 2006 
account for the distribution of the three phases in the soil mass. Therefore the value of the 

total unit weight for a given soil can vary from its maximum value when all of the voids are 

filled with water (S=100%) to its minimum value when there is no water in the voids 

(S=0%). The former value is called the saturated unit weight, γsat; the latter value is 

referred to as the dry unit weight, γd. In terms of the basic quantities shown in Figure 2-1b 

and with reference to Equation 2-5, when Ww = 0 the dry unit weight, γd, can be expressed 

as follows: 

For computations involving soils below the water table, the buoyant unit weight is frequently 

used where: 

W

γd = s 2-6

V 

γ b = γ sat − γ w 2-7

where, γw equals the unit weight of water and is defined as follows: 

In the geotechnical literature, the buoyant unit weight, γb, is also known as the effective unit 

weight, γ', or submerged unit weight, γsub. Unless there is a high concentration of dissolved 

salts, e.g., in sea water, the unit weight of water, γw, can be reasonably assumed to be 62.4 

lb/ft3

 (9.81 kN/m3). 

To compare the properties of various soils, it is often instructive and preferable to index the 

various weights and volumes to unchanging quantities, which are the volume of solids, Vs, 

and the weight of solids, Ws. A ratio of Ws to Vs, is known as the unit weight of the solid 

phase, γs, and is expressed as follows: 

The unit weight of the solid phase, γs, should not be confused with the dry unit weight of the 

soil mass, γd, which is defined in Equation 2-6 as the total unit weight of the soil mass when 

there is no water in the voids, i.e., at S = 0%. The distinction between γs and γd is very subtle, 

W

γ w = w 2-8 Vw 

W

γ S

S = 2-9 VS 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 6 December 2006 
but it is very important and should not be overlooked. For example, for a solid piece of rock 

(i.e., no voids) the total unit weight is γs while the total unit weight of a soil whose voids are 

dry is γd. In geotechnical engineering, γd is more commonly of interest than γs. 

Since the value of γw is reasonably well known, the unit weight of solids, γs, can be expressed 

in terms of γw. The concept of Specific Gravity, G, is used to achieve this goal. In physics 

textbooks, G is defined as the ratio between the mass density of a substance and the mass 

density of some reference substance. Since unit weight is equal to mass density times the 

gravitational constant, G can also be expressed as the ratio between the unit weight of a 

substance and the unit weight of some reference substance. In the case of soils, the most 

convenient reference substance is water since it is one of the three phases of the soil and its 

unit weight is reasonably constant. Using this logic, the specific gravity of the soil solids, 

Gs, can be expressed as follows: 

The bulk specific gravity of a soil is equal to γt / γw. The “bulk specific gravity” is not the 

same as Gs and is not very useful in practice since the γt of a soil can change easily with 

changes in void ratio and/or degree of saturation. Therefore, the bulk specific gravity is 

almost never used in geotechnical engineering computations. 

The value of Gs can be determined in the laboratory, but it can usually be estimated with 

sufficient accuracy for various types of soil solids. For routine computations, the value of Gs 

for sands composed primarily of quartz particles may be taken as 2.65. Tests on a large 

number of clay soils indicate that the value of Gs for clays usually ranges from 2.5 to 2.9 with 

an average value of 2.7. 

2.1.4 Determination and Use of Basic Weight-Volume Relations 

The five relationships, n, e, w, γt and Gs, represent the basic weight-volume properties of 

soils and are used in the classification of soils and for the development of other soil 

properties. These properties and how they are obtained and applied in geotechnical 

engineering are summarized in Table 2-1. A summary of commonly used weight-volume 

(unit weight) relations that incorporate these terms is presented in Table 2-2. 

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Soils and Foundations – Volume I 2 - 7 December 2006 

γ GS = S 2-10

γ w 
 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

 

Table 2-1 

Summary of index properties and their application
 

Property Symbol Units1 How Obtained 

(AASHTO/ASTM) 

Comments and Direct 

Applications 

Porosity n Dim From weight-volume 

relations 

Defines relative volume of 

voids to total volume of soil 

Void Ratio e Dim From weight-volume 

relations Volume change computations 

Moisture Content w Dim By measurement 

(T 265/ D 4959) 

Classification and in weightvolume relations 

Total unit weight 2 γt FL-3 

By measurement or 

from weight-volume 

relations 

Classification and for pressure 

computations 

Specific Gravity Gs Dim By measurement 

(T 100/D 854) Volume computations 

NOTES: 

1 F=Force or weight; L = Length; Dim = Dimensionless. Although by definition, moisture content 

is a dimensionless decimal (ratio of weight of water to weight of solids) and used as such in most 

geotechnical computations, it is commonly reported in percent by multiplying the decimal by 100. 

2 Total unit weight for the same soil can vary from “saturated” (S=100%) to “dry” (S=0%). 

Table 2-2 

Weight-volume relations (after Das, 1990)
 

Unit-Weight Relationship Dry Unit Weight (No Water) Saturated Unit Weight (No Air) 

G (1= - n)(1 + w)

S 

wG 1 + 

(1 + w) G 

= 

1+ e 

(G + Se)

= 

1+ e 

(1 + w) G 

= 

t s w

s 

s w

t 

s w

t 

s w

t 

γ γ

γ

γ

γ

γ

γ

γ

⎟ γ

⎠

⎞ ⎜

⎝

⎛

γγ

γγγ

γ

γ

γ

γ

γγ

γ

γ

γ γ

satd w

wsatd 

w 

d 

s 

s w

d 

d s w

s w

d 

t 

d 

1+ e 

e 

-= 

- n= 

(1+ e) w 

eS 

= 

S 

wG 1+ 

G 

= 

G - n)(1= 

1+ e 

G 

= 

1+ w 

= 

⎟ γ

⎠

⎞ ⎜

⎝

⎛

γγ

γγγ

⎟ γ

⎠

⎞ ⎜

⎝

⎛

⎟

⎠

⎞ ⎜

⎝

⎛

γ

γ ⎟

⎟

⎠

⎞

⎜

⎜

⎝

⎛

γ

γ γ

γ

γ

dsat w

wdsat 

sat w

s w

s 

sat 

sat s w

s w

sat 

1+ e 

e

+= 

+ n= 

1+ e 

1+ w 

w 

e 

= 

G 1+ wG 

1+ w 

= 

[(1= - n)G + n]

1+ e 

(G + e) =

 In above relations, γw refers to the unit weight of water, 62.4 pcf (=9.81 kN/m3

). 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 8 December 2006 
2.1.5 Size of Grains in the Solid Phase 

As indicated in Figure 2-1a, the solid phase is composed of soil grains. One of the major 

factors that affect the behavior of the soil mass is the size of the grains. The size of the 

grains may range from the coarsest (e.g., boulders, which can be 12- or more inches [300 

mm] in diameter) to the finest (e.g., colloids, which can be smaller than 0.0002–inches [0.005 

mm]). Since soil particles come in a variety of different shapes, the size of the grains is 

defined in terms of an effective grain diameter. The distribution of grain sizes in a soil mass 

is determined by shaking air-dried material through a stack of sieves having decreasing 

opening sizes. Table 2-3 shows U.S. standard sieve sizes and associated opening sizes. 

Sieves with opening size 0.25 in (6.35 mm) or less are identified by a sieve number 

which corresponds to the approximate number of square openings per linear inch of the 

sieve (ASTM E 11). 

To determine the grain size distribution, the soil is sieved through a stack of sieves with each 

successive screen in the stack from top to bottom having a smaller (approximately half of the 

upper sieve) opening to capture progressively smaller particles. Figure 2-2 shows a selection 

of some sieves and starting from right to left soil particles retained on each sieve, except for 

the powdery particles shown on the far left, which are those that passed through the last sieve 

on the stack. The amount retained on each sieve is collected, dried and weighed to determine 

the amount of material passing that sieve size as a percentage of the total sample being 

sieved. Since electro-static forces impede the passage of finer-grained particles through 

sieves, testing of such particles is accomplished by suspending the chemically dispersed 

particles in a water column and measuring the change in specific gravity of the liquid as the 

particles fall from suspension. The change in specific gravity is related to the fall velocities 

of specific particle sizes in the liquid. This part of the test is commonly referred to as a 

hydrometer analysis. Because of the strong influence of electro-chemical forces on their 

behavior, colloidal sized particles may remain in suspension indefinitely (particles with sizes 

from 10-3 mm to 10-6 mm are termed “colloidal.”) Sample grain size distribution curves are 

shown in Figure 2-3. The nomenclature associated with various grain sizes (cobble, gravel, 

sand, silt or clay) is also shown in Figure 2-3. Particles having sizes larger than the No. 200 

sieve (0.075 mm) are termed ”coarse-grained” while those with sizes finer than the No. 200 

sieve are termed “fine-grained.” 

The results of the sieve and hydrometer tests are represented graphically on a grain size 

distribution curve or gradation curve. As shown in Figure 2-3, an arithmetic scale is used on 

the ordinate (Y-axis) to plot the percent finer by weight and a logarithmic scale is used on the 

abscissa (X-axis) for plotting particle (grain) size, which is typically expressed in 

millimeters. 

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Soils and Foundations – Volume I 2 - 9 December 2006 
 

 

Table 2-3 

U.S. standard sieve sizes and corresponding opening dimension 

U.S. 

Standard 

Sieve No.1 

Sieve 

Opening 

(in) 

Sieve 

Opening 

(mm) 

Comment 

(Based on the Unified Soil Classification System 

(USCS) discussed in Chapter 4) 

3 0.2500 6.35 

4 0.1870 4.75 • Breakpoint between fine gravels and coarse sands 

• Soil passing this sieve is used for compaction test 

6 0.1320 3.35 

8 0.0937 2.36 

10 0.0787 2.00 • Breakpoint between coarse and medium sands 

12 0.0661 1.70 

16 0.0469 1.18 

20 0.0331 0.850 

30 0.0234 0.600 

40 0.0165 0.425 • Breakpoint between medium and fine sands 

• Soil passing this sieve is used for Atterberg limits 

50 0.0117 0.300 

60 0.0098 0.250 

70 0.0083 0.212 

100 0.0059 0.150 

140 0.0041 0.106 

200 0.0029 0.075 • Breakpoint between fine sand and silt or clay 

270 0.0021 0.053 

400 0.0015 0.038 

Note: 

1. The sieve opening sizes for various sieve numbers listed above are based on Table 1 

from ASTM E 11. Sieves with opening size greater than No. 3 are identified by their 

opening size. Some of these sieves are as follows: 

4.0 in (101.6 mm) 1-1/2 in (38.1 mm) 1⁄2 in (12.7 mm) 

3.0 in (76.1 mm)* 1-1/4 in (32.0 mm) 3/8 in (9.5 mm) 

2-1/2 in (64.0 mm) 1.0 in (25.4 mm) 5/16 in (8.0 mm) 

2.0 in (50.8 mm) 3⁄4 in (19.0 mm)** 

1-3/4 in (45.3 mm) 5/8 in (16.0 mm) 

* The 3 in (76.1 mm) sieve size differentiates between cobbles and coarse gravels. 

**The 3⁄4 in (19 mm) sieve differentiates between coarse and fine gravels. 

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Soils and Foundations – Volume I 2 - 10 December 2006 
Figure 2-2. Example of laboratory sieves for mechanical analysis for grain size 

distributions. Shown (right to left) are sieve nos. 3/8-in (9.5-mm), No. 10 (2.0-mm), No. 40 

(0.425 mm) and No. 200 (0.075 mm). Example soil particle sizes shown at the bottom of the 

photo include (right to left): medium gravel, fine gravel, medium-coarse sand, silt, and clay 

(kaolin) (FHWA, 2002b). 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 11 December 2006 
Gap Graded 

Figure 2-3. Sample grain size distribution curves. 

The logarithmic scale permits a wide range of particle sizes to be shown on a single plot. 

More importantly it extends the scale, thus giving all the grains sizes an approximately equal 

amount of separation on the X-axis. For example, a grain-size range of 4.75 mm (No.4 

sieve) to 0.075 mm (No. 200 sieve) when plotted on an arithmetic scale, will have the 0.075 

mm (No. 200 sieve), 0.105 mm (No. 140 sieve), and 0.150 mm (No. 100) particle size plot 

very close to each other. The logarithmic scale permits separation of grain sizes that makes it 

easier to compare the grain size distribution of various soils. 

The shape of the grain size distribution curve is somewhat indicative of the particle size 

distribution as shown in Figure 2-3. For example, 

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Soils and Foundations – Volume I 2 - 12 December 2006 
•	 A smooth curve covering a wide range of sizes represents a well-graded or non-uniform 

soil. 

•	 A vertical or near vertical slope over a relatively narrow range of particle sizes 

indicates that the soil consists predominantly of the particle sizes within that range of 

particle sizes. A soil consisting of particles having only a few sizes is called a poorlygraded or uniform soil. 

•	 A curve that contains a horizontal or nearly horizontal portion indicates that the soil is 

deficient in the grain sizes in the region of the horizontal slope. Such a soil is called a 

gap-graded soil. 

Well-graded soils are generally produced by bulk transport processes (e.g., glacial till). 

Poorly graded soils are usually sorted by the transporting medium e.g., beach sands by water; 

loess by wind. Gap-graded soils are also generally sorted by water, but certain sizes were not 

transported. 

2.1.6 Shape of Grains in Solid Phase 

The shape of individual grains in a soil mass plays an important role in the engineering 

characteristics (strength and stability) of the soil. Two general shapes are normally 

recognized, bulky and platy. 

2.1.6.1 Bulky Shape 

Cobbles, gravel, sand and some silt particles cover a large range of sizes as shown in Figure 

2-2; however, they are all bulky in shape. The term bulky is confined to particles that are 

relatively large in all three dimensions, as contrasted to platy particles, in which one 

dimension is small as compared to the other two, see Figure 2-4. The bulky shape has five 

subdivisions listed in descending order of desirability for construction 

•	 Angular particles are those that have been freshly broken up and are characterized by 

jagged projections, sharp ridges, and flat surfaces. Angular gravels and sands are 

generally the best materials for construction because of their interlocking 

characteristics. Such particles are seldom found in nature, however, because physical 

and chemical weathering processes usually wear off the sharp ridges in a relatively 

short period time. Angular material is usually produced artificially, by crushing. 

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Soils and Foundations – Volume I 2 - 13 December 2006 
•	 Subangular particles are those that have been weathered to the extent that the sharper 

points and ridges have been worn off. 

•	 Subrounded particles are those that have been weathered to a further degree than 

subangular particles. They are still somewhat irregular in shape but have no sharp 

corners and few flat areas. Materials with this shape are frequently found in stream 

beds. If composed of hard, durable particles, subrounded material is adequate for 

most construction needs. 

•	 Rounded particles are those on which all projections have been removed, with few 

irregularities in shape remaining. The particles resemble spheres and are of varying 

sizes. Rounded particles are usually found in or near stream beds or beaches. 

•	 Well rounded particles are rounded particles in which the few remaining irregularities 

have been removed. Like rounded particles, well rounded particles are also usually 

found in or near stream beds or beaches. 

Figure 2-4. Terminology used to describe shape of coarse-grained soils (Mitchell, 1976). 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 14 December 2006 
 

2.1.6.2 Platy Shape 

Platy, or flaky, particles are those that have flat, plate like grains. Clay and some silts are 

common examples. Because of their shape, flaky particles have a greater surface area than 

bulky particles, assuming that the weights and volumes of the two are the same. For 

example, 1 gram of bentonite (commercial name for montmorillonite clay) has a surface area 

of approximately 950 yd2

 (800 m2) compared to a surface area of approximately 0.035 yd2 

(0.03 m2

) of 1 gram of sand. Because of their mineralogical composition and greater specific 

surface area, most flaky particles also have a greater affinity for water than bulky particles. 

Due to the high affinity of such soils for water, the physical states of such fine-grained soils 

change with the amount of water in these soils. The effect of water on the physical states of 

fine grained soils is discussed next. 

2.1.7 Effect of Water on Physical States of Soils 

For practical purposes, the two most dominant phases are the solid phase and the water 

phase. It is intuitive that as the water content increases, the contacts between the particles 

comprising the solid phase will be “lubricated.” If the solid phase is comprised of coarse 

particles, e.g. coarse sand or gravels, then water will start flowing between the particles of 

the solid phase. If the solid phase is comprised of fine-grained particles, e.g., clay or silt, 

then water cannot flow as freely as in the coarse-grained solid phase because pore spaces are 

smaller and solids react with water. However, as the water content increases even the finegrained solid phase will conduct water and under certain conditions the solid phase itself will 

start deforming like a viscous fluid, e.g., like a milk shake or a lava flow. The mechanical 

transformation of the fine-grained soils from a solid phase into a viscous phase is a very 

important concept in geotechnical engineering since it is directly related to the load carrying 

capacity of soils. It is obvious that the load carrying capacity of a solid is greater than that of 

water. Since water is contained in the void space, the effect of water on the physical states of 

fine-grained soils is important. Some of the basic index properties related to the effect of 

water are described next. 

The physical and mechanical behavior of fine-grained-soils is linked to four distinct states: 

solid, semi-solid, plastic and viscous liquid in order of increasing water content. Consider a 

soil initially in a viscous liquid state that is allowed to dry uniformly. This state is shown as 

Point A in Figure 2-5, which shows a plot of total volume versus water content. As the soil 

dries, its water content reduces and consequently so does its total volume as the solid 

particles move closer to each other. As the water content reduces, the soil can no longer flow 

like a viscous liquid. Let us identify this state by Point B in Figure 2-5. The water content at 

Point B is known as the “Liquid Limit” in geotechnical engineering and is denoted by LL. 

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Soils and Foundations – Volume I 2 - 15 December 2006 
As the water content continues to reduce due to drying, there is a range of water content at 

which the soil can be molded into any desired shape without rupture. In this range of water 

content, the soil is considered to be “plastic.” 

Volume, V Shrinkage Plastic Liquid 

Limit (SL) Limit (PL) Limit (LL) A 

B 

C 

D 

Water 

Solid Semi-solid Plastic Liquid Content, w 

Phase Phase Phase Phase 

Figure 2-5. Conceptual changes in soil phases as a function of water content. 

If the soil is allowed to dry beyond the plastic state, the soil cannot be molded into any shape 

without showing cracks, i.e., signs of rupture. The soil is then in a semi-solid state. The 

water content at which cracks start appearing when the soil is molded is known as the 

“Plastic Limit.” This moisture content is shown at Point C in Figure 2-5 and is denoted by 

PL. The difference in water content between the Liquid Limit and Plastic Limit, is known as 

the Plasticity Index, PI, and is expressed as follows: 

PI = LL – PL 2-11

Since PI is the difference between the LL and PL, it denotes the range in water content over 

which the soil acts as a plastic material as shown in Figure 2-5. 

As the soil continues to dry, it will be reduced to its basic solid phase. The water content at 

which the soil changes from a semi-solid state to a solid state is called the Shrinkage Limit, 

SL. No significant change in volume will occur with additional drying below the shrinkage 

limit. The shrinkage limit is useful for the determination of the swelling and shrinkage 

characteristics of soils. 

FHWA NHI-06-088 2 – Stress and Strain in Soils 

Soils and Foundations – Volume I 2 - 16 December 2006 
The liquid limit, plastic limit and shrinkage limit are called Atterberg limits after A. 

Atterberg (1911), the Swedish soil scientist who first proposed them for agricultural 

applications. 

For foundation design, engineers are most interested in the load carrying capacity, i.e., 

strength, of the soil and its associated deformation. The soil has virtually no strength at the 

LL, while at water contents lower than the PL (and certainly below the SL) the soil may have 

considerable strength. Correspondingly, soil strength increases and soil deformation 

decreases as the water content of the soil reduces from the LL to the SL. Since the Atterberg 

limits are determined for a soil that is remolded, a connection needs to be made between 

these limits and the in-situ moisture content, w, of the soil for the limits to be useful in 

practical applications in foundation design. One way to quantify this connection is through 

the Liquidity Index, LI, that is given by: 

w − PL LI = 2-12

PI 

The liquidity index is the ratio of the difference between the soil’s in-situ water content and 

plastic limit to the soil’s plasticity index. The various phases shown in Figure 2-5 and 

anticipated deformation behavior can now be conveniently expressed in terms of LI as shown 

in Table 2-4. 

Table 2-4 

Concept of soil phase, soil strength and soil deformation based on Liquidity Index 

Liquidity 

Index, LI Soil Phase Soil Strength (Soil Deformation) 

LI ≥ 1 Liquid Low strength 

(Soil deforms like a viscous fluid) 

0 < LI < 1 Plastic 

Intermediate strength 

• at w ≈ LL, the soil is considered soft and very compressible 

• at w ≈ PL, the soil is considered stiff 

(Soil deforms like a plastic material) 

LI ≤ 0 Semi-solid to 

Solid 

High strength 

(Soil deforms as a brittle material, i.e., sudden, fracture of 

material) 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 17 December 2006 
Another valuable tool in assessing the characteristics of a fine-grained soil is to compare the 

LL and PI of various soils. Each fine-grained soil has a relatively unique value of LL and 

PI. A plot of PI versus LL is known as the Plasticity Chart (see Figure 2-6). Arthur 

Casagrande, who developed the concept of the Plasticity Chart, had noted the following 

during the First Pan American Conference on Soil Mechanics and Foundation Engineering 

(Casagrande, 1959). 

I consider it essential that an experienced soils engineer should be able to judge 

the position of soils, from his territory, on a plasticity chart merely on the basis 

of his visual and manual examination of the soils. And more than that, the 

plasticity chart should be for him like a map of the world. At least for certain 

areas of the chart, that are significant for his activities, he should be well 

familiar. The position of soils within these areas should quickly convey to him a 

picture of the significant engineering properties that he should expect. 

Figure 2-6. Plasticity chart and significance of Atterberg Limits (NAVFAC, 1986a). 

Casagrande proposed the inclusion of the A-line on the plasticity chart as a boundary 

between clay (above the A-line) and silt (below the A-line) to help assess the engineering 

characteristics of fine-grained soils. Once PI and LL are determined for a fine-grained soil at 

a specific site, a point can be plotted on the plasticity chart that will allow the engineer to 

develop a feel for the general engineering characteristics of that particular soil. The plasticity 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 18 December 2006 
chart also permits the engineer to compare different soils across the project site and even 

between different project sites. (The symbols for soil groups such as CL and CH are 

discussed later in this manual.) The plasticity chart, including the laboratory determination 

of the various limits (LL, PL and SL), are discussed further in Chapters 4 and 5. Additional 

useful terms such as “Activity Ratio” that relate the PI to clay fraction are also introduced in 

Chapter 5. 

2.2 PRINCIPLE OF EFFECTIVE STRESS 

The contacts between soil grains are effective in resisting applied stresses in a soil mass. 

Under an applied load, the total stress in a saturated soil sample is composed of the 

intergranular stress and the pore water pressure. When pore water drains from a soil, the 

contact between the soil grains increases, which increases the level of intergranular stress. 

This intergranular contact stress is called the effective stress. The effective stress, po, 

within a soil mass is the difference between the total stress, pt, and pore water pressure, 

u. The principle of effective stress is a fundamental aspect of geotechnical engineering and 

is written as follows: 

po = pt - u 2-13 

In general, soil deposits below the ground water table will be considered saturated and the 

ambient pore water pressure at any depth may be computed by multiplying the unit weight of 

water, γw, by the height of water above that depth. The total stress at that depth may be 

found by multiplying the total unit weight of the soil by the depth. The effective stress is the 

total stress minus the pore water pressure. This concept is used to construct the profile of 

pressure in the ground as a function of depth and is discussed next. 

2.3 OVERBURDEN PRESSURE 

Soils existing at a distance below ground are affected by the weight of the soil above that 

depth. The influence of this weight, known generally as overburden, causes a state of stress 

to exist, which is unique at that depth, for that soil. This state of stress is commonly referred 

to as the overburden or in-situ or geostatic state of stress. When a soil sample is removed 

from the ground, as during the field exploration phase of a project, that in-situ state of stress 

is relieved as all confinement of the sample has been removed. In laboratory testing, it is 

important to reestablish the in-situ stress conditions and to study changes in soil properties 

when additional stresses representing the expected design loading are applied. The stresses 

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Soils and Foundations – Volume I 2 - 19 December 2006 
to be used during laboratory testing of soil samples are estimated from either the total or 

effective overburden pressure. The engineer's first task is determining the total and effective 

overburden pressure variation with depth. This relatively simple task involves estimating the 

average total unit weight for each soil layer in the soil profile, and determining the depth of 

the water table. Unit weight may be reasonably well estimated from tests on undisturbed 

samples or from standard penetration N-values and visual soil identification. The water table 

depth, which is typically recorded on boring logs, can be used to compute the hydrostatic 

pore water pressure at any depth. The total overburden pressure, pt, is found by 

multiplying the total unit weight of each soil layer by the corresponding layer thickness and 

continuously summing the results with depth. The effective overburden pressure, po, at 

any depth is determined by accumulating the weights of all layers above that depth with 

consideration of the water level conditions at the site as follows: 

Soils above the water table 

•	 Multiply the total unit weight by the thickness of each respective soil layer above the 

desired depth, i.e., po = pt. 

Soils below the ground water table 

 

•	 Compute pore water pressure u as zw γw where zw is the depth below ground water 

table and γw is the unit weight of water 

•	 To obtain effective overburden pressure, po, subtract pore water pressure, u, from pt 

•	 For soils below the ground water table, pt is generally assumed to be equal to psat 

 Alternatively, the following approach can be used: 

•	 Reduce the total unit weights of soils below the ground water table by the unit weight 

of water (62.4 pcf (9.8 kN/m3

)), i.e., use effective unit weights, γ', and multiply by the 

thickness of each respective soil layer between the water table and the desired depth 

below the ground water table, i.e., po = (γt - γw ) (depth), or γ' (depth). 

In the geotechnical literature, the effective unit weight, γ', is also known as the buoyant unit 

weight or submerged unit weight and symbols, γb or γsub, respectively are used. 

An example is solved in Figure 2-7. 

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Soils and Foundations – Volume I 2 - 20 December 2006 
 

 

Example 2-1: Find po at 20 ft below ground in a sand deposit with a total unit weight of 110 pcf and 

the water table 10 ft below ground. Assume γt

 = γsat. Plot pt and po versus depth 

from 0 ft – 20 ft. 

0 ft 

γt

 = 110 pcf 

10 ft 

20 ft 

Solution: 

From Equation 2-13, po = pt – u 

pt @ 10 ft = po @ 10 ft = 10 ft × 110 pcf = 1,100 psf 

pt @ 20 ft = pt @ 10 ft + (10 ft × 110 pcf) = 2,200 psf 

u @ 20 ft = 10 ft × 62.4 pcf = 624 psf 

po @ 20 ft = pt @ 20 ft - u @ 20 ft = 2,200 psf – 624 psf = 1,576 psf 

Pressure (psf)
 

1,000 2,000 3,000 0 

10 

Depth (ft) 

20 

po = pt 

1,100 

po 

1,576 

pt 

u 

2,200 

Pressure Diagram 

A plot of effective overburden pressure versus depth is called a “po – diagram” and is used 

throughout all aspects of geotechnical testing and analysis. 

Figure 2-7. Example calculation of a po-diagram. 

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Soils and Foundations – Volume I 2 - 21 December 2006 
2.4 VERTICAL STRESS DISTRIBUTION IN SOIL DUE TO EXTERNAL 

LOADINGS 

When a load is applied to the soil surface, it increases the vertical and lateral stresses within 

the soil mass. The increased stresses are greatest directly under the loaded area and dissipate 

within the soil mass as a function of distance away from the loaded area. This is commonly 

called spatial attenuation of applied loads. A schematic of the vertical stress distribution with 

depth along the centerline under an embankment of height, h, constructed with a soil having 

total unit weight, γt, is shown in Figure 2-8. 

p = γth 

0.6p 

0.4p 

0.8p 

0.2p 

h 

Figure 2-8. Schematic of vertical stress distribution under embankment loading. 

Graphic generated by FoSSA (2003) program. 

(Note: Version 1.0 of FoSSA program is licensed to FHWA. See Appendix E for a brief 

overview of the FoSSA program).
 

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Soils and Foundations – Volume I 2 - 22 December 2006 
Estimation of vertical stresses at any point in a soil mass due to external loadings are of great 

significance in the prediction of volume change of soils (e.g., settlement) under buildings, 

bridges, embankments and many other structures. The computation of the total vertical stress 

change induced by an external loading will depend on the configuration of the external loads. 

Common examples of the external loads are as follows: 

•	 Uniform strip loads such as the load on a long wall footing of sufficient width, 

•	 Uniformly loaded square, rectangular or circular footings such as column footings of 

buildings, pier footings, footings for water tanks, mats, etc., and 

•	 Triangular and/or trapezoidal strip loads such as the loads of long earth embankments. 

The theory of elasticity is often used to compute the stresses induced within a soil mass by 

external loadings. The most widely used elastic formulae were first developed by 

Boussinesq (1885) for point loads acting at the surface of a semi-infinite elastic half-space. 

These formulae, often known as Boussinesq solutions, can be integrated to give stresses 

below external loadings acting on a finite area. The basic assumptions in these formulae are 

(a) the stress is proportional to strain, (b) the soil is homogeneous (i.e., the properties are 

constant throughout the soil mass), and (c) the soil is isotropic (i.e., the properties are the 

same in all directions through a point). Westergaard (1938) modified the Boussinesq 

solutions by assuming that the semi-infinite elastic half-space is interspersed with infinitely 

thin but perfectly rigid layers that allow vertical movement but no lateral movement. In 

reality, a soil mass never fulfills the assumptions of either of these idealized solutions. 

Nevertheless, these elastic solutions, with appropriate modifications and judgment, have been 

found to yield acceptable approximate estimates of stresses in the soil mass and are widely 

used in geotechnical engineering practice. The Boussinesq solutions are generally used in 

most situations, even those where layered soils are encountered provided the thickness of the 

layers is on the order of a few feet or more. On the other hand, the Westergaard solutions are 

usually used for varved clays where the predominant soil mass is clay interspersed with thin 

layers of sand whose thickness is on the order of inches. 

The derivations of the equations for various common loadings cited above are tedious. They 

are omitted in this manual so that the reader can concentrate on the use of published 

solutions, generally in the form of charts. The following sections contain the chart solutions 

for some of the loadings most commonly encountered in practice. Caution in the use of these 

charts is advised since they all pertain to stress increments at very well-defined points 

within the soil mass due to the applied pressures indicated. The total stress acting at a 

point of interest is equal to the stress increment at that point due to the newly applied 

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Soils and Foundations – Volume I 2 - 23 December 2006 
 

load plus existing stresses at that point due to the geostatic stress and stresses due to 

other external loads applied previously. 

2.4.1 Uniformly Loaded Continuous (Strip) and Square Footings 

A loaded area is considered to be infinitely long when its length, L, to width, B, ratio is 

greater than or equal to 10, i.e., L/B ≥ 10. The load on such an area is commonly known as a 

strip load. Figure 2-9 presents vertical pressure isobars under strip and square footings based 

on Boussinesq’s theory. An isobar is a line that connects all points of equal stress increment 

below the ground surface. In other words, an isobar is a stress increment contour. 

Each isobar represents a fraction of the stress applied at the surface and delineates the zone of 

influence of the footing such that the area contained within two adjacent isobars experiences 

stresses greater than the lower isobar and less than the upper isobar. Since these isobars form 

closed figures that resemble the form of a bulb, they are also termed bulbs of pressure or 

simply the pressure bulbs. The pressure bulb concept gives the user a feel for the spread of 

the stresses through a soil mass. 

According to linear elastic theory, the size of the pressure bulb is proportional to the size of 

the loaded area. This is a key concept in geotechnical engineering that is used to evaluate the 

depth of significant influence, DOSI, denoted by DS of an applied surface load. The depth 

DS is a finite depth below which there are no significant strains in the soil mass due to the 

loads imposed at the surface. Typically, strains are not significant once the stresses have 

attenuated to a value of 10 to 15% of those at the surface. For example, Figure 2-9a shows 

that for “infinitely long” strip footings, DS = 4 to 6B, while for square footings, Figure 2-9b 

shows that DS = 1.5 to 2B. The depths corresponding to this 10 to 15% criterion can be used 

to determine the minimum depth of field exploration for proposed strip or square footings to 

ensure that the anticipated significant depth is explored. 

It may be seen from Figure 2-9 that the effect of the vertical stresses extends laterally beyond 

the width of the loaded area, B. This observation is very useful in assessing the influence of 

one loaded area on the other. Alternatively, this observation can be used to determine an 

adequate spacing between adjacent loaded areas. It also indicates that the effect of 

construction activities may be felt beyond a specific site. Such effects should be evaluated 

before construction so that mitigation measures can be taken to avoid legal implications. 

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Soils and Foundations – Volume I 2 - 24 December 2006 
Figure 2-9. Vertical stress contours (isobars) based on Boussinesq’s theory for 

continuous and square footings (modified after Sowers, 1979; AASHTO, 2002). 

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Soils and Foundations – Volume I 2 - 25 December 2006 
2.4.2 Approximate (2:1) Stress Distribution Concept 

As an approximation to the exact solution given by the Boussinesq charts, the total load at 

the surface of the soil mass may be distributed over an area of the same shape as the loaded 

area on the surface, but with dimensions that increase with depth at a rate of one horizontal 

unit for every two vertical units. This is illustrated in Figure 2-10, which shows a rectangular 

area of dimensions B x L at the surface. At a depth, z, the total load is assumed to be 

uniformly distributed over an area (B+z) by (L+z). Since the stress is distributed at the rate 

of 2:1 (vertical:horizontal), this approximation method it is commonly known as the “2:1 

stress distribution” method. 

The relationship between the approximate distribution of stress determined by this method 

and the exact distribution is illustrated in Figure 2-10. In this figure, the vertical stress 

distribution at a depth B below a uniformly loaded square area of width B is shown along a 

horizontal line that passes beneath the center of the area and extends beyond the edges of the 

loaded area. Also shown is the approximate uniform distribution at depth B determined by 

the 2:1 stress distribution method described above. The discrepancy between the two 

methods decreases as the ratio of the depth considered to the size of the loaded area increases 

(Perloff and Baron, 1976). 

2.5 REPRESENTATION OF IMPOSED PRESSURES ON THE po DIAGRAM 

The pressure distributions computed by using the charts in Section 2.4, can be shown 

superimposed on the po diagram as shown in Figure 2-7. As discussed in the previous 

sections, an applied pressure at the surface causes stress increments within the soil mass that 

decrease with depth due to spatial attenuation. This is shown in Figure 2-11 where ∆p is 

plotted with respect to the po line that represents the existing geostatic stress distribution. As 

can been seen in Figure 2-11, ∆p approaches the po line, which indicates that at a sufficient 

depth the effect of the externally imposed loads reduces significantly. In other words, this 

means that most of the strain due to the increased stress from the applied load will be 

experienced at relatively shallow depths below the load. As noted earlier, this depth is 

known as the depth of significant influence (DOSI), DS. Also, as indicated previously, DS 

depends on the load and load configuration as demonstrated by the pressure distribution 

charts in Section 2.4. Figure 2-11 also shows that that the final stress, pf, in the soil mass at 

any depth is equal to po + ∆p. 

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Soils and Foundations – Volume I 2 - 26 December 2006 
(a) 

(b) 

“Exact” Distribution based 

on theory of elasticity 

∆p 

∆p 

Figure 2-10. Distribution of vertical stress by the 2:1 method (after Perloff and Baron, 

1976). 

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Soils and Foundations – Volume I 2 - 27 December 2006 
A chart such as that shown in Figure 2-11 is even more useful when the soil stratigraphy is 

plotted on it. Then the stress levels in various layers will be clearly identified, which can 

help the engineer determine depth of borings to collect subsurface information within DOSI 

as well as perform proper analysis. 

Example 2-2 illustrates these concepts by providing calculations of pf with depth due to stress 

increments from a strip load and presenting the results of the calculations on a po-diagram. 

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Soils and Foundations – Volume I 2 - 28 December 2006 

 

 

 

 

Pressure 

hw 

po pt 

∆p 

Legend: 

hw = depth to groundwater 

po = effective overburden pressure 

pt = total overburden pressure 

∆p = pressure due to external loads 

pf = po + ∆p 

pf 

Depth, z 

Figure 2-11. Combined plot of overburden pressures (total and effective) and pressure 

due to imposed loads. 
 

Example 2-2: For the Example 2-1 shown in Figure 2-7, assume that a 5 ft wide strip footing with a 

loading intensity of 1,000 psf is located on the ground surface. Compute the stress 

increments, ∆p, under the centerline of the footing and plot them on the po diagram 

shown in Figure 2-7 down to a depth of 20 ft. 

Solution: 

For the strip footing, use the left chart in Figure 2-9. As per the terminology of the chart in Figure 2­

9, B = 5 ft and qo= 1,000 psf. Compile a table of stresses for various depths and plot as follows: 

0 ft 

10 ft 

20 ft 

γt

 = 110 pcf 

pt 

po 

po = pt 

1,576 

1,100 

2,200 

1,000 0 

u 

20 

2,000 3,000 Pressure (psf)

Depth (ft) 10 

Pressure Diagram 

5 

15 

pf 

Figure 2-12. Example calculation of pf with stress increments from strip load on po -

diagram. 

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Soils and Foundations – Volume I 2 - 29 December 2006 

Depth 

z, ft z/B 

Isobar 

Value, x 

Stress, ∆p 

= x(qo), psf po, psf pf= po+∆p psf 

2.5 0.5 0.80 800 (110)(2.5)=275 1,075 

5.0 1.0 0.55 550 (110)(5.0)=550 1,100 

7.5 1.5 0.40 400 (110)(7.5)=825 1,225 

10.0 2.0 0.32 320 (110)(10.0)=1,100 1,420 

12.5 2.5 0.25 250 1,100+(12.5-10.0)(110-62.4)=1,219 1,469 

15.0 3.0 0.20 200 1,100+(15.0-10.0)(110-62.4)=1,338 1,538 

17.5 3.5 0.18 180 1,100+(17.5-10.0)(110-62.4)=1,457 1,637 

20.0 4.0 0.16 160 1,100+(20.0-10.0)(110-62.4)=1,576 1,736 
2.6 LOAD-DEFORMATION PROCESS IN SOILS 

When subjected to static and/or dynamic loads, soils deform mainly because of a change in 

void volume rather than through deformation of the soil solids. When the void volume 

decreases the soil is said to compress, consolidate, collapse or compact. There is an 

important distinction between these four mechanisms although conceptually they appear to 

be the same since each pertains to a reduction in volume. 

 

•	 Compression: Compression is defined as a relatively rapid decrease in void volume 

that partially saturated (unsaturated) soils undergo as air is expelled from the voids 

during loading. 

•	 Consolidation: Consolidation is generally defined as a time-dependent decrease in 

void volume that saturated and near-saturated soils undergo as water is expelled from 

the voids during loading. The conceptual process of consolidation is discussed in 

Section 2.6.1. 

•	 Collapse: Collapse is primarily related to soil structure and its response to an increase 

in water content that results in a rapid decrease in void volume. Collapse-susceptible 

soils characteristically have dry densities less than approximately 100 pcf (16 kN/m3

) 

that suggest high void ratios. Their structure is like a honeycomb with fine-grained 

“bridges” connecting coarser-grained particles. When dry, these soils are able to 

sustain externally applied loads with very little deformation. However, upon being 

wetted they tend to undergo a rapid decrease in void volume as the fine-grained 

“bridges” lose strength and the entire structure collapses. The magnitude of the 

potential collapse increases with increasing load. One of the important things to note 

is that full saturation (S=100%) is not required for these types of soils to collapse. 

Often collapse occurs at a degree of saturation of 50 to 70%. Collapse-susceptible 

soils are very common in the southwest and midwest of the United States and in 

many other parts of the world. 

•	 Compaction: Compaction is the name given to the compression that takes place 

generally under an impact-type loading (e.g., modified and standard Proctor), a static 

loading (e.g., rubber-tired or steel drum rollers) or kneading-type loading (e.g., 

sheepsfoot roller). Most commonly, the compaction processes are deliberate and 

intended to achieve a dense packing of soil particles. Regardless of the type of 

loading, the moisture content of the soil being compacted is far enough below the 

saturation moisture content that the compaction mechanism is considered to be 

related to compression (i.e., expulsion of air) rather than consolidation (i.e., expulsion 

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Soils and Foundations – Volume I 2 - 30 December 2006 
of water) from the voids. Typically, the desired moisture content in the case of 

compaction is slightly above or below the PL. If the moisture content of the soil 

being compacted gets to be too close to the saturation moisture content then 

“pumping” will occur, i.e. water in addition to air will be forced out of the soil. 

These distinctions in load-deformation processes should be kept in mind during the 

discussions that follow in subsequent sections of this chapter. 

Finally, in contrast to the above processes that involve void volume decrease, there are 

conditions under which soils may actually increase in volume. When the void volume 

increases under static and/or dynamic load, the soil is said to dilate. Dilation can occur in 

either saturated or partially saturated soils. It is a function of the initial void ratio, 

confinement stress, and the magnitude and direction of the loading/unloading imposed on the 

soil. Expansion, on the other hand, is generally considered to be due to the presence of 

expansive clay minerals, such as montmorillonite (commercially known as “bentonite”), in 

the soil and the response of these minerals to the introduction of water into the void spaces. 

The physico-chemical properties of expansive clay minerals cause inter-particle repulsions to 

take place in the presence of water so that even under considerable externally applied loads 

these soils will undergo an increase in void volume that leads to swelling. A variation of the 

expansion is heave which can occur due to various factors such as frost action or reduction in 

overburden pressure due to excavation. 

2.6.1 Time Dependent Load-Deformation (Consolidation) Process 

Deformation of a saturated soil is more complicated than that of a dry soil since water, which 

fills the voids, must be squeezed out of the soil before readjustment of the soil grains can 

occur. The more permeable a soil is, the faster the deformation under load will occur. 

However, when the load on a saturated soil is quickly increased, the increase is initially 

carried by the pore water resulting in the buildup of an excess pore water pressure, ∆u. 

Excess pore water pressure is water pressure greater than the hydrostatic pressure. As 

drainage of the water takes place more and more load is gradually transferred from the pore 

water to the soil grains until the excess pore water pressure has dissipated completely and the 

soil grains readjust to a denser configuration under the applied load. This time dependent 

process is called consolidation and results in a decreased void ratio and greater unit weight 

relative to conditions before the load was applied. To illustrate this concept, onedimensional (vertical) drainage of the water will be considered here. The process is 

analogous to loading a spring-supported piston in a cylinder filled with water. The springpiston analogy is shown schematically in Figure 2-13 and is briefly discussed below. 

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Soils and Foundations – Volume I 2 - 31 December 2006 
 

Before application of load After application of load 

Valve shut Valve open 

Water 

Spring 

Valve open Piston 

W 

W 

At time t = 0 At time t = x 

Excess water pressure = ∆u Excess water pressure = ∆u Excess water pressure, ∆u → 0 

Load in spring = weight of piston Load in spring = weight of piston Load in spring → W + weight 

of piston 

Applied Force, W W 

Force 

Spring Force 

Excess Pore Water Pressure, ∆u 

Time, t 

Figure 2-13. Spring-piston analogy for the consolidation process in fine-grained soils. 

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Soils and Foundations – Volume I 2 - 32 December 2006 
In the spring-piston model, the spring represents the solid phase of the soil and the water 

below the piston is the pore water under saturated condition in the soil mass. Before a new 

load, W, is applied to the piston, the system is assumed to be in equilibrium, i.e., the drainage 

valve is open and there is no excess pore water pressure, ∆u = 0. The spring alone is carrying 

any previously applied loads, such as the weight of the piston itself. The drainage valve is 

closed just before the new load is applied. If the valve is completely shut-off and the piston 

is leak-proof, then, there is no chance for water to escape. Such a condition represents a 

clay-water system in which the clay is very impermeable so that there is significant resistance 

to drainage of water in any direction. When the new load, W, is placed on the piston (this is 

called the initial or “time = 0” condition), the total applied pressure immediately below the 

piston, pt, which equals the load, W, divided by the area of the piston, is immediately 

transferred to the water. Since the drainage valve is closed and water is virtually 

incompressible, the water pressure increases to a value equal to the total applied pressure, 

i.e., the excess water pressure ∆u = pt. 

At “time = 0,” the spring does not carry any of the applied load W. The excess water 

pressure is analogous to the pore water pressure that would be developed in a clay-water 

system under externally applied loads, e.g., loads due to construction of an embankment on 

soft saturated clay. If the valve is now opened, the water will drain to relieve the excess 

pressure in it. With the escape of the water, a part of the pressure carried by the water is 

transferred to the spring where it induces a stress increase analogous to an effective increase 

in the inter-particle stresses, po, in a soil mass. The transfer of pressure from the water to the 

spring occurs over a period of time as shown on the bottom part of Figure 2-13, however, at 

any time during the process, the increased stress in the spring, po, plus the excess pressure in 

the water, ∆u, must equal the applied pressure, pt. This transfer of pressure from the water to 

the spring goes on until the flow stops. At that time all of the applied pressure, pt, will be 

carried by the spring, po, and none by the water, i.e., ∆u = 0, and the system will have come 

into equilibrium under the applied load. The time required to attain equilibrium depends on 

the avenue provided to the water to escape, i.e., the longest drainage path the water has to 

take to leave the system. In Figure 2-13 the longest drainage path is the length of the 

cylinder. Obviously, the system would drain quicker if there were another standpipe-type 

drain at the bottom of the cylinder. 

Regardless of the number of avenues provided for drainage, the rate of excess water pressure 

drop generally decreases with time as shown in the lower half of Figure 2-13. After the 

spring water system attains an equilibrium condition under the imposed load, the 

compression of the piston is analogous to the settlement of the clay-water system under an 

externally applied load. This process is called consolidation. 

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Soils and Foundations – Volume I 2 - 33 December 2006 
2.6.2 Comparison of Drainage Rates between Coarse-Grained and Fine-Grained Soils 

Figure 2-14 shows a comparison of excess pore water pressure dissipation in coarse-grained 

and fine-grained soils. The relatively large pore spaces in coarse-grained soils permit the 

water to drain quicker in comparison to fine-grained soils. This leads to a quick transfer of 

applied loads to the soil solids with an associated decrease in void space. This quick load 

transfer results in a displacement that is commonly termed “rapid” in contrast to the “longterm” displacement that is associated with the consolidation process in fine-grained soils. 

∆u 

Fine-grained soils 

Coarse-grained 

soils 

Maximum excess pore water pressure 

Time, t 

Figure 2-14. Comparison of excess pore water pressure dissipation in coarse-grained 

and fine-grained soils. 

2.7 LATERAL STRESSES IN FOUNDATION SOILS 

In most cases, the vertical stress at any depth in a soil mass due to its self weight is the 

summation of the simple products of the unit weight of each soil layer and its corresponding 

thickness down to the depth of interest. This vertical stress was denoted by pt and the 

effective component of this pressure was denoted by po. Due a variety of factors, including 

depositional patterns, the lateral stress, ph, in a soil mass is usually not the same as the 

vertical stress, po. Since the vertical stress is known with reasonable certainty for practical 

purposes, the lateral stress can be assumed to be a certain percentage of the vertical stress and 

can be expressed as follows: 

 

ph = K po 2-14

For an elastic solid, the value of the proportionality constant, K, can be expressed in terms of 

Poisson’s ratio, ν, as follows: 

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Soils and Foundations – Volume I 2 - 34 December 2006 
ν K = 2-15

1 − ν

Poisson’s ratio, ν, is defined as a ratio of lateral to vertical strains. The value of Poisson’s 

ratio is a function of the type of material, e.g., ν is practically zero for cork (hence its 

suitability as a bottle stopper), for concrete ν is between 0.1 and 0.2, and for steel ν is 

between 0.27 and 0.30. A theoretical upper limit of Poisson’s ratio is 0.5 (rubber comes 

close to this limiting value). In the case of soils, ν will have a different value depending 

upon the type of soil and its moisture condition. For example, for free-draining soils a 

reasonable value of ν would be in the range of 0.25 to 0.35, while for very soft saturated 

clays under rapid loading conditions the value of ν would be close to 0.5. Thus, for freedraining soils, the value of K based on elasticity theory will range from 33% to 54% 

corresponding to ν=0.25 and ν=0.35, respectively, while for soft clays the value of K ≈ 1 

since ν ≈ 0.5. 

Even though a soil mass is not an elastic body, the point to be noted here is that at any point 

within the soil mass both vertical and horizontal (or lateral) stresses exist. When external 

forces are imposed on a soil mass, they will result in an increase in vertical stresses as 

discussed in Sections 2.5 and 2.6. Equation 2-14 indicates that an increase in vertical stresses 

will in turn lead to an increase in lateral stresses. While the increase in vertical stresses is 

important in assessing vertical settlements, change in lateral stresses may affect the load 

acting, for example, against piles supporting a bridge abutment, see Figure 2-15. In this 

figure, it can be seen that the increase in vertical stress imposed by the embankment leads to 

an increase in the lateral stress in the ground that causes lateral deformation (“squeeze”) of 

the soft soil. As the soft soil spreads laterally it will have an effect on foundations. 

Therefore, it is important to evaluate the increase in lateral stresses due to vertical loadings. 

Figure 2-15. Schematic of effect of lateral stresses. 

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A two-dimensional (2-D) representation of the lateral stresses transverse to an embankment 

centerline is shown in Figure 2-16. This schematic was developed with a soft layer of soil 

under the embankment. It can be seen that significant lateral stresses are generated in the soil 

below the embankment load. Note that the vertical stresses due to the embankment can cause 

zones of tensile stresses to develop near the toes of the embankment as shown by the hatched 

zones in Figure 2-16. This means that tensile cracks are likely to develop near the toes of the 

embankments for this particular case. This knowledge can help the geotechnical specialist to 

select proper ground improvement measures rationally and to develop and implement an 

instrumentation program. The key point to understand based on the schematics shown in 

Figures 2-15 and 2-16 is that lateral deformations can be three-dimensional and can affect a 

number of facilities such as buried utilities, embankment slopes and bridge foundations. 

Lateral deformations can also affect off-site structures very easily leading to potential legal 

actions. The three-dimensional (3-D) lateral deformations coupled with vertical deformations 

due to vertical stresses can create a complex state of deformation that needs to be carefully 

considered in the design of geotechnical features. 

Similar to the estimation of vertical stresses, the theory of linear elasticity yields equations 

for lateral stress distribution. However, in these equations Poisson’s ratio is assumed to be a 

constant. Hence, the use of chart solutions in these cases is not as simple as for the vertical 

stress case since complicated equations have to be evaluated (Poulos and Davis, 1974). One 

can prepare spreadsheet solutions based on the equations or use commercially available 

computer programs that have already programmed the equations. Program FoSSA (2003) by 

ADAMA Engineering (Version 1.0 was licensed to FHWA) is an example of a program 

capable of computing the vertical and lateral stresses due to surface loading, including 

embankment and multiple footings. Figures 2-10 and 2-16 were generated using the FoSSA 

program. 

2.7.1 Effect of Shear Strength of Soils on Lateral Pressures 

Up to now the stresses in soils have been explained by using unit weights and the theory of 

elasticity. Elastic theory, when suitably modified to reflect observed phenomena in soils, 

provides a tool to obtain a reasonable first approximation to a solution for many problems in 

geotechnical engineering. However, elastic theory does not recognize the role of shear 

strength of soil in the development of lateral pressures. For example, soils have an ability to 

stand vertically or at a certain slope. The reason for this observed ability is that soil has shear 

strength and to some degree can support itself. This shear strength may come from friction 

and/or cohesion between the soil particles. It is intuitive that these components of shear 

strength should also somehow affect the lateral pressures in soils computed by use of the 

theory of elasticity. The shear strength of soils and its representation for analytical purposes 

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Soils and Foundations – Volume I 2 - 36 December 2006 
is discussed in the Section 2.8 followed in Section 2.9 by a demonstration of how the shear 

strength parameters can be used to express lateral pressures. Readers are referred to Lambe 

and Whitman (1979) or Holtz and Kovacs (1981) for detailed discussions. 

p = γth

Zone of Tensile 

Stresses 

h 

0.2p 

0.4p 

0.6p 

Soft 

Layer 

Figure 2-16. Schematic of vertical stress distribution under embankment loading. 

Graphic generated by FoSSA (2003) program. 

(Note: Version 1.0 of FoSSA program is licensed to FHWA. See Appendix E for a brief 

overview of the FoSSA program).
 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 37 December 2006 
2.8 STRENGTH OF SOILS TO RESIST IMPOSED STRESSES 

If the imposed stress in a soil mass is increased until the deformations (movements) become 

unacceptably large, a “failure” is considered to have taken place. In this case, the strength of the 

soil is considered to be insufficient to withstand the applied stress. 

The strength of geologic materials is a variable property that is dependent on many factors, 

including material properties, magnitude and direction of the applied forces and their rate of 

application, drainage conditions of the mass, and the magnitude of confining pressure. Unlike 

steel whose strength is usually discussed in terms of either tension or compression and concrete 

whose strength is generally discussed in terms of compressive strength only, the strength of soil 

is generally discussed in terms of shear strength. Typical geotechnical failures occur when the 

shear stresses induced by applied loads exceed the soil’s shear strength somewhere within the 

soil mass. 

2.8.1 Basic Concept of Shearing Resistance and Shearing Strength 

The basic concept of shearing resistance and shearing strength can be understood by first 

studying the principle of friction between solid bodies. Consider a prismatic block B resting on 

a plane surface XY as shown in Figure 2-17. The block B is subjected to two forces: 

• A normal force, Pn, that acts perpendicular to the plane XY, and 

• A tangential force, Fa, that acts parallel to the plane XY. 

Assume that the normal force, Pn, is constant and that the tangential force, Fa, is gradually 

increased. At small values of Fa, the block B will not move since the applied force, Fa, will be 

balanced by an equal and opposite force, Fr, on the plane of contact XY. The resisting force, Fr, 

is developed as a result of surface roughness on the bottom of the block B and the plane surface 

XY. The angle, θ, formed by the resultant R of the two forces Fr and Pn with the normal to the 

plane XY is known as the angle of obliquity. 

If the applied horizontal force, Fa, is gradually increased, the resisting force, Fr, will likewise 

increase, always being equal in magnitude and opposite in direction to the applied force. When 

the force Fa reaches a value that increases the angle of obliquity to a certain maximum value θm, 

the block B will start sliding along the plane. Recall that during this entire process the normal 

force, Pn, remains constant. The following terminology can now be developed: 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 38 December 2006 
Normal Force, Pn

Tangential Force, Fa 

X 

θm 

Y 

Block B 

θ

Shearing Resistance, Fr 

Fa = Pn tan θm 

Figure 2-17. Basic concept of shearing resistance and strength (after Murthy, 1989). 

•	 If the block B and the plane surface XY are made of the same material, the angle θm is 

equal to φ, which is termed the angle of friction of the material. The value tan φ is 

called the coefficient of friction. 

•	 If the block B and the plane surface XY are made of dissimilar materials, the angle θm is 

equal to δ, which is termed the angle of interface friction between the bottom of the 

block and the plane surface XY. The value tan δ is called the coefficient of interface 

friction. 

•	 The applied horizontal force, Fa, on the block B is a shearing force and the developed 

force is called frictional resistance or shearing resistance. The maximum frictional or 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 39 December 2006 
shearing resistance that the materials are capable of developing on the interface is 

(Fa)max. 

If the same experiment is conducted with a greater normal force, Pn, the maximum frictional or 

shearing resistance (Fa)max, will be correspondingly greater. A series of such experiments would 

show that for the case where the block and surface are made of the same material, the maximum 

frictional or shearing resistance is approximately proportional to the normal load Pn as follows: 

(Fa)max = Pn tan φ 2-16 

If A is the overall contact area of the block B on the plane surface XY, the relationship in 

Equation 2-16 may be written as follows to obtain stresses on surface XY: 

Fa ( )max ⎛ Pn ⎞ = ⎜ ⎟tanφ 2-17

A ⎝ A ⎠

or 

τ = σn tanφ 2-18 

The term σn is called the normal stress and the term τ is called the shear strength. A graphical 

representation of Equation 2-18 is shown in Figure 2-18a. In reality, the relationship is curved, 

but since most geotechnical problems involve a relatively narrow range of pressures, the 

relationship is assumed to be linear as represented by Equation 2-18 over that range. 

The concept of frictional resistance explained above applies to soils that possess only the 

frictional component of shear strength, i.e., generally coarse-grained granular soils. But soils 

that are not purely frictional exhibit an additional strength component due to some kind of 

internal electro-chemical bonding between the particles. This bonding between the particles is 

typically found in fine-grained soils and is termed cohesion, c. Simplistically, the shear 

strength, τ, of such soils is expressed by two additive components as follows and can be 

graphically represented as shown in Figure 2-18(b): 

τ = c + σn tanφ 2-19 

Again, in reality, the relationship is curved. But, as noted above, since most geotechnical 

problems involve a relatively narrow range of pressures, the relationship is assumed to be linear 

as represented by Equation 2-19 over that range. 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 40 December 2006 
 

 

Shear Strength, τ

Friction angle, φ

Normal Stress, σn (a) 

Shear Strength, τ

Friction angle, φ

Cohesion, c 

Normal Stress, σn 

(b) 

Figure 2-18. Graphical representation of shearing strength. 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 41 December 2006 
Equation 2-19 was first proposed by French engineer Coulomb and is used to express shear 

strength of soils. When plotted on arithmetic axes the resulting straight line is conventionally 

known as the Mohr-Coulomb (M-C) failure envelope. “Mohr” is included in “Mohr-Coulomb” 

because Equation 2-19 can also be derived based on concept of Mohr’s circle. The development 

of the Mohr-Coulomb failure envelope based on the application of Mohr’s circle is presented in 

Appendix B. 

As indicated previously, the deformation of soils occurs under effective stresses. In terms of 

effective stresses, Equation 2-19 can be re-written as follows: 

τ′ = c′ + (σn - u) tan φ′ = c′ + σ′ tan φ′	 2-20 

where c′ = effective cohesion, σ′ is the effective normal stress and φ′ is the effective friction 

angle. Further discussion on the cohesion and friction angle is presented in Chapter 4. 

In geotechnical engineering, the normal stresses are commonly expressed using the overburden 

pressure concept introduced in Section 2.3. In terms of overburden pressure, the term σn in 

above equations is the same as pt and the term σ′ is the same as po. Thus, Equations 2-19 and 2­

20 can be expressed in terms of overburden stresses as follows: 

τ = c + pt tanφ 2-21 

τ′ = c′ + (pt - u) tan φ′ = c′ + po tan φ′	 2-22 

Since this manual relates to geotechnical engineering, Equations 2-21 and 2-22 will be used to 

express the M-C failure envelope. The physical meaning of the M-C failure envelope shown in 

Figure 2-18(a) and Figure 2-18(b) may be explained as follows: 

•	 Every point on the M-C failure envelope represents a combination of normal and shear 

stress that results in failure of the soil, i.e., the Mohr failure envelope essentially defines the 

strength of the soil. In other words, any point along the M-C envelope defines the 

limiting state of stress for equilibrium. 

•	 If the state of stress is represented by a point below the M-C failure envelope then the soil 

will be stable for that state of stress. 

•	 States of stress beyond the M-C failure envelope cannot exist since failure would have 

occurred before that point could be reached. 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 42 December 2006 
2.9 STRENGTH OF SOILS RELATED TO LATERAL EARTH PRESSURES 

The concept of shear strength described in the previous section can now be used to 

understand the phenomenon of lateral earth pressure in a soil mass, which is related to 

problems of slope stability and earth retention. From a theoretical viewpoint, problems in 

these three areas (earth pressures, slope stability, and retaining structures) fall into a class of 

problems involving plasticity theory and are best solved by some form of equilibrium 

solution. Many geotechnical engineering text books (e.g., Lambe and Whitman, 1979; Holtz 

and Kovacs, 1981) deal with these solutions extensively. From a practical viewpoint, values 

of earth pressure are needed either directly or indirectly to determine: 

a) If an unrestrained slope is stable and 

b) If not, what kind of retaining structure will be required to stabilize the slope. 

The simplest consideration of earth pressure theory starts with the assessment of the vertical 

geostatic effective stress, po, at some depth in the ground (effective overburden pressure) as 

considered in Section 2.3. The lateral geostatic effective stress, ph, at this depth is given in 

general terms by Equation 2-14 where, for an ideally elastic solid, the value of the lateral 

earth pressure coefficient, K, is given by Equation 2-15. However, the behavior of real soils 

under loads is not always ideally elastic. To simplify the discussion of this topic, consider 

only dry coarse-grained cohesionless soils. The geostatic effective stress condition on a soil 

element at any depth, z, is shown in Figure 2-19a. Since the ground is “at-rest” without any 

external disturbance, this condition is commonly referred to as the “at-rest” condition with 

zero deformation. The coefficient of lateral earth pressure for this condition is labeled Ko. 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 43 December 2006 

δa δp

Wall 

ph=Ko po ph=Ko po ph=Ka po ph=Kp po

po po po z po 

(a) (b) (c) (d) 

Figure 2-19. Stress states on a soil element subjected only to body stresses: (a) In-situ 

geostatic effective vertical and horizontal stresses, (b) Insertion of hypothetical 

infinitely rigid, infinitely thin frictionless wall and removal of soil to left of wall, (c) 

Active condition of wall movement away from retained soil, (d) Passive condition of 

wall movement into retained soil. 
 

 

To relate to the lateral earth pressures acting on retaining structures, assume that a 

hypothetical, infinitely thin, infinitely rigid “wall” is inserted into the soil without changing 

the “at rest” stress condition in the soil. For the sake of discussion, assume that the 

hypothetical wall maintains the “at rest” stress condition in the soil to the right of the wall 

when the soil to the left of the wall is removed. This condition is shown in Figure 2-19b. 

Now suppose that the “at rest” condition is removed by allowing the hypothetical vertical 

wall to move slightly to the left, i.e., away from the soil element as shown in Figure 2-19c. 

In this condition, the vertical stress would remain unchanged. However, since the soil is 

cohesionless and cannot stand vertically on its own, it actively follows the wall. In this 

event, the horizontal stress decreases, which implies that the lateral earth pressure coefficient 

is less than Ko since the vertical stress remains unchanged. When this occurs the soil is said 

to be in the “active” state. The lateral earth pressure coefficient at this condition is called the 

“coefficient of active earth pressure,” Ka, and its value at failure is expressed in terms of 

effective friction angle, φ', as follows: 

1 − sin φ′ Ka = 2-23 1 + sin φ′

Returning to the condition shown in Figure 2-19b, now suppose that the “at rest” condition is 

removed by moving the hypothetical vertical wall to the right, i.e., into the soil element as 

shown in Figure 2-19d. Again, the vertical stress would remain unchanged. However, the 

soil behind the wall passively resists the tendency for it to move, i.e., the horizontal stress 

would increase, which implies that the lateral earth pressure coefficient would become 

greater than Ko since the vertical stress remains unchanged. When this occurs the soil is said 

to be in the “passive” state. The lateral earth pressure coefficient at this condition is called 

the “coefficient of passive earth pressure,” Kp, and its value at failure is expressed in terms 

of effective friction angle, φ', as follows: 

1 + sin φ′ Kp = 2-24 1 − sin φ′

When failure occurs during either of the two processes described above, “Rankine” failure 

zones form within the soil mass. The details of how the failure zones develop are described 

in most geotechnical engineering textbooks and will not be treated here. The so-called 

“Rankine” failure zones and their angles from the horizontal are shown in Figure 2-20. 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 44 December 2006 
Wall Movement 

Figure 2-20. Development of Rankine active and passive failure zones for a smooth 

retaining wall. 

2.9.1 Distribution of Lateral Earth and Water Pressures 

The earth pressure coefficients, Ka and Kp, can be substituted into Equation 2-14 to obtain 

equations for active and passive lateral earth pressures, respectively as follows: 

pa = Ka po 2-25a 

pp = Kp po 2-25b 

It can be seen from Equations 2-25a and 2-25b that the lateral pressures pa and pp are a 

certain fraction of the vertical effective overburden pressure po. Thus, active and passive 

lateral earth pressures are effective pressures and their distribution will be same as that for po. 

The overburden pressure increases in proportion to the unit weight and is typically triangular 

for a given geomaterial. The general distribution of the active and passive pressures along 

with the configuration of active and passive failure surfaces is shown in Figure 2-21a and 2­

21b, respectively. 

In cases where ground water exists, the lateral pressure due to the water at any depth below 

the ground water level is equal to the hydrostatic pressure at that point since the friction angle 

of water is zero and use of either Equation 2-23 or 2-24 leads to a coefficient of lateral 

pressure for water, Kw equal to 1.0. The computation of the vertical water pressure was 

demonstrated previously in Example 2-1. Since Kw=1, the same computation applies for the 

lateral pressure as well. The lateral earth pressure is computed by using the vertical effective 

overburden pressure po at any depth and applying Equations 2-25a and 2-25b. The lateral 

earth pressure is added to the hydrostatic water pressure to obtain the total lateral pressure 

acting on the wall at any point below the ground water level. For a typical soil friction angle 

of 30 degrees, Ka = 1/3. Since Kw = 1, it can be seen that the lateral pressure due to water 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 45 December 2006 
is approximately 3 times that due the active lateral earth pressure. A general case for 

the distribution of combined active lateral earth pressure and lateral water pressure is shown 

in Figure 2-22. As will be discussed in Chapter 10 (Earth Retaining Structures), this 

disparity in lateral pressures has serious consequences when the stability of walls is 

considered and is the reason why drainage behind walls is so important. 

(a) (b) 

Active pressure at depth z: pa = Ka γ z Passive pressure at depth z: pp = Kp γ z 

Active force within depth z: Pa = Ka γ z

2

/2 Passive force within depth z: Pp = Kp γ z

2

/2 

Figure 2-21. Failure surfaces, pressure distribution and forces (a) active case, (b) 

passive case. 

Figure 2-22. General distribution of combined active earth pressure and water 

pressure. 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 46 December 2006 
2.9.2 Deformations Associated with Lateral Pressures 

 

The active and passive pressures are predicated on the development of a certain amount of 

lateral deformation in the soil. The magnitudes of these lateral deformations and their effect 

on the development of earth pressures at failure are discussed in Chapter 10 (Earth Retaining 

Structures). 

2.10 UNSATURATED SOIL MECHANICS 

As discussed in this Chapter, soil is three phase system that consists of solid particles, liquid 

and gas. Classical soil mechanics concentrates primarily on the behavior of saturated or dry 

soils, i.e., a two phase system. For soils in a saturated state, the principle of effective stress is 

invoked to quantify stress and strain in the soil mass. For soils in a dry state, pore water 

pressure does not exist and the total stress and effective stress are the same. In reality, all the 

pore space in soil within the depth of significant influence of geotechnical features is rarely 

occupied by liquid or gas alone. This is particularly true for soils above the ground water 

table and soils that are mechanically compacted as in the case of earthworks. In such soils 

the degree of saturation is generally intermediate between 0% (dry soil) and 100% (saturated 

soil). Under these conditions, negative pore pressures, i.e., suction, may exist within the soil 

mass depending upon the type of soil and its grain size distribution. An example of the 

presence of negative pore pressures is the capillary rise often encountered above the water 

table. Such negative pore pressures affect all aspects of soil behavior ranging from volume 

change and shear strength to seepage. Consequently, unsaturated soil behavior impacts a 

broad array of engineering issues ranging from foundation design and performance to flow 

through earth embankments and the engineering of facilities on or in expansive, collapsible 

and compacted soils (ASCE 1993, 1997). 

To date the tendency in engineering practice has often been to apply a total stress approach 

where the effects of negative pore pressures are not properly simulated. In the last couple of 

decades significant progress has been made to model such negative pore pressures and that 

field of study is often called “unsaturated soil mechanics.” Discussion of the engineering 

behavior of unsaturated soils is beyond the scope of this manual. At this stage, it is important 

simply to realize that advanced studies beyond those discussed in this manual may be 

required on projects where unsaturated state can significantly affect the engineering behavior 

of soils. The interested readers are directed to the work by Fredlund and Rahardjo (1993), 

who provide a comprehensive treatment of unsaturated soils. 

FHWA NHI-06-088 2 – Stress and Strain in Soils
 

Soils and Foundations – Volume I 2 - 47 December 2006 
[THIS PAGE INTENTIONALLY BLANK] 
 

CHAPTER 3.0 

SUBSURFACE EXPLORATIONS 

To perform properly, a structure must interact favorably with the soil on which it rests. The 

modern geotechnical specialist, who often must build in areas that were considered too poor to 

build upon a few years ago, must be well versed in the fundamentals of soil mechanics. This 

knowledge will be used in the design of structural foundations and earthworks to answer the 

following questions. Will settlements be excessive? Can the structure tolerate settlements? 

Will the proposed foundation type perform better than another type? Can the foundation soils 

safely support the imposed embankment or footing loads? Will the proposed cut or fill slopes 

have adequate stability? Are the foundation and earthwork designs cost-effective? 

The engineer should have adequate knowledge of the subsurface conditions at a site before 

attempting to answer these questions. A site- and project-specific subsurface model must be 

developed for the cost-effective engineering design of a facility. Figure 3-1 shows a flow 

chart that identifies a recommended process for developing a subsurface model for 

engineering design. The investment of a few tens of thousands of dollars in a systematic 

approach as outlined in Figure 3-1 could result in design and construction savings of 

hundreds of thousands of dollars by preventing costly failures or overly conservative designs. 

The process shown in Figure 3-1 is logical and is generally followed on many projects. In 

many cases, however, old “rules-of-thumb” and “status quo” approaches can result in an 

unconscious “by-passing” of critical steps. In particular, selection of the correct tests to 

determine the relevant engineering properties, the interpretation of the results of those tests, 

and summarization of data are often poorly performed. Rigorous attention to the rational 

process in Figure 3-1 is required to assure efficient and thorough exploration and testing 

programs, especially since many projects are fragmented to the extent that drilling, testing, 

and design are performed by different parties. This document provides guidance on all the 

items presented in Figure 3-1. The three major steps in the flow chart in Figure 3-1 and the 

applicable chapters in this document are as follows: 

Step 1: Subsurface Exploration and Field Testing (this Chapter) 

Step 2: Laboratory Testing and Test Interpretation (Chapters 4 and 5) 

Step 3: Engineering Design (Chapters 6 to 10 and Appendix A) 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 1 December 2006 
Site Investigation and Field Testing

Laboratory Testing and

Test Interpretation

Engineering

Design

 

 

 

Phase II Investigation (if needed)

Subsurface Exploration and Field Testing

Laboratory Testing and Test Interpretation 

Review available subsurface information and develop preliminary model of subsurface conditions Review available subsurface information and develop preliminary model of subsurface conditions 

Conduct laboratory testing 

Identify material properties required for design and constructability and estimate scope of field program Identify material properties required for design and constructability and estimate scope of field program 

Plan site exploration and field test program Plan site exploration and field test program 

Conduct field investigations and field testing Conduct field investigations and field testing 

Perform sample descriptions and laboratory index tests Perform sample descriptions and laboratory index tests 

Summarize basic soil/rock data and develop subsurface profile Summarize basic soil/rock data and develop subsurface profile 

Are results 

consistent with 

preliminary 

model? 

Review design objectives and initial results Review design objectives and initial results

Are there additional 

data needs 

Select representative soil/rock samples and details of laboratory testing 

Yes 

No 

Review quality of laboratory test data and summarize 

Select material properties and finalize subsurface model Select material properties and finalize subsurface model 

Are results 

consistent and valid 

Is a Phase II 

Investigation 

necessary? 

Yes 

No 

Perform design and consider constructability issues 

Yes 

No 

Yes 

No 

Engineering

Design 

Phase II Investigations (if needed) 

Figure 3-1. Recommended process for developing subsurface model for engineering 

design (FHWA, 2002a). 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 2 December 2006 
3.01 Primary References 

The primary references for this Chapter as well as Chapters 4 and 5 are as follows: 

FHWA (2002a). Geotechnical Engineering Circular 5 (GEC5) - Evaluation of Soil and 

Rock Properties. Report No FHWA-IF-02-034. Authors: Sabatini, P.J, Bachus, R.C, Mayne, 

P.W., Schneider, J.A., Zettler, T.E., Federal Highway Administration, U.S. Department of 

Transportation. 

FHWA (2002b). Subsurface Investigations (Geotechnical Site Characterization). Report No. 

FHWA NHI-01-031, Authors: Mayne, P. W., Christopher, B. R., and DeJong, J., Federal 

Highway Administration, U.S. Department of Transportation. 

AASHTO (2006). Standard Specifications for Transportation Materials and Methods of 

Sampling and Testing, Parts I and II, American Association of State Highway and 

Transportation Officials, Washington, D.C. 

ASTM (2006). Annual Book of ASTM Standards – Sections 4.02, 4.08, 4.09 and 4.13. 

ASTM International, West Conshohocken, PA. 

3.1 PREPARING FOR SUBSURFACE EXPLORATION 

The initial step in any highway project must include consideration of the soil or rock on which 

the highway embankment and structures are to be supported. The extent of the site exploration 

will depend on many factors, not the least of which will be the project scheduling, general 

subsurface conditions, and the nature of the loads to be supported. In any event, certain basic 

steps should be followed before exploration equipment is mobilized to the project site. The 

first step in the exploration is to collect and analyze all existing data. A review of available 

information prior to the field reconnaissance will help establish what to look for at the site. In 

the Eighth Rankine Lecture, Glossop (1968) stated the following truism regarding site 

exploration: "If you do not know what you should be looking for in a site investigation, 

you are not likely to find much of value." For a highway project, basic sources of 

geotechnical information should be reviewed to determine landform boundaries and to provide 

a basis for outlining the project subsurface exploration program. Those sources and functional 

uses are identified in Table 3-1. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 3 December 2006 
Table 3-1 

Sources of historical site data (after FHWA, 2002a) 

Source Functional Use Location Examples 

Utility Maps 

•••

Identifies buried utility locations Identifies access restrictions Prevents damage to utilities 

Local agencies/utility companies 

Power line identification priexploration prevents extensiexpensive repairs, and bodil

Aerial Photographs 

••••

Identifies manmade structures Identifies potential borrow source areas 

Provides geologic and hydrological information which can be used as a basis for site reconnaissance Track site changes over time 

Local Soil Conservation Office, United States Geological Survey (USGS), local library, local & national aerial survey companies 

Evaluating a series of aerial an area on site which was fireviewed 

Topographic Maps 

••••

Provides good index map of site area Allows for estimation of site 

topography Identifies physical features in the site area Can be used to assess access restrictions 

USGS, State Geological Survey 

Engineer identifies access arareas of potential slope instacut/fill capacity before visiti

Existing Subsurface Exploration Report 

• May provide information on nearby 

soil/rock type; strength parameters; hydrogeological issues; foundation types previously used; environmental concerns 

USGS, United States Environmental Protection Agency (USEPA), State/local agencies, developers, etc. 

A five year old report for a nproject provides geologic, hgeotechnical information foscope of the exploration 

Geologic Reports and Maps 

• Provides information on nearby 

soil/rock type and characteristics; 

hydrogeological issues, environmental concerns 

USGS and State Geological Survey 

A twenty year old report on identifies earth fissure rock and orientation data) and gr

Water/Brine Well Logs 

•••

Provide stratigraphy of the site and/or regional area Varied quality from state to state Groundwater levels 

State Geological Survey/Natural Resources, Department of water resources 

A boring log of a water supp

the site area shows site stratevaluations of required dept

FHWA NHI-06-088 3 Soils and Foundations 3 - 4 
Source Functional Use Location Examples 

Flood Insurance 

Maps 

•

••

Identifies 100 and 500 yr floodplains near water bodies Caution against construction in a floodplain Provide information for evaluation 

Federal Emergency Management Agency (FEMA), USGS, state/local 

agencies 

Prior to exploration, the floois in a 100 yr floodplain and 

moved to a new location 

of scour potential 

Soil Survey 

•

••

Identifies site soil types Permeability of site soils Climatic and geologic information 

Local Soil Conservation 

Service 

The local soil survey providsurface soils to facilitate preevaluation 

Sanborn Fire Insurance Maps 

•

••

••

Useful in urban areas Maps for many cities are continuous for over 100 years. Identifies building locations and type 

Identifies business type at a location (e.g., chemical plant) 

May highlight potential environmental problems at an urban site 

State library/Sanborn Company (www.sanborncompany.com) 

A 1929 Sanborn map of St. 

smelter was on site for 10 yprevents an exploration in a 

Table 3-1 (Continued) 

Sources of historical site data (after FHWA, 2002a) 

FHWA NHI-06-088 3 Soils and Foundations 3 - 5 
A necessary part involved in review of existing data is to identify the major geologic processes 

that occurred at the project site because this will permit the geotechnical specialist to develop 

an understanding of how the local soil and rock formations may have developed. The soil 

formation process and consideration of landforms in designs of geotechnical features is 

discussed briefly followed by discussions of subsurface exploration programs. 

3.1.1 Soil Formations and Landforms 

Soils are a result of the weathering of rocks. In general, rocks are classified as igneous, 

sedimentary, and metamorphic. Igneous rocks are products of melts (magma) generated an 

unknown distance below the earth’s surface. Sedimentary rocks are cemented and/or 

compressed materials derived from pre-existing sediments deposited in layers by water or by 

air. A metamorphic rock is any rock that originates by a process of change from what it was 

previously. Any former igneous, sedimentary, or metamorphic rock can be metamorphosed 

(changed) into a new metamorphic rock by an increase in temperature and/or pressure and/or 

by reaction with surrounding hot fluids and gases. Regardless of the type of rock, most 

weathering takes place near the ground surface. Rock weathering can occur due to 

mechanical (physical) and/or chemical processes as follows: 

•	 Mechanical or physical process refers to the process whereby the intact rock breaks into 

smaller fragments. Physical weathering may be caused by expansion resulting from 

unloading (e.g., exfoliation or spalling off of the exterior surface of the exposed rock), 

abrasion, temperature changes (e.g., freeze/thaw), erosion by wind or rain, crystal 

growth (e.g., ice and other crystals such as salt crystals that form as the result of the 

capillary action of water containing salts in solution), and organic activity (e.g., forces 

exerted by growing plants and roots in voids and crevasses of rock). 

