NOTICE

This document is disseminated under the sponsorship

of the Department of Transportation in the interest

of information exchange. The United States Government

assumes no liability for its contents or use.

The contents of this report reflect the views of

Goldberg-Zoino and Associates, Inc., who are responsible

for the facts and the accuracy of the data presented

herein. The contents do not necessarily reflect the

official views or policy of the Department of Transportation. This report does not constitute a standard,

specification, or regulation.

The United States Government does not endorse products

or manufacturers. Trade or manufacturers’ names

appear herein only because they are considered essential to the object of this document.

FHWA DISTRIBUTION

. J

Sufficient copies of this report are being distributed

by FHWA Bulletin to provide two copies to each regional

office, one copy to each division office, and two

copies to each State highway department. Direct

distribution is being made to the division offices.
Technical Report Documentation Page

1. Report No. 2. Government Accession No.

-HWA-RD-75-,128 I

4. Title and Subtitle

LATERAL SUPPORT SYSTEMS AND UNDERPINNING

Volume I. Design and Construction

7. Author(s) Donald T. Goldberq, Walter E. Jaworski, --

and M. Daniel Gordon

9. Performing Orgonizotion Nome and Address

GoldbergiZoino & Associates, Inc.

30 Tower Road

Newton Upper Falls, Massachusetts 02164

12. Sponsoring Agency Name and Address

Offices of Research and Development

FHWA

U. S. Department of Transportation

Washinqton. D. C. 20590

3. Recipient’s Catalog No.

6. Performing Organirotion Code

8. Performing Organization Report No.

1363

10. Work Unit NO. (TRAIS)

FCP 35 B

11. Contract or Grant No.

DOT-FH-11-8499

13. Type of Report and Period covered

Final report

14. Sponsoring Agency Code

1

15. Supplementary Notes

Contract manager for this study: J. R. Sallberg (HRS-11)

16. Abstract

This volume is a convenient reference on the design and construction of lateral

support sys#tems and underpinning which are often required in conjunction with

cut-and-cover or soft ground tunneling. The design recommendations and

construction methods described herein are a summary of the more detailed

information1 presented in the companion volumes of this study. Included in

this volume are discussions of displacements, lateral earth pressure, ground

water, pass'ive resistance, stability analysis, bearing capacity, soldier

piles, steel sheeting, diaphragm walls, bracing, tiebacks, underpinning,

grouting, and freezing. An overview compares the relative costs of the

construction methods used in lateral support systems and underpinning.

Other reports developed from the study are FHWA-RD-75-129, Volume II, Design

Fundamentals; FHWA-RD-75-130, Volume III, Construction Methods; and FHWA-RD75-131, Concepts for Improved Lateral Support Systems.

17. Key Words 18. Distribution Statement

Bracing, Ground Support, Excavation, No restrictions. This document is availUnderpinning, Cut-and-Cover Construction. able to the public from the National

Technical Information Service, Springfield

Virginia 22161.

19. Security Clorsif. (of this report) 20. Security Classif. (of this page) 21. No. of Pages 22. Price

Unclassified Unclassified 324

Form DOT F 1700.7 (E-72) Reproduction of completed page authorized
PREFACE

This volume summarizes the information presented in Volumes II

and III reporting the results of the study. The purpose of this volume

AS to provide the design engineer and/or contractor with a convenient

reference for everyday use in cut-and-cover tunneling. The basic

design concepts presented in Volume II (Design FundamentaLs) are

presented with a minimum of discussion on the development of these

concepts and design recommendations. The specific design considerations and construction methods for the various wall types, bracing

types, and special techniques are also presented (summary of Volume

III, Construction Methods). Volumes II and III provide greater detail

on development of design recommendations and a more detailed

description of the construction techniques and their performance with

pertinent references listed.

The reports present information gathered from a state-of-the-art

review of Lateral support systems and underpinning. The study was

perfor,med through a contract with the Federal Highway Administration

as part: of their sponsored research program. The voL,umes reporting

the results are designed to aid the practicing engineer and contractor

participating in the design or construction of Lateral support systems

or underpinning.

‘ii
ACKNOWLEDGEMENTS

The authors acknowledge with gratitude: The assistance

of Melvin Febesh of Urban Foundation Company, New York, who

contributed so heavily to the underpinning chapter and whose insight

and construction experience assured proper direction to the effort.

The contributions of Dr. James P. Gould, of Mueser,

Rutledge, Wentworth and Johnston, New York, through his review of

the chapters in Volume I dealing with displacements, earth pressure,

and design aspects of land cofferdams. Thanks also to:

To Mr. John Shuster of Terrafreeze Corporation,

Bethesda, Maryland for the review and contributions for the section

on the techniques of ground freezing.

To Mr. John Jones of Schnabel Foundation, Co., Washington, D. C. for his review and comments on Volume III, “Construction

Methods ‘I.

To Mr. D. Maishman of Foraky, Ltd., Great Britain

for allowing the authors the opportunity to review his yet unpublished

manuscript on ground freezing.

To those many engineers, manufacturers, and contractors who graciously provided comments, data, and photographs.

To Mr. John Dunnicliff of Soil & Rock Instrumentation,

Inc., of Newton Upper Falls, Massachusetts for the chapter on

Construction Monitoring.

To those individuals associated with Goldberg-Zoino

& Associates, Inc. who contributed greatly to the various chapters

in these manuals: Dr. Stephen A. Alsup on Ground Freezing;

Mr. Peter Riordan on Bearing Capacity; Mr. Jackson Ho for the

outstanding effort on computer work; the efforts of Ms. Andrea Wizer

and Mrs. Joan Jennings for their patience and endurance in the

typing of these .manuscripts.

TO Mr. Gardner L. Hayward, Jr. for a job well done

on the drafting of the figures.

And finally, to Mr. J. R. Sallberg, contract manager

for the Department of Transportation and Federal Highway Administration, for his patience, support, and welcome criticisms throughout

the program.

. . .

111
TABLEOFCONTENTS

INTRODUCTORY SECTION

Page Number

TITLE PAGE

PREFACE

ACKNOWLEDGEMENTS

TABLEOFCONTENTS

CONVEIRSION FACTORS

SYMBO:LS

TEXT

CHAPT:ER 1 DISPLACEMENTS

1.10 General

1.20 Characteristics of Wall Deformation

1.30 Magnitude of, Displacements

1.40 Parametric Studies

1.50 Distribution of Deformations

1.60 Lateral Deformations in Adjacent

Soil Mass

1.70 Effect of Construction Procedures

1. 80 Estimating Settlements

CHAPTER 2 GROUND WATER IN OPEN CUTS

2.10 Ground Water Cutoff

2.20 Seepage Pattern to Excavation Face

CHAPTER 3 LATERAL EARTH PRESSURE 3 2

3. 10 Basic Considerations 3 2

3.20 Internally Bra ted Coffer dams 36

3. 30 Tiebacks 42

3.40 Surcharge Loading 44

i

ii

. . .

111

i v

. . .

Vlll

X

Page Number

1

1

1

4

15

20

22

27

28

29

29

30

i v
TABLE OF CONTENTS (don’t1

Page Number

CHAPTE:R 4 PASSIVE RESISTANCE BELOW BASE

OF EXCAVATION

4. 10 General

4.20 Soil Parameters

4. 30 Analysis of Passive Resistance

4.40 Over cut Design Details

4.50 Berms

51

5 1

51

5 3

5 6

5 6

CHAPTE:R 5 DESIGN ASPECTS OF LATERAL PRESSURE 57

5. 10 Analysis of Wales and Support Walls 57

5.20 Bracing and Tiebacks 5 9

5.30 Depth of Penetration Below Cut 6 3

5.40 Example Situations 6 6

5.50 Finite Element Analyses 67

CHAPTER 6 STABILITY ANALYSIS OF SHEETED 7 3

EXCAVATIONS

6. lo General 7 3

6.20 Bottom Heave 7 3

6. 30 Local Failure 7 6

6.40 Deep Seated Failures 7 6

CHAPTER 7 BEARING PRESSURE OF DEEP FOUNDATIONS 8 3

7. 10 General 8 3

7.20 Presumptive Bearing Value 8 3

7. 30 Bearing Values Based on Shear Failure 8 3

7.40 Bearing Values Based on Sett.Lement 8 7

CHAPTElR 8 OVERVIEW OF CONSTRUCTION METHODS

8. 10 Purpose and Scope

8.20 General Conclusions Concerning

Displacements

8. 30 Wall Type

8.40 Support Method

8. 50 Underpinning

8.60 Stabilization Methods

8. 70 SoiL and Ground Water Conditions

8. 80 costs

98

98

98

100

101

103

104

105

106

CHAPTER 9 SOLDIER PILE WALLS 110

9.10 Introduction 110

9.20 Types of Soldier Pile Walls 110

9. 30 Design Considerations 113

9.40 Construction Considerations 124

V
TABLE OF CONTENTS (con’t)

Page Number

CHAPTER 10 STEEL SHEET PILING 128

10.10 Introduction 128

10.20 Design Considerations 128

10.30 Construction Considerations 133

CHAPTER 11 CONCRETE DIAPHRAGM WALLS

11.10 Introduction

11.20 Properties of Bentonite Slurry

11.30 SLur r y Tr en ch Stability

11.40 Practical Aspects of Slurry

Stabilized Excavations

11.50 Structural Aspects of Cast-in-PLace

Walls

11.60 Excavation of Slurry Trenches

11.70 Diaphragm Walls Other Than

Continuous Cast-in-Place Concrete

136

136

136

142

145

151

153

159

CHAPTER 12 INTERNAL BRACING

12.10 Introduction

12.20 Design Considerations

12.30 Ins ta.llation

12.40 Temperature Effects

12.50 Strut Removal and Rebra cing

CHAPTER 13 TIEBACKS

13.10 Introduction

13.20 Design and Theoretical Considerations

13.30 Construction Consider ations for

Tiebacks

13.40 Field Testing

169

169

169

173

176

177

178

178

178

218

221

CHAPTER 1 4 UNDERPINNING 240

14.10 Introduction 240

14.20 Design and Theoretical Considerations 240

14.30 CLassical Underpinning Procedures 242

14.40 Grouted Piles 256

14.50 Tunneling Below Structures 258

14.60 Load Transfer 263

14.70 Temporary Support or “Shoring” 268

14.80 Performance 270

v i
TABLE OF CONTENTS (con’t)

Page Number

CHAPTER 15 GROUTING 276

15.10 Introduction 276

15.20 Design and Theoretical Considerations 276

15.30 Construction Considerations 287

15.40 Field Testing and Quality Control 293

CHAPTER 1 6 GROUND FREEZING 295

16.10 Introduction 295

16.20 Design and Theoretical Considerations 295

16.30 Construction Procedure 307

vii
LIST OF CONVERSIONS

The List of conversions is designed to aid in converting from

British units of measure to metric units. This section has been divided

into two parts; general notation and arithmetic conversion.

General Notation

BTU

cm

2

cm

3

cm , cc

cfs

ft

ft2

ft3

fPs

gal

gpm

g, gr

h r

in

XII2 .

i n

3

k

kg

m

2

m

3

min

British Thermal Unit

centimeter

square centimeter

cubic centimeter

cubic feet per second

feet

square feet

cubic feet

feet per second

gallon

gallons per minute

grams

hour

in the s

square inches

cubic inches

kilo (thousand)

kil0gra.m

meters

square meters

cubic meters

minute

. . . Vlll
m m

2

m m

3

m m

ml

N

Lbs

I?Cf

Plf

Ps f

psi

set

British IJnits

1 BTU

lin

1 in2

1 in3

1 ft

1 ft2

1 ft3

1 pcf (lbs/ft3)

1 psf (Lb&t’)

1 ksf (kips/ft’)

1 psi (lbs /in’)

1 l b

lin-lb

millimeters

square millimeters

cubic millimeters

milliliters

Newton

pounds

pounds per cubic foot

pounds per lineal foot

pounds per square foot

pounds per square inch

second

Conversions

Metric Units

0.2520 kg - calories

107.5 kg - meters

2.540 cm = 25.4 mm

6.452 cm2

16.103 cm3

30.48 cm = 0.3048 m

929 cm

2

= 0, 0929 m2

28,317 cm3 = 0.0283 m3

16.02 kg/m3 = 0.01602 g/cm3

4.883 kg/m2 = 47.9N/m2

4.45 N

0.1127 N-m

i x
List Of Svmbols

The following list of symbols has been prepared to aid the interpretation of symbol use in the text. This list identifies only the ,major

symbols used in the text and their general meaning. Each symbol (with

subscripts) is defined in the text for its particular usage. This list is

not a complete List of all symbols or all symbol usage. in the text but

is a summary of major symbols and their usage.

OLsymb Represents

A general symbol for area

B, b general symbols for width

C cohesion intercept

c heat capacity

D, d general symbols for distance

and diameter

E

f

F. s.

H

K

KO

general symbol for modulus

general symbol for stress

factor of safety

depth of excavation; also

genera.L symbol for height

genera.L symbol for coefficient

of lateral earth pressure

coefficient of lateral earth pressure

at rest

Ka

K

KP

J-J, I.

coefficient of active earth pressure

coefficient of passive earth pressure,

thermal conductivity Volume I, Chapter 16

Vo.Lume III, Chapter 9

general symbols for Length

or distance

N general symbol for stability

number or standard penetration

resistance

OCR over consolidation ratio

Reference

Volume I, Chapter 16

Volume III, Chapter 9

X
Symb o 1. Reoresents Reference

P

P

PH

R, r

s, s

S

U

8,(max)

6, (max:)

E

ll

8m

IIsub

VW

P

V

Q

general symbol for Load or force

general symbol for pressure

ne gative logarithm of effe ctive

hydrogen ion concentration

general symbols for radius

genera.1 symbols for shear resistance

or shear strength

undrained shear strength

pore pressure

general symbol for weight

general symbo.1 for water content

general symbo.1 for displacement

or movement; also angle of wall

friction

vertical displacement (maximum)

horizontal displacement (maximum)

general symbol for strain

general symbol for unit weight;

total unit weight of soil unless

other wise specified

dry unit weight of soil

total unit weight of soil

bouyant unit weight of soil

unit weight of water

Pois son’s Ratio

Pois son’s Ratio

general symbol for friction

angle of soil

xi
bvm

t?

Represents

general symbol for settlement

general symbol for stress

tota. vertical stress

(effective vertical stress)

total horizontal stress

(effective horizontal stress)

maximum past vertical.

consolidation pr e s sure

(effective stress)

general symbol for shear stress

or shear resistance

Reference

Note: Line over symbols indicate’s effective stress para.meters are to

be used. (e. g. cv = vertical effective stress).

xii
CHAPTER 1 - DISPLACEMENTS

1.10 GENERAL

The purpose of this section is to provide insight into displacements

occurring adjacent to deep excavations -- specifically, into those factors

influencing displacements and into the manner in which displacements

occur.

W&le the magnitude of settlement is a useful indicator of potential

damage to structures, the amount of settlement change with horizontal

distance (settlement profile) is actually of greater significance.

Horizontal displacements have proven to be a source of severe

damage, even in the presence of underpinned structures.