•	 Chemical process refers to the process whereby the minerals in the rock are altered into 

new compounds. Chemical weathering is usually preceded by hydration and hydrolysis 

and may be caused by, oxidation (e.g., chemical reaction with rainwater), solution (e.g., 

dissolution of limestone) and/or leaching (e.g., dissolution of the cementing agent in the 

rock). Chemical weathering commonly occurs by fluids seeping into the fractures 

caused by mechanical (physical) weathering processes. These fluids are chiefly acids 

created as rainwater dissolves carbon dioxide from the atmosphere and more carbon 

dioxide and organic acids from the soil. Most chemical weathering processes result in 

an increase in volume (that causes an increase in stress within the rock mass), lower 

density materials (e.g., soils), smaller particle sizes (e.g., clay sizes), and more stable 

minerals (that may decrease the rate of chemical weathering). 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 6 December 2006 
The combined effects of the mechanical and chemical weathering processes vary considerably 

with climate and the mineralogy of the parent rock. The chemical reactions proceed most 

rapidly and completely in humid tropics and subtropics and least effectively in cold or arid 

climates (Goodman, 1993). Thus, in the Arctic regions and deserts, the mechanical processes 

of physical weathering act virtually alone to gradually breakup the rock into a fractured or 

rubbled mass whereas, in the tropics, the two weathering processes work together rapidly first 

to break up the rock and then to alter newly exposed rock surfaces during a project’s life. 

Once the intact rock is broken into fragments, the rate of weathering depends on the particle 

size and the climate. In general, small particles weather at a faster rate than large ones due to 

their larger surface area. The weathering processes can result in particle sizes that are not 

distinguishable by the naked eye (e.g., colloidal particle size) and can be identified only by 

equipment such as electron microscopes. Based on particle size, the principal terms used by 

civil engineers to describe soils are gravel, sand, silt and clay. These terms were discussed in 

Chapter 2 as a function of the particle sizes they represent and some of their physical 

characteristics. For example, silt and clay particles are finer than the No. 200 sieve (0.075 mm) 

and exhibit varying properties in presence of water. 

Soils created by a particular geologic process assume characteristic topographic features, called 

landforms, which can be readily identified by the geotechnical specialist. A landform contains 

soils with generally similar engineering properties and typically extends irregularly over wide 

areas of a project alignment. Early identification of landforms can be used to optimize the 

subsurface exploration program. Landforms may be described according to the method of 

formation as residual soil landforms or transported soil landforms. Soils commonly 

associated with these two types of landforms are briefly described as follows: 

3.1.1.1 Residual Soils 

A residual soil landform is one that was formed in its present location through weathering of 

the parent (or bed) rock. Residual soils tend to be characterized by angular to subangular 

particles, mineralogy similar to parent rock, and the presence of large angular fragments within 

the overall soil mass. Because residual soil weathers from parent bedrock, its profile with 

depth represents a history of the weathering process. Figure 3-2 shows a typical weathering 

profile for metamorphic and igneous rocks. In Figure 3-2, the weathering profile is divided into 

three zones: residual soil, weathered rock, and unweathered rock. Deere and Patton (1971) 

present 12 other weathering-profile classification systems proposed by workers from around 

the world. Regardless of the weathering-profile classification, the following are some of the 

properties for such profiles: 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 7 December 2006 
(a) Metamorphic Rocks (b) Intrusive Igneous Rock

ZONE 

Colluvium of other transported soils; “pebble marker” common here IA 

A horizon 

IB 

B horizon 

IC 

C horizon 

(Saprolite) 

IIA 

Transition 

from 

Saprolite to 

Weathered 

Rock 

IIB 

Partly 

Weathered 

Rock 

III 

Unweathered 

Rock 

II 

Weathered 

Rock 

I 

Residual 

Soil 

Figure 3-2. Typical weathering profile for metamorphic and igneous rocks (Deere and Patton

FHWA NHI-06-088 3 Soils and Foundations – Volume I 3 - 8 
•	 The permeability and shear strength gradually change with depth. These two 

parameters control both the amount of rainfall infiltration and the location of the shear 

surface when external loads are applied on or in these soils. 

•	 Soil profile thickness and properties depend upon parent bedrock, discontinuities, 

topography, and climate. Because these factors vary horizontally, the profile can vary 

significantly over relatively short horizontal distances. 

•	 Deep profiles form in tropical regions where weathering agents are especially strong 

and advanced stages of chemical weathering form cemented soils called laterites. The 

technical literature often refers to residual soils as tropical soils. 

•	 The material in the transitory zone between residual soil and unweathered rock is called 

saprolite. Saprolites are generally unsaturated, weakly bonded and heterogeneous soils 

with relict joint systems (Lambe, 1996). Saprolites have widely varying void ratios and 

widely varying mineralogy and shear strength (Vaughn, et al., 1988). 

3.1.1.2 Transported Soils 

A transported soil is one that was formed from rock weathering at one location and transported 

by some exterior agent to another location. The transporting agent may be water (principal 

agent), a glacier, wind, and/or gravity. Often the deposits of transported soils are given names 

indicative of the mode of transportation causing the deposit, e.g., alluvial deposits, glacial till, 

etc. Transported soils are characterized by subrounded to rounded particles and a wide variety 

of particle sizes. Table 3-2 summarizes commonly encountered landforms composed of 

transported soils, their primary formational process, and their engineering significance. 

3.1.1.3 Area Concept for Explorations Based on Landforms 

Knowledge of the landforms and the engineering properties of the soils and rocks enables the 

designer to determine the most economical location for a highway alignment and grade, to 

evaluate design problems for each type of soil or rock, and to determine sources of granular 

borrow material. Once a landform is identified, the geotechnical specialist can develop an 

“area” concept for exploration. In this concept, the lateral extents of landforms in the 

immediate vicinity of the footprint of the proposed facility are first identified. Then, a limited 

exploration program (e.g., geologic mapping, geophysical work and some preliminary cone 

penetration tests and borings) is implemented at strategic locations such that the general 

characteristics of the landforms are identified. The exploration program can then be refined to 

obtain specific information on soil types of interest with respect to the footprint of the proposed 

facility and the anticipated loadings. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 9 December 2006 
Table 3-2 

Common landforms of transported soils and their engineering significance 

Agent Landform Formational Process and General Engineering Significance for Study 

Water Flood Plain • Formed in valleys that are nearly flat and near the high water level of streams. At flood stage 

essentially a “flood plain” that is susceptible to widespread shallow flooding. 

• Generally poor construction site with fine-grained soils and water problems. Potential scour a

design below ground will probably require undercut, low foundation pressure and scour prote

foundations probable. Additional shallow explorations required along footing length to determ

meandering channels. Historic high water levels should be used in design. 

Coastal Plain 

•

•

Formed similar to flood plain but in coastal areas. 

Consider spread footings for moderate loads except for high water areas. Potential scour ar

possible for friction piles (see Chapter 9). 

Terraces •

•

Formed when a stream or water body cuts into a previously deposited sediment or as the strea

geological periods due to normal erosion or to crustal deformations. Terraces are also known 

Consider spread footings for lightly loaded foundations. 

Lakebed (Lacustrine, Varves) 

••

Formed by sedimentation in lake (fresh water) environments. Varves are a particular type of lduring glacial periods from seasonal ice melting, which temporarily increased the runoff velocprecipitated sand layers alternate with layers of precipitates such as silt or silt-clay made at low Suitable only for spread footings to support light loads and even then settlement may be expecprobable and often deep. Obtain undisturbed tube samples for laboratory testing. Consider drrather than casing. Long-term water observations necessary to determine static water level du

Potential scour area. 

Deltas •

•

Formed by sediments precipitated at the mouths of rivers or streams into bays, oceans, or lake

The use of spread footings must be carefully studied as poor soils often underlie deltaic sands 

parent material is capable of sustaining high spread footing loads. Piles may be required to peand poor soil. Use casing of adequate size to obtain undisturbed samples of poor soil. Potenti

Alluvial Fans. Filled Valleys (Basin Deposits) 

••

Formed similar to delta deposits, but typically found in arid areas where mountain stream runovalleys or on to the plains at the mouths of streams. In arid climates, alluvial fans can become 

in the ground by evaporating water or by dropping groundwater. Cemented soils can be loose 

caliche) or open-graded (collapsible). Consider spread footings for low to moderate loads except at lower elevation of alluvial fans w

is possible. In case of collapsible soils, either treat the soils such that collapse potential is miti

foundations to bypass such soils. If the caliche is firm to hard, spread foundations can be used

FHWA NHI-06-088 3 Soils and Foundations – Volume I 3 - 10 
Table 3-2 (continued) 

Common landforms of transported soils and their engineering significance 

Agent Landform Formation and General Engineering Significance for Study 

Ice (Glacier) and meltwater associated 

Moraines (Terminal, lateral) 

• Formed by soil deposits pushed into ridges around the periphery of a glacier. Terminal moraine

material scraped or bulldozed to the front of a glacier; lateral moraines develop along the sides 

moraine may not be a single nicely rolled ridge, but rather a highly serrated, above ground level

• Advisable to use spread footings for all foundation loads. Piles should not be used due to very d

boulders. Core all rock to 10 ft (3.0 m) in case boulders are encountered. 

with ice Glacial • Glacial Till (also termed ground moraine or simply till) is the deposit of ice-suspended material 

Till the glacier. As glaciers melted, materials suspended in the ice precipitated onto the underlying s(ground glacial till. Till deposits are characterized by all sizes of particles with no obvious arrangement. 

moraine) US has glacial till. 

• Where till is unsorted, dense and contains considerable sand and gravel, it is advisable to use spr

foundation loads. Piles should not be used due to difficult driving conditions and boulders. Cor

to depth of 10 ft (3.0 m) as large boulders may be encountered. Long-term water observations nstatic water level due to soil density. 

Outwash • Sediments precipitated from glacial melts in the discontinuities between ridges in moraines. Sm

temporarily form in depressions behind ridges, producing lacustrine (fresh water) sediments. 

• Spread footing normally used to support moderate to heavy foundation loads. Piles, if required, 

large diameter sample spoon to permit representative sample to be obtained as average particle s

(35.3 mm) sample spoon. Standard penetration test may be erratically high due to large particle value as sand and gravel sources since the material often contains very little amount of fines, i.ethan No. 200 (0.075 mm) sieve. 

Eskers • Eskers are deposits (usually as ridges) formed by precipitation of water-suspended material flow

• Advisable to use spread footings for all loads as soil contains much gravel and is dense. Piles no

Large diameter sample spoon recommended as above for outwash. Commercial value as sand a

since the material often contains very little amount of fines, i.e., particle size less than No. 200 (

Drumlins • Drumlins are isolated mounds of glacial debris varying from about 35 (10 m) ft to 230 ft (70 m) 

m) to 2600 ft (800 m) long. Most drumlins are of the order of 100 ft (30 m) or less in height and 

less in length. They often occur in groups called drumlin fields (several). 

• Suitable for spread footing design with moderate to heavy loads. Piles seldom used due to dens

subsoil. Commercial value as sand and gravel sources since the material often contains very litti.e., particle size less than No. 200 (0.075 mm) sieve. 

FHWA NHI-06-088 3 Soils and Foundations – Volume I 3 - 11 
Table 3-2 (continued) 

Common landforms of transported soils and their engineering significance 

Agent Landform Formation and General Engineering Significance for Study 

Wind (Aeolian) 

Loess • Formed by wind blowing silt and clay with the deposit held together by a montmorillonite binder

from glacial outwash in the US. Low density (often less than 90 pcf (14 kN/m3)), low wet streng

upon water ingress), has the ability to stand on vertical cuts due to cementing agents between par

• Consider spread footings for low to moderate loads. Heavy loads should be pile supported with t

obtained below the loess deposit. Accurate ground water level determination important. 

Sand Dune • Formed by wind action blowing the sand. Transport occurs mainly along the ground until an obs

whereupon a dune (or mound) forms. Later winds may demolish the dune and redeposit the matefurther downwind. Dune sands tend to be well rounded from abrasion. 

• Consider spread footings for small foundations not subject to vibratory loading. Heavy structura

supported on friction piles. 

Gravity Colluvium • Formed by physical and chemical weathering of bedrock. The fragmented particles, given suffic

relief, tend to move down slopes under gravitational forces and accumulate as distinctive depositportions of slopes, in topographic depressions, and especially at the base of cliffs. 

• The characteristics of colluvial materials vary according to the characteristics of the bedrock sour

under which the weathering and transport occur. From an engineering viewpoint, colluvium is w

consists of a heterogeneous mixture of soil and rock fragments ranging in size from clay particles 

ft (1 m) in diameter. Because they are found along the lowest portions of valley sides, such depobe partially excavated to allow passage of transportation facilities. The resulting cut slopes are corequire constant monitoring and maintenance. Colluvial soils are prone to creep (slow movemenlandslides are common in such soils. 

Talus • Talus is colluvium composed of predominantly large fragments. Talus fragments can be huge bo(Scree) across; however, a lower size limit has not been well defined. With time, the coarse fragments m

materials may be added by wind or water transport so that these deposits slowly become infilled grained materials. The degree of infilling of these talus deposits may vary horizontally and vertic

• Rock-supported talus is often inherently unstable and may be hazardous to even walk across. Fu

structure is porous. Talus deposits are not suitable for engineering structures. Talus deposits couriprap. 

FHWA NHI-06-088 3 Soils and Foundations – Volume I 3 - 12 
The area concept is a powerful tool, particularly for linear highway facilities, as it streamlines 

the subsurface exploration program costs and provides the planning engineer with useful data 

during the design and construction phases of a project. It also permits early identification of the 

type and extent of problem soils to be encountered during construction and therefore allows 

construction costs to be estimated more accurately. 

3.2 FIELD RECONNAISSANCE 

Application of the area concept requires the use of proper subsurface exploration equipment 

and techniques. In particular the use of wide area exploration techniques such as remote 

sensing and geophysical techniques can provide insight of general subsurface conditions in 

the project area economically. An adequate site exploration can be accomplished only under 

the direction of a geotechnical specialist who knows the general limitations of the 

exploration equipment as well as the general demands of the project. A field reconnaissance, 

preferably with the bridge designer, roadway designer, and project manager, is recommended 

to assess subsurface conditions prior to establishing a subsurface exploration. 

The field reconnaissance should include: 

1. Inspection of nearby structures to determine their performance with the particular 

foundation type utilized. If settlement is suspected, the original structural plans should 

be reviewed and the structure surveyed by using the original benchmark. 

2. For water crossings, inspection of structural footings and the stream banks up and down 

stream for evidence of scour. Take careful note of the stream bed material. Often large 

boulders exposed in the stream but not encountered in the borings, are an indication of 

potential subsurface obstructions to pile installation. 

3.	 Recording the location, type, and depth of any existing structures or abandoned 

foundations that may infringe on the new highway facility. 

4. Relating site conditions to the proposed boring operations. Record potential problems 

with utilities (overhead and underground), site access, private property, or obstructions. 

Figure 3-3 is an example of a field reconnaissance form used to record data pertinent to the site. 

Upon completion of the site inspection, the geotechnical specialist should prepare a terrain 

reconnaissance report in which the general suitability of the site is assessed. The report should: 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 13 December 2006 
/Cone Rig

/Cone Rig 

Figure 3-3. Typical field reconnaissance form. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 14 December 2006 
1. Flag major potential problems, which may preclude construction. 

2. Recommend beneficial shifts in location. 

3. Present a general discussion of expected subsurface conditions. 

4. Present cost estimate for extraordinary geotechnical treatments. 

5. Prepare an estimate of subsurface exploration quantities, costs, and time. 

 

This information should be transmitted to all the groups involved with the project such as the 

project manager, roadway designer, and bridge designer. 

3.3 SUBSURFACE EXPLORATION PROGRAM 

The procedures employed in any subsurface exploration program are dependent on a variety of 

factors that vary from site to site. The project design objectives and the expected subsurface 

conditions have the major influence on the subsurface explorations. Highway projects 

necessarily involve both earthwork and structural foundations. Typical boring programs for 

highways on new alignments are established such that basic information is first gathered along 

the entire highway alignment and subsequent detailed borings are taken as required at the 

locations of structures or in problem earthwork areas as disclosed by the initial basic program. 

Subsurface explorations for widening or improvements of existing highways generally are done 

in one stage as the location is predetermined. 

Consideration should be given, particularly for large or complex projects, to performing 

geologic mapping and geophysical explorations after the field reconnaissance and prior to 

any invasive explorations such as borings. Geologic mapping and geophysical explorations 

can be quick and provide a much larger coverage of the project area as compared to invasive 

explorations. The information from field reconnaissance, geologic mapping and geophysical 

explorations can then be used to setup the conventional subsurface exploration and testing 

program. Geophysical explorations are discussed in Section 3.15. 

After the field reconnaissance and geophysical explorations are completed, “invasive” 

explorations using drilled borings and in-situ tests must be performed to obtain in-situ 

properties and physical samples for identification and testing. The sampling techniques and 

tools are discussed in Section 3.4 and other sections of this chapter. The objectives for such 

explorations are as follows: 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 15 December 2006 
1. Determine stratigraphy. 

a.	 physical description and extent of each stratum. 

b.	 thickness and elevation of top and bottom of each stratum. 

2. For fine-grained soils (each stratum) determine: 

a.	 natural moisture contents. 

b. Atterberg limits. 

c. stiffness. 

d.	 presence of organic materials. 

e.	 evidence of desiccation or previous soil disturbance, shearing, or slickensides. 

f.	 swelling characteristics. 

g.	 unconfined compressive strength - typically estimated from Standard 

Penetration Tests or Cone Penetration Tests. 

h. shear strength. 

i.	 compressibility. 

3. For coarse-grained soils (each stratum) determine: 

a.	 in-situ density (average and range) typically determined from Standard 

Penetration Tests or Cone Penetration Tests. 

b.	 grain-size distributions (gradation). 

c.	 presence of organic materials. 

4. Determine depth to ground water (for each aquifer if more than one is present). 

a.	 piezometric surface over site area: existing, past, and probable range in future 

(observe at several times). 

b.	 perched water table. 

5. Determine depth to bedrock. 

a.	 depth over entire site. 

b.	 type of rock. 

c.	 extent and character of weathering. 

d.	 joints, including distribution, spacing, whether open or closed, and joint infilling. 

e. faults. 

f.	 solution effects in limestone or other soluble rocks. 

g.	 core recovery and soundness (RQD). 

h.	 hardness and strength. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 16 December 2006 
3.4 SAMPLING TECHNIQUES AND TOOLS 

The purpose of this section is to provide information on various in-situ testing methods that 

are currently used to establish site-specific soil and rock properties for design and 

construction. The execution of a conventional subsurface exploration and testing program 

usually includes rotary drilling, Standard Penetration Testing (SPT), disturbed and 

undisturbed sample recovery, and laboratory testing. Although procedures for these 

commonly performed activities are described in AASHTO and ASTM standards and are well 

known to most geo-professionals, important testing details are sometimes overlooked that 

can result in data having marginal quality. This section discusses the importance of carefully 

selecting and properly conducting the appropriate field and/or laboratory testing method. 

In-situ testing methods are increasingly being used on transportation projects, however 

testing procedures and test limitations are not as well understood as those of the more 

conventional methods of subsurface exploration and testing such as the use of drilled 

borings. In this chapter, procedures for various in-situ and laboratory testing methods are 

presented as they relate to obtaining high quality data for the evaluation of engineering 

properties of soils and rocks for transportation projects. Information on equipment 

calibration, measured test parameters, quality control, and the range of ground conditions 

that apply to each test is also presented. 

Several in-situ tests define the geostratigraphy and provide direct measurements of soil 

properties and geotechnical parameters. The common in-situ tests include: Standard 

Penetration Test (SPT), Cone Penetration Test (CPT), Piezocone Test (CPTu), Flat Plate 

Dilatometer Test (DMT), Pressuremeter Test (PMT), and Vane Shear Test (VST). Although 

the load is applied differently in each test, the purpose of each test is to measure the 

corresponding response of the soil in an attempt to evaluate the soil’s engineering properties, 

such as strength and/or stiffness. Figure 3-4 depicts these various devices and a graphical 

representation of how load is applied. 

Some state DOTs perform these tests by using agency-owned equipment. In many cases 

however, the agency may retain an outside contractor for these services either directly or as 

part of an overall project development package. Several technical reports and manuals are 

available that describe these test methods. A brief list of these references is provided in 

Table 3-3. Agencies that perform or contract for these testing services are encouraged to 

obtain the references identified in Table 3-3. In this manual, only the SPT and the CPT tests 

will be discussed since they are the most commonly used. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 17 December 2006 
SPT CPT DMT PMT VST 

Figure 3-4. Common in-situ tests for geotechnical site characterization of soils (FHWA, 

2002b). 

Boreholes are required for conducting the SPT and normal versions of the PMT and VST. 

Therefore a drill rig and crew are required for the performance of these tests. Boreholes are 

not required for the CPT, CPTu, and DMT; therefore these tests are called “direct-push” 

technologies. Although boreholes are not required for these tests, special mobilized 

equipment and data acquisition systems are required. Specialized versions of the PMT (i.e., 

full-displacement type) and VST can be conducted without boreholes. In such cases either 

standard drill rigs or mobile hydraulic systems (cone trucks) are used to push the probes to 

the required test depths. Obviously direct push test methods are not suitable in soil profiles 

that contain boulders, hard cemented layers and bedrock. For such profiles, borehole 

methods prevail as the testing device may be advanced through the hard layers by coring or 

non coring techniques. An advantage of direct-push soundings is that cuttings or spoil are 

not generated, however, this advantage is offset by a significant disadvantage, i.e., no soil 

sample is retrieved for classification or subsequent laboratory testing. Another advantage of 

the CPT and CPTu tests is that they provide a continuous record of soil response through the 

entire depth of the direct push. The other tests are performed at discrete intervals so that the 

soil’s response is measure at specific depths only. In addition, important layers can be 

missed with any of the discrete interval test methods. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 18 December 2006 
Table 3-3 

Reference publications on in-situ testing (FHWA, 2002b)
 

Test 

Method 

AASHTO/ 

ASTM 

Designation 

Reference 

SPT AASHTO T206 

ASTM D 1586 

FHWA (2002b). Subsurface Investigations (Geotechnical Site 

Characterization). Report No. FHWA NHI-01-031, Authors: 

Mayne, P. W., Christopher, B. R., and DeJong, J., Federal 

Highway Administration, U.S. Department of Transportation. 

CPT, 

CPTu, 

SCPTu 

ASTM D 3341, 

D5778 

FHWA (1992a). The Cone Penetrometer Test. Report No. 

FHWA NHI-91-043, Authors: Riaund J-L and Miran J., Federal 

Highway Administration, U.S. Department of Transportation. 

Lunne, T., Robertson, P.K., and Powell, J.J.M. (1997) Cone 

Penetration Testing in Geotechnical Practice, E & F Spon. 

DMT Suggested 

ASTM Method 

FHWA (1992b). The Flat Dilatometer Test. Report No. FHWA 

NHI-91-044, Authors: Riaund J-L and Miran J., Federal 

Highway Administration, U.S. Department of Transportation. 

PMT ASTM D 4719 FHWA (1989a). The Pressuremeter Test for Highway 

Applications. Report No. FHWA IP-89-008, Authors: Briaud JL, Federal Highway Administration, U.S. Department of 

Transportation. 

Clarke, B.G. (1995) Pressuremeters in Geotechnical Design, 

Blackie Academic & Professional. 

VST ASTM D 2573 ASTM (1988). Vane Shear Strength Testing in Soils: Field and 

Laboratory Studies, American Society for Testing and Materials, 

Committee D-18 on Soil and Rock for Engineering Purposes, 

Philadelphia, PA. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 19 December 2006 
3.5 BORING METHODS 

Geotechnical borings are a critical component of any subsurface exploration program. They 

are performed to satisfy several objectives including those listed below. 

•	 identify the subsurface distribution of materials with distinctive properties, including 

the presence and thickness of distinct layers; 

•	 retrieve samples of each layer for laboratory tests to determine engineering 

properties; 

•	 determine depth to groundwater; and 

•	 provide access for the introduction of in-situ testing devices. 

There are many types of equipment used in current practice for advancing a soil or rock 

boring. Typical types of soil borings are listed in Table 3-4(a), rock coring methods in Table 

3-4(b), and other exploratory techniques in Table 3-4(c). Detailed information on soil and 

rock boring procedures can be found in AASHTO (1988), FHWA (2002b), and ASTM D 

4700. A brief description of typical soil boring methods is provided below (Day, 1999). 

3.5.1 Auger Borings 

An auger is an apparatus with a helical shaft that can be manually or mechanically advanced 

to bore a hole into soil. Large and small diameter augers are shown in Figure 3-5. The 

practice of advancing a borehole with a mechanical auger consists of rotating the auger while 

applying a downward pressure on the auger to penetrate soil and possibly weak or weathered 

rock. The auger may be continuous, where the helix extends along the entire length of the 

shaft, or discontinuous when the auger helix is at the bottom of the drill stem. 

•	 Discontinuous or single flight auger borings and bucket auger borings. There are 

basically two types of discontinuous augers: discontinuous flight augers and bucket 

augers. Commonly available discontinuous flight augers have diameters ranging 

from 0.25 to 3 ft (0.075 to 1 m) and bucket augers have diameters ranging from 1 to 8 

ft (0.3 to 2.5 m). For discontinuous flight auger borings, the auger is periodically 

removed from the hole and the soil lodged in the grooves of the flight auger is 

removed. When a bucket auger is used, it too is periodically removed from the hole 

and the soil in the bucket removed. A casing is generally not used for discontinuous 

flight and bucket auger borings. Therefore, these methods are not recommended for 

boreholes deeper than 35 ft (10 m), or where the hole may cave-in during the 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 20 December 2006 
excavation of loose or soft soils, or when the boring is below the groundwater table. 

In firm stiff clays, discontinuous auger borings can be performed to depths in excess 

of 35 ft (10 m). 

•	 Continuous flight auger borings. As the name implies, continuous flight augers have the 

auger flights continuous along the entire length of the auger. As shown in Figure 3-6a, 

there are two types of continuous flight augers: solid stem and hollow stem. For both of 

these type augers the drill cuttings are returned to the ground surface via the auger 

flights. The solid stem auger must be removed from the borehole to allow access to the 

hole for insertion of sampling or testing devices. Because the auger must be periodically 

removed from the borehole, a solid stem auger is not appropriate in sands and soft soils 

or in soil deposits where groundwater is close to the surface. A hollow-stem auger has a 

circular hollow core that allows for sampling through the center of the auger. As shown 

in Figure 3-6c, hollow-stem augers come in a variety of diameters. The hollow-stem 

auger acts like a casing and allows for sampling in loose or soft soils or when the 

excavation is below the ground water table. A plug (Figure 3-6d) is necessary when 

hollow stem augers are advanced to prevent cuttings from migrating through the hollow 

stem. The plug is removed to permit SPT sampling. In loose sands and soft clays 

extending below the water table, drilling fluids are often used to minimize and mitigate 

disturbance effects and keep the hole open. The components of the hollow stem auger 

system are shown in Figure 3-6b. 

 (a) (b) 

Figure 3-5. (a) Large diameter auger, (b) Small diameter continuous flight auger. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 21 December 2006 
Table 3-4(a) 

Soil and soft rock boring methods (FHWA, 2002a) 

Method Procedure Applications Limitati

Auger boring (ASTM D 1452) 

Dry hole drilled with hand or power auger; samples recovered from auger flights 

In soil and soft rock; to identify geologic units and water content above water table 

Soil and rock strsample mixed wwater table 

Hollow-stem auger boring 

Hole advanced by hollow-stem auger; soil sampled below auger as in auger boring above 

Typically used in soils that would require casing to maintain an open hole for sampling 

Sample limited maintaining hydhole below wate

Wash-type boring 

Light chopping and strong jetting of soil; cuttings removed by circulating fluid and discharged into settling tub 

Soft to stiff cohesive materials and fine to coarse granular soils 

Coarse material bottom of hole; boreholes above undisturbed sam

Becker Hammer Penetration Test (BPT) 

Hole advanced using double acting diesel hammer to drive a 6.6-in (168 mm) double-walled casing into the ground. Several sizes are available. 

Typically used in soils with gravel and cobbles; casing is driven open-ended if sampling of materials is desired 

Skin friction of account for; unsrepeatability of 

Bucket Auger boring 

A 2 to 4 ft (0.6 to 1.2 m) diameter drilling bucket with cutting teeth is rotated and advanced. At the completion of each advancement, the bucket is retrieved from the boring and soil is emptied on the ground. 

Most soils above water table; can dig harder soils than above types and can penetrate soils with cobbles and boulders if equipped with a rock bucket 

Not applicable iused for obtainidisturbed samplnecessary to entobservations 

FHWA NHI-06-088 3 Soils and Foundations – Volume I 3 - 22 
Table 3-4(b) 

Rock core drilling methods (FHWA, 2002a) (1)

Method Procedure Type of sample Applications Lim

Rotary coring Outer tube with diamond (or Rock cylinder 1 in to 4 in To obtain continuous core Corof rock tungsten carbide) bit on (25 to 100 mm) in in sound rock (percent of vari(ASTM lower end rotated to cut diameter and as long as core recovered depends on prevD 2113; annular hole in rock; core 10 ft (3 m), depending on fractures, rock variability, fracAASHTO protected by stationary rock soundness. equipment, and driller skill) bedT 225) inner tube; cuttings flushed 

upward by drill fluid 

Standard coring size is 2­1/8 in (54 mm) diameter. 

but 

Rotary coring Same as ASTM D 2113, but Rock cylinder 1-1/8 in to To recover core better in Corof rock, wire core and stationary inner 3-3/8 in (28 to 85 mm) fractured rock which has variline tube retrieved from outer wide and 5 ft to 10 ft (1.5 less tendency for caving prev

core barrel by lifting device or “overshot” suspended on thin cable (wire line) through special largediameter drill rods and outer core barrel 

to 3 m) long during core removal; to 

obtain much faster cycle of core recovery and resumption of drilling in deep holes 

fracbedbut 

Rotary coring Similar to rotary coring of Soil cylinder 1-1/8 in to In soils and soft rocks that Samof swelling rock; swelling core retained 3-3/8 in (28 to 85 mm) swell or disintegrate rapidly equclay, soft rock by third inner plastic liner wide and 2 ft to 5 ft (0.6 

m to 1.5 m) long encased in plastic tube 

in air (protected by plastic tube) 

than tech

 (1) See Section 3.6.4 for additional discussion on types of core barrels (i.e., single-, double-, or triple-tube). FHWA NHI-06-088 3 Soils and Foundations – Volume I 3 - 23 
Table 3-4(c) 

Other exploratory techniques (FHWA, 2002a) 

Method Procedure Type of sample Applications Lim

Borehole camera 

Inside of core hole viewed by circular photograph or scan 

No sample, but a visual representation of the material 

To examine stratification, fractures, and cavities in hole walls 

Beswhestab

Pits and Trenches 

Pit or trench excavated to expose soils and rocks 

Chunks cut from walls of trench; size not limited 

To determine structure of complex formations; to obtain samples of thin critical seams such as failure surface 

Movequstabwalgroudiffobtashalobtasamsideback

Rotary or cable tool well drill 

Toothed cutter rotated or chisel bit pounded and churned 

Pulverized To penetrate boulders, 

coarse gravel; to identify hardness from drilling rates 

Idenrock

Percussive Method (jack hammer or air track) 

Impact drill used; cuttings removed by compressed air 

Rock dust To locate rock, soft seams, 

or cavities in sound rock 

Drilby w

FHWA NHI-06-088 3 Soils and Foundations – Volume I 3 - 24 
 

 

Solid Stem 

Auger 

Hollow 

Stem Auger 

 (a) (b)

Plug 

Cutting teeth 

 (c) (d) 

Figure 3-6. (a) Solid stem auger and hollow stem auger, (FHWA, 2002b) (b) Hollow 

stem auger components (ASTM D 4700), (c) Sizes of hollow stem auger flights (FHWA, 

2002b), (d) Outer and inner assembly of hollow stem auger (FHWA, 2002b). 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 25 December 2006 
3.5.2 Wash-type Borings 

Wash-type borings use circulating drilling fluid (e.g., water or mud) to remove cuttings from 

the borehole, Figure 3-7. Cuttings are created by the chopping, twisting, and jetting action of 

the drill bit that breaks the soil or rock into small fragments. Tri-cone bits are often used in 

dense soil or soft rock. If bentonite or a polymeric drilling mud cannot be used to maintain 

an open borehole, casings are often used to prevent cave-in of the borehole. The use of 

casing will require a significant amount of additional time and effort but will result in a 

protected borehole. When drilling mud is used during subsurface boring, it will be difficult 

to classify the soil from the auger cuttings because of contamination with the mud. Also, the 

outside of samples may become coated with drilling mud. 

The properties of the drilling fluid and the quantity of water pumped through the drill bit will 

determine the size of particles that can be removed from the boring with the circulating fluid. 

In formations containing gravels, cobbles, or larger particles, coarse material may be left at 

the bottom of the boring. In these instances, cleaning the bottom of the boring with a larger 

diameter sampler (such as the 3 in (75 mm) OD split barrel sampler) may be needed to obtain 

a representative sample of the formation. 

3.5.3 Coring in Rocks 

The previously described methods are typically used for soil exploration. The following 

methods are primarily used for rock exploration. 

•	 Rotary coring. This type of coring equipment is most commonly used for rock 

exploration when an intact core of the rock is desired. Power rotation of the drilling 

bit is accompanied with introduction of a circulating fluid to remove cuttings from the 

hole. The drilling bits are specifically designed to core rock, and inner/outer tubes or 

casings are used to capture the intact core. Table 3-4(b) lists various types of rotary 

coring techniques for rock, although many of these techniques are also applicable to 

dense or stiff soil. 

•	 Percussion drilling. This type of drilling equipment is often used to penetrate hard 

rock for subsurface exploration or for the purpose of drilling wells. The drill bit 

works much like a jackhammer, rising and falling to break up and crush the rock 

material. Air is commonly used to clean the hole and transport the cuttings to the 

ground surface. Table 3-4(c) includes a description on the use of the percussion 

drilling techniques. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 26 December 2006 
 

(a) 

 (b) (c) 

Figure 3-7. (a) Schematic of drilling rig for rotary wash methods (after Hvorslev, 1948), 

(b) Typical drilling configuration, (c) Settling basin (mud tank). 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 27 December 2006 
3.6 SAMPLING METHODS 

3.6.1 Disturbed Sampling of Soil 

Disturbed sampling of soil provides a means to evaluate stratigraphy by visual examination 

and to obtain soil specimens for laboratory index testing. Disturbed samples are usually 

collected using split-barrel samplers (Figure 3-8; AASHTO T 206, ASTM D 1586), although 

several other techniques are available for disturbed sample collection in boreholes (see Table 

3-5(a) and 3-5(b)). Shallow disturbed samples can also be obtained by using hand augers 

and test pits. Direct push methods, such as GeoProbe sampling, can be used to obtain 

continuous disturbed samples but these methods have limitations in sampling depth similar to 

those of solid stem and bucket augers (i.e., generally good for depths less than 33 feet (10 

meter) unless in firm to stiff clays). Samples obtained via disturbed sampling methods are 

generally used for index property testing in the laboratory. They should not be used to 

prepare specimens for consolidation and strength tests. 

Figure 3-8. Split barrel sampler. 

3.6.2 Undisturbed Sampling of Soil 

Undisturbed soil samples are required for performing laboratory strength and consolidation 

tests on cohesive soils having consistencies ranging from soft to stiff. High-quality samples 

for such tests are particularly important for approach embankments and for structural 

foundations and wall systems that may stress compressible strata. In reality, it is impossible 

to retrieve truly undisturbed samples since changes in the state of stress in the sample occur 

upon sampling and removal of the sample from depth. The goal of high-quality undisturbed 

sampling is to minimize the potential for: (1) alteration of the soil structure; (2) changes in 

moisture content or void ratio; and (3) changes in chemical composition of the soil. Due to 

cost and ease of use, the thin-walled Shelby tube (Figure 3-9) is the most commonly used 

sampler for obtaining relatively undisturbed samples of soft to stiff fine-grained soils. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 28 December 2006 
Table 3-5(a) 

Common samplers to retrieve disturbed soil samples (modified after NAVFAC, 1986a

Sampler Typical Dimensions Soils that Give Best 

Results 

Method of Penetration 

Cause of Low Recovery 

Rem

Split Barrel Standard is 2 in (50 

mm) outside diameter (OD) and 1-3/8 in (35 mm) inside diameter (ID) 

All soils finer than gravel size particles that allow sampler to be driven; gravels invalidate drive data; A soil retainer may be required in granular soils. 

140 lb (64 kg) hammer driven 

Gravel may block sampler 

A SPa staand samdistu

Continuous helical- Diameters range 3 in Most soils above water Rotation Hard soils, cobbles, Met

flight auger to 16 in (75 to 400 

mm; penetrations to depths exceeding 50 ft (15 m) 

table; will not penetrate hard soils or those containing cobbles or boulders 

boulders soil 

can samaccobetwand surferrosam

FHWA NHI-06-088 3 Soils and Foundations – Volume I 3 - 29 
Table 3-5(b) 

Common samplers to retrieve disturbed soil samples (modified after NAVFAC, 1986a

Sampler Typical Dimensions Soils that Give Best 

Results 

Method of Penetration 

Cause of Low Recovery 

Rem

Disc auger Up to 3.5 ft (1 m) 

diameter; usually has maximum penetration depth of 25 ft (8 m) 

Most soils above water table; will not penetrate hard soils or those containing cobbles or boulders 

Rotation Hard soils, cobbles, 

boulders 

Methsoil pcan bsampaccoubetwand asurfain est

Bucket auger Up to 4 ft (1.2 m) 

diameter common; larger sizes available; with extensions, depth over 80 ft (25 m) are possible 

Most soils above water table; can penetrate harder soils than above types and can penetrate soils with cobbles and boulders if equipped with a rock bucket 

Rotation Soil too hard to 

penetrate 

Severavailwith choppslow used 

Test boring of large samples, Large Penetration Test (LPT) 

2 in to 3 in (50 to 75 mm) ID and 2.5 in to 3.5 in (63 mm to 89 mm) OD samplers (examples, Converse sampler, California Sampler) 

In sandy to gravelly soils 

Up to 350 lb (160 kg) 350 lb hammer driven 

Large gravel, cobbles, and boulders may block sampler 

Sampdisturbe repenetequivvalue due tequip

FHWA NHI-06-088 3 Soils and Foundations – Volume I 3 - 30 
Figure 3-9. Schematic of thin-walled (Shelby) tube (after ASTM D 4700) and photo of 

tube with end caps (FHWA, 2002b). 

Thin walled Shelby tube sampling is discussed in Section 3.5.3. Depending upon the in-situ 

condition of the fine-grained soil (e.g., stiffness and whether significant granular material is 

in the soil matrix), alternative sampling devices may be used to obtain nominally undisturbed 

soil samples. These alternative samplers include: 

• Stationary piston sampler (Figure 3-10); 

• Denison sampler (Figure 3-11); 

• Pitcher samplers (Figure 3-12); 

• Hydraulic piston sampler (Osterberg Sampler). 

Summary information on these samplers is provided in Table 3-6 and detailed procedures for 

these sampling techniques are provided in FHWA (1997, 2002b). Although not common for 

typical transportation-related projects, a variety of special samplers are available to obtain 

samples of soil and soft rocks. These specialty samplers include the retractable plug sampler, 

the Sherbrooke sampler, and the Laval sampler. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 31 December 2006 
Table 3-6 

Nominally undisturbed soil samplers (modified after NAVFAC, 1986a) 

Sampler Typical Dimensions Soils that Give Best 

Results 

Method of 

Penetration 

Cause of Disturbance 

or Low Recovery 

Shelby 3 in (76 mm) OD and 2-7/8 Cohesive fine-grained Pressing with Erratic pressure applied Simplest devi

tube in (73 mm) ID most or soft soils; gravelly relatively rapid, during sampling, boring should (ASTM D common; available from 2 and very stiff soils will smooth stroke; can hammering, gravel lowered; little 1587; AASHTO T 207) 

in to 5 in (50 to 127 mm) OD; 30 in (760 mm) sampler length standard 

crimp tube be carefully hammer 

driven but this will induce additional disturbance 

particles, crimping of tube edge, improper soil types for sampler, pressing tube greater than 80% of tube length 

suitable for h

Stationary 3 in (76 mm) OD most Soft to medium clays Pressing with Erratic pressure during Piston at end 

piston common; available from 2 and fine silts; not for continuous, steady sampling, allowing fluid and con

in to 5 in (50 to 127 mm) sandy soils stroke piston rod to move heavy drill rig 

OD; 30 in (760 mm) during press, improper samples gene

sampler length standard soil types for sampler compared wit

for hard, dens

Hydraulic 3 in (76 mm) OD is most Silts and clays, some Hydraulic or Inadequate clamping of Needs only st

piston common; available from 2 sandy soils compressed air drill rods, erratic adequate hyd

(Osterberg) in to 4 in (50 to 100 mm) 

OD; 36 in (910 mm) sampler length standard 

pressure pressure activate samp

disturbed comsuitable for h

Denison 3.5 in to 7 in (89 to 177 

mm) OD, producing samples 2-3/8 in 6.3 in (60 to 160 mm); 24 in (610mm) sampler length 

Stiff to hard clay, silt, and sands with some cementation, soft rock 

Rotation and hydraulic pressure 

Improper operation of sampler; poor drilling procedures 

Inner tube factube, which rocan be adjustesamples; not soft clays 

Pitcher 4 in (100 mm) OD; uses 3 Same as Denison Same as Denison Same as Denison Differs from D

sampler in (76-mm) diameter 

Shelby tubes; sample length 24 in (610 mm) 

projection is sineffective in 

Foil Continuous samples 2 in Fine grained soils Pushed into the ground Samplers should not be Samples surro

Sampler (50 mm) wide and as long including soft sensitive with steady stroke; used in soils containing stainless steel

as 65 ft (20 m) clays, silts, and varved 

clays 

Pauses occur to add segments to sample barrel 

fragments or shells prevent conta

forced into so

FHWA NHI-06-088 3 Soils and Foundations – Volume I 3 - 32 
FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 33 December 2006 

 igure 3-11. Denison sampler (FHWA, 

1997). 

F

Figure 3-10. Stationary piston sampler 

schematic (after ASTM D 4700) and photo 

(FHWA, 2002b). 
 

Figure 3-12. Pitcher sampler (FHWA, 1997, 2002b). 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 34 December 2006 
When dealing with relatively shallow soils that are very stiff, brittle, partially cemented, or 

that contain coarse gravel or stones, the best method to obtain large relatively undisturbed 

samples is by block sampling. Block sampling involves isolating a soil column, encasing it 

in paraffin wax, and covering it with an open-ended box or tube (usually about 12 in (300 

mm) square). The bottom is cut, sealed and covered, and the sample is transported to the 

laboratory. This sampling technique is generally difficult to implement at depths greater than 

approximately 10 ft (3 m). 

3.6.3 Thin-Walled (Shelby) Tube Sampling 

The importance of appropriate sampling practice using Shelby tubes cannot be overemphasized. Poor sampling practices, exposure to extreme temperatures, and careless 

handling of samples can cause sample disturbance that may result in misleading test results 

that can lead to uneconomical or unsafe designs. 

•	 Geometry of a Thin-Walled Tube: The area ratio (AR) and the inside clearance ratio 

(ICR) are parameters that are used to evaluate the disturbance potential for different 

types of soil samplers. These parameters are defined as follows: 

D 2 −D 2 

AR = o i x 100 percent 3-1a 

D 2 

i 

D − D ICR = i e 3-1b De 

where De = diameter at the sampler cutting tip, Di = inside diameter of the sampling tube, 

and Do = outside diameter of the sampling tube. For a sample to be considered 

undisturbed, the ICR should be approximately 1 percent and the AR should be 10 percent 

or less. Using a tube with this ICR value minimizes the friction buildup between the soil 

sample and the sampler during the advancement of the sampler. Using a tube with an AR 

value less than 10 percent enables the sampler to cut into the soil with minimal 

displacement of the soil. Thin-walled tubes (e.g., Shelby tubes) are typically 

manufactured to meet these specifications, but a thicker walled tube with an ICR of zero 

is commonly used in the Gulf states (e.g., Texas, Louisiana) to sample very stiff 

overconsolidated clays. The use of the thicker walled tube minimizes buckling of the 

sampler in the stiff deposits, and the ICR of zero minimizes sample expansion within the 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 35 December 2006 
tube. Additional information on suitable geometry for thin-walled tubes is provided in 

ASTM D 1587. 

•	 Sample Tube Inspection and Storage: Tubes received from the manufacturer should be 

inspected to assure that no damage has occurred to the ends of the tubes. Plastic end 

caps, which will later be used to facilitate securing of the sample, should be placed on 

the ends of the tube at this time. 

•	 Cleaning Borehole Prior to Sampling: Depending upon the methods used, drilling and 

sampling procedures will cause some disturbance in the vicinity of the advancing face of 

a borehole. This is especially the case if a sample is overpushed, if casing is advanced 

ahead of the borehole, or during continuous sampling operations. It is recommended 

that a borehole be advanced and cleaned to two to three diameters below the bottom of 

the previous sample to minimize disturbance. Additionally, after advancement of the 

borehole, caving may occur at the bottom of the hole. Thus, the bottom of the borehole 

should be cleaned out thoroughly before the sampling device is advanced. Improper 

cleaning will lead to severe disturbance of the upper material (accumulated settled 

material), and possibly disturbance of the entire sample. Cleaning is usually performed 

by washing materials out of the hole. It should be ensured that the jet holes are not 

directed downward, for this will erode soft or granular materials to an unknown depth. 

All settled material should be removed to the edge of the casing. In deep or wide 

borings, special cleaning augers may be used to decrease time for cleaning and produce 

a cleaner hole. 

•	 Tube Advancement and Retrieval: Tubes should be advanced without rotation in a 

smooth and relatively rapid manner. The length of the sampler advancement should be 

limited to 24 in (600 mm) for a 30 in (760 mm) long tube to minimize friction along the 

wall of the sampler and allow for loose material in the hole. The amount of recovery 

should be compared to the advanced length of the sampler to assess whether material has 

been lost, the sample has swelled, or some caved material has been collected at the top 

of the tube. The possible presence of caved material should be noted at the top of the 

tube so that no laboratory moisture content or performance tests are performed on that 

material. After advancing to the target depth, the drill rod should be rotated one full turn 

to shear off the bottom of the sample. Prior to shearing, a waiting period of 5 to 15 

minutes is recommended for tubes in soft soils to permit the sample to reach equilibrium 

inside the tube and prevent the sample from falling out the bottom of the tube during 

retrieval. This waiting period may be reduced for stiffer soils. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 36 December 2006 
•	 Preparation for Shipment: Upon removal of the sample from the borehole, the ends 

should be capped with the plastic end caps and the tube should be labeled. The label 

should be written directly on the tube with a permanent marking pen, and include: (1) 

tube and boring identification number; (2) sample depth; (3) top and bottom of sample; 

(4) length of recovery; (5) sampling date; (6) job name and/or number; and (7) sample 

description. Tube samples that are intended for laboratory performance testing (i.e., 

strength, consolidation, hydraulic conductivity) should never be extruded from the tube 

in the field and stored in alternative containers. Samples should be extruded only in the 

laboratory under controlled conditions. After a thin-walled tube sample has been taken, 

slough or cuttings from the upper end of the tube should be removed by use of a 

cleanout tool. The length of sample recovered should be measured and the soil 

classified for the log. About 1 in (25 mm) of material at the bottom end of the tube 

should be removed and the cuttings placed in a properly labeled storage jar. Both ends 

of the tube should then be sealed with at least a 1 in (25 mm) thick layer of 

microcrystalline (non-shrinking) wax after a plastic disk has been placed to protect the 

ends of the sample (Figure 3-13a). The use of relatively low temperature wax will 

minimize shrinkage and potential moisture migration within the sample. The remaining 

void above the top of the sample should be filled with moist sand. Plastic end caps 

should then be placed over both ends of the tube and electrician's tape wrapped over the 

joint between the collar of the cap and the tube and over the screw holes. The capped 

ends of the tubes are then dipped in molten wax. Alternatively, O-ring packers can be 

inserted into the sample ends and then sealed (Figure 3-13b). This method of sealing the 

sample may be preferable as it is cleaner and more rapid than waxing. In both cases, the 

sample must be sealed to ensure proper preservation of the sample. The tube should be 

kept vertical, with the top of the sample in the upright position. If the sample needs to 

be inverted for purposes such as sealing, care should be taken to ensure the sample does 

not slide within the tube. Samples must be stored upright in a protected environment to 

prevent freezing, desiccation, and alteration of the moisture content (ASTM D 4220).

 Shipment: Sample tubes must to be packed upright in accordance with guidelines 

provided in ASTM D 4220, or in an equivalent sample box. Tubes should be isolated 

from other sample tubes, and fit snugly in the case to protect against vibration or shock. 

The cushioning material between the samples should be at least 1 in (25 mm) thick, and 

the cushioning on the container floor should be at least 2 in (50 mm) thick. The samples 

should not be exposed to extreme heat or cold. If possible, the geotechnical specialist 

should deliver the samples to the laboratory or use a special delivery service provider 

who offers shipping of fragile items (e.g., FedEx) to ship samples. The use of a chain of 

custody form for sample traceability records is encouraged. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 37 December 2006 
(a) (b) 

Figure 3-13. Shelby tube sealing methods, (a) Microcrystalline Wax, (b) O-Ring 

packer (FHWA, 2002b). 

3.6.4 Undisturbed Sampling of Rock (Rock Coring) 

When equipment for rock coring is being considered, the dimensions, type of core barrel, 

type of coring bit, and drilling fluid are important variables. The minimum depth of rock 

coring should be determined based on the local geology of the site and the type of structure 

to be constructed. Coring should also be performed to a depth that assures that refusal is not 

encountered on a boulder. A brief description of issues related to rock coring is provided in 

this document. Additional information on drilling rigs, methods of circulating drill cuttings 

(i.e., fluid or air), hole diameters, and casings is provided in ASTM D 2113. 

3.6.4.1 Core Barrels 

Four different types of core barrels are described in ASTM D 2113 including: 

1) Single Tube - Figure 3-14(a); 

2) Rigid Double Tube - Figure 3-14(b); 

3) Swivel Double Tube - Figure 3-14(c); and 

4) Triple Tube. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 38 December 2006 
(a) 

(b) 

(c) 

Figure 3-14. (a) Single, (b) and (c) Double tube rock core barrels (FHWA, 1997). 

Since the double core barrel isolates the rock from the drilling fluid stream to yield 

better recovery, it is the minimum standard of core barrel that should be used in 

practice when an intact core is required for testing. Figure 3-15 shows the outer and 

inner assembly for a double-tube core barrel. The inner tube of a swivel-type core barrel 

does not rotate during drilling, which results in less disturbance and better recovery in weak 

and fractured rock. Rigid type double tube core barrels should not be used where core 

recovery is a concern. Triple tube swivel-type core barrels will produce better recovery and 

less core breakage than a double tube barrel. 

Most rock coring today is done by use of the wireline method, which was introduced in the 

1960s. In this method, an inner tube containing the core is detached from the core barrel 

assembly when the core barrel is full or a blockage occurs. The tube and core contained in it 

are pulled to the surface by wire dropped down the string of drill rods. A latch or “overshot 

assembly,” which snaps on to the top of the inner tube, is used for this purpose. The inner 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 39 December 2006 
tube is then rapidly hoisted to surface within the string of drill rods. After the core is 

removed, the inner tube is dropped down into the outer core barrel and drilling resumes. 

Thus, the core is retrieved without having to pull all of the rods and production rates, 

particularly for deep cores, are therefore greater than those for conventional techniques. 

Table 3-7 lists the available core sizes. The standard size rock core, NX, has a diameter of 2­

1/8 in (54 mm). Generally larger core sizes will lead to less mechanical breakage and yield 

greater recovery, but the associated cost for drilling will be much higher. Since the size of 

the core will affect the percent recovery, the core barrel size should be clearly recorded on 

the log. Additionally, the core barrel length can increase recovery in fractured and weathered 

rock zones. In these zones a core barrel length of 5 ft (1.5 m) is recommended. Core barrel 

lengths should not be greater than 10 ft (3 m) under any conditions because of the potential 

for damage to the long cores. 

(a) (b) 

Figure 3-15. Double tube core barrel. (a) Outer barrel assembly (b) Inner barrel 

assembly (FHWA, 2002b). 

3.6.4.2 Coring Bits 

The coring bit is the bottommost component of the core barrel assembly. It is the grinding 

action of this component that cuts the core from the rock mass. The following three basic 

categories of bits are in use: diamond, carbide and sawtooth (Figure 3-16). 

Coring bits are generally selected by the driller and are often approved by the geotechnical 

specialist. Bit selection should be based on a general knowledge of drill bit performance for 

the expected formations and the proposed drilling fluid. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 40 December 2006 
Table 3-7 

Dimensions of core sizes (FHWA, 1997) 

Size Diameter of Core Diameter of Borehole 

mm (in) mm (in) 

EX,EXM 21.5 (0.846) 37.7 (1.484) 

EWD3 21.2 (0.835) 37.7 (1.484) 

AX 30.1 (1.185) 48.0 (1.890) 

AWD4, AWD3 28.9 (1.138) 48.0 (1.890) 

AWM 30.1 (1.185) 48.0 (1.890) 

AQ Wireline, AV 27.1 (1.067) 48.0 (1.890) 

BX 42.0 (1.654) 59.9 (2.358) 

BWD4, BWD3 41.0 (1.614) 59.9 (2.358) 

BXB Wireline, BWC3 36.4 (1.433) 59.9 (2.358) 

BQ Wireline, BV 36.4 (1.433) 59.9 (2.358) 

NX 54.7 (2.154) 75.7 (2.980) 

NWD4,NWD3 52.3 (2.059) 75.7 (2.980) 

NXB Wireline, NWC3 47.6 (1.874) 75.7 (2.980) 

NQ Wireline, NV 47.6 (1.874) 75.7 (2.980) 

HWD4,HXB Wireline, 61.1 (2.406) 92.7 (3.650) 

HQ Wireline 63.5 (2.500) 96.3 (3.791) 

CP, PQ Wireline 85.0 (3.346) 122.6 (4.827) 

Figure 3-16. Coring bits: Diamond (top left), Carbide (top right), and Sawtooth (bottom 

center) (FHWA, 2002b). 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 41 December 2006 
Diamond coring bits, such as surface set or impregnated-diamond types, are the most 

versatile since they can produce high-quality cores in soft to extremely hard rock materials 

(see Figure 3-10, top left). Compared to other types, diamond bits in general permit more 

rapid coring and, as noted by Hvorslev (1948), exert lower torsional stresses on the core. 

Lower torsional stresses permit the retrieval of longer cores and cores of smaller diameter. 

The wide variation in the hardness, abrasiveness, and degree of fracturing encountered in 

rock has led to the design of bits to meet specific conditions known to exist or expected to be 

encountered at given sites. Thus, wide variations in the quality, size, and spacing of 

diamonds, in the composition of the metal matrix, in the face contour, and in the type and 

number of waterways are found in bits of this type. Similarly, the diamond content and the 

composition of the metal matrix of impregnated bits are varied to meet differing rock 

conditions. 

Carbide bits use tungsten carbide in lieu of diamonds. There are of several types of carbide 

bits. The standard type carbide bit is shown in Figure 3-16, top right. Bits of this type are 

used to core soft to medium hard rock. They are less expensive than diamond bits. 

However, the rate of drilling is slower than with diamond bits. 

Sawtooth bits consist of teeth cut into the bottom of the bit (see Figure 3-10, bottom center). 

The teeth are faced and tipped with a hard metal alloy such as tungsten carbide to provide 

water resistance and thereby to increase the life of the bit. Although these bits are less 

expensive than diamond bits, they do not provide as high a rate of coring and do not have a 

salvage value. The saw tooth bit is used primarily to core overburden and very soft rock. 

An important feature in all bits is the type of waterways provided in the bits for the passage 

of drilling fluid. Bits are available with so-called “conventional” waterways, which are 

passages cut on the interior face of the bit, or with bottom discharge waterways, which are 

internal passages that discharge at the bottom face of the bit behind a metal skirt separating 

the core from the discharge fluid. Bottom discharge bits should be used when soft rock or 

rock having soil-filled joints is cored to prevent erosion of the core by the drilling fluid 

before the core enters the core barrel. 

Bit selection is based on the anticipated rock formation as well as the expected drilling fluid. 

Diamond bits are applicable in all rock types. They permit greater rates of coring than other 

types of bits. Carbide bits are less expensive than diamond bits and can be used in soft to 

medium-hard rock. Sawtooth bits are the least expensive of the three, however they have no 

salvage value. They lead to slower coring and are typically used only in soft rock. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 42 December 2006 
3.6.4.3 Drilling Fluid 

In many instances, clear water is used as the drilling fluid in rock coring. If drilling mud is 

required to stabilize collapsing holes or to seal zones when there is loss of drill water, the 

geotechnical specialist should be notified to confirm that the type of drilling mud is 

acceptable. Drilling mud will clog open joints and fractures, which adversely affects 

permeability measurements and piezometer installations. Drilling fluid should be contained 

in a settling basin to remove drill cuttings and to allow recirculation of the fluid. Generally, 

drilling fluids can be discharged onto the ground surface. However, special precautions or 

handling may be required if the material is contaminated with oil or other substances. Such 

fluids may require disposal off site. Water flow over the ground surface should be avoided 

as much as possible. Local environmental agencies should be contacted for permits because 

some drilling fluids may have adverse effects on local surface and subsurface environments. 

Certain local agencies may also require implementation of a Storm Water Pollution 

Prevention Plan (SWPPP). 

3.6.5 Observations During Rock Core Drilling 

3.6.5.1 Drilling Rate/Time 

The drilling rate should be monitored and recorded on the boring log in the units of minutes 

per 1 ft (0.3 m). Only time spent advancing the boring should be used to determine the 

drilling rate. 

3.6.5.2 Core Photographs 

Cores in the split core barrel should be photographed immediately upon removal from the 

borehole. A label should be included in the photograph to identify the borehole, the depth 

interval and the number of the core run. It may be desirable to get a "close-up" of interesting 

features in the core. Wetting the surface of the recovered core by using a spray bottle and/or 

sponge prior to photographing will often enhance the color contrasts of the core. 

A tape measure or ruler should be placed across the top or bottom edge of the core box to 

provide a scale in the photograph. The tape or ruler should be at least 3 ft (1 m) long, and it 

should have relatively large, high contrast markings to be visible in the photograph. 

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A color bar chart is often desirable in the photograph to provide indications of the effects of 

variation in film age, film processing, and the ambient light source. The photographer should 

strive to maintain uniform light conditions from day to day, and those lighting conditions 

should be compatible with the type of film selected for the project. The use of a digital 

camera is a convenient way to circumvent some of the problems associated with the use of 

film cameras for photographing rock cores. 

3.6.5.3 Rock Classification 

The rock type and its inherent discontinuities, joints, seams, and other facets should be 

documented. See Chapter 4 for a discussion of rock description and classification. 

3.6.5.4 Recovery 

The core recovery is the length of rock core recovered from a core run. The recovery ratio is 

the ratio of the length of core recovered to the total length of the core drilled on a given run, 

expressed as either a fraction or a percentage. Core length should be measured along the 

core centerline. When the recovery is less than the length of the core run, the non-recovered 

section should be assumed to be at the end of the run unless there is reason to suspect 

otherwise (e.g., weathered zone, drop of rods, plugging during drilling, loss of fluid, and 

rolled or re-cut pieces of core). Non-recovery should be marked as NCR (no core recovery) 

on the boring log, and entries should not be made for bedding, fracturing, or weathering in 

that interval. 

Recoveries greater than 100 percent may occur if core that was not recovered during a run is 

subsequently recovered in the next run. Recoveries greater than 100 percent should be 

recorded and adjustments to data should not be made in the field. 

3.6.5.5 Rock Quality Designation (RQD) 

The RQD is a quantitative measure that represents a modified core recovery percentage. By 

definition the RQD is the sum of the lengths of all pieces of sound core over 4 in (100 mm) 

long divided by the length of the core run (Deere, 1963). The correct procedure for 

measuring RQD is illustrated in Figure 3-17. The RQD is an index of rock quality. 

Problematic rock that is highly weathered, soft, fractured, sheared, and jointed typically 

yields lower RQD values than more intact rock. Thus, RQD is simply a measurement of the 

percentage of "good" rock recovered from an interval of a borehole. 

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Soils and Foundations – Volume I 3 - 44 December 2006 
It should be noted that the original definition of RQD reported by Deere (1963) was based on 

measurements made on NX-size core. Experience in recent years reported by Deere and 

Deere (1989) indicates that cores with diameters both slightly larger and smaller than NX 

may be used for computing RQD. The wire line cores using NQ, HQ, and PQ are also 

considered acceptable. Use of RQD for the smaller BQ and BX sizes is discouraged because 

of a greater potential for core breakage and loss that would result in a smaller value of RQD. 

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Soils and Foundations – Volume I 3 - 45 December 2006 

L = 8 in 

L = 7.5 in 

L = 10 in 

< 4 in 

< 4 in 

> 4 in 

10 + 7.5 + 8 

48 

Core Run Total = 48 in 

Figure 3-17. Modified core recovery as an index of rock mass quality (FHWA, 1997). 
Length Measurements of Core Pieces 

The same piece of core could be measured three ways: along the centerline, from tip to tip, 

or along the fully circular barrel section (Figure 3-18). The recommended procedure is to 

measure the core length along the centerline. This method is advocated by the International 

Society for Rock Mechanics (ISRM), Commission on Standardization of Laboratory and 

Field Tests (ISRM, 1981). The centerline measurement is preferred because: (1) it results in 

a standardized RQD not dependent on the core diameter, and (2) it avoids unduly penalizing 

the rock quality for cases where fractures run parallel to the borehole and are cut by a second 

set of fractures. 

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L=9.5 in 

L=9.5 in 

LT=12 in 

LT=14.5 in 

LFC=7 in 

LFC=4.5 in 

4 in diameter core 4 in diameter core 4 in diameter core 

2 in diameter core 2 in diameter core 2 in diameter core 

Figure 3-18. Length measurements for core RQD determination (FHWA, 1997). 
Assessment of Soundness 

Pieces of core which are not "hard and sound" should not be counted for the RQD even 

though they possess the requisite 4 in (100 mm) length. The purpose of the soundness 

requirement is to downgrade the rock quality where the rock has been altered and weakened 

either by agents of surface weathering or by hydrothermal activity. Obviously, in many 

instances a judgment decision must be made as to whether or not the degree of chemical 

alteration is sufficient to reject the core piece. 

One commonly used procedure is not to count a piece of core if there is any doubt about its 

meeting the soundness requirement as evidenced by discolored or bleached grains, heavy 

staining, pitting, or weak grain boundaries. This procedure may unduly penalize the rock 

quality, but it errs on the side of conservatism. A second procedure that is occasionally used 

includes the altered rock within the RQD summed percentage, but indicates by means of an 

asterisk (RQD*) that the soundness requirements have not been met. The advantage of this 

method is that the RQD* will provide some indication of the rock quality with respect to the 

degree of fracturing, while also noting its lack of soundness. 

Core breaks caused by the drilling process should be fitted together and counted as one piece. 

Drilling breaks are usually evidenced by rough fresh surfaces. For schistose and laminated 

rocks, it is often difficult to discern the difference between natural breaks and drilling breaks. 

When in doubt about a break, it should be considered as natural in order to be conservative in 

the calculation of RQD for most uses. Obviously, this practice would not be conservative 

when the RQD is used as part of a ripping or dredging estimate. 

3.6.5.6 Drilling Fluid Recovery 

The loss of drilling fluid during the advancement of a boring can be indicative of the 

presence of open joints, fracture zones or voids in the rock mass being drilled. Therefore, the 

volumes of fluid losses and the intervals over which they occur should be recorded. For 

example, "no fluid loss" means that no fluid was lost except through spillage and filling the 

hole. "Partial fluid loss" means that a return was achieved, but the amount of return was 

significantly less than the amount being pumped in. "Complete water loss" means that no 

fluid returned to the surface during the pumping operation. A combination of opinions from 

the field personnel and the driller on this matter will result in the best estimate. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 47 December 2006 
3.6.5.7 Core Handling and Labeling 

Rock cores from geotechnical explorations should be stored in structurally sound core boxes 

made of wood or corrugated waxed cardboard (Figure 3-19). Wooden boxes should be 

provided with hinged lids, with the hinges on the upper side of the box and a latch to secure 

the lid in a closed position. 

Cores should be handled carefully during transfer from barrel to box to preserve mating 

across fractures and fracture-filling materials. Breaks in core that occur during or after the 

core is transferred to the core box should be refitted and marked with three short parallel 

lines across the fracture trace to indicate a mechanical break. Breaks made to fit the core into 

the core box and breaks made to examine an inner core surface should be marked as such. 

These deliberate breaks should be avoided unless absolutely necessary. Cores should be 

placed in the boxes from left to right, top to bottom. When the upper compartment of the box 

is filled, the next lower (or adjoining) compartment should be filled beginning at the lefthand side, and so on the same way until the box is filled. The depths of the top and bottom 

of the core and each noticeable gap in the formation should be marked by a clearly labeled 

wooden spacer block. 

If there is less than 100 percent core recovery for a run, a cardboard tube spacer of the same 

length as the core loss should be placed in the core box either at the depth of core loss, if 

known, or at the bottom of the run. The depth of core loss, if known, or length of core loss 

should be marked on the spacer with a black permanent marker. The core box labels should 

be completed using an indelible black marking pen. An example of recommended core box 

markings is shown in Figure 3-19. The core box lid should have identical markings both 

inside and out, and both exterior ends of the box should be marked as shown in Figure 3-19. 

For angled borings, depths marked on core boxes and boring logs should be those measured 

along the axis of the boring. The angle and orientation of the boring should be noted on the 

core box and the boring log. 

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Figure 3-19. Core box for storage of recovered rock and labeling. 

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Soils and Foundations – Volume I 3 - 49 December 2006 
3.6.5.8 Care and Preservation of Rock Samples 

A detailed discussion of sample preservation and transportation is presented in ASTM D 

5079. Four levels of sample protection are identified as follows: 

a) routine care, 

b) special care, 

c) soil-like care, and 

d) critical care. 

Routine care in placing rock core in core boxes will be used for most geotechnical 

explorations. ASTM D 5079 suggests enclosing the core in a loose-fitting polyethylene 

sleeve prior to placing the core in the core box. 

Special care is considered appropriate if the moisture state of the rock core (especially shale, 

claystone and siltstone) and the corresponding properties of the core may be affected by 

exposure. Special care can also be applied if it is important to maintain fluids other than 

water in the sample. Critical care is needed to protect samples against shock and vibration or 

variations in temperature, or both. For soil-like care, samples should be treated as indicated 

in ASTM D 4220. 

3.6.6 Geologic Mapping 

Geologic mapping is the systematic collection of local, detailed geologic data, and, for 

engineering purposes, is used to characterize and document the condition of a rock mass or 

outcrop. The data derived from geologic mapping are a portion of the data required for the 

design of a cut slope or for the stabilization of an existing slope. Geologic mapping can 

often provide more extensive and less costly information than drilling. Soil and soil-like 

materials, although occasionally mapped, are not considered in this section. For a detailed 

discussion of geologic mapping, the reader is referred to the FHWA manual on rock slopes 

(FHWA, 1998a). 

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3.7 STANDARD PENETRATION TEST (SPT) 

The standard penetration test (SPT) is performed during the advancement of a soil boring to 

obtain a disturbed drive sample (split barrel type) of the soil being penetrated and an 

approximate measure of its dynamic resistance. The test was introduced by the Raymond 

Pile Company in 1902 and remains today as the most common in-situ test performed 

worldwide. The procedures for the SPT are detailed in ASTM D 1586 and AASHTO T 206. 

A summary of the important features of the test follows. 

The SPT involves the driving of a hollow thick-walled tube into the ground and measuring 

the number of blows to advance the split-barrel sampler having standard dimensions of 2 in 

(50 mm) outside diameter (OD) and 1-3/8 in (35 mm) inside diameter (ID) a vertical distance 

of 1 ft (300 mm), see Figure 3-20. A 140 pound (63.5 kilogram) hammer is repeatedly 

dropped from a height of 30 in (0.76 m) to achieve three successive 6 in (150 mm) 

increments of penetration. The first recorded increment is considered as a “seating” 

penetration, while the number of blows to advance the second and third increments are 

summed to give the N-value ("blow count") or SPT-resistance (reported in blows per foot 

(0.3 m)). 

The SPT can be halted when a total of 100 blows have been counted or if the number of 

blows exceeds 50 in any given 6 in (150 mm) increment, or if the sampler fails to advance 

during 10 consecutive blows. SPT refusal is defined by penetration resistances exceeding 

100 blows per 2 in (50 mm), although ASTM D 1586 has re-defined this limit at 50 blows 

per 1 in (25 mm). If bedrock, or an obstacle such as a boulder, is encountered, the boring 

may be advanced further by using diamond core drilling or non-core rotary methods (ASTM 

D 2113; AASHTO T 225) at the discretion of the geotechnical specialist. In certain cases, 

this SPT criterion may be utilized to define the top of bedrock within a particular geologic 

setting where boulders are not of concern or not of great impact on the project requirements. 

The advantages and disadvantages of the SPT are listed in Table 3-8. 

Table 3-8 

Advantages and disadvantages of the Standard Penetration Test (SPT)
 

Advantages Disadvantages 

• Obtain both a sample and an N-value 

• Simple and rugged 

• Suitable in many soil types 

• Can be performed in weak rocks 

• Readily available throughout the U.S. 

• Disturbed sample (index tests only) 

• N-value is a crude number for analysis 

• Not applicable in soft clays & loose silts 

• High variability and uncertainty 

• Unreliable in gravelly soils 

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Soils and Foundations – Volume I 3 - 51 December 2006 
63.5-kg Drop

Hammer

Repeatedly

Falling 0.76 m

 

0.15 

m 0.15 

m 0.15 

m

N = No. of Blows

per 0.3 meters

Anvil 

Split-Barrel 

(Drive) Sampler 

[Thick Hollow 

Tube]:

 O.D. = 50 mm

 I.D. = 35 mm 

L = 760 mm Hollow Sampler Driven in 3 Successive Increments

Seating 

Borehole 

Drill Rod 

(”N” or

“A” Type) 

Need to Correct to a Reference 

Energy Efficiency of 60% 

(ASTM D 4633) 

First Increment 

Second Increment 

Third Increment 

Note: Occasional 

Fourth Increment Used 

to provide additional 

soil material

 SPT Resistance (N-value) 

or “Blow Counts” is total 

number of blows to drive 

sampler last 300 mm (or 

blows per foot). 

} 

Standard Penetration Test (SPT)

 Per ASTM D 1586 

Standard Drill Rod Sizes 

Symbol OD, in* 

A 1-5/8 

N 2-3/8 

* 1 in = 25.4 mm 

140 lb (63.5 

kg) hammer 

falling 30 in 

(760 mm) 6 in 6 in 6 in

N = No. of blows

per ft 

Figure 3-20. Sequence of driving split-barrel sampler during the Standard Penetration 

Test (modified after FHWA, 2002b). 

The SPT is conducted at the bottom of a soil boring that has been advanced by use of either 

flight augers or rotary wash drilling methods. The borehole can be cased or uncased. At 

regular depth intervals, the drilling process is interrupted to perform the SPT. Generally, at 

depths shallower than 10 ft (3 m) the SPT is performed continuously or at intervals of 2.5 ft 

(0.75 m). Below a depth of 10 ft (3 m) the SPT is generally performed at intervals of 5 ft 

(1.5 m) to the planned end of the boring or refusal. If the borehole extends below the 

groundwater table, the head of water in the borehole must be maintained at or above the 

ambient groundwater level to avoid inflow of water and borehole instability. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 52 December 2006 
 

Liners may be placed inside the split-barrel sampler with the same inside diameter as the 

cutting shoe, see Figure 3-21a. This allows samples to remain intact during transport to the 

laboratory. The liners may be arranged in a set of 1-inch (25 mm) high rings in which case 

“ring” samples of pre-determined height may be obtained. In U.S. practice, it is normal to 

omit the liner. The resistance of the sampler to driving is altered depending upon whether or 

not a liner is used (Skempton 1986, Kulhawy and Mayne, 1990). Therefore, when the liners 

are used, their use should be clearly mentioned in the boring logs. 

Steel or plastic sampler “catchers” are often required to keep samples of clean granular soils 

in the split-barrel sampler. Figure 3-21 shows a variety of catchers. They are inserted inside 

the sampler between the cutting shoe and the sample barrel to help retain loose or flowing 

materials. These catchers permit the soil to enter the sampler during driving but upon 

withdrawal they close and thereby retain the sample. Use of sample catchers should be 

noted on the boring log. 

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(a) 

(b) 

Figure 3-21: (a) Stainless steel and brass liners, (b) Sample catchers (FHWA, 2002b). 
3.7.1 Energy Efficiency of Hammers 

In current U.S. practice, three types of drop hammers and four types of drill rods are used in 

performing the SPT. Drop hammer types are typically donut, safety, and automatic (see 

Figure 3-22). Typical drill rod sizes are N or A (see Figure 3-20 for sizes). The test results 

are highly dependent upon the type of equipment used and the experience of the operator 

performing the test. One of the more important factors for obtaining useful data from the test 

is the energy efficiency of the system. The theoretical energy of a free-fall system with the 

specified mass and drop height is 350 ft-lb (48 kg-m), but the actual energy is less due to a 

number of factors including frictional losses and eccentric loading that are specific to the 

hammer drop. The energy efficiency of the rotating cathead and rope system commonly used 

in the past depends on numerous factors including: type of hammer, number of rope turns, 

conditions of the sheaves and rotating cathead (e.g., lubricated, rusted, bent, new, old), age of 

the rope, actual drop height, vertical plumbness, weather and moisture conditions (e.g., wet, 

dry, freezing), and other variables (see for example Skempton, 1986). In the recent past the 

trend has been towards the use of automated systems for lifting and dropping the mass in 

order to minimize these factors. Automated systems provide more reliable and more 

reproducible results than the rotating cathead and rope system used in the past. 

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 (a) (b) (c) 

Figure 3-22. SPT hammer types, (a) Donut, (b) Safety, and (c) Automatic (FHWA, 

2006a). 
A calibration of energy efficiency for a specific drill rig and operator is recommended by 

ASTM D 4633. Instrumented strain gages and accelerometer measurements are used for 

these calibrations in an attempt to standardize the energy levels. The standard of practice for 

energy efficiency varies from about 35% to 85% with cathead systems using donut or safety 

hammers. The average for cathead systems in the United States is approximately 60%. The 

newer automatic trip-hammers can deliver between 80 to 100% efficiency, depending upon 

the type of commercial system being used. 

If energy efficiency (Ef) is measured, then the energy-corrected SPT N-value adjusted to 

60% efficiency (N60) is given by: 

N60 = (Ef/60) Nmeas 3-2

where Nmeas

field should be corrected to N60 for all soils, if possible. The relative magnitudes of 

corrections for energy efficiency, sampler lining, rod lengths, and borehole diameter are 

given by Skempton (1986) and Kulhawy and Mayne (1990), but only as general guidelines. 

Theoretically it is mandatory to measure Ef to get the proper correction to N60. In absence of 

data, AASHTO (2004 with 2006 Interims) recommends Ef = 60 for rope and cathead 

systems, i.e., donut and safety hammers and Ef = 80 for automatic hammer systems. 

The efficiency may be obtained by comparing either the work done (W = F.d = force times 

displacement) or the kinetic energy (KE = 1⁄2mv2

) with the potential energy of the system (PE 

= mgh), where m = mass, v = impact velocity, g = 32.2 ft/s2

 = 9.8 m/s2= gravitational 

constant, and h = drop height. Thus, the energy ratio (ER) is defined as W/PE, or ER = 

KE/PE. It is important to note that geotechnical foundation practice and engineering 

usage based on SPT correlations have been developed on the basis of the standard-ofpractice, corresponding to an average ER ≈ 60 %. Thus, it is recommended to adjust 

measured N-values (Nmeas) to N60 values. 

Figure 3-23 exemplifies the need for correcting measured N-values to a reference energy 

level where the successive SPTs were conducted by alternating the use of donut and safety 

hammers in the same borehole. The energy ratios were measured for each test and gave 34 < 

ER < 56 for the donut hammer (average = 45%) and 55 < ER < 69 for the safety hammer 

(average = 60%) at this site. The individual trends for the measured N-values from donut 

and safety hammers are quite apparent in Figure 3-23(a), whereas a consistent profile is 

obtained in Figure 3-23(b) once the data have been corrected to ER = 60%. 

 is the N-value measured in the field during the test. N-values measured in the 

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 6 

8 

Measured N-values Corrected N60
 

0 10 20 30 40 50 
 0 10 20 30 40 50 

4 4

ER = 34 (energy ratio) 

55 

60 

4

4

5 

0 

56 

63 

41 

41 

63 

63 

39 

Donut 

Saf et y 

Sequence 

64 

69 

47 

56 

16 

Donut 

Saf et y 

Tr end 

(a) (b) 

6 

Depth (meters)8 

10 

Depth (meters)

12

10 

12 

14 

16 

14 

Figure 3-23. SPT N-values from (a) Uncorrected data, and (b) Corrected to 60% 

efficiency (Data modified after Robertson and Campanella, 1983). 