1.20 CHARACTERISTICS OF WALL DEFORMATION

s!neral Mode of Deformations

Figure 1 shows the possible range of deformations for perfectly

rigid walls and for walls displaying flexure. Basically, the range of

behavior includes translation and either rotation about the base or rotation

about the top. In addition, wall deformation will include some bulging

as a result of fiexure -- the amount of bulging depending upon the stiffness

of the wal’l support system.

-

Internally Braced Walls

The upper portion of the internally braced wall is restrained from

undergoing large horizontal movement especially when braces are prestressed and are installed at or close to the surface. This produces the

typical deformation mode as shown in Figure 2a. The degree of rotation

will depend upon the toe restraint below the bottom of the excavation.

Tied-Back Walls

If the top of the tied-back wall remains fixed, then the deformation

mode is similar to that of an internally braced wall (see Figure 2b, left

panel). On the other hand, settlement of the wall, partial yielding of the

ties, gross movement of the soil mass, or shear deformation of the soil

mass may result in inward movement of the top and rotation about the

bottom as shown in Figure 2b, right panel.

-l-
(a) INFINITELY RIGID WALLS

I

i

I

I

I

I

I

+-I

I

l-RANSLATION ROTATION

ABOUT BOTTOM

ROTATION

ABOUT TOP

FIXED

(b) WALLS DtSPLAYING FLEXURE

TRANSLATION ROTATION

ABOUT BOTTOM

/

I

/ /

I

I

i i

I

I

I

\

\

_ ROTATION

ABOUT TOP

FIXED

Figure 1. General deformation modes.

-2-
(a) TYPICAL FOR INTERNAL BRACING

I

I

I

1

I

I

\

/

FIXED OR SLIGHT

TRANSLATION

ROTATION

ABOUT TOP

(b) TYPICAL FOR TIEBACKS I

I

1

\

\\

FIXED OR SLIGHT

TRANSLATION

ROTATION

ABOUT BOTTOM

Figure, 2. Typical deformation of tied- back

and internally braced walls.

-3-
Comparison of Braced Walls with Tied-Back Walls

There are insufficient data for a meaningful comparison of deformations of internally braced walls and tied-back walls. In competent

soils (e. g. granular deposits, dense cohesive sands, very stiff or hard

clays, etc#, ) displacements are small and no significant differentiation can

be made between tiebacks and bracing.

A number of factors indicate that a superior performance should

be attainedl with tiebacks in competent soils:

1. In granular soils in which soil modulus increases with stress

level, the ,prestressed soil mass, engaged by the tiebacks, is made more

rigid and therefore less deformable.

2. Tiebacks are typically prestressed to about 125 percent of the

design load and then locked-off between 75 percent and 100 percent of

the design load. Prestressing in this manner prestrains and stiffens the

soil and pu.lls the wall back toward the soil to remove any “slack” in the

contact zone.

3, Internal bracing, if prestressed, is usually to about 50 percent

of the design load. Typically, the. bracing gains in load as the excavation

deepens. :Elastic shortening of the strut continues after installation of

the member.

4. Temperature strains are more important with bracing than

with tiebac’ks because the latter are insulated in the ground.

5. Internal bracing is removed then rebraced to facilitate

construction, whereas tiebacks do not have to be removed.

1.30 MAGNITUDE OF DISPLACEMENTS

1.31 Reported Horizontal and Vertical Displacements

The summary of data in Figures 3 and 4 is an extension

of a similar procedure presented by D’Appolonia (1971).* The figures

show normalized vertical and horizontal displacements (ratio of the maximum displacements to the height of the cut) versus three general

categories of soil type and support type, The corresponding references are

listed in T,able 1. Diaphragm walls are distinguished from the

relatively more flexible soldier pile or sheet pile walls by symbol.

* Complete references are given in the Bibliography contained in Volumes

II and III of this report.

-4-

0.5

I.5

2.0

VERY STIFF TO HARD SOFT TO STIFF

SAND AND GRAVEL CLAY CLAY OTHER SOIL

SIJ~ 2000 psf Su’2000 psf CONDITIONS

TIEBACKS PRESTR. BRACING

TIE BACKS PRESTR. BRACINB PRESTR. BRACING PRESTR. BRACING

BRACING ( STAND.) BRACING (STAND.)

TIEBACKS BRAGINO (sTANn) TIE BACKS BRAGINO (sTANDJ

047 04w 041

l 28 025 OS 042

~32 1

ë.* 03

I l 30052 ~2~~uL2~~~ W I 1 I

00 ou

048 lJji;;AD,*wRm” 1 I

I I

.I3 044 036

t I

*I6

----- WALLS 033

l 4

024 l 31 I

(oRK%

I I

I I I I I I

034 I

-----

I

x I l 23b

.I?! I

I

I I

950

I

w 21 I I

oae* (OR;%&

SOILS)

I

I

I I

I 011 II

I

I I I I

035 I022

I

.I4 I

*

(ORi:NC I

m--m75% OF CASES EXPERIENCED LESS l 1 1

SILT)

MAXIMUM MOVEMENT. STEEL (SOFT CLP

WALLS 061

----- 100% OF CASES EXPERIENCED LESS

MAXIMUM MOVEMENT.

I

I- OS7

r,’

/ /ANDZKT

+ ATYPICAL OR UNUSUAL CASES. REFER

I

STEEL

TO THE TEXT AND TABLE I. WALL I I

NOTE: NUMBERS REFER TO REFERENCES

IN TABLE I. I

I

059

I 1

l + 5 % - DISPLACEMENT OFF SCALE, MAGNITUDE

EQUAL TO 5% I

I

b&2, I I

I

m--e

I I I I I I I I

I

Figure 4. Normalized horizontal movements.

O= DIAPHRAGM WALLS

. = SOLDIER PILES OR

STEEL SHEETING
Table 1. Summary of references on displacement.

Ref. k Author(sI

I Lambc, Wolfsk,ll, bbk’ b! rut, , Fill, Organic 5X’ 7” 9” Consolidatron settlements significant.

Ri Wang (I9iUl (I’r~.str<~ss~~rl) Slit, till, rock (17. 7m) (17. Xcml (22.9cm) brtll<!n,cnts of 3” (7. ocm, up to 70’

(ZI.3ml from excavation.

2 O’Rourke and SP h, 1‘11, * Ihnsc Sand and 60’ I. 5” * 9” Removal of struts increased settleCording (19i4I (I’t~~st~csscrll pravcl, Stiff (IX. 3m) (3. Hcrn) (L3cm) IIIC”, from 0. 9”(%. 3cnt) to I. 5’

IC St. Excavatton) clay (3.Xcm).

3 O’Rourkc and SP strut* D<:nsc Sand and X2’ I. 5” I. 25” Some time-dependent conaolidatlon

Cording (19741 (Prcstrcsscd) gravel, Stiff (25m) (3. km) (3. Z~tnl sc1t1cnrcnts.

(7th L G Streets) clay

4 O’Rourkc and 5P Ticbacks Dense Sand and 40’ .7” 2” Strc,ct scttlcments small \b,hilc

Cording (1974) and gravel, Stiff (lZ.dm) (1.8cml (5. Icm) solrlicr piles settled dua to do~“nWare. Mirsky, clay 2” drag Iron) tiebacks. Soldier piles

and Leuniz (19731 (5. Icm) settled 2”(5. Icmj maximum.

4th & C Strcctsl

5 Lambe, Wolfskill D W S,ruts Fill, hard to 50’ 1 ” I. 2” Minor consolidation settlements.

and Jaworski (Prcstrcssetl) medium clay, (15. Lm) (Z.SCrn) (,.“c 11,) S< hool located 5’( I. 5m) from wall.

(19721 till

6 Durland (1974) and D W Struts (Slabs Gravel and 52’ .6” I”

St. John (19741 poured as very stiff (15.9m) (1.5cm) (2.5Ll,,)

(NW Palace Car cucavation clay

PrrkI “TO< <~PdCd,

Burland (1974) and

St. John (1974)

D W Tlcbacks Very Stiff clay 26’ I. 1” 2. L” Murh of the wall movcmcnt was pure

(7.9ml (2.8cm) (5. (Kllll translation dnrl continued with time.

Evtrr~nvly small “crtical settlements

c\cept directly behind the xvall.

SP struts Dense Sand and --

(Prestrrsscrl) prnvel and stiff

Cl.XY

. 2% _- Did not report depth of excavation or

amount of settlement.

Durland (1974) and

St. John (1974)

(London YMCA)

D W Slabs and Crawl and 52’ .5” . 6”

Tiebacks “cry stiff (15.9m) (I. 3cm) (I. 5rm)

clav

I O N.C.1. (1962) SSP struts Soft to medium 19. 5’ 3” _- Consolidation ecttlements due to

(Oslo Technical clay (5. 9m) (7. Ocm) Irw~~ring of head in underlying

School) Sdlld.

II N. G. 1. (1962) SSP St ruts Soft to medium 36’ x. 9” 5. I10 Nearby undsrpinncd structure

(Vatcrland #2) (Prestressedl clay (llrnl (22. 6cm) (I 3cml scttlcd significantly.

1 2 McRostie. Burn SSP ‘Tirbacks Medium to 40’ 4.5” -4” Excessive txback prestrcssing

and M&hell stiff Clay (12.2m) (11.4cm)(-lU.Zcm) pulled wall away from excavation.

(1972) Scnsitlvc clay consolidated due to

shuarinx stresses.

13 D~Biagllo and

Roti (1’372)

See Sheet 5 for notes.

D W Floor slabs Medium clay 62’ I. 6” l-1.2” str\,rt,,rc ( 2’(0. km) Iron1 wall.

U%-tt to (18.9ml (4. Lcml(Z.S-3.0cml All scttlcmcnt appeared to be due

support wall to lateral wall dcfloction.

Sheet 1 of 5

-7-
Table 1. Summary of references on dk.placement. (Continued. 1

Shannon and

s&raze* (1970)

SP Ti”backs Very stiff clay 78’ 3” 3” Mssirnurn settlement mcasurcd at

and sand (23. Em) (7.6cm) (7. hem) \v;ti11. Scttlcnrent may bc due to

downward force exerted by ticbacks.

Swatek, Asrow,

and Seitz (1972)

SSP struts Soft to stiff 70’ 9” 2.3” Largr scttlcment attributed to

(Prestressed) clay (21.4m) (22.9cm) (5. Rem) locallzcd heavy truck traffic.

Typically settlements < 5”( 12. 7cm).

Rodriquez and

Flamand (19691

SSP St*UtS Soft to medium 37’ _- 7.9” Stag”d construction to minimize

(Prestresscd) clay (11.3ml (20. ICIll) n~o~rmcnte. Dewatered to orcvent

I8 Scott, Wilson and SSP St*UlE Dense fine 50’ _- 8” I’oor porfurn>ance attributcrl to

Bauer (1972) sands (15.3d (20. 3cm) poor construction techniques and

dewatcring problems. Nearby

structures damaged.

I9 Chapman, Cording SP Struts and Sand and 451 . 25” I ” Running soil “ncountcred in one

and Schnabel (1972) Rakrrs gravel and (13. 8m) (0.64cm) (2. 5cm) section.

(Pr”strcsscd) stiff clay

20 Boutsma and SSP struts Soft clay and 33’ 14” 6” Sornc scttlemcnt du” to extensive

Horvat (1969) soft peat (IO. Imj (35.6cml (15.2cm) dcwatcring for long time pcr~od.

AlfV< 11:d struc,urus 600’ I,““,

vxravation. Liquefaction “f backfill during uxtract~on.

21 In&y (1972) SkJ Rakers soft to mod,um 25’ _ _ 2.5” vne SCCtl”” tcstcrt to fsilurc.

clay (7. bm) (6. 4cm)

22 Tait and Taylor, SSP Struts and Soft to 45’ 6” 7.5” 1,~ rk<‘r n~“vc~n~~nts dttributcd to

(19741 Rakers medium (13.8m) (15.2cm) (19. ICI,,) Ia< k “i firm hottow for wall.

(Prestrcssed) clay Utility lines dan,?gcd; no major

danra,w lo ad>.lccnt s,ructu,Bs.

23a Hansb”, Hoiman, SSP Rakers sdrt clay 23’ 13.M” I I, 8” P”or sheet p:lv interlocking. Long

and Moseeson (7. Om) (35. Icm) (29.9~11) tit,w bctnvcn excavation of vcnt<,r

(1973) portion and bracing. D~sturbancc

rlurins ~11” driving for foundation.

23b Hansbo. Hoiman,

and Mosesson

(19731

SSP TIcbacks and Soft clay 23’ 2” 2” Improved construction techniques.

Rakers (7. Om) (5. lcm) (5. Icm)

24 Prasad, Freeman, SP Ticbacks Very stiff 45’ _- -2” ‘1 cap of wall mowd anay from

and Kla,nerman Clay (13.8m) (-5. Ic,n, L’\C.l”atl”“. Maximum movcmcnt

(1972) at top.

25 Mans”, and SP Tiebacks Very stiff to 45’ .5” .5”

Alizadeh I 1 970) hard clav 113.8ml ll.3cml ll.3cml

26 Sandqvist (1972.1 SSP Tirbacks Sand and silt 19.5’ 7.9” 2” b<~ttlcnwnt in “rpanics tlur to

with organic (5.9m) (20. Icm) (5. Icm) Iubc~ *cd ground water lcvcl. I’ilc

soils driving also caus~tl settlement.

See Sheet 5 for notes. Sheut 2 of 5

-8-
Table 1. Summary of references on displacement. (Continued. 1

Ref. % Author(s)

Wall

’ Type

Bracing

TYPO

hpth d

Soil Type

” ‘h of Cut max n,ax Comments

27 Sigourney (1971) SP Tiebacks Clayey sand 20-26’ -- ( 5”

and hard clay (6. l-7. 9m) (1.3cm)

28 Coettle, Flaig. SP Tiebacks Dense sand 25” 25”

Miller. and (7fin, (01 h4cm~ (O.‘h4cm)

Structure with footings only 2’

and grave1

bchaeier (1974)

(U.SZrn) from wall was undamaged.

29 Sigourney (1971) SP Tiebacks Very dense 35-43’ -- . I”

silty sand (IO. 7- (0. 25cm)

and gravel 13.2m)

30 Clough, Weber, and SP Tiebacks Very stiff 64’ I. 25”+ I “i

Lament (1972)

Top of wall moved away from

day (19. 6m) (3. 2,:) (2. SC;,) excavation.

31 Nelson ( 1973) SP TIebacks bandy over- 90’ 1 ” 4 ” Cracking in street indicated patcnburden, hard (27.5m) (2.5cm) (IO. Lcm) tial stability failure (i~‘h~ax[ 15. Zcm])

clay shales Mallian L Van Beveren (1974).

32 Liu and Dugan SP Tiebacks Dense sand and 55’ , 8”l I”/

(1972)

Tops of soldier piles pulled away

gravel. very (16. Em) (2. Ocrnl (2.5<E+rnl from excavation during prcstressstiff clay ing.

33 Larson, Willette, SP Tiebacks Dense sand 50’ 1 ” I 10 Hall, ancl (15.2m) (2. 5cm) (2.5rml

Gnaedmger (1972)

34 Dietrich, Chase, SP Tiebacks Silty sand 23-54’ 2.5” 1.X” l.ate~ral “l”Yerne”tS measured at

and Teul (1971) (7- 16. 5m)(6. 3cm) (4. bcmt top of wall.

35 Cunningham and D W Tiebacks Medium clay 23’

(7.Om) --

4” ‘TIcbarks anchored to deadman.

Fernandez (1972) under dense (IO. 2cm)

sand

36 Cole and Burland D W Rakers Very stiff 60’ I. 5” 2. 5’1 hlost movements occurred while ear0

(19721 clay (18.4114 (3.&m) (6.3cm) berm supported wall. Excavation in

heavily ovurconsolldated clay.

37 Tait and ‘Taylor D W Tiebacks, Medium and 45’ - -

, 9“

(1974)

M,nor settlements of nearby

prestressed soft clay (13.8m) (2.3cm) structures

Struts and

rakers

38 Armento (1973) D W struts Sand and soft 70’ 1. 7” I” borne settlement may have beon

(Prestressed) to medium (21.4m) (4.3cm) (2.5cm) <dusod by other excavattons in

clay the at-e=.

39 Cunningham and D W struts Suit and 32’ 5.5” 3.5”

Fernandez (1972)

Undcrpinninp of ncarby footinKs

medium (9.8ml (13.9cml (8.9cm) rc,quired after 5. 5“(13. 9cm) of

Clay scttlan,ent. 50.700; of “lovemcllt

during caisson construction.

_ _

40 Tan (1973) D W lhsemcnt soit clay 43’ 6”t _ _ S<,ttlcnwnt cslitnatcd on basis of

slab as (13.2m) (15.2cm) substantial damage to structure

support ‘lO’(I2. Lm) irun excavat,on.

See Sheet 5 for notes. Sheet 3 of 5

-9-
Talble 1. Summary of references on displacement. (Continued. )

Ref.X Author(s)

Wall Bracing Depth d

TYPO TYPO Soil Type of cut “max

dh

max Comments I

Thon and Harlan

Barla and Mascardi

DW Str”tS Soft to medium 78’ 1 ” 1.2”

(Prestressed, clay (23. Em) (2. 5cm, (3. Ocml

S W Tiebacks Stiff clay

(2: 9m,

-_ 2. 6” Cracking in nearby structures.

(6. bcm,

Heeb. Schurr, Bone, SP struts Sand _ _ . 8”

Henke, and Muller (2.0cm,

46 Breth and Romberg SP Tiebacks Stiff clay _ _ 5.9” Lateral movement of entire

(L972,, Rornberg and sand (14.9cml soil block.

(1973,

47 Schwarz (1972) and S W Tiebacks Clayey marl 97.5’ . 2” . 6” Many levels of tiebacks at very

Andra, Kunzl, and (stiff clay, (29.8ml (0.5 lcrn, (I. 5cmt close spacing.

Rojek (1973)

48 Corbett, Davies, D W Hakers Very stiff clay; -- -_ 2” Construction delayed after

and Langiord upper sand (0.51Crnl hole opened.

(1974, and gravel

49 Hodgson (1974, D W Ticbacks Fill, gravel 26’ _ _ I‘?” Spr,cial construction procedure

and struts very stiff clay (7.9ml (0. 3ctn) “SCd.

50 Corbett and SP Tiebacks Fill, sand - - 8” Heave observed 18m from *all.

Strau” (1974, and marl (2. ncm)

51 Littlejohn and D W Tiebacks Gravel and

MacFarlane (1974, very stiff clay (5t85*, - -

. 8”

(2. ocm,

52 Littlejohn and D W Tic backs Gravel and

MacFarlane (1974, very stiff clay (1: 4m) (;.‘;:m,

. 9”

(2.3cm,

53 Saxena (1974, D W Tiebacks Organic Silt ” __

and sand ( 156: &II,

2. 7” Tops of some wall sections moved

(6.9cml toward soil by same amount.

54 Ware (1974, D W struts Sand and _- I. 25”

Personal (Prestressed, gravel and l3.2cml

communication stiff clay

55 Goldberg-Zoino SP Tiebacks Fill, organic 45’ 1. 5” 1 ” Vertical scttlemcnts due to lagging

& Assoc. Files sand, stiff (13.8m) (3.8cm) (2. 5cml installation. Most horizontal

clay, till movement away from excavation.

56 Burland (1974, D W C.tntilcver Very stiff . 5”

and St. John Wall clay (I. 3cml

(1974, merits

See Sheet 5 for notes. Sheet 4 of 5

-lO-
Table 1. Summary of references on displacement. (Continued. )

ICI. t l Author(s)

Wall

Typo

Bracmg

TYPO

Depth J

Soil Type of Cut “max

c f

‘I,, lax Col,,nwnta

57 N. C. I. (19h2) bSP SLruts Medium and 26’ 3.9” 5.5” bipnificant ~1~~~w~nwnls alter

Telecommunications (Prestresscd) soft clay (7.9ml (9.9cmt (13.9cn,l SII’“! rcnluval,

CC”kI

5 8 N.C.I. (1962) bSP struts Medium and 26’ 4.2” 21’ L.ateral dcflrclions p*-obably

Enerhau,;en South (Prcstressed) soft clay (7.9ml (IO. 7crn) (5. ICI,,) ,,,ore than shonn.

5 9 N. G. 1. (19h2) bSP StlWtS hlcdium and 360 7. 9” ‘),L

Vatcrland I I 1 soft clay (I l.Orn) (20. IClll) (22.9‘,,1)

6 0 N.G.1, (1962) SSP Slabs as Medium to 371 7.5” -- Air prcssut’e and upside donn

Grlnland # 1 suppurt soft clay (I 1.3nl) (I9.OCrn) construction ~b~l.1.

6 I N.G.I. (1902t SSP Struts Medium and 3U’ 3.9” 5. 9”

Vaterland H 3 SOL< clay (9. Znr) (9.9cmb (14.9crnl

62 MalJian and SP Tiebacks bliff to very 110’ 3” 2” hlasimutn vcrti~al sottletn~~nt

Van Bevoren stiff clay and (33. 6m) (7. 6cml (5. Icrn) .glypical for the situ--uoually

(1974) cohcsivs santl lateral ,,,ov~nwnt fi!rua,r:r than

0 3 Jennings (rascs Tiebacks F&i-111 4X’ - - 3” I).~nng~, to uti1illt.r in strcut

reported by (I4.7mI (i. o< “1, and building across street.

Lzrtlu,ohn and --- ------ -----

MacFarl.and [ 197411 Fissured 4s’ - - 1.5” Acccotable n,ovenxents

South Africa (14. 7m) (3. Hrrnl

--- -_--

Clay 14’ - - ,l.S”

(22. IbIn) (3. XC”>)

-a-------_-

Very sttff ‘461 _- .15”

fissured clay (14.7m) (1.9clnl

--- --- A--_-

Soft Jointed 59’ _- I ‘1

rock (18.Oml (2. 5Lrn)

Notes:

Sheet 5 of 5

1. SSP : Stool shoot pihng

SP TSoldier pile nail

DW : Diaphragm wall

SW = Secant wall

2. cIh and 6” are maximum horizontal and vertical displacements.

3. Reference X represents rcfercnccs lislcd by author in Bibliography.

-ll-
Vertical and horizontal displacements in the ground outside

the excavation arise from:

1. Horizontal and vertical displacement of the wall - - in

general, this is rotation, translation, and flexure.

2. Movement of soil -- for example, loss of soil through

lagging, overcutting and improperly backpacking of lagging, spalling of

slurry trench walls, voids created from pulling of sheeting, etc. (See

Volume III, Construction Methods, for more detailed discussion under

various te&nique s) .

3. Consolidation of soil -- for example, densification of

loose granular soils from vibration or consolidation of soft cohesive

soils from lowering of ground water outside the excavation.

4. Base instability or near instability -- excessive shear

strains set up by the imbalance created by removal of removal of load

contribute to base heave and/or plastic conditions in soil.

5. Stress relief from excavation -- this reduces vertical

stress below the base and relieves the K horizontal stress (earth

pressure at rest). In turn, the possible &odes are base heave, shear

strains, and lateral strains.

The performance data indicate the following.

1. Sand and Gravel; Very Stiff to Hard Clay

Seventy-five percent of the excavations in this material

experienced horizontal movements less than 0.35 percent of the excavation depth. Generally, the performance is not significantly affected by

support method or by wall type.

One probable reason for little apparent difference

between wall type and support method is the fact that the measured

displacements are small (typically less than 0. 10 feet for a 50-foot

excavation). Many construction factors can contribute to displacement

variation o.f similar magnitude and therefore would mask such variation.

Two anomalous cases (no. 7 and no. 46, Table 1)

reveal a potential source of extraordinary lateral movement of a tied-back

wall retaining primarily very stiff or hard clays, Wgrd (1972) cites

horizontal strains as two to three times as large as vertical strains in

overconsolidated London clay.

-12-
2. Soft to Stiff Clay

Wide variations for both horizontal and vertical displacements are evident. Sixty-five percent of the cases experienced horizontal displacements which exceeded 1 percent for steel sheet pile or

soldier pile walls, whether prestressed or not.

The largest benefit is derived from concrete diaphragm

walls with prestressed bracing. Indeed, both horizontal and vertical

displacements are no different from those typical for sands and very

stiff to ha.rd clays, being about 0.25 percent or less.

Another major cause of settlements in cohesive soils is

lowering of the ground water table.

1.32 Effect of Wall Stiffness on Lateral Displacements in Clay

Wall stiffness refers not only to the structural elements

comprising the wall but also to the vertical spacing between the support

members. The measure of wall stiffness is defined as the inverse of

Rowe’s flexibility number for walls EI .

73

where:

E = modulus of elasticity of wall

I = moment of inertia/foot of wall

L = vertical distance between support levels or

between support level and excavation base

A plot of observed displacements for stability number

y H:k

(N = ‘s,- ) and stiffness factor ($$) is developed on Figure 5. The

stability number , which considers both overburden stress ( 8 H) and the

undrained shear strength (Su), is a measure of the relative strength

or deformtability of the soil.

The data demonstrate what is intuitively obvious -- that

deformations are functions of soil strength and wall stiffness. The

contour lines of maximum lateral wall movements show this trend clearly.

These dat’a allow one to examine qualitatively the relative change in

anticipated lateral displacement for a given change in wall stiffness and/

or stability number of the soil.

*Ratio of overburden stress to undrained shear strength.

-13-
-14-
L* 33 Wall Movement Versus Settlement

Comparison for all Cases

Figure 6 compares observed maximum horizontal

and vertical displacements for all types of soils, support systems, and

wall types. The absolute magnitude is shown in panel (a) and the frequency

distribution of the ratio of the movements in panel (b). The figure shows

that practically all the vertical displacements fall within a range of

l/2 to l-1/2 times the horizontal displacements, with most of them lying

in the range of 2/3 to 1- l/3 times the horizontal movement.

Soft to Medium Clay

Figure 7 compares displacements for soft to medium

clays. The average curve shows that the vertical displacements are

generally well in excess of the horizontal displacements and that the

range of displacements increases with the magnitude of the displacements.

The difference is believed to be directly attributable

to consolidation settlements which are usually the result of changes in

water levels adjacent to the excavation.

Very Stiff to Hard Clays

Figure 8 compares the displacements of these soils.

As mentioned in a previous discussion, comparatively large lateral

displacements have been reported in several tieback projects.

1.40 PARAMETRIC STUDIES

The results of a finite element study for evaluating the effect of

wall stiffness on reducing deformations in various soil conditions is

shown in Figure 9. Also shown for comparison are the lines defining

deformation limits from Figure 5.

The finite element computer program used to develop thesedata

considered only cohesive soils and internally braced excavations; See

Volume IK (Design Fundamentals) for soil properties and methods of analysis,

Figure 9 shows that the predicted lateral displacements are less

than the observed values for a given condition. This difference is

related to the inherent movements which are a function of the costruction

process. Nonetheless, the theoretical results show a trend similar to

that described by the field observations; that is, the stiffer walls result

in lower movements for a given soil condition.

-15 -
T 026

0

//

*012

1

50 0%6, /

0 SOFT TO MEDIUM CLAY (SOME STIFF CLAY) ~~~~~~~T

2 4 6 8 IO 12

sh max. inches

(a)

MAXIMUM VERTICAL AND HORIZONTAL

DEFORMATIONS

I I

) 0.4 0.6 1.2 1.6 2.0

&nox

sv mox

( b)

FREQUENCY DISTRIBUTION OF RATIO

OF MOVEMENTS

NOTE: NUMBERS REFER TO CASE STUDIES

LISTED IN TABLE I.

Figure 6. Comparison of maximum

vertical and horizontal displacements,
1 4

12

IO

is

.-

-

::E

s

6

4

2

AVERAGE FOR

./ SOFT TO MEDJ

+ SOFT TO MEDIUM CLAYS

FOR OTHER SYMBOLS, SEE FIGURE 6

4 12

NOTE: NUMBERS REFER TO CASE

STUDIES LISTED IN TABLE I.

Figure, 7. Comparison of vertical and horizontal

displacements for soft to medium clays,

-17-
+ VERY STIFF TO HARD CLAYS

*

+31

FOR OTHER SYMBOLS, SEE FIGURE 6

2 4 6 8 IO i 2

6 h mm. inches

NOTE: NUMBERS REFER TO CASE

STUDIES LISTED IN TABLE I.

Figure 8. Comparison of vertical and horizontal

displacements for very stiff to hard clays.

-18-
-

-

0

-

-

7

I

- -/--

- FLEXIBLE El

/

KSF/FT

I*

4

1

-

-

-

-

-

7

I

,I

-

-

-

STIFF .-

b

--- FINITE ELEMENT ANALYSIS

Figure 9. OBSERVED DATA Comparison of lateral movements from finite element

analysis and observed movements.
The results of the finite element analysis should not be taken in

the quantitative sense. The intent is that such an analysis should be used

as a guide in the design and in the consideration of various options for

a bracing system.

1.50 DISTRIBUTION OF DEFORMATIONS

Currently, many engineers rely on judgment and experience in

predicting deformation patterns adjacent to sheeted excavations. This

section provides some information to aid the engineer in evaluating what

deformation patterns might be expected adjacent to a cofferdam.

1.51 Vertical Deformations

Figure 10 illustrates how the observed maximum settlement

patterns behind a wall varies with the soil conditions. The pattern of

movements indicates that maximum movements occur immediately

adjacent to the excavation. Also, one might expect significant movements

a distance from the cut equal to twice the depth of the cut. At present,

there are insufficient data to define any significant difference in settlement

pattern based on soil type or support wall.

Comparing the settlement patterns of sand versus cohesive

soil, the sands show essentially no settlement beyond twice the depth

of the excavation whereas the cohesive soils do. This is most likely

attributable to the consolidation experienced in the more compressible

soils caused by lowering of the ground water table.