3.7.2 Effect of Overburden Stress on N-values 

Since N-values of similar materials increase with increasing effective overburden stress, the 

corrected blow count (N60) is often normalized to 1-atmosphere (1 tsf or about 100 kPa) 

effective overburden stress by using overburden normalization schemes. The energycorrected blow count normalized for overburden is referred to as N160, and is equal to: 

N160=CN N60 3-3

where CN is the overburden correction factor (or stress normalization parameter) calculated 

as (Peck, et al., 1974): 

CN = [0.77 log10 (20/po)], and CN < 2.0 

po = vertical effective pressure at the depth where the SPT test is performed (tsf) 

N60 = SPT blow count corrected for hammer efficiency (blows/ft) – refer to Equation 3-2. 

Note that the constants in Equation 3-3 are unit dependent therefore the units of po must be tsf. 

Figure 3-24 presents the overburden correction factor as a function of vertical effective stress. 

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Soils and Foundations – Volume I 3 - 56 December 2006 
Overburden Correction Factor, CN 

0.0 0.2 0.4 0.6 0.8 1.0 1.2 1.4 1.6 1.8 2.0
 

0 

1 

2 

3 

4 

5 

6 

7 

8 

9 

10 

Vertical Effective Pressure, tsf 

Figure 3-24. Variation of overburden correction factor, CN, as a function of vertical 

effective stress. 

Caution should be exercised in applying the overburden correction factor to indurated cemented 

soils, e.g., hard caliche soils encountered in the desert southwest. In such soils, the overburden 

pressure may not be a direct function of the depth of the soil. Therefore, the overburden 

correction is not recommended for such soils since it may lead to overly conservative designs. 

FHWA NHI-06-088 3 – Subsurface Explorations 

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3.7.3 Correlation of SPT N-Values with Basic Soil Characteristics 

SPT N-values are an indication of the relative density of cohesionless soils and the consistency 

of cohesive soil. Table 3-9 shows N-value ranges correlated to the relative density of sands and 

the consistency of fine-grained soils. It is emphasized that these correlations are unreliable for 

gravels, silts and clays and should serve only as crude estimates for these materials. 

Table 3-9 

Soil properties correlated with Standard Penetration Test values (after Peck, et al., 1974) 

Sands (Reliable) Silts and Clays (Unreliable) 

N60 Relative Density N60 Consistency 

0-4 Very loose Below 2 Very soft 

5-10 Loose 2-4 Soft 

11-30 Medium Dense 5-8 Medium 

31-50 Dense 9-15 Stiff 

Over 50 Very dense 16-30 Very stiff 

Over 30 Hard 

3.7.3.1 Applicability of SPTs in Gravelly Soils 

The SPT can be performed in a wide variety of soil types as well as weak rocks, however the 

SPT is not particularly useful in the characterization of gravelly soils. Since the split-spoon 

inside diameter is 1- 3⁄8 in (35 mm), gravel sizes larger than 1-3⁄8 in (35 mm) will not enter the 

spoon. Therefore, soil descriptions may not reflect actual gravel content of the deposit. Also, 

gravel pieces may plug the end of the spoon and cause the SPT blow count to be erroneously 

large. Thus, the SPT in such cases produces refusal blow counts (i.e., > 50 blows per 1 in (25 

mm)) that are misleading and lead to unconservative designs. In this case, “Large 

Penetration Tests” (LPTs), such as the Becker Penetration Test (BPT), are more suitable. 

The LPTs consist of driving a pipe (casing) larger than the standard split spoon sampler into 

the ground with a pile-driving hammer. While the pipe is being driven, the driving resistance 

or blow count/ft of penetration is recorded. Unlike the SPT, the LPT blow count is nonstandardized and is a function of the drill rod size, pipe (sampler) size, hammer type, and 

hammer efficiency. Careful energy calibrations are required to correlate the LPT blow 

counts to SPT N-values However, this effort may be worthwhile considering that the results 

of SPT in gravelly soils are unreliable and misleading. Daniel, et al. (2003) present methods 

for evaluating LPT blow counts. 

Since the gravel content cannot be measured by the SPT, it is recommended that 

consideration be given to obtaining bulk samples by drilling large diameter borings with 

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augers similar to the one shown in Figure 3-5a. The bulk samples obtained from such 

borings will also help evaluate whether the soil deposit is indeed a gravel deposit or gravels 

are larger particles floating in a softer soil matrix. The bulk samples will also permit an 

accurate determination of the Unified Soil Classification System (USCS) designation (see 

Chapter 4), which will be useful from design as well as constructability considerations. 

3.7.4 SPT Test Errors 

Although the procedures for conducting the SPT test have been standardized, several errors can 

creep into the test. The most common errors are: 

1.	 Effect of overburden pressure. Soils of the same density will give smaller blow counts 

near the ground surface. The overburden stress normalization parameter (CN) can be 

used to correct for this factor. 

2.	 Variations in the 30 in (770 mm) free fall of the drive weight. The drop height is often 

gauged by eye with the older rotating cathead and rope system Newer hammer 

systems automatically release the weight at a height of 30 inches. The energy 

correction factor accounts for this factor. 

3.	 Interference with the free fall of the drive weight by the guides or the hoist rope 

required in the rotating cathead and rope system. Newer automatic hammer systems 

eliminate rope interference. The energy correction factor accounts for this factor. 

4.	 Use of a drive shoe that is damaged or worn from too many "refusal" blow counts 

(Nmeas ≥ 100 blows/foot). 

5.	 Failure to seat the sampler properly on undisturbed material in the bottom of the boring. 

6.	 Inadequate cleaning of loosened material (slough) from the bottom of the boring. 

7.	 Failure to maintain sufficient hydrostatic pressure in the borehole during drilling below 

the groundwater table. Unbalanced hydrostatic pressures between the borehole drill 

water and the groundwater table can cause the test zone to become "quick." This can 

happen when a continuous-flight auger is used with the end plugged and with a water 

level in the hollow stem below that in the hole. 

8.	 Effect of gravel size as discussed in Section 3.7.3.1. 

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Soils and Foundations – Volume I 3 - 59 December 2006 
9.	 Samples retrieved from dilatant soils (fine sands, sandy silts) that exhibit unusually high 

blow counts should be examined in the field to determine if the sampler drive shoe is 

plugged. Poor sample recovery is usually an indication of plugging. 

10. Careless work on the part of the drill crew. 

The use of qualified and experienced drillers cannot be overemphasized. Agencies that 

maintain their own drilling personnel and equipment generally achieve much more reliable 

and consistent results than those that routinely let boring contracts to the lowest bidder. 

Soil type, density, and overburden pressure are the most significant factors affecting SPT Nvalues (assuming good workmanship and equipment). Table 3-10 lists factors affecting the 

SPT and SPT results. 

Regardless of the impressive list of shortcomings, the SPT is not likely to be abandoned for 

several reasons: 

1.	 The test is very economical in terms of cost per unit of information. 

2.	 The test results provides soil samples, which can be tested for index properties and 

visually examined. 

3.	 Long service life of the enormous amount of equipment in use. 

4.	 The accumulation of a large SPT database that is continually expanding. 

5.	 The results of the SPT have been correlated with a number of soil properties to provide 

estimates of the values of those properties. The estimated values are often used for 

preliminary designs in lieu of values obtained from tests run specifically to determine 

those properties. 

6.	 The fact that other methods can be readily used to supplement the SPT when the 

borings indicate more refinement in sample/data collection. 

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Soils and Foundations – Volume I 3 - 60 December 2006 
Table 3-10 

Factors affecting the SPT and SPT results (after Kulhawy and Mayne, 1990) 

Cause Effects Influence on 

SPT N-value 

Inadequate cleaning of hole SPT is performed in loose slough. 

Therefore soil may become trapped in 

sampler and may be compressed as 

sampler is driven, reducing recovery 

Increases 

Failure to maintain adequate 

head of water in borehole when 

test is performed below 

groundwater level 

Bottom of borehole may become 

“quick” 

Decreases 

Careless measure of hammer 

drop 

Hammer energy varies (generally 

variations cluster on low side) 

Increases 

Hammer weight inaccurate Hammer energy varies (driller supplies 

weight; variations of 5 - 7 percent are 

common) 

Increases or 

decreases 

Hammer strikes drill rod collar 

eccentrically 

Hammer energy reduced Increases 

Lack of hammer free fall 

because of ungreased sheaves, 

new stiff rope on weight, more 

than two turns on cathead, 

incomplete release of rope each 

drop 

Hammer energy reduced Increases 

Sampler driven above bottom 

of casing 

Sampler driven in disturbed, artificially 

densified soil 

Increases 

greatly 

Careless counting of hammer 

blows 

Inaccurate results Increases or 

decreases 

Use of non-standard sampler Correlations with SPT sampler invalid Increases or 

decreases 

Coarse gravel or cobbles in soil Sampler becomes clogged or impeded Increases 

Use of bent drill rods Inhibited transfer of energy of sampler Increases 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 61 December 2006 
3.8 LOG OF BOREHOLE INFORMATION (“BORING LOGS”) 

The importance of accurate field notes and good logging of boreholes cannot be 

overemphasized. The logger must realize that a good field description must be recorded. The 

field-boring log is the major portion of the factual data used in the analysis of foundation 

conditions. 

The boring log is a record that should contain all of the information obtained from a boring 

whether or not it may seem important at the time of drilling. It is important to record the 

maximum amount of information accurately. This record is the "field" boring log, as opposed 

to the "finished" boring log used in the preparation of the geotechnical data report. The 

finished log is drawn from the data presented in the field log supplemented by the results of 

visual identifications of samples and classification tests made in the laboratory. A typical 

boring log form is shown on Figures 3-25. The form presented in Figures 3-25 can be used for 

recording field data as well. 

3.8.1 Boring Log Format 

A wide variety of boring log forms are used by various agencies. The specific log to be used 

for a given type of boring will depend on local practice. The log in Figures 3-25 is just one 

example of a log used by geotechnical specialists. For detailed information on boring logs, the 

reader can refer to FHWA (1997, 2002b). The boring log shown in Figures 3-25 is used in this 

document simply to present the reader with an idea of the basic information that should be 

included in a boring log. Specific projects will likely require more detailed logs. Often separate 

logs are used for logging information from borings in soils and rocks unlike the log shown in 

Figures 3-25, which combines this information. 

3.8.2 Duties of the Logger 

The technical background and experience of the person who logs the field information will vary 

by organization. Some organizations will have a geotechnical engineer, an engineering 

geologist, a geologist, or a trained technician to accompany the drill crew, while others may 

train the drill crew foreman to log the borehole. In order to obtain the maximum amount of 

accurate data, the logger should work closely with the driller and be alert for changes in 

materials and operations while drilling is being performed. The logger is generally responsible 

for recording the following basic information on the field boring log: 

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Soils and Foundations – Volume I 3 - 62 December 2006 
Figure 3-25a. Example subsurface exploration log (0 – 35 ft depth). 

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Soils and Foundations – Volume I 3 - 63 December 2006 
Figure 3-25b. Example subsurface exploration log (35 - 60 ft depth). 

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Soils and Foundations – Volume I 3 - 64 December 2006 
1.	 General description of each soil and rock stratum, and the depth to the top and bottom 

of each stratum. As noted before, the log demonstrated in Figure 3-25 is intended to be 

a field log. On the final log, the description of the soil should be much more detailed 

and follow a specified soil classification system. Soils in the geotechnical engineering 

community are most often classified according to the Unified Soil Classification 

System (USCS). For example, the soils between depths 15 ft to 39 ft have been simply 

described in the field as “GR. SILTY CLAY, MOIST-PLASTIC.” The full 

classification as per USCS on the final log may read as follows (the detailed description 

and classification of soils and the USCS are discussed in Chapter 4): 

Soft, wet, gray, high plasticity CLAY, with Silt; Fat CLAY (CH); (Alluvium) 

2.	 The depth to groundwater at the time it is first encountered and afterwards at the end of 

each day, at completion of boring, and, if possible, at least 24 hours after completion of 

the boring. 

3.	 The depth at which each sample is taken, the type of sample taken, its number, and any 

loss of samples taken during extraction from the hole. 

4.	 The depths at which field tests are made and the results of the test. 

5.	 Information generally required by the log format, such as: 

•	 Boring number and location. 

•	 Date of start and finish of the hole. 

•	 Name of driller (and of logger, if applicable). 

•	 Elevation at top of hole. 

•	 Depth of hole and reason for termination. 

•	 Diameter of any casing used. 

•	 Size of hammer and free fall used on casing (if driven). 

•	 Blows per foot to advance casing (if driven). 

•	 Description and size of sampler. 

•	 Size of drive hammer and free fall used on sampler in dynamic field tests. 

•	 Blow count for each 6 in (150 mm) to drive sampler. (Sampler should be driven 

three 6 in (150 mm) increments or to a total of 100 blows). 

•	 Type of drilling rig used. 

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Soils and Foundations – Volume I 3 - 65 December 2006 
•	 Type and size of core barrel used. 

•	 Length of time to drill each core run or foot of core run. 

•	 Length of each core run and amount of core per run. 

•	 Recovery of sample in inches and RQD of rock core. 

•	 Project identification. 

6.	 Notes regarding any other pertinent information and remarks on miscellaneous 

conditions encountered, such as: 

•	 Depth of observed groundwater, elapsed time from completion of drilling, 

conditions under which observations were made, and comparison with the elevation 

noted during reconnaissance (if any). 

•	 Artesian water pressure. 

•	 Obstructions encountered. 

•	 Difficulties in drilling (caving, coring boulders, surging or rise of sands in casing, 

caverns, etc.). 

•	 Loss of circulating water and addition of extra drilling water. 

•	 Drilling mud and casing as needed and why. 

•	 Odor of recovered sample. 

•	 Sampler plugged. 

•	 Poor recovery. 

7.	 Any other information the collection of which may be required by agency policy (e.g., 

names and associations of visitors to the site, etc.). 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 66 December 2006 
3.9 CONE PENETRATION TESTING (CPT) 

The history of field cone penetrometers began with a design by the Netherlands Department 

of Public Works in 1930. This "Dutch" cone penetrometer was a mechanical operation using 

a manometer to read loads. Paired sets of inner and outer rods are pushed into the ground in 

8 in (200 mm) intervals. In 1948, electric cones permitted continuous measurements to be 

taken downhole. In 1965, the addition of sleeve friction measurement devices allowed an 

indirect means for identifying soil types. Later, in 1974, the electric cone was combined with 

a piezoprobe to form the first piezocone penetrometer. Most recently, additional sensors 

have been added to form specialized devices such as the resistivity cone, acoustic cone, 

seismic cone, vibrocone, cone pressuremeter, and lateral stress cone. 

The cone penetration test (CPT) was first introduced in the U.S. in 1965. Since that time, the 

CPT has developed into one of the most popular in-situ testing methods because it is fast, 

economical, and provides continuous profiling of the geostratigraphy and allows for 

continuous in-situ evaluation of soil properties. Depending upon equipment capability as 

well as soil conditions, 330 to 1150 ft (100 to 350 m) of penetration testing may be 

completed in one day. 

As shown in Figure 3-26, the CPT involves the hydraulic push of an instrumented steel probe 

into the soil at a constant rate to obtain continuous vertical profiles of stresses and/or other 

measurements. No borehole, cuttings, or spoil are produced by this test. Testing is 

conducted in accordance with ASTM D 5778. 

The CPT can be used in very soft clays to dense sands. It is not suitable for use in highly 

indurated or cemented soils or in soils containing significant amounts of gravel and boulders. 

The advantages and disadvantages of the CPT are listed in Table 3-11. 

Table 3-11 

Advantages and disadvantages of the Cone Penetration Test (CPT) (FHWA, 2002b) 

Advantages Disadvantages 

• Fast and continuous profiling 

• Economical and productive 

• Results not operator-dependent 

• Strong theoretical basis in interpretation 

• Particularly suitable for soft soils 

• High capital investment 

• Requires skilled operator to run 

• Electronic drift, noise, and calibration 

• No soil samples are obtained 

• Unsuitable for gravel/boulder deposits* 

*Note: Except where special rigs are provided and/or additional drilling support is available. 

FHWA NHI-06-088 3 – Subsurface Explorations 

Soils and Foundations – Volume I 3 - 67 December 2006 
Figure 3-26. Procedures and components of the Cone Penetration Test (FHWA, 2002b). 

Although the test provides continuous logging of the in-situ response of the soil, which can 

lead to more accurate and reliable analyses, no soil samples are available for laboratory 

testing. For that reason the CPT provides an excellent complement to the more conventional 

soil test boring with SPT measurements and subsequent laboratory testing on retrieved 

samples. 

3.9.1 Equipment Description and Operation 

Electronic cones are now the dominant cone type used in cone penetration testing. 

Therefore, mechanical cones are not discussed in this document. Electronic cones may be 

further divided into three primary types: (a) the standard friction cone (CPT), (b) the piezocone (PCPT or more commonly CPTu), and (c) the seismic cone piezo-cone (SCPTu). Each 

of these cones is briefly described here. To assist in following the brief descriptions, the 

standard terminology regarding the cone penetrometer is shown in Figure 3-27. 

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Soils and Foundations – Volume I 3 - 68 December 2006 
Figure 3-27. Cone penetrometer terminology (from Robertson and Campanella, 1989). 

3.9.2 The Standard Cone Penetration Test (CPT) 

The equipment necessary for performing a standard CPT includes a penetrometer, cone rod 

or drill rod, an electrical cable, a data acquisition system, and a hydraulic actuator attached to 

equipment that has sufficient reaction mass to advance the penetrometer. The equipment that 

provides the reaction mass can be a conventional drilling rig, however, a dedicated CPT 

truck commonly weighing 20 to 25 tons (200 to 250 kN) is more commonly used. 

A standard cone penetrometer is a 1.4 in (35.7 mm) diameter cylindrical probe with a 60o 

conical apex at the tip. The tip has a projected area of 1.6 in2

 (10 cm2). The surface area of 

the sleeve above the cone is 23.3 in2

 (150 cm2). More robust penetrometers are available 

with a 1.7 in (44 mm) diameter body, a 2.3 in2

 (15 cm2) projected tip area, and a 31 to 35 in2 

(200 to 225 cm2

) sleeve surface area. A penetrometer having a projected cone area of 2.3 in2 

(15 cm2

) will generally provide the same response as one having a projected cone area of 1.6 

in2

 (10 cm2). The “size” of a cone is defined by its projected tip area, e.g. a 1.6 in2 (10 cm2) 

cone or a 2.3 in2

 (15 cm2) cone. Figure 3-28 shows a number of different cone penetrometers 

and piezocones. 

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Soils and Foundations – Volume I 3 - 69 December 2006 
Figure 3-28. Cone and piezocone penetrometers (note the quarter for scale) (FHWA, 

2002b). 

A section of standard cone rod is typically 3.3 ft (1 m) in length with a 1.4 in (35.7 mm) 

outer diameter and a 0.9 in (22 mm) inner diameter. Alternatively, the penetrometer can be 

pushed with standard AW drill rod (13⁄4 in (44.4 mm) OD; 11⁄4 in (31.8mm) ID) or EW drill 

rod (13⁄8 in (34.9 mm) OD; 15/16 in (23.8 mm) ID). 

The cone cable runs through the hollow cone/drill rods and attaches to an electronic data 

acquisition system at the ground surface. The data acquisition system generally consists of 

an analog signal conditioner, an analog to digital (A-D) converter, and a computer processor. 

Current data acquisition systems are attached to one or two computer monitors so the 

operator and engineer can observe data recorded during the sounding in real time. Real time 

monitoring allows for decisions to be made in the field with respect to the sounding. This is 

helpful if auxiliary tests, such as a pore pressure dissipation test, are to be performed in 

certain soil layers, or if the test is to be terminated once a certain layer is encountered. 

Printers can be attached to the computer processor to obtain a real-time printout of the data. 

Printed data are a good backup in case an unforeseen incident causes the computer to crash 

resulting in the loss of the electronically stored data. Data are typically recorded every 3⁄4 to 

2 in (20 to 50 mm) of vertical penetration. 

The test procedure for the standard cone penetration test and the nature of the data acquired 

during the test are described in Sections 3.9.5 and 3.9.6, respectively. 

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Soils and Foundations – Volume I 3 - 70 December 2006 
3.9.3 The Piezo-cone Penetration Test (CPTu) 

The piezo-cone (CPTu) is essentially the same as the standard electronic friction cone except 

that it includes porous filter piezo-elements that may be located at the cone tip, on the cone 

face, behind the cone tip, or behind the friction sleeve. These porous filter elements are used 

to measure pore water pressure during penetration. Saturation of the porous element and 

cavity is essential to obtain reliable pore water pressure measurements. 

3.9.4 The Seismic Piezocone Penetration Test (SCPTu) 

For the seismic piezocone test (SCPTu), a geophone is located approximately 1.6 feet (500 

millimeters) uphole from the cone tip. The geophone detects shear waves generated at the 

ground surface at intervals of approximately 3 or 5 ft (1 or 1.5 m), corresponding to 

successive rod additions. If necessary, adjustments should be made if AW or EW rods are 

used to advance the cone since they typically come in longer lengths. 

3.9.5 Test Procedures 

The test procedure for the CPT consists of hydraulically pushing the cone at a rate of 0.8 in/s 

(20 mm/s) in accordance with ASTM D 5778 by using either a standard drill rig or a 

specialized cone truck as the reaction mass (see Figure 3-29). The advance of the probe 

requires the successive addition of rods at approximately 3 or 5 ft (1 or 1.5 m) intervals. 

Readings of tip resistance (qt), sleeve friction (fs), inclination (i), and pore water pressure 

(um) are taken at least every 2 in (50 mm) (i.e., at approximately 2.5-sec intervals). For the 

seismic cone test, shear wave arrival times (ts) are typically recorded at rod breaks 

corresponding to 3 or 5 ft (1 or 1.5 m) intervals. 

Figure 3-29. Cone penetration testing from cone truck. 

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Soils and Foundations – Volume I 3 - 71 December 2006 
3.9.6 CPT Profiles 

The results of the individual channels of a piezocone penetration test are plotted with depth, 

as illustrated in a typical plot shown in Figure 3-30. Since soil samples are not obtained with 

the CPT, an indirect assessment of Soil Behavioral Type (SBT) is inferred by an examination 

of the readings. The numbers can be processed for use in empirical chart classification 

systems, or the raw readings can be easily interpreted for soil strata changes. A simplified 

soil classification chart for a standard electric friction cone is presented in Figure 3-31. The 

sleeve friction, often expressed in terms of a friction ratio Rf = fs/qt, also is a general 

indicator of soil type. For example, in sands, usually 0.5% < Rf < 1.5 %; and in clays, 

normally 3% < Rf < 10%. In the lower half of the Figure 3-31, the center column shows an 

approximate relationship between the SPT N-value and the cone tip resistance, qc. The SPT 

N-value obtained from this relationship should be considered to be equivalent to N160. 

3.9.7 CPT Profile Interpretation 

The CPT sounding shown in Figure 3-30 was taken in the immediate vicinity of the boring 

recorded in the boring log shown in Figures 3-25. These two logs permit an interesting 

comparison to be made of the SPT and CPT procedures. In the sounding shown in Figure 3­

30, a clayey and sandy stratum (clay, clayey silt, silt, silty sand and sandy silt) occurs from 

the ground surface to a depth of 10 ft (3 m). These strata are underlain by a thick layer of 

sand and sandy silt to depth of approximately 20 ft (6 m), which in turn is underlain by a clay 

layer extending down to a depth of approximately 45 ft (14 m). Finally, a dense gravelly 

sand layer is encountered, which the cone penetrometer could not penetrate. The SPT tests 

could however be performed in this dense layer since it was possible to drill into this layer. 

Figure 3-30 is a good example that demonstrates the advantage of continuous sounding 

compared to the samples obtained at discrete intervals using SPT procedures. For example, 

depending on the sampling interval in the SPT test, the silt layer within the clay layer may 

not have been undetected. Even if it had been detected, it would not have been possible to 

estimate its thickness accurately between the locations of the SPT samples because the SPT 

samples are commonly retrieved at 5 ft (1.5 m) intervals. The implications of this 

shortcoming can be significant in design. For example, silt consolidates faster than clay. If 

the designer is not aware of the silt layer, then he/she might design a wick drain surcharge 

system that, based on a clay layer 25 ft (7.5 m) thick, will take longer to consolidate than 

what might actually be the case. In fact, the CPT profile in Figure 3-30 shows that only the 

15 ft (4.5 m) portion of the clay layer below the silt layer has excess pore pressures, which 

suggests that it will be the primary source of consolidation settlements. 

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Figure 3-30. Piezo-cone results for Apple Freeway Bridge. 

FHWA NHI-06-088 3 Soils and Foundations – Volume I 3 - 73 
It is important for the reader to understand that CPT procedures do not allow retrieval 

of physical samples that can be tested in the laboratory to characterize various 

phenomena such as consolidation and shear strength. Thus, it is most beneficial to use 

the CPT with another method, such as the boring technique used in the SPT, that 

allows the retrieval of physical samples for laboratory testing. Performing the CPT 

before sampling in borings will permit identification of the specific depths where 

disturbed and undisturbed physical samples should be obtained for laboratory tests. 

Figure 3-31. A commonly used simplified soil classification chart for standard electronic 

friction cone (after Robertson, et al., 1986). 

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Soils and Foundations – Volume I 3 - 74 December 2006 
3.10 DILATOMETER TEST (DMT) 

The dilatometer is an in-situ testing device that was developed in Italy in the early 1970s and 

first introduced in the U.S. in 1979. Like the cone penetrometer, the dilatometer is generally 

hydraulically pushed into the ground although it may also be driven. When the dilatometer 

can be pushed into the ground with tests conducted at 8 in (200 mm) increments, 100 to 130 

ft (30 to 40 m) of soundings may be completed in a day. The primary utilization of the 

dilatometer test (DMT) in pile foundation design is the delineation of subsurface stratigraphy 

and interpreted soil properties. However, it would appear that the CPT/CPTu is generally 

better suited to this task than the DMT. The DMT may be a potentially useful test for the 

design of piles subjected to lateral loads. Design methods in this area show promise, but are 

still in the development stage. For design of axially loaded piles, the DMT has limited direct 

value. A picture of the DMT equipment is presented in Figure 3-32. 

Figure 3-32. Dilatometer test equipment and procedure (FHWA 2002b). 

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3.11 PRESSUREMETER TEST (PMT) 

The pressuremeter is an in-situ device used to evaluate soil and rock properties. The 

pressuremeter has been used in Europe for many years and was introduced into the U.S. in 

the mid 1970s. The pressuremeter imparts lateral pressures to the soil, and the soil shear 

strength and compressibility are determined by interpretation of a pressure-volume 

relationship. The pressuremeter test (PMT) allows a determination of the load-deformation 

characteristics of soil in axisymmetric conditions. Deposits such as soft clays, fissured clays, 

sands, gravels and soft rock can be tested with pressuremeters. A pressuremeter test 

produces information on the elastic modulus of the soil as well as the at rest horizontal earth 

pressure, the creep pressure, and the soil limit pressure. A schematic of the pressuremeter 

test is presented in Figure 3-33. 

Low er P robe 

Into Pre-Bored Hole 

and Expan d w ith 

Pre ssuriz ed W ater 

Pre ssure meter 

Pro be 

At-rest Pressure 

Cre ep Pressu re 

Lim it Pressure 

Elastic Mo dulus 

Drill Rod 

Preb ored Hole 

Pressuremeter 

Test (PMT)

 ASTM D 4719 

Sc rew Pump: 

1. Each Full Rota tion of 

Piston Cylinder Forces 

an In cremental Volume of 

Water (or Gas or Oil) 

Into the PMT P ro be. 

2. Me asure Corresponding 

Pres sure at each increment. 

Plot Pre ssure versus 

Vo lume Change )V (or

alte rn atively, Volumetric

Strain or Cavity Strain) to

Find Pressure me ter Parameters: 

Tubing 

Temporary

C asin g 

Rubber Memb ra ne o f Probe 

Expands as a right cylinder. 

Evaluated per Cylindric al 

Cavity Exp ansion Theory. 

Gage 

Figure 3-33. Pressuremeter test schematic (FHWA, 2002b). 

The utilization of test results is based upon semi-empirical correlations from a large number 

of tests and observations on actual structures. For piles subjected to lateral loads, the 

pressuremeter test is a useful design tool and can be used for determination of p-y curves. 

For design of vertically loaded piles, the pressuremeter test has limited value. Pile design 

procedures using pressuremeter data have been developed and may be found in FHWA 

(1989a). Details on test procedures may be found in ASTM D 4719. 

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Soils and Foundations – Volume I 3 - 76 December 2006 
3.12 VANE SHEAR TEST (VST) 

The vane shear test is an in-situ test for determining the undrained shear strength of soft to 

medium clays. Figure 3-34 is a schematic drawing of the essential components and test 

procedure. The test consists of forcing a four-bladed vane into undisturbed soil and rotating 

it until the soil shears. Two shear strengths are usually recorded, the peak shearing strength 

and the remolded shearing strength. These measurements are used to determine the 

sensitivity of clay, which is defined as the ratio of the peak undrained shearing strength to 

the remolded undrained shearing strength. Sensitivity, St, allows analysis of the soil 

resistance to be overcome during pile driving in clays which is useful for pile driveability 

analyses. It is necessary to measure skin friction along the steel connector rods which must 

be subtracted to determine the actual shear strength. The VST generally provides the most 

accurate undrained shear strength values for clays with undrained shear strengths less than 1 

ksf (50 kPa). The test procedure has been standardized in AASHTO T 223-74 and ASTM D 

2573. 

It should be noted that the sensitivity of a clay determined from a vane shear test provides 

insight into the set-up potential of the clay deposit. However, the sensitivity value is a 

qualitative and not a quantitative indicator of soil set-up. Classification of clayey soils based 

on sensitivity values is presented in Table 3-12. 

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Torquemete r Van e Push in Vane
 Rods
 at Bottom of 

Borehole 

Lower Vane
 

to Bottom of
 

Prebored Hole
 

Four -Bladed
 

Vane Shear
 

Device:
 

 bla de

 height

4 borehole
 

diameter
 

borehole diamet ers 

 blade width 

 blade thickne ss 

1. Ins ertion of Va ne 2. Within 1 minute, rotate 3. Perform an 4. Measure residual 

vane at 6 deg./minute; additional 8 to torque for 

 Measure peak torque. 10 revolutions remolded case

Figure 3-34. Vane shear test equipment and procedure (after FHWA, 2002b). 
Table 3-12 

Classification of Sensitivity Values (Mitchell, 1976)
 

Classification Sensitivity, St 

Insensitive ~ 1.0 

Slightly sensitive clays 1 – 2 

Medium sensitive clays 2 – 4 

Very sensitive clays 4 – 8 

Slightly quick clays 8 – 16 

Medium quick clays 16 – 32 

Very quick clays 32 – 64 

Extra quick clays > 64 

3.13 GROUNDWATER MEASUREMENTS 

Observations of the groundwater level and pore water pressure are an important part of all 

geotechnical explorations. The identification of groundwater conditions should receive the 

same level of care given to soil descriptions and samples. Measurements of water entry 

during drilling and measurements of the groundwater level at least once following drilling 

should be considered a minimum effort to obtain water level data, unless alternate methods, 

such as installation of observation wells, are defined by the geotechnical specialist. Detailed 

information regarding groundwater observations can be obtained from ASTM D 4750 and 

ASTM D 5092. 

3.13.1 Information on Existing Wells 

Many states require the drillers of water wells to file logs of the wells they have drilled. 

These are good sources of information of the materials encountered and water levels 

recorded during well installation. The well owners, both public and private, may have 

records of the water levels after installation, which may provide extensive information on 

fluctuations of the water level. This information may be available at state agencies 

regulating the drilling and installation of water wells, such as the Department of 

Transportation, the Department of Natural Resources, State Geologist, Hydrology 

Department, Department of Environmental Quality, and Division of Water Resources. 

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Soils and Foundations – Volume I 3 - 78 December 2006 
3.13.2 Open Borings 

The water level in open borings should be measured after any prolonged interruption in 

drilling, at the completion of each boring, and at least 12 hours (preferably 24 hours) after 

completion of drilling. Additional water level measurements should be made at the 

completion of the field exploration and at other times designated by the engineer. The date 

and time of each observation should be recorded. 

If the borehole has caved, the depth to the collapsed region should be recorded on the boring 

record as the collapse may have been caused by groundwater conditions. The elevations of 

the caved depths of certain borings may be consistent with groundwater table elevations at 

the site. This consistency may become apparent once the subsurface profile is constructed 

(see Chapters 4 and 11). 

Drilling mud obscures observations of the groundwater level owing to filter cake action and 

the greater specific gravity of the drilling mud compared to that of the water. If drilling 

fluids are used to advance borings, the drill crew should be instructed to bail the hole prior to 

making groundwater observations. 

3.13.3 Observation Wells 

The observation well, also referred to as a piezometer, is the fundamental means for 

measuring water head in an aquifer and for evaluating the performance of dewatering 

systems. In theory, a “piezometer” measures the water pressure in a confined aquifer or at a 

specific horizon of the geologic profile, while an “observation well” measures the level of a 

water table in an aquifer (Powers, 1992). In practice, however, the two terms are often used 

interchangeably to describe any device for determining static water head. 

The term “observation well” is applied to any well or drilled hole used for the purpose of 

long-term studies of groundwater levels and pressures. Existing wells and bore holes in 

which casing is left in place are often used to observe groundwater levels. These, however, 

are not considered to be as satisfactory as wells constructed specifically for the purpose of 

measuring groundwater conditions. The latter may consist of a standpipe installed in a 

previously drilled exploratory hole or a hole drilled solely for use as an observation well. 

Details of typical observation well installations are shown in Figure 3-35. The simplest type 

of observation well is formed by a small-diameter polyvinyl chloride (PVC) pipe set in an 

open hole. The bottom of the pipe is slotted and capped, and the annular space around the 

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slotted pipe is backfilled with clean sand. The area above the sand is sealed with bentonite, 

and the remaining annulus is filled with grout, concrete, or soil cuttings. A surface seal, 

which is sloped away from the pipe, is commonly formed with concrete in order to prevent 

the entrance of surface water. The top of the pipe should also be capped to prevent the 

entrance of foreign material; there should be a small vent hole in the top of the removable 

cap. In some localities, regulatory agencies may stipulate the manner for installation and 

closure of observation wells. 

Figure 3-35. Representative details of observation well installations. (a) Drilled-in

place stand-pipe piezometer, (b) Driven well point (FHWA, 1997). 

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Driven or pushed-in well points are another common type of observation well for use in 

granular soil formations and very soft clay (Figure 3-35b). The well is formed by a stainless 

steel or brass well point threaded to a galvanized steel pipe. In granular soils, an open boring 

or rotary wash boring is advanced to a point several inches above the measurement depth and 

the well point is driven to the desired depth. A seal is commonly required in the boring 

above the well point with a surface seal at the ground surface. Note that observation wells 

may require development (see ASTM D 5092) to minimize the effects of installation, drilling 

fluids, etc. Minimum pipe diameters should allow introduction of a bailer or other pumping 

apparatus to remove fine-grained materials in the well to improve the response time. 

Local jurisdictions may impose specific requirements on “permanent” observation wells, 

including closure and reporting of the location and construction that must be considered in 

the planning and installation. Licensed drillers and special fees may be required. 

Piezometers are available in a number of designs. Commonly used piezometers are of the 

pneumatic and the vibrating wire type. Interested readers are directed to Dunnicliff (1988) 

and FHWA (1997) for a detailed discussion of the various types of piezometers. 

3.13.4 Water Level Measurements 

A number of devices have been developed for sensing or measuring the water level in 

observation wells. Following is a brief presentation of three methods that are commonly 

used to measure the depth to groundwater. In general, common practice is to measure the 

depth to the water surface by using the top of the casing as a reference, with the reference 

point at a common orientation (often north) marked or notched on the well casing. 

3.13.4.1 Chalked Tape 

In this method a short section at the lower end of a metal tape is chalked. The tape with a 

weight attached to its end is then lowered until the chalked section has passed slightly below 

the water surface. The depth to the water is determined by subtracting the depth of 

penetration of the line into water, as measured by the water line in the chalked section, from 

the total depth from the top of casing. This is probably the most accurate method, and the 

accuracy is useful in pump tests where very small drawdowns are significant. The method is 

cumbersome, however, when a series of rapid readings is taken since the tape must be fully 

removed each time. An enameled tape is not suitable unless it is roughened with sandpaper 

so it will accept chalk. The weight on the end of the tape should be small in volume so it 

does not displace enough water to create an error in the water level. 

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3.13.4.2 Tape with a Float 

In this method, a tape with a flat-bottomed float attached to its end is lowered until the float 

hits the water surface and the tape goes slack. The tape is then lifted until the float is felt to 

touch the water surface and the tape is just taut; the depth to the water surface is then 

measured. With practice this method can give rough measurements, but its accuracy is poor. 

A refinement is to mount a heavy whistle, open at the bottom, on a tape. When it sinks in the 

water, the whistle will give an audible beep as the air within it is displaced. 

3.13.4.3 Electric Water-Level Indicator 

This battery operated indicator consists of a weighted electric probe attached to the lower 

end of a length of electrical cable that is marked at intervals to indicate the depth. When the 

probe reaches the water a circuit is completed. This condition is registered by a meter 

mounted on the cable reel. Various manufacturers produce the instrument, utilizing a neon 

lamp, a horn, or an ammeter as the signaling device. The electric indicator has the advantage 

that it may be used in extremely small holes. 

The instrument should be ruggedly built, since some degree of rough handling can be 

expected. The distance markings must be securely fastened to the cable. Some models are 

available in which the cable itself is manufactured as a measuring tape. The sensing probe 

should be shielded to prevent shorting out against metal risers. When the water is highly 

conductive, erratic readings can develop in the moist air above the actual water level. 

Sometimes careful attention to the intensity of the neon lamp or the pitch of the horn will 

enable the reader to distinguish the true level. A sensitivity adjustment on the instrument can 

be useful. If oil or iron sludge has accumulated in the observation well, the electric probe 

will give unreliable readings. 

3.13.4.4 Data Loggers 

When timed and frequent water level measurements are required, as for a pump test or slug 

test, data loggers are useful. Data loggers are in the form of an electric transducer near the 

bottom of the well that senses changes in water level as changes in pressure. A data 

acquisition system is used to acquire and store the readings. A data logger can eliminate the 

need for onsite technicians on night shifts during extended field permeability testing. A 

further significant saving is in the technician’s time back in the office. The preferred models 

of the data logger not only record the water level readings but permit the data to be 

downloaded into a personal computer and, with appropriate software, to be quickly reduced 

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and plotted. These devices are also extremely useful for cases where measurement of 

artesian pressures is required or where data for tidal corrections during field permeability 

tests are necessary. 

3.14 GUIDELINES FOR MINIMUM SUBSURFACE EXPLORATION 

In regard to the scope of the subsurface exploration program for a structure or a geotechnical 

feature, one must consider the small cost of the borings in relation to the foundation costs. The 

knowledge gained from a thorough subsurface exploration program will allow the geotechnical 

specialist to evaluate various candidate foundation schemes and provide recommendations for 

those that can be built most efficiently and economically on the project site. Without an 

adequate subsurface exploration program, the result is generally an extremely conservative 

foundation recommendation. 

Planning a subsurface exploration program should include determining the location and depth 

of borings, test pits, or other procedures to be used and establishing the methods of soil 

sampling and testing to be employed. Usually, the extent of the work is estimated based on 

available geological studies, earlier explorations, or records of existing structures. The number 

of borings and their locations in a site area will depend on the proposed structure, design 

parameters, access issues, geologic constraints, and expected stratigraphy. 

Although no rigid rules apply universally to geotechnical explorations, certain general 

principles are usually followed in practice. Recommended guidelines for the minimum 

number of exploration points and their spacing are provided in Table 3-13. This table was 

developed based on a number of FHWA documents. This table should be used only as a first 

step in estimating the minimum number of borings for a particular design, as actual boring 

spacings will be dependent upon the project type and geologic environment. In all cases, it is 

recommended that the depth of the exploration should be such that the depth of significant 

influence (DOSI) is explored. For a given configuration of loading, the DOSI may exceed 

the minimum guidelines in Table 3-13 in which case the depth of exploration should be 

increased accordingly. Some other general guidance in addition to that in Table 3-13 is as 

follows: 

•	 In areas underlain by heterogeneous soil deposits and/or rock formations, it will 

probably be necessary to exceed the minimum guidelines presented in Table 3-13 to 

capture variations in soil and/or rock type and to assess consistency across the site. 

 

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Table 3-13 

Guidelines for minimum number of exploration points and depth of exploration (modified after FH

FHWA NHI-06-088 3 Soils and Foundations – Volume I 3 - 84 

Application 

Minimum Number of Exploration Points 

and Location of Exploration Points Minimum Depth of Explo

Retaining walls (1) 

(2) (3) (4) 

A minimum of one exploration point for each retaining 

wall. 

For retaining walls more than 100 ft (30 m) in length, 

exploration points spaced every 100 to 200 ft (30 to 60 m) with locations alternating from in front of the wall to behind the wall. 

For anchored walls, additional exploration points in the 

anchorage zone spaced at 100 to 200 ft (30 to 60 m). 

For soil-nail walls, additional exploration points at a 

distance of 1.0 to 1.5 times the height of the wall behind the wall spaced at 100 to 200 ft (30 to 60 m). 

(1) Investigate to a depth below bottom of w

the wall height or a minimum of 10 ft (3 

(2) Exploration depth should be great enoug

highly compressible soils (e.g. peat, orga

soils) into competent material of suitable stiff to hard cohesive soil, compact dense bedrock). 

Embankment Foundations 

(1) A minimum of one exploration point every 200 ft (60 m) 

(erratic conditions) to 400 ft (120 m) (uniform conditions) of embankment length along the centerline of the embankment. 

(2) At critical locations, (e.g., maximum embankment heights, 

maximum depths of soft strata) a minimum of three exploration points in the transverse direction to define the existing subsurface conditions for stability analyses. 

(3) For bridge approach embankments, at least one exploration 

point at abutment locations. 

(1) Exploration depth should be, at a minimu

embankment height unless a hard stratum this depth. 

(2) If soft strata are encountered extending t

twice the embankment height, the explorgreat enough to fully penetrate the soft stmaterial (e.g., stiff to hard cohesive soil, 

cohesionless soil, or bedrock). 

Cut Slopes (1) A minimum of one exploration point every 200 ft (60 m) 

(erratic conditions) to 400 ft (120 m) (uniform conditions) of slope length. 

(2) At critical locations (e.g., maximum cut depths, maximum 

depths of soft strata) a minimum of three exploration points in the transverse direction to define the existing subsurface conditions for stability analyses. 

(3) For cut slopes in rock, perform geologic mapping along the 

length of the cut slope. 

(1) Exploration depth should be, at a minimu

the minimum elevation of the cut unless encountered below the minimum elevatio

(2) Exploration depth should be great enoug

through soft strata into competent matericohesive soil, compact to dense cohesion

(3) In locations where the base of cut is belo

increase depth of exploration as needed tunderlying pervious strata. 
Table 3-13 (Continued) 

Guidelines for minimum number of exploration points and depth of exploration (after FHWA

Application Minimum Number of Exploration Points and 

Location of Exploration Points 

Minimum Depth of Explorati

Shallow Foundations 

(1) For substructure (e.g., piers or abutments) widths 

less than or equal to 100 ft (30 m), a minimum of one exploration point per substructure. 

(2) For substructure widths greater than 100 ft (30 

m), a minimum of two exploration points per substructure. 

(3) Additional exploration points should be provided 

if erratic subsurface conditions or sloping rock surfaces are encountered. 

Depth of exploration should be: (1) great enough to fully penetrate unsuitable founda

organic silt, soft fine grained soils) into competenbearing capacity (e.g. stiff to hard cohesive soil, c

cohesionless soil or bedrock); and 

(2) at least to a depth where stress increase due to 

is less than 10% of the applied stress at the bas

(3) in terms of the width of the footing: at least 2 tim

and 4 times for strip footing (interpolate for interm

(4) if bedrock is encountered before the depth require

achieved, exploration depth should be great enouminimum of 10 ft (3 m) into the bedrock, but rock sufficient to characterize compressibility of infill horizontal to horizontal discontinuities. 

Deep Foundations (1) For substructure (e.g., bridge piers or abutments) 

widths less than or equal to 100 ft (30 m), a minimum of one exploration point per substructure. 

(2) For substructure widths greater than 100 ft (30 

m), a minimum of two exploration points per substructure. 

(3) Additional exploration points should be provided 

if erratic subsurface conditions are encountered. 

(4) Due to large expense associated with construction 

of rock-socketed shafts, conditions should be confirmed at each shaft location. 

(1) In soil, depth of exploration should extend below 

shaft tip elevation a minimum of 20 ft (6 m), or a the maximum pile group dimension, whichever is should extend through unsuitable strata such as uhighly organic materials, soft fine-grained soils, asoils to reach hard or dense materials. 

(2) For piles bearing on rock, a minimum of 10 ft (3 

obtained at each exploration point location to ver

not terminated on a boulder. 

(3) For shafts supported on or extending into rock, a 

of rock core, or a length of rock core equal to at ldiameter for isolated shafts or two times the maxidimension, whichever is greater, shall be extendeshaft tip elevation to determine the physical charathe zone of foundation influence. 

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•	 In the case of embankments, the guidance provided in Table 3-13 for the depth of 

exploration is the minimum recommended for any transportation facility, i.e., 2 times 

the height of the embankment. The minimum guidance may not be the same as the 

DOSI, which is a function of the geometry (crest width, height, configuration of side 

slopes, and the base width) of the embankment. For the same height of embankment, 

the DOSI increases as the base width of an embankment increases, and may vary 

from 4 to 6 times the height of the embankment. An example of this is shown in 

Figures 2-8 in Chapter 2, which shows that the DOSI can extend to depths much 

deeper than 2 times the height of the embankment. This may be particularly critical 

in cases where there are soft soils in the subsurface. Thus, for such situations, the 

geotechnical specialist should use tools such as the FoSSA program or published 

elastic solutions (Poulos and Davis, 1974) to determine the depth of exploration. 

•	 For situations where large-diameter rock-socketed shafts will be used or where drilled 

shafts are being installed in karstic formations, it may be necessary to advance a 

boring at the location of each shaft. 

•	 In a laterally homogeneous area, drilling or advancing a large number of borings may 

be redundant, since each sample tested would exhibit similar strength and 

compressibility properties. 

•	 In all cases, it is necessary to understand how the design and construction of the 

geotechnical feature will affect the soil and/or rock mass in order to optimize the 

exploration. 

During exploration, each exploration point (e.g., drill hole or CPT sounding) should be 

designated by a unique identification number to prevent duplication during subsequent 

explorations. For example, it is not unusual to find projects where borehole numbering was 

done by only single numbers so that the same designations were used during one or more 

subsequent explorations. A suggested method to avoid duplication is to designate that all 

bridge holes begin with the letter "B,” followed by the initials of the highway or river being 

crossed and finally a sequential number. For example, the first boring for a structure on 

Apple Freeway would be designated DH-BAF-1, where the DH means a “drill hole” where 

SPTs were performed as opposed to CPT-BAF-1, where the CPT refers to the first CPT 

sounding performed for a bridge structure on Apple Freeway. 

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The geotechnical specialist should plot the proposed boring locations on a site topographic 

map prior to initiation of drilling. Notes taken during the site visit should be reviewed before 

boring locations are selected so that site access restrictions can be considered. Boring 

locations should never be selected arbitrarily or randomly. Alternate boring locations should 

be considered and a contingency plan should be developed in case a boring needs to be 

relocated due to access restrictions or unexpected geologic conditions (e.g., locate a 

replacement boring within a maximum of 15 ft (4.5 m) from the location of a boring that 

could not be drilled at a particular location). Field personnel unfamiliar with the objectives 

and rationale behind the planning of the site exploration should maintain contact with the 

person in responsible charge of the field exploration during field activities and discuss issues 

such as the relocation of a boring with that person. Arbitrary or random boring selection will 

increase the chances of boring relocation, confusion, and wasted time in the field. Final 

boring locations should be surveyed and recorded as part of the permanent project record. 

Elevations and northing and easting should be provided for each boring. 

3.14.1 Recommendations for Sampling Depth Intervals in Soils 

It is difficult to establish a prescriptive drilling, sampling, and testing protocol that is 

applicable to all sites. To be most effective the geotechnical specialist should: 

(1) apply conventional guidelines with project-specific requirements and constraints; 

(2) recognize the advantages and limitations of the available sampling devices and in-situ 

testing methods. 

Some general recommendations for minimum sampling depth intervals are as follows: 

•	 For preliminary screening, disturbed samples might be taken continuously in the 

upper 10 ft (3 m), at 5 ft (1.5 m) intervals up to 100 ft (30 m), and possibly every 

10 ft (3 m) at depths greater than 100 ft (30 m). 

•	 Disturbed samples should be taken at every abrupt change in stratum as indicated 

by a noticeable change in the drilling pressure. 

•	 Where footings are to be placed on natural soil, continuous spoon samples are 

recommended for a depth equal to 15 ft (4.5 m) or 1.5 times the width of the 

footing, whichever is greater, as measured below the anticipated footing base 

elevation. 

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•	 For characterization and assessment of design properties in fine-grained soils, a 

minimum of one undisturbed sample should be taken in each stratum, with 

additional samples taken at 10- to 20-ft (3 to 6 m) intervals with depth. 

•	 Undisturbed Shelby tube samples should be obtained at 5 ft (1.5 m) intervals in at 

least one boring in cohesive soils. For cohesive deposits greater than 30 ft (9 m) 

in depth, the tube sample interval can be increased to 10 ft (3 m). Undisturbed 

samples may not need to be taken in each boring if the deposit is relatively 

homogeneous within closely spaced borings. 

These minimum guidelines and intervals may need to be increased depending upon the 

project requirements and site geologic conditions. The sampling interval may need to be 

increased when soil/rock conditions change frequently with depth; however, these changes 

need to be considered in the context of the design. Therefore, ongoing communication 

between field personnel and the office/design engineer is absolutely essential. Once the site 

stratigraphy has been established, it may not be necessary to sample every time there is a 

change in stratigraphy if the changes have no impact on design. For example, it may not be 

necessary to sample alternating layers of coarse-grained deposits where settlement is of 

concern, and for designs concerned with bearing capacity. Similarly, although samples 

below the anticipated extent of the area influenced by the load may be reduced, samples 

should be obtained in case the type of foundation changes between preliminary and final 

design. 

The sampling interval will vary between individual projects and regional geologies. If soils 

are anticipated to be difficult to sample or trim in the laboratory due to defects (e.g., bent 

tubes, improper handling, etc.), the frequency of sample collection should be greater than 

average to offset the number of samples that may be unusable in the laboratory for 

performance property evaluation (e.g., shear strength). When borings are widely spaced, it 

may be appropriate to retrieve undisturbed samples in each boring. For closely spaced 

borings or in deposits of lateral uniformity, undisturbed samples may be needed only in 

select borings. If a thin clay seam is encountered during drilling and not sampled, the boring 

may need to be offset and re-drilled to obtain a sample. 

It is often quite helpful to combine in-situ soundings with conventional disturbed/undisturbed 

sampling. For example, by performing CPT or CPTu soundings prior to conventional 

drilling and sampling, it may be possible to target representative and/or critical areas where 

samples can be obtained later. This concept was illustrated previously by the discussion in 

Section 3.9.7 and Figure 3-30. The use of precursor soundings may reduce some of the 

potential drilling redundancy in heterogeneous environments. Geophysical methods can also 

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Soils and Foundations – Volume I 3 - 88 December 2006 
be used to provide useful information on conditions between and even beyond boring 

locations. 

3.14.2 Recommendations for Sampling Depth Intervals in Rocks 

For explorations for slopes and foundations within rock, it is important to consider structural 

geology in addition to the information obtained as part of a rock-coring program. For 

example, the orientation and characteristics of a clay-filled discontinuity are critical since 

they can be used to judge whether a rock slope will be stable or unstable or whether a 

structural foundation will undergo minor or significant settlement. A detailed structural 

geologic assessment may provide enough information to limit the scope of a rock-coring 

program significantly or even preclude such a program. For example, drilling and coring 

may not be required where applied loads are significantly less than the bearing capacity of 

the rock, where there is no possibility of sliding instability in a rock slope, or where there are 

extensive rock outcrops from which information can be obtained to establish the subsurface 

conditions confidently for design and constructability assessments (Wyllie, 1999). 

3.14.3 Recommendations for Water Level Monitoring in Borings 

The water level in each boring should be observed and the depth below the top of hole 

recorded on the drill log with the date and time of the reading for each of the following 

situations: 

a) Water seepage or artesian pressure encountered during drilling. Artesian pressure 

may be measured by extending drill casing above the ground until flow stops. 

Report the pressure as the number of ft (m) of head above ground. 

b) Water level at the end of each day and at completion of boring. 

c) Water level 24 hours (minimum) after hole completion. Long term readings may 

require installation of a perforated plastic tube before abandoning the hole. 

A false indication of water level may be obtained when water is used in drilling and adequate 

time is not permitted after the boring is completed for the water level to stabilize. In low 

permeability soils, such as clays, more than one week may be required to obtain accurate 

readings. Proper safety precautions should be taken if a hole is allowed to remain open for 

such an extended period of time. 

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3.15 GEOPHYSICAL TESTS 

As indicated in Section 3.3, geophysical testing can be used as part of the initial site 

exploration to provide supplementary information to data collected by other means (i.e., 

borings, test pits, geologic surveys, etc.). Geophysical testing can be used for establishing 

stratification of subsurface materials, the profile of the top of bedrock, the depth to 

groundwater, the boundaries of various types of soil deposits, the rippability of hard soil and 

rock, and the presence and depth of voids, buried pipes, and existing foundations. Data from 

geophysical testing should always be correlated with information from the direct methods of 

exploration discussed previously. 

3.15.1 Types of Geophysical Tests 

There are a number of different types of geophysical in-situ tests that can be used to obtain 

stratigraphic information from which engineering properties can be estimated. Table 3-14 

provides a summary of the various geophysical methods that are currently available in U.S. 

practice. Further information on the procedures used for these methods is provided in 

FHWA (2003). Additional general discussion regarding the major test methods listed in 

Table 3-14 is presented below, with particular emphasis on potential applications to highway 

engineering. 

1.	 Seismic Methods: These methods are becoming increasingly popular for 

geotechnical engineering practice because they have the potential to provide data 

regarding the compression and shear wave velocities of the subsurface materials. The 

shear wave velocity is directly related to small-strain material stiffness, which, in 

turn, is often correlated to compressive strength and soil/rock type. These techniques 

are often used for assessing the vertical stiffness profile in a soil deposit and for 

assessing the location at depth of the interface between soil and rock. Seismic 

refraction method involves measurement of time of arrival of the initial ground 

motion generated by the energy source while the seismic reflection method involves 

measurement of the energy arrival after the initial ground motion. 

2.	 Electrical Methods: These methods are usually used to locate voids or locally 

distinct materials. With regards to highway applications, these procedures may be 

used to assess the potential for karst activity along a planned transportation corridor, 

or for locating large underground voids and/or specific underground anomalies such 

as storage drums and/or tanks. Electrical methods provide qualitative information 

only and are usually part of a two- or three-phased exploration program. 

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Table 3-14 

Geophysical testing techniques (modified after FHWA, 2002a) 

FHWA NHI-06-088 3 Soils and Foundations – Volume I 3 - 91 

Method Basic Field Procedures Applications Limitations 

SEISMIC METHODS 

Seismic Impact load is applied to the ground surface. • depth to bedrock • does not work 

Refraction Seismic energy refracts off soil/rock layer 

interfaces and the time of arrival is recorded 

on the ground surface using several dozen geophones positioned along a line or performing repeated events using a single geophone. 

••

depth to water table thickness and relative stiffness soil/rock layers •

with depth or ilayer works best whdiscontinuity i

SpectralAnalysis-ofSurface-Waves 

(SASW) 

Impact load is applied to the ground surface. 

Surface waves propagate along ground surface 

and are recorded on the ground surface with 

two geophones positioned along a line. 

•

•

•

•

depth to bedrock 

measurement of shear wave velocity 

thickness and stiffness of surface 

pavement layer qualitative indicator of cracking in pavement 

•

•

•

resolution decr

increasing dep

accurate interp

significant amointerpretation ioverlies a soft properties are d

ELECTRICAL METHODS 

DC Resistivity DC current is applied to the ground by 

electrodes. Voltages are measured at different 

points on the ground surface with other electrodes positioned along a line. 

•••••

depth to water table inorganic groundwater contamination groundwater salinity soil layer thickness delineation of certain vertical features (e.g., sinkholes, contamination plumes, waste trenches) 

•••

slow; must insthe ground resolution decrincreasing depresolution is diheterogeneous 

Electromagnetics Electrical coils are held over the ground. 

Current passing through the coils induces a magnetic field in the ground, which is measured with receiver coils. 

••••

groundwater salinity inorganic groundwater contamination detection of buried metal objects delineation of certain vertical features (e.g., sinkholes, contamination plumes, waste trenches) 

••

extra effort is rdepth of target resolution decrincreasing dep

Ground Penetrating Radar (GPR) 

Electromagnetic energy is pulsed into the ground. This energy reflects off boundaries between different soil layers and is measured at the ground surface. 

••••

depth to water table identification of buried objects thickness of pavement layers void detection 

••

not effective bclay depth of penetr30 ft (10 m) 
Table 3-14 

Geophysical testing techniques (modified after FHWA, 2002a) 

FHWA NHI-06-088 3 Soils and Foundations – Volume I 3 - 92 

Method Basic Field Procedures Applications Limitations 

GRAVITY AND MAGNETIC METHODS Gravity The Earth’s gravitational field is measured 

at the ground surface. 

••

identification of large subsurface voids identification of large objects 

possessing unusually high or low densities 

••

results are nonone subsurface same result) primarily, largtool; applicatioengineering ar

Magnetics The Earth’s magnetic field is measured at 

the ground surface. 

••

identification of ferrous materials identification of soil/rock containing 

large amounts of magnetic minerals 

••

results are nonone subsurface same results) primarily a largtool; applicatioengineering ar

NEAR-SURFACE NUCLEAR METHODS Neutron Moisture Instrument is placed on the ground surface • estimate of water content in • limited explora

Content and neutrons are emitted into the ground. 

Energy of returning neutrons is related to the moisture content in the ground (hydrogen atoms decrease the energy of the neutrons detected at the sensor). 

••

compacted soil estimate of asphalt content in asphalt concrete can be quantitative if properly calibrated to site conditions 

••

inches) possible health operators not pwill detect hydin non-water b

Gamma Density Instrument is placed on the ground surface 

and gamma radiation is emitted into the ground. Returning gamma energy is a function of material density (denser materials absorb more gamma energy so less is detected at the sensor) 

• estimate of density of soil or asphalt 

concrete 

•••

limited explorafoot); exploration defew inches if gpenetrated possible health operators not p
Table 3-14 (continued) 

Geophysical testing techniques (modified after FHWA, 2002a) 

FHWA NHI-06-088 3 

Soils and Foundations – Volume I 3 - 93 

Method Basic Field Procedures Applications Limitations BOREHOLE METHODS Crosshole/ Energy sources and geophones are placed • measurement of wave velocities for • requires one or 

Downhole in boreholes and/or on ground surface; 

interval travel times are converted into seismic wave velocity as a function of depth in the borehole. 

•••

seismic site response analysis depth to water table correlation of lithologic units with 

surface seismic identification of thin layers at depth 

significant supp

Suspension Field instrument is placed in a fluid-filled • measurement of wave velocities for • requires boreho

Logger borehole and used to measure P- 

(compression) and S-(shear) wave velocities in surrounding soil or rock. 

••

seismic site response analysis correlation of lithologic units with surface seismic identification of thin layers at depth 

•

field equipmen

borehole must b

Electrical Logging Field instrument is placed in a borehole. 

Electrical fields are directly applied or electromagnetically induced into surrounding soil or rock and electrical resistivity is measured. 

•••

estimate of soil/rock permeability or 

porosity identification of inorganic contaminant plumes or saltwater intrusion identification of thin layers at depth 

••••

requires borehofield equipmengenerally cannoborehole may require fluresults may be mud salinity 

Nuclear Logging Field instrument is placed in a borehole. 

Surrounding soil or rock is irradiated with neutrons particles and/or gamma energy. Energy and neutrons returning to the instrument are measured and related to rock density, porosity and pore fluid type. 

••

estimate of soil/rock type, density, 

porosity, and pore fluid density 

identification of thin layers at depth 

••

requires borehofield equipmenpossible health operators are n

Lithology Field instrument is placed in a borehole; • classification of soil or rock type • requires boreho

Logging naturally occurring electrical fields and 

radiation levels are related to soil or rock type. 

• identification of thin layers at depth 

•

•

field equipmenmay require fluresults are depeconditions and/
3.	 Gravity and Magnetic Methods: These methods are similar to electrical methods, 

except that they rely on correlations between the potential gravitational and/or 

magnetic influence of voids and subsurface anomalies and measured differences in 

the earth’s micro-gravitational and/or magnetic fields, rather than on changes in 

electrical fields. These methods provide measurements at specific points unlike 

seismic and electrical methods that provide measurements over large areas. 

4.	 Near-surface nuclear methods: These methods have been used for several years for 

compaction control of fills in the field. Through careful calibration, it is possible to 

assess the moisture content and density of compacted soils reliably. These methods 

have been widely adopted as reliable quantitative methods. 

5. Borehole Methods: Downhole geophysical methods provide reliable indications of 

a wide range of soil properties. For example, downhole/crosshole methods provide 

reliable measures of shear wave velocity. As indicated previously, shear wave 

velocity is directly related to small-strain stiffness and is correlated to strength and 

soil/rock type. Although downhole logging methods have seen little use in highway 

construction, they have been the mainstay for deep geologic characterization in oil 

exploration. The principal advantage of downhole logging is the ability to obtain 

several different geophysical tests/indicators by “stringing” these tools together in a 

deep borehole. 

3.15.2 Advantages and Disadvantages of Geophysical Tests 

As with the other methods of exploration, geophysical testing offers some advantages and 

some disadvantages that should be considered before these techniques are recommended for 

a specific application. These are summarized as follows: 

3.15.2.1 Advantages of Geophysical Tests 

1.	 Many geophysical tests are non-invasive. Therefore such tests offer significant 

benefits in cases where conventional drilling, testing, and sampling are difficult (e.g., 

deposits of gravel, talus deposits, etc.) or where potentially contaminated soils may 

occur in the subsurface. 

2.	 In general, geophysical testing can cover a relatively large geographical area thereby 

providing the opportunity to characterize large areas with relatively few tests. 

Geophysical testing is particularly well-suited to projects that have large longitudinal 

extent such as new highway construction. 

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3.	 Geophysical measurements are used to assess the properties of soil and rock at very 

small strains, typically on the order of 0.001 percent, thereby providing information 

on truly elastic properties. 

4.	 For the purpose of obtaining information on the subsurface, geophysical methods are 

relatively inexpensive considering the large area over which they provide 

information. 

3.15.2.2 Disadvantages of Geophysical Tests 

1.	 Most methods work best for situations in which there is a large difference in the 

property being measured between adjacent subsurface units. In seismic methods, it is 

difficult to develop good stratigraphic profiling if the general stratigraphy consists of 

hard material overlying soft material. 

2.	 Each geophysical method has limitations that may be associated with equipment, 

signal noise, unfavorable site and subsurface conditions, and processing constraints. 

3.	 Results can be non-unique and are generally interpreted qualitatively. Therefore 

useful results can be obtained only through analyses performed by a geotechnical 

specialist experienced with the particular testing method. 

4.	 Specialized and more electronically sophisticated equipment is required as compared 

to the more conventional subsurface exploration tools. 

3.15.3 Examples of Uses of Geophysical Tests 

The following are a few examples where geophysical testing could be used on highway 

projects to compliment conventional exploration. 

1.	 Highly Variable Subsurface Conditions: In several geologic settings, the subsurface 

conditions along a transportation corridor may be expected to be variable. This 

variability could be from underlying karst development above limestone; alluvial 

deposits, including buried terrace gravels, across a wide floodplain; buried boulders 

in a talus slope, etc. For these cases, conventional exploration techniques may be 

very difficult and if “refusal” is encountered at one depth, there is a strong likelihood 

that different materials could underlie the region. Development of a preliminary 

subsurface characterization profile by using geophysical testing could prove 

advantageous in designing future focused explorations. 

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2.	 Regional Studies: Along a transportation corridor it may be necessary to assess the 

depth to (and through) rippable rock or highly cemented caliche. Alternative 

alignments may or may not be possible, but the cost implications may be significant. 

Therefore, it is important to obtain a profile related to rock/soil stiffness. 

Geophysical testing is a logical consideration for this application as a precursor to 

invasive explorations. 

3.	 Settlement Sensitive Structures: The prior two examples related to cases where the 

geophysical testing served as the front-end of a multi-phase project. In the case 

where a settlement-sensitive structure is to be founded on deposits of sands, 

knowledge of the in-situ modulus of the sand deposit is critical. After the 

characteristics of the site are assessed, it may be helpful to quantify the deformation 

modulus by the use of geophysical testing at the specific foundation site. 

These examples demonstrate that geophysical testing can play a potentially important role in 

the subsurface characterization of soils and rocks. As with the other investigative “tools” 

described in this document, the particular selection of the appropriate technology is very 

much a function of the site conditions and the goals of the characterization program. 

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CHAPTER 4.0 

ENGINEERING DESCRIPTION, CLASSIFICATION AND CHARACTERISTICS 

OF SOILS AND ROCKS 

The geotechnical specialist is usually concerned with the design and construction of some 

type of geotechnical feature constructed on or out of a geomaterial. For engineering 

purposes, in the context of this manual, the geomaterial is considered to be primarily rock 

and soil. A geomaterial intermediate between soil and rock is labeled as an intermediate 

geomaterial (IGM). These three classes of geomaterials are described as follows: 

•	 Rock is a relatively hard, naturally formed solid mass consisting of various minerals and 

whose formation is due to any number of physical and chemical processes. The rock 

mass is generally so large and so hard that relatively great effort (e.g., blasting or heavy 

crushing forces) is required to break it down into smaller particles. 

•	 Soil is defined as a conglomeration consisting of a wide range of relatively smaller 

particles derived from a parent rock through mechanical weathering processes that 

include air and/or water abrasion, freeze-thaw cycles, temperature changes, plant and 

animal activity and by chemical weathering processes that include oxidation and 

carbonation. The soil mass may contain air, water, and/or organic materials derived from 

decay of vegetation, etc. The density or consistency of the soil mass can range from very 

dense or hard to loose or very soft. 

•	 Intermediate geomaterials (IGMs) are transition materials between soils and rocks. The 

distinction of IGMs from soils or rocks for geotechnical engineering purposes is made 

purely on the basis of strength of the geomaterials. Discussions and special design 

considerations of IGMs are beyond the scope of this document. 

The following three terms are often used by geotechnical specialists to describe a 

geomaterial: identification, description and classification. For soils, these terms have the 

following meaning: 

•	 Identification is the process of determining which components exist in a particular soil 

sample, i.e., gravel, sand, silt, clay, etc. 

•	 Description is the process of estimating the relative percentage of each component to 

prepare a word picture of the sample (ASTM D 2488). Identification and description are 

accomplished primarily by both a visual examination and the feel of the sample, 

particularly when water is added to the sample. Description is usually performed in the 

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field and may be reevaluated by experienced personnel in the laboratory. 

•	 Classification is the laboratory-based process of grouping soils with similar engineering 

characteristics into categories. For example, the Unified Soil Classification System, 

USCS, (ASTM D 2487), which is the most commonly used system in geotechnical work, 

is based on grain size, gradation, and plasticity. The AASHTO system (M 145), which is 

commonly used for highway projects, groups soils into categories having similar load 

carrying capacity and service characteristics for pavement subgrade design. 

It may be noted from the above definitions that the description of a geomaterial necessarily 

includes its identification. Therefore, as used in this document, the term “description” is 

meant to include “identification.” 

The important distinction between classification and description is that standard AASHTO or 

ASTM laboratory tests must be performed to determine the classification. It is often 

unnecessary to perform the laboratory tests to classify every sample. Instead soil technicians 

are trained to identify and describe soil samples to an accuracy that is acceptable for design 

and construction purposes. ASTM D 2488 is used for guidance in such visual and tactile 

identification and description procedures. These visual/tactile methods provide the basis for 

a preliminary classification of the soil according to the USCS and AASHTO system. 

During progression of a boring, the field personnel should describe only the soils 

encountered. Group symbols associated with classification should not be used in the field. It 

is important to send the soil samples to a laboratory for accurate visual description and 

classification by a laboratory technician experienced in soils work, as this assessment will 

provide the basis for later testing and soil profile development. Classification tests can be 

performed in the laboratory on representative samples to verify the description and assign 

appropriate group symbols based on a soil classification system (e.g., USCS). If possible, the 

moisture content of every sample should be determined since it is potentially a good indicator 

of performance. The test to determine the moisture content is simple and inexpensive to 

perform. 

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4.01 Primary References 

The primary references for this Chapter are as follows: 

ASTM (2006). Annual Book of ASTM Standards – Sections 4.02, 4.08, 4.09 and 4.13. 

ASTM International, West Conshohocken, PA. 

AASHTO (2006). Standard Specifications for Transportation Materials and Methods of 

Sampling and Testing, Parts I and II, American Association of State Highway and 

Transportation Officials, Washington, D.C. 

FHWA (2002a). Geotechnical Engineering Circular 5 (GEC5) - Evaluation of Soil and Rock 

Properties. Report No FHWA-IF-02-034. Authors: Sabatini, P.J, Bachus, R.C, Mayne, P.W., 

Schneider, J.A., Zettler, T.E., Federal Highway Administration, U.S. Department of 

Transportation. 

4.1 SOIL DESCRIPTION 

Soil description/identification is the systematic naming of individual soils in both written and 

spoken forms (ASTM D 2488, AASHTO M 145). Soil classification is the grouping of soils 

with similar engineering properties into a category by using the results of laboratory-based 

index tests, e.g., group name and symbol (ASTM D 2487, AASHTO M 145). It is important 

to distinguish between a visual description of a soil and its classification in order to minimize 

potential conflicts between general visual evaluations of soil samples in the field and more 

precise laboratory evaluations supported by index tests. 

The soil's description should include as a minimum: 

•	 Apparent consistency (e.g., soft, firm, etc. for fine-grained soils) or density adjective 

(e.g., loose, dense, etc. for coarse-grained soils); 

•	 Water content condition adjective (e.g., dry, moist, wet); 

•	 Color description (e.g., brown, gray, etc.); 

•	 Main soil type name, often presented in all capital letters (e.g. SAND, CLAY); 

•	 Descriptive adjective for main soil type (e.g., fine, medium, coarse, well-rounded, 

angular, etc. for coarse-grained soils; organic, inorganic, compressible, laminated, 

etc., for fine-grained soils); 

•	 Particle-size distribution adjective for gravel and sand (e.g., uniform, well-graded, 

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gap-graded); 

•	 Plasticity adjective (e.g., high, low) and soil texture (e.g., rough, smooth, slick, waxy, 

etc.) for inorganic and organic silts or clays; 

•	 Descriptive term for minor type(s) of soil (with, some, trace, etc.); 

•	 Minor soil type name with "y" added if the fine-grained minor component is less than 

30 percent but greater than 12 percent or the coarse-grained minor component is 30 

percent or more (e.g., silty for fine grained minor soil type, sandy for coarse-grained 

minor soil type); 

•	 Descriptive adjective “with” if the fine-grained minor soil type is 5 to 12 percent 

(e.g., with clay) or if the coarse-grained minor soil type is less than 30 percent but 15 

percent or more (e.g., with gravel). Note: some practices use the descriptive 

adjectives “some” and “trace” for minor components; 

•	 Inclusions (e.g., concretions, cementation); 

•	 Geological name (e.g., Holocene, Eocene, Pleistocene, Cretaceous), if known, in 

parenthesis or in notes column. 

The various elements of the soil description are generally stated in the order given above. 

For example, a soil description might be presented as follows: 

Fine-grained soils: Soft, wet, gray, high plasticity CLAY, with f. Sand; (Alluvium) 

Coarse-grained soils: Dense, moist, brown, silty m-f SAND, with f. Gravel to c. Sand; 

(Alluvium) 

When minor changes occur within the same soil layer (e.g., a change in apparent density), the 

boring log should indicate a description of the change, such as “same, except very dense.” 

4.1.1 Consistency and Apparent Density 

The consistency of fine-grained soils and apparent density of coarse-grained soils can be 

estimated from the energy-corrected SPT N-value, N60. The consistency of clays and silts 

varies from very soft to firm to stiff to hard. The apparent density of coarse-grained soil 

ranges from very loose to dense to very dense. Suggested guidelines for estimating the inplace apparent density or consistency of soils are given in Tables 4-1 and 4-2, respectively. 

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Table 4-1 

Evaluation of the apparent density of coarse-grained soils (after Peck, et al., 1974)
 

N60 Apparent Density Relative Density, % 

0 – 4 Very loose 0 – 20 

>4 - 10 Loose 20 – 40 

>10 - 30 Medium dense 40 – 70 

>30 - 50 Dense 70 – 85 

>50 Very Dense 85 – 100 

The above guidance may be misleading in gravelly soils. 

Table 4-2 

Evaluation of the consistency of fine-grained soils (after Peck, et al., 1974)
 

N60 Consistency 

Unconfined 

Compressive Strength, 

qu, ksf (kPa) 

Results of Manual Manipulation 

<2 Very soft 

< 0.5 

(<25) 

Specimen (height = twice the diameter) sags 

under its own weight; extrudes between 

fingers when squeezed. 

2 - 4 Soft 

0.5 – 1 

(25 – 50) 

Specimen can be pinched in two between the 

thumb and forefinger; remolded by light 

finger pressure. 

4 - 8 Medium stiff 

1 – 2 

(50 – 100) 

Can be imprinted easily with fingers; 

remolded by strong finger pressure. 

8 - 15 Stiff 

2 – 4 

(100 – 200) 

Can be imprinted with considerable pressure 

from fingers or indented by thumbnail. 

15 - 30 Very stiff 

4 – 8 

(200 – 400) 

Can barely be imprinted by pressure from 

fingers or indented by thumbnail. 

>30 Hard 

> 8 

>400 

Cannot be imprinted by fingers or difficult to 

indent by thumbnail. 

Note that N60-values should not be used to determine the design strength of fine grained soils. 

The apparent density or consistency of the soil formation can vary from these empirical 

correlations for a variety of reasons. Judgment remains an important part of the visual 

identification process. Field index tests (e.g., smear test, dried strength test, thread test) 

which will be described in the next section are suggested as aids in estimating the 

consistency of fine grained soils. 

In some cases the sampler may pass from one layer into another of markedly different 

properties; for example, from a dense sand into a soft clay. In attempting to identify apparent 

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density, an assessment should be made as to what part of the blow count corresponds to each 

layer since the sampler begins to reflect the presence of the lower layer before it actually 

reaches it. 

4.1.2 Water Content (Moisture) 

The relative amount of water present in the soil sample should be described by an adjective 

such as dry, moist, or wet as indicated in Table 4-3. 

Table 4-3 

Adjectives to describe water content of soils (ASTM D 2488) 

Description Conditions 

Dry No sign of water and soil dry to touch 

Moist Signs of water and soil is relatively dry to touch 

Wet Signs of water and soil definitely wet to touch; granular soil exhibits some free water 

when densified 

4.1.3 Color 

The color must be described when the sample is first retrieved in the field at the as-sampled 

water content since the color may change with changes in the water content. Primary colors 

should be used (brown, gray, black, green, white, yellow, red). Soils with different shades or 

tints of basic colors are described by using two basic colors; e.g., gray-green. Some agencies 

may require use of the Munsell color system (USDA, 1993). When the soil is marked with 

spots of color, the term “mottled” can be applied. Soils with a homogeneous texture but 

having color patterns that change and are not considered mottled can be described as 

“streaked.” 

4.1.4 Type of Soil 

The constituent parts of a given soil type are defined on the basis of texture in accordance 

with particle-size designators separating the soil into coarse-grained, fine-grained, and highly 

organic designations. Soil with more than 50 percent by weight of the particles larger than 

the U.S. Standard No. 200 sieve (0.075 mm) is designated coarse-grained. Soil (inorganic 

and organic) with 50 percent or more by weight of the particles finer than the No. 200 sieve 

(0.075 mm) is designated fine-grained. Soil primarily consisting of less than 50 percent by 

volume of organic matter, dark in color, and with an organic odor is designated as organic 

soil. Soil with organic content more than 50 percent is designated as peat. The soil type 

designations used by FHWA follow ASTM D 2487; i.e., gravel, sand, silt, clay, organic silt, 

organic clay, and peat. 

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4.1.4.1 Coarse-Grained Soils (Gravel and Sand) 

Coarse-grained soils consist of a matrix of either gravel or sand in which more than 50 

percent by weight of the soil is retained on the No. 200 sieve (0.075 mm). Coarse-grained 

soils may contain fine-grained soil, i.e., soils passing the No. 200 sieve (0.075 mm), but the 

percent by weight of the fine-grained portion is less than 50 percent. The gravel and sand 

components are defined on the basis of particle size as indicated in Table 4-4. The particlesize distribution is identified as well graded or poorly graded. Well graded coarse-grained 

soil contains a good representation of all particle sizes from largest to smallest, with #12 

percent fines. Poorly graded coarse-grained soil is uniformly graded, i.e., most of the coarsegrained particles are about the same size, with # 12 percent fines. Gap graded coarse grained 

soil can be either a well graded or poorly graded soil lacking one or more intermediate sizes 

within the range of the gradation. 

Gravels and sands may be described by adding particle-size distribution adjectives in front of 

the soil type in accordance with the criteria given in Table 4-5. Based on correlation with 

laboratory tests, the following simple field identification tests can be used as an aid in 

identifying granular soils. 

Table 4-4 

Particle size definition for gravels and sands (after ASTM D 2488) 

Component Grain Size Determination 

Boulders* 12” + 

(300 mm +) Measurable 

Cobbles* 3” to 12” 

(300 mm to 75 mm) Measurable 

Gravel 

Coarse 

Fine 

3⁄4” – 3” 

(19 mm to 75 mm) 

3⁄4” to #4 sieve (3⁄4” to 0.187”) 

(19 mm to 4.75 mm) 

Measurable 

Measurable 

Sand 

Coarse 

Medium

 Fine 

#4 to #10 sieve (0.19” to 0.079”) 

(4.75 mm – 2.00 mm) 

#10 to #40 sieve (0.079” to 0.017”) 

(2.00 mm – 0.425 mm) 

#40 to #200 sieve (0.017” to 0.003”) 

(0.425 mm- 0.075 mm) 

Measurable and visible to the eye 

Measurable and visible to the eye 

Measurable but barely discernible to the 

eye 

*Boulders and cobbles are not considered soil or part of the soil's classification or description, except 

under miscellaneous description; i.e., with cobbles at about 5 percent (volume). 

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Table 4-5 

Adjectives for describing size distribution for sands and gravels (after ASTM D 2488)
 

Particle-Size Adjective Abbreviation Size Requirement 

Coarse c. < 30% m-f sand or < 12% f. gravel 

Coarse to medium c-m < 12% f. sand 

Medium to fine m-f < 12% c. sand and > 30% m. sand 

Fine f. < 30% m. sand or < 12% c. gravel 

Coarse to fine c-f > 12% of each size1 

1 12% and 30% criteria can be modified depending on fines content. The key is the shape of 

the particle-size distribution curve. If the curve is relatively straight or dished down, and 

coarse sand is present, use c-f, also use m-f sand if a moderate amount of m. sand is 

present. If one has any doubts, determine the above percentages based on the amount of 

sand or gravel present. 

Feel and Smear Tests: A pinch of soil is handled lightly between the thumb and fingers to 

obtain an impression of the grittiness (i.e., roughness) or softness (smoothness) of the 

constituent particles. Thereafter, a pinch of soil is smeared with considerable pressure 

between the thumb and forefinger to determine the degrees of grittiness (roughness), or the 

softness (smoothness) of the soil. The following guidelines may be used: 

•	 Coarse- to medium-grained sand typically exhibits a very gritty feel and smear. 

•	 Coarse- to fine-grained sand has less gritty feel, but exhibits a very gritty smear. 

•	 Medium- to fine-grained sand exhibits a less gritty feel and smear that becomes softer 

(smoother) and less gritty with an increase in the fine sand fraction. 

•	 Fine-grained sand exhibits a relatively soft feel and a much less gritty smear than the 

coarser sand components. 

•	 Silt components less than about 10 percent of the total weight can be identified by a 

slight discoloration of the fingers after smear of a moist sample. Increasing silt 

increases discoloration and softens the smear. 

Sedimentation Test: A small sample of soil is shaken in a test tube filled with water and 

allowed to settle. The time required for the particles to fall a distance of 4-inches (100 mm) 

is about 1/2 minute for particle sizes coarser than silt. About 50 minutes would be required 

for particles of 0.0002 in (0.005 mm) or smaller (often defined as "clay size") to settle out. 

For sands and gravels containing more than 5 percent fines, the type of inorganic fines (silt or 

clay) can be identified by performing a shaking/dilatancy test. See fine-grained soils section. 

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Visual Characteristics: Sand and gravel particles can be readily identified visually, but silt 

particles are generally indistinguishable to the eye. With an increasing silt component, 

individual sand grains become obscured, and when silt exceeds about 12 percent, the silt 

almost entirely masks the sand component from visual separation. Note that gray finegrained sand visually appears to contain more silt than the actual silt content. 

4.1.4.2 Fine-Grained Soils 

Fine-grained soils are those having 50 percent or more by weight pass the No. 200 sieve. 

The so-called fines are either inorganic or organic silts and/or clays. To describe finegrained soils, plasticity adjectives and soil-type adjectives should be used to further define 

the soil's plasticity and texture. The following simple field identification tests can be used to 

estimate the degree of plasticity of fine-grained soils. 

Shaking (Dilatancy) Test (Holtz and Kovacs, 1981). Water is dropped or sprayed on a 

portion of a fine-grained soil sample mixed and held in the palm of the hand until it shows a 

wet surface appearance when shaken or bounced lightly in the hand or a sticky nature when 

touched. The test involves lightly squeezing the wetted soil sample between the thumb and 

forefinger and releasing it alternatively to observe its reaction and the speed of the response. 

Soils that are predominantly silty (nonplastic to low plasticity) will show a dull dry surface 

upon squeezing and a glassy wet surface immediately upon release of the pressure. This 

phenomenon becomes less and less pronounced in soils with increasing plasticity and 

decreasing dilatancy, 

Dry Strength Test (Holtz and Kovacs, 1981). A relatively undisturbed portion of the sample 

is allowed to dry out and a fragment of the dried soil is pressed between the fingers. 

Fragments which cannot be crumbled or broken are characteristic of clays with high 

plasticity. Fragments which can be disintegrated with gentle finger pressure are 

characteristic of silty materials of low plasticity. Thus, in generally, fine-grained materials 

with relatively high dry strength are clays of high plasticity and those with relatively little dry 

strength are predominantly silts. 

Thread Test (After Burmister, 1970). Moisture is added to or worked out of a small ball 

(about 1.5 in (40 mm) diameter) of fine grained soil and the ball kneaded until its consistency 

approaches medium stiff to stiff (compressive strength of about 2,100 psf (100 kPa)). This 

condition is observed when the material just starts to break or crumble. A thread is then 

rolled out between the palm of one hand and the fingers of the other to the smallest diameter 

possible before disintegration of the sample occurs. The smaller the thread achieved, the 

higher the plasticity of the soil. Fine-grained soils of high plasticity will have threads smaller 

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than 0.03 in (3/4 mm) in diameter. Soils with low plasticity will have threads larger than 

0.12 in (3 mm) in diameter. 

Smear Test (FHWA, 2002b). A fragment of soil smeared between the thumb and forefinger 

or drawn across the thumbnail will, by the smoothness and sheen of the smear surface, 

indicate the plasticity of the soil. A soil of low plasticity will exhibit a rough textured, dull 

smear while a soil of high plasticity will exhibit a slick, waxy smear surface. 

Table 4-6 identifies field methods to approximate the plasticity range for the dry strength, 

thread, and smear tests. 

Table 4-6 

Field methods to describe plasticity (FHWA, 2002b)
 

Plasticity 

Range Adjective Dry Strength Smear Test 

Thread Smallest 

Diameter, in 

(mm) 

0 Nonplastic none - crumbles into powder with 

mere pressure 

gritty or 

rough ball cracks 

1 - 10 low 

plasticity 

low - crumbles into powder with 

some finger pressure 

rough to 

smooth 

1/4 – 1/8 

(6 to 3) 

>10 - 20 medium 

plasticity 

medium - breaks into pieces or 

crumbles with considerable finger 

pressure 

smooth and 

dull 

1/16 

(1.5) 

>20 - 40 high 

plasticity 

high - cannot be broken with 

finger pressure; spec. will break 

into pieces between thumb and a 

hard surface 

Shiny 

0.03 

(0.75) 

>40 very plastic very high - can’t be broken 

between thumb and a hard surface 

very shiny 

and waxy 

0.02 

(0.5) 

4.1.4.3 Highly Organic Soils 

Colloidal and amorphous organic materials finer than the No. 200 sieve (0.075 mm) are 

identified and classified in accordance with their drop in plasticity upon oven drying (ASTM 

D 2487). Further identification markers are: 

1. dark gray and black and sometimes dark brown colors, although not all dark colored 

soils are organic; 

2. most organic soils will oxidize when exposed to air and change from a dark 

gray/black color to a lighter brown; i.e., the exposed surface is brownish, but when 

the sample is pulled apart the freshly exposed surface is dark gray/black; 

3. fresh organic soils usually have a characteristic odor that can be recognized, 

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particularly when the soil is heated; 

4. compared to inorganic soils, less effort is typically required to pull the material apart 

and a friable break is usually formed with a fine granular or silty texture and 

appearance; 

5. workability of organic soils at the plastic limit is weaker and spongier than an 

equivalent inorganic soil; 

6. the smear, although generally smooth, is usually duller and appears more silty than an 

equivalent inorganic soil’s; and 

7. the organic content of organic soils can also be determined by the combustion test 

method (AASHTO T 267, ASTM D 2974). 

Fine-grained soils, where the organic content appears to be less than 50 percent of the 

volume (about 22 percent by weight), should be described as soils with organic material or as 

organic soils such as clay with organic material or organic clays etc. If the soil appears to 

have an organic content greater than 50 percent by volume it should be described as peat. 

The engineering behavior of soils below and above the 50 percent dividing line is entirely 

different. It is therefore critical that the organic content of soils be determined both in the 

field and in the laboratory (AASHTO T 267, ASTM D 2974). Simple field or visual 

laboratory identification of soils as organic or peat is neither advisable nor acceptable. 

It is very important not to confuse topsoil with organic soils or peat. Topsoil is the relatively 

thin layer of soil found on the surface composed of partially decomposed organic materials, 

such as leaves, grass, small roots etc. Topsoil contains many nutrients that sustain plant and 

insect life and should not be used to construct geotechnical features or to support engineered 

structures. 

4.1.4.4 Minor Soil Type(s) 

Two or more soil types may be present in many soil formations,. When the percentage of the 

fine-grained minor soil type is less than 30 percent but greater than 12 percent, or the total 

sample or the coarse-grained minor component is 30 percent or more of the total sample, the 

minor soil type is indicated by adding a "y" to its name (e.g., f. gravelly, c-f. sandy, silty, 

clayey). Note the gradation adjectives are given for granular soils, while the plasticity 

adjective is omitted for the fine-grained soils. 

When the percentage of the fine-grained minor soil type is 5 to 12 percent or for the coarsegrained minor soil type is less than 30 percent but 15 percent or more of the total sample, the 

minor soil type is indicated by adding the descriptive adjective “with” to the group name 

(i.e., with clay, with silt, with sand, with gravel, and/or with cobbles). 

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Some local practices also use the descriptive adjectives “some” and “trace” for minor 

components as follows: 

•	 "trace" when the percentage is between 1 and 12 percent of the total sample; or 

•	 "some" when the percentage is greater than 12 percent and less than 30 percent of the 

total sample. 

4.1.4.5 Inclusions 

Additional inclusions or characteristics of the sample can be described by using "with" and 

the descriptions described above. For example: 

•	 with petroleum odor 

•	 with organic matter 

•	 with foreign matter (roots, brick, etc.) 

•	 with shell fragments 

•	 with mica 

•	 with parting(s), seam(s), etc. of (give soil’s complete description) 

4.1.4.6 Other Descriptors 

Depending on local conditions, the soils may be described based on reaction to HCl acid, and 

type and degree of cementation. ASTM D 2488 provides guidance for such descriptors. 

4.1.4.7 Layered Soils 

Soils of different types can be found in repeating layers of various thickness. It is important 

that all such formations and their thicknesses are noted. Each layer is described as if it is a 

non-layered soil by using the sequence for soil descriptions discussed above. The thickness 

and shape of layers and the geological type of layering are noted according to the descriptive 

terms presented in Table 4-7. The thickness designation is given in parentheses before the 

type of layer or at the end of each description, whichever is more appropriate. 

Examples of descriptions for layered soils are: 

•	 Medium stiff, moist to wet 0.2 to 0.75 in (5 to 20 mm) interbedded seams and layers 

of gray, medium plastic, silty CLAY and lt. gray, low plasticity SILT; (Alluvium). 

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•	 Soft moist to wet varved layers of gray-brown, high plasticity CLAY (0.2 to 0.75-in 

(5 to 20 mm)) and nonplastic SILT, trace f. sand (0.4 to 0.6 in (10 to 15 mm)); 

(Alluvium). 

Table 4-7 

Descriptive terms for layered soils (NAVFAC, 1986a) 

Type of Layer Thickness Occurrence 

Parting < 1/16” 

(< 1.5 mm) 

Seam 

1/16 to 1⁄2” 

(1.5 mm to 12 mm) 

Layer 

1⁄2” to 12” 

(12 mm to 300 mm) 

Stratum 

> 12” 

(>300 mm) 

Pocket Small erratic deposit 

Lens Lenticular deposit 

Varved (also 

layered) 

Alternating seams or layers of silt and/or clay 

and sometimes fine sand 

Occasional One or less per 12” (300 mm) of thickness or 

laboratory sample inspected 

Frequent More than one per 12” (300 mm) of thickness 

or laboratory 

4.1.4.8 Geological Name 

The soil description should include the geotechnical specialist’s assessment of the origin of 

the soil unit and the geologic name, if known. This information is generally placed in 

parentheses or brackets at the end of the soil description or in the field notes column of the 

boring log. Some examples include: 

a. Washington, D.C.-Cretaceous Age Material with SPT N-values between 30 and 100: 

Very hard gray-blue silty CLAY (CH), moist [Potomac Group Formation] 

b. Newport News, VA-Miocene Age Marine Deposit with SPT N-values around 10 to 

15: Stiff green sandy CLAY (CL) with shell fragments, calcareous [Yorktown 

Formation]. 

c. Tucson, AZ – Holocene Age Alluvial Deposit with SPT N-values around 35: 

Cemented clayey SAND (SC), dry [Pantano Formation]. 

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4.2 SOIL CLASSIFICATION 

As previously indicated, final identification with classification is best performed in the 

laboratory. This process will lead to more consistent final boring logs and avoid conflicts 

with field descriptions. The Unified Soil Classification System (USCS) group name and 

symbol (in parenthesis) appropriate for the soil type in accordance with AASHTO M 145 (or 

ASTM D 3282) or ASTM D 2487 is the most commonly used system in geotechnical work 

and is covered in this section. For classification of highway subgrade material, the AASHTO 

classification system (see Section 4.2.2) is used. The AASHTO classification system is also 

based on grain size and plasticity. 

4.2.1 Unified Soil Classification System (USCS) 

The Unified Soil Classification System (ASTM D 2487) groups soils with similar 

engineering properties into categories base on grain size, gradation and plasticity. Table 4-8 

provides a simplification of the group breakdown based on percent passing No. 200 sieve 

(0.075 mm) and Table 4-9 provides an outline of the complete laboratory classification 

method. The procedures, along with charts and tables, for classifying coarse-grained and 

fine-grained soils follow. 

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Table 4-8 

Basic USCS soil designations based on percent passing No. 200 sieve (0.075 mm) (after 

ASTM D 2487; Holtz and Kovacs, 1981) 

(“Clean”)

 (“Clean/Dirty”) 

(“Dirty”) 
 

 

 

 

Table 4-9 

Soil classification chart (laboratory method) (after ASTM D 2487)
 

Criteria for Assigning Group Symbols and Group Names 

Using Laboratory Testsa 

COARSE-GRAINED SOILS (Sands and Gravels) - more than 

50% retained on No. 200 (0.075 mm) sieve 

FINE-GRAINED (Silts and Clays) - 50% or more passes the 

No. 200 (0.075 mm) sieve 

Soil Classification 

Group 

Symbol 

Group 

Nameb 

GRAVELS 

More than 

50% of 

coarse 

Fraction 

retained on 

No. 4 

Sieve 

CLEAN 

GRAVELS 

< 5% fines 

Cu ≥ 4 and 1 ≤ Cc ≤ 3e

 GW Well-graded 

gravelf 

Cu < 4 and/or 1 > Cc > 3e

 GP Poorly-graded 

gravelf 

GRAVELS 

WITH FINES 

> 12% of finesc 

Fines classify as ML or MH GM Silty gravelf,g,h 

Fines classify as CL or CH GC Clayey 

gravelf,g,h 

SANDS 

50% or more 

of coarse 

fraction 

passes No. 4 

Sieve 

CLEAN 

SANDS 

< 5% finesd 

Cu ≥ 6 and 1 ≤ Cc ≤ 3e SW Well-graded 

Sandi 

Cu < 6 and/or 1 > Cc > 3e SP Poorly-graded 

sandi 

SANDS WITH 

FINES 

> 12% finesd 

Fines classify as ML or MH SM Silty sandg,h,i 

Fines classify as CL or CH SC Clayey 

sandg,h,i 

SILTS AND 

CLAYS 

Liquid limit 

less than 50 

Inorganic 

PI > 7 and plots on or above 

"A" linej CL Lean clayk,l,m 

PI < 4 or plots below "A" linej

 ML Siltk,l,m 

Organic < 0.75 

Liquid limit - not dried 

Liquid limit - overdried OL 

Organic 

clayk,l,m,n 

Organic 

siltk,l,m,o 

SILTS AND 

CLAYS 

Liquid limit 

50 or more 

Inorganic PI plots on or above "A" line CH Fat clayk,l,m 

PI plots below "A" line MH Elastic siltk,l,m 

Organic < 0.75 

Liquid limit - not dried 

Liquid limit - ove n dried 

OH 

Organic 

clayk,l,m,p 

Organic 

siltk,l,m,q 

Highly 

fibrous 

organic soils 

Primary organic matter, dark in color, and 

organic odor Pt Peat 

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Table 4-9 (Continued) 

Soil classification chart (laboratory method) (after ASTM D 2487)
 

NOTES: 

a Based on the material passing the 3 in (75 mm) sieve. 

b If field sample contained cobbles and/or boulders, add “with cobbles and/or boulders” 

to group name. 

c Gravels with 5 to 12% fines require dual symbols: 

GW-GM, well-graded gravel with silt 

GW-GC, well-graded gravel with clay 

GP-GM, poorly graded gravel with silt 

GP-GC, poorly graded gravel with clay 

d Sands with 5 to 12% fines require dual symbols: 

SW-SM, well-graded sand with silt 

SW-SC, well-graded sand with clay 

SP-SM, poorly graded sand with silt 

SP-SC, poorly graded sand with clay 

D (D )

2 

e C = 60 

u C 30 

c = D 10 (D 10 ) (D 60) 

 [Cu: Uniformity Coefficient; Cc: Coefficient of Curvature] 

f If soil contains ≥ 15% sand, add “with sand” to group name. 

g If fines classify as CL-ML, use dual symbol GC-GM, SC-SM. 

h If fines are organic, add “with organic fines” to group name. 

i If soil contains ≥ 15% gravel, add “with gravel” to group name. 

j If the liquid limit and plasticity index plot in hatched area on plasticity chart, soil is a 

CL-ML, silty clay. 

k If soil contains 15 to 29% plus No. 200 (0.075 mm), add “with sand” or “with gravel,” 

whichever is predominant. 

l If soil contains ≥ 30% plus No. 200 (0.075mm), predominantly sand, add “sandy” to 

group name. 

m If soil contains ≥ 30% plus No. 200 (0.075 mm), predominantly gravel, add “gravelly” 

to group name. 

n PI ≥ 4 and plots on or above “A” line. 

o PI < 4 or plots below “A” line. 

p PI plots on or above “A” line. 

q PI plots below “A” line. 

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Figure 4-1: Flow chart to determine the group symbol and group name for coarse-grained soils (A

FHWA NHI-06-088 4 – ESoils and Foundations – Volume I 4 - 17 
4.2.1.1 Classification of Coarse-Grained Soils 

Coarse-grained soils are defined as those in which 50 percent or more by weight are retained 

on the No. 200 sieve (0.075 mm). The flow chart to determine the group symbol and group 

name for coarse-grained soils is given in Figure 4-1. This figure is identical to Figure 3 in 

ASTM D 2487 except for the recommendation to capitalize the primary soil type; e.g., 

GRAVEL. 

•	 The shape of the grain-size distribution (GSD) curve or “gradation curve” as it 

is frequently called, is one of the more important aspects in a soil classification 

system for coarse-grained soils. The shape of the gradation curve can be 

characterized by a pair of “shape” parameters called the coefficient of uniformity, Cu, 

and the coefficient of curvature, Cc, to which numerical values may be assigned. By 

assigning numerical values to such shape parameters it becomes possible to compare 

grain-size distribution curves for different soils without having to plot them on the 

same diagram. In order to define shape parameters certain characteristic particle sizes 

must be identified that are common to all soils. Since the openings of a sieve are 

square, particles of many different shapes are able to pass through a sieve of given 

size even though the abscissa on the gradation curve is expressed in terms of particle 

“diameter,” which implies a spherical-shaped particle. Therefore, the “diameter” 

shown on the gradation curve is an effective diameter so that the characteristic 

particle sizes that must be identified to define the shape parameters are in reality 

effective grain sizes (EGS). 

A useful EGS for the characterizing the shape of the gradation curve is the grain size 

for which 10 percent of the soil by weight is finer. This EGS is labeled D10. This 

size is convenient because Hazen (1911) found that the ease with which water flows 

through a soil is a function of the D10. In other words, Hazen found that the sizes 

smaller than the D10 affected the permeability more than the remaining 90 percent of 

the sizes. Therefore, the D10 is a logical choice as a characteristic particle size. Other 

convenient sizes were found to be the D30 and the D60, which pertain to the grain size 

for which thirty and sixty percent, respectively, of the soil by weight is finer. These 

EGSs are used as follows in the Unified Soil Classification System (USCS) for the 

classification of coarse grained soils. 

•	 Slope of the gradation curve: The shape of the curve could be defined relative to an 

arbitrary slope of a portion of the gradation curve. Since one EGS has already been 

identified as the D10, the slope of the gradation curve could be described by 

identifying another convenient point (EGS) that is “higher” on the curve. Hazen 

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selected this other convenient size as the D60 that indicates the particle size for which 

60 percent of the soil by weight is finer. The slope between the D60 and the D10 can 

then be related to the degree of uniformity of the sample through a parameter called 

the “Coefficient of Uniformity” or the “Uniformity Coefficient,” Cu, which is 

expressed as follows: 

 

•	 Curvature of the gradation curve: The second “shape” parameter is used to 

evaluate the curvature of the gradation curve between the two arbitrary points, D60 

and D10. A third EGS, D30, that indicates the particle size for which 30 percent of the 

soil by weight is finer, is chosen for this purpose. The curvature of the slope between 

the D60 and the D10 can then be related to the three EGS’ through a parameter called 

the “Coefficient of Curvature” or the “Coefficient of Concavity” or the “Coefficient 

of Gradation,” Cc, which is expressed as follows: 

By use of the two “shape” parameters, Cu and Cc, the uniformity of the coarse-grained soil 

(gravel and sand) can now be classified as well-graded (non-uniform), poorly graded 

(uniform), or gap graded (uniform or non-uniform). Table 4-10 presents criteria for such 

classifications. 

Table 4-10 

Gradation based on Cu and Cc parameters 

D C 60 	 u	 = 4-1 D10 

D2

C 30

c =	 4-2 

D 60 x D10 

Gradation Gravels Sands 

Well-graded Cu ≥ 4 and 1 < Cc < 3 Cu ≥ 6 and 1 < Cc < 3 

Poorly graded Cu < 4 and 1 < Cc < 3 Cu < 6 and 1 < Cc < 3 

Gap graded* Cc not between 1 and 3 Cc not between 1 and 3 

*Gap-graded soils may be well-graded or poorly graded. In addition to the Cc value it is 

recommended that the shape of the GSD be the basis for definition of gap-graded. 

Cu and Cc are statistical parameters and provide good initial guidance. However, the plot of 

the GSD curve must always be reviewed in conjunction with the values of Cu and Cc to 

avoid incorrect classification. Examples of the importance of reviewing the GSD curves 

are presented in Figure 4-2 and discussed subsequently. 

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Curve B Curve A

Curve C 

D60 = 

0.6 mm 

D10 = 

0.075 mm 

D30 = 

0.2 mm 

Curve D10 

(mm) 

D30 

(mm) 

D60 

(mm) Cu Cc Gradation 

A 0.075 0.2 0.6 8.0 0.9 Well graded (1) 

B 1 1.5 2 2.0 1.12 Poorly graded - Gap graded (2) 

C 19 25 27 1.4 1.2 Poorly graded 

(1) 

(2) 

Soil does not meet Cu and Cc criteria for well-graded soil but GSD curve clearly 

indicates a well-graded soil 

The Cu and Cc parameters indicate a uniform (or poorly) graded material, but the 

GSD curve clearly indicates a gap-graded soil. 

Note: For clarity only the D10, D30, and D60 sizes for Curve A are shown on the figure. 

Figure 4-2. Evaluation of type of gradation for coarse-grained soils. 

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Discussion of Figure 4-2: Curve A in Figure 4-2 has Cu = 8 and Cc = 0.9. The soil 

represented by Curve A would not meet the criteria listed in Table 4-10 for well-graded soil, 

but yet an examination of the GSD curve shows that the soil is well-graded. Examination of 

the GSD curve is even more critical for the case of gap graded soils because the largest 

particle size evaluated by parameters Cu and Cc is D60 while the gap grading may occur at a 

size larger than D60 size as shown for a 2/3:1/3 proportion of gravel: sand mix represented by 

Curve B in Figure 4-2. Based on the criteria in Table 4-10, the soil represented by Curve B 

would be classified as a uniform or poorly graded soil which would be an incorrect 

classification. Such incorrect classifications can and do occur on construction sites where the 

contractor may (a) simply mix two stockpiles of uniformly graded soils leftover from a 

previous project. (b) use multiple sand and gravel pits to obtain borrow soils, and/or (c) mix 

soils from two different seams or layers of poorly graded material in the same gravel pit. 

Figure 4-2 is an illustration on the importance of evaluating the shape of the GSD curve in 

addition to the statistical parameters Cu and Cc. Practical aspects of the engineering 

characteristics of granular soils are discussed in Section 4.4. 

4.2.1.2 Classification of Fine-Grained Soils 

Fine-grained soils, or “fines,” are those in which 50 percent or more by weight pass the No. 

200 (0.075 mm) sieve, The classification of fine-grained soils is accomplished by use of the 

plasticity chart (Figure 4-3). For fine-grained organic soils, Table 4-11 may be used. 

Inorganic silts and clays are those that do not meet the organic criteria as given in Table 4-11. 

The flow charts to determine the group symbol and group name for fine-grained soils are 

given in Figure 4-4a and 4-4b. These figures are identical to Figures 1a and 1b in ASTM D 

2487 except that they are modified to show the soil type capitalized; e.g., CLAY. Dual 

symbols are used to classify organic silts and clays whose liquid limit and PI plot above the 

"A"-line, for example, CL-OL instead of OL and CH-OH instead of OH. To describe the 

fine-grained soil types more fully, plasticity adjectives and soil types used as adjectives 

should be used to further define the soil type's texture, plasticity, and location on the 

plasticity chart (see Table 4-12). Examples using Table 4-11 are given in Table 4-12. An 

example description of fine-grained soils is as follows: 

Soft, wet, gray, high plasticity CLAY, with f. Sand; Fat CLAY (CH); (Alluvium) 

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Figure 4-3. Plasticity chart for Unified Soil Classification System (ASTM D 2487). 

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Figure 4-4a. Flow chart to determine the group symbol and group name for fine-grained soils (AS

FHWA NHI-06-088 4 – ESoils and Foundations – Volume I 4 - 23 
Figure 4-4b. Flow chart to determine the group symbol and group name for organic soils (ASTM 

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Table 4-11 

Soil plasticity descriptors (based on Figures 4-3, 4-4a and 4-4b)
 

Plasticity 

Index Range 

Plasticity 

Adjective 

Adjective for Soil Type, Texture, and Plasticity 

Chart Location 

ML & 

MH 

(Silt) 

CL & CH 

(Clay) 

OL & OH 

(Organic Silt or Clay)1 

0 nonplastic - - ORGANIC SILT 

1 - 10 low plasticity - silty ORGANIC SILT 

>10 - 20 medium 

plasticity Clayey silty to no adj. ORGANIC clayey SILT 

>20 - 40 high plasticity Clayey - ORGANIC silty CLAY 

>40 very plastic Clayey - ORGANIC CLAY 

Soil type is the same for above or below the “A”-line; the dual group symbol (CL-OL or 

CH-OH) identifies the soil types above the “A”-line. 

Table 4-12 

Examples of description of fine-grained soils (based on Figures 4-3, 4-4a and 4-4b)
 

Group 

Symbol PI Group Name Complete Description For Main Soil Type (FineGrained Soil) 

CL 9 lean CLAY low plasticity silty CLAY 

ML 7 SILT low plasticity SILT 

ML 15 SILT medium plastic clayey SILT 

MH 21 elastic SILT high plasticity clayey SILT 

CH 25 fat CLAY 

high plasticity silty CLAY or high plasticity CLAY, 

depending on smear test (for silty relatively dull and not 

shiny or just CLAY for shiny, waxy) 

OL 8 

ORGANIC 

SILT 

low plasticity ORGANIC SILT 

OL 19 

ORGANIC 

SILT 

medium plastic ORGANIC clayey SILT 

CH >40 fat CLAY very plastic CLAY 

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4.2.2 AASHTO Soil Classification System 

The AASHTO soil classification system is shown in Table 4-13. The AASHTO 

classification system is useful in determining the relative quality of the soil material for use 

in earthwork structures, particularly embankments, subgrades, subbases and bases. 

According to this system, soil is classified into seven major groups, A-1 through A-7. Soils 

classified under groups A-1, A-2 and A-3 are granular materials where 35% or less of the 

particles pass through the No. 200 sieve (0.075 mm). Soils where more than 35% pass the 

No. 200 sieve (0.075 mm) are classified under groups A-4, A-5, A-6 and A-7. Soils where 

more than 35% pass the No. 200 sieve (0.075 mm) are mostly silt and clay-size materials. 

The classification procedure is shown in Table 4-13. The classification system is based on 

the following criteria: 

i Grain Size: The grain size terminology for this classification system is as follows: 

Gravel: fraction passing the 3 in (75 mm) sieve and retained on the No. 10 

(2 mm) sieve. 

Sand: fraction passing the No. 10 (2 mm) sieve and retained on the No. 

200 (0.075 mm) sieve 

Silt and clay: fraction passing the No. 200 (0.075 mm) sieve 

ii Plasticity: The term silty and clayey are used as follows: 

Silty: use when the fine fractions of the soil have a plasticity index of 10 

or less. 

Clayey: use when the fine fractions have a plasticity index of 11 or more. 

iii. If cobbles and boulders (size larger than 3 in (75 mm)) are encountered they are 

excluded from the portion of the soil sample on which the classification is made. 

However, the percentage of material is recorded. 

To evaluate the quality of a soil as a highway subgrade material, a number called the group 

index (GI) is also incorporated along with the groups and subgroups of the soil. The group 

index is written in parenthesis after the group or subgroup designation. The group index is 

given by Equation 4-3 where F is the percent passing the No. 200 (0.075 mm) sieve, LL is 

the liquid limit, and PI is the plasticity index. 

GI = (F-35)[0.2+0.005(LL-40)] + 0.01(F-15) (PI-10) 4-3 

 

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Table 4-13 

AASHTO soil classification system based on AASHTO M 145 (or ASTM D 3282)

GENERAL 

CLASSIFICATION 

GRANULAR MATERIALS 

(35 percent or less of total sample passing No. 200 sieve (0.075 mm) 

SILT-CLA

(More than 3

sample passing N

GROUP 

CLASSIFICATION 

A-1 

A-3 

A-2 

A-4 A-5 

A-1-a A-1-b A-2-4 A-2-5 A-2-6 A-2-7 

Sieve analysis, percent passing: No. 10 (2 mm) No. 40 (0.425 mm) No. 200 (0.075 mm) 

50 max. 30 max. 15 max. 

50 max. 25 max. 

51 min. 

10 max. 35 max. 35 max. 35 max. 35 max. 36 min. 36 min

Characteristics of fraction passing No 40 (0.425 mm) Liquid limit  Plasticity index 6 max. NP 

40 max. 10 max. 

41 min. 

10 max. 

40 max. 

11 min. 

41 min. 11 min. 

40 max. 10 max. 

41 min

10 max

Usual significant constituent materials 

Stone fragments, 

gravel and sand 

Fine 

sand Silty or clayey gravel and sand Silty soils 

Group Index** 0 0 0 4 max. 8 max. 12 maxClassification procedure: With required test data available, proceed from left to right on chart; correct group will be found by process of elimination. left into which the test data will fit is the correct classification. *Plasticity Index of A-7-5 subgroup is equal to or less than LL minus 30. Plasticity Index of A-7-6 subgroup is greater tha4-5). **See group index formula (Eq. 4-3). Group index should be shown in parentheses after group symbol as: A-2-6(3), A-4(5

FHWA NHI-06-088 4 – ESoils and Foundations – Volume I 4 - 27 
Figure 4-5. Range of liquid limit and plasticity index for soils in groups A-2, A-4, 

A-5, A-6 and A-7 per AASHTO M 145 (or ASTM D 3282).
 

The first term of Equation 4-3 is the partial group index determined from the liquid limit. 

The second term is the partial group index determined from the plasticity index. Following 

are some rules for determining group index: 

•	 If Equation 4-3 yields a negative value for GI, it is taken as zero. 

•	 The group index calculated from Equation 4-3 is rounded off to the nearest whole 

number, e.g., GI=3.4 is rounded off to 3; GI=3.5 is rounded off to 4. 

•	 There is no upper limit for the group index. 

•	 The group index of soils belonging to groups A-1-a, A-1-b, A-2-4, A-2-5, and A-3 

will always be zero. 

•	 When the group index for soils belonging to groups A-2-6 and A-2-7 is calculated, 

the partial group index for PI should be used, or 

GI=0.01(F-15) (PI-10) 4-4 

In general, the quality of performance of a soil as a subgrade material is inversely 

proportional to the group index. 

A comparison of the USCS and AASHTO system is shown in Figures 4-6 and 4-7. 

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Figure 4-6. Comparison of the USCS with the AASHTO soil classification system (after 

Utah DOT – Pavement Design and Management Manual, 2005).
 

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Figure 4-7. Comparison of soil groups in the USCS with the AASHTO Soil Classification Systems (Holtz and 

FHWA NHI-06-088 4 – ESoils and Foundations – Volume I 4 - 30 
4.3 ENGINEERING CHARACTERISTICS OF SOILS 

The major engineering characteristics of the main soil groups discussed in the previous 

section as related to foundation design are summarized as follows. A discussion on the 

practical aspects of the engineering characteristics is presented for granular and fine-grained 

soils following these summaries. 

4.3.1 Engineering Characteristics of Coarse-Grained Soils (Sands and Gravels) 

• Generally very good foundation material for supporting structures and roads. 

• Generally very good embankment material. 

• Generally the best backfill material for retaining walls. 

• Might settle under vibratory loads or blasts. 

• Dewatering may be difficult in open-graded gravels due to high permeability. 

• Generally not frost susceptible. 

4.3.2 Engineering Characteristics of Fine-Grained Soils (Inorganic Clays) 

• Generally possess low shear strength. 

• Plastic and compressible. 

• Can lose part of shear strength upon wetting. 

• Can lose part of shear strength upon disturbance. 

• Can shrink upon drying and expand upon wetting. 

• Generally very poor material for backfill. 

• Generally poor material for embankments. 

• Can be practically impervious. 

• Clay slopes are prone to landslides. 

4.3.3 Engineering Characteristics of Fine-Grained Soils (Inorganic Silts) 

• Relatively low shear strength. 

• High capillarity and frost susceptibility. 

• Relatively low permeability. 

• Frost heaving susceptibility 

• Difficult to compact. 

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4.3.4 Engineering Characteristics of Organic Soils 

The term organic designates those soils, other than topsoil, that contain an appreciable 

amount of vegetative matter and occasionally animal organisms in various states of 

decomposition. Any soil containing a sufficient amount of organic matter to influence its 

engineering properties is called an organic soil. The organic matter is objectionable for three 

main reasons: 

1. Reduces load carrying capacity of soil. 

2. Increases compressibility considerably. 

3. Frequently contains toxic gasses that are released during the excavation process. 

Generally organic soils, whether peat, organic clays, organic silts, or even organic sands, are 

not used as construction materials. 

4.4 PRACTICAL ASPECTS OF ENGINEERING CHARACTERISTICS OF 

COARSE-GRAINED SOILS 

Grain size distribution is the single most important element in the design of structures on, in, 

or composed of granular soils. As discussed in Chapter 2, grain size distribution is 

determined by sieving a dried soil sample of known weight through a nest of U.S. Standard 

sieves with decreasing mesh opening sizes. Figures 2-3 and 4-2 presented sample grain size 

distribution curves, also known as gradation curves, and introduced the terminology “well 

graded,” “poorly graded,” and “gap graded.” 

Much can be learned about a soil’s behavior from the shape and location of the curve. For 

instance, the “well graded” curve shown in Figure 4-2 represents a non-uniform soil with a 

wide range of particle sizes that are evenly distributed. Densification of a well-graded soil 

causes the smaller particles to move into the voids between the larger particles. As the voids 

in the soil are reduced, the density and strength of the soil increase. Specifications for select 

structural fill should contain required ranges of different particle sizes so that a dense, noncompressible backfill can be achieved with reasonable compactive effort. For example, the 

well-graded soil represented by Curve A shown in Figure 4-2 could be specified by providing 

the gradation limits listed in Table 4-14. 

As shown by Curve C in Figure 4-2, a poorly graded or uniform soil is composed of a narrow 

range of particle sizes. When compaction is attempted, inadequate distribution of particle 

sizes prevents reduction of the volume of voids by infilling with smaller particles. Such 

uniform soils should be avoided as select fill material. However, uniform soils do have an 

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important use as drainage materials. The relatively large and permanent void spaces act as 

conduits to carry water. Obviously, the larger the average particle size the larger the void 

space. The "French drain” is an example of the engineering use of a coarse uniform soil. 

Table 4-15 presents a typical specification for drainage materials having a narrow band of 

particle sizes. For material specifications related to drain material, it is important to specify 

that gap-graded materials shall not be acceptable. This is because gap-graded materials have 

variable permeabilities that may cause malfunction of the drain with associated damage to 

the geotechnical feature associated with the drain. 

Table 4-14 

Example gradation limits of well-graded granular material 

(see Curve A in Figure 4-2)
 

Sieve Size Percent Passing by Weight 

2′′ (50.8 mm) 100 

#10 (2 mm) 75-90 

#40 (0.425 mm) 40-60 

#200 (0.075 mm) 0 – 15 

Table 4-15 

Example gradation limits of drainage materials 

(see Curve C in Figure 4-2)
 

Sieve Size Percent Passing by Weight 

2′′ (50.8 mm) 100 

1 1⁄2 ′′ (37.5 mm) 90-100 

3⁄4 ′′ (19 mm) 0-15 

4.5 PRACTICAL ASPECTS OF ENGINEERING CHARACTERISITICS OF 

FINE-GRAINED SOILS 

As indicated in Chapter 2, the plasticity index (PI) is the difference between the liquid limit 

(LL) and the plastic limit (PL). The PI represents the range of water content over which the 

soil remains plastic. In general, the greater the PI, the greater the amount of clay particles 

present and the more plastic the soil. The more plastic a soil, the more likely it will be to 

have the following characteristics: 

1. Be more compressible. 

2. Have greater potential to shrink upon drying and/or swell upon wetting. 

3. Be less permeable. 

In addition to the PI, the Liquidity Index (LI) is a useful indicator of the engineering 

characteristics of fine-grained soils. Table 2-4 in Chapter 2 identifies the strength and 

deformation characteristics of fine-grained soils in terms of the LI. 

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4.6 DESCRIPTION OF ROCK 

When providing rock descriptions, geotechnical specialists should use technically correct 

geological terms. Local terms in common use may be acceptable if they help describe 

distinctive characteristics. Rock cores should be logged when wet for consistency of color 

description and greater visibility of rock features such as hairline fractures. The guidelines 

presented in the ISRM (1981), should be reviewed for additional information regarding 

logging procedures for core drilling. 

The rock's lithologic description should include as a minimum the following items: 

• Rock type 

• Color 

• Grain size and shape 

• Texture (stratification/foliation) 

• Mineral composition 

• Weathering and alteration 

• Strength 

• Other relevant notes 

The various elements of the rock's description should be stated in the order listed above, for 

example: 

"Limestone, light gray, very fine-grained, thin-bedded, unweathered, strong" 

The rock description should include identification of discontinuities and fractures. The 

description should also include a drawing of the naturally occurring fractures and mechanical 

breaks. 

4.6.1 Rock Type 

Rocks are classified according to their origin into three major divisions: igneous, 

sedimentary, and metamorphic (see Table 4-16). These three groups are subdivided into 

types according to mineral and chemical composition, texture, and internal structure. For 

some projects a library of hand samples and photographs representing lithologic rock types 

present in the project area should be maintained. 

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Table 4-16 

Rock groups and types (FHWA, 1997)
 

Igneous 

Intrusive 

(Coarse Grained) 

Extrusive 

(Fine Grained) 

Pyroclastic 

Granite 

Syenite 

Diorite 

Diabase 

Gabbro 

Peridotite 

Pegmatite 

Rhyolite 

Trachyte 

Andesite 

Basalt 

Obsidian 

Pumice 

Tuff 

Sedimentary 

Clastic (Sediment) Chemically Formed Organic Remains 

Shale 

Mudstone 

Claystone 

Siltstone 

Sandstone 

Conglomerate 

Limestone, oolitic 

Limestone 

Dolomite 

Gypsum 

Halite 

Chalk 

Coquina 

Lignite 

Coal 

Metamorphic 

Foliated Non-foliated 

Slate 

Phyllite 

Schist 

Gneiss 

Quartzite 

Amphibolite 

Marble 

Hornfel 

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4.6.2 Color 

Colors should be consistent with a Munsell Color Chart (USDA, 1993) and recorded for both 

wet and dry conditions as appropriate. 

4.6.3 Grain Size and Shape 

The grain size description should be classified according to the terms presented in Table 4­

17. Table 4-18 is used to classify the shape of the grains. The grain size descriptions are 

consistent with those used in the USCS for soil particles. 

Table 4-17 

Terms to describe grain size (typically for sedimentary rocks) 

Description Grain Size (mm) Characteristic of Individual Grains 

Very coarse grained #4 (> 4.75) Can be easily distinguished by eye 

Coarse grained #10 to #4 (2.00 -4.75) Can be easily distinguished by eye 

Medium grained #40 to #10 (0.425 -2.00) Can be distinguished by eye 

Fine grained #200 to #40 (0.075-0.425) Can be distinguished by eye with 

difficulty 

Very fine grained < #200 (< 0.075) Cannot be distinguished by unaided eye 

Table 4-18 

Terms to describe grain shape (for sedimentary rocks)
 

Description Characteristic 

Angular Showing very little evidence of wear. Grain edges and corners are sharp. Secondary 

corners are numerous and sharp. 

Subangular Showing some evidence of wear. Grain edges and corners are slightly rounded off. 

Secondary corners are slightly less numerous and slightly less sharp than in angular 

grains. 

Subrounded Showing considerable wear. Grain edges and corners are rounded to smooth curves. 

Secondary corners are reduced greatly in number and highly rounded. 

Rounded Showing extreme wear. Grain edges and corners are smoothed off to broad curves. 

Secondary corners are few in number and rounded. 

Wellrounded Completely worn. Grain edges or corners are not present. No secondary edges or 

corners are present. 

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4.6.4 Stratification/Foliation 

Significant non-fracture structural features should be described. The thickness should be 

described by using the terms in Table 4-19. The orientation of the bedding/foliation should 

be measured from the horizontal or with respect to the core axis. 

Table 4-19. Terms to describe stratum thickness 

Descriptive Term Stratum Thickness in (mm)* 

Very Thickly bedded 

Thickly bedded 

Thinly bedded 

Very Thinly bedded 

Laminated 

Thinly Laminated 

(> 1 m) 

(0.5 to 1.0 m) 

(50 mm to 500 mm) 

(10 mm to 50 mm) 

(2.5 mm to 10 mm) 

(< 2.5 mm) 

* Conventionally measured in m or mm. (1 m = 3.28 ft; 25.4 mm = 1 in) 

4.6.5 Mineral Composition 

The mineral composition should be identified by a geologist based on experience and the use 

of appropriate references. The most abundant mineral should be listed first, followed by 

minerals in decreasing order of abundance. For some common rock types, the mineral 

composition need not be specified (e.g. dolomite, limestone). 

4.6.6 Weathering and Alteration 

Weathering and alteration is due to the weathering processes discussed in Chapter 3, e.g., 

physical, chemical and thermal mechanisms. Terms and abbreviations used to describe 

weathering and alteration are presented in Table 4-20. 

4.6.7 Strength 

The point load test described in Chapter 5 is recommended for the measurement of sample 

strength. The point-load index, Is, obtained from the point load test should be converted to 

uniaxial compressive strength. Categories and terminology for describing rock strength 

based on the uniaxial compressive strength are presented in Table 4-21. Table 4-21 also 

presents guidelines for common qualitative assessments of strength that can be performed 

with the aid of a geologist’s hammer and a pocket knife while the geotechnical specialist is 

mapping or doing primary logging of core at the drill rig site. The field estimates should be 

confirmed where appropriate by comparison with selected laboratory tests. 

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Table 4-20
 

Terms to describe rock weathering and alteration (ISRM, 1981)
 

Grade (Term) Description 

I 

(Fresh) 

Rock shows no discoloration, loss of strength, or other effects of 

weathering/alteration 

II 

(Slightly 

Weathered/Altered) 

Rock is slightly discolored, but not noticeably lower in strength than 

fresh rock 

III 

(Moderately 

Weathered/Altered) 

Rock is discolored and noticeably weakened, but less than half is 

decomposed; a minimum 2 in (50 mm) diameter sample cannot be 

broken readily by hand across the rock fabric 

IV 

(Highly 

Weathered/Altered) 

More than half of the rock is decomposed; rock is weathered so that a 

minimum 2 in (50 mm) diameter sample can be broken readily by 

hand across the rock fabric 

V 

(Completely 

Weathered/Altered) 

Original minerals of rock have been almost entirely decomposed to 

secondary minerals even though the original fabric may be intact; 

material can be granulated by hand 

VI 

(Residual Soil) 

Original minerals of rock have been entirely decomposed to secondary 

minerals, and original rock fabric is not apparent; material can be 

easily broke by hand 

Table 4-21
 

Terms to describe rock strength (ISRM, 1981)
 

Grade (Description) Field Identification Approximate Range of Uniaxial 

Compressive Strength, psi (kPa) 

R0 

(Extremely Weak Rock) 

Can be indented by 

thumbnail 

35 

(250) - 150 

(1,000) 

R1 

(Very Weak Rock) 

Can be peeled by pocket 

knife 

150 

(1,000) - 725 

(5,000) 

R2 

(Weak Rock) 

Can be peeled with 

difficulty by pocket knife 

725 

(5,000) - 3,500 

(25,000) 

R3 

(Medium Strong Rock) 

Can be indented 3/16 in (5 

mm) with sharp end of pick 

3,500 

(25,000) - 7,000 

(50,000) 

R4 

(Strong Rock) 

Requires one blow of 

geologist’s hammer to 

fracture 

7,000 

(50,000) - 15,000 

(100,000) 

R5 

(Very Strong Rock) 

Requires many blows of 

geologist’s hammer to 

fracture 

15,000 

(100,000) - 36,000 

(250,000) 

R6 

(Extremely Strong Rock) 

Can only be chipped with 

blows of geologist’s 

hammer 

> 36,000 

(>250,000) 

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4.6.8 Hardness 

Hardness is commonly assessed by the scratch test. Descriptions and abbreviations used to 

describe rock hardness are presented in Table 4-22. 

Table 4-22 

Terms to describe rock hardness (FHWA, 2002b) 

Description (Abbr) Characteristic 

Soft (S) Reserved for plastic material alone. 

Friable (F) Easily crumbled by hand, pulverized or reduced to powder. 

Low Hardness (LH) Can be gouged deeply or carved with a pocket knife. 

Moderately Hard 

(MH) 

Can be readily scratched by a knife blade; scratch leaves a heavy 

trace of dust and scratch is readily visible after the powder has 

been blown away. 

Hard (H) Can be scratched with difficulty; scratch produces little powder 

and is often faintly visible; traces of the knife steel may be visible. 

Very Hard (VH) Cannot be scratched with pocket knife. Leave knife steel marks on 

surface. 

4.6.9 Rock Discontinuity 

Discontinuity is the general term for any mechanical break in a rock mass that has zero or 

low tensile strength. Discontinuity is the collective term used for most types of joints, weak 

bedding planes, weak schistosity planes, weakness zones, and faults. The spacing between 

discontinuities is defined as the perpendicular distance between adjacent discontinuities. The 

spacing should be measured perpendicular to the planes in the set. Table 4-23 presents 

guidelines to describe discontinuity spacing. 

Discontinuities should be described as closed, open, or filled. Aperture is the term used to 

describe the perpendicular distance separating the adjacent rock walls of an open 

discontinuity in which the intervening space is air- or water-filled. Width is the term used to 

describe the distance separating the adjacent rock walls of filled discontinuities. The terms 

presented in Table 4-24 should be used to describe apertures. Terms such as "wide,” 

"narrow" and "tight" are used to describe the width of discontinuities such as thickness of 

veins, fault gouge filling, or joints openings. Guidelines for use of such terms are presented 

in Tables 4-23 and 4-24. 

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Table 4-23. Terms to describe discontinuities (after ISRM, 1981) 

Discontinuity Type Amount of Infilling Discontinuity Spacing (m)* 

F - Fault Su - Surface Stain EW - Extremely Wide (>6) 

J - Joint Sp - Spotty VW - Very Wide (2-6) 

Sh - Shear Pa - Partially Filled W - Wide (0.6-2) 

Fo - Foliation Fi - Filled M - Moderate (0.2-0.6) 

V - Vein No - None C - Close (0.06-0.2) 

B - Bedding VC 

EC 

-

-

Very Close (0.02-0.06) 

Extremely close (<0.02) 

Discontinuity Width (mm)* 

W 

MW 

N 

VN 

T 

-

-

-

-

-

Wide (12.5-5.0) 

Moderately Wide (2.5-12.5) 

Narrow (1.25-2.5) 

Very Narrow (<1.25) 

Tight (~ 0) 

Surface Shape of Joint 

Wa 

Pl 

St 

Ir 

-

-

-

-

Wavy 

Planar 

Stepped 

Irregular 

Type of Infilling 

Cl 

Ca 

Ch 

Fe 

Gy 

H 

No 

Py 

Qz 

Sd 

-

-

-

-

-

-

-

-

-

-

Clay 

Calcite 

Chlorite 

Iron Oxide 

Gypsum/Talc 

Healed 

None 

Pyrite 

Quartz 

Sand 

Roughness of Surface 

Slk - Slickensided (surface has smooth, glassy finish with 

visual evidence of striations) 

S - Smooth (surface appears smooth and feels so to the 

touch) 

SR - Slightly Rough (asperities on the discontinuity surface 

are distinguishable and can felt) 

R - Rough (some ridges and side-angle steps are evident; 

asperities are clearly visible, and discontinuity surface 

feels very abrasive) 

V 

R 

- Very Rough (near-vertical steps and ridges occur on 

the discontinuity surface 

* Conventionally measured in m or mm. (1 m = 3.28 ft; 1 in = 25.4 mm) 

Table 4-24. Terms to classify discontinuities based on aperture size (ISRM, 1981) 

Aperture (mm)* Description 

<0.1 

0.1 - 0.25 

0.25 – 0.5 

Very tight 

Tight 

Partly open 

"Closed Features" 

0.5 - 2.5 

2.5 - 10 

> 10 

Open 

Moderately open 

Wide 

"Gapped Features" 

1-100 

100-1000 

>1 m 

Very wide 

Extremely wide 

Cavernous 

"Open Features" 

* Conventionally measured in mm, cm or m. (1 m = 3.28 ft; 1 in = 25.4 mm) 

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For faults or shears that are not thick enough to be represented on the boring log, the 

measured thickness is recorded numerically in millimeters. 

Discontinuities are further characterized by the surface shape of the joint and the roughness 

of its surface in addition to the fill material separating the adjacent rock walls of the 

discontinuities. Filling is characterized by its type, amount, width (i.e., perpendicular 

distance between adjacent rock walls) and strength. If non-cohesive fillings are identified, 

then the filling should be identified qualitatively, e.g., fine sand. Refer to Table 4-23 for 

guidelines to characterize these features. 

4.6.10 Fracture Description 

Naturally occurring fractures are numbered and described by using the same terminology that 

is used for discontinuities. The number of naturally occurring fractures observed in each 1 ft 

(0.5 m) of core should be recorded as the fracture frequency. Mechanical breaks, thought to 

have occurred during drilling, are not counted. The following criteria can be used to identify 

natural breaks: 

1. A rough brittle surface with fresh cleavage planes in individual rock minerals 

suggests an artificial fracture. 

2. A generally smooth or somewhat weathered surface with soft coating or infilling 

materials, such as talc, gypsum, chlorite, mica, or calcite indicates a natural 

discontinuity. 

3. In rocks showing foliation, cleavage or bedding it may be difficult to distinguish 

between natural discontinuities and artificial fractures when the discontinuities are 

parallel with the incipient weakness planes. If drilling has been carried out carefully 

then the questionable breaks should be counted as natural features to be on the 

conservative side. 

4.	 Depending upon the drilling equipment, part of the length of core being drilled may 

occasionally rotate with the inner barrels in such a way that grinding of the surfaces 

of discontinuities and fractures occurs. In weak rock types it may be very difficult to 

decide if the resulting rounded surfaces represent natural or artificial features. When 

in doubt, conservatively assume that they are natural. 

The fracture description can be strongly time dependent and moisture content dependent in 

the case of certain varieties of shales and mudstones that have relatively weakly developed 

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diagenetic bonds. A diagenetic bond is the bond that is formed in a deposited sediment by 

chemical and physical processes during its conversion to rock. A frequent problem is 

"discing,” in which an initially intact core separates into discs on incipient planes. The 

process generally becomes noticeable perhaps within a few minutes of core recovery. This 

phenomenon is experienced in several different forms: 

1. Stress relief cracking and swelling by the initially rapid release of strain energy in 

cores recovered from areas of high stress, especially in the case of shaley rocks. 

2. Dehydration cracking experienced in the weaker mudstones and shales that may 

reduce RQD values from 100 percent to 0 percent in a matter of minutes. The initial 

integrity might possibly have been due to negative pore water pressure. 

3. Slaking and cracking experienced by some of the weaker mudstones and shales when 

they are subjected to wetting and drying. 

Any of these forms of “discing” may make logging of fracture frequency unreliable. 

Whenever such conditions are anticipated, core should be logged by a geotechnical specialist 

as it is being recovered and at subsequent intervals until the phenomenon is predictable. 

4.6.11 Rock Mass Classification 

In determining the rock strength for transportation facilities constructed in, on, or of rock, it 

is most important to account for the presence of discontinuities, such as joints, faults or 

bedding planes. Therefore, for most conditions, the rock mass strength properties, rather 

than the intact rock properties must be determined for use in design. The rock mass is the insitu, fractured rock that will almost always have significantly lower strength than the intact 

rock because of discontinuities that divide the rock mass into blocks. Therefore, the strength 

of the rock mass will depend on such factors as the shear strength of the surfaces of the 

blocks, the spacing and continuous length of the discontinuities and their alignment relative 

to the direction of loading. These factors were identified in the previous sections. Using 

these factors, Bieniawski (1989) proposed a method for estimating rock mass properties from 

an index that characterizes the overall properties of the rock mass quality. This index is 

known as the rock mass rating (RMR). Originally developed for tunnel support design, the 

RMR has been adopted by AASHTO (2004 with 2006 Interims) because the RMR is 

determined from readily measurable parameters. Table 4-25 identifies the following five 

measurable parameters and assigns relative ratings to each parameter: 

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Table 4-25 

Geomechanics classification of rock masses (AASHTO 2004 with 2006 Interims) 

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PARAMETER RANGES OF VALUES 

1 

Strength of intact rock material 

Point load strength index 

>1,200 psi 600 to 1,200 

psi 

300 to 600 

psi 

150 to 300 

psi 

For thicompr

Uniaxial compressive strength 

>30,000 psi 15,000 to 

30,000 psi 

7,500 to 

15,000 psi 

3,600 to 

7,500 psi 

1,500 

3,600 

Relative Rating 15 12 7 4 2 

2 Drill core quality RQD 90% to 100% 75% to 90% 50% to 75% 25% to 50% 

Relative Rating 20 17 13 8 

3 Spacing of joints >10 ft 3 to 10 ft 1 to 3 ft 2 in. to 1 foo

Relative Rating 30 25 20 10 

4 

Condition of joints 

••••

Very rough surfaces Not continuous No separation Hard joint wall rock 

•••

Slightly rough 

surfaces 

Separation <0.05” Hard joint wall 

rock 

•

••

Slightly rough 

surfaces 

Separation <0.05” Soft joint wall rock 

•

•

••

Slickensided su

- or -

Gouge <0.2 in t

orJoints open 0.05

Continuous join

Relative Rating 25 20 12 6 

5 Ground water conditions 

(use one of the three evaluation criteria as appropriate to the method of exploration) 

Inflow per 30 ft tunnel length 

None <400 gallons/hr 400 to 2,000 gall

Ratio= joint water pressure/ major principal stress 

0 0.0 to 0.2 0.2 to 0.5 

General Conditions Completely Dry Moist only 

(interstitial water) 

Water under mod

pressure 

Relative Rating 10 7 4 

Note: 1 psi = 6.895 kPa; 1 in = 25.4 mm 
1. Strength of intact rock material. 

2. Drill core quality as expressed by RQD. 

3. Spacing of joints. 

4. Condition of joints. 

5. Ground water conditions. 

The RMR is determined as the sum of the five relative ratings. The RMR should be adjusted 

in accordance with the criteria in Table 4-26. The rock classification should be 

determined in accordance with Table 4-27 where RMR refers to the adjusted value. 

Table 4-26 

Geomechanics rating adjustment for joint orientations 

(after AASHTO 2004 with 2006 Interims) 

Orientations of joints Very 

favorable Favorable Fair Unfavorable Very 

Unfavorable 

Ratings 

Tunnels 0 -2 -5 -10 -12 

Foundations 0 -2 -7 -15 -25 

Slopes 0 -5 -25 -50 -60 

Table 4-27 

Geomechanics rock mass classes determined from total ratings 

(AASHTO 2004 with 2006 Interims) 

RMR 

(Note 1) 

100 to 81 80 to 61 60 to 41 40 to 21 <20 

Class No. I II III IV V 

Description Very good 

rock Good rock Fair rock Poor rock Very poor 

rock 

Note 1: RMR is adjusted for structural application and rock joint orientation as per Table 4­

26 prior to evaluating the Class No. 

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4.7 SUBSURFACE PROFILE DEVELOPMENT 

The mark of successfully accomplishing a subsurface exploration is the ability to draw a 

subsurface profile of the project site complete with soil types, rock interfaces, and the 

relevant design properties. The subsurface profile is a visual display of subsurface conditions 

as interpreted from all of the methods of explorations and testing described previously. 

Uncertainties in the development of a subsurface exploration usually indicate the need for 

additional explorations or testing. Because of the diverse nature of the geologic processes 

that contribute to soil formation, actual subsurface profiles can be extremely varied both 

vertically and horizontally, and can differ significantly from interpreted profiles developed 

from boring logs. Therefore, subsurface profiles developed from boring logs should contain 

some indication that the delineation between strata do not necessarily suggest that distinct 

boundaries exist between the strata or that the interpolations of strata thickness between 

borings are necessarily correct. The main purpose of subsurface profiles is to provide a 

starting point for design and not necessarily to present an accurate description of subsurface 

conditions. 

In the optimum situation, the subsurface profile is developed in stages. First, a rough profile 

is established from the driller’s logs by the geotechnical specialist. The object is to discover 

any obvious gaps or question marks while the drill crew is still at the site so that additional 

work can be performed immediately. Once a crew has left the site, a delay of months may 

occur before their schedule permits them to reoccupy the site, not to mention the additional 

cost to remobilize/demobilize. The drilling inspector or crew chief should be required to call 

the project geotechnical specialist when the last scheduled boring has begun to request 

instructions for any supplemental borings. 

When all borings are completed and laboratory visuals and moisture content data received, 

the initial subsurface profile should be revised. Estimated soil layer boundaries and accurate 

soil descriptions should be established for soil deposits. Estimated bedrock interfaces should 

be identified. Most importantly, the depth to perched or regional groundwater should be 

indicated. The over-complication of the profile by noting minute variations between adjacent 

soil samples can be avoided by: 

1. Reviewing the geologic history of the site, e.g., if the soil map denotes a lakebed 

deposit overlying a glacial till deposit, do not subdivide the lakebed deposit because 

adjacent samples have differing amounts of silt and clay. Realize before breaking 

down the soil profile that probably only two layers exist and variations are to be 

expected within each. Important variations such as the average thickness of silt and 

clay varves can be noted adjacent to the visual description of the layer. 

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2. Remembering that the soil samples examined are only a minute portion of the soil 

underlying the site and must be considered in relation to adjacent samples as well as 

adjacent borings. 

A few simple rules should be followed at this stage to interpret the available data properly: 

1. Review the USDA Soil Survey map for the county and determine major surface and 

near-surface deposits that can be expected at the site. 

2. Examine the subsurface log containing SPT results and the laboratory visual 

descriptions with accompanying moisture contents. 

3. Review representative soil samples to check laboratory identifications and to 

calibrate your interpretations with those of the laboratory technicians who 

performed the visual description. 

4. Establish rational mechanics for drawing the soil profile. For example: 

 a. Use a vertical scale of 1 in equals 10 ft or 20 ft; generally, any smaller scale 

tends to compress data visually and prevent proper interpretation. 

 b. Use a horizontal scale equal to the vertical scale, if possible, to simulate actual 

relationships. However, the total length should be kept within 36 inches (920 

millimeter) to permit review in a single glance. 

When the subsurface layer boundaries and descriptions have been established, determine the 

extent and details of laboratory testing. Do not casually read the driller’s log and randomly 

select certain samples for testing. Plan the test program intelligently from the subsurface 

profile and for the proposed feature. Identify major soil deposits and assign appropriate tests 

for the design project under investigation. 

The final subsurface profile is the geotechnical specialist’s best interpretation of all available 

subsurface data. The final subsurface profile should include the following: 

•	 interpreted boundaries of soil and rock 

•	 the average physical properties of the soil layers, e.g., unit weight, shear strength, etc. 

•	 a visual description of each layer including USCS symbols for soil classification 

•	 location of the ground water level, and 

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• notations for special items such as boulders, artesian pressure, etc. 

If the inclusion of all of the information listed above clutters the subsurface profile, then 

complementary tables containing some of that information should be developed to 

accompany the profile. Figures 4-8 and 4-9 show a typical boring location plan and an 

interpreted subsurface profile. Note that the interpreted boundaries of rock and 

groundwater profiles are for internal agency use. Such interpretations should not be 

presented in bid documents. Another example of boring location plan and subsurface 

profile is presented in Chapter 11 (Geotechnical Reports). 

Figure 4-8. Example boring location plan (FHWA, 2002a). 

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Soils and Foundations – Volume I 4 - 47 December 2006
 
Figure 4-9. Example interpreted subsurface profile (FHWA, 2002a). 

4.7.1 Use of Historical Data in Development of Subsurface Profile 

Data from historical boring logs from the area can be used to supplement data provided by 

the current boring logs in developing a subsurface profile, however, such historical logs need 

to be reviewed carefully well in advance of drilling activities to ensure that the data are 

accurate. In some cases, boring log locations are referenced to the center alignment of a 

roadway without the location of the borehole having been actually surveyed. It is imperative 

to ensure that a consistent coordinate system is used to establish the correct relative location 

of all borings. Since borings would have likely been performed over an extended period of 

time or for different contracts along a roadway alignment (i.e., project centerlines are 

commonly changed during project development), it is possible that coordinate systems will 

not be consistent. Simply stated, if a historical boring cannot be located confidently on a site 

FHWA NHI-06-088 4 – Engineering Characteristics 

Soils and Foundations – Volume I 4 - 48 December 2006
 
plan, then the boring has limited usefulness for establishing stratigraphy. Also, it is likely 

that different drill rigs with different operators and different energy efficiencies were used in 

the collection of SPT data on historical boring logs. This factor must also be recognized 

when an attempt is made to correlate engineering properties to SPT blow count values. 

However, the geotechnical specialist should realize that while there may be potential 

limitations in the use of historical borings, it is necessary to review these borings relative to 

the design under consideration. As an example, a historical boring may indicate a thick layer 

of very soft clay as evidenced by the description “weight of rod/weight of hammer” in the 

SPT recording box of the log at a large number of test depths. While shear strength and 

consolidation properties cannot be reliably estimated based on SPT blow count values, the 

historical boring may provide useful information concerning the depth to a firm stratum. 

Most DOTs have collected large amounts of subsurface data from previous investigations 

within their states. Unfortunately, much of these data are archived with related project data 

once the project has been completed, and thus may not be readily available or accessible for 

use during future projects. Additionally, the subsurface data may not be fully utilized if the 

locations of the borings are not identified properly or if the plan drawing of the project site is 

not maintained with the boring logs. To overcome this problem, many DOTs currently use 

longitude and latitude to identify the boring locations, in lieu of or in conjunction with the 

conventional positioning format that uses station and offset. Unfortunately, the vast majority 

of the historical subsurface boring information is available only on paper. Therefore, a 

considerable amount of work is required to convert that data into electronic form before it 

can be fully appreciated and used to establish an electronic database of the subsurface 

information. 

Several DOTs have recently commenced using electronic boring records for their projects. 

Not only does the use of electronic boring records provide a redundancy to compliment the 

paper copy, but it also preserves data in a way that has the potential for automated electronic 

data management. One method of electronic data management increasingly used by DOTs 

involves the use of a centralized electronic database in conjunction with Geographic 

Information System (GIS) techniques to locate and identify borings on a plan. In its most 

simplistic form, the electronically stored data are managed and assessed visually by using 

GIS software, where each boring location is identified on a plan map. An appropriately 

developed database and GIS can be used to great advantage by the DOT. Specifically, in 

addition to the previously mentioned advantages of having electronic data records 

compliment paper logs, it is possible to: 

1. catalog borings that were conducted previously; 

2. inventory data regarding specific problematic formations across the state; and 

FHWA NHI-06-088 4 – Engineering Characteristics 

Soils and Foundations – Volume I 4 - 49 December 2006
 
3.	 develop cross sections that depict subsurface conditions across a site or within a 

region. 

This type of application of electronic boring records and data base accessibility can facilitate 

the development of subsequent subsurface investigations that are appropriately focused and 

that optimize the utility of existing data. 

FHWA NHI-06-088 4 – Engineering Characteristics 

Soils and Foundations – Volume I 4 - 50 December 2006
 
CHAPTER 5.0 

LABORATORY TESTING 

FOR GEOTECHNICAL DESIGN AND CONSTRUCTION 

Laboratory testing of soils and rocks is a fundamental element of geotechnical engineering. 

The complexity of testing required for a particular project may range from a simple moisture 

content determination to sophisticated triaxial strength testing. A laboratory test program 

should be well-planned to optimize the test data for design and construction. The 

geotechnical specialist, therefore, should recognize the project=s issues ahead of time so as to 

optimize the testing program, particularly strength and consolidation testing. 

Laboratory testing of samples recovered during subsurface investigations is the most 

common technique to obtain values of the engineering properties necessary for design. A 

laboratory-testing program consists of “index tests” to obtain general information on 

categorizing materials, and “performance tests” to measure specific properties that 

characterize soil behavior for design and constructability assessments (e.g., shear strength, 

compressibility, hydraulic conductivity, etc.). This chapter provides information on common 

laboratory test methods for soils and rocks including testing equipment, general procedures 

related to each test, and parameters measured by the tests. 

5.01 Primary References 

The primary references for this Chapter are as follows: 

ASTM (2006). Annual Book of ASTM Standards – Sections 4.02, 4.08, 4.09 and 4.13. 

ASTM International, West Conshohocken, PA. 

AASHTO (2006). Standard Specifications for Transportation Materials and Methods of 

Sampling and Testing, Parts I and II, American Association of State Highway and 

Transportation Officials, Washington, D.C. 

FHWA (2002a). Geotechnical Engineering Circular 5 (GEC5) - Evaluation of Soil and Rock 

Properties. Report No FHWA-IF-02-034. Authors: Sabatini, P.J, Bachus, R.C, Mayne, P.W., 

Schneider, J.A., Zettler, T.E., Federal Highway Administration, U.S. Department of 

Transportation. 

FHWA NHI-06-088 5 – Laboratory Tests 

Soils and Foundations – Volume I 5 - 1 December 2006 
5.1 QUALITY ASSURANCE FOR LABORATORY TESTING 

Laboratory testing will be required for most projects. Therefore, it is necessary to select the 

appropriate types and quantities of laboratory tests to be performed. A careful review of all 

data obtained during the field investigation and a thorough understanding of the preliminary 

design of geotechnical, structural and hydraulic features of the project are essential to 

develop an appropriately scoped laboratory testing program. In some cases owners may hire 

external testing laboratories to perform select tests. It is necessary that testing requests be 

clear and sufficiently detailed. Unless specialized testing is required, the owner should 

require that all testing be performed in accordance with the appropriate specifications for 

laboratory testing such as those codified in AASHTO and/or ASTM. Several tables are 

presented in this chapter that summarize various common tests for soils and rocks per 

AASHTO and ASTM standards. In order to assure that the results of laboratory testing are 

representative, several precautions must be taken before the tests themselves are performed. 

These precautions include: sample tracking, sample storage, sample handling to prevent 

sample disturbance, and sample selection. Discussion of each of these precautions follows. 

5.1.1 Sample Tracking 

Whether the laboratory testing is performed in-house or is subcontracted, samples will likely 

be assigned a laboratory identification number that differs from the identification number 

assigned in the field. A list should be prepared to match the laboratory identification number 

with the field identification number. This list can also be used to provide tracking 

information to ensure that each sample arrived at the lab. When laboratory testing is 

requested, both the field identification number and the laboratory identification number 

should be used on the request form. An example request form is shown in Figure 5-1. A 

spreadsheet or database program is useful to manage sample identification data. 

5.1.2 Sample Storage 

Undisturbed soil samples should be transported and stored so that that the moisture content is 

maintained as close as possible to the natural conditions (AASHTO T 207, ASTM D 4220 

and D 5079). Samples should not be placed, even temporarily, in direct sunlight. Shelby 

tubes should be stored in an upright position with the top side of the sample up in a humid 

room with relative humidity above 90%. 

Long-term storage of soil samples in Shelby tubes is not recommended. As storage time 

increases, moisture will migrate within the tube. Potential for disturbance and moisture 

migration within the sample will increase with time, and samples tested 30 days after their 

retrieval should be noted on the laboratory data sheet. Excessive storage time can lead to 

FHWA NHI-06-088 5 – Laboratory Tests 

Soils and Foundations – Volume I 5 - 2 December 2006 
additional sample disturbance that will affect strength and compressibility properties. 

Additionally, stress relaxation, temperature changes, and storage in a room with humidity 

below 90 percent may have detrimental effects on the samples. Long-term storage of soil 

samples should be in temperature- and humidity-controlled environments. The temperature 

control requirements may vary from sub freezing to ambient and above, depending on the 

environment of the parent formation. The relative humidity for soil storage normally should 

be maintained at 90 percent or higher to prevent moisture evaporation from the samples. 

Figure 5-1. Sample laboratory test request form (Note: only some tests are included in 

this sample form). 

FHWA NHI-06-088 5 – Laboratory Tests 

Soils and Foundations – Volume I 5 - 3 December 2006 
 

Long-term storage of soil samples in Shelby tubes is not recommended for another reason. 

During long term storage the sample tubes may corrode. Corrosion accompanied by 

adhesion of the soil to the tube may result in the development of such a large sidewall 

resistance that some soils may experience internal failures during extrusion. Often these 

failures cannot be seen by the naked eye; x-ray radiography (ASTM D 4452) will likely be 

necessary to confirm the presence of such conditions. If these samples are tested as 

“undisturbed” specimens, the results may be misleading. 

5.1.3 Sample Handling 

Careless handling of nominally undisturbed soil samples after they have been retrieved may 

cause major disturbances that could influence test results and lead to serious design and 

construction consequences. Samples should always be handled by experienced personnel in 

a manner that ensures that the sample maintains structural integrity and its natural moisture 

condition. Saws and knives used to prepare soil specimens should be clean and sharp. 

Preparation time should be kept to a minimum, especially where the maintenance of the 

moisture content is critical. Specimens should not be exposed to direct sun, freezing, or 

precipitation. 

5.1.4 Effects of Sample Disturbance 

As a soil sample is removed from the ground during a conventional soil investigation, its insitu effective stress condition is being changed. In addition, nominally undisturbed 

specimens taken from samples obtained from drilled boreholes will become disturbed as a 

result of the drilling itself, sampling, sample extrusion, and sample trimming to form a 

specimen for testing. These processes will also change the effective stress condition in the 

soil sample, i.e., the effective stress in the soil at the time after a sample is trimmed and 

prepared for testing is different from that of the same soil in the ground. Therefore the 

utmost care should be taken to minimize the effect of these processes in order for the results 

of laboratory tests to represent the in-situ soil behavior accurately. 

5.1.5 Specimen Selection 

The selection of representative specimens for testing is one of the most important aspects of 

sampling and testing procedures. Selected specimens must be representative of the formation 

or deposit being investigated. The geotechnical specialist should study the boring logs, 

understand the geology of the site, and visually examine the samples before selecting the test 

specimens. Samples should be selected on the basis of their color, physical appearance, 

structural features and an understanding of the disturbance of the samples. Specimens should 

be selected to represent all types of materials present at the site, not just the worst or the best. 

FHWA NHI-06-088 5 – Laboratory Tests 

Soils and Foundations – Volume I 5 - 4 December 2006 
Samples with discontinuities and intrusions may fail prematurely in the laboratory. The first 

inclination would be to test these samples. However, if these features are small and 

randomly located, they may not necessarily cause such failures in the field. Therefore 

samples having such local features should be noted, but not necessarily selected for testing 

since such samples may not be representative of the stratum in terms of its response to 

applied loads. 

Certain considerations regarding laboratory testing, such as when, how much, and what type, 

can be decided only by an experienced geotechnical specialist. The following minimal 

criteria should be considered when the scope of the laboratory testing program is being 

determined: 

•	 Project type (bridge, embankment, building, reconstruction or new construction, etc.) 

•	 Size of the project (geographic extent). 

•	 Loads to be imposed on the foundation soils (geometry, type, direction and 

magnitude). 

•	 Performance requirements for the project (e.g., settlement and lateral deformation 

limitations). 

•	 Vertical and horizontal variations in the subsurface profile as determined from boring 

logs and visual identification of subsurface material types in the laboratory. 

•	 Known or suspected peculiarities of subsurface strata at the project location (e.g., 

swelling soils, collapsible soils, organics, etc.) 

•	 Presence of visually observed intrusions, slickensides, fissures, concretions, etc. 

The selection of tests should be considered preliminary until the geotechnical specialist is 

satisfied that the test results are sufficient to develop reliable subsurface profiles and provide 

the parameters needed for design. 

5.2 LABORATORY TESTING FOR SOILS 

Table 5-1 provides a listing of commonly-performed soil laboratory tests. Tables 5-2 and 5-3 

provide a summary of typical soil index and performance tests, respectively. Additional 

information on these tests is provided in subsequent sections. 

FHWA NHI-06-088 5 – Laboratory Tests 

Soils and Foundations – Volume I 5 - 5 December 2006 
Table 5-1 

Commonly performed laboratory tests on soils (after FHWA, 2002a)

Test 

Category Name of Test 

A

Visual Identification 

Practice for Description and Identification of Soils (Visual-Manual Procedure) 

Practice for Description of Frozen Soils (Visual-Manual Procedure) 

Index Properties Test Method for Determination of Water (Moisture) Content of Soil by Direct Heating Method 

Test Method for Specific Gravity of Soils 

Method for Particle-Size Analysis of Soils Test Method for Classification of Soils for Engineering Purposes Test Method for Amount of Material in Soils Finer than the No. 200 (0.075 mm) Sieve Test Method for Liquid Limit, Plastic Limit, and Plasticity Index of Soils T 

Compaction Test Method for Laboratory Compaction Characteristics of Soil Using Standard Effort (12,375 ft. lbs/ft3) 

Test Method for Laboratory Compaction Characteristics of Soil Using Modified Effort (56,250 ft.lbs/ft3) 

Strength Properties 

Test Method for Unconfined Compressive Strength of Cohesive Soil Test Method for Unconsolidated, Undrained Compressive Strength of Cohesive Soils in Triaxial Compression 

Test Method for Consolidated, Undrained Compressive Strength of Cohesive Soils in Triaxial Compression 

Method for Direct Shear Test of Soils under Consolidated Drained Conditions 

Test Methods for Modulus and Damping of Soils by the Resonant-Column Method 

Test Method for Laboratory Miniature Vane Shear Test for Saturated Fine-Grained Clayey Soil Test Method for CBR (California Bearing Ratio) of Laboratory-Compacted Soils 

Test Method for Resilient Modulus of Soils 

Test Method for Resistance R-Value and Expansion Pressure of Compacted Soils 

Consolidation, Swelling, Collapse Properties 

Test Method for One-Dimensional Consolidation Properties of Soils 

Test Method for One-Dimensional Consolidation Properties of Soils Using Controlled-Strain Loading 

Test Methods for One-Dimensional Swell or Settlement Potential of Cohesive Soils 

Test Method for Measurement of Collapse Potential of Soils 

Permeability Test Method for Permeability of Granular Soils (Constant Head) 

Test Method for Measurement of Hydraulic Conductivity of Saturated Porous Materials Using a Flexible Wall Permeameter 

Corrosivity (Electrochemical) 

Test Method for pH for Peat Materials Test Method for pH of Soils Test Method for pH of Soil for Use in Corrosion Testing Test Method for Sulfate Content 

Test Method for Resistivity Test Method for Chloride Content 

Organic Content Test Methods for Moisture, Ash, and Organic Matter of Peat and Other Organic Soils 

FHWA NHI-06-088 Soils and Foundations – Volume I 5 - 6 
Table 5-2 

Methods for index testing of soils (after FHWA, 2002a) 

FHWA NHI-06-088 Soils and Foundations – Volume I 5 - 7 

Test Procedure ASTM and/or AASHTO Applicable Soil Types Applicable Soil Properties 

Moisture content, wn 

Dry soil in oven at 100 + 5 oC D 2216 

T 265 

Gravel, sand, silt, clay, peat 

eo, γ 

Unit weight and density 

Extract a tube sample; measure dimensions 

and weight; 

D 2216 

T 265 

Soils where 

undisturbed samples can be taken, i.e., silt, clay, peat 

γt, γdry, ρtot, ρdry, pt 

Atterberg 

limits, LL, 

PL, PI, SL, LI 

LL – Moisture content associated with 

closure of the groove at 25 blows of 

specimen in Casagrande cup PL – Moisture content associated with crumbling of rolled soil at 1/8-in (3mm) 

D 4318 

T 89 

T 90 

Clays, silts, peat; 

silty and clayey 

sands to determine whether SM or SC 

Soil classification and used in 

consolidation parameters 

Mechanical Place air dry material on a series of D 422 Gravel, sand, silt Soil classification 

sieve successively smaller screens of known 

opening size and vibrate to separate particles of a specific equivalent diameter 

T 88 

Wash sieve Flush fine particles through a U.S. No. 200 

(0.075 mm) sieve with water. 