Reviewing Figure 10, it appears that both soft clays

and the granular soils experience a significant agnular distortion

outside a distance equal to the excavation depth (D/H = 1). The average

lines of settlement ratio versus normalized distance, shown as dashed on

the figure, may be used as a basis of comparison of this distortion.

On the other hand, the stiffer clays (S> 2000 psf) seem to experience a more

gentle distortion slope, even though the zone of influence extends further

back from the excavation face.

1.52 Parametric Study On Zone Of Influence

Finite element studies were performed on several of

the deformation modes shown in Figure 1. These analyses were aimed

at obtaining some qualitative information on the settlement profile one

might expect adjacent to the excavation.

Details of the finite element analysis are given in

Volume II ( Design Fundamentals).

- 20-
D/H

6”

b“rnax

&l

s

“IlXIX

6”

b Vmax

kERA6E

DISTORTION / /_fD/H

I.5 2.0-I 2.5

0.5

I P

r-r;10

-A-A- 1 4

I I

__----- 5

- - -13

------I2

r

.-

‘LEGEND

6” = SETTLEMENT AT DISTANCE D

b

“max = MAXIMUM SETTLEMENT ADJACENT

TO THE CUT.

1

D = DISTANCE FROM EXCAVATION FACE.

H = EXCAVATION DEPTH

-v-4- 30

-o-o- 15

-v-v- 7

-A-A-6

---- 9

AVERAGE ANGULAR -X-Y- 283

-o-o- 8

-IaOSNUMBERS REFER TO PROJECTS

SUMMARIZED IN TABLE I.

Figure 10. Normalized settlements adjacent to a wall.
Figures 11 and 12 show the wall deformations assumed

and the corresponding settlement profiles predicted by the finite element

program.

Two soil conditions were analyzed:

(a) Normally consolidated clay with both the soil strength

and soil modulus increasing with depth.

(b) Elastic medium where the soil was assigned a constant

modulus with depth.

Figure 11 illustrates settlement profiles for the ideal cases of

tilting about base, rotation about top, and pure flexure. The first two

conditions may be considered representative of rigid wall behavior,

whereas the bulging cases represent deformations associated with a

flexible wall.

The results indicate that for tilting about base and flexure

the settlements are concentrated within a distance one-half the excavation

depth. On the other hand, when rotation is the predominant mode of

deformation, significant deformations may occur at distances up to

1.5 times the excavation depth from the excavation face.

Figure 12 shows the settlement profiles for wall deformations which are a combination of rigid wall displacement plus flexural

deformations. The results show the zone of .influence is greatly affected

by the nature and volume encompassed by the horizontal wall movement.

The zone of influence demonstrated by finite element

analysis ranged between 0.5H and 2. OH from the excavation face. This

is consistent with data from field measurements (Figure 10) with the

exception that field data are influenced by consolidation of softer cohesive

soils. Consolidation settlements, which extend the zone of influence out

further, are not accounted for in a finite element anal.ysis.

The effect wall movement has on the zone of influence is

another significant trend. Figure 11 and Figure-12 both show the

importance of minimizing movement below the excavation base.

1.60 LATERAL DEFORMATIONS IN ADJACENT SOIL MASS

Tied-Back Walls in Heavily Overconsolidated Clay

Normalized contour plots of horizontal deformations are presented

in Figures 13 and 14,

-22-
D/H

WALL DEFORMATION

6” = 1.9”

max

Cl) TILTING

D/H

1.0 1.5 2.0, 2.5

I ’

I

WALL DEFORMATION

rn DATA BEYOND REFLECTED

ELASTIC STRAINS AND IS

- I

D/H

- 9 q,s I.9 I,5 2.0 2.5

E&MATED FIELD PERFORMANCE

WALL DEFORMATION

I---D

Figure 11. Finite element prediction of surface settlement

profiles for normally consolidated clays,

-23-
D/H

0 0.5 1.0 1.5 2.0 2.5 0 I

0

y GATA BEYOND REFLECTS

/

’ t-+- ELASTIC STRAINS AND IS

NOT REPRESENTATIVE

OF FIELD BEHAVIOR.

WALL DEFORMATION

Q5.O

d

6 6h mox~ Bin

a) LATERAL TRANSLATION

D/H

PLUS FLEXURE

0,5 I;0 I.,5 I 2,.0 2,5

I

/

I

WALL DEFORMATION

-0

TILTING PLUS FLEXURE

D/H

0 0.5

0 ’

I.,0 I.,5 2.0 2;5

I I 1

I %I&TIMATE~ FIELD PERFORMANCE

- -

-

WALL DEFORMATION

l---+-D

6h max= 6 in..

C) ROTATION PLUS FLEXURE

Figure 12. Finite element prediction of surface settlement

profiles for normally consolidated clays.

-24-
WALLA/L DEFLECTION OEF LECTION O/H

ghJ 13” 05 1.0 I.5 2.0

mqx

\

\

\

\

I 0.5

I -7

i ii

I

I 1.0

I

I

FROM SHANNON AND STRAZER

(1970)

D/H - NORMALIZED DISTANCE FROM CUT.

z/H - NORMALIZED DEPTH OF CUT.

D/H

0.5 1.0 I.5 2.0 25

Figure 13. Normalized lateral movement for

tied-back excavation in heavily overconsolidated clays.

-25-
WALL DEFLECTION

Figure 14. Normalized later al movements from finite element

analysis for normally consolidated cLa.ys,

-26-
The aforementioned field data suggest two trends. First, the

pattern ‘of the lateral movement follows closely with the deflected shape

of the sheeting. Second, the lateral movements can extend a substantial

distance from the excavation face, and may involve general movement of

the soil mass embodied by the tiebacks. Discussion of several case

historties is made in Volume II (Design Fundamentals).

In overconsolidated clays and shales the movement is believed

to be associated with lateral expansion following stress relief from the

excavation. A weak layer below the excavation would add to the magnitude

of movement.

JValls in Normally Consolidated Clay

There are little field data available regarding the distribution of

horizont,al displacements for excavations in a normally consolidated

clay for comparison with the observed data for the heavily overconsolidated

clays. Therefore, the results of the finite element studies used to develop

Figure 12 were reduced to provide some insight to the distribution

which might be expected for ideal conditions. These results, shown

in Figure 14, indicate that the zone of significant movement is confined

to an area described by a 1 on 1 slope from the base of the sheeting.

As expected, it is within the theoretical yield zone. The movements

are largely controlled by the sheeting displacement, with the zone of

significant movements increasing with depth in the same pattern as the

sheeting movements,

1.70 E:FFECT OF CONSTRUCTION PROCEDURES

It is well known that construction procedures can have a significant

effect on the performance of excavations.

L,owering of the ground water level either by pumping or by

seepage into the excavation can result in significant settlements. These

settlements could be associated with consolidation of the soil or, in the

case of granular soils, the piping of soil into the excavation.

Poor installation techniques for tiebacks or struts can lead to

surface settlements. Tiebacks should be carefully drilled to minimize

the soil removed from holes, Also, any voids remaining after the tieback

is installed should be filled with grout. Struts, rakers, and wales should

be tightly wedged and preloaded to prevent movement of the wall. Earth

berms when used to provide temporary support before installing a strut

have been observed to be of little value in preventing wall movement.

-27-
Even though the entire support system may be in place, the sides

of the excavation may continue to creep inward with time. This problem

appears to be particularly acute in tied-back walls in very stiff to hard

clays. There is also some evidence to indicate that lagging in soldier

pile walls tends to pick up more load with time in all soils. Excessive

bulging or even failure of some lagging has been observed.

1.80 ESTIMATING SETTLEMENTS

The data presented in this section may be used to obtain rough

estimates of the ground movements which might occur adjacent to a

support wall. The reason for making this estimate is to provide some

additional input to aid in the decision of whether or not to underpin

adjacent structures or utilities.

Settlements may be estimated using both Figure 3 and Figure 10.

Once the soil type and excavation geometry are defined, an estimate

of the maximum settlement may be made from Figure 3. Figure 10

provides a means of estimating the angular distortion and zone of

influence of the ground movements. In the case of cohesive soils,

Figure 5 may be used to estimate the wall stiffness necessary to limit

the settlements.

-28-
CHAPTER 2 - GROUND WATER IN OPEN CUTS

2.10 GROUND WATER CUTOFF

Lienera

Cutoff walls are used for the following purposes:

1. To avoid or to minimize dewatering of the excavation.

2. To lessen or to prevent lowering of ground water level outs

the excavation.

ide

3. Because it may be impractical to place lagging in soils that

are extremely difficult to dewater in advance of excavation (such as

silts and./or dilatant clayey sands).

4,. To cut off pervious water bearing strata within or just below

the bottom of the excavation; thus, protecting against the possibility

of a l’blolwt’ condition or other source of ground loss.

Soldier Pile Wall

Inherently, a soldier pile and lagging wall is not watertight. If

ground water is to be controlled it must be done by dewatering or by

grouting or freezing. In “running” soils it is essential to maintain the

ground water level below the working face in order to prevent in flow

and subsequent ground loss.

Interlocked Sheeting

Provided that the steel sheet pile wall remains intact and

penetrates into an underlying impervious stratum, the effectiveness of

sheet piling as a cutoff will be very significant in pervious sands and

gravel. On the other hand, in granular soils of low permeability (for

example:, silty or clayey sands) interlocked sheeting will have relatively

little effect on the flow into the excavation. In all cases, however,

sheeting effectively cuts off flow in pervious interbedded layers, which in

the case of soldier pile walls, may lead to ground loss at the face.

With regard to maintaining ground water level outside the

excavation, interlocked sheeting is effective in pervious granular soils.

For relatively impervious soils (such as clayey sands, silts, and clays)

the sheet piling is essentially equivalent to the permeability of the soil

and therefore, will have little or no effect on the seepage pattern toward

the excavation or on lowering of piezometric levels.

-29-
The above discussion applies only to intact sheeting. The presence

of boulders, difficult driving conditions, or obstructions can lead to

ripping of the sheeting and/or jumping out of interlocks which will seriously

impair if not destroy the effectiveness of the cutoff wall.

Another common problem is when the effectiveness of a cutoff in

pervious soil depends upon achieving a tight seal on rock. This situation

may be especially acute when rock occurs within the depth of excavation

because of the threat of ground loss below the tips of the sheeting.

Concrete Diaphragm Walls

For all practical purposes, a well constructed concrete diaphragm

wall is essentially impermeable. It will effectively cut off flow and

prevent ground water lowering outside the excavation provided there

is penetration into an underlying impe-rv-io-us formation.

2.20 SEEPAGE PATTERN TO EXCAVATION FACE

As mentioned previously, interlocked steel sheeting has relatively

little influence on the seepage pattern in impervious soils. As a result,

when cuts are made below ground water there will be flow to the face of

the excavation. In clays, such a flow will be so small that it may not even

be noticeable.

An example of a flow net for this type of situation is shown in

Figure 15. During the initial process of excavation, deformation in the

soil will generate sheas strains and cause pore pressure changes.

Eventually, these pore pressures will be dissipated and a steady state

seepage pattern will develop as shown in the figure.

The equipotential lines shown in the figure demonstrate the

changes in hydrostatic stress. Such changes in hydrostatic stress

lead to a time dependent equivalent change in effective stress and

consolidation of the soil. In soft normally consolidated clays or organic

soils the associated amount of consolidation can be significant and will

contribute to displacements behind the excavation.

The foregoing case is important because if illustrates that steel

sheeting may not be effective in preventing consolidation of normally

consolidated soils within depth of cut. Soil compressibility and rate of

consolidation must be considered.

It would not be possible to recharge the ground water level in the

cohesive soils of this example. Diaphragm walls should be considered

in cases where there is a need to prevent displacement.

-3o-
INTERLOCKED STEEL SHEETING

ROCK (IMPERVIOUS)

(

0.50h

xi >

- INDICATES CHANGE IN TOTAL HEAD

FROM INITIAL CONDITION TO STEADY STATE

SEEPAGE CONDITION.

Figure 15. Change in pressure head for cut in impervious

soil.

-31-
CHAPTER - 3 LATERAL EARTH PRESSURE

3.10 BASIC CONSIDERATIONS

Earth Pressure at Rest

The ratio of the geostatic horizontal to vertical effective stress

of a natural soil formation is defined as:

where:

KO

= coefficient of earth pressure at rest

3

= horizontal effective stress

c$ = vertical effective stress

For granular soils Terzaghi and Peck (1968) suggest K. values of

0.5 for loose deposits and 0.4 for dense soils. Generally, K, can be

estimated for normally loaded soil deposits as:

K, = 1 - sin 2

where:

K

z

= coefficient of earth pressure at rest

= angle of internal friction in terms of effective stress.

For cohesive soils, K. is primarily dependent on the overconsolidation ratio (OCR). Normally consolidated clays typically have K, values

of 0.5 to 0.6; lightly overconsolidated clays (OCR ,C 4) have K, values

up to 1; heavily overconsolidated clays (OCR - 16) K, may range up to

a value of 2.

Active Earth Pressure

Lateral displacement (as shown in Figure 16) transforms the

state of stress in the ground from the at-rest condition to the active

condition. The mechanics of this process are the mobilization of

full shear resistance within the soil mass -- a state of stress referred to as “plastic equilibrium”.

Figure 16 shows the active earth pressure distribution associated

with displacement modes. The fully active state stems from lateral

translation, from rotation about the bottom, or from a combination

-32-
(a) FULLY ACTIVE

ACTIVE

KJE

T

\

\

\

\

(b) ARCHING ACTIVE

ARCHING ACTIVE

Figure 16. Earth pressure distributions for active and arching

active conditions.

-33-
of both. ‘The earth pressure distribution is triangular and the resultant occurs at the third height of the wall.

The direction and magnitude of active pressure depends upon

whether or not there is wall friction. The particular case of horizontal

surface and zero wall friction is the Rankine fully active condition,

shown in Figure 16. For this case, the active stress’acts horizontally

on a vertical waLL. The Rankine coefficient of active earth pressure

Ka is the ratio of the effective stress.

=-G

6?

= K,

For sands, Ka = tan’ (450 - J/Z)

For cohesive soils,

General case (3, c):

K a = tan2 (45’ - !z/2) - e tan (45’

Speciat case (d = 0, c=Su):

Ka = 1 - 2su

H

wher e :

Ka F coefficient of active pressure

i, c = friction angle and cohesion intercept

@V = vertical effective stress

*h = horizontal effective stress

sa = active earth pressure (horizontal)

S

U

= undrained shear strength (B = 0 case)

According to the Rankine expression, the pressure distribution

for cohesive soils is theoretically in tension in the upper part of the

wall as shown on Figure 17a. Frequently, adhesion simply does not

-34-
rH

I

(a) RANKINE ACTIVE PRESSURE DISTRIBUTION

IN COHESIVE SOILS

TENSION

N = *

- 2% H

(I- ;I

I

=‘dH(I-2S, 1 ’

<p,IjH ‘6HI

t b) TRIANGULAR PRESSURE DISTRIBUTION

EQUIVALENT TO NET RANKINE FORCE

(N =

pn

= 4 H- (‘6H-4 S, 1

= & %H*- 2SuH

= & YH*(lFigure 17 Earth pressure distribution for cohesive soil (q!~ = 0).