C 117 

D 1140 

T 88 

Sand, silt, clay Soil classification 

Hydrometer Allow particles to settle, and measure 

specific gravity of the solution with time. 

D 422 

D 1140 

T 88 

Fine sand, silt, clay 

Soil classification 

Sand Equivalent 

Sample passing No. 4 (4.75 mm) sieve is separated into sand and clay size particles 

D 2419 

T 176 

Gravel, Sand, silt, clay 

Aggregate classification Compaction 

Specific gravity of solids 

The volume of a known mass of soil is compared to the known volume of water in a calibrated pyncnometer 

D 854 

D 5550 

T 100 

Sand, silt, clay, peat 

Used in calculation of eo 

Organic content 

After performing a moisture content test at 110 oC (230o F), the sample is ignited in a 

muffle furnace at 440 oC (824o F) to measure 

the ash content. 

D 2974 

T 194 

All soil types 

where organic matter is suspected to be a concern 

Not related to any specific 

performance parameters, but samples high in organic content will likely have high compressibility. 

Symbols used in Table 5-2 

eo: in-situ void ratio 

ρdry:dry density 

γdry:dry unit weight 

ρtot:total density 

γ: unit weight γt:total unit weight p

t: total vertical stress 
Table 5-3 

Methods for performance testing of soils (after FHWA, 2002a) 

Test Procedure Applicable Soil Types Soil Properties Limitatio

1-D oedometer Incremental loads are applied to a soil 

specimen confined by a rigid ring; deformation values are recorded with time; loads are typically doubled for each increment and applied for 24 hours each. 

Primarily clays and silts; granular soils can be tested, but typically are not. 

pc, OCR, 

Cc, Ccε, Cr, 

Crε, Cα, 

Cαε, cv, k 

Recommended for fine be useful index to othe

Constant rate of strain oedometer 

Loads are applied such that ∆u is between 3 and 30 percent of the applied vertical stress during testing 

Clays and silts; not applicable to free draining granular soils. 

pc, Cc, 

Ccε, Cr, 

Crε, cv, k 

Requires special testinreduce testing time sig

Unconfined compression (UC) 

A specimen is placed in a loading apparatus and sheared under axial compression with no confinement. 

Clays and silts; cannot be performed on granular soils or fissured and varved materials 

su,UC Provides rapid means t

shear strength, but distand moisture migration 

Unconsolidated undrained (UU) triaxial shear 

The specimen is not allowed to consolidate under the confining stress, and the specimen is loaded at a quick enough rate to prevent drainage. 

Clays and silts su,UU Sample must be nearly 

disturbance and rate efstrength. 

Isotropic consolidated drained compression (CIDC) 

The specimen is allowed to consolidate under the confining stress, and then is sheared at a rate slow enough to prevent build-up of pore water pressures. 

Sands, silts, clays φ′, c′, E Can be run on clay spe

consuming. Best triaxdeformation properties

Isotropic consolidated undrained compression (CIUC) 

The specimen is allowed to consolidate under the confining stress with drainage allowed, and then is sheared with no drainage allowed, but pore water pressures measured. 

Sands, silts, clays, peats φ′, c′, 

su,CIUC, E 

Recommended to meatest. Useful test to assparameters. Not recomdeformation properties

Direct shear The specimen is sheared on a forced 

failure plane at a constant rate, which is a function of the hydraulic conductivity of the specimen. 

Compacted fill materials; sands, silts, and clays 

φ′, φ′r Requires assumption o

Relatively easy to perf

FHWA NHI-06-088 Soils and Foundations – Volume I 5 - 8 
Test Procedure Applicable Soil Types Soil Properties Limitatio

Flexible Wall The specimen is encased in a Relatively low k Recommended for fine Permeameter membrane, consolidated, 

backpressure saturated, and 

permeability materials (k < 1x10-5 cm/s); clays & 

Backpressure saturatiostress needs to be prov

measurements of flow with time are silts must be at least an ordrecorded for a specific gradient. than that of the specim

inflow and outflow to 

Rigid Wall The specimen is placed in a rigid wall Relatively high k Need to control gradiePermeameter cell, vertical confinement is applied, permeability materials; grained soils. Monitor 

and flow measurements are recorded with time under constant head or 

sands, gravels, and silts 

falling head conditions. 

Symbols used in Table 5-3 φ′: peak effective stress friction angle OCR: overconsolidation ratio Ccε: modified com

φ′r residual effective stress friction angle cv: vertical coefficient of consolidation Cr: recompression c′: effective stress cohesion intercept E: Young’s modulus Crε: modified recosu: undrained shear strength k: hydraulic conductivity Cα: secondary com

pc: preconsolidation stress Cc: compression index Cαε: modified seco

FHWA NHI-06-088 Soils and Foundations – Volume I 5 - 9 
5.3 LABORATORY INDEX TESTS FOR SOILS 

5.3.1 General 

Data generated from laboratory index tests provide an inexpensive way to assess soil 

consistency and variability among samples collected from a site. Information obtained from 

index tests is used to select samples for engineering property testing as well as to provide an 

indicator of general engineering behavior. For example, a soil with a high plasticity index 

(PI) can be expected to have high compressibility, low hydraulic conductivity, and high swell 

potential. Common index tests discussed in this section include moisture content, unit 

weight (wet density), Atterberg limits, particle size distribution, visual classification, specific 

gravity, and organic content. Index testing should be conducted on each type of soil material 

on every project. Information from index tests should be assessed prior to a final decision 

regarding the specimens selected for subsequent performance testing. 

5.3.2 Moisture Content 

The moisture (or water) content test is one of the simplest and least expensive laboratory 

tests to perform. Moisture content is defined as the ratio of the weight of the water in a soil 

specimen to the dry weight of the specimen. Natural moisture contents (wn) of sands are 

typically 0 ≤ wn ≤ 20 %, whereas for inorganic and insensitive silts and clays, the typical 

range is 10 ≤ wn ≤ 40 %. However, for clays it is possible to have more water than solids 

(i.e., wn > 100%), depending upon the mineralogy, formation environment, and structure of 

the clay. Therefore, soft and highly compressible clays, as well as sensitive, quick, or 

organically rich clays, can exhibit water contents in the range of 40 ≤ wn ≤ 300 % or more. 

Moisture content can be tested a number of different ways including: (1) a drying oven 

(ASTM D 2216); (2) a microwave oven (ASTM D 4643); or (3) a field stove or blowtorch 

(ASTM D 4959). While the microwave or field stove (or blowtorch) methods provide a 

rapid evaluation of moisture content, potential errors inherent with these methods require 

confirmation of results obtained by using ASTM D 2216. The radiation heating induced by 

the microwave oven and the excessive temperature induced by the field stove may release 

water entrapped in the soil structure (adsorbed water) that would normally not be released at 

230o

 F (110o C), the maximum temperature specified by ASTM D 2216. Therefore, the 

microwave oven and field stove methods may yield greater values of moisture content than 

would occur from ASTM D 2216. 

Field measurements of moisture content often rely on a field stove or microwave due to the 

speed of testing. For control of compacted material, it is common to use a nuclear gauge 

FHWA NHI-06-088 5 – Laboratory Tests 

Soils and Foundations – Volume I 5 - 10 December 2006
 
(ASTM D 3017) in the field to assess moisture contents rapidly. Nuclear gage readings may 

indicate widely varying moisture contents for micaceous soils, i.e., soils containing a 

significant amount of mica particles. Results from nuclear techniques should be “calibrated” 

or confirmed by using the drying oven method (ASTM D 2216). 

Moisture contents of soils as determined from in-situ moisture content tests may be altered 

during sampling, sample handling, and sample storage. Because the top end of the sample 

tube may contain water or collapse material from the borehole, moisture content tests should 

not be performed on material near the top of the tube. Also, as storage time increases, 

moisture will migrate within a specimen and lead to altered values of moisture content. If the 

sample is not properly sealed, moisture loss through drying of the sample will likely occur. 

5.3.3 Unit Weight 

The terms density (ρ) and unit weight (γ) are often incorrectly used interchangeably. The 

correct usage is that density implies mass while unit weight implies weight measurements. 

Density and unit weight are related through the gravitational constant (g) as follows: γ = ρg. 

In this document they will be referenced as “density (unit weight)” if the usage is 

independent of the specific definition. 

In the laboratory, soil unit weight and mass density are easily measured on tube (undisturbed) 

samples of natural soils. The moist (total) mass density is ρt = Mt/Vt, where Mt is the total 

mass of the soil sample including the mass of the moisture in the pores and Vt is the total 

volume of the soil sample. Similarly the dry mass density is given by ρd = Ms/Vt , where Ms 

is the mass of the solid component of the soil sample and Vt is the total volume of the soil 

sample. Likewise, the moist unit weight is γt = Wt/Vt, where Wt is the total weight including 

the weight of the water in the pores and Vt is the total volume of the soil sample. Similarly, 

the dry unit weight is defined as γd = Ws/Vt where Ws is the weight of the solid component of 

the soil sample and Vt is the total volume of the soil sample. The relationship between the 

total and dry mass density and unit weight in terms of natural moisture content, w, is given 

by: 

ρt ρd = 5-1 1 + w 

Since γ = ρg the relationship between total and dry unit weight is given by: 

FHWA NHI-06-088 5 – Laboratory Tests 

Soils and Foundations – Volume I 5 - 11 December 2006
 
γ t γ d = 5-2

1 + w 

 

Field measurements of soil mass density (unit weight) are generally restricted to shallow 

surface samples such as those obtained during placement of compacted fills. In those cases, 

field measurements of soil mass density (unit weight) can be accomplished by using drive 

tubes (ASTM D 2937), the sand cone method (ASTM D 1556), or a nuclear gauge (ASTM D 

2922). To obtain unit weights or mass densities with depth, either high-quality thin-walled 

tube samples must be obtained (ASTM D 1587), or relatively expensive geophysical logging 

by gamma ray techniques (ASTM D 5195) can be employed. 

Table 5-4 presents typical unit weights along with a range of void ratios for a variety of soils. 

5.3.4 Particle Size Distribution 

Particle size distributions by mechanical sieve and hydrometer analyses are useful for soil 

classification purposes. Procedures for grain size analyses are contained in ASTM D 422 

and AASHTO T88. Testing is accomplished by shaking air-dried material through a stack of 

sieves having decreasing opening sizes. Table 2-3 in Chapter 2 listed U.S. standard sieve 

sizes and their associated opening sizes. Each successive screen in the stack has a smaller 

opening to capture progressively smaller particles. The amount retained on each sieve is 

collected, dried and weighed to determine the percentage of material passing that sieve size. 

An example of how to determine the grain size distribution from sieve data is shown in 

Figure 5-2. The grain size distribution curve corresponding to the data in Figure 5-2 is 

presented in Figure 5-3. 

Testing of the finer grained particles is accomplished by suspending the chemically dispersed 

particles in a water column and measuring the change in the specific gravity of the liquid as 

the particles fall from suspension. This part of the test is commonly referred to as a 

hydrometer analysis. 

Obviously, obtaining a representative specimen is an important aspect of this test. When soil 

samples are dried or washed for testing, it may be necessary to break up the soil clods. Care 

should be taken to avoid crushing of soft carbonate or sand particles. If the soil contains a 

substantial amount of fibrous organic materials, these may tend to plug the sieve openings 

during washing. The material settling over the sieve during washing should be constantly 

stirred to avoid plugging. 

FHWA NHI-06-088 5 – Laboratory Tests 

Soils and Foundations – Volume I 5 - 12 December 2006
 
Table 5-4 

Typical particle sizes, uniformity coefficients, void ratios and unit weights (from Kulhawy and M

FHWA NHI-06-088 Soils and Foundations – Volume I 5 - 13 

Soil Type 

Approximate 

Particle Size, mm 

Uniformity 

Coefficient 

Void Ratio 

Norm

Dry 

γdry/γ

Dmax min 60 60/D10 max min Min 

Uniform granular soil D

Equal spheres (theoretical) - D - - D 1.0 e 0.92 e 0.35 - Standard Ottawa sand 0.84 0.59 0.67 1.1 0.80 0.50 1.47 

Clean, uniform sand - - - 1.2 to 2.0 1.00 0.40 1.33 Uniform, inorganic silt 0.05 0.005 0.012 1.2 to 2.0 1.10 0.40 1.28 

Well-graded granular soil 

Silty sand 2.0 0.005 0.02 5 to 10 0.90 0.30 1.39 Clean, fine to coarse sand 2.0 0.05 0.09 4 to 6 0.95 0.20 1.36 

Micaceous sand - - - - 1.20 0.40 1.22 

Silty sand and gravel 100 0.005 0.02 15 to 300 0.85 0.14 1.43 

Silty or sandy clay 2.0 0.001 0.003 10 to 30 1.80 0.25 0.96 

Gap-graded silty clay with gravel or larger 250 0.001 - - 1.00 0.20 1.35 

Well-graded gravel, sand, silt, and clay 250 0.001 0.002 25 to 1,000 0.70 0.13 1.60 

Clay (30 to 50% < 2μ size) 0.05 0.5μ 0.001 - 2.40 0.50 0.80 

Colloidal clay (over 50% < 2μ size) 0.01 10Å - - 12.00 0.60 0.21 

Organic silt - - - - 3.00 0.55 0.64 

Organic clay (30 to 50% < 2μ size) - - - - 5.40 0.70 0.48 

Note: γw= 62.4 pcf (9.80 kN/m3); μ = 10-3 mm; Å : Angstrom = 10-7 mm 
Figure 5-2. Example grain size distribution based 

on sieve analysis (Jumikis, 1962). 

Figure 5-3. Grain size distribution curve based on 

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Particle size testing is relatively straightforward, but the results can be misleading if 

procedures are not performed correctly and/or if equipment is not maintained in good 

working condition. If the sieve screen is distorted, large particles may be able to pass 

through sieve openings that typically would retain the particles. Material lodged within the 

sieve from previous tests could become dislodged during shaking, thereby increasing the 

weight of material retained on the following sieve. Therefore, sieves should be cleaned 

thoroughly after each test. A wire brush may distort finer sieve meshes during cleaning, so a 

plastic brush should be used to clean the U.S. No. 40 (0.425 mm) sieve and finer. Openings 

of fine mesh No. 200 sieve (0.075 mm)) are easily distorted as a result of normal handling 

and use. Therefore, fine-mesh sieves should be replaced often. A simple way to determine 

whether sieves should be replaced is to examine the stretch of the sieve fabric on its frame 

periodically. The fabric should remain taut; if it sags, it has been distorted and should be 

replaced. A common cause of serious errors is the use of “dirty” sieves. Some soil particles, 

because of their shape, size or adhesion characteristics, have a tendency to lodge in the sieve 

openings. This is especially true of the fine mesh sieves. 

Representative samples of fine-grained soils (i.e., samples containing more than 50% of 

particles with diameter less than the U.S. No. 200 sieve size (075 mm ) should not be oven 

dried prior to testing because some particles may cement together leading to a calculated 

lower fines content from mechanical sieve analyses than is actually present. When finegrained particles are a concern, the wash sieve method (ASTM D 1140) should be performed 

to assess the fines content. 

If the clay-size content is an important parameter, hydrometer analyses should be performed 

even though the hydrometer test provides only approximate results due to oversimplified 

assumptions. However, the results can still be used as a general index of silt and clay-size 

content. Depending upon the chemical makeup of the fine grained particles, the traditional 

sodium hexametaphosphate solution used to disperse the clay-size particles may not provide 

adequate dispersion. If the clay-size particles are not dispersed, the hydrometer data leads to 

the interpretation of a lower than actual clay-size content. In some cases the concentration of 

the dispersing agent may need to be increased or a different dispersing agent may need to be 

used. If the sieve and hydrometer analyses are performed correctly, the gradation curve 

should be continuous over a range that includes all particle sizes. 

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5.3.4.1 Sand Equivalent 

The sand equivalent test is a rapid test to show the relative proportions of fine dust or 

claylike materials in aggregate (or soils). A sample of aggregate passing the No. 4 sieve 

(4.75-mm) sieve and a small amount of flocculating solution are poured into a graduated 

cylinder and are agitated to loosen the claylike coatings from the sand particles. The sample 

is then irrigated with additional flocculation solution forcing the claylike material into 

suspension above the sand. After a prescribed sedimentation period, the height of flocculated 

clay and height of sand are determined. The sand equivalent is determined from the ratio of 

the height of the sand to height of the clay and expressed as a percentage. Cleaner aggregates 

will have higher sand equivalent values. For asphalt pavements, agencies often specify a 

minimum sand equivalent around 25 to 35 (Roberts, et al., 1996). Higher values are used in 

case of compacted structural fill which may support structures (see Section 8.6). 

5.3.5 Atterberg Limits 

The Atterberg limits of a fine grained soil represent the moisture content at which the 

physical state of the soil changes. The tests for the Atterberg limits are referred to as index 

tests because they serve as an indication of several physical properties of the soil, including 

strength, permeability, compressibility, and shrink/swell potential. These limits also provide 

a relative indication of the plasticity of the soil, where plasticity refers to the ability of a silt 

or clay to retain water without changing state from a semi-solid to a viscous liquid. In 

geotechnical engineering practice, the Atterberg limits generally refer to the liquid limit (LL), 

plastic limit (PL), and shrinkage limit (SL). The limits were defined and discussed in 

Chapter 2. In this chapter the definition is extended further in terms of quantifiable 

parameters that permit their measurements in the laboratory. These quantifiable definitions 

are as follows: 

•	 Liquid Limit (LL) - This limit represents the moisture content at which any increase 

in moisture content will cause a plastic soil to behave as a viscous liquid. The LL is 

defined as the moisture content at which a standard groove cut in a remolded sample 

will close over a distance of 1⁄2-inch (13 mm) at 25 blows of the liquid limit device 

(Figure 5-4). The test is performed on material passing a US Standard No. 40 sieve 

(0.425 mm). During the test the material is brought to various moisture contents, 

usually by adding water. The plot of moisture contents vs. blows required to close 

the groove is called a “flow curve” and the value of the liquid limit moisture content 

is obtained from the flow curve at 25 blows. 

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Figure 5-4. Some of the equipment used for Atterberg limits testing of soil. 

•	 Plastic Limit (PL) - This limit represents the moisture content at which the transition 

between the plastic and semisolid state of a soil occurs. The PL is defined as the 

moisture content at which a thread of soil just crumbles when it is carefully rolled out 

by hand to a diameter of 1/8-inch (3 mm). 

•	 Shrinkage Limit (SL) – This limit represents the moisture content corresponding to 

the change between the semisolid to solid state of the soil. The SL is also defined as 

the moisture content at which any further reduction in moisture content will not result 

in a decrease in the volume of the soil. 

Based on the above index values, there are two useful related indices, namely, the Plasticity 

Index (PI) and the Liquidity Index (LI), which were defined in Chapter 2 as follows: 

PI = LL - PL 2-11 

w − PL LI =	 2-12

PI 

where w is the natural (in-situ) water content of the soil. Numerous engineering correlations 

have been developed that relate PI and LI to clay soil properties, including undrained and 

drained strength to PI and compression index to LI. 

Another useful index proposed by Skempton (1953) based on the proportion of clay and PI is 

known as the “Activity Index.” The activity index of a clay soil is denoted by A and is 

generally defined as follows: 

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PI A = 5-3

CF 

where CF is the clay fraction is usually taken as the percentage by weight of the soil with a 

particle size less than 0.002 mm. Clays with 0.75 < A < 1.25 are classified as “normal” clays 

while those with A < 0.75 are “inactive” and A > 1.25 are “active.” Values of activity index, 

A, can be correlated to the type of clay mineral that, in turn, provides important information 

relative to the expected behavior of a clay soil. A clay soil that consists predominantly of the 

clay mineral montmorillonite behaves very differently from a clay soil composed 

predominantly of kaolinite. Figure 5-5 also shows the activities of various clay minerals and 

their location on the Casagrande’s plasticity chart. The symbol for the activity index (A) in 

Figure 5-5 should not be confused with the “A-line” also shown in the figure. 

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A=1.5 (Calcium) 

A=4 to 7 (Sodium) 

A=0.3 to 0.5 

A=0.1 (hydrated) 

A=0.5 (dehydrated) 

A=0.5 to 1.3 

Activity Index, A, of other minerals 

Attapulgite 0.5-1.2 Mica (muscovite) 0.2 

Allophane 0.5-1.2 Quartz 0 

Calcite 0.2 

Figure 5-5. Location of clay minerals on the Casagrande Plasticity Chart and Activity 

Index values (after Skempton, 1953, Mitchell, 1976, Holtz and Kovacs, 1981). 
Modified Activity Index, Am: Based on their studies regarding the swell potential of 

compacted natural and artificial clay soils, Seed et al. (1962) proposed that for natural clay 

soils compacted as per the requirements of ASTM D 698 and Atterberg limits determined by 

ASTM D 4318 (AASHTO T 89, T 90), a Modified Activity Index, Am, defined as follows is 

more appropriate: 

PI Am = 5-4

CF − 5 

The above definition is used to define the swell potential of soils (see Section 5.7). 

5.3.5.1 Significance of the “A-line” and “U-line” on Plasticity Chart 

As shown in Figure 4-3 in Chapter 4, the equation for the A-line and U-line are: 

A − line : PI = 0.73 (LL − 20 ) 5-5 

U − line : PI =0.9(LL−8) 5-6 

The A-line generally separates soils whose behavior is more claylike (points plotting above 

the A-Line) from those that exhibit a behavior more characteristic of silt (points plotting 

below the A-line). The A-line also separates organic (below) from inorganic (above) soils. 

The LL = 50 line generally represents the dividing line between silt, clay and organic 

fractions of the soil that exhibit low plasticity (LL<50) and high plasticity (LL>50). The Uline shown in Figure 5-5 represents the upper range of PI and LL coordinates that have been 

found for soils. When the limits of any soil plot above the U-line, the results should be 

considered spurious and the tests should be rerun. Note that in Figure 5-5 the clay mineral 

montmorillonite plots well above the A-line and just below the U-line. If a soil plots in this 

range, it probably contains a significant amount of the clay mineral montmorillonite that 

expands in presence of water. 

5.3.6 Specific Gravity 

The specific gravity of solids (Gs) is a measure of solid particle density and is referenced to 

an equivalent volume of water. Specific gravity of solids is defined as Gs = (Ms/Vs)/ ρd 

where Ms is the mass of the soil solids and Vs is the volume of the soil solids and ρd is the 

mass density of water = 1,000 kg/m3

 or 1 Mg/m3 . This formulation represents the 

theoretically correct definition of specific gravity and can be rewritten as Gs = ρs / ρd. 

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However, since γ = ρg the gravitational constant appears in both the numerator and 

denominator of the expression and the equation for Gs can also be given as Gs = γs/γw where 

γs = unit weight of solid particles in the soil mass and γw = unit weight of water = 62.4 pcf 

(1,000 kg/m3

 or 1 Mg/m3). 

The typical values of specific gravity of most soils lie within the narrow range of Gs = 2.7 ± 

0.1. Exceptions include soils with appreciable organics (e.g., peat), ores (e.g., mine tailings), 

or calcareous (high calcium carbonate content) constituents (e.g., caliche). It is common to 

assume a reasonable Gs value within the range listed above for preliminary calculations. 

Laboratory testing by AASHTO T100 or ASTM D 854 or D 5550 can be used to confirm the 

magnitude of Gs, particularly on projects where little previous experience exists and 

unusually low or high unit weights are measured. 

5.3.7 Organic Content 

A visual assessment of organic materials may be very misleading in terms of engineering 

analysis. Laboratory test method AASHTO T194 or ASTM D 2974 should be used to 

evaluate the percentage of organic material in a specimen where the presence of organic 

material is suspected based on field information or from previous experience at a site. The 

test involves weighing and heating a previously dried sample to a temperature of 824°F 

(440°C) and holding this temperature until no further change in weight occurs. At this 

temperature, the organics in the sample turn to ash and the sample is re-weighed. Therefore, 

with the assumption that the weight of the ash is negligible, the percentage of organic matter 

is the ratio of the difference in weight before and after heating the sample to 824°F (440°C) 

to the weight of the original dried sample. The sample used for the test can be a previously 

dried sample from a moisture content evaluation. Usually organic soils can be distinguished 

from inorganic soils by their characteristic odor and their dark gray to black color. In 

doubtful cases, the liquid limit should be determined for an oven-dried sample (i.e., dry 

preparation method) and for a sample that is not pre-dried before testing (i.e., wet preparation 

method). If drying decreases the value of the liquid limit by about 30 percent or more, the 

soil may usually be classified as organic (Terzaghi, et al., 1996). 

Soils with relatively high organic contents have the ability to retain water. Water retention 

may result in higher moisture content, higher primary and secondary compressibility, and 

potentially higher corrosion potential. Organic soils may or may not be relatively weak 

depending on the nature of the organic material. Highly organic fibrous peats can exhibit 

high strengths despite having a very high compressibility. In some instances such soils may 

even exhibit tensile strength. 

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5.3.8 Electro Chemical Classification Tests 

Electro chemical classification tests provide the geotechnical specialist with quantitative 

information related to the aggressiveness of the soil conditions with respect to corrosion and 

the potential for deterioration of typical foundation materials. Electro chemical tests include 

determination of pH, resistivity, sulfate ion content, sulfides, and chloride ion content. 

Depending on the application, limits of these electro chemical properties are established 

based on various factors such as corrosion rates for metals and disintegration rates for 

concrete. Tests to characterize the aggressiveness of a soil environment are important for 

design applications that include metallic elements, especially for ground anchors comprised 

of high strength steel and for metallic reinforcements in mechanically stabilized earth walls. 

ASTM and AASHTO test procedures are listed under “Corrosivity (Electrochemical)” in 

Table 5-1. 

5.3.9 Laboratory Classification 

In addition to field identification (ASTM D 2488), soils should be classified in the laboratory 

by using the Unified Soil Classification System (USCS) in accordance with ASTM D 2487 

or by the AASHTO soil classification system in accordance with AASHTO T 145. These 

two systems were discussed in Chapter 4. The USCS will be used throughout the remainder 

of this document. Classification in the laboratory occurs in a controlled environment and 

more time can be spent on this classification than the identification exercise performed in the 

field. Laboratory and/or field identification is also important so that defects and features of 

the soil can be recorded that would not typically be noticed from index testing or standard 

classification. Some of the features include degree of calcium carbonate cementation, mica 

content, joints, and fractures. 

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5.4 CONSOLIDATION TESTING 

5.4.1 Process of Consolidation 

As discussed in Chapter 2, consolidation is a time-dependent decrease in the volume of a soil 

mass under applied loading. In highway design, static loading is represented by the 

permanent load placed on the soil by embankments and structures. Depending on the 

configuration of the load and the subsurface conditions, the stress increase due to the 

externally applied loads may extend below the water table where all the voids are filled with 

water. An applied load will cause the soil grains to readjust to a more compact position to 

carry the load. This readjustment cannot take place until the water, which is incompressible, 

escapes from the voids. 

As discussed in Chapter 2, the rate of the readjustment of the soil particles is a function of the 

void size, which controls the rate at which the water can escape from the voids. The 

settlement associated with the readjustment of the soil particles due to migration of water out 

of the voids is known as primary consolidation. 

The amount of primary consolidation will depend on the initial void ratio of the soil. The 

greater the initial void ratio, the more water that can be squeezed out, and the greater the 

primary consolidation. The rate at which primary consolidation occurs is dependent on the 

rate at which the water is squeezed out of the soil voids. Secondary compression occurs 

after primary consolidation is complete. Secondary compression occurs under constant load. 

It is caused by the soil particles reorienting or deforming under constant load at a very slow 

rate. This process is known as “creep” and it occurs in most soils when they are subjected to 

long-term applied loads. Therefore, secondary compression is also a time-dependent process. 

However, secondary compression is not dependent on water being squeezed out of the soil as 

is consolidation. That is why it is called “secondary compression” and not “secondary 

consolidation.” Primary consolidation accounts for the major portion of settlement in 

saturated fine-grained soils. Primary consolidation and secondary compression both 

contribute significantly to settlements in organic soil. 

Some natural deposits of fine-grained soils experienced compression in geologic history due 

to the weight of glaciers, due to the weight of overlying soil that has been eroded, or due to 

desiccation. Since their void ratios were substantially reduced in the past by these processes, 

these soils are less compressible today. Such soils are called “preconsolidated” or 

“overconsolidated” since they have been subjected to greater stresses in the past than exist at 

present. This concept is important because overconsolidated soils can be reloaded such as by 

the load from an embankment or bridge substructure without settling appreciably until the 

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currently applied load exceeds the preconsolidation load. Saturated fine-grained soil 

deposits, which have never consolidated under loads other than the current loads, are called 

“normally consolidated.” On the other end of the spectrum, soils whose present loading 

induces stresses in the soil that are greater than the maximum effective stress they have 

experienced in the past are called “under consolidated.” This means that the consolidation 

process under the existing loading is on-going and the soil will continue to consolidate until 

that process is complete, even if no additional loads are applied. 

5.4.2 Consolidation Testing 

To predict the amount of consolidation in saturated fine-grained and organic soils, adequate 

testing must be performed. An undisturbed soil sample should be obtained in the field with a 

Shelby tube sampler. The oedometer or one-dimensional consolidometer is the primary 

laboratory equipment used to evaluate consolidation and settlement potential of fine-grained 

soils. A consolidation test is typically performed on a specimen obtained from an 

undisturbed sample retrieved from the deposit of fine-grained soils to evaluate the 

consolidation characteristics of the soil and define the settlement-time relationship of the insitu soils under proposed foundation loads. The equipment for a consolidation test includes: 

1.	 A loading device that applies a vertical load to the soil specimen, 

2.	 A metal ring (fixed or free) that laterally confines the soil specimen and restricts 

deformation to the vertical direction only (i.e., only one-dimensional compression is 

modeled), 

3.	 Porous discs placed on the top and bottom of the sample to allow the sample to drain, 

4.	 A dial indicator or linear variable differential transducer (LVDT) to measure vertical 

displacement. Properly calibrated, each device should provide the same accuracy, but 

the electronic output of an LVDT can be incorporated into an automated recording 

system for quicker, more efficient, and higher resolution readings. 

5.	 A timer to assess the duration of loading increments. Monitoring of time for manual 

systems can be accommodated by use of a wall clock with a second hand. The 

internal clock of a computer is used for automated systems 

6.	 A surrounding container to permit the specimen to remain submerged during the test. 

Figure 5-6 shows a schematic of a consolidation test. The consolidation-loading device may 

be a weighted lever arm as shown in Figure 5-7b, a pneumatic device, or an automated 

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loading frame as shown in Figure 5-7c. Automated loading frames are recommended for use 

in production testing because they provide the most flexibility in testing options. The 

pneumatic device provides flexibility in loads and load increment ratios (LIR) that can be 

applied during testing. A weighted lever arm provides a robust, relatively simple system for 

consolidation testing, however, because data are generally recorded manually, it is difficult to 

expedite testing or vary the loading schedule since data reduction cannot typically be 

performed in real time. 

Legend:
 

Sides fixed against 

lateral movement Represents top drainage 

plate or disc. Often 

there is also a bottom Sample 

Axial Stress 

drainage plate or disc 

Figure 5-6. Schematic of a consolidation test. 

Consolidation cells may be either fixed ring or floating ring. Friction and drag are created in 

the ring as the specimen compresses in relation to the ring. In a fixed ring test the sample 

compresses from the top only, potentially resulting in high incremental side shear forces. In 

a floating ring test the sample compresses from the top and bottom thus providing the 

advantage of minimizing drag forces. However, the floating ring method has the following 

disadvantages: it is more difficult to set up; it has the potential for sidewall leakage that 

would result in an inaccurate assessment of the rate of consolidation, and soil may squeeze 

out near the junction of the sidewall and the bottom porous disc. Because of these 

disadvantages, the fixed ring method is most commonly used. 

5.4.3 Procedures 

The consolidation properties of fine-grained soils are evaluated in the laboratory by using the 

one-dimensional consolidation test. The most common laboratory method is the incremental 

load (IL) method (ASTM D 2435). The weighted lever arm oedometer shown in Figure 5-7b 

is commonly used for performing the procedure. The automated load-frame apparatus shown 

in Figure 5-7c provides higher quality test results compared to the weighted lever apparatus. 

High-quality undisturbed samples obtained by using Shelby tubes (ASTM D 1587), piston 

samplers, or other special samplers are preferred for laboratory consolidation tests. 

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 (a) (b) 

(c) 

Figure 5-7. (a) Components of consolidation test equipment, (b) Weighted lever arm - 

incremental load consolidation apparatus, (c) Automated load-frame and computerized 

consolidation apparatus (Photographs courtesy of GeoComp Corporation). 

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5.4.4 Presentation and Understanding the Consolidation Test Results 

The consolidation test should be run in such a way that sufficient time is allowed for the 

applied pressure (total stress) increment to be transmitted from the pore water, where it acts 

initially as a excess pore water pressure, to the soil structure where it ultimately becomes an 

applied effective stress increment. The time it takes for this transfer to occur is the basis for 

the process being called “consolidation” and not “compression.” Therefore, the effective 

stress corresponding to the applied pressure is generally plotted versus void ratio. The 

resulting “consolidation curve” permits an evaluation of the preconsolidation pressure and 

values for other parameters pertaining to the consolidation characteristics of the soil sample. 

Plots of void ratio versus effective pressure on arithmetic and logarithmic scales are shown in 

Figure 5-8. The semi log plot is more widely used in practice and will be used in subsequent 

sections of this manual. The consolidation curve on the void ratio versus semi log pressure 

plot is commonly referred to as the “e-log p” relationship. As discussed in detail in Chapter 

7 and as shown on Figure 5-8, the slope of the loading portion of the e-lop p curve is called 

the compression index, which is denoted by the symbol Cc. The slope of the re-load portion 

of the e-log p curve is called the re-compression index; it is denoted by the symbol Cr. 

Some geotechnical specialists prefer to use a plot of percent strain versus log of pressure 

instead of the e-log p plot. In this case the slope of the virgin compression portion of the 

consolidation curve is called the modified compression index denoted by the symbol Ccε and 

the slope of the rebound portion of the curve is called the modified recompression index 

denoted by the symbol Crε. The modified indices reflect the relationship between strain and 

void ratio, i.e., strain (ε) = ∆e/(1+eo). Therefore, to convert the strain-based indices (Ccε and 

Crε) to the void-ratio-based indices (Cc and Cr) multiply the strain based values by (1 + eo). 

Void-ratio-based values (e-log p) will be used in the remainder of this manual. 

Analysis of consolidation test data allows the engineer to determine: 

1. Initial Void Ratio (eo) 

The value of the initial void ratio is very important because it defines the amount of 

void space at the start of the loading. It is this initial void space that will be reduced 

as the water is squeezed out of the voids with time. The initial void ratio eo is a key 

parameter used in settlement computations to determine the magnitude of settlement. 

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av 

Figure 5–8. Consolidation test relationships (after NAVFAC, 1986a). 

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2. Preconsolidation Pressure (pc) 

The e-log p relationship generally displays a distinct break at approximately the 

maximum past effective stress (pc). The graphical technique developed by 

Casagrande (1936) is generally used to determine the value of pc, which is known as 

preconsolidation pressure. The Casagrande procedure is included in the middle 

portion of Figure 5-8 and is discussed in detail in Chapter 7. 

The maximum effective stress to which a soil has been loaded in the past will have a 

major influence on the amount of settlement to be expected under a proposed 

loading. In fact, 10 times more settlement may occur in a normally consolidated soil 

than a preconsolidated soil for equal load increments up to the preconsolidation 

pressure. Values of preconsolidation pressure should be carefully established for the 

entire depth of the fine-grained soil deposit under consideration. Normally, a 

minimum, maximum and most probable value of pc will be determined from 

laboratory test results and plotted as a range with depth. 

3. Compression Index (Cc) 

The slope of the consolidation curve beyond pc is called the compression index (Cc). 

It is a measure of the load-deformation characteristic of the soil during “virgin” 

compression. 

4. Recompression Index (Cr) 

An unload/reload segment of the consolidation curve is also shown in Figure 5-8. 

The slope of the reload curve is called the recompression index (Cr). It is a measure 

of the load-deformation characteristic of the soil upon reloading after some amount 

of load release. As is obvious in Figure 5-8, the slope of the reload portion of the 

consolidation curve is not as steep as the slope of the virgin portion of the curve 

since the void ratio change accompanying the virgin loading is unrecoverable. 

Figure 5-8 also shows that if, upon reloading, the applied pressure exceeds the 

pressure from which the soil was unloaded, the slope of the reload curve reverts back 

to the virgin compression slope, Cc. In general, Cc ≈ 10 Cr. 

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5. Coefficient of Consolidation (cv) 

 The coefficient of consolidation is an indicator of the rate of drainage during 

consolidation. The value may be determined by the t50 (log time) method or the t90 

(square root of time) method. Both of these methods are described in Chapter 7 

(Approach Roadway Deformations). As shown in the bottom portion of Figure 5-8, 

the compression-log time curve for a given load increment is used to determine the 

coefficient of consolidation (cv), which is a measure of the time rate of primary 

consolidation. The value of cv is determined for each load increment. These values 

are sometimes plotted on a separate axis below the consolidation curve . 

6. Secondary Compression Index (Cα) 

Of great importance in organic materials, secondary compression may account for 

the majority of settlement that takes place over a long period of time in such soils. 

The compression-log time curve for a given load increment is used to determine the 

secondary compression index (Cα), which is basically the slope of the curve over one 

log cycle beyond the time required for primary consolidation (t100) as shown in the 

bottom portion of Figure 5-8. 

7. Effects of Sample Disturbance on Consolidation Test Results 

 

 The influence of sample disturbance on consolidation test results is shown on Figure 

5-8 by the dashed lines. The dash lines indicate that disturbance: 

a.	 Eliminates the distinct break in the e-log p curve at the preconsolidation pressure 

(pc). 

b.	 Lowers the estimated value of the preconsolidation pressure (pc) and the 

measured value of the compression index (Cc). 

c.	 Decreases the measured values of cv. 

d.	 Increases the recompression index (Cr). 

e.	 Decreases the secondary compression index (Cα). 

 

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 The general effects of disturbance are (a) under- or over-prediction of the 

magnitude of expected settlement and (b) over-prediction of the time for its 

occurrence. 

The importance of the consolidation test results as applied to design is summarized below. 

The test results may be applied to project design after a series of tests have been completed to 

represent the total depth of the fine-grained soil deposit. The two most important predictions 

are: 

1. The amount of settlement. The value is determined by analyzing the consolidation 

curve between the existing overburden pressure and the final pressure induced by the 

highway load at various depths. The amount of settlement may vary dramatically 

depending upon the maximum past pressure to which the soil has been loaded. The 

total amount of long-term settlement should include an estimate of settlement due to 

secondary compression, especially for times past the time for 100% primary 

consolidation if that is less than the design life of the constructed facility. 

 2. The time for settlement. The time for primary consolidation to occur may be 

estimated from the results of the compression versus time plots at loads between the 

overburden pressure and final pressure induced by the applied load. The important 

factors in the settlement-time relationship are: 

 (a) Time required is proportional to the square of the longest distance required for 

water to drain from the deposit. This distance is the thickness of the layer if 

water drains in one direction only (generally vertically upward to the surface), 

and one-half the layer thickness if more permeable soils exist above and 

below the consolidating layer. 

 

 (b) Time required for consolidation varies inversely with the coefficient of 

consolidation. 

 

 (c) Rate of settlement decreases as time increases. 

Settlement computations based on consolidation test results are demonstrated in Chapter 7 

(Approach Roadway Deformations). 

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5.4.5 Comments on the Consolidation Tests 

The consolidation test results are necessary to assess the consolidation properties of the soil. 

As will be shown in subsequent sections of this document, the consolidation test is one of the 

most important tests for fine-grained soil as it provides data regarding stress history and 

compressibility. It is important to consider all laboratory testing variables and their potential 

effects on the values of soil properties computed from the test results. Information that will 

need to be provided to a laboratory for a consolidation test includes the loading schedule (i.e., 

magnitude and duration of loads). It is important to evaluate the loading schedule to be used, 

especially the duration of loading since time is required for the applied total stress increment 

to be transferred from the pore water to the soil structure so that it becomes an effective 

stress acting on the soil mass. Important issues related to consolidation tests are discussed 

below. 

•	 Loading Sequence: The loading sequence selected for a consolidation test will 

depend on the type of soil being tested and the particular application being considered 

for the project (e.g., embankment, shallow foundation). The selection of a loading 

sequence should never be left to the discretion of the laboratory. As an example, if 

the clay soil is heavily overconsolidated, it is possible that a laboratory-determined 

maximum load for the consolidation test will not be sufficient to exceed pc. 

•	 Range of Applied Loads: The range of applied loads for the test should well exceed 

the effective stresses that are required for settlement analyses. This range should 

cover the smallest and largest effective stresses anticipated in the field and will 

depend on depth, foundation loads, and excavations. The anticipated 

preconsolidation stress should be exceeded by at least a factor of four during the 

laboratory test. If the preconsolidation stress is not significantly exceeded during the 

loading schedule, pc, and Cc (or Ccε) may be underestimated due to specimen 

disturbance effects. 

•	 Load Increment Ratio (LIR): By definition the LIR= ∆σ/σinitial where ∆σ is the 

incremental stress and σinitial is the previous stress. A LIR=1 corresponds to a 

doubling of the vertical stress applied to the specimen at each successive load 

increment during a consolidation test. A LIR of 1 is commonly used for most tests. 

Experience with soft sensitive soils suggests that as the stress approaches the value of 

pc, a smaller LIR will facilitate a better estimate of pc. Typically, laboratories provide 

a unit cost for a consolidation test that may be based on 6 to 8 load increments with a 

separate cost for each additional increment. 

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•	 Unload-Reload Cycle: It is recommended that an unload-reload cycle be performed, 

especially for cases where accurate settlement predictions are required, specifically to 

obtain a value for Cr. Since most samples will inevitably be somewhat disturbed, a Cr 

value based on the initial loading of a consolidation test sample will be greater than 

that for an undisturbed sample, resulting in an overestimation of settlements in the 

overconsolidated region. A value of Cr based on an unload-reload cycle is more 

likely to be representative of the actual behavior of the soil in the overconsolidated 

region. 

•	 Duration of Load Increment: The duration of each load increment should be 

selected to ensure that the sample is approximately 100 percent consolidated prior to 

application of the next load increment. For relatively low to moderate plasticity silts 

and clays, durations of 3 to 12 hours will be appropriate for loads in the normally 

consolidated range. For fibrous organic materials, primary consolidation may be 

completed in 15 minutes for each load increment. For high plasticity materials, the 

duration for each load increment may need to be 24 hours or more to ensure complete 

primary consolidation and to evaluate secondary compression behavior. Conversely, 

primary consolidation may occur in less than 3 hours for loads less than pc. 

 If the time period is too short for a given load increment (i.e., the sample is not 

allowed to achieve approximately 100 percent consolidation before the next load 

increment is applied), then values of Cc may be underestimated and values of cv may 

be overestimated. The duration of time required, however, can be optimized by using 

pneumatic, hydraulic, or electro-mechanical loading systems that include automated 

loading and data acquisition systems. Continuous deformation versus time 

measurements and the square root of time method described in Chapter 7 (Approach 

Embankment Deformations) can be used to estimate the beginning and end of 

primary consolidation during the test. Once the end of primary consolidation is 

detected, the system can automatically apply the next load increment. Alternatively, 

some laboratories can provide real-time deformation versus log time plots to enable 

the engineer to evaluate whether 100 percent primary consolidation has been 

achieved. 

•	 Secondary Compression: In cases where secondary compression is important (e.g., 

organic soils), secondary compression should be assessed on the basis of the 

deformation versus log-time response. The consolidation test for each load increment 

should be run long enough to establish a linear trend between vertical displacement 

and log time. 

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5.4.6 Useful Correlations between Consolidation Parameters and Index Values 

This section presents some useful correlations between consolidation parameters and other 

index values. These correlations can be used by the designer to check the validity of the 

laboratory tests results or to develop a prediction of the range of values of consolidation 

parameters that can be expected from yet-to-be performed consolidation tests. It must be 

emphasized that predictions based on correlations should never be substituted for 

proper testing and that any assumptions regarding consolidation parameters should 

always be verified through testing. 

5.4.6.1 Compression Index, Cc 

Over 70 different equations have been published for correlating Cc with the index properties 

of clays. Table 5-5 lists some of the more useful correlations. Figure 5-9 shows correlations 

between natural water content and Cc for fine-grained soils, peats and shales Note that the 

coordinates in Figure 5-9 are both logarithmic so that values of Cc can vary by as much 

as a factor of 5 with respect to the average trend line in these empirical correlations. 

Values of Cc obtained from Table 5-5 or Figure 5-9 should not be used for final design. 

Table 5-5 

Correlations for Cc (modified after Holtz and Kovacs, 1981) 

Correlation Soil 

Cc=0.156 eo + 0.0107(1) 

Cc.0.5Gs(PI/100) (2) All Clays 

Cc=0.30 (eo - 0.27) Inorganic, silt, silty clay 

Cc=0.009 (LL-10) (3) Clay of medium to slight sensitivity 

(St < 4, LL<100)(4) 

Cc=0.0115 wn

 (5) Organic Soils, Peat 

Cc=0.75 (eo - 0.50) Low plasticity clays 

(1) eo = initial void ratio, (2) PI = Plasticity Index, (3) LL=Liquid Limit, (4) St = sensitivity 

=Undisturbed undrained shear strength/Remolded undrained shear strength (see Table 3-12 

in Chapter 3, (5) wn = natural water content 

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Figure 5-9. Empirical correlation between compression index and natural (in-situ) 

water content (from Terzaghi, et al., 1996). 

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5.4.6.2 Recompression Index, Cr 

The ratio of Cr / Cc typically ranges from 0.02 to 0.20 (Terzaghi, and Peck, 1967). The low 

value is typical of highly structured and bonded soft clay or silt, while the largest ratio 

corresponds to micaceous silts and fissured stiff clays and shales. In reality, the value of Cr 

depends on whether loading or unloading is occurring, since some hysteresis effects develop 

when the soil is subjected to cycles of loading and unloading. 

 

Generally, it is sufficiently accurate to assume Cr is constant for most clay deposits. It may 

not be adequate to rely on a single value of Cr for loading and unloading in the case of highly 

structured soft clays or stiff clay shales. In the case of highly structured soft clays the initial 

value of Cr is steep as a result of flocculation (edge to face structure of clays) and bonding 

that allows the soil to be stable at high void ratios until the stress exceeds the 

preconsolidation pressure (Terzaghi and Peck, 1967). The subsequent rebound slope can be 

significantly different from the initial Cr. 

5.4.6.3 Coefficient of Vertical Consolidation, cv 

Because of the wide range of permeabilities that exist in soils (see Table 5-10), the 

coefficient of consolidation can itself vary widely, from less than 10 ft2

/yr (≈1 m2/yr) for 

clays of low permeability to 10,000 ft2

/yr (≈1,000 m2/yr) or more for very sandy clays, 

fissured clays and weathered rocks. Some typical values for clays are given in Table 5-6 and 

an approximate correlation with liquid limit is shown in Figure 5-10. 

Just as permeabilities in the horizontal and vertical directions can be significantly different 

due to variations in soil particle orientation, non-homogeneity, etc., so too can the in situ 

coefficient of horizontal consolidation, ch, be much different from the coefficient of vertical 

consolidation cv measured in the laboratory for the same reasons. For example, the in situ 

coefficient of horizontal consolidation, ch, for clays containing fissures or fine bands of sand, 

may often be much greater than cv measured in the laboratory for the clay alone. In such 

cases, the in-situ ch may govern the actual rate of consolidation under field loading 

conditions. 

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Table 5-6 

Typical values of coefficient of vertical consolidation, cv
 

(after Carter and Bentley, 1991)
 

Soil 

cv 

(cm2/s x 10-4) (m2/yr) (ft2/yr) 

Boston Blue Clay (CL) 40"20 12"6 135"70 

Organic silt (OH) 2-10 0.6-3 7-34 

Glacial lake clays (CL) 6.5-8.7 2.0-2.7 22-30 

Chicago silty clay (CL) 8.5 2.7 29 

Swedish medium sensitive clays (CL-CH) 

1. laboratory 0.5-0.7 0.1-0.2 1.7-2.4 

2. field 0.7-3.0 0.2-1.0 2.4-10.2 

San Francisco Bay Mud (CL) 2-4 0.6-1.2 6.8-13.6 

Mexico City clay (MH) 0.9-1.5 0.3-0.5 3.1-5.1 

Figure 5-10. Approximate correlations between cv and LL (NAVFAC, 1986a). 

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5.4.6.4 Coefficient of Secondary Compression, Cα

Table 5-7 presents typical values of Cα in terms of Cc for various geomaterials. As shown 

below by Equations 5-7, the coefficient, Cα, may be expressed either in units of strain (Cαε) or 

void ratio (Cαe) per log cycle of time. As indicated previously, to convert void-ratio-based 

consolidation curve indices to strain-based indices divide the void-ratio-based values by (1 + 

eo). 

dε de C C = ; C = ; C = αe

αε αe αε 5-7 d(log t) d(log t) (1 + e o )

Cαe is usually assumed to be related to Cc with values of Cαe/Cc typically in the range 0.025­

0.006 for inorganic soils and 0.035-0.085 for organic soils. Figure 5-11 presents a 

correlation between Cαe and natural water content. 

Table 5-7 

Typical Values of Cαe/Cce (Terzaghi, et al., 1996)
 

Soil Cα/Cc 

Granular soils including rockfill 0.02 ± 0.01 

Shale and mudstone 0.03 ± 0.01 

Inorganic clays and silts 0.04 ± 0.01 

Organic clays and silts 0.05 ± 0.01 

Peat and muskeg 0.06 ± 0.01 

Figure 5-11. Correlation between Cαe and natural water content (NAVFAC, 1986a). 

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Soils and Foundations – Volume I 5 - 37 December 2006 
5.5 SHEAR STRENGTH OF SOILS 

The shear strength of soils is extremely important to foundation design. In addition, slopes 

of all kinds, including hills, river banks, and man-made cuts and fills, stay in place only 

because of the shear strength of the material of which they are composed. Knowledge of the 

shear strength of soil is important for the design of structural foundations, embankments, 

retaining walls, pavements, and cuts. Table 5-8 provides a summary of specific issues 

related to the design and construction of typical highway design elements that should be 

considered in developing and implementing a laboratory and in-situ testing program for 

evaluating soil shear strength. 

5.5.1 Concept of Frictional and Cohesive Strengths 

The concept of shear strength was introduced in Chapter 2 where it was shown to be 

comprised of two components, friction (φ) and cohesion (c). In terms of the classification of 

soils introduced in Chapter 3, these two components of the shear strength can be generalized 

as follows: 

1. Coarse-grained soils, such as gravel and sand, and fine-grained silt, derive strength 

primarily from friction between particles. Therefore they are considered to be 

“cohesionless” or “frictional” soils and are often denoted as “φ-soils.” 

2. Fine-grained soils, composed mainly of clay, derive strength primarily from the 

electro-chemical attraction, or bond, between particles. Therefore they are 

considered to be “cohesive” soils and are often denoted as “c-soils.” 

3. Mixtures of cohesionless and cohesive soils derive strength from both interparticle 

friction and bonding. Such soils are commonly denoted as “c-φ soils.” 

5.5.1.1 Strength Due to Friction 

The strength due to friction between soil particles is dependent on the stress state of the soil 

(e.g., overburden pressure) and the angle of internal friction (φ) between the particles. As 

discussed in Chapter 2, the frictional resistance of soil is equal to the normal stress, σn, times 

the tangent of friction angle, φ. The tangent of φ is equal to the coefficient of friction (μ) 

between the soil particles. 

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Table 5-8 

Summary of issues relevant to shear strength evaluation in support of the design of 

typical geotechnical features (after FHWA, 2002a) 

Design 

Element Issues Relevant to Shear Strength Evaluation 

Shallow • Soil shear strength information required for depths up to 2 times the width of the 

Foundation footing, unless weak zones are found below this depth. The depth of bottom of 

footing will be based, in part, on requirements with respect to frost penetration depths 

and scour depths. 

Drilled Shaft • The excavation of a hole to construct a drilled shaft results in stress relief and 

disturbance in the soil that ultimately results in a reduction in shear strength from that 

corresponding to in-situ (i.e., before construction) conditions. The magnitude of the 

stress relief and disturbance will depend upon the method of construction, soil type, 

saturation condition, and type of strength (e.g., side shear or end bearing). 

Driven Pile • The shear strength of the soil may vary significantly between the time when the pile 

(or pile group) is first driven and tested to the time when the superstructure loads are 

applied to the pile (or pile group). The time-dependent phenomena of strength 

increase is referred to as “pile set-up” and is often observed for driven piles in 

saturated normally consolidated (NC) to moderately over consolidated (OC) clay and 

fine-grained material. A decrease in strength with time is referred to as “relaxation” 

and is often observed for heavily OC clays, dense silts, dense fine sands, and weak 

laminated rock. Shear strengths should therefore be evaluated for both long and short 

term conditions. 

• Changes in site conditions that affect the in-situ the effective stress state may increase 

or decrease shear strength and pile capacity. These may include site dewatering or 

additional surface loading from an embankment. 

• An increase in granular soil strength may occur due to densification during driving. 

This increased strength will need to be considered such that an appropriate pile driving 

system can be selected for construction. 

Retaining • The analysis of non-gravity cantilevered and anchored walls requires an evaluation of 

Walls earth pressures on the active side and passive side of the excavation. For undrained 

loadings in some clayey soils, particularly low to medium plasticity materials, there 

can be a large difference in undrained strength between the strength used for the active 

side and the passive side of the excavation. 

• For soils that may exhibit peak, fully softened, and residual conditions, an estimate of 

the tolerable deflection of the wall system needs to be made and this deflection used to 

select the appropriate strength condition for analysis. 

Slopes • The shear strength of discontinuities (e.g., fissures) in soil (and rock) needs to be 

evaluated since it may represent the critical (i.e., lowest) shear strength for design. 

• Weathering and other physiochemical reactions may occur at a quick enough rate to 

weaken soil bonds and reduce shear strength. 

• Strength loss may occur in cut slopes due to soil softening (in presence of water) and 

continuing deformations. Large deformation residual strengths should be used for 

long-term analyses. 

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The equation for frictional resistance, τ, is written in terms of normal stress, σn, as follows: 

τ = σn tan φ	 5-8 

The coefficient of friction, tan φ, between individual particles depends on both their mineral 

hardness and surface roughness. However, the measured friction angle of a soil sample or 

deposit also depends on the density of the mass caused by interlocking of particles. For a 

detailed discussion of factors affecting frictional resistance, the reader is referred to 

textbooks such as by Holtz and Kovacs (1981). 

5.5.1.2 Strength Due to Cohesion 

The concept of cohesive strength is more difficult to explain and often misunderstood. The 

designer must develop a good understanding of this concept otherwise there will be a 

disconnect between reality and the design of some structures, e.g., the first bench cut in 

shoring. 

There are two types of cohesion in soils: true cohesion and apparent cohesion. These are 

briefly discussed as follows (after Mitchell, 1976): 

1.	 True cohesion may result from chemical cementation (just like in rocks) and/or 

forces of attraction (e.g., electrostatic and electromagnetic attractions) between 

colloidal (10-3 mm to 10-6 mm) clay particles. True cohesion is stress-independent 

unlike frictional resistance that is a function of normal stress. 

2.	 Apparent cohesion may develop because of capillary stresses and mechanical 

interlocking as follows: 

•	 Capillary stresses develop between particles in a partially saturated soil due to 

surface tension in the water. The surface tension (negative pressure) in the water 

produces an equal and opposite effective stress between the soil particles, which 

results in an apparent cohesion since it too is stress-independent. The magnitude 

of this type of apparent cohesion can be extremely large, especially in finegrained soils. Such capillary stresses can be overcome by an increase in the 

degree of saturation. 

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•	 Apparent mechanical forces are often exhibited by the interlocking of rough 

(angular) soil particles. The interlock between the soil particles can offer some 

resistance to shear stresses even in the absence of a normal stress. This type of 

apparent cohesion is often the cause of cohesion measured in compacted soils. 

However, such apparent mechanical forces are susceptible to significant reduction 

by vibrations and other types of mechanical disturbance. 

Figure 5-12 presents a graphical representation of the potential contribution of various 

mechanisms of cohesion. It can be seen that true cohesion in soils exists only when the 

particle size is colloidal. Unless the complete soil sample is composed of colloidal particles, 

true cohesion due to interparticle attraction cannot be relied on. Cementation by deposition 

is often observed in arid environments (e.g., desert southwest), but it is difficult to quantify. 

As indicated above, capillary stresses can provide a large apparent cohesion, but such 

cohesion can be overcome by saturation. Since cohesion cannot be defined with confidence, 

its contribution to long-term shear strength in c-φ soils is often disregarded or greatly 

minimized by using only a small value such as 100 to 500 psf (5 to 25 kPa). For purely 

cohesive soils, the designer should be careful in evaluating the cohesion for long-term design 

purposes. Further discussion on apparent cohesion in the context of compacted soils is 

included in Section 5.8.4.1. 

 

Figure 5-12. Potential contributions of various bonding mechanisms to cohesive 

strength (after Ingles, 1962). 

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5.5.1.3 Simplified Expression for Shear Strength of Soils 

As indicated in Chapter 2, the shear strength, τ, of soils is expressed simplistically by two 

additive components as follows: 

τ = c + σn tan φ 5-9 

In terms of effective stresses, the effective shear strength, τ′, can be re-written as follows: 

τ′ = c′ + (σn - u) tan φ′ = c′ + σ′ tan φ′ 5-10 

where u is the pore water pressure, c′ is the effective cohesion, σ′ is the effective normal 

stress, and φ′ is the effective friction angle. 

The shear strength soil is influenced by many factors including the effective stress state, 

mineralogy, packing arrangement of the soil particles, soil hydraulic conductivity, rate of 

loading, stress history, sensitivity, and other variables. As a result, the shear strength of soil 

is not a unique property. The following sections present and discuss various laboratory 

tests to determine the shear strength for various types of construction and loading conditions. 

Typical laboratory strength tests are introduced including the unconfined compression test 

(AASHTO T208; ASTM D 2166), the triaxial test (AASHTO T234; ASTM D 4767), and the 

direct shear test (AASHTO T236; ASTM D 3080). A detailed discussion on testing 

equipment and procedures is beyond the scope of this document. The interested reader 

should review the AASHTO and ASTM standards for detailed information on testing 

equipment and procedures. The following sections also describe information that must be 

conveyed to a laboratory testing firm to ensure that the strength testing is performed 

consistent with the requirements imposed by the design (e.g., selection of confining pressures 

consistent with the imposed loads). 

5.5.2 Strength Testing of Soils in the Laboratory 

The shear strength of a soil is the maximum shear stress that the soil structure can resist 

before failure. Failure is generally defined as continuing displacement without an increase in 

applied stress. Since the water filling the pores has no shear strength, shear stresses are 

carried by the structure of soil grains. However, the shear strength of the soil structure is 

indirectly dependent on the pressure in the pore water, which influences the friction term as 

shown by the excess pore water pressure term, u, in Equation 5-10. Foundation designers 

must consider the effects of expected construction operations on the subsoils when planning 

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Soils and Foundations – Volume I 5 - 42 December 2006 
a test program. For example, when a highway embankment or structural footing is suddenly 

placed on a soft clay deposit, the pore water initially carries all the load and the available 

shear strength does not increase until drainage begins and the excess pore water pressure 

decreases. In planning a test program for such a situation the designer should request 

unconsolidated undrained (UU) triaxial tests to determine the undrained shear strength of the 

soil which, in this case, would be the critical strength value, i.e., the initial shear strength 

before consolidation begins. Additional consolidated undrained (CU) or consolidated 

drained (CD) tests would also be used to determine the increase in shear strength as 

consolidation occurs and excess pore water pressures dissipate. These results can be used to 

determine alternate methods of applying the loads safely, especially if the undrained strength 

is insufficient to sustain the proposed loading. Stage construction involves placement of an 

increment of load and a waiting period to allow strength gain through excess pore water 

pressure dissipation so the soil deposit can safely support the next load increment. 

The majority of strength tests are conducted on cohesive soils since obtaining undisturbed 

samples of non-cohesive soils is difficult. Strength tests on cohesive soils are usually 

conducted on high quality undisturbed samples obtained from thin wall sampling tubes. The 

preferred test for most projects where cohesive soils are involved is the triaxial compression 

test. The number and types of tests must be selected by the designer to suit the project 

conditions. For each test the designer should clearly indicate the consolidation or confining 

pressure to be used. These pressures are determined from the po diagram for each specific 

project (refer to Chapter 2 for discussion of the po diagram). The range usually extends from 

the effective overburden pressure to the pressure induced by the highway loading. The 

program objective should be to establish a profile of soil strength with depth. Soil strength 

parameters are frequently expressed as a ratio of shear strength over the effective overburden 

pressure (τ/po). 

The most common laboratory soil strength tests are: 

• Unconfined compression test 

• Triaxial compression test, and 

• Direct shear test 

Each of these tests is briefly discussed below. For the triaxial compression and the direct 

shear tests, it is important that each test be performed on a new sample. The practice of 

performing multi-stage shear strength tests on a single sample is not recommended. 

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5.5.2.1 Unconfined Compression (UC) Tests 

The unconfined compression test is a quick, relatively inexpensive means to obtain an 

estimation of the undrained shear strength of cohesive specimens. In this test a cylindrical 

specimen of the soil is loaded axially as shown in Figure 5-13 without any lateral 

confinement to the specimen, at a sufficiently high rate to prevent drainage. Since there is no 

confinement, residual negative pore pressures that may exist in the sample following sample 

preparation generally control the state of effective stress in the sample. The shear stresses 

induced in the specimen by the axial load result in a shear failure. The magnitude of the 

shear stress at the moment of failure represents the shear strength of the soil under these 

conditions of loading and drainage. Therefore, the shear strength obtained from this test is 

called the “undrained shear strength (su).” In most cases, the value of undrained shear 

strength obtained from an unconfined compression test is conservative. The maximum axial 

compressive stress measured at failure represents the compressive strength of the soil under 

these conditions of loading, drainage, and confinement. Therefore, the compressive strength 

obtained from this test is called the “unconfined compressive strength (qu).” These two 

strengths terms should not be confused; one is a shear strength the other a compressive 

strength. It can be shown graphically by a Mohr’s circle construction (Appendix B) that the 

undrained shear strength (su) is equal to one-half the unconfined compressive strength (qu). 

Axial Stress 

Cylindrical 

Soil Sample 

Axial Stress 

Figure 5-13. Schematic of an unconfined compression test. 

The unconfined compression test cannot be performed on granular soils, dry or crumbly 

soils, silts, peat, or fissured or varved materials. Because there is no control over the 

effective stress state of the specimen, this test is not recommended for evaluating strength 

properties for compressible clay soils subjected to embankment or structural foundation 

loads. The reliability of this test decreases with increasing sampling depth because the 

sample tends to swell after removal from the ground due to confining stress release. The 

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swelling results in greater particle separation and reduced shear strength. Testing the full 

diameter extruded specimen as soon as possible after removal from the tube can: 

• minimize swelling 

• reduce disturbance 

• preserve the natural moisture content. 

Unfortunately, despite these shortcomings, this test is commonly used in practice because of 

its simplicity and low cost. 

5.5.2.2 Triaxial Tests 

The triaxial test is very versatile in the sense that the shear strength can be evaluated under 

compression as well as extension loading conditions. A schematic of triaxial compression 

test is shown in Figure 5-14 where the axial stress is greater than the confining stress. Lateral 

pressures at various depths below the ground surface can be simulated by confining 

pressures. Note that the confining pressures acts on the entire sample and is equal to the 

axial stress before the application of an axial stress increment. Typically, failure of the 

sample is caused by increasing the axial stress (compression) until a shear failure takes place. 

In an extension test, the confining pressures are increased while keeping the axial stress 

constant. Pore water pressures during the test can be measured. 

Axial Stress = 

Confining Stress + 

Axial stress increment 

Confining 

(lateral) Stress 

Axial Stress = 

Confining Stress + 

Axial stress increment 

Figure 5-14. Schematic of a triaxial compression test (Lambe and Whitman, 1979). 

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 (a) (b) 

Figure 5-15. (a) Failure of a loose sand specimen in a triaxial cell; and 

(b) Load frame, pressure panel, and computerized data acquisition system 

(Photographs courtesy of GeoComp Corporation). 

 

Equipment – Triaxial systems today use electronic instrumentation to provide continuous 

monitoring and periodic acquisition of test data (see Figure 5-15b). Force is measured by 

using a force transducer or load cell that is typically mounted outside the triaxial cell. More 

advanced systems have the transducer incorporated within the testing cell to reduce the 

effects of rod-friction. Linear variable differential transducers (LVDTs) are used to monitor 

deformations. Additionally, volume measurements can be taken with a device that makes 

use of an LVDT to measure the rise or fall of a bellofram cylinder. This change in 

movement is calibrated to the volume of water taken in or pushed out of the sample. 

Pressure transducers are mounted on the base of the test cell to monitor the confining 

pressure within the cell and the pore water pressure within the sample. 

Unconsolidated-Undrained (UU) Test 

In the UU test, no drainage or consolidation is allowed during either the application of the 

confining pressure or the application of the axial load that induces shear stress. The shear 

stresses induced in the specimen by the axial load result in failure. As indicated in Section 

5.5.2, the UU test models the response of a soil that has been subjected to a rapid application 

of an axial load such as that due to construction of an embankment. It is difficult to obtain 

repeatable results for UU testing due to the effects of sample disturbance. The accuracy of 

the UU test is dependent on the soil sample retaining its original structure until testing 

occurs. The undrained shear strength of the soil, su, is measured in this test. 

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Consolidated-Drained (CD) Test 

In the CD test, the specimen is allowed to consolidate completely under the confining 

pressure prior to the application of axial load, i.e., the confining pressure acts as an effective 

stress throughout the soil specimen. The axial load is applied at a rate slow enough to allow 

drainage of pore water so that there is no buildup of excess pore water pressures, i.e., the 

stresses imposed by the axial load are effective stresses. The shear stresses induced in the 

specimen by the axial load result in failure. The time required to conduct this test in low 

permeability soil may be as long as several months. Therefore it is not common to conduct a 

CD test on low permeability soils. The CD test models the long-term (drained) condition in 

soil. Effective stress strength parameters (i.e., φ′ and c′) are evaluated from the results of the 

CD test. 

Consolidated-Undrained (CU) Test 

The initial part of the CU test is similar to the CD test in that the specimen is allowed to 

consolidate under the confining pressure. However, unlike the CD test, the axial load is 

applied with the drainage lines closed in the CU test. Thus, during shearing there is continual 

development (+ or -) of excess pore water pressure. The rate of axial load application for this 

test is more rapid than that for a CD test. Pore pressures are typically measured during the 

CU test so that both total stress and effective stress strength parameters can be obtained. 

Recall that total stress equals effective stress + pore water pressure as expressed by Equation 

2-13. The pore water pressure may be + or – depending upon whether the specimen dilates 

or compresses during application of the axial load. The shear stresses induced in the 

specimen by the axial load result in failure. The effective stress parameters evaluated for 

most soils based on CU testing with excess pore water pressure measurements will be similar 

to those obtained from CD testing, thus making CD tests unnecessary for typical 

applications. 

During triaxial testing, the confining pressure, which acts uniformly over the entire 

specimen, is considered to be the minor principal stress. By definition, a principal stress is 

one that acts on a plane where shear stress is zero. The interface between the soil and the 

membrane isolating it from the fluid in the chamber is assumed to be frictionless during the 

entire test, i.e. no shear stresses develop along the circumference of the specimen. Likewise, 

the applied axial load causes a normal stress to act on the top and bottom of the specimen. 

As shown in Figure 5-14, this vertical normal stress acts in addition to the confining pressure. 

Therefore, the combined vertical stress acting on the top and bottom of the specimen during 

the triaxial test is considered to be the major principal stress not only because the plane 

(horizontal) on which it acts is orthogonal to the minor principal stress plane (vertical), but 

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mainly because the interface shear stress between the specimen and the top and bottom caps 

is assumed to be equal to or close to zero. To assure this condition, the end caps are usually 

coated with a lubricant to make them virtually frictionless. Because of these boundary stress 

conditions, the specimen is free to shear on a plane consistent with the directions of the major 

and minor principal stress planes and its inherent shear strength as expressed by c and φ. 

Figure 5-16. Typical stress-strain curves from CU test. 

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5.5.2.3 Direct Shear Tests 

The oldest form of shear test upon soil is the direct shear test, first used by Coulomb (1776). 

A schematic of the essential elements of the direct shear apparatus are shown in Figure 5-17. 

The soil is held in a box that is split across its middle; the bottom portion of the box is 

usually fixed against lateral movement. A confining normal force, N, is applied, and then a 

tangential shear force, T, is applied so as to cause relative displacement between the two 

parts of the box. The magnitude of the shear force is recorded as a function of the shear 

displacement, and usually, the change in thickness of the soil sample is also recorded. 

Although it is widely used in practice, the direct shear device lacks a number of features that 

limit is applicability. For example, there is no way to control the confining pressure. Also, 

since there is no way to measure excess pore water pressures generated during shearing of 

saturated clay specimens, use of the direct shear test is generally limited to cohesionless soils. 

Predetermined Normal Force, N 

failure (shear) plane 

Sample Tangential 

Shear Force, T 

Side of bottom box 

fixed against lateral 

movement 

Figure 5-17. Schematic of the Direct Shear Test. 

Equipment – The apparatus and procedures for direct shear testing are discussed in ASTM D 

3080. A specimen is prepared in a split square or circular box. Figure 5-18 shows a circular 

specimen. The specimen is sheared as one part of the box is displaced horizontally with 

respect to the other. Generally, the lower part of the box is fixed against lateral movement 

and the shear force is applied to the upper part of the box through a loading frame as shown 

in Figure 5-18. The central two of the six screws visible in the top portion of the box extend 

into the bottom portion and are used to hold the assembly together while the specimen is 

being prepared. This shear box assembly is then placed in a reservoir which could be filled 

with water to allow saturation of the specimen prior to shearing. Before the test is begun, the 

two central screws are removed and the four corner screws, which rest on the top surface of 

the bottom portion of the box, are turned to slightly raise the top part of the box so that there 

is no contact between it and the bottom part of the box. This is done to prevent the error that 

would result from the frictional resistance between the two boxes at their contact. Load cells 

are used to monitor the shear force and LVDTs are used to monitor both horizontal and 

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vertical deformation. By use of this instrumentation, as well as a loading frame that provides 

a constant rate of horizontal deformation, it is possible to automate the direct shear test. 

Figure 5-18. Direct shear testing box (Photograph courtesy of GeoComp Corporation). 

Figure 5-19. Soil sample mounted in direct shear testing apparatus (Photograph 

courtesy of GeoComp Corporation). 

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Test Procedures – In the direct shear test, the soil is first consolidated under an applied 

normal stress. After consolidation is completed, which will be virtually instantaneous in 

cohesionless soils, the specimen is sheared directly at a constant rate. The rate of shear is 

typically selected as a function of the hydraulic conductivity of the specimen. Direct shear 

testing is commonly performed on compacted materials used for embankment fills and 

retaining structures. Direct shear testing can also be performed on natural materials. 

However, the lack of control on soil specimen drainage makes the evaluation of undrained 

strength unreliable. The direct shear test can also be used to evaluate the drained strength of 

natural materials by shearing the sample at a rate slow enough to ensure that no significant 

pore water pressures develop, however there is no way to verify this condition by 

measurement. 

In addition to providing data that allows the determination of the peak effective stress friction 

angle (φ′), the direct shear test data can also be used for the evaluation of effective stress 

residual strengths (c′r ≈ 0; φ′r). The effective stress residual strength parameters are 

necessary for stability and landslide analyses. Residual strengths are associated with very 

large shear strains along a predefined or preferential slip surfaces that result in very large 

deformations. Data from a reversing direct shear test can also be used to evaluate residual 

shear strengths. In a reversing direct shear test, the direction of shearing in the test is 

reversed several times thereby causing the accumulation of displacements at the slip surface. 

A characteristic of the direct shear test that distinguishes it from the triaxial test is that the 

shear failure in the direct shear device is forced to occur on a horizontal plane so that the 

orientations of the major and minor principal stress planes are not apparent. Ordinarily this 

characteristic is considered to be a disadvantage of the direct shear test. However this 

characteristic is advantageous for designs involving geosynthetics where the shearing 

resistance of the interface between the soil and the geosynthetic or between two pieces of 

geosynthetic is often required. Direct shear machines have been modified to test the 

interface shear strength between various types of engineering materials, as described in 

ASTM D 5321. 

5.5.3 Factors Affecting Strength Testing Results 

It is important for the designer to realize that all laboratory tests on soils must be carefully 

performed. This is particularly important for strength testing since the use of strength 

parameters are a key to successful foundation design. The following seven factors in 

particular affect the results of strength testing: 

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1. Sample disturbance 

2. Mode of shearing 

3. Confining pressures 

4. Specimen size 

5. Saturation 

6. Displacement at failure 

7. Rate of shearing and strain required to reach peak strength 

A detailed discussion of each of these factors is presented in FHWA (2002a). In addition to 

recognizing the effect of the factors that can affect the strength testing results, it is extremely 

important for the designer to perform the proper tests depending on project requirements. 

The selection of an appropriate test to be used to provide relevant information for a particular 

geotechnical structure should consider, at a minimum, the following questions: 

(1) How fast will construction occur relative to the hydraulic conductivity of the soil 

(i.e., should drained or undrained strength tests be performed)? 

(2) How does the direction of applied load affect measured shear strengths and the 

appropriate strength to be used for an analysis? 

(3) How do the expected levels of deformation for the geotechnical structure affect the 

selection of shear strength? and 

(4) How does the manner in which the feature is constructed affect the shear strength to 

be used in analysis? 

5.5.4 Comparison of Laboratory and Field Strengths 

Soil samples are obtained from the ground for laboratory testing by sampling from boreholes 

and sealing and transporting these samples to the laboratory. The degree of disturbance 

affecting the samples will vary according to the type of soil, sampling method and the skill of 

the driller. At best some disturbance will occur from the removal of in-situ stresses during 

sampling and from the preparation of specimens in the laboratory for testing. In general, 

disturbance tends to reduce the shear strength obtained from unconfined or unconsolidated 

tests and increase the shear strength obtained from consolidated tests. There is, therefore, 

considerable merit in measuring the in-situ shear strength. The field vane shear test 

discussed in Chapter 3 is the most commonly used field test for direct measurement of the 

undrained shear strength of soft to medium clays. In reviewing the different types of field 

and laboratory tests available to determine the undrained shear strength in clays, the designer 

should expect the field vane shear test to provide the most accurate value of su, with UC and 

UU tests yielding lower results and CU tests yielding slightly higher results. 

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It is important that the results of in-situ tests be interpreted carefully and calibrated with 

laboratory tests. Without careful calibrations, the in-situ tests can yield inaccurate results. 

5.5.5 Selection of Design Shear Strength 

Frequently, on a large project the designer will receive a large quantity of undrained shear 

strength test results from both the field and laboratory. This data must be synthesized to 

permit rational interpretation of the results. The tests should be analyzed on a hole-by-hole 

basis. All tests from one hole should be reviewed and the results for each type of test should 

be plotted separately versus depth to determine the pattern of strength variation for each test 

type and to assess the reliability of the data, e.g., a CU test result that is lower than a UC test 

result at the same depth should be considered suspect. The general pattern of shear strength 

results should show an increase in strength with depth in a normally consolidated clay 

deposit. Overconsolidated clays may exhibit this pattern only at greater depths since the 

amount of preconsolidation increases shear strength in the upper portions of the soil deposit. 

Section 5.14 presents values of the coefficient of variation of measured soil properties that 

should be taken into account while selecting the final design shear strength. 

5.5.6 Correlations of Shear Strength Parameters with Index Parameters 

This section presents some useful correlations between shear strength parameters and other 

index values or field conditions. These correlations may be used by the designer to check the 

general validity of the laboratory test results or to develop a preliminary assessment about the 

shear strength characteristics of the soils on the project site. In the latter case, it must be 

emphasized that predictions based on correlations alone should never be used for design and 

that any assumptions regarding shear strength parameters must always be verified through 

testing. 

5.5.6.1 Undrained Shear Strength of Cohesive Soils 

For most saturated clays tested under quick undrained conditions, the angle of shearing 

resistance, φu, is zero. This means that the shear strength of the clay is a fixed value and is 

equal to the apparent cohesion, cu, at a specific moisture content and preconsolidation 

pressure. A value for the undrained shear strength may be crudely estimated for a sample for 

which uncorrected SPT N-values are known by molding a specimen of the clay between the 

fingers and by applying the observations indicated in Table 4-2 in Chapter 4. However, as 

noted in the footnote of Table 4-2, the values of SPT blow count listed there should not be 

used to determine the design strength of fine grained soils. 