-35-
(or cannot) develop and therefore tension cannot occur. However,

the net total lateral force on the wall is equivalent to that described

by subtracting the “negative” pressure at the top from the positive

pressure at the bottom. Assuming that this net force increases linearly with depth of wall, it can be represented by a net pressure diagram

with a triangular distribution of the same force magnitude as shown

on Figure 17b. The ordinate at the base of the wall is:

‘a = dH-4S

U

3.20 INTERNALLYBUCEDCOFFERDAMS

General

Initially, the interna bracing is set near or at the top, thus

restraining inward displacement. With each stage of excavation and

bracing there will be progressive inward displacement below previously

placed braces. The net displacement profile typically takes the form

shown in Figure 18 (after Bjerrum, et al, 1972).

Characteristically, there will always be some inward

rotation about the top, at least in the upper portion of the cut. The

degree of bulging and displacement. below the cut depends upon several

factors -- the distance between braces, the stiffness of the waU, and

the stiffness of soils near the base of the wall. In general, the

resulting deformation pattern most closely resembles the arching

active condition. Therefore, a parabolic, rather than triangular,

pressure distribution is most Likely to act on the wall.

Figure 19 shows the conventional procedure for

analyzing empirical load data. The resulting apparent earth pressure

diagrams are used to develop an envelope encompassing the maximum

distributed pressures. This design envelope then represents the

maximum strut load that can be anticipated at any stage of construction.

Design Earth Pressure Diapram

Apparent earth press’ure diagrams suggested by Terzaghi

and Peck (1968) for design of braced walls are shown on Figure 20.

Strut loads for a given Level are determined by reversing the procedure

used for development of the diagram. A strut is designed to support a

load described by the area between the mid points of the adjacent upper

and lower support levels.

The following discussion does not include the effect of surcharge

(see Section 3.40).

-36-
/

/- ., l l * í *

/í

I

,I

-y-+ d 1

\

--A

\ - 6

I

-1

I

-----A

\

\

\

., *:.._

. \

..’ ‘., ..’ 1. o

-6

DENSE

STRATUM

// ’ /’

DENSE

STRATUM

Fi.gure 18 Mode of deformation of internally.braced cofferdam

(after Bjerrum, et al, 1972).

-37-
RD

lf , D

w LD b

L4 I

.

. .

. .

REACTIONS

%*ë8*fE l %

APeARENT EARTH

PRESSURES

R

P =-A

A LA

l etc.

Figure 19 Conventional procedure for development of

earth pressure diagram,

-38-
PRESSURE DISTRIBUTION

:

a) Sands a) Sands

‘\‘\ lrGf---\ lrGf---\ KA = tan2(45 - $/2) - $m \\

H

-Rankine Rankine Active Acti ve L

!

f\

- 0.65 0.65 KAyH KAyH + ’

b) Soft to Medium Clays

W> 6)

For clays, base the selection

on N =sr”

U

Equivalent Rankine Active

KA = 1 -mA= lm4 4 s

YH N

m = 1.0 except

where cut is underlain by deeF

soft normally consolidated cla

Stiff-Clays

For Nd 4

For 44 Nh 6, use the larger

of diagrams (b) and (c).

TOTALFORCE

P, = Trapezoid

P

A

= Rankine

Pt = .65 KAyH2

PA= . 50 KAyH2

Pt

- = 1.30

pA

m = 1.0

Pt = . 875y~‘(l - R)

PA= . 50yH2(l - R,

Pt

- = 1.75

pA

2

Pt = . 15yH to . 30yH2

pA

-=4, PAZ0 N

Ne 4, PAL 0

NOTE: Equivalent

Rankine Active = 0

Figure 20 Design earth pressure diagram for internally braced

flexible walls (sands, soft to medium clays, stiff

fissured clays), from Terzaghi and Peck (1968).

-39-
a. Sands : This diagram, which was developed from dewatered sites applies to cohesionless soils. If the soils outside the

excavations remain submerged, then the earth pressure should be

computed using the buoyant unit weight of the soi.L. Hydrostatic pressures

are treated separately and added to the effect of the earth pressure.

b. Soft to Medium Clays: The recommended earth pressure

diagram for these soils is shown in Figure 20b.

The value of *m’ used in the determination of the ordinate for

earth pressure applies to situations where the cut is underlain by a

deep deposit of soft clay. Its value can only be determined by empirical means from measurements and performance of an actual excavation. Experience thus far, reported by Peck (1969) from cases in

Mexico City and in Oslo, Norway, leads to the conclusion that the

value of ‘ml is on the order of 0.4 for sensitive clays. For insensitive

clays the value of ‘m’ may be taken as 1.0.

c. Stiff Clays : The recommended apparent earth pressure diagram for stiff clays is used when the stability number, N, is Less

than 4. This empirical diagram is independent of the value of shear

strength; the lateral earth pressure is a function of the gravity forces

only.

d. Heavily Over consolidated Very Stiff Fissured Clay: Several

cases have been reported which suggest that stress relief from excavation leads to lateral deformation of these soils toward the excavation. Soil behavior would suggest that the potential for expansion

increases with increasing overconsolidation ratio, increasing pLasticity of clay, and for cuts below the water table with the intensity of

fissuring in the soil. For strutted excavations this condition may lead

to build up of strut load with time.

Design criteria for cases involving potentially laterally expansive

soils are as yet undeveloped. Therefore, a Laboratory test program

(possibly stress-path triaxia.1) should be undertaken toaid in evaluating

the magnitude of the problem. Prototype test sections with construction monitoring are also recommended.

e. Dense Cohesive Sand; Very Stiff, Sandy Clay: Recommended

design diagrams for dense cohesive sands and very stiff sandy clays

are shown in Figure 21. The minimum pressure Line is associated

with cuts less than about 30 or 40 feet deep, with reasonably consistent

spacing between wale levels and relatively uniform soil conditions.

The maximum pressure line is recommended to cover uncertainty

-40 -
lH

L

TH

1

(a) RELATIVELY UNIFORM C-QOtu-TIONS

Pt =0.112h-12 TO 0.1881(H2

RANGE

4

0.25h-l 1

0.80H :I0.20H

(b) UPPER THIRD OF CUT DOMINATED BY

COHESIONLESS SOIL

C

‘T Pt =O. 135 YH2 TO 0.225ifH2

Figure 21. Proposed pressure diagram for internally braced

flexible walls (dense cohesive sands, very stiff sandy clays).

-4l-
regarding the effect of weak strata within the depth of cuts, contingencies arising from construction (for example, over -excavation below

support level, or ineffective toe berms), and cuts in excess of 60 or

70 feet deep.

Cohesive soils near the top of the cut will justify pressure

reduction as shown in Figure Zla. Absence of cohesive soils near

the top of cut will require the higher pressures associated with Figure

21b.

f. Stratified Soils :

idealizedoil profiles.

The aforementioned cases apply to readily

Actual soil conditions may have a stratigraphy

which does not conveniently match these simplified cases. Moreover ,

an irregular ground surface or surcharge may complicate the analysis.

Under such circumstances, one approach is to determine the

Lateral thr,ust either on the basis of classic active earth pressure or

on the basis of trial planar sliding surfaces and wedge stability anaiysis.

In this latter case the most critical wedge is used to determine the

lateral thrust (see Chapter 6). In such cases, hydrostatic forces are

treated separately.

Once the Lateral thrust is determined, it should be increased

by the most appropriate value of P,/P

pirical diagram to the force determine%

(ratio of force from the emfrom the analysis of active

earth pressure or wedge equilibrium). The designer must choose

the most appropriate ratio based upon a comparison of the actual case

to one of the simplified cases presented in this section,

The final question is one of pressure distribution. Initially,

the designer must compare the actual case with the simplified cases.

Serious qu.estions may need field measurements to provide data input

during construction.

3.30 TIEBACKS

Background

Many practitioners have successfully applcied the empirical rules

developed for internally braced walls to tiebacks; others make variations

for tied-back installations. In any event, at the present time there are

no empirical methods for tied-back walls that have been accepted as

universally as Peck’s rules for internally braced flexible walls.

-42 -
Recommendations for Tiebacks

The following discussion does not include the effect of surcharge

(see Section 3.40).

Only limited documentation is available to quantify conclusions concerning the relative magnitudes of appropriate pressure

envelopes for tiebacks and bracing. Accordingly, there is no present

justification for a major departure from loading on internal bracing.

In general, the force magnitude of the followzing proposals is similar,

or the same, but the distribution has been changed slightly.

Soil type classifications are the same as for internal bracing,

namely: sands , soft to medium clays, stiff clays, and fina.lly, dense

cohesive sands or very stiff sandy clays. A triangular pressure

distribution, increasing linearly with depth, is recommended for

soft to medium clay; a uniform pressure distribution is recommended

for all other cases.

a. Sands: Where deformations are critical, and it is intended to prestress to 100 percent of design load, compute force using

Ke. For dense sands K, = 0.4; for loose sands K, = 0.5. Thus,

the uniform ordinate will vary from:

Uniform Pressure, p = 0.20 # H to 0.25 I H

Force, P, = 0.201H2 to 0.25jJH2

c

Where deformations are not cirtical, use K

Ko+ Ka

=

avg

, that

3

is, a coefficient midway between active and at rest. A similar procedure

was used by Hanna and Matallana (1970).

b. Stiff to Very Stiff Clays: ‘Use. a uniform pressure ordinate

varying from 0.15 II H to 0.30 rH to produce the same force magnitude as

that for braced excavations. The higher value is associated witka

stability number of about 4. The lower number is associated with

very stiff clays where the stability number is less than 4. The force

varies as follows:

Stiff clays, Pt = 0.30 II I-l2

Very stiff clays, Pt = 0. 151H2

-43-
C. Cohesive Sand, Very Stiff Sandy C.Lays: Compute the total

force associated with the diagram for braced excavations (Figure 2) and

distribute uniformly with depth. Relatively uniform conditions :

Force, Pt = 0.112 ifH2 to 0.188 #H2

Uniform Pressure, p = 0.112 XH to 0.188dH

Upper third of cut dominated by cohesionless soil:

Force, Pt = 0.135 rH2 to 0.225 4 H2

Uniform Pressure, p = 0.135 ZH to 0.225 i$ H

d. Soft Clays: It is unlikely that tiebacks would be used unless

they could be embedded in an underlying denser stratum of soil or

or in rock. Design with a triangular earth pressure diagram assuming

at rest conditions and a K. value between 0.5 and 0.6.

Force, Pt = 0.25 I H2 to 0.30 3H2

In normally consolidated, sensitive clays, excessive prestressing should

be avoided because of the potential for induced consolidation.

e. Stratified Soils: As with braced excavation, an approach

based upon active earth pressure or wedge equilibrium should be investigated.

3.40 SURCHARGE LOADING

General Backvround

Surcharge near excavations may be the result of many differ -

ent types of loading conditions including footings, structures, storage

of construction materials, or traffic. The lateral pressure caused

by a surcharge load on a retaining wall has been investigated for a

variety of different loading and soil conditions (Spangler, 1940; Newmark,

1942; Terzaghi, 1954b). This pressure is in addition to the normal

earth and water pressure.

Theoretical Considerations

The four basic loading conditions for which solutions of the

lateral stresses in an elastic medium are readily available are:

1. Point Loading

-44-
2. Uniform line loading

3. Irregular area loading

4. Uniform area loading

Practical Considerations

With regard to surcharge loading from construction operations, it is common to take a distributed surface surcharge on the order

of 300 psf to cover storage of construction materials and general

equipment. Usually, this surcharge should be considered within

a rather limited work area on the order of 20 feet to 30 feet from the

coffer dam wall.

A second major consideration is the question of concentrated

loads from heavy equipment (concrete truck, cranes, etc.). Lateral

thrust from such equipment would easi.ly be covered within the 300 psf

sur charge, provided that the equipment were more than approximately

20 feet from the wall. On the other hand, such equipment within a

few feet of the wall may create a concentrated surcharge loading

which would be of far greater significance than a uniform surcharge

loading. This must be accounted for separately.

Point Load and Line Load

Solutions, summarized by Terzaghi (1954b) are shown in Figures

22 and 23.

Irregular Area Loading

Figure 24 shows an influence chart for evaluating the lateral

stresses acting on a rigid wall due to a rectangular loading (Sandhu,

1974). These charts assume a Poisson’s ratio of 0.5 for the soil

mass. Using the influence charts for point loadings, the lateral stress

due to an irregular sur char ge loading can be calculated more easily.

Uniform Area Loading

The solution for lateral stresses on a rigid wall is presented

in Figure 24. An example of the stress effect with depth is shown

in Figure 25. Note that the stress influence below a depth of about

1.5B is negligible.

-45-
co

II

C

2

0.6

3

0.8

I

I h I I

0 0.5 3

VALUE OF 0-h (!$)

P

(a)

SECTION a-a

P h (TOTAL FORCE)

FOR m5 0.4:

(b)

FOR ma.4:

H’

ch $1 u

l.77m2n2

(rn4+#j3

NOTE:

me+

n=+-

Figure 22 Lateral stresses on the face of an unyielding

wall from a point loading (NAVFAC, 1971

and Terzaghi, 1954b).

-46-
$ 0.4 0.4

II

E

!i! 0.6”” 0.6

0.8 0.8 I

0 cl2 0.4 0.6 0.8 . 1.0

m-n,

-.-

0.2 0.2

1.0 ’

I 1111 Y I 1 I I I I

VALUE OF(& ( tL)

’ 1.0

LINE LOAD QL (a)

FOR m 5 0.4:

P

ph = 0.55 QL

RESULTANT P = o’64 ”

.h (ma+ I )

Figure 23 Lateral stresses acting on an unyielding

wall from a uniform line loading

NAVFAC, 1971 and Terzaghi, 1954b).

-47-
mt -

B L

2 ’ “=z, p=q x Ip

CL

A( t 0.5

q = SURCHARGE

L = LENGTH PARALLEL TO WALL

B = LENGTH PERPENDICUL

TO WALL V.-t

0.3

0.2

0.1

n

Figure 24 Lateral stresses on an unyielding wall due

to irregular surface loading (Sandhu, 1974).

-48-
0.58

tu

- I.OE

z

1.5F

INTENSITY OF LATERAL STRESS

u”h=K*n

\RRlGlD WALL

Figure 25. Lateral stress on rigid wall from surcharge of

widlth B and infinitely long (solution from Sandhu, 1974).

-49-
A second approach is to apply an earth pressure coefficient,

K, to the surcharge loading and to consider the surcharge effective

within some portion of the cut. The magnitude of this coefficient

will range from K, (active earth pressure) to K, (earth pressure at rest).

-5o-
CHAPTER 4 - PASSIVE RESISTANCE BELOW BASE OF EXCAVATION

4.10 GENERAL

The design should provide that the soils below the base of an excavation mobilize sufficient passive resistance for force equilibrium or

for limiting movement. The performance of the wall will depend upon

the spacing of the support levels since the greater the spacing, the

greater the passive resistance (and movement) that will be required

below the lowermost support level. Figure 26 illustrates the case of

a wall in which the passive resistance of the soil is insufficient to

limit exce s sive wall movement.

This section deals with the selection of soil parameters and

methods used to evaluate passive resistance. It does not deal with

the depths of penetration required to maintain overall stability of the

earth mass or to limit displacements in the earth mass.

4.20 SOIL PARAMETERS

Granular Soil

Granular soils are free draining and cannot sustain positive or

negative pore pressures generated by strain or load changes for even

a short period of time. Therefore, analyses of the stability of granular

soils are performed on the basis of drained strength parameters and

effective stresses in the ground. The appropriate soil strength parameter is the angle of internal friction, i. For design, granular soils

are assumed to have no cohesive strength component,

C ohe sive Soil

Because of the load decrease from excavation, soils in the

passive zone just below the excavation will initially experience a

pore pressure decrease, Pore pressure may become negative. With

time, the pore pressure will rise. This may be accompanied by heave,

caused by swelling of the soil.

Limiting case strength parameters for passive pressure computation are:

a. Immediate Condition: Pore pressures generated by unloading

and strain do not have time to dissipate. Use undrained strength of soil

S, at natural water content. Conventionally, this is determined from

vane shear, unconfined compression, or unconsolidated-undrained

compression tests.

-51-
Figure 26 Movement at wall base due to

insufficient passive resistance.

-52-
b. Ultimate Condition: Pore pressures generated by unloading

and strain are dissipated by drainage. Effective stresses can be

computed on the basis of static water levels.

from the efffective stress envelope, ‘c and 3.

Use strength parameters

General recommendations for strength relationships are to

use undrained strength for the “during excavation” stage, and effective

stress strength parameters for the final construction condition. Greater

accuracy in determining strength values can be obtained by measuring

pore press’ures during construction and by appropriately modifying

the strength values (either undrained strength or drained strength).

For overconsolidated soils, the undrained strength at natural

water content may be greater than the drained strength. Therefore,

indiscriminate use of undrained strength without regard to pore pressure

dissipation may be unsafe.

4.30 ANALYSIS OF PASSIVE RESISTANCE

Several articles and texts address the problem of passive pressures

that can develop behind a continuous wall (Terzaghi, 1954b; NAVFAC,

1971). In cohesionless soil, wall friction modifies both the direction

and magnitude of the, passive resistance . Typically, the resultant of

the passive pressure acts at an angle 6 equal to l/2 to 2/3 of the angle

of inte ma1 friction. The following table (from Terzaghi, 1954b)

summarizes values of d and 6 .

6 6=0 6= Q/2 6= 213 d

25’ 2.46 3,oo 3.20

3o” 3.00 4.20

35O

4.80

3.70 6,50 7. 30

4o” 4.60 9.20 11.00

The passive pressure for drained loading or in terms of effective

stress at depth, z, will be:

$

= (Tvtan2 (45O - J/2) + 2; tan (45O t J/2)

whe:re:

8P

= passive pressure (effective stress)

6,

= vertical effective stress = llmZ-U

5 = angle of internal friction (effective stress envelope)

C = cohesion intercept

-53-
For the above drained condition, in which by definition there is

no excess pore pressure, the total stress at any depth, z, will be

%

=iipt 1 z

W

where:

‘h = lateral stress

1 w = unit weight of water

The passive resistance of cohesive soils in an undrained condition

should be evaluated on the basis of the undrained shear strength, S,,

and the in situ total vertical stress, CT . For a continuous wall, the

passive pressure at a given depth will &ual.

5

P

= cTvt2s

U

= Yzt2s

U

where:

oV

= total vertical stress = #m z

S

U

= undrained shear strength of the soil

In this case, the water pressure is not added separately because

pore pressure is, already accounted for in the determination of the

undrained strength, S,.

Soldier pile walls are not continuous walls, therefore the passive

earth pressure coefficients must be modified from those used for

continuous walls. Broms (1965) showed the passive resistance of

laterally loaded piles based on pile width and on K values for continuous

walls was too conservative, His study showed tha: soil arching and nonplane strain conditions increase the capacity of individual piles. Broms.!

recommendations are given in the charts shown in Figure 27. It should

be noted that for cohesive soils the lateral resistance of the soil

should be neglected to a depth of 1. 5 Pile diameters. In cohesionless

soils where the depth of penetration is greater than one pile diameter,

soil arching causes an effective increase of 3.0 in the value of K .

P

A factor of safety of 1. 5 is recommended for use in passive

pressure calculations.

-54 -
D

v- h

w 3UvbKp

-7-Lb

- -

D

1

UNDRAINED STRENGTH

(COHESIVE)

EFFECTIVE STRESS

PARAMETERS

(COHESIONLESS or COHESIVE)

Figure 27 Passive pressure for soldier piles

(after Broms, 1965). (Modified. )

-55-
4.40 OVERCUT DESIGN DETAILS

Over-excavation below the required support level depth is

co.mmon either to obtain working room or to muck up the bottom.

During intermediate excavation phases assume a minimum

of two feet of overt-ut before strut placement. At final depth assume

a minimum of one foot of overcut.’

4.50 BERMS

Lateral resistance of berms will, of course, be lower than

the lateral resistance of a horizontal plane at the top elevation of the

berm. One method of analysis is wedge or logarithmic spiral force

equilibrium of trial failure surfaces. Another procedure is to replace the berm with an equivalent sloping plane and assign the appropriate passive coefficient (Terzaghi and Peck, 1968; NAVFAC,

1971).

-56-
CHAPTER 5 - DESIGN ASPECTS OF LATERAL PRESSURE

5.10 A.NALYSIS OF WALES AND SUPPORT WALLS

General

Deflection of structural members supporting soil causes

arching of earth resulting in a reduction of pressure near the center

of spans and a concentration of pressure at the supports. Hence, the

actual bending moments in wall elements and wales is less than that

which would be computed assuming a uniform loading on these flexural

members.

The approach used herein, for moment computation in wales

and support walls, is to apply a uniform pressure equal to 80 percent of the loading diagram. For evaluation of loads in internal bracing

and tiebacks, the full loading diagram (100 percent) is used. (See

Figure 28).

When rigid walls support the earth, arching will be minimal;

therefore, structural design of the wall as well as other elements

should be based on the full pressure diagram.

gontinuous Members

The following expression should be used for computing moments

over continuous members (either wall member or wale) with uniformly

applied loads :

M=Cwl’

where:

M = moment

C = moment coefficient

w = distributed load on span

1 = span length

Since construction methods greatly influence the position of the

elastic line of members (especially vertical members), there is no

practical way that the moment can be pre>cisely analyzed. Therefore,

a coefficient of C = 0.10 is recommended for continuous members

supporting a uniform distributed load.

-57-
Rc ---

AREA A

AREA B

AREA C

AREA 0

AREAE

STRUT LOAD PER LINEAL FOOT OF WALL. IS ,EQUAL TO DESIGNATEDAREA

EXAMPLE: R&V <h+ 5 ) 2 2

Figure 28. Load determination from apparent earth pressure diagram.

-58-
Discontinuous Wales

The moment in the wale wiL.1 depend on the splice detail.

For splices which occur at a strut and which tie the wale together

with a steel strap, which transfers shear but not moment, zero moment

should be assumed at that point.

Wales supporting uniform load with moment splices over

less than three spans should not be considered continuous. Three

spans or :more should be considered continuous using a moment

coefficien.t, C 2 0. 10.

The moment in wales supporting concentrated loads

(as from soldier piles or tiebacks) should be calculated on the’basis

of statics. Assume full continuity where moment splices are used;

assume zero moment in other splices.

Member Connections

It is common to design splices for the full structural

capacity of the member (both shear and moment). This is often done

with a combination of fully penetrating butt welds and cover plates.

Figures 29, 30, and 31 show some typical details

for splices and wale to strut connections. For splices that

are butt welded it is often assumed that the butt weld is only 50 percent

to 75 percent effective since the beveled edges at the splice are field

cut. Hence, the cover plates are designed to carry 25 percent to

50 percent of the member capacity. In designing a strut to wale

connection, stiffness must be provided to prevent web crippling.

Also, if raked struts are used, a knee brace is required at the strut

to prevent buckling of the wale from the vertical component of load.

Lagging

The determination of lagging size is largely based on

the past experience. See Chapter 9 (SoMier Pile Walls).

5.20 BRACING AND TIEBACKS

Bracing and tieback loads must be determined for the most

critical construction condition. This may be at an intermediate depth of

cut or at full depth.

-59-
SOLDIER PILE

2 k’s F2 FL’S

GAS CUT ENOS OF SOLDIER PILE

FULLY PENETRATING BUTT

WELD

Figure 29. Typical splice with butt welding,

-6O-
PLAl

-

u I’kLEAR SPACE v

IF NO BUTT

WELDING

‘F ‘=y

, M , PLATE

/‘- 1 .

w-w- --ah a SHEAR

PLATE

STIFFENER AT SOLDIER PILE

\

STIFFENER

AT STRUT

Figure 30, Plan view of typical wale splice and strut connection.

-61-
l----lb ---.-_ -- -

y CONTACT LAGGING

,~~A+‘---;kkS&;FFENERS 1/

SPACER p--.-c

USED WITH

--_ -_--. .--__.-

SOLDIER

PILE

Figure 31. Typical strut-wale-soldier pile

connections (elevation view).

-62-
For bracing:

a. At final depth, use allowable stresses by AISC Code.

b. For temporary conditions at intermediate depth of excavation

use AISC -t 20 percent.

For tiebacks

6, (Tiebacks).

use the stress values stated in Volume III, Chapter

5.30 DEPTH OF PENETRATION BELOW CUT

Lateral Resistance

When use is made of the design pressure diagrams, a reaction

at the base of the cut is assumed to exist which is equal to the lowest

area shown in Figure 28. This reaction is provided by the passive

resistance of the soil beneath the cut.

Figure 32 illustrates the method for determining the depth of

penetration in competent soils that are capable of developing adequate

passive resistance. Soils satisfying this condition are medium dense

to dense granular soils and stiff to hard clays. The general method

of analysis is:

a , Compute the equivalent reaction at the base of the cut (RE).

b. Determine the depth required to satisfy force equilibrium on the

hokizontal plane.

c. Check the maximum moment at or below RD against overstressing of the support wall.

d. Drive sheeting to a depth 20 percent greater than that required for force equilibrium.

In cases where the soils below the base of the cut are soft

clays the passive may never equal the active pressure, no matter how

deeply it is driven. Since the passive resistance from the weak layer is

small the sheeting acts much like a cantilever member; thus, a large

load is developed in the lowest strut. For these conditions, where the

base is stable against bottom heave, little is gained from driving the

sheeting deeply below the bottom of the cut (see Figure 33),

-63-
pt (FROM DESIGN EARTH PRESSURE

FIGURE 28)

BOTTOM OF CUT

V = SHEAR AT TIP OF SHEETING *O

Kp= COULOMB EARTH PASSIVE COEFFICIENT

(SEE SECTION 6.30 1

1. Compute RE = 0. 5 p

t Ld-e

2. Compute depth x such that: Pp = R + P

E A K

IJsc minimum F. S. = 1. 5 for pass.ivedoefficienL, KL = e

.

3. Check M

max

5 yield moment of sheeting

4 . Drive to depth D = 1. 2x

Figure 32. Procedure for determining depth of penetration

in relatively uniform competent soil conditions.

-64-
BOTTOM OF

PASSIVE

Su = i& = 0.208H

RANKINE ACTIVE=)JH-2(.20ijH)=

0.601/H AT L = H

RANKINE PASSIVE

AT z=H=O t 2(.20ifH) = 0.401H

1 . Theoretical passive resistance is not available

below bottom of cut to develop horizontal reaction.

In fact, the net force below cut is theoretically

toward excavation, based on active and passive

pressure.

2. IJse nominal penetration of 0. 2 FI or 5 feet whichever

is greater,or penetration to cut off pervious layers.

3 . Check base stability (see Chapter 6).

4. Design for cantilever condition below E.

Figure 33. Method for analyzing sheeting with

weak underlying Layer.

-65-
Therefore, a minimal penetration of five feet or 20 percent of

the excava.tion depth, whichever is greater, is recommended. In

situations where the base is unstable, consideration may be given to

deeper penetration and stiffer sheeting as a possible means to prevent

bottom heave.

Bearing Capacity Considerations

Load capacity must be evaluated when there is a downward

component of load, as is the case for inclined tiebacks. This may

be accomplished by pile driving formulas or by the empirical and

semi-empirical methods outlined in Chapter 7 (Bearing Capacity).

5.40 EX.AMPLE SITUATIONS

Case I - Homogeneous Soil Profile

Case I is the analysis of a homogeneous soil profile which

provides a basis for comparison of required penetration depth and

strut load variations. It represents, most ideally, the conditions where

the design envelope is appropriate. The method for analyzing soldier

piles set in concrete-filled pre-augered holes is also presented,

Case II - Soft Soil Stratum to Base of Excavation Underlain

byDense Stratum

Let F represent an unyielding passive support at the base of

the excavat:ion, Let E represent the first wale level above the base.

Let D repr’esent the second wale level above the base.

-66 -
As the excavation proceeds below level D to level E, little

passive resistance is provided because of the soft soil above F; hence,

the wall d.eflects inward. Effectively, the wall spans from level D to

F (the excavation base) with full active pressure applied and negligible

passive resistance above F. The deformation of the sheeting is such

that during this excavation stage it resists essentially the same load

over the mrpan D to F whether or not strut level E is installed. This

would be particularly true in the stiffer diaphragm walls. The effect

of this large unsupported length is twofold:

a . Since the sheeting has already assumed an elastic line

such that it resists the full active load, little load is transferred to

strut level E, Hence, strut level D effectively takes a disproportionate

share of the load.

b. The moment in the sheeting is greatly increased by the long

unsupported length.

-Case III - Soft Layer Underlying More Competent Soil

Design sheeting as cantilever below lowest strut with normal

penetration below base (see discussion Section 5.30).

5.50 FINITE ELEMENT ANALYSES

-

Came Studies

Patrametric studies by finite element analyses are powerful

tools to examine qualitatively the effects of wall stiffness and soil

conditions: on strut loads.

This section presents a brief summary. Details are contained in

Volume II (De sign Fundamentals). Four soil conditions were analyzed:

Ca,se la. Homogeneous soil profile of soft, normally consolidated

clays.

Case lb. Homogeneous soil profile of medium-stiff clay.

Case 2. A soft soil stratum to base of excavation underlain

by a stiff stratum.

Case 3. A soft soil layer underlying a more competent stiff soil.

-67-
Distribution of Earth Pressures

Figures 34 and 35 show normalized apparent earth pressure

diagrams predicted by the finite element analysis for the four soil

conditions .

Cornparing Case la with Case lb, the analysis shows that walls

in the soft clay are expected to experience relatively higher pressures

near the base of the cut than the wall in the medium-stiff clay. This

trend is more obvious for the stiffer concrete walls. As Case lb shows,

this behavior becomes Iess pronounced as the soil becomes stiffer, One

possible remedy for reducing this effect in soft soils would be to prestress

the second lowest strut and lock in a high residual compressive force,

On Figure 35, Case 2 (soft clay overlying stiff clay) shows

an opposite effect to that experienced in the homogeneous soil mass.