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Undrained Shear Strength 

For most normally consolidated clays the undrained shear strength (su) is proportional to the 

effective overburden pressure (po). For such soils, Skempton (1957) proposed the 

relationship shown in Figure 5-20 between the su/po and plasticity index (PI). Figure 5-20 

also includes results obtained by a number of other researchers. As can be seen in the figure, 

the composite of all findings varies so much that such relations should be used with 

caution. However, such correlations, particularly the correlation by Skempton (1957), are 

useful for obtaining preliminary estimates and for checking laboratory results of projectspecific tests performed on normally consolidated clays. 

 

 

, i.e., pc o > p

, i.e., pc = po 

(normally consolidated) 

(overconsolidated) 

Figure 5-20. Relationship between the ratio of undrained shear strength to effective 

overburden pressure and plasticity index for normally consolidated and 

overconsolidated clays (after Holtz and Kovacs, 1981). 

The shear strength of undisturbed clays depends on the consolidation history of the clay as 

well as its fabric characteristics. In general, the undrained strength ratio, su/po, increases with 

increasing overconsolidation as measured by the overconsolidation ratio, OCR. It is 

recommended that laboratory tests be performed to determine the undrained strength 

ratio for overconsolidated clays rather than relying on any published correlations. In 

practical terms, it is more straightforward to measure the undrained shear strength of 

overconsolidated clays than to predict it from other indices. 

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5.5.6.2 Drained and Effective Shear Strength of Cohesive Soils 

It is often important to carry out stability calculations in terms of effective stresses. The soil 

strength parameters, c′ and φ′, used in these calculations should be obtained from either 

drained direct shear box or drained triaxial tests or from CU triaxial tests with pore water 

pressure measurements (giving φ′cu and c′cu). Generally, there is a minor difference in the 

results obtained from these two tests for saturated clays because the soil is being tested under 

different boundary conditions and stress paths. In-situ tests such as CPTs can also be used to 

estimate the drained and effective shear strength parameters of cohesive soils. 

For clays, empirical correlations have been developed to relate φ′ to the plasticity 

characteristics of the soil. Figure 5-21 shows a slight trend of φ′ decreasing with increasing 

PI. The existence of these relationships arises because both PI and shear strength reflect the 

clay mineral composition of the soil; as the clay mineral content increases, the PI increases 

and the strength decreases. From Figure 5-21, it can be seen that the drained friction angle 

values can be ± 8° in variance with respect to the dashed trend line. 

Considering the overall importance of φ′ in stability calculations, foundations design, and 

landslide analyses, it is essential to assess φ′ directly by means of consolidated drained direct 

shear tests, consolidated drained triaxial tests, or consolidated undrained triaxial tests with 

pore water pressure measurements. The consequences of merely estimating φ′ can be 

economically unwise. As an example, in stability analyses for relatively long, shallow slip 

surfaces that may be associated with a landslide, the required forces that would need to be 

resisted by some form of stabilization system (e.g., retaining wall, micropiles) would vary 

significantly depending on the drained friction angle of the soil. It is highly recommended 

that geotechnical designers develop historical data summaries of φ′ versus PI to check the 

validity of future test results. 

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Soils and Foundations – Volume I 5 - 55 December 2006 
20 

40 

50 

0 

10 

Soft Clays 

Mexico City 

Clay 

Attapulgite 

Soft and Stiff Clays 

Shales 

Clay Minerials 

30 

Fricttion Angle, (degrees) ’ 

0 20 40 60 80 100 200 400 600 1000 

Plasticity Index, PI (%) 

Figure 5-21. Relationships between φ and PI. (after Terzaghi, et al., 1996). 

5.5.6.3 Shear Strength of Cohesionless Soils 

Because of their high permeability, pore water pressures do not build up significantly when 

cohesionless soils are subjected to shearing forces. The complication of total and effective 

stresses is therefore avoided and the phenomenon of apparent cohesion, or undrained shear 

strength does not occur. Consequently, the shear strength of cohesionless soils is defined 

exclusively in terms of frictional resistance between the grains, as measured by the angle of 

shearing resistance, φ. Typical values of φ for sands and gravels are given in Figure 5-22 as a 

function of dry unit weight and relative density. The material types indicated in the figure 

relate to the Unified Classification System (USCS). 

Figure 5-22 requires determination of relative density. A reasonable estimate of relative 

density can be obtained from Figure 5-23. Figure 5-23 was originally developed based on 

data obtained using rope and cathead operated hammers. Thus, it is recommended that an 

energy corrected SPT N-value, i.e., N60, be used as shown in Figure 5-23. However, note 

that Figure 5-23 is a function of both N-value and the vertical effective overburden pressure, 

po. Therefore, N60-value should not be corrected for overburden pressure, i.e., CN=1.0 (see 

Section 3.7.2) while using Figure 5-23. 

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Note: Use caution in the shaded portion of the chart due to the potential for unreliable SPT N-values in gravels 

Figure 5-22. Correlation between relative density, material classification and angle of 

internal friction for coarse-grained soils (NAVFAC, 1986a). 

Vertical effective 

overburden pressure, po, ksf 

Energy corrected SPT N-value, N60 

1 ksf = 47.9 kPa 

Figure 5-23. Correlation between relative density and SPT resistance (NAVFAC, 

1986a). 

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5.6 PERMEABILITY 

5.6.1 General 

Permeability, also known as hydraulic conductivity, is one of the major parameters used in 

selecting soils for various types of construction. In some cases, it may be desirable to place a 

high-permeability fill immediately under a pavement surface or behind a wall to facilitate the 

removal of water. In other cases, such as retention pond dikes, it may be detrimental to use 

high-permeability materials. Permeability also significantly influences the choice of backfill 

materials for various elements such as trenches and retaining walls. 

Unlike the fill soils discussed in the previous paragraph, the permeability of a natural soil is 

strongly influenced by its macroscopic structure, e.g., clays containing fissures or fine bands 

of sand will have permeabilities that are many times greater than that of the clay material 

itself. Also, stratified soils often have horizontal permeabilities that are many times the 

vertical permeability. Because of the small size of laboratory specimens and the way they 

are obtained and prepared, large scale in-situ features are absent and test results do not give a 

true indication of field values in soils with a pronounced macro-structure. Moreover, 

laboratory tests usually constrain water to flow vertically through the specimen whereas the 

horizontal permeability may be much greater, and hence of overriding importance, so far as 

site conditions are concerned. Field permeability tests overcome these shortcomings, but, 

since the pattern of water flow from a well used to determine in situ permeabilities can only 

be estimated, interpretation of the field test results is difficult and uncertain. Thus, one set of 

problems is exchanged for another. 

5.6.2 Equipment 

Laboratory permeability testing is performed to determine the permeability or hydraulic 

conductivity (k) of a soil specimen. For natural soils, tests are conducted on specimens from 

tube samples. For fill and borrow soils, tests are performed on compacted materials. Two 

types of tests are commonly performed, the rigid wall test (AASHTO T215; ASTM D 2434) 

and the flexible wall test (ASTM D 5084). 

The equipment for the rigid wall test includes a rigid wall permeameter, a water tank, a 

vacuum pump, and manometer tubes (see Figure 5-24). The permeameter must be large 

enough to minimize sidewall leakage. Therefore, the diameter of the rigid wall device should 

be at least 8 to 12 times that of the largest soil particle in the sample being tested. 

Frequently, the side wall of the cylinder is coated with petroleum jelly to prevent sidewall 

leakage. Porous stones and filter paper placed on the top and bottom of the test specimen to 

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Soils and Foundations – Volume I 5 - 58 December 2006 
prevent soil particle erosion by retaining fine particles must not restrict flow through the soil 

otherwise the permeability of the porous stones or filter paper will be measured. A vacuum 

pump is used to remove air from the sample and to help saturate the specimen prior to 

testing. In a rigid wall test saturation is performed from the bottom of the specimen upward 

(see ASTM D 2434). Manometer outlets should be available on the sides of the cell to 

measure head loss over the specimen. 

Rigid wall permeameters are not recommended for low permeability (i.e., k ≤ 2 x 10-5 ft/min 

(1 x 10-6 cm/s)) soils due to the potential for sidewall leakage. Rigid wall permeameters are 

typically used for sandy and gravelly soils (ASTM D 2434) with a permeability greater than 

2x10-2 ft/min (1x10-3 cm/s). Rigid wall systems are used for granular materials because the 

permeability of the flexible wall system (valves, pore stones, tubing, etc.) may be less than 

that of the specimen. Therefore the flexible wall system itself may affect the permeability 

value of granular soils and a measured value of permeability lower than that of the specimen 

may result. 

Figure 5-24. Rigid wall permeameter (Photograph courtesy of GeoComp Corporation). 

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Soils and Foundations – Volume I 5 - 59 December 2006 
Figure 5-25. Flexible wall permeameter (Photograph courtesy of GeoComp 

Corporation). 

The equipment for a flexible wall test includes a flexible wall permeameter cell, a cell 

reservoir, a headwater reservoir, a tailwater reservoir, top and base caps, a flexible 

membrane, porous stones, and filter paper (see Figure 5-25). The specimen can be tested 

over a range of confining stresses under backpressure saturation. The separate headwater and 

tailwater reservoirs can be monitored, and falling head or constant head tests can be 

performed. Since the flexible membrane encases the specimen, side leakage is prevented. 

Flexible wall permeameter cells consist of influent and effluent lines as well as porous stones 

and filter paper. The hydraulic conductivity of the system should be tested before a soil 

specimen is tested to ensure that the system’s conductivity is at least one order of magnitude 

greater than that anticipated for the soil. The triaxial cell can be used as a flexible wall 

permeameter. In fact, permeability measurements are often made as part of a drained triaxial 

(CD) test. 

The determination of permeability by testing is predicated on the validity of Darcy’s Law for 

laminar flow through porous media. If any of the assumptions of Darcy’s Law are violated, 

the results of permeability testing will be invalid. 

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5.6.3 Procedures 

Flexible wall permeameters (ASTM D 5084) are used for materials with a hydraulic 

conductivity (k) less than or equal to 2x10-2 ft/min (1x10-3 cm/sec). The flexible membrane 

used to encase the specimen prevents sidewall leakage for fine-grained soils that are likely to 

occur in a rigid wall system. The confining stress of the hydraulic conductivity test should 

be specified to the laboratory. As confining stress increases, the hydraulic conductivity of 

fine grained soils will typically decrease due to consolidation of the specimen and reduction 

of void ratio. The confining stress should be equal to the anticipated effective stress-state in 

the soil. 

The hydraulic gradient, defined as the difference in hydraulic head across the specimen 

divided by the length of the specimen, should also be specified to the laboratory. Typical 

hydraulic gradients in field situations are less than 5, however the use of such a small 

gradient in the laboratory will result in extremely long testing times for materials having 

hydraulic conductivities less than 2x10-5 ft/min (1x10-6 cm/sec). If the hydraulic gradient 

across the specimen is too high, turbulent flow will occur and Darcy’s Law will be violated 

with the result that the measured hydraulic conductivity will be less than that which will 

occur in the field. Suggested values of hydraulic gradient, as presented in Table 5-9, are a 

function of the anticipated hydraulic conductivity. 

Saturation of the specimen is necessary to achieve accurate results. A hydraulic conductivity 

test should be ended when steady flow is occurring. The flow through the permeameter is 

considered to be steady when four or more consecutive hydraulic conductivity measurements 

fall within ±25 percent of the average k value if k is greater than 1 x 10-8 cm/sec, or if four or 

more measurements fall within ±50 percent of the average if k is less than 1 x 10-8 cm/sec. 

Table 5-9 

Recommended maximum hydraulic gradient for permeability testing 

Hydraulic Conductivity, cm/sec* Recommended Maximum Hydraulic Gradient 

1 x 10-3 to 1 x 10-4 2 

1 x 10-4 to 1 x 10-5 5 

1 x 10-5 to 1 x 10-6 10 

1 x 10-6 to 1 x 10-7 20 

less than 1 x 10-7 30 

* Conventionally expressed in cm/sec. [1 cm/sec ≈ 2 ft/min = 0.6 m/min] 

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5.6.4 Useful Correlations of Permeability with Index Values 

The typical range of permeability values for various soil types and USCS groups is presented 

in Table 5-10, which is based on information originally presented by Casagrande and Fadum 

(1940). Superimposed on Table 5-10 are “typical soil groups” identified by their USCS 

symbols (Carter and Bentley, 1991). The range of permeability values corresponding to 

those groups is typical for compacted soils of that type where compaction is according to 

ASTM D 1557. Typical permeability values for highway construction materials are given in 

Table 5-11. 

Numerous correlations of permeability with grain size can be found in the literature. Figure 

5-26 presents logarithmic plots of k versus D10, based on experimental results. Figure 5-26 

includes the well known Hazen’s formula. All the correlations shown in Figure 5-26 were 

developed for sands and gravels. The great range of particle size present in most clays and 

the effects of clay mineralogy make such correlations of limited use for clays. 

Hazen’s equation is the most common correlation equation used to estimate permeability for 

sands (k > 10-3 cm/sec). This equation is written as: 

where: k is the permeability in cm/s; 

C = a coefficient ranging from 0.4 to 1.2 depending on sand size/sorting; and 

D10 = effective grain size in mm at 10% passing by weight as determined from 

sieve analysis. 

Hazen’s equation should be used with caution since it provides very approximate estimates 

of k applicable only to clean sands having less than 5% passing the No. 200 sieve (0.075 

mm) and with D10 sizes between 0.1 and 3.0 mm (Holtz and Kovacs, 1981). Hazen’s 

equation is valid only for k > 10-3 cm/sec. 

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k = C(D10)

2 5-14
Table 5-10 

Typical permeability values in soils (after Carter and Bentley, 1991) 

Table 5-11 

Typical permeability values for highway materials (after Krebs and Walker, 1971)
 

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Materials Permeability (cm/sec)

Uniformly graded coarse aggregate 40 - 4x10

Well-graded aggregate without fines 4x10-1 - 4x10-3 

Concrete sand, low dust content 7x10-2 - 7x10-4 

Concrete sand, high dust content 7x10-4 - 7x10-6 

Silty and clayey sands 10-5 - 10-7 

Compacted silt 7x10-6 - 7x10-8 

Compacted clay Less than 10-7 

Bituminous concrete (new pavements)* 4x10-3 - 4x10-6 

Portland cement concrete less than 10-8 

-1 

* Values as low as 10-8 have been reported for sealed, traffic compacted highway pavement. 
cm/sec 

Figure 5-26. The permeability of sands and gravels (after NAVFAC, 1986a). 

Note: In Figure 5-26, correlations shown are for remolded compacted sands and sandgravel mixtures with Cu values as indicated. 

2 

5 

2 

1 

0.5 

0.2 

0.1 

0.05 

0.02 

0.01 

0.005 

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5.7 VOLUME CHANGE PHENOMENA DUE TO LOADING AND MOISTURE 

Depending on mineralogy and depositional patterns, natural soils can exhibit either swell 

(expansion) or collapse under various degrees of loading and moisture ingress. Moisture 

may be in liquid or frozen form. For foundation design, it is very important to recognize and 

evaluate the potential for soils to swell or collapse. It is important to realize that these 

phenomena happen in both natural and compacted soils. Every year millions of dollars are 

spent dealing with the consequences of swelling (expanding) and collapsing soils. This 

section briefly discusses these two mechanisms and the tests that can be performed to 

evaluate the swell (expansion) and collapse potentials. 

5.7.1 Swell Potential of Clays 

Swelling is a characteristic reaction of some clays to water ingress. The potential for swell 

depends on the mineralogical composition of the soil fines. While montmorillonite (smectite) 

exhibits a high degree of swell potential, illite has no to moderate swell potential, and 

kaolinite exhibits almost none. The percentage of volumetric swell of a soil depends on the 

amount and type of clay, its relative density, the compaction moisture content and dry 

density, permeability, location of the groundwater table, the presence of vegetation and trees, 

overburden pressure, etc. Expansive soils are found throughout the U.S., however, damage 

caused by expansive clays is most prevalent in certain parts of California, Wyoming, 

Colorado, and Texas, where the climate is considered to be semi-arid and periods of intense 

rainfall are followed by long periods of drought. This pattern of wet and dry cycles results in 

periods of extensive near-surface drying and desiccation crack formation. During intense 

precipitation, water enters the deep cracks causing the soil to swell; upon drying, the soil will 

shrink. This weather pattern results in cycles of swelling and shrinking that can be 

detrimental to the performance of pavements, slabs on-grade, and retaining walls built on or 

in such soils. 

Deep-seated volume changes in expansive soils are rare. More common are volume changes 

within the upper 3-10 feet (1- 3 m) of a soil deposit. These upper few feet are more likely to 

be affected by seasonal moisture content changes due to climatic changes. The zone over 

which volume changes are most likely to occur is defined as the active zone. The active zone 

can be evaluated by plotting the moisture content with depth for samples taken during the 

wet season and for samples taken during the dry season at the same location. The depth at 

which the moisture content becomes nearly constant is the limit of the active zone depth, 

which is also referred to as the depth of seasonal moisture change. The active zone is an 

important consideration in foundation design. In the design of piles or drilled shafts, it is 

important to recognize that full side friction resistance may not be realized in this zone. As 

FHWA NHI-06-088 5 – Laboratory Tests 

Soils and Foundations – Volume I 5 - 65 December 2006 
 

 

the soil undergoes cycles of shrinking and swelling, it may lose contact with the pile or shaft. 

Alternatively, as the soil swells, it may impose significant uplift pressures on the foundation 

element. 

In the field, the presence of surface desiccation cracks and/or fissures in a clay deposit is an 

indication of expansion potential. Experience has indicated that the most problematic 

expansive near-surface soils are typically highly plastic, stiff, fissured, overconsolidated 

clays. Several classification methods are used to identify expansive soils in the laboratory. 

Currently, there is not a standard classification procedure; different methods are used in 

various locations across the U.S. Typically, methods include the use of Atterberg limits 

and/or clay size percentage to describe a soil qualitatively as having low, medium, high, or 

very high expansion potential. Generally, soils with a plasticity index less than 15 percent 

will not exhibit expansive behavior. For soils with a plasticity index greater than 15 percent, 

the clay content of the soil should be evaluated in addition to the Atterberg limits. Figure 5­

27 shows the swelling potential of natural soils and soils compacted to standard Proctor 

procedures (ASTM D 698) as a function of modified activity index, Am, (Equation 5-4) and 

clay fraction. 

Modified Activity Index, Am 

Figure 5-27. Classification of swell potential for soils (after Seed et al., 1962). 

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Soils and Foundations – Volume I 5 - 66 December 2006 
5.7.1.1 Evaluation of Expansion (Swell) Potential 

For situations where it is necessary to construct a facility in and around expansive soils, it 

will be necessary to estimate the magnitude of swell, i.e., surface heave, and the 

corresponding swelling pressures that may occur if the soil becomes wetted. The swelling 

pressure represents the magnitude of pressure that would be necessary to resist the tendency 

of the soil to swell. A one-dimensional swell potential test can be performed in an oedometer 

on undisturbed or recompacted samples according to AASHTO T256 or ASTM D 4546. In 

this test, the swell potential is evaluated by observing and measuring the swell of a laterally 

confined specimen when it is lightly surcharged and flooded with water. Alternatively, if the 

swelling pressure is to be measured, the height of the specimen is kept constant by adding 

load after the specimen is inundated. The swelling pressure is then defined as the vertical 

pressure necessary to maintain zero volume change. Swelling pressures in some expansive 

soils may be so large that the loads imposed by lightweight structures or pavements do little 

to counteract the swelling. 

The use of the one-dimensional swell potential test to evaluate in-situ swell potential of 

natural and compacted clay soils has limitations including: 

•	 Lateral swell and lateral confining pressure are not simulated in the laboratory. The 

calculated magnitude of swell in the vertical direction may not be a reliable estimate 

of soil expansion for structures that are not confined laterally (e.g., bridge 

abutments); 

•	 The rate of swell calculated in the laboratory will not likely be indicative of the rate 

of swell experienced in the field. Laboratory tests cannot simulate the actual 

availability of water in the field. 

It should be noted that there is a lack of a standard definition of swell potential in the 

technical literature based in part on variations in the test procedures, e.g., the condition of the 

test specimen (remolded or undisturbed), the magnitude of the surcharge, etc. Therefore the 

geotechnical specialist must be sure that the conditions used in the laboratory swell test 

simulate those expected in the field. In general, soils classified as CL or CH according to the 

USCS and A-6 or A-7 according to the AASHTO classification system should be considered 

potentially expansive. 

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5.7.2 Collapse Potential of Soils 

There are several types of soils that can experience collapse under moisture ingress. 

Examples of such soils are wind-blown deposits such as loess, or alluvial soils deposited in 

arid or semi-arid environments where evaporation of soil moisture takes place at such a rapid 

rate that the deposits do not have time to consolidate under their self weight or where the 

deposits are cemented by precipitated salts. Such soils are predominantly composed of silts 

and some clay. Typically, the structure of such soils is flocculated and the soil particles are 

held together by “clay bridges” or some other cementing agent such as calcium carbonate. 

In both cases disturbed samples obtained from these deposits are generally classified as silt. 

When dry or at low moisture content the in-situ material gives the appearance of a stable 

deposit. At elevated moisture contents these soils generally undergo sudden changes in 

volume and collapse. Full saturation is not required to realize collapse of such soils; 

often collapse of the soil structure occurs at moisture contents corresponding to precollapse degrees of saturation between 50 to 70%. Such soils, unlike other non-cohesive 

soils, will stand on almost a vertical slope until inundated. Collapse-susceptible soils 

typically have a low relative density, a low unit weight and a high void ratio. Figure 5-28 is 

a useful tool for assessing whether a soil is collapsible or not based on LL and dry unit 

weight. 

Figure 5-28. Chart for evaluation of collapsible soils (after Holtz and Hilf, 1961). 

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Soils and Foundations – Volume I 5 - 68 December 2006 
Structures founded on such soils may be seriously damaged if the soils are inundated and 

collapse. Therefore, if a soil is suspected to be collapse susceptible, then it is of primary 

importance to estimate the magnitude of potential collapse that may occur if the soil becomes 

wetted. To do this, a one-dimensional collapse potential test can be performed in an 

oedometer on undisturbed or recompacted samples according to ASTM D 5333. For this 

test, a sample is placed in an oedometer at it natural or compacted moisture content and the 

vertical pressure on the sample is increased in increments to the anticipated final loading in 

the field. Readings of vertical deformation are taken during the loading sequence. At the 

anticipated final load level, water is introduced to the sample and the resulting deformation 

due to collapse is recorded. The percent collapse (%C) is defined as: 

%C 100 ∆Hc = 5-15 Ho 

where ∆Hc is the change in height upon wetting and Ho is the initial height of the specimen. 

Conceptually, C is a strain. Therefore, for a soil layer with a given thickness, H, the 

settlement due to collapse, scollapse, if the entire thickness is inundated may be calculated as: 

⎛(%C)⎞

scollapse = H ⎜ ⎟ 5-16 ⎝ 100 ⎠

The collapse potential (CP) is calculated as the percent collapse (%C) of a soil specimen 

subjected to a total load of 4 ksf (200 kPa) as measured by using procedures specified in 

ASTM D 5333. The CP is an index value used to compare the susceptibility of various soils 

to collapse. Table 5-12 provides a relative indication of the degree of severity for various 

values of CP. 

Table 5-12 

Qualitative assessment of collapse potential (after ASTM D 5333) 

Collapse Potential (CP) Severity of Problem 

0 None 

0.1 to 2% Slight 

2.1 to 6% Moderate 

6.1 to 10% Moderately Severe 

>10% Severe 

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Soils and Foundations – Volume I 5 - 69 December 2006 
5.7.3 Expansion of Soils due to Frost Action 

The expansion of soils due to frost action is commonly known as frost heave. Three 

conditions are required for frost heave to occur. These are (a) freezing surface, (b) source of 

water, and (c) fine grained soils in which capillary rise can occur. Frost action in soils can 

have important engineering consequences as follows (after Holtz and Kovacs, 1981): 

•	 The volume of the soil can immediately increase about 10% just due to the volumetric 

expansion of water upon freezing. 

•	 The formation of ice crystals and lenses in the soil can cause heaving and damage to 

light surface structures such as small buildings, and highway pavements. Frost action 

can also displace retaining walls due to increased lateral pressures. 

When water in saturated fine-grained soils freezes, it forms lenses of ice oriented roughly 

parallel to the surface exposed to low temperature (Terzaghi, et al., 1996). Due to the 

inherent variability in the distribution of pore space, soils affected by frost action do not 

freeze and expand uniformly. Therefore, just as with swelling and collapsing soils, 

differential movement occur, and causes structural damage. Upon thawing, the moisture 

content of the soil increases which leads to reduction in shear strength and a consequent 

reduction in bearing capacity. Thus, the freeze-thaw cycle results in significant distress to 

structures and in particular highway pavements. 

Only fine-grained soils are susceptible to frost action. However, the critical grain size 

marking the boundary between soils that are subject to ice-lens formation and those that are 

not depends on the uniformity of the soil, i.e., the distribution of pore space. The following 

conditions noted by Terzaghi, et al. (1996) may be used as a general guide for evaluating the 

frost susceptibility of soils: 

•	 In perfectly uniform soils, i.e., a single particle size soils, ice lenses do not develop 

unless the grains are smaller than 0.01 mm. 

•	 Uniform soils must contain at least 10% of grains smaller than 0.02 mm. 

•	 In mixed-grain soils, ice lenses form when grains with a size less than 0.02 mm 

constitute at least 3% of the total aggregate. 

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•	 In soils with less than 1% of grains smaller than 0.02 mm, ice lenses are not formed 

under any conditions which may be encountered in the field. 

Common frost susceptible soils include silts (ML, MH), silty sands (SM), and low plasticity 

clays (CL, CL-ML). One of the most common methods to mitigate the detrimental effects of 

frost is to place the foundations below the anticipated frost depth. The depth of frost action 

depends primarily on air temperature below freezing and duration, soil permeability and soil 

water content. Figure 5-29 can be used for a preliminary estimate of the frost depth. More 

positive measures to mitigate damage due to frost action include lowering of the ground 

water table and, depending on the depth of the frost penetration, removal of the frost 

susceptible soils in the subgrade or foundation. Use of impervious membranes, chemical 

additives, and even foamed insulation (Styrofoam) under highways, buildings, and railroads 

have been successfully employed (Holtz and Kovacs, 1981). 

Figure 5-29. Approximate frost depth map for United States (Bowles, 1996). 

(1 m = 3.28 ft) 

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Soils and Foundations – Volume I 5 - 71 December 2006 
5.8 COMPACTION CHARACTERISTICS OF SOIL 

5.8.1 Concept of Compaction 

In the construction of highway embankments, earth dams, retaining walls, structural 

foundations and many other facilities, loose soils must be compacted to increase their 

densities. Compaction is the process of densifying soil under controlled moisture conditions 

by application of a given amount and type of energy. Compaction increases the density of 

the soil, which generally leads to: 

•	 an increase in the strength and stiffness characteristics of the soil, 

•	 a decrease in the amount of undesirable settlement of structures under both static and 

dynamic loads, 

•	 a reduction in soil permeability, and 

•	 an increase in the stability of slopes and embankments. 

Unless compaction is properly controlled, there is a potential that the volume change 

phenomena described in Section 5.7 (swell, collapse and frost heave) can occur. 

The density of compacted soils is measured in terms of the dry unit weight, γd, of the soil. 

The dry unit weight is a measure of the amount of solid materials present in a unit volume of 

soil. The greater the amount of solid materials, the stronger and more stable the soil will be. 

Pertinent parameters for evaluating the results of laboratory and field compaction tests are: 

•	 dry “density” or dry “unit weight.” 

•	 compaction water content. 

•	 type of energy input, e.g., impact, static, vibratory, kneading. 

•	 amount of energy input expressed in ft-lbs/ft3 

. 

Table 5-13 presents a summary of the characteristics of the most commonly used laboratory 

compaction tests. Figure 5-30 shows a typical hammer and a mold which is used for 

performing compaction tests in the laboratory. A comparison of the various values in Table 

5-13 reveals that the energy level in the Modified Proctor compaction (MPC) test is 4.5 times 

that for the Standard Proctor compaction (SPC) test. 

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Table 5-13 

Characteristics of laboratory compaction tests
 

Common 

Name 

ASTM 

(AASHTO) 

Designation 

Mold 

Dimensions Hammer No. of 

Layers 

Blows/ 

Layer 

Energy 

(ft-lbs/ft3 Diam. )

(in) 

Height 

(in) 

Vol. 

(ft3

) 

Wt. 

(lbs) 

Drop 

Ht. (in) 

Standard 

Proctor 

D 698 

(T 99) 

4 41⁄2 1/30 5.5 12 3 25 12,375 

Modified 

Proctor 

D 1557 

(T 180) 

4 41⁄2 1/30 10 18 5 25 56,250 

Note: Both tests are performed on minus No. 4 (4.75 mm) fraction of the soil. 

Hammer 

Mold 

12” 

Figure 5-30. Hammer and mold for laboratory compaction test 

(tape measure is for scale purpose only). 

5.8.2 Test Procedures 

At least 3 (preferably 5) samples of the same type of soil are prepared at various water 

contents and compacted according to the requirements listed in Table 5-13. Following 

compaction, the moist unit weight of the compacted soil (γt) in the mold is easily calculated 

as the weight of the soil (measured) divided by the volume of the mold (constant = 1/30 ft3

). 

The water content (w) is determined as per ASTM D 2216-05 and the dry unit weight is then 

calculated as (see Table 2-2 in Chapter 2): 

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γ γ	 d = t 	 5-17 (1 + w) 

The dry unit weight (pcf (kN/m3

)) for each compacted sample is plotted versus its 

compaction moisture content (%). The resulting curve is called a compaction curve. Figure 

5-31 shows compaction curves for the same soil using Standard Proctor compaction (SPC) 

test parameters and Modified Proctor compaction (MPC) test parameters as listed in Table 5­

13. The typical compaction curves as presented in Figure 5-31 have the following 

characteristics: 

•	 Maximum dry density (γd-max) is the dry density corresponding to the peak of the 

compaction curve for a given type and amount of input energy. Note from Figure 5­

31, that the SPC γd-max is less than the MPC γd-max. Note from Table 5-13 that 

although the type of energy (impact) is the same for both SPC and MPC, the amount 

of energy in the MPC test is 4.5 times that of the SPC test. 

•	 Optimum moisture content (wopt) is the compaction water content at which the soil 

attains its maximum dry density for a given input energy. Note from Figure 5-31, 

that the SPC wopt is greater than the MPC wopt. Dry unit weight (density), 

γd , pcf (Mg/m3 ) 

a given Gs 

Compaction Moisture Content, w (%) 

Figure 5-31. Compaction curves (after Holtz and Kovacs, 1981). 

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Soils and Foundations – Volume I 5 - 74 December 2006 
•	 Zero air voids curve is the curve that corresponds to S=100% regardless of the 

amount or type of energy input. The importance of the zero-air-voids curve is that it 

denotes the limits of compaction, i.e., if the moisture content of a fill is too high for a 

given amount of input energy, the compacted fill may begin to “pump” as its voids 

become fully saturated with moisture. This can happen even at low moisture contents 

if the input energy is very large as may be the case with too many passes of a too 

heavy a piece of compaction equipment. Points on the zero air voids curve are 

calculated from the basic equation for dry unit weight given by Equation 5-18 by 

setting S=1 and choosing arbitrary values of compaction moisture content within the 

range of the compaction curve. 

G	 s γ G γ γ = w = s w

d	 (1 + e) ⎛ wG 1 + s ⎞ 5-18 ⎜ ⎟ ⎝ S ⎠

 

 where: Gs = specific gravity of solid particles 

 γw = unit weight of water 

 e = void ratio 

 w = water content expressed as a decimal 

 S = degree of saturation expressed as a decimal. 

Note that the S=100% (zero air voids) curve is calculated for a specific value of Gs, in 

this case 2.7. Curves corresponding to other degrees of saturation can be calculated 

in the same way by setting S=80 for the 80% saturation curve, S=60 for the 60% 

saturation curve and so forth. The saturation curve for a degree of saturation less than 

100% is often useful for developing compaction specifications for silty soils since 

such soils frequently have sharply peaked compaction curves. Therefore, they can 

begin to “pump” even though the degree of saturation is less then 100%. 

•	 Line of optimums - As its name suggests the “line of optimums” is obtained by 

passing a curve through the peaks of the compaction curves that were developed for a 

certain type of soil compacted at various energy input levels. Testing laboratories 

frequently develop such curves for various types of soil based on information in their 

job files. The line of optimums can be used as a guide for developing compaction 

specifications where no laboratory test data are available. 

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The above observations are true for all types of soils and apply to all methods of compaction. 

The most important concept about compaction curves as discussed above is that an increase 

in the amount of compaction (more energy) results in an increase in the maximum dry 

density and a corresponding decrease in the optimum moisture content. Therefore, this 

concept should be recognized when the geotechnical specialist is required to develop 

specifications for field compaction of soils. 

5.8.3 Implication of Laboratory Tests on Field Compaction Specifications 

With reference to Figure 5-31 it is obvious that for a given compaction curve the same dry 

unit weight can be obtained at two different compaction moisture contents, one below 

optimum and the other above optimum. For fine-grained soils this difference in moisture 

contents relates to a difference in soil structure that may affect engineering properties such as 

shear strength and permeability. 

It is very important that compaction specifications be given in terms of three parameters: the 

compaction energy (Standard or Modified Proctor), the desired dry density expressed as a 

percentage of the maximum dry density, and the compaction moisture content expressed as a 

range (+ or -) with respect to the optimum moisture content. For example, since the input 

energy of Modified Proctor is greater than the input energy of Standard Proctor (see Table 5­

13) the Modified Proctor curve plots above the Standard Proctor curve so that 95% of MPC 

γd-max may be greater than 100% of SPC γd-max. Likewise, a compaction moisture content of 1 

or 2% above optimum for modified Proctor compaction may be below the standard Proctor 

optimum moisture content. 

Unfortunately, laboratory compaction curves mainly serve as guidelines for field compaction. 

This approach is inconsistent because the impact type of energy input in the laboratory is not 

the same as the type of energy delivered by the equipment commonly used in construction. 

Figure 5-32 illustrates this point by presenting the types of compactive effort (static, 

vibratory, kneading) corresponding to the equipment typically used in practice. Note that 

none of the compaction processes in Figure 5-32 involves impact type of energy that is used 

to determine the compaction characteristics of the soils in a laboratory SPC or MPC test. 

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Soils and Foundations – Volume I 5 - 76 December 2006 
Figure 5-32. Compactors recommended for various types of soil and rock (Schroeder, 

1980). 

Due to the obvious disconnect between the types of energy in the laboratory and the field, 

some method is needed to express the laboratory-measured compaction parameters, i.e., 

maximum dry unit weight (γd-max) and optimum moisture content (wopt), in terms of field 

compaction. Most commonly, this relationship is achieved by so-called performance based 

or end-product specifications wherein a certain relative compaction, RC, also known as 

percent compaction, is specified. The RC is simply the ratio of the desired field dry unit 

weight, γd field, to the maximum dry density measured in the laboratory, γd max, expressed in 

percent as follows: 

The relative compaction, RC, is not the same as relative density, Dr, that was defined in 

Chapter 2. Relative density applies only to granular soils with fines less than 12% (ASTM D 

2049), while relative compaction is used across a wide variety of soils. Lee and Singh (1971) 

published the following relationship between RC and Dr based on a statistical evaluation of 

47 different granular soils compacted by using Modified Proctor energy (Wright, et al., 

2003). 

 Dr = 0% for RC = 80% 

 Dr = 100% for RC = 100% 

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γd field RC = x100% 5-19 

γd max 
Assuming a linear interpolation, the above relationship can be expressed as follows: 

Dr (%) = 5[RC( %) − 80] 5-20 

or 

Dr (%) RC (%) = 80 + 5-21

5 

In terms of Standard Proctor, Equations 5-20 and 5-21 are approximately as follows: 

Dr (%) = 5[RC ( %) − 85] 5-22 

or 

D (%) r RC(%) = 85 + 5-23

5 

Figure 5-33 presents the above equations in a graphical format. Table 5-14 presents the 

values of Dr for values of RC values ranging from 85% to 100% for MPC and from 90% to 

105% for SPC. 

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Soils and Foundations – Volume I 5 - 78 December 2006 Relative Density, Dr (%)

100 

Relative Density, Dr (%) 

95 

90 

85 

80 

75 

70 

65 

60 

55 

50 

45 

40 

35 

30 

25 

20 

15 

10 

5 

0 

Modified 

Proctor 

Standard 

Proctor 

At Dr = 0%, e = emax 

At Dr = 100%, e = emin 

75 80 85 90 95 100 

Relative Compaction, RC (%) 

Figure 5-33. Relative density, relative compaction and void ratio concepts. 

105 
Table 5-14 

Some values of Dr as a function of RC 

based on Modified and Standard Proctor Compaction Test 

RC (%) 

MPC (SPC)* 

Dr (%) RC (%) 

MPC (SPC)* 

Dr (%) RC (%) 

MPC (SPC)* 

Dr (%) 

85 (90) 25 90 (95) 50 95 (100) 75 

86 (91) 30 91 (96) 55 96 (101) 80 

87 (92) 35 92 (97) 60 97 (102) 85 

88 (93) 40 93 (98) 65 98 (103) 90 

89 (94) 45 94 (99) 70 99 (104) 95 

 100 (105) 100 

* MP: Modified Proctor; SP: Standard Proctor 

Figure 5-33 and Table 5-14 indicates that for every 1% increase in RC, the increase in Dr is 

5% regardless of compaction energy. This is rather significant when it is realized that the 

shear strength parameter, φ, of granular soils is a direct function of relative density as shown 

in Figure 5-22 and as illustrated by the following simple computations: 

•	 Based on Figure 5-22, the angle of internal friction for well-graded sands (SW soils) 

for values of Dr between 50% and 100% varies from 33o to 41o. From Table 5-14, 

values of Dr between 50% and 100% correspond to RC values of 90 and 100%, 

respectively for Modified Proctor and 95 and 105%, respectively for Standard 

Proctor. In other words, for SW soils, for every 1% increase in RC, the angle of 

internal friction increases by 0.8o. 

•	 Alternatively, the increase in the coefficient of friction, tan φ, would be tan (41o)/ tan 

(33o) = 1.33 or a 33% increase over a 10% change in RC. In other words, there is a 

3.3% increase in shear strength for every 1% increase in RC. 

Select materials are often specified in the construction of transportation facilities such as 

embankments, foundations, and pavement sub-bases and bases. The select materials are 

granular soils as discussed in Chapters 6 (Slope Stability), 7 (Approach Roadway 

Deformations), 8 (Shallow Foundations) and 10 (Earth Retaining Structures). The above 

simple example illustrates the importance of carefully specifying RC for such materials. RC 

values of 90 to 100% of standard Proctor values are commonly used. Based on Table 5-14, 

this range of RC corresponds to a Dr between 25% and 75%. 

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For most transportation applications, the RC value is prescribed in performance based 

specifications. In this case, it does not matter which equipment or type of compaction 

energy the contractor chooses to use as long as the end-product meets the specified RC. The 

prudent contractor would choose the equipment according to the type of soil. Often the 

contractor chooses to use the equipment he/she owns or is cheapest to lease or rent. 

Unfortunately, this equipment may not always be the most efficient equipment for the work. 

Figure 5-32 can be used as a preliminary guide in selecting the type of equipment and mode 

of compaction energy as a function of soil type. In Figure 5-32 the “100%” above the word 

clay on the left and the word sand on the right indicates boundaries for the range of soils 

types in between, e.g. 100% clay means that the soil to be compacted is all fine grained, 

therefore use of a sheepsfoot roller is recommended. The figure also suggests that a 

sheepsfoot roller can be used for various soil mixtures consisting of up to approximately 35% 

fine grained soils and 70 % coarse grained soils. 

An example of the influence of the choice of compaction equipment and energy is shown in 

Figure 5-34. Assume that Curve 1 is obtained from laboratory tests to develop the 

compaction curve for a borrow material that the contractor has identified for a given project. 

Further assume that the specification for the project requires that RC = 90%. If M represents 

the point of maximum dry density, γdmax, then RC=90% would mean that Points P and Y 

represent the limits of Curve 1 within which the contractor has to operate. In other words, 

the contractor cannot use compaction moisture contents less than a or c on the compaction 

moisture content axis. 

To properly evaluate the choice of the compaction equipment, the contractor should perform 

compaction tests at various RC values in the laboratory to develop a line of optimums and a 

family of curves similar to Curve 2 and 3 shown in Figure 5-34. Once this data is developed, 

then it can be observed from Figure 5-34, that the most economical water content would be 

that corresponding to point R along the line of optimums, i.e., the moisture content given by 

Point b on the X-axis. Point R represents the minimum compactive effort to attain RC=90%. 

To avoid inadequate compaction and risk failed field quality control tests, a prudent 

contractor usually aims to achieve somewhat higher dry density. Thus, the contractor often 

chooses to select a target curve similar to Curve 2 and aim to maintain moisture content in 

Zone B. 

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Soils and Foundations – Volume I 5 - 80 December 2006 
Dry unit weight (density), 

γd , pcf (Mg/m3 ) N 

X 

Gs = 2.7 

γdmax 

Y 

M 

P Q 

R S 

γd at 

RC=90% 

Compaction Moisture Content, w (%) 

Figure 5-34. Example evaluation of economical field compaction conditions (after 

Bowles, 1979) 

5.8.4 Engineering Characteristics of Compacted Soils 

Typical values for the engineering characteristics of compacted soils are given in Table 5-15. 

The values of the engineering properties refer to soils compacted to maximum dry density by 

using the standard Proctor test. The data in Table 5-15 are based on more than 1500 soil tests 

performed by the Bureau of Reclamation in Denver, CO. The large majority of soils were 

from 17 western states in the U.S. with some foreign soils (USBR, 1960). The background 

information for the values in Table 5-15 is given in the notes section of the table. 

FHWA NHI-06-088 5 – Laboratory Tests 

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Table 5-15 

Average engineering properties of compacted inorganic soils (after USBR, 1960)
 

USCS 

Standard Proctor Compaction 

(ASTM D 698/AASHTO T 99) 

As 

Compacted 

Cohesion, 

c′

psi (kPa) 

Saturated 

Cohesion, 

c′sat 

psi (kPa) 

Friction 

Angle, φ′

(deg) 

Void Ratio, e 

[Permeability, 

k (ft/yr)] Maximum Dry 

Density, pcf 

(kN/m3

) 

Optimum 

Moisture 

Content (%) 

GW >119 

(>18.7) <13.3 * * >38 * 

[27,000±13,000] 

GP >110 

(>17.3) <12.4 * * >37 * 

[64,000±34,000] 

GM >114 

(>17.9) <14.5 * * >34 * 

[>0.3] 

GC >115 

(>18.1) <14.7 * * >31 * 

[>0.3] 

SW 119"5 

(18.7"0.8) 13.3"2.5 5.7"0.6 (39"4) * 38"1 0.37"* 

[*] 

SP 110"2 

(17.3"0.3) 12.4"1.0 3.3"0.9 (23"6) * 37"1 0.50"0.03 

[>15.0] 

SM 114"1 

(17.9"0.2) 14.5"0.4 7.4"0.9 (51"6) 

2.9"1.0 

(20"7) 34"1 0.48"0.02 

[7.5±4.8] 

SM-SC 119"1 

(18.7"0.2) 12.8"0.5 7.3"3.1 (50"21) 

2.1"0.8 

(14"6) 33"4 0.41"0.02 

[0.8±0.6] 

SC 115"1 

(18.1"0.2) 14.7"0.4 10.9"2.2 (75"15) 

1.6"0.9 

(11"6) 31"4 0.48"0.01 

[0.3±0.2] 

ML 103"1 

(16.2"0.2) 19.2"0.7 9.7"1.5 (67"10) 

1.3"* 

(9"*) 32"2 0.63"0.02 

[0.59±0.23] 

ML-CL 109"2 

(17.1"0.3) 16.8"0.7 9.2"2.4 (63"17) 

3.2"* 

(22"*) 32"3 0.54"0.03 

[0.13±0.07] 

CL 108"1 

(17.0"0.2) 17.3"0.3 12.6"1.5 (87"10) 

1.9"0.3 

(13"2) 28"2 0.56"0.01 

[0.08±0.03] 

MH 82"4 

(12.9"0.6) 36.3"3.2 10.5"4.3 (72"30) 

2.9"1.3 

(20"9) 25"3 1.15"0.12 

[0.16±0.10] 

CH 94"2 

(14.8"0.3) 25.5"1.2 14.9"4.9 (103"34) 

1.6"0.86 

(11"6) 19"5 0.80"0.04 

[0.05±0.05] 

Notes: 

1. The entry ± indicates 90 percent confidence limits of the average value; * denotes insufficient data. 

2. For permeability, 1 ft/yr ≈ 10-6 cm/sec. 

3. All shear strengths, void ratios and permeabilities were determined on samples prepared at Standard 

Proctor maximum dry density and optimum moisture content. 

4. The values of cohesion, c′, and friction angle, φ′, are based on a straight-line Mohr strength envelope on 

an effective stress basis. The value c′sat, was obtained by saturating the sample and shearing it to failure. 

Consolidated-undrained (CU) triaxial tests were used to determine all the shear strengths. 

5. Since all laboratory tests, except large-sized permeability tests, were performed on the minus No. 4 (4.75 

mm) fraction of soil, data on average values for gravels are not available for most properties. However, 

an indication as to whether these average values will be greater than or less than the average values for 

the corresponding sand group are given in the table (note entries with > or < symbol). 

6. Void ratio was derived from the maximum dry density and specific gravity of the soil. 

7. In USCS, there are no upper boundaries of liquid limit of MH and CH soils. The maximum limits for 

MH and CH soils tested by USBR (1960) were 81% and 88%, respectively. Soils with higher liquid 

limits than these will have inferior engineering properties. 

FHWA NHI-06-088 5 – Laboratory Tests 

Soils and Foundations – Volume I 5 - 82 December 2006 
5.8.4.1 Effect of Increase in Moisture Content on Shear Strength of Compacted Soils 

The cohesion values, c′ (as-compacted) and c′sat (after saturation of compacted soil) listed in 

Table 5-15 are instructive in the context of the apparent cohesion concept discussed in 

Section 5.5.1.2. In the soil’s compacted state at optimum moisture content (OMC), the 

capillary stresses and the apparent mechanical forces assume their peak values at that 

particular compaction energy. Capillary stress, as discussed in Section 5.5.1.2, is due to 

surface tension in the water between individual soil grains. The magnitude of capillary stress 

is larger in fine-grained soils than coarse-grained soils as demonstrated by the increasing 

values of c′ in Table 5-15 as the soil type changes from granular to fine-grained. 

The same trend is observed with the c′sat values. However, the values of c′sat are 

approximately 10% (for CH soils) to 40% (for SM soils) of the corresponding c′ values. This 

drastic reduction in cohesive strength is attributable to the effect of capillary stresses being 

significantly reduced by the increase in moisture content required to reach saturation 

resulting in much lower apparent cohesive strengths. The reduction may also represent loss 

of apparent mechanical forces due to reduction in the interlocking of the particles because of 

the lubricating effect of water. 

Based on the above discussions, it is important to ensure that compacted soils are 

protected against increases in moisture content because the strength of such soils will 

decrease with associated detrimental effects on the facilities they support. 

5.9 ELASTIC PROPERTIES OF SOILS 

The stress-strain behavior of soils and rocks is highly nonlinear or inelastic. However, as 

indicated in Chapter 2, elastic theory provides a convenient first order approximation to 

stresses and strains induced in soils by external loads. A pair of elastic constants is required 

when elastic theory is used to solve such problems, e.g., elastic modulus (E) and Poisson’s 

ratio (ν), or shear modulus (G) and bulk modulus (B), or some other pair of elastic constants. 

The pair of E and ν is most widely used since both parameters are readily measurable. 

Consequently, many of the elastic equations in geotechnical engineering are formulated with 

this pair. Therefore, typical values of E and ν for soils are presented in this section. 

The elastic properties of soils may be measured from laboratory stress-strain curves such as 

those shown in Figure 5-16. The elastic properties, Es and ν, of a soil may be estimated from 

empirical relationships presented in Table 5-16 for preliminary design or for final design 

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Soils and Foundations – Volume I 5 - 83 December 2006 
where the prediction of deformation is not critical to the performance of the structure, i.e., 

when the structural design can tolerate the potential inaccuracies inherent in the correlations. 

The definition of Es is not always consistent for the various correlations and methods of insitu measurement. FHWA (2002a) provides additional details regarding the definition and 

determination of Es. Where evaluation of elastic settlement is critical to the design of the 

foundation or selection of the foundation type, in-situ methods such as pressuremeter or 

dilatometer tests should be used for evaluating the modulus of the impacted strata. 

The modulus of elasticity for normally consolidated cohesionless soils tends to increase with 

depth. An alternative method of defining the soil modulus for granular soils is to assume that 

the modulus, Es, increases linearly with depth, starting at zero at the ground surface, in 

accordance with the following equation: 

Es (tsf) = nh x z 5-24 

where: nh = rate of increase of soil modulus with depth as defined in Table 5-17 (tsf/ft) 

 z = depth in feet below the ground surface (ft) 

The formulation provided in Equation 5-24 is used primarily for analysis of lateral response 

or buckling of deep foundations. 

5.10 COMMON SENSE GUIDELINES FOR LABORATORY TESTING OF SOILS 

Sampling and testing of soils is one of the first and most important steps in the design and 

construction of all types of structures. Omissions or errors introduced here, if undetected, 

will be carried through the process of design and construction and will often result in costly 

and possibly unsafe facilities. Table 5-18 lists topics that should be considered for proper 

handling of samples, preparation of test specimens, and laboratory test procedures. Table 5­

18 should in no way be construed as being a complete list of guidelines to avoid possible 

errors and omissions in handling or testing of soil specimens; there are more. These are just 

some of the more common ones. 

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Table 5-16 

Elastic constants of various soils (after AASHTO 2004 with 2006 Interims)
 

Soil Type Typical Range of Young’s 

Modulus Values, Es (tsf) Poisson’s Ratio, ν 

Clay: 

Soft sensitive 

Medium stiff to stiff 

Very stiff 

25-150 

150-500 

500-1,000 

0.4-0.5 (undrained) 

Loess 

Silt 

150-600 

20-200 

0.1-0.3 

0.3-0.35 

Fine Sand: 

Loose 

Medium dense 

Dense 

80-120 

120-200 

200-300 

0.25 

Sand: 

Loose 

Medium dense 

Dense 

100-300 

300-500 

500-800 

0.20-0.36 

0.30-0.40 

Gravel: 

Loose 

Medium dense 

Dense 

300-800 

800-1,000 

1,000-2,000 

0.20-0.35 

0.30-0.40 

Estimating Es from SPT N-value 

Soil Type Es (tsf) 

Silts, sandy silts, slightly cohesive mixtures 

Clean fine to medium sands and slightly silty sands 

Coarse sands and sands with little gravel 

Sandy gravel and gravels 

4 N160 

7 N160 

10 N160 

12 N160 

Estimating Es (tsf) from qc static cone resistance 

Sandy soils 2qc where (qc is in tsf) 

Note: 1 tsf = 95.76 kPa 

Table 5-17 

Rate of increase of soil modulus with depth nh (tsf/ft) for sand 

(AASHTO 2004 with 2006 Interims)
 

Consistency Dry or Moist Submerged 

Loose 30 15 

Medium 80 40 

Dense 200 100 

Note: 1 tsf/ft = 314.7 kPa/m 

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Table 5-18 

Common sense guidelines for laboratory testing of soils
 

1. Protect samples to prevent moisture loss and structural disturbance. 

2.	 Carefully handle samples during extrusion; samples being extruded should be properly 

supported upon their exit from the tube. 

3.	 Avoid long term storage of soil samples in Shelby tubes. 

5.	 Properly number and identify samples. 

5.	 Store samples in properly controlled environments. 

6. Visually examine and identify soil samples after removal of smear from the sample surface. 

7.	 Use pocket penetrometer or miniature vane only for an indication of consistency not strength. 

8. Carefully select Arepresentative@ specimens for testing. 

9. Have a sufficient number of samples to select from. 

10.	 Always consult the field logs for proper selection of samples. 

11. Recognize disturbances caused by sampling, the presence of cuttings, drilling mud or other 

foreign matter. 

12. Do not depend solely on the visual identification of soils for classification. 

13.	 Always perform organic content tests when classifying soils as peat or organic. Visual 

classifications of organic soils may be very misleading. 

15.	 Do not dry soils in overheated or underheated ovens. 

15. Discard old worn-out equipment; old sieves for example, particularly fine (<No. 40) mesh ones 

need to be inspected and replaced often; worn compaction molds or compaction hammers 

should be checked and replaced if needed. An error in the volume of a compaction mold is 

amplified 30x when translated to unit volume. 

16. Performance of Atterberg limits tests requires carefully adjusted drop height of the liquid limit 

machine and proper rolling of plastic limit specimens. 

17. Do not use tap water for tests where distilled water is specified. 

18. Properly cure stabilization test specimens. 

19. Never assume that all samples are saturated as received. 

20. Perform saturation by applying properly staged back pressures of adequate magnitude. 

21. Use properly fitting o-rings, membranes, etc. in triaxial or permeability tests. 

22.	 Evenly trim ends and sides of undisturbed samples. 

23.	 Be careful to identify and report slickensides and natural fissures. 

25. Do not mistakenly identify failures due slickensides as shear failures. 

25. Do not use stress-strain curves from unconfined compression test results to determine elastic 

moduli. 

26.	 Incremental loading of consolidation tests should be performed only after the completion of the 

primary stage. 

27.	 Use proper loading rate for strength tests. 

28. Do not guesstimate e-log p curves from accelerated, incomplete consolidation tests. 

29.	 Avoid "reconstructing" soil specimens, disturbed by sampling or handling, for undisturbed 

testing. 

30. Correctly label all laboratory test specimens. 

31.	 Do not take shortcuts by using non-standard equipment or non-standard test procedures. 

32.	 Periodically calibrate testing equipment and maintain calibration records. 

33. Always test a sufficient number of samples to obtain representative results in variable material. 

34 Take proper precautions to assure the safety of personnel when performing any test procedure. 

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5.11 LABORATORY TESTS FOR ROCK 

5.11.1 Introduction 

This section provides information on common laboratory test methods for rock including 

testing equipment, general procedures related to each test, and parameters measured by the 

tests. Table 5-19 provides a list of commonly performed laboratory tests for rock associated 

with typical projects for highway applications. Although other laboratory test methods for 

rock are available including triaxial strength testing, rock tensile strength testing, and 

durability testing related to rock soundness, most design procedures for structural 

foundations and slopes on or in rock are developed based on empirical rules related to RQD, 

degree of fracturing, and the unconfined compressive strength of the rock. The use of more 

sophisticated laboratory testing for rock properties is usually limited to the most critical 

projects. Details on other laboratory testing procedures for rock are provided in FHWA, 

1997). Table 5-20 provides summary information on the typical rock index and performance 

tests. 

Table 5-19 

Common rock tests performed in the laboratory 

Test 

Category Name of Test ASTM Test Designation 

Point Load 

Strength Suggested method for evaluating point-load strength D 5731 

Compressive 

Strength 

Compressive strength of intact rock core specimen (in 

unconfined compression) D 2938 

Direct Shear 

Strength 

Laboratory direct shear strength tests for rock 

specimens under constant normal stress D 5607 

Durability Slake durability of shales and similar weak rocks D 4644 

StrengthDeformation Elastic moduli of intact rock core specimens in 

uniaxial compression D 3148 

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Table 5-20 

Summary information on laboratory test methods for rock (FHWA, 2002a) 

Test Procedure Applicable 

Rock Types 

Applicable 

Rock 

Properties 

Limitations / 

Remarks 

Point-Load Rock specimens in the form of Generally not Provides an Can be performed in 

Strength Test core, cut blocks, or irregular lumps 

are broken by application of 

concentrated load through a pair of 

spherically truncated, conical 

platens. 

appropriate for 

rock with 

uniaxial 

compressive 

strength less 

than 520 ksf (25 

MPa) 

index of 

uniaxial 

compressive 

strength 

the field with 

portable equipment 

or in the laboratory; 

in soft or weak rock, 

test results need to be 

adjusted to account 

for platen indentation 

Unconfined A cylindrical rock specimen is Intact rock core Uniaxial Simplest and fastest 

Compressive placed in a loading apparatus and compressive test to evaluate rock 

Strength of sheared under axial compression strength strength; fissures or 

Intact Rock with no confinement until peak other anomalies will 

Core load and failure are obtained. often cause 

premature failure 

Laboratory A rock specimen is placed in the Used to assess Peak and May need to perform 

Direct Shear lower half of the shear box and peak and residual in-situ direct shear 

Test encapsulated in either synthetic 

resin or mortar. The specimen 

must be positioned so that the line 

of shear force lies in the plane of 

the discontinuity to be 

investigated. The specimen is then 

mounted in the upper shear box 

and the normal load and shear 

force are applied. 

residual shear 

strength of 

discontinuity 

shear 

strength 

test if design is 

controlled by 

potential slip along a 

discontinuity filled 

with very weak 

material 

Elastic Procedure is similar to that for Intact rock core Modulus Modulus values (and 

Moduli of unconfined compressive strength and Poisson’s ratio) vary 

Intact Rock of intact rock. Lateral strains are Poisson’s due to nonlinearity of 

Core also measured ratio stress-strain curve. 

Slake Dried fragments of rock are placed Shale or other Index of 

Durability in a drum made of wire mesh that 

is partially submerged in distilled 

water. The drum is rotated, the 

sample dried, and the sample is 

weighed. After two cycles of 

rotating and drying, the weight loss 

and the shape of size of the 

remaining rock fragments are 

recorded. 

soft or weak 

rocks 

degradation 

potential of 

rock 

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5.11.2 Point-Load Strength Test 

The point load strength test is used to estimate the unconfined compressive strength of rock. 

Both core samples and fractured rock samples can be tested. The test is conducted by 

compressing a piece of the rock between two points on cone-shaped platens (see Figure 5-35) 

until the rock specimen breaks in tension between these two points. Each of the cone points 

has a 1/5 in (5 mm) radius of curvature and the cone bodies themselves include a 60° apex 

angle. The equipment is portable, and tests can be carried out quickly and inexpensively in 

the field. Because the point load test provides an index value for the compressive strength, 

usual practice is to calibrate the results with a limited number of uniaxial compression tests 

on prepared core samples. The point load test is also used with other index values to assess 

the degradation potential of shales. 

Figure 5-35. Point load strength test equipment (Wyllie, 1999). 

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If the distance between the contact points of the platens is D and the breaking load is P, then 

the point load strength, Is, is calculated as: 

P Is = 2 5-25 

D e 

where De is the equivalent core diameter given by: 

(1) De = D2

 for diametral tests; or 

(2) De = 4×A for axial, block, or lump tests where A = W×D. The area A is the minimum 

cross-sectional area of a lump sample for a plane through the platen contact points 

where W is the specimen width. 

The size-corrected point load strength index, Is(50) of a rock specimen is defined as the value 

of Is that would have been measured by a diametral test with D = 2 in (50 mm). For tests 

performed on specimens other than 2 in (50 mm) in diameter, the results can be standardized 

to the size-corrected point load strength index according to: 

Is(50) =kPLT Is 5-26 

The value of the size correction factor, kPLT, is given by: 

⎛ D ⎞

0.45 

k PLT =⎜ ⎟ (Din mm) 5-27 ⎝ 50 ⎠

It has been found that, on average, the uniaxial compressive strength, σc, is about 20 to 25 

times the point load strength index, with a value of 24 commonly used, i.e., 

σc = 24 Is(50) 5-28 

However, tests on many different types of rock show that the σc / Is(50) ratio can vary between 

15 and 50, especially for anisotropic rocks. Consequently, the most reliable results are 

obtained if a series of uniaxial calibration tests are carried out. Point load test results are not 

acceptable if the failure plane lies partially along a pre-existing fracture in the rock, or is not 

coincident with the line between the platens. For tests in weak rock where the platens indent 

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the rock, the test results should be adjusted by measuring the amount of indentation and 

correcting the distance D (Wyllie, 1999). 

5.11.3 Unconfined Compressive Strength of Intact Rock Core 

The unconfined compressive strength of intact rock core can be evaluated by using ASTM D 

2938. In this test, rock specimens of regular geometry, generally rock cores, are used. The 

rock core specimen is cut to length so that the length to diameter ratio is 2.5 to 3.0 and the 

ends of the specimen are machined flat. ASTM D 2938 provides tolerance requirements 

related to the flatness of the ends of the specimen, the perpendicularity of the ends of the 

specimens, and the smoothness of the length of the specimen. The specimen is placed in a 

loading frame, see Figure 5-36a. Axial load is then continuously applied to the specimen at a 

uniform rate until peak load and failure are obtained. The unconfined or uniaxial 

compressive strength of the specimen is calculated by dividing the maximum load carried by 

the specimen during the test by the initial cross-sectional area of the specimen. 

This test is more expensive than the point load strength test, but it is also more accurate with 

respect to in situ strength. Careful consideration of the design requirements should be made 

before deciding which test to perform, the unconfined compression test, a performance test, 

or the simpler point load strength test, an index test. 

5.11.4 Elastic Modulus of Intact Rock Core 

The test to determine the elastic modulus of intact rock is performed similarly to the 

unconfined compressive test discussed previously, except that deformation is monitored 

during application of load. This test is performed when it is necessary to estimate both the 

elastic modulus and the Poisson’s ratio of the intact rock core. Because of this dual purpose, 

it is common to measure both axial (or vertical) and lateral (or diametral) strain during 

compression. It is preferable to use strain gauges glued directly to the rock surface (see 

Figure 5-36b) as compared to LVDTs mounted on the platens since slight imperfections at 

the contact between the platens and the rock may lead to movements that are not related to 

strain in the rock (Wyllie, 1999). 

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 (a) (b) 

Figure 5-36. (a) Unconfined compression strength test on intact rock core, (b) Use of 

strain gage on intact rock core sample for measurement of stress-strain characteristics. 

(Photographs courtesy of Geomechanics Laboratory, University of Arizona). 

5.11.5 Laboratory Direct Shear Test 

The apparatus and procedures for direct shear testing are discussed in ASTM D 5607. The 

direct shear test is typically used to evaluate the shear strength of a rock discontinuity. 

Overall, the equipment for the direct shear test on rock is similar to that for soil including a 

direct shear testing machine, a device for applying normal pressure, and vertical and 

horizontal displacement monitoring devices. A schematic of the test set up is shown in 

Figure 5-37. For testing rock specimens, an encapsulating material such as a high strength 

gypsum cement is poured around the specimen in the upper and lower holding ring. The 

specimen is sheared as one holding ring is displaced horizontally with respect to the other 

such that the discontinuity surface is exactly parallel to the direction of the shear load. Load 

cells are used to monitor the shear force and LVDTs or dial gauges are used to monitor both 

horizontal and vertical deformation. Multiple LVDTs should be used to monitor vertical 

deformation and potential rotation of the specimen in the vertical plane. 

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Typically, the results of a direct shear test on rock are presented on two separate plots: one a 

plot of shear stress versus shear displacement and the other a plot of normal displacement 

versus shear displacement. Normal stresses should be adjusted to account for potential 

decreases in the shear contact area. After the sample is sheared, the sample is reset to its 

original position, the normal load is increased, and another test is performed. Each test will 

produce a pair of shear stress and normal stress values for both peak and residual conditions. 

The friction angle of the discontinuity surface can be evaluated from this data. 

Figure 5-37. Laboratory direct shear testing equipment for rock (Wyllie, 1999). 

FHWA NHI-06-088 5 – Laboratory Tests 

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5.12 ELASTIC PROPERTIES OF ROCKS 

Preliminary estimates of the elastic modulus of intact rock can be made from Table 5-21. 

Note that some of the rock types identified in the table are not present in the U.S. As 

discussed in Chapter 4, it is extremely important to use the elastic modulus of the rock 

mass for computation of in-situ displacements of rock under applied loads. Use of the 

intact modulus will result in unrealistic and unconservative estimates of displacement. 

Section 5.12.1 presents some guidance for estimating the elastic modulus of a rock mass. 

Poisson’s ratio for rock should be determined from tests on intact rock core. Where tests on 

rock core are not practical, Poisson’s ratio may be estimated from Table 5-22. 

Table 5-21 

Summary of elastic moduli for intact rock (AASHTO 2004 with 2006 Interims).
 

Rock Type 

No. of 

Values 

No. of 

Rock 

Types 

Elastic Modulus, Ei 

(psi ×106) 

Standard 

Deviation 

(psi ×106 Maximum Minimum Mean )

Granite 26 26 15.5 0.93 7.64 3.55 

Diorite 3 3 16.2 2.48 7.45 6.19 

Gabbro 3 3 12.2 9.8 11.0 0.97 

Diabase 7 7 15.1 10.0 12.8 1.78 

Basalt 12 12 12.2 5.20 8.14 2.60 

Quartzite 7 7 12.8 5.29 9.59 2.32 

Marble 14 13 10.7 0.58 6.18 2.49 

Gneiss 13 13 11.9 5.13 8.86 2.31 

Slate 11 2 3.79 0.35 1.39 0.96 

Schist 13 12 10.0 0.86 5.97 3.18 

Phyllite 3 3 2.51 1.25 1.71 0.57 

Sandstone 27 19 5.68 0.09 2.13 1.19 

Siltstone 5 5 5.76 0.38 2.39 1.65 

Shale 30 14 5.60 0.001 1.42 1.45 

Limestone 30 30 13.0 0.65 5.7 3.73 

Dolostone 17 16 11.4 0.83 5.22 3.44 

Note: 1 psi = 6.895 kPa 

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Table 5-22 

Summary of Poisson's ratio for intact rock (AASHTO 2004 with 2006 Interims)
 

Rock Type No. of 

Values 

No. of 

Rock 

Types 

Poisson's Ratio, ν Standard 

Deviation Maximum Minimum Mean 

Granite 22 22 0.39 0.09 0.20 0.08 

Gabbro 3 3 0.20 0.16 0.18 0.02 

Diabase 6 6 0.38 0.20 0.29 0.06 

Basalt 11 11 0.32 0.16 0.23 0.05 

Quartzite 6 6 0.22 0.08 0.14 0.05 

Marble 5 5 0.40 0.17 0.28 0.08 

Gneiss 11 11 0.40 0.09 0.22 0.09 

Schist 12 11 0.31 0.02 0.12 0.08 

Sandstone 12 9 0.46 0.08 0.20 0.11 

Siltstone 3 3 0.23 0.09 0.18 0.06 

Shale 3 3 0.18 0.03 0.09 0.06 

Limestone 19 19 0.33 0.12 0.23 0.06 

Dolostone 5 5 0.35 0.14 0.29 0.08 

5.12.1 Elastic Modulus of Rock Mass 

The elastic modulus of a rock mass (Em) shall be taken as the lesser of the intact modulus of a 

sample of rock core (Ei) or the modulus computed from one of the following equations: 

E (RMR-1 m = 145000 x [10 0)/40] 5-29

where: Em = Elastic modulus of the rock mass (psi) 

 RMR = Rock Mass Rating (see Chapter 4) 

Note that in almost all cases, the elastic modulus of the rock mass, Em, is less than the elastic 

modulus of the intact rock, Ei. 

The elastic modulus of the rock mass can also be determined from the following equation: 

Em = Em/Ei x Ei 5-30

 

where Ei is the elastic modulus of the intact rock. Em/Ei is basically a reduction factor to 

account for discontinuities in the rock mass and can be determined by using the guidance in 

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Soils and Foundations – Volume I 5 - 95 December 2006 
Table 5-23. In using Table 5-23, it is important that the elastic modulus for the intact rock, 

Ei, be determined from tests rather than by using the data in Table 5-21. For critical or large 

structures, determination of rock mass modulus (Em) by in-situ tests may be warranted. A 

discussion of suitable in-situ tests can be found in FHWA (2002a). 

Table 5-23 

Estimation of Em based on RQD (AASHTO 2004 with 2006 

Interims). 

RQD 

(Percent) 

Em/Ei 

Closed Joints Open Joints 

100 1.00 0.60 

70 0.70 0.10 

50 0.15 0.10 

20 0.05 0.05 

Note: Refer to Chapter 3 for guidance on determination of RQD and 

a description of rock joints. 

5.13 COMMON SENSE GUIDELINES FOR LABORATORY TESTING OF 

ROCKS 

As with soils, omissions or errors introduced during laboratory testing of rock, if undetected, 

will be carried though the process of design and construction and will often result in costly 

and possibly unsafe facilities. Table 5-24 lists topics that should be considered and given 

proper attention so that a reasonable assessment of the rock properties will be assured and an 

optimization of the geotechnical investigation can be realized in terms of economy, 

performance, and safety. Guidance in proper handling and storage of rock cores may be 

found in ASTM D 5079. 

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Table 5-24 

Common sense guidelines for laboratory testing of rocks
 

1. Provide protection of samples to avoid moisture loss and structural disturbance. 

2. Clearly indicate proper numbering and identification of samples. 

3. Store samples in controlled environments to prevent drying, overheating & freezing. 

5. Take care in the handling and selection of Arepresentative@ specimens for testing. 

5. Consult the field logs while selecting test specimens. 

6. Recognize disturbances and fractures caused by coring procedures. 

7. Maintain trimming and testing equipment in good operating condition. 

8. Use properly fitting, platens, o-rings and membranes in triaxial, uniaxial, and shear tests. 

9. Maintain tolerances in trimming of ends and sides of intact cores. 

10. Document frequency, spacing, conditions and infilling of joints and discontinuities. 

11. Periodically calibrate instruments used to measure load, deflection, temperatures and 

time. 

12. Use a properly-determined loading rate for strength tests. 

13. Photo document samples cores, fracture patterns and test specimens for possible use in a 

report. 

15. Carefully align and level all specimens in directional loading apparatuses and test frames. 

15. Record initial baselines, offsets, and eccentricities prior to testing. 

16. Save remnant rock pieces after destructive testing by uniaxial, triaxial and direct shear 

tests. 

17. Conduct nondestructive tests (i.e., porosity, unit weight, ultrasonics) prior to destructive 

strength tests (compression, tensile, shear). 

18. Take proper precautions to assure the safety of personnel when any test procedure is 

performed. 

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5.14 PRACTICAL ASPECTS FOR LABORATORY TESTING 

A poor understanding sometimes exists among geologists, structural engineers, and some 

foundation engineers about the type and amount of laboratory testing required for design of 

geotechnical features whether they happen to be structural foundations or earthwork. This 

weakness may render subsequent analyses useless. Organizations that have neither the 

proper testing facilities nor trained soils laboratory personnel should contract testing to 

competent AASHTO/ASTM certified private testing firms. This solution can be effective 

only if the project foundation designer can confidently request the necessary testing and 

review the results to select design values. A fair estimate of the costs associated with a 

private testing laboratory may be obtained by assuming the following number of person-days 

(pd) per test and multiplying by current labor costs: 

•	 visual description of an SPT sample including moisture content (0.05 pd), 

•	 visual description of a tube sample including moisture content and unit weight (0.1 

pd), 

•	 classification tests (0.7 pd), 

•	 undrained triaxial test (0.9 pd), 

•	 drained triaxial test (2.0 pd), 

•	 consolidation test (2.0 pd). 

These values include all work required to present a completed test result to the foundation 

designer. Alternatively, most private testing laboratories provide a schedule of services and 

associated costs that can be used to obtain a more accurate estimate of the cost of a proposed 

laboratory test program. 

Blanket consultant contracts "to perform testing necessary for design" usually result in 

unnecessarily large quantities of testing being performed, much of which does not apply to 

the project geotechnical issues. For example, if a multi span structure is crossing an area 

having a soft clay surface deposit underlain by sands, inordinate amounts of time and money 

should not be spent to determine the strength and consolidation parameters of the soft clay 

layer at pier and abutment locations in great detail. Generally a pile foundation will be 

designed by using SPT N-values and the only laboratory testing that may be needed in the 

soft clay layer may be to estimate drag forces on the piles. Also, non-standard strength 

testing such as torvanes, penetrometers, etc., which are not covered by ASTM or AASHTO 

standards, should not be permitted. Such devices should be used only for field index tests to 

determine consistency. 

FHWA NHI-06-088 5 – Laboratory Tests 

Soils and Foundations – Volume I 5 - 98 December 2006 
5.15 VARIABILITY OF MEASURED PROPERTIES 

Chapters 3, 4 and 5 presented methodologies to perform subsurface explorations, to describe 

identify, and classify subsurface materials, and to implement and interpret a laboratory test 

program. The topics covered in these chapters are presented in the temporal sequence in 

which they are performed on an actual project. Therefore, the geotechnical specialist should 

continuously evaluate the results of laboratory tests with respect to the initial subsurface 

model prepared as part of Step 1 of the flow chart presented in Figure 3-1 of Chapter 3. 

Table 5-25 presents the values of the coefficient of variation of measured properties that 

should be taken into consideration as the subsurface model is finalized for engineering 

design. The data in Table 5-25 can be used by the geotechnical specialist as follows: 

•	 Perform sensitivity (parametric) studies to evaluate the effect of variability in 

properties with respect to the subsurface profile and the type, magnitude and direction 

of the anticipated loading and establish the best-case and worst-case scenarios, 

•	 Evaluate the need to perform additional explorations, 

•	 Exercise judgment with respect to the results of engineering analyses and designs and 

convey the uncertainty to the project team, in particular to the structural engineer, and 

•	 Establish the need for instrumentation to monitor the performance of the facility 

during and after construction . 

The finalization of the subsurface model should be performed by a geotechnical specialist 

who is experienced in the design and construction aspects of the proposed facility. Active 

input should be sought from the field inspectors and the laboratory personnel who were 

actually involved in the collection of the data. At this stage, it is recommended that the 

geotechnical specialist seek a peer review from one or more qualified and experienced 

geotechnical specialist(s) with the specific purpose of having them evaluate whether or not 

the collected data and subsurface model is adequate to permit a cost-effective design of the 

facility. The experienced geotechnical specialist(s) can also provide information on potential 

value analysis alternatives (value engineering) for the design of the facility based on the 

collected data. The importance of such peer reviews cannot be overemphasized. Finally, the 

owner of the facility should be informed of the findings so that the owner can make decisions 

such as authorizing more field explorations and/or laboratory tests or modifying the facilities 

as appropriate based on the results of the geotechnical investigation thus far. 

FHWA NHI-06-088 5 – Laboratory Tests 

Soils and Foundations – Volume I 5 - 99 December 2006 
 

Table 5-25 

Values of coefficient of variation, V, for geotechnical properties and in situ tests 

(after Duncan, 2000) 

Measured or interpreted parameter value Coefficient of Variation, V 

(Note 1) 

Unit weight, γ 3 to 7 % 

Buoyant unit weight, γb 0 to 10 % 

Effective stress friction angle, φ′ 2 to 13 % 

Undrained shear strength, su 13 to 40 % 

Undrained strength ratio (su/po) 5 to 15 % 

Compression index, Cc 10 to 37 % 

Preconsolidation pressure, pc 10 to 35 % 

Hydraulic conductivity of saturated clay, k 68 to 90 % 

Hydraulic conductivity of partially-saturated clay, k 130 to 240 % 

Coefficient of consolidation, cv 33 to 68 % 

Standard penetration blow count, N 15 to 45 % 

Electric cone penetration test, qc 5 to 15 % 

Mechanical cone penetration test, qc 15 to 37 % 

Vane shear test undrained strength, suVST 10 to 20 % 

Note 1: Coefficient of Variation, V, is defined as standard deviation divided by the 

average (mean) value expressed as a percentage. 

FHWA NHI-06-088 5 – Laboratory Tests 

Soils and Foundations – Volume I 5 - 100 December 2006 
CHAPTER 6.0 

SLOPE STABILITY 

Ground stability must be assured prior to consideration of other foundation related items. 

Embankment foundation problems involve the support of the embankment by natural soil. 

Problems with embankments and structures occasionally occur that could be prevented by 

initial recognition of the problem and appropriate design. Stability problems most often 

occur when the embankment is to be built over soft soils such as low strength clays, silts, or 

peats. Once the soil profile, soil strengths, and depth of ground water table have been 

determined by field explorations and/or field and laboratory testing, the stability of the 

embankment can be analyzed and a factor of safety estimated. If the embankment is found to 

be unstable, measures can then be taken to stabilize the foundation soils. 

As illustrated in Figure 6-1, there are four major types of instability that should be considered 

in the design of embankments over weak foundation soils. Recommendations on how to 

recognize, analyze, and solve each of the first three problems are presented in this chapter. 

Lateral squeeze is more closely related to the evaluation of foundation deformation and is 

discussed in Chapter 7 (Approach Roadway Deformations). 

The stability problems illustrated in Figure 6-1 can be classified as “internal” or "external." 

"Internal" embankment stability problems generally result from the selection of poor quality 

embankment materials and/or improper placement of the embankment fills and/or improper 

placement requirements. The infinite slope failure mode is an example of an “internal” 

stability problem; often such a failure is manifested as sloughing of the surface of the slope. 

Internal stability can be assured through project specifications by requiring granular materials 

with minimum gradation and compaction requirements. An example of a typical 

specification for approach roadway construction is presented in Chapter 7. The failure 

modes shown in Figure 6-1b, c and d, can be classified as “external” stability problems. 

6.01 Primary Reference 

The primary reference for this chapter is as follows: 

FHWA (2001a). Soil Slope and Embankment Design Reference Manual. Report No. FHWA 

NHI-01-026, Authors: Collin, J. G., Hung, J. C., Lee, W. S., Munfakh, G., Federal Highway 

Administration, U.S. Department of Transportation. 

. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 1 December 2006 
(a) 

Embankment Fill 

Firm Soil 

Shallow translational failure 

(Infinite Slope condition) 

(b) 

(c) 

(d) 

Figure 6-1. Embankment failures: (a) Infinite slope failure in embankment fill, (b) 

Circular arc failure in embankment fill and foundation soil, (c) Sliding block failure in 

embankment fill and foundation soil, and (d) Lateral squeeze of foundation soil. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 2 December 2006 
 

6.1 EFFECTS OF WATER ON SLOPE STABILITY 

Very soft, saturated foundation soils or ground water generally play a prominent role in 

geotechnical failures in general. They are certainly major factors in cut slope stability and in 

the stability of fill slopes involving both “internal” and “external” slope failures. The effect 

of water on cut and fill slope stability is briefly discussed below. 

• Importance of Water 

Next to gravity, water is the most important factor in slope stability. The effect of 

gravity is known, therefore, water is the key factor in assessing slope stability. 

• Effect of Water on Cohesionless Soils 

In cohesionless soils, water does not affect the angle of internal friction (φ). The effect 

of water on cohesionless soils below the water table is to decrease the intergranular 

(effective) stress between soil grains (σ' n), which decreases the frictional shearing 

resistance (τ'). 

• Effect of Water on Cohesive Soils 

Routine seasonal fluctuations in the ground water table do not usually influence either 

the amount of water in the pore spaces between soil grains or the cohesion. The 

attractive forces between soil particles prevent water absorption unless external forces 

such as pile driving, disrupt the grain structure. However, certain clay minerals do react 

to the presence of water and cause volume changes of the clay mass. 

An increase in absorbed moisture is a major factor in the decrease in strength of 

cohesive soils as shown schematically in Figure 6-2. Water absorbed by clay minerals 

causes increased water contents that decrease the cohesion of clayey soils. These effects 

are amplified if the clay mineral happens to be expansive, e.g., montmorillonite. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 3 December 2006 
 Cohesive 

Strength 

Water 

Content 

Figure 6-2. Effect of water content on cohesive strength of clay. 

• Fills on Clays 

Excess pore water pressures are created when fills are placed on clay or silt. Provided 

the applied loads do not cause the undrained shear strength of the clay or silt to be 

exceeded, as the excess pore water pressure dissipates consolidation occurs, and the 

shear strength of the clay or silt increases with time. For this reason, the factor of safety 

increases with time under the load of the fill. 

• Cuts in Clay 

As a cut is made in clay the effective stress is reduced. This reduction will allow the 

clay to expand and absorb water, which will lead to a decrease in the clay strength with 

time. For this reason, the factor of safety of a cut slope in clay may decrease with time. 

Cut slopes in clay should be designed by using effective strength parameters and the 

effective stresses that will exist in the soil after the cut is made. 

• Slaking - Shales, Claystones, Siltstones, etc. 

Sudden moisture increase in weak rocks can produce a pore pressure increase in trapped 

pore air accompanied by local expansion and strength decrease. The "slaking" or 

sudden disintegration of hard shales, claystones, and siltstones results from this 

mechanism. If placed as rock fill, these materials will tend to disintegrate into a clay 

soil if water is allowed to percolate through the fill. This transformation from rock to 

clay often leads to settlement and/or shear failure of the fill. Index tests such as the jarslake test and the slake-durability test used to assess slaking potential are discussed in 

FHWA (1978). 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 4 December 2006 
6.2 DESIGN FACTOR OF SAFETY 

A minimum factor of safety as low as 1.25 is used for highway embankment side slopes. 

This value of the safety factor should be increased to a minimum of 1.30 to 1.50 for slopes 

whose failure would cause significant damage such as end slopes beneath bridge abutments, 

major retaining structures and major roadways such as regional routes, interstates, etc The 

selection of the design safety factor for a particular project depends on: 

• The method of stability analysis used (see Section 6.4.5). 

• The method used to determine the shear strength. 

• The degree of confidence in the reliability of subsurface data. 

• The consequences of a failure. 

• How critical the application is. 

6.3 INFINITE SLOPE ANALYSIS 

A slope that extends for a relatively long distance and has a consistent subsurface profile may 

be analyzed as an infinite slope. The failure plane for this case is parallel to the surface of 

the slope and the limit equilibrium method can be applied readily. 

6.3.1 Infinite Slopes in Dry Cohesionless Soils 

A typical section or “slice” through the potential failure zone of a slope in a dry cohesionless 

soil, e.g., dry sand, is shown in Figure 6-3, along with its free body diagram. The weight of 

the slice of width b and height h having a unit dimension into the page is given by: 

W = γ b h 6-1 

where γ is the effective unit weight of the dry soil. For a slope with angle β as shown in 

Figure 6-3, the normal (N) and tangential (T) force components of W are determined as 

follows: 

N = W cos β and 6-2 

T = W sin β 6-3 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 5 December 2006 
T 

W 

b 

h 

Slope 

Surface 

Failure 

Surface 

S 

N 

β

W 

S 

N β

Force Polygon 

N 

Figure 6-3. Infinite slope failure in dry sand. 

The available shear strength along the failure plane is given by: 

 

S = N tan φ 6-4 

The factor of safety (FS) is defined as the ratio of available shear strength to strength 

required to maintain stability. Thus, the FS will be given by: 

 

S N tan φ (W cos β) tan φ tan φ FS = = = = 6-5 T W sin β W sin β tan β

For an infinite slope analysis, the FS is independent of the slope depth, h, and depends only 

on the angle of internal friction, φ, and the angle of the slope, β. The slope is said to have 

reached limit equilibrium when FS=1.0. Also, at a FS = 1.0, the maximum slope angle will 

be limited to the angle of internal friction, φ. 

6.3.2 Infinite Slopes in c-φ Soils with Parallel Seepage 

If a saturated slope in a c-φ soil has seepage parallel to the surface of the slope as shown in 

Figure 6-4, the same limit equilibrium concepts may be applied to determine the FS, which 

will now depend on the effective normal force (N'). In the following analysis, effective shear 

strength parameters, c' and φ' are used. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 6 December 2006 
 

 

h cos2β

Seepage Flow 

Failure Surface 

Slope Surface 

h 

β

W h

N'+U 

Pore Water Force 

U = γwbh cosβ

b 

T 

Figure 6-4. Infinite slope failure in a c-φ soil with parallel seepage. 

From Figure 6-4, the pore water force acting on the base of a typical slice having a unit 

dimension into the page is: 

 

where h is any depth less than or equal to the depth of saturation and b is a unit width. 

The available frictional strength, S, along the failure plane will depend on φ' and the effective 

normal force, N' =N-U, where N is the total normal force. The equation for S is: 

The factor of safety for this case will be: 

By substituting W = γsat b h into the above expression and rearranging terms, the FS is given 

by: 

b U = ( γ 2 = γ w h cos β ) w b h cos β 6-6 

cos β

 

b S = c' + ( N - U ) tan φ' 6-7 

cos β

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 7 December 2006 

S (c' b/cos β ) + ( N - U ) tan φ' FS = = 6-8 T W sin β
c' + h ( γsat - γ cos 2 ) ( β φ FS = 

w ) tan ' 6-9

γ h sin β cos β sat 

where γ' = (γsat - γw). 

For c' = 0, the above expression may be simplified to: 

 

γ ' tan φ' FS = 6-10

γ tanβ sat 

From Equation 6-10 it is apparent that for a cohesionless material with parallel seepage, the 

FS is also independent of the slope depth, h, just as it is for a dry cohesionless material as 

given by Equation 6-5. The difference is that the FS for the dry material is reduced by the 

factor γ'/γsat for saturated cohesionless materials to account for the effect of seepage. For 

typical soils, this reduction will be about 50 percent in comparison to dry slopes. 

The above analysis can be generalized if the seepage line is assumed to be located at a 

normalized height, m, above the failure surface where m =z/h. In this case, the FS is: 

 

c' + h cos 2 β [ ( 1- m ) γ γ m + m ' ] tan φ' FS = 6-11

h sin β cos β [ ( 1- m ) γ m + m γ sat ]

and γsat and γm are the saturated and moist unit weights of the soil below and above the 

seepage line. The above equation may be readily reformulated to determine the critical depth 

of the failure surface in a c'-φ' soil for any seepage condition. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 8 December 2006 
6.4 CIRCULAR ARC FAILURE 

Experience and observations of failures of embankments constructed over relatively deep 

deposits of soft soils have shown that when failure occurs, the embankment sinks down, the 

adjacent ground rises and the failure surface follows a circular arc as illustrated in Figure 6-5. 

LW Center 

Fill Weight 

Force 

Fill Surface LS 

After Failure 

Failure 

Case 

Fill 

Direction 

Soft Clay of	 Failure 

Movement 

Sum of Shear Strength Resistance Along Arc Force 

Figure 6-5. Typical circular arc failure mechanism. 

At failure the driving and resistance forces act as follows: 

•	 The force driving movement consists of the embankment weight. The driving moment is 

the product of the weight of the embankment acting through its center of gravity times 

the horizontal distance from the center of gravity to the center of rotation (LW). 

•	 The resisting force against movement is the total shear strength acting along the failure 

arc. The resisting moment is the product of the resisting force times the radius of the 

circle (LS). 

The factor of safety against slope instability is equal to the ratio of the resisting moment to 

driving moment. 

 

Failure takes place when the factor of safety is less than 1, i.e., the driving moment > 

resisting moment. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 9 December 2006 

Total Shear Strength × L S Resisting Moment Factor of Safety = = 6-12 Weight Force × L W Driving Moment
 

6.4.1 Simple Rule of Thumb for Factor of Safety 

A rule of thumb based on simplified bearing capacity theory can be used to make a 

preliminary "guestimate" of the factor of safety (FS) against circular arc failure for an 

embankment built on a clay foundation without presence of free water. The rule of thumb is 

as follows: 

 

6 c FS ≅ 6-13

γ Fill Fill × H

Where: c = unit cohesion of clay foundation soil (psf) 

γFill = unit weight fill (pcf) 

HFill = height of fill (feet) 

Since the rule of thumb assumes that there is no influence from groundwater, c and γFill are 

effective stress parameters. 

For example, the factor of safety for the proposed embankment illustrated in Figure 6-6 can 

be computed as follows: 

(6)(1,100 psf) FS = = 1.69 Use Rule of Thumb 6-13 (130 pcf)(30 ft) 

Figure 6-6. Example proposed embankment. 

The factor of safety computed by using this rule of thumb should never be used for final 

design. This simple equation obviously does not take into account such factors as fill 

strength or fill slope angle and does not identify the location of a critical failure surface. If 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 10 December 2006 
the factor of safety computed by using the rule of thumb is less than 2.5, a more 

sophisticated stability analysis is required. 

However, this rule of thumb can be helpful very early in the design stage to make a quick 

preliminary check on whether stability may be a problem and if more detailed analyses 

should be conducted. It can also be of use in the field while borings and sampling are being 

performed. For example, if in-situ vane shear tests are being carried out as part of the field 

investigation for a proposed embankment, the geotechnical specialist can use the vane 

strength with Equation 6-13 to estimate the FS in the field. This estimate can aid in directing 

the drilling, sampling, and testing program while the drill crew is at the site and help insure 

that critical strata are adequately explored and sampled. Finally, the FS calculated by the 

rule of thumb can be used to check for gross errors in computer output or input. 

6.4.2 Stability Analysis Methods (General) 

There are several available methods that can be used to perform a circular arc stability 

analysis for an approach embankment over soft ground. The simplest basic method is known 

as the Normal or Ordinary Method of Slices, also known as Fellenius’ method (Fellenius, 

1936) or the Swedish circle method of analysis. The Ordinary Method of Slices can easily be 

performed by hand calculations and is also a method by which the computation of driving 

and resisting forces is straightforward and easily demonstrated. For this method, the failure 

surface is assumed to be the arc of a circle as shown in Figure 6-7 and the factor of safety 

against sliding along the failure surface is defined as the ratio of the moment of the total 

available resisting forces on the trial failure surface to the net moment of the driving forces 

due to the embankment weight, that is: 

Note that since the method consists of computing the driving and resisting forces along the 

failure arc, the moment arm R is the same for both the driving and resisting forces. Thus, 

Equation 6-14 reduces to: 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 11 December 2006 

Sum of Resisting Forces × Moment Arm (R) FS = 6-14 Sum of Driving Forces × Moment Arm (R) 

Sum of Resisting Forces FS = 6-14a Sum of Driving Forces 
 

 

 

 

 

 

 

 

Figure 6-7. Geometry of Ordinary Method of Slices. 

For slope stability analysis, the mass within the failure surface is divided into vertical slices 

as shown in Figures 6-7 and 6-8. A typical vertical slice and its free body diagram is shown 

in Figure 6-9 for the case where water is not a factor. The case with the presence of water is 

shown in Figure 6-10. The following assumptions are then made in the analysis using 

Ordinary Method of Slices: 

1. The available shear strength of the soil can be adequately described by the 

Mohr-Coulomb equation: 

τ = c + (σ – u) tan φ 6-15

where: 

τ = effective shear strength 

c = cohesion component of shear strength 

(σ - u) tan φ = frictional component of shear strength 

σ = total normal stress on the failure surface at the base of a slice due to 

the weight of soil and water above the failure surface 

u = water uplift pressure against the failure surface 

φ = angle of internal friction of soil 

tan φ = coefficient of friction along failure surface 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 12 December 2006 
2. The factor of safety is the same for all slices. 

3. The factors of safety with respect to cohesion (c) and friction (tan φ) are equal. 

4. Shear and normal forces on the sides of each slice are ignored. 

5. The water pressure (u) is taken into account by reducing the total weight of 

the slice by the water uplift force acting at the base of the slice. 

Equation 6-15 is expressed in terms of total strength parameters. The equation could easily 

have been expressed in terms of effective strength parameters. Therefore, the convention to 

be used in the stability analysis, be it total stress or effective stress, should be chosen and 

specified. In soil problems involving water, the engineer may compute the normal and 

tangential forces by using either total soil weights and boundary water forces (both buoyancy 

and unbalanced hydrostatic forces) or submerged (buoyant) soil weights and unbalanced 

hydrostatic forces. The results are the same. When total weight and boundary water forces 

are used, the equilibrium of the entire block is considered. When submerged weights and 

hydrostatic forces are used, the equilibrium of the mineral skeleton is considered. The total 

weight notation is used herein as this method is the simplest to compute. 

6.4.3 Ordinary Method of Slices - Step-By-Step Computation Procedure 

To compute the factor of safety for an embankment by using the Ordinary Method of Slices, 

the step-by-step computational procedure is as follows: 

Figure 6-8. Example of dividing the failure mass in slices. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 13 December 2006 
Step 1. Draw a cross-section of the embankment and foundation soil profile on a scale 

of either 1-inch = 10 feet or 1-inch = 20 feet scale both horizontal and vertical. 

Step 2. Select a circular failure surface such as shown in Figure 6-7. 

Step 3. Divide the circular mass above the failure surface into 10 - 15 vertical slices as 

illustrated in Figure 6-8. 

 

 To simplify computation, locate the vertical sides of the slices so that the bottom of 

any one slice is located entirely in a single soil layer or at the intersection of the 

ground water level and the circle. 

 Locate the top boundaries of vertical slices at breaks in the slope. The slice widths 

do not have to be equal. For convenience assume a one-foot (0.3 m) thick section 

of embankment. This unit width simplifies computation of driving and resisting 

forces. 

 Also, as shown in Figure 6-9 and 6-10 the driving and resisting forces of each slice 

act at the intersection of a vertical line drawn from the center of gravity of the slice 

to the failure circle to establish a centroid point on the circle. Lines (called rays) 

are then drawn from the center of the circle to the centroid point on the circular 

arc. The α angles are then measured from the vertical to each ray. 

 When the water table is sloping, use Equation 6-16 to calculate the water pressure 

on the base of the slice: 

u = hw γw cos2 αw 6-16

 where: αw = slope of water table from horizontal in degrees. 

 hw = depth from ground water surface to the centroid point on the circle. 

Step 4: Compute the total weight (WT) of each slice. 

For illustration, the resisting and driving forces acting on individual slices with and 

without water pressure are shown on Figures 6-9 and 6-10. 

 To compute WT, use total soil unit weight, γt, both above and below the water 

table. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 14 December 2006 
C = Cohesion Along 

Slice Base 

Tan φ = Coefficient of 

Friction Along Slice 

Base 

WT = Total Slice Weight 

N = WT Cos α 

T = WT Sin α 

Figure 6-9. Forces on a slice without water effect. 

T = WT Sin α 

C = Cohesion Along 

Slice Base 

Tan φ = Coefficient of 

Friction Along Slice 

Base 

WT = Total Slice Weight 

(Soil + Water) 

N = WT Cos α - ul 

Figure 6-10. Forces on a slice with water effect. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 15 December 2006 
WT = γt × Average Slice Height × Slice Width 6-17

 For example: Assume
 

 γt = 120 pcf (18.9 kN/m3

) 

 Average height of slice = 10 ft (3 m) 

 Slice width = 10 ft (3 m) 
 

 Then for a unit thickness into the plane of the paper, WT = (120 pcf) (10 ft) (10 ft) 

(1 ft) = 12,000 lbs (53.3 kN) 

Step 5: Compute frictional resisting force for each slice depending on location of 

ground water table. 

N = WT cos α 6-18a

N′ = WT cos α – ul 6-18b 

 N = total normal force acting against the slice base 

N′ = effective normal force acting against the slice base 

 WT = total weight of slice (from Step 4 above) 

α = angle between vertical and line drawn from circle center to midpoint 

(centroid) of slice base (Note: α is also equal to the angle between the 

horizontal and a line tangent to the base of the slice) 

 u = water pressure on the base of the slice = average height of water, hw × γw. 

Use γw = 62.4 pcf (9.8 kN/m3

) 

l = arc length of slice base. To simplify computations, take l as the secant to 

the arc. 

 u l = water uplift force against base of the slice per unit thickness into the 

plane of the paper. 

 φ = internal friction angle of the soil. 

 tan φ = coefficient of friction along base of the slice. 

 Note that the effect of water is to reduce the normal force against the base of 

the slice and thus reduce the frictional resisting force. To illustrate this 

reduction, take the same slice used in Step 4 and compute the friction resistance 

force for the slice with no water and then for the ground water table located 5 feet 

above the base of the slice. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 16 December 2006 
 Assume: φ = 25o α = 20o l = 11 ft (3.3 m) 

 If there is no water in the slice, u l = 0 and Equation 6-18b reverts to Equation 6­

18a and the total frictional resistance can be computed as follows: 

N = W cos α = (12,000 lbs) (cos 20o T ) = 11,276 lbs (50.18 kN) 

 N tan φ = (11,276 lbs) (tan 25o

) = 5,258 lbs (23.4 kN) 

 If there is 5-ft of water above the midpoint of the slice, Equation 6-18b is used 

directly and the effective frictional resistance is computed as follows: 

 u l = (hw)(γw)(l) = (5 ft)(62.4 pcf))(11 ft)(1 ft) = 3,432 lbs (15.3 kN) 

N′ = WT cos α - u l = 11,276 lbs - 3,432 lbs = 7,844 lbs (34.9 kN) 

N′ tan φ = (7,844 lbs) (tan 25°) = 3,658 lbs (16.3 kN) 

Step 6: Compute cohesive resisting force for each slice. 

c = cohesive soil strength 

 l = length of slice base 

 Example: c = 200 psf (9.6 kPa) 

l = 11 ft (3.6 m) 
 

 cl = (200 psf)(11 ft)(1 ft) = 2,200 lbs (9.8 kN) 

Step 7: Compute tangential driving force, T, for each slice. 

T = WT sin α 6-19

 T is the component of total weight of the slice, WT, acting tangent to the slice base. 

T is the driving force due to the weight of both soil and water in the slice.

 Example: Given WT = 12,000 lbs (53.3 kN) 

 α = 20o

 T = WT sin α = (12,000 lbs)(sin 20o

) = 4,104 lbs (18.2 kN) 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 17 December 2006 
Step 8: Sum resisting forces and driving forces for all slices and compute factor of 

safety. 

 Tabular computation forms for use in performing a method of slices stability 

analysis by hand are included on Figures 6-11 and 6-12. 

 

∑ Resisting Forces ∑ N′ tan φ + ∑ c 1 FS =	 = 6-20 ∑ Driving Forces ∑ T 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 18 December 2006 

 

 

 

 

 

γi = unit weight of layer i 

hi = height of layer at center of slice 

Wi = partial weight = b hi γi 

∑ Wi = total weight of slice WT 

Slice No. b hi γi Wi ∑Wi = WT 

Figure 6-11a. Tabular form for computing weights of slices. 
Slice 

No. 

WT 

(from 

Table 

6-11a) 

l α c φ u ul WTcosα N′ = 

WTcosα -ul 

N′tanφ 

Σ 

Σ ( W T cos α − ul ) tan φ + Σ cl Σ N ′ tan φ + Σ cl FS = = =

. 

Σ W sin T α Σ W sin T α

WT 

α 

l 

Legend: Refer to Figure 6-10 for definition of various slice quantities 

W

T = Total weight of Slice (soil + water) 

l 

= 

Base length of the slice 

c 

= 

Cohesion at base of slice 

φ = angle of internal friction 

= 

pore water pressure at base of slice 

Figure 6-11b. Tabular form for calculating factor of safety by Ordinary Method of Slice

FHWA NHI-06-088 u Soils and Foundations – Volume I 6 - 19 
6.4.4 Recommended Stability Methods 

The basic static forces on a typical slice are shown in Figure 6-12. The limit equilibrium 

method of slices is based on the principles of statics, i.e., summation of moments, vertical 

forces, and horizontal forces. The Ordinary Method of Slices ignores both interslice shear 

(IS) and interslice normal (IN) forces and satisfies only moment equilibrium. There are many 

other methods available for performing a slope stability analysis besides the Ordinary 

Method of Slices. These include the Bishop Method (Bishop, 1955), the Simplified Janbu 

Method (Janbu, 1954) and the Spencer Method (Spencer, 1967). These methods are 

primarily variations and refinements of the Ordinary Method of Slices. The differences 

among these more refined methods lie in the assumptions made regarding the interslice shear 

and normal forces acting on the sides of slices. The Bishop Method, also known as the 

Simplified Bishop Method, includes interslice normal forces (IN) but ignores interslice shear 

(IS) forces. Again, Bishop’s method satisfies only moment equilibrium. The Simplified 

Janbu Method is similar to the Bishop Method in that it includes the interslice normal (IN) 

forces and ignores the interslice shear (IS) forces. The difference between the Bishop Method 

and the Simplified Janbu Method is that the Simplified Janbu Method satisfies only 

horizontal force equilibrium, as opposed to moment equilibrium. The Spencer Method 

considers both normal and shear interslice side forces as well as moments. Therefore the 

Spencer Method is theoretically more rigorous than the other methods. 

 

 

 

 

WT Total weight of slice 

N Normal force at base of slice 

IN T Tangential force at base of slice 

IS Interslice shear (vertical) force 

IN Interslice normal (horizontal) force IS 

IS 

IN 

WT 

N 

T 

Legend: 

Figure 6-12. Typical static forces on a slice of sliding mass without seepage. 

The Ordinary Method of Slices is more conservative and gives unrealistically lower factors 

of safety than the Bishop Method or the other more refined methods. The only reason for 

inclusion of the Ordinary Method of Slides here is to demonstrate the principles of slope 

stability. For purely cohesive soils the Ordinary Method of Slices and Bishop’s method give 

identical results. For soils that have frictional strength, the Bishop Method should be used as 

a minimum. While none of the methods is 100 percent correct theoretically, currently 

available procedures such as Bishop’s method, Janbu’s Simplified method or Spencer’s 

method are sufficiently accurate for practical analysis and design. For more information on 

these and other slope stability methods, the reader is referred to FHWA (2001a). 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 20 December 2006 
 

The method of analysis that should be used to determine a factor of safety depends on the 

soil type, the source of the soil strength parameters, the level of confidence in the values, and 

the type of slope that is being designed. Slope stability analyses should be performed only 

by qualified and experienced geotechnical specialists. Guidelines recommended for the 

analysis of slope stability are given in Table 6-1. 

Table 6 -1. Slope stability guidelines for design 

Foundation 

Soil Type 

Type of 

Analysis 

Source of Strength Parameters 

(see Chapter 5) 

Remarks 

(see Note 1) 

Short-term 

(embankments 

on soft clays – 

immediate end 

of construction 

– φ = 0 

analysis). 

• UU or field vane shear test or 

CU triaxial test. 

• Use undrained strength 

parameters at po 

Use Bishop Method. An angle 

of internal friction should not be 

used to represent an increase of 

shear strength with depth. The 

clay profile should be divided 

into convenient layers and the 

appropriate cohesive shear 

strength assigned to each layer. 

Stage Use Bishop Method at each 

Cohesive 

construction 

(embankments 

on soft clays – 

build 

embankment in 

stages with 

waiting periods 

to take 

advantage of 

• CU triaxial test. Some 

samples should be 

consolidated to higher than 

existing in-situ stress to 

determine clay strength gain 

due to consolidation under 

staged fill heights. 

• Use undrained strength 

stage of embankment height. 

Consider that clay shear strength 

will increase with consolidation 

under each stage. Consolidation 

test data needed to estimate 

length of waiting periods 

between embankment stages. 

Piezometers and settlement 

devices should be used to 

clay strength parameters at appropriate po monitor pore water pressure 

gain due to for staged height. dissipation and consolidation 

consolidation). during construction. 

Long-term • CU triaxial test with pore Use Bishop Method with 

(embankment water pressure measurements combination of cohesion and 

on soft clays or CD triaxial test. angle of internal friction 

and clay cut • Use effective strength (effective strength parameters 

slopes). parameters. from laboratory test). 

Existing 

failure planes 

• Direct shear or direct simple 

shear test. Slow strain rate 

and large deflection needed. 

• Use residual strength 

parameters. 

Use Bishop, Janbu or Spencer 

Method to duplicate previous 

shear surface. 

Granular All types 

• Obtain effective friction angle 

from charts of standard 

penetration resistance (SPT) 

versus friction angle or from 

direct shear tests. 

Use Bishop Method with an 

effective stress analysis. 

Note 1: Methods recommended represent minimum requirement. More rigorous methods such as 

Spencer’s method should be used when a computer program has such capabilities. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 21 December 2006 
6.4.5 Remarks on Safety Factor 

For side slopes of routine highway embankments, a minimum design safety factor of 1.25 as 

determined by the Ordinary Method of Slices is used. For slopes that would cause greater 

damage upon failure, such as end slopes beneath bridge abutments, major retaining 

structures, and major roadways such as regional routes, interstates, etc., the design safety 

factor should be increased to at least 1.30 to 1.50. For cut slopes in fine-grained soils, which 

can lose shear strength with time, a design safety factor of 1.50 is desirable. 

6.5 CRITICAL FAILURE SURFACE 

The step-by-step procedure presented in the preceding section illustrates how to compute the 

factor of safety for one selected circular arc failure surface. The complete analysis requires 

that a large number of assumed failure surfaces be checked in order to find the critical one, 

i.e., the surface with the lowest factor of safety. This task would obviously be a tedious and 

time consuming operation if done by hand. Therefore a computer program becomes a 

valuable tool for performing such computations. Any method for stability analysis is easily 

adapted to computer solution. For critical circle methods a grid of possible circle centers is 

defined, and a range of radius values established for each. The computer can be directed to 

perform stability analyses for each circle center over the range of radii and then to print out 

all the safety factors or just the minimum one and its radius. A plot of minimum safety factor 

for each circle center in the form of contours can be used to define the location of the most 

critical circle and the minimum safety factor as shown in Figure 6-13. The radius of the most 

critical surface can be used to locate the intersection points of the circle with the ground 

surface above and below the slope. This is useful in identifying structures above and below 

the slope that may be potentially impacted by slope instability. 

Figure 6-13 shows just one of several ways that computer programs can be used to search for 

the most critical failure surface. It is beyond the scope of this manual to discuss these in 

detail. However, the following points should be noted as one uses a computer program for 

locating the most critical failure surface: 

1.	 Check multiple circle center locations and compare the lowest safety factors. There 

may be more than one “local” minimum and a single circle center location may not 

necessarily locate the lowest safety factor for the slope. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 22 December 2006 
 

Figure 6-13. Location of critical circle by plotting contours of minimum safety factors 

for various trial circles. 

2.	 Search all areas of the slope to find the lowest safety factor. The designer may find 

multiple areas of the slope where the safety factors are low and comparable. In this 

case, the designer should try to identify insignificant failure modes that lead to low 

safety factors for which the consequences of failure are small. This is often the case 

in cohesionless soils, where the lowest safety factor is found for a shallow failure 

plane located close the slope face. 

3.	 Review the soil stratigraphy for “secondary” geological features such as thin 

relatively weak zones where a slip surface can develop. Often, circular failure 

surfaces are locally modified by the presence of such weak zones. Therefore 

computer software capable of simulating such failures should be used. Some of the 

weak zones may be man-made, e.g., when new fills are not adequately keyed into 

existing fills for widening projects. 

4.	 Conduct stability analyses to take into account all possible loading and unloading 

schemes to which the slope might be subjected during its design life. For example, if 

the slope has a detention basin next to it, then it might be prudent to evaluate the 

effect of water on the slope, e.g., perform an analysis for a rapid drawdown condition. 

5.	 Use the drained or undrained soil strength parameters as appropriate for the 

conditions being analyzed 

6.	 Use stability charts to develop a “feel” for the safety factor that may be anticipated. 

Stability charts are discussed in the next section. Such charts may also be used to 

verify the results of computer solutions. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 23 December 2006 
6.6 DESIGN (STABILITY) CHARTS 

Slope stability charts are useful for preliminary analysis to compare alternates that may be 

examined in more detail later. Chart solutions also provide a quick means of checking the 

results of detailed analyses. Engineers are encouraged to use these charts before performing 

a computer analysis in order to determine the approximate value of the factor of safety. The 

chart solution allows some quality control and a check for the subsequent computergenerated solutions. 

Slope stability charts are also used to back-calculate strength values for failed slopes, such as 

landslides, to aid in planning remedial measures. In back-calculating strength values a factor 

of safety of unity is assumed for the conditions at failure. Since soil strength often involves 

both cohesion and friction, there are no unique values that will give a factor of safety equal to 

one. Therefore, selection of the most appropriate values of cohesion and friction depends on 

local experience and judgment. Since the friction angle is usually within a narrow range for 

many types of soils and can be obtained by laboratory tests with a certain degree of 

confidence, it is generally fixed for the back-calculations in practice and the value of 

cohesion is varied until a factor of safety of one is obtained. 

The major shortcoming in using design charts is that most charts are for ideal, homogeneous 

soil conditions that are not typically encountered in practice. Design charts have been 

devised with the following general assumptions: 

1. Two-dimensional limit equilibrium analysis. 

2. Simple homogeneous slopes. 

3. Slip surfaces of circular shapes only. 

It is imperative that the user understands the underlying assumptions for the charts 

before using them for the design of slopes. 

Regardless of the above shortcomings, many practicing engineers use these charts for nonhomogeneous and non-uniform slopes with different geometrical configurations. To do this 

correctly, one must use an average slope inclination and weighted averages of c, φ, or c', φ' 

or cu calculated on the basis of the proportional length of slip surface passing through 

different relatively homogeneous layers. Such a procedure is extremely useful for 

preliminary analyses and saves time and expense. In most cases, the results are checked by 

performing detailed analyses using more suitable and accurate methods, for example, one of 

the methods of slices discussed previously. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 24 December 2006 
6.6.1 Historical Background 

Some of the first slope stability charts were published by Taylor (1948). Since then various 

charts were developed by many investigators. Two of the most common stability charts are 

presented in this manual. These were developed by Taylor (1948) and Janbu (1968). 

6.6.2 Taylor’s Stability Charts 

Taylor’s Stability Charts (Taylor, 1948) were derived from solutions based on circular failure 

surfaces for the stability of simple, homogeneous, finite slopes without seepage (i.e., 

condition of effective stress). The general equations that Taylor developed as the basis for 

his stability charts are relationships between the height (H) and inclination (β) of the slope, 

the unit weight of the soil (γ), and the values of the soil’s developed (mobilized) shear 

strength parameters, cd and φd. These developed (mobilized) quantities are as follows: 

Where Fc is the average factor of safety with respect to cohesion and Fφ, is the average factor 

of safety with respect to friction angle, i.e., φd = arctan (tan φ′/ Fφ). As an approximation, the 

following equation may be used for the developed friction angle: 

However, for soils possessing both frictional and cohesive components of strength, the factor 

of safety in slope stability analyses generally refers to the overall factor of safety with respect 

to shearing strength, FS, which equals τ/τd where τ = shear strength and τd = the developed 

(mobilized) shear strength. Therefore, the general Mohr-Coulomb expression for developed 

shear strength in terms of combined factors of safety is: 

Equation 6-23 can be re-written in terms of developed shear strength parameters as follows: 

c ′ tan φ ′ c d = ; tan φ d = 6-21 F c Fφ

φ ′ φ d ≈ 6-22 Fφ

τ c ′ tan φ τ = = + σ ′ d 6-23 FS F c Fφ

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 25 December 2006 

τ d = c d + σ ′tan φ d 6-24 
There are an unlimited number of combinations of Fc and Fφ that can result in a given value 

of FS. However, Equations 6-21 and 6-23 suggest that for the case where the value of Fc = 

Fφ, the factor of safety with respect to shearing strength, FS, also equals that value. The 

importance of this condition will be illustrated in Section 6.6.2.1 by an example problem. 

To simplify the determination of the factor of safety, Taylor calculated the stability of a large 

number of slopes over a wide range of slope angles and developed friction angles, φd. He 

represented the results by a dimensionless number that he called the “Stability Number,” Ns, 

which he defined as follows: 

Equation 6-25 can be rearranged to provide an expression for Fc as a function of the Stability 

Number and three variables, c′, H and γ, as follows: 

Taylor published his results in the form of curves that give the relationship between Ns and 

slope angle, β, for various values of developed friction angles, φd, as shown in Figure 6-14. 

Note that factors of safety do not appear in the chart. The chart is divided into two zones, A 

and B. As shown in the inset for Zone A, the critical circle for steep slopes passes through 

the toe of the slope with the lowest point on the failure arc at the toe of the slope. As shown 

in the inset for Zone B, for shallower slopes the lowest point of the critical circle is not at the 

toe, and three cases must be considered as follows: 

•	 Case 2: For shallow slope angles or small developed friction angles the critical 

circle may pass below the toe of the slope. This condition corresponds to Case 2 in 

the inset for Zone B. The values of Ns for this case are given in the chart by the long 

dashed curves. 

c c ′ N d =	 s = 6-25

γH Fc γH 

c ′ F =	 c 6-26 N s γH

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 26 December 2006 
See Figure 6-15 

Slope Angle, β

Figure 6-14. Taylor's chart for soils with friction angle (after Taylor, 1948). 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 27 December 2006 
β = 53o 

For β > 53o, use Figure 6-14 

Figure 6-15. Taylor's chart for φ' =0 conditions for slope angles (β) less than 54° (after 

Taylor, 1948). 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 28 December 2006 
•	 Case 1: Where long dashed curves do not appear in the chart, the critical circle 

passes through the toe. This condition corresponds to Case 1. Stability numbers for 

Case 1 are given by the solid lines in the chart both when there is and when there is 

not a more dangerous circle that passes below the toe, i.e., the curves for Case 1 are 

an extension of the curves that correspond to a toe circle failure in Zone A. In both 

Case 1 and Case 2 the failure circle passes through the soil below the toe of the slope. 

The depth ratio, D, which is a multiple of the slope height H, is used to define the 

depth (DH) from the top of the slope to an underlying strong material through which 

the failure circle does not pass. 

•	 Case 3: This case corresponds to the condition where there is an underlying strong 

layer at the elevation of the toe (D=1). This case is represented by short dashed lines 

in the chart. 

Comment on φd = 0 condition: The condition of φd = 0 in Taylor’s Stability Chart is 

somewhat misleading since, as noted previously, Taylor’s charts were derived for simple 

slopes without seepage, i.e., for an effective stress analysis. The condition of “φd = 0” was 

used by Taylor to simplify the analysis and permit generation of the stability charts by 

assuming that shear strength is constant with depth. Basically, in the Mohr-Coulomb 

equation, Taylor assumed an average intergranular pressure, σ′avg instead of an actual value 

of σ′ which varies with depth. Since stability analyses are much simpler to perform when the 

shear strength is constant, he introduced this concept into his stability charts by considering 

the effective cohesion to be the average shear strength and by considering the friction angle 

to be zero. Thus the condition where φd = 0 is merely an example of substitution of an 

average value for a variable quantity. However, in the context of Taylor’s definition of φd = 

0, the stability charts are often used in practice for estimating the factor of safety and location 

of the critical circle in a homogeneous saturated clay in undrained shear. 

As shown in Figure 6-14, the critical circle for the “φd = 0” case passes below the toe for 

slopes with inclinations less than 53o. In practice the depth to which the failure circle extends 

is limited by an underlying strong material. Thus, the value of Ns for this case is greatly 

dependent on this limiting value of depth. The chart shown in Figure 6-15 is used 

exclusively for the “φd = 0” case and supplements the curves shown in Figure 6-14 for that 

condition. The coordinates in Figure 6-15 allow the chart to be used easily and enable the 

user to evaluate a number of parameters that may be of interest in practice. For example, the 

chart can be used to determine nH, which is the distance from the toe of the slope to where 

the critical failure surface passing below the toe may be expected to emerge. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 29 December 2006 
6.6.2.1 Determination of the Factor of Safety for a Slope 

As indicated in Section 6.6.2, in order to use Taylor’s charts to determine the minimum 

overall factor of safety with respect to shear strength for a slope of given height H and 

inclination β having soil properties γ, c′ and φ′, the condition FS = Fc = Fφ must be satisfied. 

The general computational approach is as follows: 

1.	 Assume a reasonable value for the common factors of safety FS = Fc = Fφ. 

2.	 Use Equations 6-21 to calculate the corresponding values of cd and φd. 

3.	 For the given value of β and the calculated value of φd, read the corresponding value 

of the stability number Ns from Figure 6-14. 

4.	 Use an inverted form of Equation 6-26 to calculate the slope height H corresponding 

to the assumed factor of safety. 

5.	 If the calculated value of H is within an acceptable distance of the actual height, e.g., 

± 0.5 feet, the assumed value of the common factor of safety represents the minimum 

overall factor of safety of the slope with respect to shear strength, Fs. 

6.	 If the calculated value of H is not within the desired acceptable range, the process is 

repeated with a new assumed value of the common factor of safety until the 

recomputed value of H falls within that range. 

7.	 The new assumed value of the common factor of safety for subsequent iterations is 

generally less than the previously chosen value if the calculated value of H is less 

than the actual value of H. Conversely, a. larger value of the new common factor of 

safety is assumed if the calculated value of H is greater than the actual value of H. 

The use of Taylor’s chart is illustrated by the following example. 

Example 6-1: Determine the factor of safety for a 30 ft high fill slope. The slope angle is 

30o. The fill is constructed with soil having the following properties: 

 Total unit weight, γ = 120 pcf; Effective cohesion, c′ = 500 psf 

 Effective friction angle, φ′ = 20o 

Solution: 

First assume a common factor of safety of 1.6 for both cohesion and friction angle so that Fc 

= Fφ = 1.6. Since Fφ = 1.6, the developed friction angle, φd, can be computed as follows: 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 30 December 2006 

⎛ tan φ ′ ⎞ ⎛ tan 20 o ⎞

φ	 = arctan ⎜ ⎟ = ⎜ ⎟ o

d arctan = 12.8 ⎜ F	 ⎟ ⎜ 1.6 ⎟ ⎝ φ ⎠ ⎝ ⎠
For φd = 12.8o, and β = 30o, the value of the stability number Ns from Figure 6-14 is 

approximately 0.065. Thus, from Equation 6-26 

or 

Since computed height H = 40.1 ft is greater than the actual height of 30-ft, the value of the 

common safety factor must be greater than 1.6. Assume Fc = Fφ = 1.9 and recompute as 

follows: 

If Fφ = 1.9, then φd = 10.8o and Ns from Figure 6-14 is approximately 0.073 based on which 

the recomputed value of H is as follows: 

The height of 30.04 ft is virtually identical to the correct height of 30 ft. Therefore, the 

minimum factor of safety with respect to shearing strength is approximately 1.9. 

Alternate Graphical Approach 

An alternate graphical approach for determining the minimum factor of safety with respect to 

shearing strength is also available. The procedure is as follows: 

1.	 Assume a reasonable value for Fφ and calculate φd. 

2.	 For the given value of β and the calculated value of φd read the corresponding value 

of the stability number Ns from Figure 6-14. 

3.	 Use Equation 6-26 to calculate Fc 

4.	 Repeat the process for at least two other assumed values of Fφ over a range of 

expected factors of safety so that at least three pairs of Fφ and Fc are obtained. 

5.	 Plot the calculated points on Fc versus Fφ coordinates and draw a curve through the 

points. 

6.	 Draw a line through the origin that represents Fc = Fφ. 

7.	 The minimum overall factor of safety of the slope with respect to shear strength is the 

value of factor of safety at the intersection of the calculated with the Fc = Fφ line. 

500 psf 0.065 = (1.6) (120 pcf) (H) 

500 psf H = = 40.1 ft 

(1.6) (120 pcf) (0.065) 

500 psf H = = 30.04 ft 

(1.9) (120 pcf) (0.073) 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 31 December 2006 
Example 6-1a: Solve Example 6-1 by using the alternate graphical approach. 

Solution: 

Set up a table for ease of computation (steps 1 through 4) as follows: 

Assumed Fφ Calculated φd Ns from Figure 6-14 Calculated Fc 

1.0 20 0.026 5.3 

1.5 14 0.055 2.5 

2.0 10 0.075 1.9 

2.5 8 0.087 1.6 

As shown in the figure below, plot the data as per steps 5 and 6 of the procedure. Read the 

value of Fc = Fφ = 1.95 at the intersection of plotted curves. This value is the minimum 

factor of safety with respect to shearing strength, FS. This value is close to the value of 1.9 

calculated in Example 6-1. This example problem is also solved by the use of a computer 

program, ReSSA, in Appendix D and the computer analysis yielded a FS =1.96 which is 

close to the values computed by the use of stability chart. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 32 December 2006 

6 

Fc 

5 

4 

3 

2 

1 

0 

Fc = Fφ

0 1 2 3 4 5 6 

Fφ 
6.6.3 Janbu’s Stability Charts 

Janbu (1968) published stability charts for slopes in soils with uniform strength for φ = 0 and 

φ > 0 conditions. These charts are presented in Figures 6-16 through 6-19. This series of 

charts accounts for several different conditions and provides factors for surcharge loading at 

the top of the slope, submergence, and tension cracks that can be expected to influence the 

design of typical highway slopes. 

The stability chart for slopes in soils with uniform shear strength throughout the depth of the 

layer and with φ = 0 is shown in Figure 6-16. Charts for correction factors for the conditions 

when surcharge loads, submergence and tension cracks are present are shown in Figures 6-17 

through 6-19. Step-by-step guidance for the use Janbu’s charts follows. 

Steps for using Janbu’s Charts on Figures 6-16 through 6-19, for φ = 0 material. 

Step 1. Use the chart at the bottom of Figure 6-16 to determine the position of the center 

of the critical circle, which is located at a coordinate point defined by Xo, Yo with 

respect to a cartesian coordinate system whose origin is at the toe of the slope. 

Following are some guidelines that can be used to identify the critical center: 

o	 For slopes steeper than 53°, the critical circle passes through the toe. For 

slopes flatter than 53°, the critical circle passes below the toe. 

o	 In addition to the toe circle, at least four circles with different depths below 

the toe, D, should be analyzed to ensure that the actual minimum factor of 

safety and the actual critical circle have been found. The following 

suggestions may be used to select the circles (Duncan and Wright, 2005): 

 If there is water outside the slope, a circle passing above the water may 

be critical. 

 If a soil layer is weaker than the one above it, the critical circle may 

extend into the lower (weaker) layer. This applies to layers both above 

and below the toe. 

 If a soil layer is stronger than the one above it, the critical circle may 

be tangent to the top of the stronger layer. 

 For each of the assumed circles, perform Steps 2 to 6. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 33 December 2006 
Figure 6-16. Stability charts for φ = 0 soils (Janbu, 1968). 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 34 December 2006 
Step 2. Using the assumed critical circle as a guide, estimate the average value of 

strength, c, by calculating the weighted average of the strengths along the failure 

surface. The number of degrees intersected along the arc by each soil layer as a 

percentage of the entire angle subtended by the arc is used as the weighting factor. 

Step 3. Calculate the depth factor, d where d = D/H. (Note that the depth factor, d, for 

Janbu’s charts is different from the depth ratio D for Taylor’s chart.) 

Step 4. Calculate Pd by using the following equation: 

Pd = (γH + q - γwHw)/( μt μqμw) 6-27

 where: q = surcharge load 

 γw = unit weight of water 

 Hw = depth of water outside the slope 

 μt = tension crack correction factor (Figure 6-17) 

 μq = surcharge correction factor (Figure 6-18, top) 

 μw = submergence correction factor (Figure 6-18, bottom) 

Step 5. Use the chart at the top of Figure 6-16 to determine the value of the stability 

number, No, which depends on the slope angle β, and the value of d. 

Step 6. Calculate the factor of safety (FS) by using the following equation: 

FS = Noc/Pd 6-28

Step 7. Repeat Steps 2 to 6 for all the circles assumed in Step 1. Compare the FS to 

obtain the most critical circle as the circle with the lowest FS. If it appears that 

the minimum FS is for a circle close to the toe, i.e., d=0, then it is prudent to 

check if the critical failure surface is within the height of the slope, H. In this 

case, the toe of the slope is adjusted to the point of intersection of assumed circle 

with the slope and all dimensions, (i.e., D, H, and Hw) are adjusted accordingly in 

the calculations and steps 1 to 6 are repeated (Duncan and Wright, 2005). 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 35 December 2006 
Crack filled with water, i.e., full 

hydrostatic pressure in crack 

No water in crack i.e., no 

hydrostatic pressure in crack 

Figure 6-17. Reduction factors to account for tension cracks to be used with stability 

charts for φ=0 and φ > 0 soils (Janbu, 1968). 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 36 December 2006 
Figure 6-18. Reduction factors to account for surcharge (upper) and submergence 

and/or seepage (lower) to be used with stability charts for φ=0 and φ > 0 soils (Janbu, 

1968). 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 37 December 2006 
 

 

 

 

 

 

 

 

 

 

Steps for using Janbu’s Charts on Figures 6-17 through 6-19, for φ > 0 materials. 

Step 1.	 Use judgment to estimate the location of the critical circle. For most conditions 

of simple slopes in uniform soils with φ > 0, the critical circle passes through the 

toe of the slope. The critical stability numbers given in Figure 6-19 were 

developed from analyses of toe circles. 

Where conditions are not uniform and there is a weak layer beneath the toe of the 

slope, a circle passing beneath the toe may be more critical than a toe failure. 

Figure 6-19 may be used to calculate the factor of safety for such cases provided 

the values of c and φ used in the analysis represent the correct average values for 

the circle considered. 

If there is a weak layer above the toe of the slope, a circle passing above the toe of 

the slope may be more critical. Similarly, if there is water outside the toe of the 

slope, a circle passing above the water may be more critical. When these types of 

circles are analyzed, the value of H should be equal to the height from the base of 

the weak layer, or the water level, to the top of the slope. 

Step 2.	 Use the estimated circle in Step 1 as a guide to estimate the average values of c 

and φ. This can be done by calculating the weighted average values of c and φ. 

The number of degrees intersected along the arc by each soil layer as a percentage 

of the entire angle subtended by the arc is used as the weighting factor for each 

parameter. 

Step 3.	 Calculate Pd by using Equation 6-27. 

Step 4.	 Calculate Pe by using the following equation: 

Pe = (γH + q - γwH′w)/(μqμ′w) 6-29

 where: H′w = height of water within the slope (Figure 6-18, bottom) 

μq = surcharge correction factor (Figure 6-18, top) 

μ′w = seepage correction factor (Figure 6-18, bottom) 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 38 December 2006 
Figure 6-19. Stability charts for φ > 0 (Janbu, 1968). 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 39 December 2006 
 

 

 

 

 

Step 5.	 Calculate the dimensionless parameter λCφ by using the following equation: 

λCφ = Pe tanφ/c 6-30 

For c=0, λCφ is infinite therefore skip to Step 6. 

Step 6.	 Use the chart in Figure 6-19 to determine the value of the critical stability 

number, Ncf, which is dependent on the slope angle, β, and the value of λCφ. 

Step 7. Calculate the factor of safety for the slope as follows: 

 

 

For c > 0 FS = Ncf c/ Pd 6-31 

For c = 0 FS = Pe b tan φ/ Pd 6-32 

Step 8.	 Determine the actual location of the critical circle by using the chart on the right 

side of Figure 6-19. The center of the circle is located at a coordinate point 

defined by Xo, Yo with respect to a cartesian coordinate system whose origin is at 

the toe of the slope. The circle passes through the toe of the slope (the origin), 

except for slopes flatter than 53°, where the critical circle passes tangent to the top 

of firm soil or rock. If the critical circle is much different from the one assumed 

in Step 1 for the purpose of determining the average strength, Steps 2 through 8 

should be repeated. 

If a slope contains more than one soil layer, it may be necessary to calculate the factor of 

safety for circles at more than one depth. If the underlying soil layer is weaker than the layer 

above it, the critical circle will extend into the lower layer, and either a toe circle or a deep 

circle within this layer will be critical. If the underlying soil layer is stronger than the layer 

above it, the critical circle may or may not extend into the lower layer, depending on the 

relative strengths of the two layers. Both possibilities should be examined (Duncan and 

Wright, 2005). 

The use of Janbu’s charts is illustrated by the following example. 

Example 6-2:	 Figure 6-20 shows a 35 ft high slope with a grade of 1.5H:1V. The soil 

properties within the slope and under it are shown on the figure. 

Groundwater is immediately under the slope. Calculate the factor of safety 

for a toe circle by using total stress analysis based on the soil properties 

shown. (Note: The circle in Figure 6-20 is plotted in Step 5 of the solution.) 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 40 December 2006 
 

 

 

 

 

 

35 ft 

25 ft 

δ2 = 29o δ1 = 54o 

γt = 120 pcf 

φu = 20o 

cu = 500 psf 

γt = 120 pcf 

φu = 0o 

cu = 1,000 psf 

Figure 6-20. Data for Example 6-2. 

Solution: 

The correction factors μt, μq, μw and μ′w are all equal to 1.0 since there is no tension crack (Ht 

= 0), no surcharge on the slope (q = 0), no water above the toe of the slope (γwHw= 0), and no 

seepage out of the slope (γwH′w= 0). 

1.	 Calculate Pd by using Equation 6-27 as follows: 

Pd = (γH + q - γwHw)/(μt μqμw) 

Pd = (120 pcf)(35 ft)/[(1)(1)(1)] = 4,200 psf 

2.	 Calculate Pe by using Equation 6-29 as follows: 

Pe = (γH + q - γwH′w)/(μqμ′w) 

Pe = (120 pcf)(35 ft)/[(1)(1)] = 4,200 psf 

3.	 For a toe circle, it is likely that a segment of the circle will pass through the soil 

below the toe and the average shear strength parameters along the circle will be 

different than those for the two layers. However, since at this stage the length of the 

segment passing through the soil below the toe is unknown, assume that the shear 

strength values of the soil within the slope height are representative and calculate the 

parameter λCφ by using Equation 6-30 as follows: 

λCφ = Pe tanφ/c = (4,200 psf) (tan 20o)/(500 psf) = 3.06 

4.	 From Figure 6-19 obtain the approximate center coordinates of the critical circle by 

using b=1.5 and λCφ = 3.06 as follows 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 41 December 2006 
xo ≈ 0.4 yo ≈ 1.6 Thus, Xo = (H)(xo) = (35 ft) (0.4) = 14 ft 

Yo = (H)(yo) = (35 ft) (1.6) = 56 ft 

5.	 Plot the critical circle on the given slope, as shown in Figure 6-20. Note that the 

subtended angles for the failure circle within the slope and the foundation are δ1= 54o 

and δ2 = 29 degrees, respectively. 

6.	 Calculate cav, tan φav and λCφ based on the angular distribution of the failure surface 

within the slope and foundation soil using δ1 and δ2 as follows: 

cav = [(54o) (500 psf) + (29o) (1,000 psf)] / (54o+29o) = 674.7 psf 

tan φav = [(54o) (tan 20o) + (29o) (tan 0o)] / (54o+29o) = 0.236 (or φav = 13.3o) 

Thus, according to Equation 6-30; 

λCφ = Pe tanφ/c = (4,200 psf) (0.236)/(674.7 psf) = 1.47 ≈ 1.5 

7.	 From Figure 6-19 obtain the center coordinates of the critical circle by using b=1.5 

and λCφ = 1.5 as follows 

xo ≈ 0.5 yo ≈ 1.55 Thus, Xo = (H)(xo) = (35 ft) (0.5) = 17.5 ft 

Yo = (H)(yo) = (35 ft) (1.55) = 54.3 ft 

This circle is close to the circle obtained in the previous iteration, so retain λCφ = 1.5 

and cav = 674.7 psf . 

8.	 From Figure 6-19, obtain Ncf = 10.0 for b = 1.5 and λCφ = 1.5. 

9.	 Calculate the factor of safety, FS, by using Equation 6-28 as follows : 

FS = Noc/Pd = (10) (674.7 psf) / (4,200 psf) 

FS = 1.61 

This calculation sequence is only for a given circle. This sequence is repeated for 

several circles and the resulting FS compared to find the minimum FS. With the 

advent of computer programs, this method is now more often used to verify the 

results of the computer generated most critical circle rather than computing the 

minimum FS by repeating the above sequence of calculations for several circles. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 42 December 2006 
6.7 SLIDING BLOCK FAILURE 

A "sliding block" type failure can occur where: 

1.	 the foundation soil contains thin seams of weak clay or organic soils, 

2.	 a shallow layer of weak soil exists at the ground surface and is underlain by firm soil, 

and 

3.	 the foundation soil contains thin sand or silt lenses sandwiched between less 

permeable soil. The weak layers or lenses provide a plane of weakness along which 

sliding can occur. In the case of sand or silt lenses trapped between less permeable 

soils, the mechanism that can cause sliding is as follows. As the fill load is placed, 

the water pressure is increased in the sand or silt lense. Since the water cannot escape 

due to the impermeable soil above and below, the sand or silt loses frictional strength 

as a result of the intergranular effective stress between soil grains being decreased 

due to the excess pore water pressure. 

Typical "sliding block" type failures are illustrated in Figure 6-21. When sliding occurs, an 

active wedge type failure occurs through the fill and a passive wedge type failure occurs 

below the fill toe as soil in the toe area is pushed out of the way. The sliding mass moves 

essentially as a block, thus the term "sliding block." These concepts are illustrated in Figure 

6-22. 

Figure 6-21. Sliding block failure mechanism. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 43 December 2006 
 

 

 

 

6.7.1 Sliding Block – Hand Method of Analysis 

A simple sliding block analysis to estimate the factor of safety against sliding is 

straightforward and can be performed easily and quickly by hand. For the analysis, the 

potential sliding block is divided into three parts; (1) an active wedge at the head of the slide, 

(2) a central block, and (3) a passive wedge at the toe as shown in Figure 6-22. 

Figure 6-22. Geometry and force components for sliding block analysis. 

Passive 

Wedge Active Wedge Central Block 

W 

Soft 

12′′ Clay Seam 

Pa 

cL 

Pp 

Fill 

Sand 

Sand L 

For the problem illustrated in Figure 6-22, the factor of safety would be computed by 

summing forces horizontally, to give: 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 44 December 2006 

Horizontal Resisting Forces Pp + cL 

FS = = Horizontal Driving Forces Pa 

6-33 

where: Pa = Active force (driving) 

Pp = Passive force (resisting) 

cL = Resisting force due to cohesion of clay 

The assumption is made that the loading is rapid so that there is no frictional component of 

resistance. For convenience of computation of a 1 ft thick slice of embankment is assumed. 

Several trial locations of the active and passive wedges must be checked to determine the 

minimum factor of safety. Note that since wedge type failures occur at the head and toe of 

the slope, similar to what occurs behind retaining walls, the active and passive forces are 
 

 

 

 

 

 

 

 

assumed to act against vertical planes that are treated as "imaginary" retaining walls, and the 

active and passive forces are computed the same as for retaining wall problems. 

6.7.1.1 Computation of Forces - Simple Sliding Block Analysis 

For the simple sliding block problem illustrated Figure 6-22 the forces used to compute the 

factor of safety can be calculated by using the Rankine approach as follows: 

Driving Force – Rankine Active Force 

Where: Pa = active force (kips) (kN) 

γ = unit weight of soil in the active wedge (kcf) (kN/m3

) 

H = height of soil layer in active wedge (ft) (m) 

Ka = active earth pressure coefficient for level ground surface 

Ka = (1-sinφ)/(1+sinφ) = tan2

 (45o - φ/2) (see Chapter 2) 

φ = angle of internal friction of soil in the active wedge. 

Resisting Force – Rankine Passive Force 

Where: Pp = passive force (kips) (kN) 

γ = unit weight of soil in the passive wedge (kcf) (kN/m3

) 

H = height of soil layer in passive wedge (ft) (m) 

Kp = passive earth pressure coefficient for level ground surface 

Kp = (1+sinφ)/(1-sinφ) = tan2

 (45o + φ/2) (see Chapter 2) 

φ = angle of internal friction of soil in the passive wedge. 

Resisting Force (kips or kN) = Clay cohesion (c in ksf or kPa) x 

Length of central wedge (L in ft or m) 

Computation Tips: 

The following design tips should be kept in mind when a sliding block analysis is performed. 

1.	 Be aware that the active or passive wedge can pass through more than one soil type 

with different strengths or unit weights. If that is the case then the active or passive 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 45 December 2006 

 

Pa = 1/2 γ H2

 Ka 6-34

 

Pp = 1/2 γ H2

 Kp 6-35
pressure distribution changes at the boundary between the different soils. This abrupt 

change in pressure is due to a change in either the angle of internal friction that 

affects the value of the earth pressure coefficient Ka or Kp and/or a change in the unit 

weight of the soil. The easiest way to handle this condition is to compute the active 

or passive earth pressure distribution diagram for each soil. There may be a 

discontinuity in the pressure diagram at the boundary between the two different soil 

layers. Then compute the active or passive force for each segment of the pressure 

distribution diagram from the area of each segment. 

2.	 When the active or passive pressure is being computed for soils below the ground 

water table, the buoyant (effective) unit weight of the soil must be used. 

The step-by step procedure for the Sliding Block Method of Analysis is illustrated by the 

following numerical example. 

Example 6-3: Find the safety factor for the 20 ft high embankment illustrated in Figure 6-23 

by using the simple sliding block method and Rankine earth pressure 

coefficients. Consider a 1 ft wide strip of the embankment into the plane of 

the paper. 

 

1′

10′

Soft Clay Layer c = 400 psf 

20′

2 

1 

γt = 110 pcf 

φ = 30°

γt = 110 pcf 

φ = 30°

Firm Material 

Figure 6-23. Example simple sliding block method using Rankine pressure coefficients. 

Solution 

Step 1: Compute driving force, Pa, by using Equation 6-34 

•	 Active Driving Force (Pa) by using Equation 6-34 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 46 December 2006 
1 Pa

	 = γt H2 K 

2 a (use γt as the water table is below the failure plane) 
 

⎛ ⎞ ⎛	 φ⎞ 30 o

K	 a	 = tan 2 ⎜45 − = tan 2 ⎜	 45 − ⎟ ⎟ = 0.33 ⎜ ⎟ ⎝ 2 ⎠ 2 ⎝	 ⎠

Step 2: Compute resisting forces 

•	 Central Block Resistance 

 kips cl = (0.400 ksf)(40ft)(1 ft) = 16.0 (71.1 kN) 

•	 Passive Resisting Force by using Equation 6-35 

Step 3: Compute factor of safety by using Equation 6-33 

6.7.2 Computation of Forces - Complicated Sliding Block Analysis 

The Rankine approach is a useful tool to portray the mechanism of a planar failure condition. 

However a general force diagram applicable to a more difficult sliding block type problem 

can account for the effects of water pressure, cohesion, friction, and a sloping failure plane in 

1	 Pa	 = (0.110 kcf )(30ft) 2

(0.33)(1ft) =16.5 kips 2 

1 Pp = γ H 2

K

2
 t p

⎛	 ⎞ ⎛	 φ⎞ tan2 30o 

K = 2 	 45 ⎜45 + ⎟

p ⎜ + ⎟ = tan = 3.0 ⎜ ⎝ 2 ⎠ 2 ⎟ ⎝ ⎠
 

⎛ 1 ⎞ Pp = ⎜ ⎟(0.110 kcf )(10ft) 2 (3)(1ft) =16.5 kips ⎝ 2 ⎠

cl + P 16.0kips +16.5kips Safety Factor = p =	 = 1.97 

Pa 16.5kips 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 47 December 2006 
the analysis. This analysis procedure, which is described in FHWA (2001a), can be used 

both to estimate the factor of safety for assumed failure surfaces in design or to "backanalyze" sliding block landslide problems. 

Computer solutions are also available for failure modes defined by planar and non-circular 

surfaces. However most of those solutions do not use the simplified Rankine block approach 

but rather a more complex failure plane such as that used in Janbu’s method. In general a 

computer solution is preferred for these planar failure problems because of the flexibility 

they offer in handling a variety of conditions that result in a more complex failure plane. 

6.8 SLOPE STABILITY ANALYSIS USING COMPUTER PROGRAMS 

Slope stability procedures are well suited to computer analysis due to the interactive nature of 

the solution. Also, the simplified hand solution procedures do not properly account for 

interslice forces, irregular failure surfaces, seismic forces, and external loads such as line 

load surcharges or tieback forces. Several user-friendly computer programs exist to analyze 

two-dimensional slope stability problems. One of the advantages of a computer program is 

that it allows parametric studies to be performed by varying parameters of interest, e.g., shear 

strength parameters. More complex computer programs are available for three dimensional 

slope stability analysis. As a minimum, a basic two-dimensional slope stability program is 

recommended for routine use. 

Desirable geotechnical features of such a program should include: 

•	 Multiple analysis capability 

 a. Circular arc (Bishop) 

 b. Non-circular (Janbu) 

 c. Sliding block 

•	 Variable input parameters to account for specific conditions 

 a. Heterogeneous soil systems 

 b. Pseudo-static seismic loads 

 c. Ground anchor forces 

 d Piezometric levels 

•	 Random generation of multiple failure surfaces with an option to analyze a specific 

failure surface. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 48 December 2006 
Desirable software features include: 

•	 User-friendly input screens including a summary screen that shows the cross section 

and soil boundaries in profile. 

•	 Help screens and error tracking messages. 

•	 Expanded output options for both resisting forces in friction, cohesion or tieback 

computations and driving forces in static or dynamic computations. 

•	 Ordered output and plotting capability for the failure surface of 10 minimum safety 

factors. 

•	 Documentation of program. 

A major problem for software users is technical support, maintenance and update of 

programs. Slope stability programs are in a continual process of improvement that can be 

expected to continue indefinitely. Highway agencies should implement only software that is 

documented and verified and for which the seller agrees to provide full technical support, 

maintenance and update. The following web page for the FHWA National Geotechnical 

Team contains links to distributors of FHWA software: 

http://www.fhwa.dot.gov/engineering/geotech/index.cfm 

Similar services are provided for commercially available slope stability programs such as the 

ReSSA (2001), SLOPE/W, SLIDE, STABL series (e.g., PCSTABL, XSTABL, GSTABL), 

and UTEXAS. Appendix D provides an overview of use of the ReSSA program. 

Finally, it is extremely important for the designer to understand that the design is only as 

good as the input parameters. Therefore, the designer should put major emphasis where it 

belongs, which is on: 

• Investigation 

• Sampling 

• Testing 

• Development of soil profile 

• Design soil strengths 

• Ground water table location 

Computer programs are only tools that aid in the design. The answers are only as good as the 

input data. Don't get carried away with plugging in the numbers and examining the results. 

You may learn the "garbage in - garbage out" principle the hard way. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 49 December 2006 
6.9 IMPROVING THE STABILITY OF EMBANKMENTS 

There are usually several technically feasible solutions to a stability problem. The chosen 

solution should be the most economical considering the following factors: 

1. Available materials. 

2. Quantity and cost of materials. 

3. Construction time schedules. 

4. Line and grade requirements. 

5. Right-of-way issues. 

6.9.1 Embankment Stability Design Solutions 

Table 6-2 presents a summary of practical solutions to mitigate embankment stability 

problems. Figures 6-24 to 6-26 illustrate some of the mitigation methods listed in Table 6-2. 

One of the solutions listed in Table 6-2 is the use of ground improvement. This solution can 

be used for cases where the internal stability of the embankment is not an issue due to the use 

of competent embankment materials, but the foundation materials are weak enough to affect 

the stability of the embankment slope. By improving the ground under the embankment, the 

resistance along the failure surface within the foundation is improved, thereby increasing the 

safety factor against slope failure. Relatively poor soils can be reinforced with geosynthetics 

to offset their low shear strength so that acceptable embankments can be constructed. 

Another solution is related to reinforcement of the embankment soils themselves. This 

solution can be used where the foundation is adequate but the locally available soils may not 

be suitable for construction of embankments at the desired slope angles. In this case, the 

embankment soils may be strengthened by the inclusion of reinforcements. Such slopes are 

called reinforced soil slopes (RSS). The RSS technology can be used to construct slopes at 

angles up to 69-degrees from horizontal. The RSS design method is discussed here as an 

example of a remediation method. Only the basics of the RSS design method are presented 

herein. Detailed design procedures for RSS technology can be found in FHWA (2001b). 

6.9.2 Design Approach for Reinforced Soil Slopes 

The design of internal reinforcement for safe, steep slopes requires a rigorous analysis. The 

design of the reinforcement for this application is critical, as failure of the reinforcement 

would result in failure of the slope. The overall design requirements for reinforced slopes are 

similar to those for unreinforced slopes. The factor of safety must be adequate for both the 

short-term and long-term conditions and for all possible modes of failure. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 50 December 2006 
 

 

 

 

 

Table 6-2 

Practical design solutions to mitigate embankment stability problems
 

*1. Relocate highway

 alignment. 

A line shift of the highway to an area having better soils may be the most 

economical solution. 

*2. Reduce grade line. 

 (flatten slope) 

A reduction in grade line will decrease the weight of the embankment and 

will improve stability (Figure 6-24). 

3. Counterweight 

berms. 

A counterweight berm outside of the center of rotation, as illustrated in 

(Figure 6-25), provides an additional resisting moment that increases the 

factor of safety. Berms should be built concurrently with the embankment. 

The embankment should never be completed prior to berm construction since 

the critical time for shear failure is at the end of embankment construction. 

The top surface of a berm should be sloped to drain water away from the 

embankment. Also, care should be exercised in selection of materials and 

compaction specifications to assure the design unit weight will be achieved 

for berm construction. 

4. Excavation of soft 

soil and replacement 

with shear key. 

The strength of soft soils is often insufficient to support embankments. In 

such cases, the soft soils are excavated and replaced with granular material 

that acts like a shear key (Figure 6-26). 

5. Displacement of For deep soft deposits, excavation is difficult. The soft soil can be displaced 

 soft soil. by generating continuous shear failures along the advancing fill front until 

the embankment is on firm bottom. The mudwave forced up in front of the 

fill must be excavated to insure continuous displacement and prevent large 

pockets of soft soil from being trapped under the fill 

6. Slow rate or stage Many weak subsoils will tend to gain strength during the loading process as 

consolidation occurs and pore water pressures dissipate. For soils that 

consolidate relatively fast, such as some silts and silty clays, this method is 

practical. Proper instrumentation is desirable to monitor the state of stress in 

the soil during the loading period to insure that loading does not proceed so 

rapidly that a shear failure occurs. Typical instrumentation consists of slope 

inclinometers to monitor stability, piezometers to measure excess pore water 

pressure, and settlement devices to measure the amount and rate of 

settlement. Planning of the instrumentation program and data interpretation 

should be done by a qualified and experienced geotechnical engineer. This 

option could also be used if weak subsoils are pretreated with wick drains 

7. Lightweight In some areas of the country lightweight materials such as blast furnace slag, 

 embankment. shredded rubber tires, or expanded shale are available. The slag material 

weighs about 80 pcf (12.6 kN/m3

). Sawdust fill weighs about 50 pcf (7.9 

kN/m3

) and has a friction angle of 35o or more. Expanded Polystyrene Foam 

(EPS) is available throughout the country and weighs 1 to 3 pcf (0.15 to 0.5 

kN/m3

). Use of such materials decreases the driving force. Typical 

advantages and disadvantages of the use of such materials, and specifications 

for lightweight fills are included in FHWA (2006b). 

8. Ground improvement Recently developed techniques such as stone columns, soil mixing, 

geosynthetics, soil nailing, ground anchors, and grouting can be used to 

increase resisting forces. Specialty contractors should be considered for 

these design solutions. 

9. Reinforcement of 

embankment soils. 

The embankment soils can be strengthened by incorporating reinforcements 

with the compacted soil. The reinforcement generally permits steeper slopes 

compared to unreinforced embankments. 

*Always consider these solutions first since they are relatively simple and inexpensive. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 51 December 2006 
Figure 6-24. Reduction of grade line to improve slope stability. 

Figure 6-25. Use of counterweight berm to improve slope stability. 

Figure 6-26. Use of shear key to improve slope stability. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 52 December 2006 
As illustrated in Figure 6-27, there are three possible failure modes for reinforced slopes: 

1.	 Internal - the failure plane passes through the reinforcing elements. 

2.	 External - the failure surface passes behind and underneath the reinforced mass. The 

reinforced mass is the mass of soil that contains the reinforcements. 

3.	 Compound - the failure surface passes behind and through the reinforced soil mass. 

Figure 6-27. Failure modes for Reinforced Soil Slopes. 

In some cases, the calculated minimum safety factor can be approximately equal in two or 

even all three modes if the reinforcement strengths, lengths, and vertical spacing are 

optimized (FHWA, 2001b). FHWA (2001b) contains a detailed discussion of the analysis 

and design of RSS’. A convenient chart solution is presented in this manual for preliminary 

feasibility-level design of the RSS. 

6.9.2.1 Preliminary Feasibility Design of RSS 

A preliminary design for a feasibility evaluation can be easily made by the use of design 

charts. These charts can also be used for the final design of low slopes, i.e., slope height less 

than 20 ft (6 m), where the consequences of failure are not critical. Figure 6-28 is a widely 

used chart that presents a simplified method based on a two-part, wedge-type failure surface. 

Use of the chart is limited by the assumptions noted on the figure. Figure 6-28 is not 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 53 December 2006 
intended to be a single design tool. Other design charts available from the literature could 

also be used, e.g., FHWA (2001b), Leshchinsky and Perry (1987). 

The procedure for using the charts shown in Figure 6-28 is as follows: 

1.	 For an assumed (desired) safety factor, F, determine the factored friction angle, φ′f, in 

degrees as follows (Note: this is similar to the factored friction angle in Taylor’s 

stability chart): 

2. Using φ′f read the force coefficient K from Part A and determine TS-MAX as follows: 

⎛ tan φ′ ⎞ φ′

f = arctan	 ⎜ ⎟ ⎝ F ⎠

T = 0.5 K γ (H′)

2 

S-MAX f 

where H′ = H + q/γ is the effective height, q = surcharge, and γf = fill unit weight. 

3.	 Determine the length of the reinforcement at the top, LT, and bottom, LB, of the slope 

from Part B. 

4.	 Determine the distribution of reinforcement: 

•	 For low slope heights (H ≤ 20 ft) assume a uniform reinforcement distribution, 

and use TS-MAX to determine the spacing or the required tension, TMAX,, for each 

reinforcement layer. 

•	 For high slope heights (H > 20 ft), divide the slope into two or three 

reinforcement zones of equal height, and use a factored TS-MAX in each zone for 

spacing or design tension requirements. 

For 2 zones: For 3 zones:

 TBottom = 3/4 TS-MAX TBottom = 1/2 TS-MAX

 TTop = 1/4 TS-MAX TMiddle = 1/3 TS-MAX

 TTop = 1/6 TS-MAX 

The force is assumed to be uniformly distributed over the entire zone. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 54 December 2006 
Part A: Reinforcement 

Force Coefficient 

Part B: Reinfo

Ratios 

K 

Chart assumptions: (1) extensible reinforcement, (2) slopes constructed with uniform cohesionless soils (c=0), (3) no pore water pres

(4) competent, level foundation soils, (5) no seismic forces, (6) uniform surcharge, q, not greater than 0.2γfH, (7) 

soil/reinforcement interface friction angle = 0.9φ′ (may not be appropriate for some geotextiles). 

Figure 6-28. Chart solution for determining the reinforcement strength requirements (after Schmertm

FHWA NHI-06-088 Soils and Foundations – Volume I 6 - 55 
•	 Determine the requirements for vertical spacing of the reinforcement, Sv, or the 

maximum design tension, TMAX, for each reinforcement layer. 

•	 For each zone, calculate TMAX for each reinforcing layer in that zone based on an 

assumed Sv or, if the allowable reinforcement strength is known, calculate the 

minimum vertical spacing and number of reinforcing layers, N, required for each 

zone based on Equation 6-36 and the use of consistent units. 

where: 

Ta = sum of available tensile force per width of reinforcement for all 

reinforcement layers. 

Rc = coverage ratio of the reinforcement that equals the width of the 

reinforcement, b, divided by the horizontal spacing Sh. 

Sv = vertical spacing of reinforcement; multiples of compacted layer thickness 

for ease of construction. 

Tzone 	 = maximum reinforcement tension required for each zone. 

 = TS-MAX for low slopes (H< 20 ft) 

 Hzone 	 = height of zone. 

 = Ttop, Tmiddle, and TBottom for high slopes (H > 20 ft) 

N = number of reinforcement layers. 

•	 In general, use short (4 - 6.5 ft (1.2 – 2 m)) lengths of reinforcement layers to 

maintain a maximum vertical spacing of 16 in (400 mm) or less for face stability and 

compaction quality. This short reinforcement should be placed in continuous layers 

and need not be as strong as the primary load bearing reinforcement, but it must be 

strong enough to survive construction (e.g., minimum survivability requirements for 

geotextiles in road stabilization applications in AASHTO M-288) and provide 

localized tensile reinforcement to the surficial soils. 

For detailed analyses required for final design, refer to FHWA (2001b). The computer 

program ReSSA (2001) noted earlier, can perform analysis and design of reinforced soil 

slopes using the methods described in FHWA (2001b). 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 56 December 2006 

Tzone S v T T zone 	 R a c = T d = = 6-36 Hzone N 
6.10 IMPROVING THE STABILITY OF CUT SLOPES 

The two most common types of cut slope failures are deep-seated and shallow surface 

failures. Both of these types of failure and their mitigation are discussed in this section. 

6.10.1 Deep Seated Failure 

A deep seated failure usually occurs in slopes cut into clay. The clay has insufficient shear 

strength to support the slope, and shear failure generally occurs along a circular arc. If the 

clay contains water-bearing silt or sand layers, seepage forces will also contribute to the 

instability. Figure 6-29 shows an example of a deep seated failure and a possible design 

solution. Table 6-3 lists typical design solutions to potential cut slope stability problems in 

clay. 

Figure 6-29. Deep seated slope failure (left) and bench slope design (right) to prevent 

slope failure. 

Table 6-3 

Typical design solutions to mitigate cut slope stability problems
 

Design Solution Effect on Stability 

a. Flatten slope. Reduces driving force. 

b. Bench slope. Reduces driving force. 

c. Buttress toe. Increases resisting force. 

d. Lower water table. Reduces seepage force. 

e. Reinforcement (e.g., nails) Increases resisting force 

The design of cut slopes in clay should NOT be based on the undrained strength of the 

clay determined by tests on samples obtained before the cut is made. Designs based on 

undrained strength will be unconservative since the effective stress is reduced when the cut is 

made because load is removed. This decrease in effective stress allows the clay to swell and 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 57 December 2006 
 

 

lose strength if water is made available to the clay as shown in Figure 6-30. Therefore, the 

design of cut slopes in clays should be based on effective strength parameters so that the 

reduction in effective stress resulting from the excavation can be taken into account. It is 

important to remember that an undrained clay in a cut gradually weakens and may fail long 

after construction. 

Figure 6-30: Typical cut slope failure mechanism in clay soils. 

6.10.2 Shallow Surface Failures 

Shallow surface failures (sloughs) are most common in cut slopes in layered clay or silt. 

This type of failure may involve either an entire slope or local areas in the slope. The prime 

cause of shallow surface failures is water seepage. Water seepage reduces the strength of the 

surface soils, causing them to slide or flow. 

Sloughing of slopes due to ground water seepage can often be remedied by placing a 2-3 ft 

(0.6-1 m) thick rock or gravel blanket over the critical area. The blanket reduces the seepage 

forces, drains the water, and acts as a counter-weight on the unstable soil. The blanket 

should be "keyed" into the ditch at the toe of the slope. The key should extend about 4 feet 

(1.2 m) below the ditch line and be about 4 ft (1.2 m) wide. A geotextile should be placed 

both under the key and against the slope before placement of the gravel blanket. 

Construction of the gravel blanket should proceed from the toe upwards. The most effective 

placement is by a dozer that will track over and compact the lower areas of the gravel blanket 

while the upper areas are being constructed. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 58 December 2006 
6.10.3 Factor of Safety - Cut Slopes 

As indicated previously, a minimum design safety factor of 1.25 is used for routine highway 

embankment side slopes. A minimum factor of safety against sliding of 1.50 is recommended 

for the stability of cut slopes in fine-grained soils. The greater factor of safety for cut slopes 

is based upon the knowledge that cut slopes may deteriorate with time as a result of natural 

drainage conditions that embankments generally do not experience. In addition, there is a 

greater degree of uncertainty about the homogeneity of the soils in cut slopes than in 

embankment slopes that are engineered and constructed under controlled conditions. 