This stiffer layer provides an adequate reation for the wall, restricting its inward deflection in the overlying soft clay. This leads to a

larger stmt load in the second to last strut and a reduction in the

load received by the lowest strut. This results because the wall has

already deflected inward close to its maximum amount before the

last strut is installed and final excavation completed. Therefore, this

last excavation stage results in little load transfer to the lowest strut.

Stiffer wal.ls push the center of gravity higher (Case 2, right panel).

Fo:r Case 3, where the soils within the depth of cut are

stiff (N< 411 and soft soils exist immediately below the base of the

excavation, the results show that the strut loads are greatest in the

lower two struts. This occurs for the same reasons given for Case

la, that is, lack of support below the excavation base. For this soil

profile, the pattern of pressure distribution appears independent of

wall rigidity since both give essentially the same normalized pressure

diagram.

Magnitude of Strut Loads

Figure 36 shows the magnitudes of the predicted loads for Cases

2 and 3. 1:n both cases, the diaphragm wall receives much greater

apparent pressures, on the order of 2 to 4 times that of the more

flexible PZ-38 steel sheeting.

The higher apparent pressures in the concrete wall are attributed to snnaller lateral deformations, hence, less mobilization of

shear strength in the soil adjacent to the wall. This behavior is

particular1.y acute in the heavily overconsolidated soils such as those

assumed for Case 3.

-68-
P/P max

O&

4’

A-h

I

Pmax = 121K

PZ-38 SHEETlNG

P’P max

Oo:5O

LA A

LB

;I- ‘;

B+ ‘B

C

/e

STRUT LOAD P

Pe= LB

mox = 20sK

48” CONC. WALL

N’ 6.8 AT BASE OF CUT

p’pmax

CASE la - HOMOGENIOUS SOFT CLAY

P’prnax

0

l---=-GO Oo;s

1x1 /A\

A-ft A++

B-k e-‘w

C-f+ CYf+

1

D--f, Dt,

E-f+ E*

i

- pmx=26K = MAXIMUM STRUT LOAD ‘51 Pmax ~61~

PZ -38

48” CONC. WALL

N = 4.8 AT BASE OF CUT

CASE lb - HOMOGENIOUS MEDIUM-STIFF CLAY

Figure 34. Normalized apparent earth pressure

diagrams predicted by finite element analysis.

-69-
P/P mox

0

+--z--A”

r -

P/P max

0

a0

- r I I /&

*g?

LA LAB

2

B L

i FJ

B PB

B $g- -,:

C

2

C-+

D D--H

I

E E++

/p ’

Pmox = 5oK STRUT LOAD

/#(A

PB= P

LB

max s 73K

PZ-38 SHEETING 48”CONC. WALL

tj= 2.3 AT BASE OF CUT

CASE 2 - SOFT OVERLYING STIFF CLAY

o+--=4° Oq.5

,

/o /7+

AT+ A+p

B-h B-H

CT+ C-f+

D-h* D-p+

E-H E-+b

/& I

Pmox = 17K = MAXIMUM STRUT LOAD

/h\

Pmax = 62K

PZ -38 48” CONC. WALL

N= 6.4 AT BASE OF CUT

CASE 3 - STIFF OVERLYING SOFT CLAY

Figure 35. Normalized apparent earth pressure diagrams

predicted by finite element analysis.

-7o-
APPARENT LATERAL PRESSURE ( psf)

0’ I 2,000 , I 4,000, I 6,000 1 0woo

A

\

\

B \

C

Es

I

D

I

E

_ft

I

I

t+

-roMJti+f

PZ -:38 WALL I-- l.OK,IIH 4

48”CONC. WALL

AT BOTTOM OF CUT: N= 2.3

CASE.2 (SOFT’OVERLYING STIFF CLAY)

I ’ , I I 1 1

0 2,000 4,000 6,000

I ’ 1 1

0

I ;

Epoo 4&o 6@00

\

A--

\

_s

DESIGN PRESSURE DIAGRAMS \

FROM FIGURE 28. A---+ ’

I

\

B- 0-m

I

1

1

C- I C--+ I

ID e

I

I

Dj 1.

I

/’

/

E-+ E-c 1

1’

i/

7//F/N/ +0.4q 7//m77’

e 0.4 #Hj

PZ-38 WALL 48”CONC. WALL

NOTE: SEE TEXT FOR DISCUSSION

AT BOTTOM OF CUT: N = 6.4 OF LOADS.

CASE. 3 (STIFF OVERLYING SOFT CLAY 1

Figure 36. Comparison of predicted apparent

earth pressures with finite element analysis on

stratified soils.

-71-
There is scant field evidence to support the trends illustrated

by the finite element analyses. One cannot accept these implications

literally, but nevertheless, pending further advances in the state of

the art, they are a matter of concern. Therefore, when overconsolidated

soils are present, one should be aware that loads may build up on the

support system causing overloading, especially if a relatively rigid

wall is used which restricts the lateral swelling of the soil.

The Use of the Finite Element Method in Design

The finite element method (FEM) is a more realistic mathematical modeling of the complex soil profile and the soil structure system,

thus making it a powerful tool in the analysis of supported excavations.

However, it should be used with great care and only by experienced

engineers with a strong soil mechanics background.

Until substantially more experience is gained with FEM as a

design tool, it should be used primarily as an aid to guide engineering

judgement.

-72-
CHAPTER 6 - STABILITY ANALYSIS OF SHEETED EXCAVATIONS

6.10 GE:NERAL

The three primary modes of instability for excavations in clay

are shown in Figure 37. Bottom heave and deepseated failure (Figures

3’7a and 37b) are related to the overall stability of the excavation and

may even dictate the construction procedure.

Loc:al failures (Figure 37~) are of concern where it is necessary

to limit inward sheeting deformations. Failures of this type occur

below the excavation level immediately adjacent to the sheeting, resulting

in partial loss of lateral support.

6.20 BOTTOM HEAVE

Bottom heave is a problem primarily in soft to medium clays

where the strength of the soil is nearly constant with depth below the

base of the excavation. The failure is analogous to a bearing capacity

failure; the difference being that stress is relieved. This mode of

failure should be analyzed (Bjerru.m and Eide, 1956) using the stabi.lity

chart given in Figure 38. The factor of safety against a bottom heave

is determ,ined as:

F.S. = Ncb (

,Hs;q ) = ;cb

where:

N = stability number =

IHtq

sU

N

c b = bearing capacity factor from Figure 38

S

U

= the undrained strength of the clay

b’ = total unit weight of the soil

H = depth of excavation

q = uniform surcharge loading on the area adjacent

to the excavation

Where the soil is stratified within the depth of excavation and

below, a weighted average of undrained strength should be used for

-73-
I

I

I

I 1

H

:v

GTTLEMENT

I xm I

1

I

I

I

I

------ I

I

1 S”

,/ ‘\

BOTTOM HEAVE FAILURE- -

RESISTING STRUT

b) DEEP SEATED FAILURE

INTERNALLY BRACED

EXCAVATION

. ROTATION QcFNTER

c) DEEP SEATED FAILURE

TIED BACK

EXCAVATION

Figure 37. Potential failure surfaces.

-74-
II0

9

8

Ncb 7

6

5

H

Y

Hmax .(‘u Ncb - 9)

.

8 m

4

SQUARE@

i.

L= LENGTH OF TRENCH

(FROM BJERRUM & EIDE, 1956) -

H/B

Figure 38. Bearing capacity factors for

bottom stability analy si s.

-75-
sue This average should be taken over a zone described between -

below the excavation base and 2.5B above the base.

6.30 LOCAL FAILURE

Lateral pressure on the retaining wall coupled with the stress

relief from the excavation can be of sufficient magnitude to cause

local yielding of the soil immediately adjacent to the inside of the

sheeting, This localized overstressing results in loss of passive

resistance which in turn leads to uncontrolled inward movements of

the sheeting, often amounting to about 50 percent of total movement.

Figures 39 and 40 can be used to estimate when local failure is

imminent in cohesive soils where flexible sheeting is used. Figure 39

shows the factor of safety against bottom heave necessary to prevent

local yield as a function of excavation geometry and the shear stress

ratio.

The potential for local yielding is most prominent in the overconsolidated soils, that is, soils with a high value of Ko(Bho/8 vo),

Excavation in these soils relieves lateral stress which in turn leads

to an extension type local failure near the base. Note, for example

in Figure 39, that the ratio Ncb/Nc required to prevent local failure

will increase with increasing K,, ‘all other variables being constant,

Figure 41 shows the effect sheeting stiffness has on reducing local

yielding in normally consolidated soils. Stiffening the sheeting reduces

the factor of safety required to prevent local failure,

6.40 DEEP SEATED FAILURES

6.41 Internally Braced Excavations

Circular Arc Analysis

One way to analyze the stability is the classical circular

arc analysis as illustrated in Figure 42, This involves a series of

trial centers of rotation and failure surfaces to find the most critical

condition.

The sum of the strut forces necessary to maintain a stable

excavation should be compared to those predicted from the lateral

earth pressure diagram. The greater of the two total loads should be

used to establish the ordinate of the design earth pressure diagram.

In the cases where the retaining wall extends through

a weak layer into a highly competent soil, the structural resistance of

-76-
CURVES PREPARED FROM

RESULTS OF FINITE ELEMENT

COMPUTER PROGRAM- BRACE

-- --_- -.-.--- ---- -__-----I--

( FROM D’APPOLONIA, 1971)

-1.0 -0.5 0 +0.5 + 1.0

SHEAR STRESS RATIO, f =

I-K,

2- s,

8 ,

Figure 39. Factor of safety required to prevent

local yield below bottom of excavation in clay.

-77-
-0.5 0 to.5 + 1.0 t1.s

SHEAR STRESS RATIO, f

Figure 40. Shear stress ratio vs. overconsolidation ratio.

-78 -
d

0 9

Niq’N :A13jVS JO tlO&d

-79-
BRACED WITH EITHER

CROSS LOT OR RAKED

-COMPETENT

SOIL LAYER

Consider overall stability:

Moments around center of rotation

Forces to consider:

1)

2)

of

3)

4)

Weight of driving mass (WT)

Resisting strut loads (P,,

support load.)

P2) (Horizontal component

Resisting shear capacity of wall (Hs) from competent soil layer.

Shear #strength of soil, frictional component (T). and

Cohesion, (c)

Note: If rakers used, kicker must be located outside failure

mass for PI and P

2

to be considered in analysis.

Z”R (N Safety Factor = x f w a

tan @I t CL) R

D

T w - P1ll - P212 - HsR

Figure 42. Stability of internally braced cut (circular arc method).

-8O-
the retaining wail (Hs) should be considered in the analysis. The soil

shear resistance should be taken equal to the passive force determined

in accordance with Chapter 4, (Passive Resistance).

Wedge Stability AnaIysis

Figure 43 shows a wedge stability analysis based upon

pIanar failure surfaces. Like the circular arc method, this involves

a series of trial planes to find the most critical failure surface.

6..42 Tied-Back Walls

Detailed procedures for analyzing the stability of tiedback walls by a variety of methods employing trial planar surfaces and

wedges are presented in Volume III. By and large, these methods place

emphasis on failure surfaces passing through the zone of tiebacks.

As such, the techniques may be used as a design tool for establishing

the appropriate length of tiebacks.

-81-
1: ACTIVE WEDGES PASSIVE

1. WGES .J

Stratum III

For general solution varya(, 6, and&angles to obtain minimum value

for factor of safety.

Method of Analysis:

1. Assume a<, p,6; angles.

2. Sequentially analyze the active and passive segment for loads

pIII and P V’ Include water pressure.

3. Sum forces in horizontal direction for factor of safety

i.e. F.S. =

Pl +P 2

pIII - p v

P

wII

= horizontal water pressu

uplift water force on Wedge II

re

Typical Analysis of Wedge

(Wedge II)

Figure 43. Wedge stability analysis for braced cut.

-82-
CHAPTER 7 - BEARING PRESSURE OF DEEP FOUNDATIONS

7.10 GENERAL

This section is directed toward those basic considerations used

to establish bearing values for elements involved in cut-and-cover

operations. Typically, the bearing stratum is deep--that is, it lies

at great depth relative to the width of the bearing area. A ccor dingly,

design rules developed for shallow foundations will be overly conservative.

Fundamentally, a.llowable bearing value must recognize two

governing criteria --first, adequate safety against shear failure of the

foundation and second, a limitation of settlement. UsuaLly, as will be

apparent further in this discussion, it is the former which controls

for clays and it is the latter which controls for sands.

7.20 PRESUMPTIVE BEARING VALUE

Tab.le 2 presents a summary of the range of allowable bearing

values for building foundations resting on a variety of soil types. This

tabulation is not intended to represent a recommendation for design

but rather to aid in assessing the relative competency of different materials

and to provide a crude initial guide.

apply to shallow foundations

Because the values typically

, acceptable values for deep foundations

will be somewhat higher.

7.30 BEARING VALUES BASED ON SHEAR FAILURE

7.31 General

The following represents a summary of theoretical

procedures for calculating net ultimate bearing capacity using shear

strength parameters, 8, o co f h esionless soil, and undrained shear

strength, Su, of cohesive soils. A factor of safety of 2 to 3 should

be applied depending upon risk and confidence leve.1 in data.

7. 32 Sand

For deep piers in sand the end bearing load capacity is

generally expressed as:

-83-
Table 2. Abstract of presumptive bearing capacity, ksf.

A B C D E

Mass. State New York Atlanta National Board BOCA

Code (1974) City (1968) (1950) of Fire Under- (1970)

writers (1955)

Glacial Till’:c

Hardpan*

Gravel, wellgraded sand and

gravel::’

Coarse sand+

Medium sand::

Fine sand

Hard clay

Stiff clay

Medium clay

20

- -

10

6

4

2 - 4

10

- -

2

16 - 24 20 20 20

8 - 20 8 -12l 8 - 12l 8 - 121

6 - 122

--

4- 83

10

--

4

-- -- --

--

- -

- -

- -

4

- -

6- 81 6- 8l

- - 4(100 se)

4- 61 - -

- - - -

5 - -

5 - -

Massachusetts and New York Code allow 5 percent increase in bearing value per foot of additional

embedment, but not more than twice tabulated value.

1 - Range reflects compactness, gradation, and/or silt content

2 - 0. 1 x N, but not less than 6 ksf nor more than 12 ksf (where N= no. of blows in SPT)

3 - 0. 1 x N, but not less than 4 ksf nor more than 8 ksf (where N= no. of blows in SPT)
where:

N

Q

= dimensionless bearing capacity factor that is a

function of the shear strength parameter, 3, of

the bearing material and shape of the loaded area

q = effective stress in the soil at the bearing surface

au = ultimate bearing capacity (load per unit area)

derivation.

Values of Nq vary depending upon assumptions made in the

Vesic (1965) presents ranges for the values as shown in

the Figure 44. l[n general, a safety factor of 3 is applied to these ultimate

values.

7. 33 Clay

In clays the undrained strength, S,, rather than drained

strength will control the bearing capacity of a foundation element.