FHWA NHI-06-088 6 – Slope Stability 

Soils and Foundations – Volume I 6 - 59 December 2006 
[THIS PAGE INTENTIONALLY BLANK] 
CHAPTER 7.0 

APPROACH ROADWAY DEFORMATIONS 

Often roadways are constructed on embankment fills to meet the requirements of the vertical 

grade of a roadway alignment. Fills placed to accommodate the vertical profile as the 

roadway approaches a bridge are often referred to as “approach embankment fills” or 

“approach roadway fills.” Typical elements of a bridge approach system are shown in Figure 

7-1. The abutment configuration may vary as shown in Figure 7-2. An abutment fill slope is 

also referred to as an “end-slope.” The common element to all types of abutments is an 

approach fill. Deformation, both vertical and lateral, of approach fills is the most prevalent 

foundation problem in highway construction. The embankment deformation near a bridge 

structure, leads to the ubiquitous “bump at the end of the bridge.” Figure 7-3 shows some of 

the problems leading to the existence of the bump. 

Approach slabs are often used by most state agencies to provide a smooth transition between 

the bridge deck and the roadway pavement. The slab usually is designed to withstand some 

embankment settlement and a reduction in subgrade support near the abutment. Joints must 

be provided to accommodate cyclic thermal movements of the bridge deck, abutment and 

roadway pavement. Figure 7-1b shows one common joint set. However, if the approach 

embankments are not properly engineered, the approach slab merely moves the bump at the 

end of the bridge to the approach slab-roadway interface. Unlike stability problems, the 

results of approach embankment deformations are seldom catastrophic but the cost of 

perpetual maintenance of continuing deformation can be immense. The difficulty in 

preventing these problems is not so much a lack of technical knowledge as a lack of 

communication between personnel involved in the roadway design and those involved in the 

structural design and construction. 

7.1 TYPICAL APPROACH ROADWAY DEFORMATION PROBLEMS 

Roadway designers allow use of inexpensive available soils for approach fills to reduce 

project costs. The bridge structures are necessarily designed for little or no deformation to 

maintain specified highway clearances and to insure integrity of structural members. In most 

agencies the responsibility for approach embankment design is not defined as a structural 

issue, which results in roadway embankment requirements being used up to the structure. In 

reality, the approach embankment requires special materials and placement criteria to 

prevent internal deformations and to mitigate external deformations. A discussion of 

the types of deformation associated with approach fills follows. 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 1 December 2006
 
Traffic 

Lanes 

Construction joint 

Expansion joint 

Shoulder 

Wingwall 

(b) Bridge Approach slab Pavement 

(a) 

15 ft 

Pneumatic adjustable sleeper footing Wingwall Access Port 

Shoulder 

Note: This plan detail is only one way of handling the bridge/fill interface. 

An approach slab with expansion between the superstructure and the 

approach slab without a sleeper slab is another. 

Figure 7-1. (a) Elements of a bridge approach system, (b) Plan view of an approach 

system (modified after NCHRP, 1997). 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 2 December 2006
 
(“True” bridge abutment) 

Figure 7-2. Types of abutments (modified after NCHRP, 1990). 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 3 December 2006
 
15 ft 

• Incorrect bearings and 

associated movements 

Figure 7-3. Problems leading to the existence of a bump (modified after NCHRP, 1997

FHWA NHI-06-088 7 – ApproaSoils and Foundations – Volume I 7 - 4 
Most state agencies, as noted earlier, use bridge approach slabs to provide a transitional 

roadway between the pavement on the approach embankment and the actual structure of the 

bridge. Due to the deformation of the approach embankment fills for various reasons shown 

in Figure 7-3, these slabs can settle and/or rotate creating problems for the abutment as well 

as the joints. Depending on the configuration of the approach slab, e.g., how the slab is 

connected to the abutment and/or the wing walls, voids may develop under the slab as the 

approach fill settles. Such voids can then fill with water, which can further compound the 

problem, e.g., water pressures acting against structural elements, softening of the soils with 

associated reduction in strength, freeze-thaw issues, etc. Due to the above considerations, 

design problems with approach roadway embankments are classified as follows: 

• Internal deformation within the embankment 

• External deformation in native soils below the embankment 

As mentioned previously, it is important to realize that the deformation considerations for the 

embankment include both vertical as well as lateral deformations. Vertical deformations are 

commonly referred to as “settlements.” Lateral deformations can result in rotation of the 

structure that is commonly referred to as “tilting.” 

Internal deformation is a direct result of compression of the materials used in the 

construction of the embankment fill. The importance of adequate drainage with respect to 

the internal behavior of the embankment cannot be overemphasized. Poor drainage can (a) 

cause softening of the embankment soils leading to vertical and lateral deformations, (b) 

reduce the stability of soils near the slope leading to lateral deformations and associated 

vertical deformations near the crest of the slope, and (c) potentially lead to migration of fill 

material and creation of voids or substantial vertical and lateral deformations. 

External deformation is due to the vertical and lateral deformation of the foundation soils 

on which the embankment is placed. Furthermore, deformation of foundation soils may 

include both immediate and consolidation deformations depending on the type of foundation 

soils. Lateral squeeze of the foundation soils can occur if the soils are soft and if their 

thickness is less than the width of the end slope of the embankment. Consolidation 

settlement and lateral squeeze are not an issue within embankment fills since coarse-grained 

soils placed under controlled compaction conditions are generally used. 

This chapter discusses internal as well as external deformations of approach fills. Design 

solutions to mitigate the detrimental effect of these deformations are presented. Guidelines 

for construction monitoring are also provided. 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 5 December 2006
 
7.2 INTERNAL DEFORMATION WITHIN EMBANKMENTS 

Internal deformation within embankments can be easily controlled by using fill materials that 

have the ability to resist the anticipated loads imposed on them. A well constructed soil 

embankment will not excessively deform internally if quality control is exercised with regard 

to material and compaction. Standard specifications and construction drawings should be 

prepared for the approach embankment area, normally designated to extend 50 ft (15 m) 

behind the wingwall. The structural designer should have the responsibility for selecting the 

appropriate cross section for the approach embankment depending on selection of the 

foundation type. A typical approach embankment cross section is shown in Figure 7-4. 

Special attention should be given to the interface area between the structure and the approach 

embankment, as this is where the "bump at the end of the bridge" occurs. The reasons for the 

bump are (a) poor compaction of embankment material near the structure, (b) migration of 

fine soil into drainage material, and (c) loss of embankment material due to poor drainage 

details as discussed earlier. Poor compaction is usually caused by restricted access of 

standard compaction equipment. Proper compaction can be achieved by optimizing the soil 

gradation in the interface area to permit compaction to maximum density with minimum 

effort. Figure 7-5 shows a detail for placement of drainage material. Considerations for the 

specification of select structural backfill and underdrain filter material to minimize the 

“bump” problem are included in the next two sections. Similar drainage results can be 

obtained by the use of prefabricated geocomposite drains that are attached to the backwall 

and connected to an underdrain. 

7.2.1 General Considerations for Select Structural Backfill 

Select structural backfill is usually placed in relatively small quantities and in relatively 

confined areas. Structural backfill specifications must be designed to ensure construction of 

a durable, dense backfill. Table 7-1 lists considerations for the specification of select 

structural backfill. 

7.2.2 General Considerations for Drainage Aggregate 

The drainage aggregate, such as that used for underdrain filters, should consist of crushed 

stone, sand, gravel or screened gravel. Suggested gradation for drainage aggregate is 

provided in Table 7-2. The AASHTO standard gradation No. 57 or 67 should be equally 

suitable. 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 6 December 2006
 
15 m (50') min

1.5 m (5')

mm (6") 

Backfill abutment to a point 1.5 m

(5') behind the wingwall with

Material per Note 1 to subgrade

1.5 m (5 ) Pad of Select

Material placed beneath

abutments on spread

footings

Minimum breakpoint of berm

and end slope may be located

0.6 m (2 ) above the top of the

footing and 1.2 m (4') out from

the front edge

 

1 

1 

3 min 

1 

Place embankment to this line 

prior to abutment construction 

or pile driving 

W orking lines drawn perpendicular to projected 

' 

' 

See Note 1 

Projection of abutment wingwall 

5 ft (1.5 m) pad of select 

material placed beneath 

abutments on spread 

footings 

5 ft (1.5 m)

 50 ft (15 m) min 

Minimum breakpoint of berm 

and end slope may be located 

2 ft (0.6 m) above the top of 

the footing and 4 ft (1.2 m) out 

from the front edge 

centerline of bearing to determine end limit for Backfill abutment to a point 5 ft

placement of Select Material and Highway (1.5 m) behind the wingwall 

Embankment Material with material as per Note 1 to 

subgrade 

Select Structural Fill
 

(Minimum 100% compaction, T99)
 

Highway Embankment Material, 150 Max. Topsize 6 in (150 mm) 
 

(Minimum 95% compaction, T180) 

Highway Embankment Material 

(Minimum 90% compaction, T180) 

Note 1: Highway embankment material and select material shall be placed simultaneously of the 

vertical payment line 

Figure 7-4. Suggested approach embankment details. 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 7 December 2006
 
Heel Projection + 3’- 0” 

Heel Projection + 3’ – 0” 

(Table 7-2) 

(Table 7-1) 

Figure 7-5. Structural backfill placement limits for porous drainage aggregate. 

(1 ft = 0.3 m; 1 in = 25.4 mm) 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 8 December 2006
 
Table 7-1 

General considerations for specification of select structural backfill 

Consideration Comment 

Lift Thickness Limit to 6" to 8" (150 mm to 200 mm), so compaction is possible 

with small equipment. 

Topsize (largest 

particle size) Limit to less than 3⁄4 of lift thickness. 

Gradation/Percent 

Fines 

Use well graded soil for ease of compaction. Typical gradation is 

as follows: 

Sieve Size Percent Passing (by weight) 

4-in (100 mm) 100 

No. 40 (0.425 mm) 0 to 70 

No. 200 (0.075 mm) 0 to 15 

The limitation on percent fines (particles smaller than No. 200 

sieve) is to prevent piping and allow gravity drainage. For rapid 

drainage, consideration may be given to limiting the percent fines 

to 5%. 

Plasticity Index The plasticity index (PI) should not exceed 10 to control long-term 

deformation. 

Durability This consideration attempts to address breakdown of particles and 

resultant settlement. The material should be substantially free of 

shale or other soft, poor-durability particles. Where the agency 

elects to test for this requirement, a material with a magnesium 

sulfate soundness loss exceeding 30 should be rejected. 

T99 Density Control Small equipment cannot achieve AASHTO T180 densities. 

Minimum of 100 percent of standard Proctor maximum density is 

required. 

Compatibility Particles should not move into voids of adjacent fill or drain 

material 

Table 7-2 

Suggested gradation for drainage aggregate 

Sieve Size Percent Passing (by weight) 

1-in (25.4 mm) 100 

1⁄2-in (12.7 mm) 30 to 100 

No. 3 (6.3 mm) 0 to 30 

No. 10 (2.00 mm) 0 to 10 

No. 20 (0.85 mm) 0 to 5 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 9 December 2006
 
As with the select backfill, the soundness of the drainage aggregate should be tested. The 

drainage aggregate should have a loss not exceeding 20 percent by weight after four (4) 

cycles of the magnesium sulfate soundness test. 

The maximum loose lift thickness for the drainage aggregate should not exceed 6 in (150 

mm). Placement and compaction operations should be conducted in a manner so as to insure 

that the top surface of each lift of the drainage aggregate should not be contaminated by the 

adjacent backfill materials. Compaction of the drainage aggregate is commonly achieved by 

two passes of a vibratory compactor approved by the engineer. No compaction control tests 

are normally required for the drainage aggregate. 

7.2.3 Use of Geosynthetics to Control Internal Deformations 

In geographic areas where select materials are not available, the use of geosynthetic materials 

to reinforce the abutment backfill and approach area can reduce the bump at the end of the 

bridge. Such reinforced fills can be designed by using the principles of Reinforced Soil 

Slopes (RSS) discussed in Chapter 6 (Slope Stability) 

It is suspected that high dynamic loads are routinely induced in the abutment backfill due to 

vehicle impact loads. Poorly designed or constructed drainage layers or non-durable 

drainage aggregate can cause either piping of fines or accelerated pavement subsidence due 

to breakdown of aggregates. As indicated previously, the use of geotextiles or geocomposite 

drains can be an effective method of minimizing internal embankment deformation and the 

resulting “bump at the end of the bridge.” 

7.3 EXTERNAL DEFORMATION IN FOUNDATION SOILS BELOW 

EMBANKMENTS 

Once the issue of internal deformation within fills has been addressed, the designer must 

concentrate on the evaluation of the deformation of foundation soils and any engineered soils 

on which the fills will be placed. As explained in Chapter 2, deformations in foundation 

soils under embankments occur due to the pressure imposed by the embankments. 

Depending on the type of foundation soils, one or both of the following deformations may 

occur: 

• Immediate (elastic) deformation 

• Consolidation (or long-term) deformation 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 10 December 2006
 
Immediate or elastic deformations occur in all soils regardless of whether they are cohesive 

or cohesionless. Consolidation deformations typically occur in fine-grained soils that are 

saturated at the time additional loads are applied. Many and varied procedures exist for 

computation of these types of deformations. Two methods are presented in this chapter; one 

each for cohesionless and cohesive soils. However, there is a critical first step that is 

common to both modes of deformation. This first step involves the estimation of the stress 

distribution within the foundation soils due to the pressures imposed by the embankment 

fills. This step is discussed next. 

7.3.1 Procedure for Estimating Stress Distribution in Foundation Soils under Fills 

The basic steps involved in estimating stresses in native soils under fills are as follows: 

1. Develop a soil profile including soil unit weights, SPT results (N160), moisture contents 

and interpreted consolidation test values. 

2. Draw effective overburden pressure (po) diagram with depth. 

3. Plot total embankment pressure (pf) on the po diagram at ground surface level. 

4. Distribute the total embankment pressure with depth by using the appropriate pressure 

coefficient charts presented in Figure 7-6. 

(Note: The charts in Figure 7-6 are limited to only two locations, Section B-B and 

Section C-C, and assume that the end and side slopes have the same grades. Programs 

such as FoSSA (2003) may be used for case of unequal end and side slopes, or if 

pressure coefficients at locations other than along Section B-B or C-C are desired.) 

The principles to remember are: (1) stresses induced in the soil from an embankment load 

are distributed with depth in proportion to the embankment width, and (2) the additional 

stresses in the soil decrease with depth. 

Following is a step-by-step procedure to use the chart in Figure 7-6. A worked example is 

presented afterwards to illustrate the use of the chart numerically: 

Step 1. Determine the distance bf from the centerline of the approach embankment to the 

midpoint of side slope. Multiply the numerical value of "bf" by the appropriate 

values shown to the right of the chart to develop the depths at which the distributed 

pressures will be computed, e.g., 0.2bf, 0.4bf, etc. 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 11 December 2006
 
CHART LEGEND: 

In each chart, the upper line gives the 

pressure only under the embankment 

centerline B-B while the lower line gives 

the pressure only under the center of 

the side slope along Section C-C. 

A 

B 

C 

B 

C 

CHART ASSUMPTION 

Side slope = End slope 

bf 

Section B-B 

Point X 

Figure 7-6. Pressure coefficients beneath the end of a fill (after NYSDOT). 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 12 December 2006
 
 

 

Step 2. Select the point X on the approach embankment where the vertical stress prediction 

is desired, normally at the intersection of the centerline of the embankment and the 

abutment. In this case the side slope is called the end slope. Measure the distance 

from X to the midpoint of the end slope. Return to the chart and scale that distance 

as a multiple of bf on the horizontal axis from the appropriate side of the midpoint 

centerline line of the end slope. 

Step 3. Read vertically up or down from the plotted distance on the horizontal axes to the 

various curves corresponding to depth below surface. The "Kf" value on the left 

vertical axis should be read and recorded on a computation sheet with the 

corresponding depth. Note that the upper line gives the pressure under the 

embankment centerline (Section B-B) while the lower line gives the pressure under 

the mid of the side slope (Section C-C). 

Step 4. Multiply each "Kf" value by the value of total embankment pressure (γfhf) to 

determine the amount of the pressure increment (∆p) transmitted to each depth, 

where γf is the unit weight of the embankment fill soil and hf is height of the 

embankment fill. 

The application of this step-by-step procedure and the charts shown on Figure 7-6 to a 

typical embankment problem is illustrated by the following worked example problem. 

Example 7-1: The geometry of a fill slope is as follows: 

 Fill height hf = 30 ft; Fill unit weight γf = 100 pcf 

End and side slopes (2H:1V); Embankment top width = 100 ft 

 

1 

2 

hf 

bf 

Point X 

bf = 80′ 30′

60′ 100′

Find: The stress increase (∆p) under the proposed abutment centroid (Point X) at a depth 

of 0.8 bf below the base of the fill. 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 13 December 2006
 
 

Solution: 

Figure 7-6 will be used to determine the stress increase. To use the chart first compute the 

following quantities: 

• Distance from midpoint of end slope to Point X = 30 ft. 

• Distance from centerline to mid point of side slope bf = (100 ft/2) + (60 ft/2) = 80 ft. 

Enter stress distribution chart for a depth of 0.8bf = (0.8)(80 ft) = 64 ft and a distance 

measured from the midpoint of the end slope to Point X expressed as a multiple of bf = (30 

ft/80 ft) bf = 0.38 bf. Enter the plot with this value to the left of the value of zero on the 

abscissa, i.e., upslope from the midpoint on the end slope. 

In Figure 7-6 read Kf = 0.7 from the chart for 0.8bf. Therefore, at a depth of 64 ft below the 

embankment at Point X 

∆p = Kf γf hf 

∆p = (0.7) (100 pcf) (30 ft) = 2,100 psf 

Repeat the above steps for distances to other points along the centerline of the embankment 

expressed as a multiple of bf and measured (+ and -) from the midpoint of end slope to 

develop the horizontal distribution of vertical stress increases due to the embankment at a 

depth of 64 ft below and beyond the base of the end slope along the embankment centerline. 

Horizontal Distribution of Vertical Stress Increases Below and Beyond the End Slope at 

a Depth of 64-ft Below the Embankment 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 14 December 2006
 Pressure Coefficient Kf 0.5 

0 

0.38bf Mid Point of 

1.0 

0.7 

1.0bf 0.5bf End Slope 0.5bf 1.0bf 

Centerline 

0.8bf = 64 ft

Depth Below Surface

Mid Point of 

Side Slope 
7.4 COMPUTATION OF IMMEDIATE SETTELEMENT 

All geomaterials, whether cohesionless or cohesive, will experience settlements immediately 

after application of loads. Whether or not the settlements will continue with time after the 

application of the loads will be a function of how quickly the water can drain from the voids 

as explained in Chapter 2. Long-term consolidation-type settlements are generally not 

experienced in cohesionless soils where pore water can drain quickly or in dry or slightly 

moist cohesive soils where significant amounts of pore water are not present. Therefore, 

embankment settlements caused by consolidation of cohesionless or dry cohesive soil 

deposits are frequently ignored as they are much smaller compared to immediate settlements 

in such soils. Consolidation type settlement for saturated cohesive soils is discussed in 

Section 7.5. 

Many methods have been published in the geotechnical literature for the computation of 

immediate settlements in soils or rocks. These methods vary from the use of rules of thumb 

based on experience to the use of complex nonlinear elasto-plastic constitutive models. All 

methods are based on some form of estimate of soil compressibility. In the geotechnical 

literature, soil compressibility is expressed using several different terms such as “bearing 

capacity index,” “compression index,” “elastic modulus,” “constrained modulus,” etc. 

For computing external embankment settlements, the method by Hough (1959) as modified 

by AASHTO (2004 with 2006 Interims) can be used since it is simple and provides a firstorder conservative estimate of immediate settlements. The original Hough method (Hough, 

1959) was based on uncorrected SPT N-values and included recommendations for 

cohesionless as well as cohesive soils such as sandy clay and remolded clay. AASHTO 

modified the Hough (1959) method for use with N160 values and eliminated the 

recommendations for sandy clay and remolded clay. Since the method presented here is 

AASHTO’s version of the Hough method, it will be referred to as the “Modified Hough” 

method. 

Even after the modifications, the settlements estimated by Modified Hough method are 

usually overestimated by a factor of 2 or more based on the data in FHWA (1987). While 

such conservative estimates may be acceptable from the viewpoint of the earthwork 

quantities (see discussion regarding compaction factor in Section 7.4.1.1), they may be 

excessive with respect to the behavior of the structures founded within, under or near the 

embankment. In cases where structures are affected by embankment settlement, more 

refined estimates of the immediate settlements are warranted. For more refined estimates 

of immediate settlements it is recommended that the designer use either the modified 

method of Schmertmann, et al. (1978), which takes into account the strain distribution 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 15 December 2006
 
with depth, or the D’Appolonia (1968, 1970) method, which takes into account the effect 

of preconsolidation. Both methods provide equally suitable results. Schmertmann’s 

modified method is presented in Chapter 8 (Shallow Foundations). 

7.4.1 Modified Hough Method for Estimating Immediate Settlements of 

Embankments 

The following steps are used in Modified Hough method to estimate immediate settlement: 

Step 1. Determine the bearing capacity index (C′) by entering Figure 7-7 with N160 value 

and the visual description of the soil. 

Step 2. Compute settlement by using the following equation. Subdivide the total thickness 

of the layer impacted by the applied loads into 10 ft ± (3 m ±) increments and sum 

the incremental solutions: 

where: ∆H = settlement of subdivided layer (ft) 

 H = thickness of subdivided soil layer considered (ft) 

C′ = bearing capacity index (Figure 7-7) 

 po = existing effective overburden pressure (psf) at center of the subdivided 

layer being considered. For shallow surface deposits, a minimum value 

of 200 psf should be used to prevent unrealistic settlement predictions. 

 ∆p = distributed embankment pressure (psf) at center of the subdivided layer 

being considered 

Note that the term po + ∆p represents the final pressure applied to the foundation subsoil, pf. 

A key point is that the logarithm term in Equation 7-1 incorporates the fundamental feature 

of dissipation of applied stress with depth. The use of Modified Hough method is illustrated 

numerically in Example 7-2. 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 16 December 2006
 

 

⎛ 1 ⎞ p + ∆p ∆H = H ⎜ ⎟log o 

10 7-1 ⎝ C′⎠ po 
CORRECTED SPT N-VALUE, N160 

(Note: The “Inorganic SILT” curve should generally not be applied to soils that exhibit plasticity 

because N-values in such soils are unreliable) 

Figure 7-7. Bearing capacity index (C') values used in Modified Hough method for 

computing immediate settlements of embankments (AASHTO, 2004 with 2006 Interims; 

modified after Hough, 1959). 

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Example 7-2: For the geometry shown in the following figure, determine the settlement at 

the center of a wide embankment placed on a silty sand layer by using 

Modified Hough method and the po diagram. 

 

γt = 120 pcf 

20 ft 

Silty Sand 10 ft 

γt = 120 pcf, N160 = 20 

Solution: 

The original overburden pressure at the center of the 10 ft thick silty sand deposit can be 

computed as po = (10 ft/2) (120 pcf) = 600 psf. Since, the embankment is “wide” the stress 

does not practically dissipate with depth. Therefore, increase in the stress at this depth due to 

the 20 ft high wide embankment can be computed as ∆p = (20 ft) (120 pcf) = 2,400 psf. The 

po diagram based on these values of po and ∆p is shown below. 

Pressure (psf) 

pf po 

0 

Depth (ft) 

1,000 2,000 3,000 4,000

10 

5 600 

3,000 

∆p = 2,400 

From Figure 7-7, find C′ for “silty sand.” Using N160 = 20 and the “silty sand” curve, C′ ≈

58. Find immediate settlement using Equation 7-1 as follows: 

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⎛ 1 ⎞ p o + ∆p ∆H = H⎜ ⎟ log10 ⎝ C′⎠ po 

⎛ 1 ⎞ 600 psf + 2,400 psf ∆H = 10 ft⎜ ⎟log 10	 = 0.12 ft =1.44 in ⎝ 58 ⎠ 600 psf 
7.4.1.1 Comments on the Computed Settlement of Embankments 

The implication of the amount of embankment settlement is that when the embankment is 

completed, additional fill will be required to bring the top of the embankment to the design 

grade. For example, a 1 in (25 mm) settlement on a 60-ft (18 m) wide, 1-mile (5,280 ft or 

1,610 m) long embankment will result in a need for approximately 1,000 yd3

 (~750 m3) of 

additional fill. Some state agencies refer to such settlement estimates as the “compaction 

factor” and note it in the contract plans so that the contractor can make appropriate 

allowances in the bid price to accommodate the additional embankment fill material needed 

to achieve the required design grades. It is in this regard the conservative estimate of the 

settlement resulting from the Modified Hough method may be acceptable and may even be 

preferable to prevent construction change orders. 

7.5 COMPUTATION OF CONSOLIDATION (LONG-TERM) SETTLEMENTS 

Unless the geomaterial is friable, consolidation settlements in fine-grained saturated soils 

occur over a period of time as a function of the permeability of the soils. This concept was 

introduced in discussed in Chapter 2 by using the spring-piston analogy. The features of the 

laboratory consolidation test were discussed in Chapter 5. In this chapter the data obtained 

from the consolidation test are used to demonstrate the computation of long-term settlements 

due to the consolidation phenomena, i.e., primary consolidation and secondary compression. 

Theoretically, a necessary condition for consolidation settlement is that the soil must be 

saturated, i.e., degree of saturation, S = 100%. While the laboratory test for moisture content 

of a soil is inexpensive and relatively straightforward to perform and generally yields 

reliable, reproducible results, there are a number of parameters in consolidation analysis that 

cannot be determined with confidence as indicated by the data in Table 5-25. Therefore, 

depending on the magnitude and configuration of the load with respect to the size and 

moisture content of the compressible soil layer, it is possible that consolidation settlements 

may occur in soils that are judged to be “nearly saturated” but not “fully saturated.” This is 

because such nearly saturated soils may approach full saturation after application of a load of 

sufficient magnitude to cause the pore spaces filled with air to compress (immediate 

settlement) to the extent that the degree of saturation is virtually 100%. Therefore, the 

geotechnical specialist should carefully evaluate the in-situ degree of saturation with respect 

to the degree of saturation of the soil sample at the beginning and end of the consolidation 

test. The geotechnical specialist should also carefully evaluate the reliability of other 

parameters determined during the performance of the consolidation test to make an informed 

judgment regarding the potential for consolidation settlements to occur. Unnecessarily 

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conservative assumptions regarding the magnitude and time rate of consolidation settlements 

may lead to recommendations for deep foundations or for unnecessary implementation of 

costly ground improvement measures. 

Settlement resulting from primary consolidation may take months or even years to be 

completed. Furthermore, because soil properties may vary beneath the location of loading, 

the duration of the primary consolidation and the amount of settlement may also vary with 

the location of the applied load, resulting in differential settlement. If such settlements are 

not within tolerable limits the geotechnical feature as well as a structure founded on or in it 

may be damaged. In the case of embankments, differential settlements that occur along the 

longitudinal axis of the embankment because of changes in thickness and/or consolidation 

properties of underlying clays can cause transverse cracking on the surface of the 

embankment where pavement structures are usually constructed. 

When the areal extent of the applied load is wide compared to the thickness of the 

compressible layer beneath it, a large portion of the soil will consolidate vertically (onedimensionally) with very little lateral displacement because of the constraining forces 

exerted by the neighboring soil elements. However, when the areal extent of the applied load 

is smaller than the thickness of the compressible layer or when there is a finite soft layer at a 

certain depth below the loaded area, significant lateral stresses and associated deformations 

can occur as shown earlier in Figure 2-16 in Chapter 2. Back-to-back retaining walls and a 

narrow embankment for an approach ramp on soft soils are examples of this condition. Due 

to the potential for significant lateral stresses and associated lateral deformations, the 

geotechnical specialist should carefully evaluate the loading geometry with respect to 

subsurface conditions and ascertain whether the problem is 1-D or 3-D. This type of 

evaluation is important because 3-D deformations can affect a number of facilities such as 

buried utilities, bridge foundations, and the stability of embankment slopes. 

The determination of the vertical component of 3-D consolidation deformation is commonly 

based on the one-dimensional consolidation test (ASTM D 2435). Typically, the results of 

the one-dimensional consolidation test are expressed in an e-log p plot which is the so-called 

“consolidation curve.” As indicated in Chapter 5, settlement due to consolidation can be 

estimated from the slope of the consolidation curve. This procedure is generally used in 

practice despite the fact that not all of the points beneath the embankment undergo onedimensional consolidation. However, before the laboratory test results are used, it is very 

important to correct the consolidation curves for the effects of sampling. Thus, before 

proceeding with the discussion of computing consolidation settlements, the correction of the 

laboratory consolidation curves is discussed. 

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7.5.1 Correction of Laboratory One-Dimensional Consolidation Curves 

As indicated in Chapter 3, the process of sampling soils will cause some disturbance no 

matter how carefully the samples are taken. This sampling disturbance will affect virtually 

all measured physical properties of the soil. The sampling disturbance will usually cause the 

“break” in the laboratory consolidation curve to occur at a lower maximum past vertical 

pressure (pc) than would be measured for a truly undisturbed specimen. The effect of 

disturbance from the sampling procedure is illustrated in Figure 7-8 where, for the sake of 

comparison, the vertical strain rather than void ratio (e) is plotted versus the logarithm of the 

vertical effective stress. 

Figure 7-8 shows three consolidation curves for a red-colored plastic clay from Fond du Lac, 

Wisconsin. Samples were taken alternately with 3 in (75 mm) and 2 in (50 mm) thin walled 

samplers. The 3 in (50 mm) sampler apparently caused less disturbance than the 2 in (50 

mm) sampler. The curve for the remolded sample is the flattest curve without a well defined 

break between reloading and virgin compression. 

Figure 7-8. Effect of sample disturbance on the shape of the one-dimensional 

consolidation curve (Reese, et al., 2006). 

Even for good quality samples, it is still necessary to “correct” the e-log p curve since no 

sampling technique is perfect. There are several methods available to correct the 

consolidation curve. The laboratory curve can be corrected according to Figures 7-9a and 7­

9b for normally consolidated and overconsolidated soils, respectively. Table 7-3 presents the 

reconstruction procedures. 

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po pc pvs pf

 po = pc pf 

pvs 

pvs 

Cr 

Cr 

Figure 7-9. Construction of field virgin consolidation relationships 

(adapted from USACE, 1994). 

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Table 7-3 

Reconstruction of virgin field consolidation curve (modified from USACE, 1994). 

Step Description 

a. Normally Consolidated Soil (Figure 7-9a) 

1 By eye choose the point B at the point of minimum radius of curvature (maximum 

curvature) of the laboratory consolidation curve. 

2 Plot point C by the Casagrande construction procedure: (1) Draw a horizontal line through 

point B; (2) Draw a line tangent to the consolidation curve at point B; (3) Draw the 

bisector between the horizontal and tangent lines; and (4) Draw a line tangent to the 

“virgin” portion of the laboratory consolidation curve. Point C is the intersection of the 

tangent to the virgin portion of the laboratory curve with the bisector. Point C indicates the 

maximum preconsolidation (past) pressure pc. 

3 Plot point E at the intersection of a horizontal line through eo and the vertical extension of 

point C, that corresponds to pc as found from Step 2. The value of eo is given as the initial 

void ratio prior to testing in the consolidometer. 

4 Plot point D on the laboratory virgin consolidation curve at a void ratio e = 0.42eo. Extend 

the laboratory virgin consolidation curve to that void ratio if necessary. On the basis of 

many laboratory tests, Schmertmann (1955) found that the laboratory curve for various 

degrees of disturbance intersects the field virgin curve at a value of e= 0.42eo. 

5 The field virgin consolidation curve is the straight line determined by points E and D. 

6 The field compression index, Cc, is the slope of the line ED. 

b. Overconsolidated Soil (Figure 7-9b) 

1 Plot point B at the intersection of a horizontal line through the given eo and the vertical line 

representing the initial estimated in situ effective overburden pressure po. 

2 Draw a line through point B parallel to the mean slope, Cr, of the rebound laboratory curve. 

3 Plot point D by using Step 2 in Table 7-3a for normally consolidated soil. 

4 Plot point F by extending a vertical line through point D up through the intersection of the 

line of slope Cr extending through B. 

5 Plot point E on the laboratory virgin consolidation curve at a void ratio e = 0.42eo. 

6 The field virgin consolidation curve is the straight line through points F and E. The field 

reload curve is the straight line between points B and F. 

7 The field compression index, Cc, is the slope of the line FE. 

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7.5.2 Computation of Primary Consolidation Settlements 

Depending upon the magnitude of the existing effective stress relative to the maximum past 

effective stress at a given depth, in-situ soils can be considered normally consolidated, 

overconsolidated (preconsolidated), or underconsolidated. The behavior of in-situ soils to 

additional loads is highly dependent upon the stress history. The overconsolidation ratio, 

OCR, which is a measure of the degree of overconsolidation in a soil is defined as pc/po. The 

value of OCR provides a basis for determining the effective stress history of the clay at the 

time of the proposed loading as follows: 

•	 OCR = 1 - the clay is considered to be “normally consolidated” under the existing load, 

i.e., the clay has fully consolidated under the existing load (pc = po). 

•	 OCR > 1 - the clay is considered to be “overconsolidated” under the existing load, i.e., 

the clay has consolidated under a load greater than the load that currently exists (pc > po). 

•	 OCR < 1 – the clay is considered to be “underconsolidated” under the existing load, i.e., 

consolidation under the existing load is still occurring and will continue to occur under 

that load until primary consolidation is complete, even if no additional load is applied (pc 

< po). 

The manner in which primary settlements are computed for each of these three conditions 

varies as will be discussed in the following sections. 

7.5.2.1 Normally Consolidated Soils 

The settlement of a geotechnical feature or a structure resting on n layers of normally 

consolidated soils (pc = po) can be computed from Figure 7-10a where n is the number of 

layers into which the consolidating layer is divided: 

where: Cc = compression index 

eo = initial void ratio 

Ho = layer thickness 

po = initial effective vertical stress at the center of layer n 

pf = po+ ∆p = final effective vertical stress at the center of layer n. 

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n Cc ⎛ p ⎞ Sc = ∑ Ho log ⎜ f ⎟	 10 ⎜ ⎟ 7-2 

i 1 + eo ⎝ po ⎠
The final effective vertical stress is computed by adding the stress change due to the applied 

load to the initial vertical effective stress. The total settlement will be the sum of the 

compressions of the n layers of soil. 

Figure 7-10. Typical consolidation curve for normally consolidated soil, (a) Void ratio 

versus vertical effective stress and (b) Vertical strain versus vertical effective stress. 

Normally the slope of the virgin portion of the e-log p curve is determined from the corrected 

one-dimensional consolidation curve measured on specimens taken from each relevant soil in 

the stratigraphic profile. The procedure for determining the corrected curve is presented in 

Table 7-6a. Common correlations for estimating Cc were presented in Section 5.4.6.1 of 

Chapter 5 and can be used to check laboratory results. 

Sometimes the consolidation data is presented in terms of vertical strain (εv) instead of void 

ratio. In this case the slope of the virgin portion of the modified consolidation curve is called 

the modified compression index and is denoted as Ccε as shown in Figure 7-10b. Settlement 

is computed by using Equation 7-3 for normally consolidated soils where all of the other 

terms are defined as for Equation 7-2. 

 n ⎛ p ⎞ Sc = ∑ HoCcε log10 ⎜

⎜ f ⎟

⎟ 7-3 

1 p⎝ o ⎠

By comparing Equations 7-2 and 7-3, it can be seen that Ccε = Cc / (1+eo) 

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7.5.2.2 Overconsolidated (Preconsolidated) Soils 

If the water content of a clay layer below the water table is closer to the plastic limit than the 

liquid limit, the soil is likely overconsolidated, i.e., OCR >1. This means that in the past the 

clay was subjected to a greater stress than now exists. Preconsolidation could have occurred 

because of any number of factors including but not limited to the weight of glaciers which is 

especially prevalent in the northern tier of states and in the northeast, the weight of a natural 

soil deposit that has since eroded away, the weight of a previously placed fill that has since 

been removed, loads due to structures that have since been demolished, desiccation, etc. 

As a result of preconsolidation, the field state of stress will reside on the initially flat portion 

of the e-log p curve. Figures 7-11a and 7-11b illustrate the case where a load increment, ∆p, 

is added so that the final stress, pf, is greater than the maximum past effective stress, pc. For 

this condition, the settlements for the case of n layers of overconsolidated soils will be 

computed from Equation 7-4 or Equation 7-5 that correspond to Figure 7-11a and 7-11b, 

respectively. 

 n Ho ⎛ pc pf ⎞ S = ∑ ⎜

⎜Crlog10 + Cc log10 ⎟

⎟ 7-4 1+ eo ⎝ po p 1 c ⎠

 n ⎛ p p ⎞ S = ∑ Ho ⎜

⎜Crεlog10 

c + Ccε log10 

f ⎟

⎟ 7-5 

p p 1 ⎝ o c ⎠

Figure 7-11. Typical consolidation curve for overconsolidated soil, (a) Void ratio versus 

vertical effective stress and (b) Vertical strain versus vertical effective stress. 

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The total settlement is computed by summing the settlements computed from each 

subdivided compressible layer within the zone of influence (ZI). The assumption is made 

that the initial and final stress calculated at the center of each sublayer is representative of the 

average stress for the sublayer, and the material properties are reasonably constant within the 

sublayer. The sublayers are typically 5 ft (1.5 m) to 10 ft (3 m) thick in highway 

applications. In cases where the various stratigraphic layers represent combinations of both 

normally and overconsolidated soils, the settlement is computed by using the appropriate 

combinations of Equations 7-2 through 7-5. 

7.5.2.3 Underconsolidated Soils 

Underconsolidation is the term used to describe the effective stress state of a soil that has not 

fully consolidated under an existing load, i.e., OCR < 1. Consolidation settlement due to the 

existing load will continue to occur under that load until primary consolidation is complete, 

even if no additional load is applied. This condition is shown in Figure 7-12 by ∆po. 

Therefore, any additional load increment, ∆p, would have to be added to po. Consequently, if 

the soil is not recognized as being underconsolidated, the actual total primary settlement due 

to ∆po+∆p will be greater than the primary settlement computed for an additional load ∆p 

only, i.e., the settlement may be under-predicted. As a result of under-consolidation, the field 

state of stress will reside entirely on the virgin portion of the consolidation curve as shown in 

Figure 7-12. The settlements for the case of n layers of under-consolidated soils are 

computed by Equation 7-6 or Equation 7-7 that correspond to Figure 7-12a and 7-12b, 

respectively. 

Figure 7-12. Typical consolidation curve for under-consolidated soil – (a) Void ratio 

versus vertical effective stress and (b) Vertical strain versus vertical effective stress. 

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n H ⎛ p ⎞ o po f S = ∑ ⎜ C log + C log ⎟

1 ⎜ c 10 c 10 ⎟ 7-6+ 1 eo ⎝ pc po ⎠

n ⎛ p p ⎞ S = ∑ H ⎜ o ⎜C o 

cεlog10 + Ccε log10

f ⎟

⎟ 7-7

p p 1 ⎝ c o ⎠

7.5.3 Consolidation Rates (Time Rate of Consolidation Settlement) 

The rate of consolidation should be considered for the design of geotechnical features and 

structures on compressible clay. For example, a geotechnical feature such as an embankment 

will settle relative to a bridge foundation supported on piles, creating an undesirable “bump 

at the end of the bridge.” Hence, time rate of consolidation, as well as differential 

settlements between the bridge and embankment, is important. The concept of time rate of 

consolidation is explained with respect to Figure 7-13. 

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Drainage 

ust

a c

C1

 C2

Consolidating 

Layer 

us ∆u ∆po

b d usb ∆ui = ∆pt

Drainage 

Definitions: ust = hydrostatic pore water pressure at top of layer 

usb = hydrostatic pore water pressure at bottom of layer 

us = hydrostatic pore water pressure at any depth 

∆ui = initial excess pore water pressure 

∆u = excess pore water pressure at any depth after time t 

ut = us + ∆u = total pore water pressure at any depth after time t 

Figure 7-13. Diagram illustrating consolidation of a layer of clay between two pervious 

layers (modified after Terzaghi, et al. 1996).
•	 The initial hydrostatic pore water pressure distribution, us, is assumed to be linear in a 

layer of saturated clay. Line a-b in Figure 7-13 shows the initial hydrostatic pore water 

pressure distribution through a clay layer at a certain depth below the ground water 

elevation where ust is the pore water pressure at the top of the clay layer and usb is the 

pore water pressure at the bottom of the clay layer. Experimental measurement of pore 

pressures in saturated clays subjected to one-dimensional loading indicate that when a 

load is applied the pore water pressure will instantaneously increase an amount equal to 

the total vertical stress increment, ∆pt, uniformly throughout the entire thickness of the 

consolidating layer as shown by a-c-d-b in Figure 7-13. The initial increase in the pore 

water pressure, ∆ui, above the static value is called the initial excess pore water pressure 

and it is equal to ∆pt. The total initial pore water pressure which is the sum of the 

hydrostatic pressure and the initial excess pore water pressures is shown as line c-d in 

Figure 7-13. 

•	 With time, water will drain out of the consolidating layer to relieve the excess pore water 

pressure and the applied total vertical stress increment, ∆pt, will be slowly transferred to 

the soil particles, i.e., at any given time after application of the load, the initial excess 

pore water pressure will decrease at all depths to an excess pore water pressure having a 

value less than of ∆ui. The pattern of the excess pore water pressure at any given time is 

not parallel to line c-d, but is curvilinear similar to curve C1 in Figure 7-13. Curves such 

as C1 and C2 are known as iscochrones because they are lines of equal time. The 

difference between the line a-b and curve C1, for example, represents the excess pore 

water pressure, ∆u, at any point within the consolidating layer at any time after 

application of the vertical load stress increment, ∆pt. 

•	 If the clay layer is confined between two sand layers that are more permeable, the initial 

excess pore water pressure will drop immediately to zero at the drainage boundaries as 

shown in Figure 7-13 and the total vertical stress increment ∆pt and will be equal to the 

effective vertical stress increment, ∆po. The rate of this transfer with depth depends upon 

the boundary drainage conditions. With time, the vertical distribution of excess pore 

water pressure within the consolidating layer will evolve from the initial distribution (ac-d-b), to the C1 distribution, to the C2 distribution, and finally to the initial distribution 

of the hydrostatic pressure represented by line a-b. 

•	 At any depth, the difference between a pore water pressure isochrone, such as C1, and the 

initial excess pore water pressure c-d is equal to the effective vertical stress increment, 

∆po, i.e., the amount of ∆pt that has been transferred to the soil structure. Since the 

isochrone C1 develops after a certain period of time, the difference between C1 and c-d 

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also represents the distribution of the effective stress increments with depth at a given 

time after application of load. 

•	 Note that the distribution shown in Figure 7-13 pertains only to the specific boundary 

drainage condition where a more permeable material exists above and below the 

consolidating clay layer. In this case the clay layer is considered to be “doubly drained” 

with the longest distance to a drainage boundary being half the layer thickness. If the 

clay layer is underlain by a less permeably material (e.g., rock), drainage will occur in 

only one direction and the isochrones at a given time will be different from those shown 

for double drainage in Figure 7-13. In this case the clay layer is considered to be “singly 

drained” with the longest distance to a drainage boundary being the entire layer 

thickness. During the consolidation process the principle of effective stress will be in 

operation at every depth, i.e., ∆pt = ∆u + ∆po and settlement will be occurring due to the 

effective stress increment ∆po. The drainage boundary condition will affect the time it 

takes for settlement to occur, but it has no effect on the magnitude of settlement, which is 

determined by use of the equations presented previously in which settlement is a function 

of ∆po only. 

7.5.3.1 Percent Consolidation 

As indicated previously, immediately after application of load, ∆u, will drop to zero at the 

drainage boundaries because the water will drain immediately into the more pervious layers. 

Since the excess pore water pressure is zero at the drainage boundaries, the soil there has 

undergone 100% consolidation. However, at interior points, the pore water pressure 

dissipates more slowly with time depending on the permeability of the compressible soil. At 

any time after application of a load, the actual degree or percentage of consolidation at a 

given depth is defined as (∆ui-∆u)/∆ui, where ∆u is the excess pore water pressure at that 

depth at that time and ∆ui = the initial excess pore water pressure which, as indicated 

previously, equals the total stress increment ∆pt. Thus, where ∆ui = ∆u (i.e., at the instant of 

loading), the percent consolidation is zero. When ∆u= 0 (i.e., at the end of consolidation), 

the percent consolidation is 100. This relationship is valid at any depth within the 

consolidating layer at any time from the instant of loading to the completion of primary 

consolidation. 

While plots of the type shown in Figure 7-13 give an indication of the pore pressure variation 

within the consolidating layer at any time and are useful to explain the theory of 

consolidation, from a practical viewpoint it is usually more beneficial to obtain the average

degree or percent of consolidation, U, within the entire layer to indicate when the entire clay 

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layer has undergone a certain average amount of consolidation of say 10, 50, or 80 percent. 

With reference to Figure 7-13, the average degree of consolidation at any time is defined as 

the difference between the area under the initial excess pore water pressure curve (a-c-d-b) 

and the area under the isochrone at that time, e.g., the cross hatched area under isochrone C2 

divided by the area under the initial excess pore water pressure curve (a-c-d-b). The result is 

expressed as a percentage. Therefore, at the instant ∆pt is applied the area under the 

isochrone is exactly equal to the area (a-c-d-b) as indicated above and the average percent 

consolidation (U) equals zero. At the end of primary consolidation all excess pore water 

pressures have dissipated and the area under the isochrone is zero. Thus, the average percent 

consolidation (U) equals 100. Since, according to the principle of effective stress, ∆pt = ∆u + 

∆po, the amount of settlement at any time after the application of load is directly related to 

the amount of consolidation that has taken place up to that time. As a practicality the 

average degree of consolidation at any time, t, can be defined as the ratio of the settlement at 

that time, St, to the settlement at the end of primary consolidation, Sultimate, when excess pore 

water pressures are zero throughout the consolidating layer, i.e., U= St/Sultimate. This 

relationship is used to develop a so-called “settlement-time curve” as will be discussed later. 

Table 7-4 shows the average degree of consolidation (U) corresponding to a normalized time 

expressed in terms of a time factor, Tv, where: 

 

2 c t v T Hv d Tv = 2 , which can be written as t = H c d v

7-8 

where: cv = coefficient of consolidation (ft2/day) (m2/day) 

Hd = the longest distance to a drainage boundary (ft) (m) 

t = time (day). 

Any consistent set of units can be used in Equation 7-8 since Tv is dimensionless. As 

indicated previously, the longest drainage distance of a soil layer confined by more 

permeable layers on both ends is equal to one-half of the layer thickness. When confined by 

a more permeable layer on one side and an impermeable boundary on the other side, the 

longest drainage distance is equal to the layer thickness. The value of the dimensionless time 

factor Tv may be determined from Table 7-4 for any average degree of consolidation. U. The 

actual time, t, it takes for this percent of consolidation to occur is a function of the boundary 

drainage conditions, i.e., the longest distance to a drainage boundary, as indicated by 

Equation 7-8. By using the normalized time factor, Tv, settlement time can be computed for 

various percentages of settlement due to primary consolidation, to develop a predicted 

settlement-time curve. A typical settlement-time curve for a clay deposit under an 

embankment loading is shown in Figure 7-14. 

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Table 7-4 

Average degree of consolidation, U, versus Time Factor, Tv, 

for uniform initial increase in pore water pressure 

U % Tv 

0 0.000 

10 0.008 

20 0.031 

30 0.071 

40 0.126 

50 0.197 

60 0.287 

70 0.403 

80 0.567 

90 0.848 

93.1 1.000 

95.0 1.163 

98.0 1.500 

99.4 2.000 

100.0 Infinity 

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 Settlement 

Time, t 

Figure 7-14. Typical settlement-time curve for clay under an embankment loading. 
7.5.3.2 Step-by-Step Procedure to Determine Amount and Time for Consolidation 

The step-by-step process for determining the amount of and time for consolidation to occur 

for a single-stage construction of an embankment on soft ground is outlined below: 

1.	 From laboratory consolidation test data determine the e-log p curve and estimate the 

change in void ratio that results from the added weight of the embankment. Create 

the virgin field consolidation curve by using the guidelines presented in Table 7-3. 

2.	 Determine if the foundation soil is normally consolidated, overconsolidated or underconsolidated. 

3.	 Use Equations 7-2 to 7-7 to compute the primary consolidation settlement for 

normally consolidated, overconsolidated and under-consolidated foundation soils. 

4.	 Determine cv from laboratory consolidation test data. Two graphical procedures are 

commonly used for this determination are the logarithm-of-time method (log t) 

proposed by Casagrande and Fadum (1940) and the square-root-of-time method 

( t ) proposed by Taylor (1948). These methods are shown in Figures 7-15 and 7­

16, respectively. Because both methods are different approximations of theory, they 

do not give the same answers. Often the t method gives slightly greater values of cv 

than the log t method. 

5.	 Use Equation 7-8 to calculate the time to achieve 90% - 95% primary consolidation. 

For a more detailed discussion on the consolidation theory, the reader is referred to Holtz and 

Kovacs (1981). An alternative approach to hand calculations is the use of a computerized 

method. For example, program FoSSA (2003) by ADAMA Engineering, Version 1.0 

licensed to FHWA, which was introduced in Chapter 2, calculates the time rate of settlement 

for various boundary conditions including the effects of staged construction and strip drains 

in addition to calculating the stresses and settlements. FoSSA (2003) also allows for 

simulation of multiple layers undergoing simultaneous consolidation. In any event, the stepby-step hand calculations can serve to verify the correctness of benchmark cases and thereby 

be used to ascertain the correctness of any computerized procedure. 

 

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Primary 

Consolidation 

Secondary 

Compression t100 

Step-by-step procedure: 

1.	 Plot the dial readings for sample deformation for a given load increment against time on a 

semi-log paper. 

2.	 Plot two points, P and Q on the upper portion of the consolidation curve which 

correspond to time t1 and t2, respectively. Note that t2 = 4 t1. 

3.	 The difference of the dial readings between P and Q is equal to x. Locate point R, which 

is at a distance x above point P. 

4.	 Draw the horizontal line RS. The dial reading corresponding to this line is d0, which 

corresponds to 0% consolidation. 

5.	 Project the straight-line portions of the primary consolidation and the flatter portion 

towards the end of the consolidation curve to intersect at T. The dial reading 

corresponding to T is d100, i.e., 100% primary consolidation. The sample deformation 

beyond t100 is due to secondary compression (see Section 7.5.4). 

6.	 Determine the point V on the consolidation curve which corresponds to a dial reading of 

(d0+d100)/2 = d50. The time corresponding to the point V is t50, i.e., 50% consolidation. 

7.	 Determine cv from Equation 7-8 for desired U. Example: For U=50% the value of Tv for 

is 0.197 from Table 7-4. Thus, cv can be determined as follows: 

2 0.197 Hd c	 v = t50 

Figure 7-15. Logarithm-of-time method for determination of cv. 

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Step-by-step procedure: 

1. Plot the dial reading and the corresponding square-root-of-time, t . 

2. Draw the tangent PQ to the early portion of the plot. 

3. Draw a line PR such that OR = (1.15) (OQ). 

4.	 The abscissa of the point S (i.e., the intersection of PR and the consolidation curve) will 

give t90 , i.e., the square-root-of-time for 90% consolidation. 

5.	 Determine cv from Equation 7-8 for U=90%. From Table 7-4, the value of Tv for U=90% 

is 0.848. Thus, cv can be determined as follows: 

0.848 H2

c d 	 v = t90 

Figure 7-16. Square-root-of-time method for determination of cv. 

Comments on cv value: The value of cv is determined for a given load increment. It 

varies from increment to increment and is different for loading and unloading. Moreover, 

cv, usually varies considerably among samples of the same soil. Therefore, if the actual 

rate of consolidation is critical to the design, as in certain stability problems where 

excess pore water pressures must be known accurately, pore pressures must 

actually be measured in the field as construction proceeds. 

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Regardless of whether hand-calculations or computerized methods are used, the important 

factors to remember are: 

•	 the time required for consolidation is proportional to the square of the longest 

distance required for water to drain from the deposit and, 

•	 the rate of settlement decreases as time increases. 

The maximum length of vertical drainage path, Hd, bears further explanation. This term 

should not be confused with the H term in the equation for the computation of the settlement 

magnitude. H is an arbitrarily selected value usually representing a portion of the total 

compressible layer thickness. For calculating the magnitude of settlement the sum of the 

sublayer H values must equal the total thickness of the clay layer. For calculating the time 

rate of settlement, the Hd term in Equation 7-8 is the maximum vertical distance that a water 

molecule must travel to escape from the compressible layer to a more permeable layer. In 

the case of a 20 ft (6 m) thick clay layer bounded by a sand layer on top and a virtually 

impermeable rock stratum on the bottom, the Hv term would equal to 20 ft (6 m). The water 

molecule must travel from the bottom of the layer to the top of the layer to escape, i.e., single 

drainage. However, if the clay layer was bounded top and bottom by more permeable sand 

deposits, the Hv distance would be 10 ft (3 m). The water molecule in this case, needs only 

to travel from the center of the layer to either boundary to escape, i.e., double drainage. 

However, regardless of the boundary drainage conditions, the sum of the sublayer H values 

must equal 20 ft (6 m) in the settlement computations. 

The mechanism for determining the maximum horizontal path for escape of a water molecule 

is similar. The influence of horizontal drainage may be significant if the width of the loaded 

area is small. For instance, during consolidation under a long, narrow embankment, a water 

molecule can escape by traveling a distance equal to one half the embankment width. 

However, for very wide embankments the beneficial effect of lateral drainage may be small 

as the time for lateral escape of a water molecule increases as the square of one-half the 

embankment width. 

The concepts of consolidation settlement and time rates of consolidation with reference to an 

embankment loading are illustrated by the following example. 

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Example 7-3: Determine the magnitude of and the time for 90% consolidation for the 

primary settlement of a “wide” embankment by using the po diagram. 

γt = 120 pcf 

20′

Clay (normally consolidated) γt = 120 pcf, 10′ Cc = 0.5, eo = 1.0, cv = 0.2 ft2

/day 

Rock 

0 

Depth (ft.) 

Pressure (psf) 

1,000 2,000 3,000 4,000

10 

5 600 3,000 

∆p = 2,400 

po pf 

Solution: 

Since the embankment is “wide,” the vertical stress at the base of the embankment is 

assumed to be the same within the 10-ft thick clay layer. Since soil is normally consolidated, 

use Equation 7-2 to determine the primary consolidation settlement as follows: 

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C p + ∆p c o ∆H = H log10 1 + e p o o 

⎛ 0.5 ⎞ 600 psf + 2,400 psf ∆H = 10 ft ⎜ ⎟log10 = 1.75 ft = 21 inches (0.53 m) ⎝1 + 1.0 ⎠ 600 psf 
Find time for 90% consolidation use Tv = 0.848 from Table 7-4. Assume single vertical 

drainage due to impervious rock underlying clay layer and use Equation 7-8 to calculate the 

time required for 90% consolidation to occur. 

2 THd t 90 = cv 

(0.848)(10 ft)2 

t90 =	 = 424 days

0.2 ft2 / day 

7.5.4 Secondary Compression of Cohesive Soils 

Secondary compression is the process whereby the soil continues to displace vertically after 

the excess pore water pressures are dissipated to a negligible level i.e., primary compression 

is essentially completed. Secondary compression is normally evident in the settlement-log 

time plot when the specimen continues to consolidate beyond 100 percent of primary 

consolidation, i.e., beyond t100, as shown in Figure 7-15. An example is shown in Figure 7­

17, where secondary compression occurs beyond t100 = 392 mins. There are numerous 

hypotheses as to the reason for the secondary compression. The most obvious reason is 

associated with the simplifications involved in the theory of one-dimensional consolidation 

derived by Terzaghi. More rigorous numerical solutions accounting for the simplifications 

can often predict apparent secondary compression effects. 

The magnitude of secondary compression is estimated from the coefficient of secondary 

compression, Cα, as determined from laboratory tests by using Equation 7-9 that is derived 

from Figure 7-17. 

∆e Cα = ⎛ t ⎞ ⎜ 2 lab	 ⎟ 7-9 log 10⎜ t ⎟ 1lab ⎝ ⎠

where: t1 lab = time when secondary compression begins and is typically taken as the time 

when 90 percent of primary compression has occurred 

t2 lab = an arbitrary time on the curve at least one log-cycle beyond t90 or the time 

corresponding to the service life of the structure 

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Figure 7-17: Example time plot from one-dimensional consolidometer test for 

determination of secondary compression (USACE, 1994). (1 in = 25.4 mm) 

The settlement due to secondary compression (Ss) is then determined from Equation 7-10. 

where: t1 = time when approximately 90 percent of primary compression has occurred for 

the actual clay layer being considered as determined from Equation 7-8. 

t2 = the service life of the structure or any other time of interest. 

The values of Cα can be determined from the dial reading vs. log time plots associated with 

the one-dimensional consolidation test as shown in Figure 7-17. Typical ranges of the ratio 

of Cα/Cc presented in Section 5.4.6.4 of Chapter 5 can be used to check laboratory test 

results. 

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C α ⎛ t ⎞ S 2 ⎟

⎟

	 s = Hc log ⎜ 10 ⎜ 7-10 1+ e o ⎝ t1 ⎠
7.6 LATERAL SQUEEZE OF FOUNDATION SOILS 

When the geometry of the applied load is larger than the thickness of the compressible layer 

or when there is a finite soft layer within the depth of significant influence (DOSI) below the 

loaded area, significant lateral stresses and associated lateral deformations can occur as 

shown earlier in Figure 2-16 in Chapter 2. For example, as shown in Figure 7-18, if the 

thickness of a soft soil layer beneath an embankment fill is such that it less than the width, be, 

of an end or side slope, then the soft soil may squeeze out. 

be 

Soft Soil 

Fill 

Figure 7-18. Schematic of lateral squeeze phenomenon. 

The lateral squeeze phenomenon is due to an unbalanced load at the surface of the soft soil. 

The lateral squeeze behavior may be of two types, (a) short-term undrained deformation that 

results from a local bearing capacity type of deformation, or (b) long-term drained, creeptype deformation. Creep refers to the slow deformation of soils under sustained loads 

over extended periods of time and can occur at stresses well below the shear strength of 

the soil. As discussed in Section 5.4.1, secondary compression is a form of creep 

deformation while primary consolidation is not. 

The lateral squeeze phenomenon can be observed in the field. For example, field 

observations and measurements have shown that some bridge abutments supported on piles 

driven through compressible soils tilted toward the embankment fill. Many of the abutments 

experienced large horizontal movements resulting in damage to the structure. The cause of 

this problem is attributed to the unbalanced fill load, which "squeezes" the soil laterally as 

discussed previously. This "lateral squeeze" of the soft foundation soil can apply enough 

lateral thrust against the piles to bend or even shear the piles. This problem is illustrated in 

Figure 7-19. The bridge abutment may tilt forward or backward depending on a number of 

factors including the relative configuration of the fill and the abutment, the relative stiffness 

of the piles or shafts and the soft deposit, the strength and thickness of the soft layer, rate of 

construction of the fill, and depth to bearing layer. 

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(+) (-) (+) 

Figure 7-19. Examples of abutment tilting due to lateral squeeze (FHWA, 2006a). 

7.6.1 Threshold Condition for Lateral Squeeze 

Experience has shown that lateral squeeze of the foundation soil can occur and abutment 

tilting may result if the surface load applied by the weight of the fill exceeds 3 times the 

undrained shear strength, su, of the soft foundation soil, i.e., 

(γ)(H) > 3su 7-11

where, γ is the unit weigh of the fill and H is the height of the fill. The possibility of 

abutment tilting can be evaluated in design by using the above relationship. Whether the 

lateral squeeze will be short-term or long-term can be determined by evaluating the 

consolidation rate of settlement with respect to the rate of application of the load. For 

practical purposes, the unit weight of an embankment fill can be assumed to be 

approximately 125 pcf (19.7 kN/m3

). The undrained shear strength, su, of the foundation soil 

can be determined either from in-situ field vane shear tests or from laboratory triaxial tests 

on high quality undisturbed Shelby tube samples. 

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7.6.2 Calculation of the Safety Factor against Lateral Squeeze 

The safety factor against failure by squeezing, FSSQ, may be calculated by Equation 7-12 

(Silvestri, 1983). The geometry of the problem and the forces involved are shown in Figure 

7-20. 

⎡ 2su ⎤ ⎡4.14su ⎤ FSSQ = ⎢ ⎥ + ⎢ ⎥ 7-12 ⎣ γD tan S θ⎦ ⎣ γH ⎦

where: θ = angle of slope 

γ = unit weight of the fill 

DS = depth of soft soil beneath the toe of the end slope or side slope of the fill 

H = height of the fill 

su = undrained shear strength of soft soil beneath the fill 

Caution is advised when Equation 7-12 is used. It was found that when FSSQ < 2, a rigorous 

slope stability analysis and possibly advanced numerical analysis, e.g., finite element 

analysis should be performed. When the depth of the soft layer, DS, is greater than the base 

width of the end slope, b=H/tanθ, general slope stability behavior governs the design. In that 

case, the methods described in Chapter 6 (Slope Stability) may be used to evaluate the 

stability of the foundation soils. 

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θ

σz = (2τx)/Ds 

x 

τ = su 

τ = su 

H 

Ds 

Fill 

Soft Soil 

Figure 7-20. Definitions for calculating safety factor against lateral squeeze (after 

Silvestri, 1983). 
7.6.3 Estimation of Horizontal Movement of Abutments 

The amount of horizontal movement the abutment may experience can also be estimated in 

design. Information from case histories for nine structures where measurements of abutment 

movements occurred is documented in Table 7-5. 

The data presented in Table 7-5 provides a basis for estimating horizontal movement for 

abutments under similar conditions, provided a reasonable estimate of the post-construction 

fill settlement is made by using data from consolidation tests on high quality undisturbed 

Shelby tube samples. Note that the data for the abutments listed in Table 7-5 shows the 

horizontal movement (tilt) to range from 6 to 33% of the vertical fill settlement, with the 

average being 21%. Therefore, as a first approximation, it can be said that if the fill load 

exceeds the 3su limit prescribed by Equation 7-11, then the horizontal movement (tilt) of an 

abutment can be reasonably estimated as approximately 25% of the vertical fill settlement for 

the conditions listed in Table 7-5. 

Table 7-5 

Summary of abutment movements (Nicu, et al., 1971) 

Foundation Fill 

Settlement 

(inches) 

Abutment 

Settlement 

(inches) 

Abutment 

Tilting 

(inches) 

Ratio of Abutment 

Tilting to Fill 

Settlement 

Steel H-piles 16 Unknown 3 0.19 

Steel H-piles 30 0 3 0.10 

Soil bridge 24 24 4 0.17 

Cast-in-place pile 12 3.5 2.5 0.19 

Soil bridge 12 12 3 0.25 

Steel H-piles 48 0 2 0.06 

Steel H-piles 30 0 10 0.33 

Steel H-piles 5 0.4 0.5 to 1.5 0.1 to 0.3 

Timber Piles 36 36 12 0.33 

7.7 DESIGN SOLUTIONS - DEFORMATION PROBLEMS 

Both the magnitude and time rate of settlement can affect fill structures, which in turn may 

affect the performance of other structures such as bridge abutments that are built within or in 

the vicinity of the fills. There are various methods to reduce the magnitude and time rate of 

settlement. All of these methods can be considered as ground improvement and are 

discussed in detail in FHWA (2006b). Two of these methods are briefly discussed in this 

manual. The reader is referred to FHWA (2006b) for further details. Solutions to prevent 

abutment tilting due to lateral squeeze are discussed in Section 7.7.3. 

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7.7.1 Reducing the Amount of Settlement 

Settlement can be reduced by either increasing the resistance or reducing the load. Several 

ground improvement methods that are particularly suitable for reducing the amount of 

settlement are noted below. 

7.7.1.1 Category 1 - Increasing the Resistance 

Common ground improvement techniques that increase resistance include the following: 

•	 Excavation and recompaction. 

•	 Excavation and replacement. 

•	 Vertical inclusions such as stone columns, shafts and piles. Embankments supported 

in this way are known as column supported embankments. 

•	 Horizontal inclusions such as geosynthetics. 

•	 Grouting, e.g., soil mixing, jet grouting. 

•	 Dynamic compaction. 

7.7.1.2 Category 2 - Reducing the Load 

Common load reduction techniques include the following: 

•	 Reduce grade line (reduction in height and/or flattening the slope) 

•	 Use lightweight fill material, e.g., expanded shale, foamed concrete, geofoam. 

•	 Bypass the soft layer with a deep foundation. Deep foundations may be used in 

conjunction with a load transfer platform (see FHWA 2006b). 

7.7.2 Reducing Settlement Time 

Often the major design consideration related to a settlement problem is the time for the 

settlement to occur. Low permeability clays and silty clays can take a long time to 

consolidate under an applied load. The settlement time is critical on most projects because it 

has a direct impact on construction schedules and delays increase project costs. Settlement 

time is also important to the maintenance personnel of a highway agency. The life cycle cost 

of annual regrading and resurfacing of settling roadways is usually far greater than the cost 

of design treatments to eliminate settlement before or during initial construction. 

The two most common methods used to accelerate settlement and reduce settlement time are: 

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1. Application of surcharge. 

2. Installation of vertical drains in the foundation soils. 

Note that both of the above techniques lead to an increase in the resistance. These 

techniques are briefly discussed below and their use is illustrated in the Apple Freeway 

example in Appendix A. 

7.7.2.1 Surcharge Treatment 

An embankment surcharge is constructed to a predetermined height, usually 1 to 10 ft (0.3 to 

3 m) above final grade elevation based on settlement calculations. The surcharge is 

maintained for a predetermined waiting period (typically 3 to 12 months) based on 

settlement-time calculations. Depending upon the strength of the consolidating layer(s) the 

surcharge may have to be constructed in stages. The actual dimensions of the surcharge and 

the waiting period for each stage depend on the strength and drainage properties of the 

foundation soil as well as the initial height of the proposed embankment. The length of the 

waiting period can be estimated by using laboratory consolidation test data. The actual 

settlement occurring during embankment construction is then monitored with geotechnical 

instrumentation. When the settlement with surcharge equals the settlement originally 

estimated for the embankment, the surcharge is removed, as illustrated in Figure 7-21. 

If the surcharge is not removed after the desired amount of settlement has occurred, then 

additional settlement will continue to occur. Note that the stability of a surcharged 

embankment must be checked as part of the embankment design to ensure that an adequate 

short term safety factor exists. The stability is often field verified by monitoring with 

instrumentation such as inclinometers, piezometers and settlement points as discussed later. 

Figure 7-21. Determination of surcharge time required to achieve desired settlement. 

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7.7.2.2 Vertical Drains 

Primary consolidation of some highly plastic clays can take many years to be completed. 

Surcharging alone may not be effective in reducing settlement time sufficiently since the 

longest distance to a drainage boundary may be significant. In such cases, vertical drains can 

be used to accelerate the settlement, either with or without surcharge treatment. The vertical 

drains accelerate the settlement rate by reducing the drainage path the water must travel to 

escape from the compressible soil layer to half the horizontal distance between drains, as 

illustrated in Figure 7-22. In most applications, a permeable sand blanket, 2 to 3 ft (0.6 to 1 

m) thick, should be placed on the ground surface to permit free movement of water away 

from the embankment area and to create a working platform for installation of the drains. 

The drains are installed prior to placement of the embankment. The applied pressure from 

the embankment generates excess pore water pressure. 

Recall that the consolidation time is proportional to the square of the length of the longest 

drainage path. Thus if the length of the drainage path is shortened by 50%, the consolidation 

time is reduced by a factor of four. Vertical drains and sand blankets should have high 

permeability to allow the water squeezed out of the subsoil to travel relatively quickly 

through the drains and the blanket. 

Figure 7-22. Use of vertical drains to accelerate settlement (NCHRP, 1989). 

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Wick drains are small prefabricated drains consisting of a plastic core that is wrapped with 

geotextile. Wick drains are typically 4 inches (100 mm) wide and about 1/4 inch (7 mm) 

thick. The drains are produced in rolls that can be fed into a mandrel. Wick drains are 

installed by pushing or vibrating a mandrel into the ground with the wick drain inside. When 

the bottom of the compressible soil is reached, the mandrel is withdrawn and the trimmed 

portion of the wick drain left in the ground. To minimize smear of the compressible soil, the 

cross-sectional area of the mandrel is recommended to be limited to a maximum of about 10 

in2

 (6,450 mm2). Predrilling of dense soil deposits may be required in some cases to reach 

the design depth. Use of wick drains in the United States began in the early 1970s. Design 

and construction guidance on the use of wick drains is provided in FHWA (1986, 2006b). 

The feasibility of a surcharge solution should always be considered first since vertical drains 

are generally more expensive. 

7.7.3 Design Solutions to Prevent Abutment Tilting 

A recommended solution to minimize abutment-tilting is to induce settlement of the fill 

before the abutment piles or shafts are installed. If the construction time schedule or other 

factors do not permit pre-consolidation of the foundation soils before the piles or shafts are 

installed, then abutment tilting issues can be mitigated by the following design provisions: 

1.	 Use sliding plate expansion shoes large enough to accommodate the anticipated 

horizontal movement. 

2.	 Make provisions to fill in the bridge deck expansion joint over the abutment by 

inserting either metal plate fillers or larger neoprene joint fillers. 

3.	 Design the deep foundations for downdrag forces due to settlement. This solution 

does not improve the horizontal displacement effects. 

4.	 Use backward battered piles at the abutment and particularly at the wingwalls. 

5.	 Use lightweight fill materials to reduce driving forces 

Displacements should also be monitored during and after construction so that the predicted 

movements can be compared to actual displacements. Displacements should be monitored 

by survey monuments or protected prisms installed on the face of the abutment and 

wingwalls and should be tied into permanent benchmarks. 

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7.8 PRACTICAL ASPECTS OF EMBANKMENT SETTLEMENT 

Few engineers realize the influence of embankment construction on the response of subsoils. 

The total weight of an embankment has an impact on the type of foundation treatment that 

may be appropriate. For instance, a relatively low height embankment of 10 ft (3 m) may be 

effectively surcharged because the additional surcharge weight could be 30 to 40 percent of 

the proposed embankment weight. However, when the embankment height exceeds 50 ft (15 

m) the influence of a 5 or 10 ft (1.5 or 3 m) trapezoid of soil on top of this heavy 50 ft (15 m) 

mass is small and probably not cost-effective. Conversely, as the embankment height 

increases, the use of a shallow foundation for support of the abutment becomes more 

attractive. A 30 ft (9 m) high, 50 ft (15 m) long approach embankment weighs about 15,000 

tons (130 MN) compared to the insignificant weight of a total (stub type) abutment loading 

that may equal 1,000 tons (9 MN). The width of an embankment also has an effect on total 

settlement. Wider embankments cause a pressure increase deeper into the subsoil. As might 

be expected, wider embankments may also cause more immediate and consolidation 

settlement and increase the time for consolidation to occur. 

Recent developments in computer software readily permit computer analysis of approach 

embankment settlement. Programs such as FoSSA (2003), discussed in Chapter 2, allow the 

user to compute settlements along abutments and to evaluate the effects of settlements on 

pipes buried in end slopes or pipes placed diagonally under approach fills. 

7.9 CONSTRUCTION MONITORING AND QUALITY ASSURANCE 

Approach embankment construction should be clearly defined in standard drawings as to 

materials and limits of placement. Such standards assure uniformity in construction due to 

the familiarity of the construction personnel with the operations being performed and results 

expected. Designers should attempt to use standard details wherever possible. Attempts at 

small changes in materials or limits are generally counterproductive to good construction 

where repetition of good practice is an important factor. 

The philosophy of approach embankment details is to insure adequate bearing capacity for 

abutments or piers placed in the embankment and to minimize settlement of the pavement or 

footing. Typical highway embankments require compaction to 90 percent of maximum dry 

density (AASHTO T180) to control pavement settlement. Designers should specify 

materials and compaction control as shown in Figure 7-4, to limit differential 

settlement between the structure and approach fill. If piles are used to support footings in 

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fill, the largest particle size of embankment material should be limited to 6 in (150 mm) to 

ease pile installation either by driving or pre-drilling. If spread footings are used, a minimum 

of 5 ft (1.5 m) of select material compacted to 100 percent of maximum dry density 

(AASHTO T99) should be placed beneath the footing and extended beyond the wingwalls. 

This layer provides uniform support for the footing and a rigid transition between the 

structure and the fill to minimize differential settlement. Construction control is usually 

referenced to percent compaction on the standard design drawings. 

7.9.1 Embankment Construction Monitoring by Instrumentation 

The observational approach to design involves monitoring subsoil behavior during early 

construction stages to verify design and to predict responses to subsequent construction. 

Basic soil mechanics concepts can be used to predict future subsoil behavior accurately if 

data from instrumentation are analyzed after initial construction loads have been placed. 

Occasionally a design problem arises that is unique or extremely critical and that can be 

safely solved only by utilizing the observational approach. 

Embankment placement must be carefully observed and monitored on projects where 

stability and/or settlement are critical. The monitoring should include visual observation by 

the construction inspection staff and the use of instrumentation. Without the aid of various 

forms of instrumentation, it is impossible to determine accurately what is happening to the 

foundation. Instrumentation can be used to warn of imminent failure or to indicate whether 

settlement is occurring as predicted. The type of instruments to be used and where they will 

be placed should be planned by a qualified and experienced geotechnical specialist. Actual 

interpretation and analysis of the data should also be performed by someone with a 

background in soil mechanics; however, the project engineer and inspector should 

understand the purpose of each type of instrumentation and how the data are to be used. 

7.9.1.1 Inspector's Visual Observation 

In areas of marginal embankment stability, the inspector should walk the surface of the 

embankment daily looking for any sign of cracking or movement. Hairline cracks often 

develop at the embankment surface just prior to failure. If the inspector discovers any such 

features, all fill operations should cease immediately. All instrumentation should be read 

immediately. The geotechnical specialist should be notified. Subsequent readings will 

indicate when it is safe to resume operations. Unloading by removal of fill material or other 

mitigation methods are sometimes necessary to prevent an embankment failure. 

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7.9.1.2 Types of Instrumentation 

The typical instrumentation specified to monitor foundation performance on projects where 

stability and settlement are critical consists of: 

1. Slope Inclinometers are used to monitor subsurface lateral deformation. A slope 

inclinometer typically consists of a 3 in (75 mm) internal diameter (ID) plastic tube with 

four grooves cut at 90-degree intervals around the inside. The slope inclinometer tube 

is installed in a borehole. The bottom of the slope inclinometer tube must be founded in 

firm soil or rock. A readout probe that fits into the grooves is lowered down the tube 

and angular deflection of the tube is measured. The amount and location of horizontal 

movement in the foundation soil can then be measured. For embankments built over 

very soft subsoils, telescoping inclinometer casing should be used to account for vertical 

consolidation. In soft ground conditions, several inches of lateral movement due to 

squeeze may occur without shear failure as the embankment is built. Therefore, from a 

practical construction control standpoint, the rate of movement rather than the amount is 

the better indicator of imminent failure. Slope inclinometer readings should be made 

often during the critical embankment placement period, daily if fill placement is 

proceeding rapidly, and readings should be plotted immediately on a movement versus 

time plot. Fill operations should cease if a sudden increase in the rate of movement 

occurs. 

2. Piezometers indicate the amount of excess pressure build-up within the water-saturated 

pores of the soil. There are critical levels to which the water pressure in the subsoil will 

increase just prior to failure. The geotechnical specialist can estimate the critical water 

pressure level during design. Normally, the primary function of piezometers during fill 

placement is to warn of failures. Once the embankment placement is complete, the 

piezometers are used to measure the rate of consolidation. There are several different 

types of piezometers. The simplest is the open standpipe type, which is essentially a 

well point with a metal or plastic pipe attached to it. The pipe is extended up through 

the fill in sections as the fill height increases. This type of open well piezometer has the 

disadvantage that the pipes are susceptible to damage if hit by construction equipment. 

Also, the response time of open well piezometers is often too slow in soft clays to warn 

of potential embankment failure. There are several types of remote piezometers that 

eliminate the requirement for extending a pipe up through the fill. The remote units 

consist of a piezometer transducer that is sealed in a borehole with leads carried out 

laterally under the base of the embankment to a readout device that records the pore 

water pressure measured by the transducer. Pneumatic or vibrating wire piezometers 

FHWA NHI-06-088 7 – Approach Roadway Deformations 

Soils and Foundations – Volume I 7 - 50 December 2006
 
have a more rapid response to changes in pore water pressure than open-stand pipe 

piezometers. 

3. Settlement devices are used to measure the amount and rate of settlement of the 

foundation soil due to the load from the embankment. Typically they are installed on or 

just below the existing ground surface before any fill is placed. The simplest settlement 

device is a settlement plate usually a 3 or 4 ft (0.9 to 1.2 m square plywood mat or steel 

plate with a vertical reference rod (usually 3⁄4 in (19 mm) pipe) attached to the plate. 

The reference rods are normally added 4 ft (1.2 m) at a time as the height of the 

embankment increases. The elevation of the top of the reference rod is surveyed 

periodically to measure the foundation settlement. Remote pneumatic settlement 

devices are also available. As with the remote piezometer devices, the remote 

settlement devices have the advantage of not having a reference rod extending up 

through the fill. 

7.9.1.3 Typical Locations for Instruments 

Instrument installations should be spaced approximately 250 to 500 ft (75 to 150 m) along 

the roadway alignment in critical areas. Typical locations of instruments for an embankment 

over soft ground are shown in Figure 7-23: 

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Soils and Foundations – Volume I 7 - 51 December 2006
 

 

 

 

S.I. 

H/4 

H/2 

3⁄4 H H 

S.I. 

Original Ground 

Soft Clay 

Fill 

S.P 

Firm Soil 

Piezometers Settlement Plate (S.P.) S.I. Slope Inclinometer 

Figure 7-23. Typical locations for various types of monitoring instruments for an 

embankment constructed over soft ground. 
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