Skempton (1951) presents bearing capacity factors NC for net ultimate

bearing capacity on clays. In this case, “net” means pressure in

excess of the effective overburn stress on the bearing level.

where:

au

= net ultimate bearing capacity (.load per unit area).

NC

= dimensionless bearing capacity factor that is a

function of the shape of the loaded area.

S

U

= undrained shear strength of soil.

For deep foundations (at depth greater than 4 to 5 times

the breadth of the loaded area), values of NC are as follows:

Circle: N = 9

C

Strip:

NC

= 7.5

Rectangle: NC = 7.5 (ltO.ZB/L)

where: B = breadth

L = length

-85-
2Y 30° 350 400 450 500

10

ANGLE OF INTERNAL FRICTION 4

FROM VESIC (1965)

Figure 44 Bearing capacity factors for deep

circular foundations.

-86-
Note that for clays the net ultimate bearing pressure is

independent of depth (and therefore overburden stress). It is a function

only of the shape of the loaded area and undrained shear strength of the

soil.

In addition to the load bearing capacity at the base, the

side friction may be determined on the basis of the embedded area and

adhesion along the shaft. In soft clays, the adhesion is equal to or

only slightly less than the undrained shear strength. However, in

stiff to hard clays the adhesion is typically less than one-half the

undrained strength.

The practice is to apply a reduction factor, d , to the

undrained strength to estimate adhesion. Thus :

S

eff =

&S

U

where:

d = reduction factor

S

U

= undrained shear strength, psf

S

eff = adhesion along shaft, psf

Figure 45 (after Peck, et al, 1974) shows that cx decreases as the shear strength of clay increases.

7.40 BEARING VALUES BASED ON SETTLEMENT

7. 4Ll Soils Havinp Constant Modulus of Deformation with Depth

Surface Loading

Theoretical procedures for determination of settlements

have been ‘developed based on integration of the Mindlin solution for

a point LoaId within an elastic half space. At a depth equal to zero,

the Mindlin solution is identical to the familiar Boussinesq solution.

These solutions a.ll have the general form

P

BIP =q E (1 - 1) 2, Eq. 7.41.1

-87-
0.6 -

d

AA

1.5

AVERAGE

0(x S”

isf

I

FROM PECK, HANSON 8 THORNBURN

(1974)

Figure 45. R.eduction factor in S from observed

capacity of friction pile\.

-88-
where:

P

= settlement

q = distributed load

B = least dimension of foundation unit

E = modulus of deformation

0 = Poisson’s Ratio

Ip = influence factor which depends on rigidity of

footing, shape of footing, and depth of footing

A simplified method for determining settle.ment at

the surface is based upon a coefficient of subgrade reaction, defined as

fo.ll.ows :

P

=“+-- Eq. 7.41.2

where:

P

and q are defined as above

k = coefficient of subgrade reaction in general units

of pressure per unit deflection

The value of the coefficient of subgrade reaction is

commonly determined by plate loading tests or by correlation with in

situ soil indices such as relative density and standard penetration

resistance. By comparison of Eq. 7.41.L and 7.41.. 2, the coefficient

of subgrade reaction is related to the theoretical settlement as follows:

k =

E

B (I-V

Eq. 7.41.3

For a constant footing shape and depth and constant

material properties, the coefficient of subgrade reaction for a footing

of size B is therefore related to a footing of size B’ as follows:

k

B’ = kg ( Eq. 7.41.4

-89-
It is common to express the coefficient of subgrade

reaction in terms of the value for a 1 foot square plate (kl) as this is

the size for conventional plate loading tests. Therefore,

kl kg = 7

Typical values for kl are shown in Figure 46.

Rectangular Footins

Terzaghi (1955) has proposed the following empirical

relationship for rectangular footings :

k = kg (1 t 0.5 B/L)

LxB 1.5 Eq. 7.41.5

where:

k

LxB

= coefficient of subgrade reaction for footings

of length, L, and width, B

kB = coeffi.cient of subgrade reaction for square

footing of dimension, B

See Figure 47 for comparison of Terzaghi’s empirical

equation and elastic theory. Terzaghi’s equation is recommended.

Effect of Depth

For a footing with constant loading, shape, and

material properties, the subgrade modulus of that footing is inversely

proportional to the influence factor (see Eq. 7.41.3). Thus, when

the influence factor varies with depth, the ratio of subgrade modulus

at the surface to the subgrade modulus at depth may be computed as

foLlows:

7 =

B

where:

Eq. 7.41.6

coefficient of subgrade reaction for a footing

(breadth B) at the surface

-9o-
FINE GRAINED SOIL

STIFF I VERY STIFF

?

I

2 I { 3 I

“NCONFlINED COMPRESSIVE STRENGTH

L

G3

200

E

z

150

at100

50

k, =COEFFlClENT OF SUBGRADE

REACTION FOR I FOOT

SQUARE BEARING PLATE

AT GROUND SURFACE.

COARSE GRAINED SOIL

Figure 46. Coefficient of subgrade reaction vs.

in situ soil indices (NAVFAC, 1971).

-91-
1.0

0.9

0.8

0.7

.v) 0.6

IL

0.2

0.1

0

FOR A CIRCLE OF

DIAMETER B, 5 ‘0.89

LENGTH TO

BREADTH RATIO

VS

SHAPE FACTOR

t

TERZAGHI (1955)

ELASTIC THEORY

K

Lx8

= COEFFICIENT OF SUBGRADE REACTION FOR

RECTANGULAR FOOTING OF WIDTH B, LENGTH L.

KS = COEFFICIENT OF SUBGRADE REACTION FOR

SOUARE FOOTING.

I 1.0

L/e

IO.0 100

LENGTH To BREADTH RATIO

Figure 47. Shape factor for rectangular footings.
kD

B =

coefficient of subgrade reaction for a footing

(breadth B) at depth D

= influence factor for footing at depth D

$ = influence factor for footing at surface

Elastic theory demonstrates that footings will undergo Less

settlement with depth. This is illustrated in Figure 48 which shows that

the subgrade modulus increases with depth below the surface.

7.42 Soils Havink Modulus of Deformation that Increases

with Depth

Surface Loading

Terzaghi (1955) has proposed the following empirical

relationship to convert the coefficient of subgrade reaction for a

1 foot square area to an area B x B square.

2

kB = kl( B2;1 1 Eq. 7.42. 1

Once k

%

is determined for a square footing, the coefficient

of subgrade reaction,

from Figure 47,

L x B, for a rectangular footing may be obtained

Depth Effects

Taylor (1948) has proposed an embedment correction to

account for the increase in modulus of deformation with depth as

follows :

kDB= k; (1 t 2 D/B) Eq. 7.42. 2

where:

D is the depth of footing

B is the minimum footing dimension

A second approach is based on Janbu (1963) who

demonstrated that the initial tangent modulus increases as a power

function of confining stress.

-93-
1.0

.95

.90

.85

I .&I

ii0 .70

x” .65

.60

.55

-50

DEPTH RATIO

VS

DEPTH FACTOR I

’ I I I \I I I I

0 0.2 0.4 0.6 0.8 1.0 0.8 Q6 ’ 0.4 0.2 0

I D I l

l - KS= COEl=rlClENT $F SUBGRADE REACTION AT SURFACE.

l KD=

I I I I ” DEPTH, D

. LxB = FOOTING DIMENSIONS

RECTANGULAR AREAS FROM FOX ( 1948

CIRCULAR AREAS FROM WOODWARD et

11

al (1972)

NOTE: THIS FIGURE BASER UPON THEORY WITH

CONSTANT MODULUS OF ELASTICITY

(DEFORMATIONJ

Figure 48. Influence of depth on coefficient of subgrade reaction (based on

modulus of deformation that is constant with depth).
E it is proportional to ( 83)”

where:

E

it

= initial tangent modutus

U3 = lateral effective stress

n = 0.3 for gravel and 0.5 for sands

Fro,m the assumption that k is proportional. to Eit:

ksB

kDB

f

-s n

d3 [ 1iT3D

=

FDG Eq. 7.42. 3

In normally consolidated deposits, e3, is proportional

to the overburden stress and therefore the depth.

The Taylor and Janbu methods for determining depth

effects in soils with varying modulus of deformation are presented in

Figure 49. Note that a limitation of FDG q 0.5 has been set for the

Taylor expression.

Water Table Effects

The presence of ground water in granular soils will

effect the modulus of deformation by reducing the lateral effective

stress. The plots in Figure 49 would therefore require corrections

based upon reduction of effective stress level from submergence.

7.43 Recommended Procedure ‘for Determination of Settlements

of Deep Foundations

Clays

Assume modulus of deformation is constant with depth.

Compute settlement for Eq. 7.41.2

P = q/k

-95-
0.9

fn P

YY

0.8

i 0 7.

DEPTH RATIO

DEPTH FACTOR

BASED ON JANBU(l963)

BASED ON

JANBU(1963)

K S = COEFFICIENT OF SUBGRADE REACTION AT SURFACE.

KD = COEFFICIENT OF SUBGRADE REACTION AT DEPTH, D

1.0

Figure 49. Influence of depth on coefficient of subgrade reaction

for granular soils (based on modulus of deformation that increases with depth).
where:

q = load in tsf

k = coefficient of subgrade reaction in tsf/ft

P

= settlement in feet

Determine k by first determining kl at the surface of

the soil from Figure 46 or from plate load test. Modify kl as follows:

k =

kl FS

B (F ) D

where:

FS = shape factor from Figure 47

F

D

= depth factor from Figure 48

B = least dimension of bearing area in feet

Sands

Assume modulus of deformation increases with depth.

Compute settlement from Eq. 7.41.2 as above. Determine k by first

determining kl as above. Modify kl as follows to account for size,

shape, and depth.

kl x FS Btl 2

k =

FD x FDG 2B

where:

FS’ FD and B as defined above

FDG =

depth factor for granular soil from Figure 49

-97-
CHAPTER 8 - OVERVIEW OF CONSTRUCTION METHODS

8.10 PURPOSE AND SCOPE

This section is a synthesis of the main conclusions concerning

the performance of underpinning and of various techniques for supporting

open excavations. The general applicability of each of the various

techniques is discussed, and comparisons are made, when appropriate,

in evaluating the influence of such variables as soil type, wall type,

and method of lateral support. An attempt has been made to identify

key operational contingencies which may contribute to excessive

horizontal and vertical displacexnents in the adjacent ground. Finally,

some general guidelines are provided on costs.

8.20 GENERAL CONCLUSIONS CONCERNING DISPLACEMENTS

8.21 Lateral Support Methods

“Competent Soils” (granular soils, very stiff clays, etc.)

a . For these soil types the displacements reported in the

literature on well- constructed, well-documented projects are of insufficient magnitude to distinguish variations that may be caused by

wall type or method of lateral support. Nevertheless, there is strong

evidence to suggest that use of concrete diaphragm wal.ls will result

in .less displacement than other wall types and some evidence that walls

supported by tiebacks will perform better than internally braced walls

b. Maximum displacements are typically 0.25 percent to 0.35

percent of wall height. The lower range is associated with granular

soils; the upper range is associated with cohesive soils.

c. Typically, maximum horizontal and vertical displacements are about equal.

“Weaker Soils” (soft to medium clays, organic soils, etc. ),

d. Maximum displacements typically exceed 1 percent of

the depth of the cut for flexible wa.lls. The use of concrete diaphragm

walls reduces the magnitude of displacements’to about 0.25 percent

of the depth of the cut -- or about the same as those observed for

competent soils.

e. Typically, the maximum vertical displacements exceed

maximum horizontal displacements.

-98-
f. When the excavation is in deep deposits of weak

soils, the cumulative total of all displacements occurring below the

last placed strut Level amounts to about 60 percent of the total

measured movement.

“Wall Type”

!iz* With concrete diaphragm walls, displacements are

typically less than 0.25 percent of wall height, regardless of soil type.

h, Wall stiffness can be increased by using rigid concrete walls or by reducing spacing between support levels. It is believed that

comparable watt stiffness (defined as ET-) will result in comparable

L 4

performance provided that the installations are carefully performed.

i. A comparison from observational data between soldier

pile walls and sheet pile walls (of comparable stiffness) is not possible

in very stiff to hard clays and dense granular soils because comparative

data are not available. Sheet pile walls are rarely used in these soil

types because of the hard driving conditions.

Effect of Wall Stiffness in Cohesive Soil

j. The influence of wall stiffness and of stability number

of cohesive soil (defined as N =

)IH- ) was exa.mined in some detail.

%

The data show increasing displacementswith weaker soils and with more

flexible walls. Displacements with sheet piling may exceed 4 to 5

inches, but in similar cases diaphragm walls would control displacements to Less than 1 -l/2 inches.

8.22 Underpinning

The underpinning process has an inherent source

of deformation resulting from the transfer of Load from the existing

foundation to the new foundation. Well-executed construction procedures

can normally control this vertical displacement to l/2 inch or less.

The underpinning elements may also be influenced

by the adjacent excavation because the underpinning elements will be

installed within the zone of vertical and horizontal displacements.

This *may result in additional displacements and/or additional Load on

the underpinning elements, Experience has shown that horizontal

movements cause more damage than vertical movements.

-99-
8.30 WALL TYPE

8. 31 Concrete Diaphragm Walls

8, 31. 1 Applicability

Diaphragm walls can be used in virtually every

soil condition with the possible exception of very soft clays, peat, or

cohesive hydraulic fi.ll. They are used frequently to minimize displacements behind the wall. It is common in European practice to incorporate

the diaphragm wall into the permanent structure, whereas in the United

States diaphragm walls have generally been used as a method of

ground support without being incorporated into the permanent structure.

8.31.2 Operational Considerations

Soil and water conditions can adversely affect

diaphr a grn wall cons tr u&ion. Of particular concern are excavations

in very pervious soils (fluid loss), contamination of the fluid (adverse

pH, high salinity, or high calcium content), and spalling of the trench

wall. Spalling of the trench wall may be caused by unstable soils or

loose fill, particularly when containing miscellaneous rubble or old

foundations. It is believed that most of the problems can be identified

during initial investigation and controlled during construction.

8. 32 Soldier Pile Walls

8. 32. 1 Applicability

Soldier piles can be used in all soils except

perhaps soft to medium clays and loose or soft dilatant soils of low

plasticity below the water table. These latter soils have a tendency

to run after exposure.

8. 32.2 Operational Considerations

The following items may cause displacements:

deflection of lagging, overcut behind lagging, ground loss caused by

surface and ground water flow, and ground loss associated with preexcavation for soldier piles. Broken water mains or flooding may

cause heavy water flow toward the excavation. This water flow is an

additional risk in soldier pile walls.

Pre-draining of saturated soils is essential,

especially those which may have a tendency to run (silt or silty fine

-lOO-
sand for example). A common, difficult situation is when such soils

are underlain by rock or by impervious soils within the depth of the

excavation. This sequence makes it extremely difficult to fully dewater

to the lowest extent of the water-bearing formation.

8. 33 Steel Sheet Pile Walls

8.33. 1 Applicability

Sheet pile walls are most generally used in soil

types that are inappropriate for soldier pile walls, such as soft cLays,

organic soiLs, and dilatant soils of Low plasticity. Sheeting is also

used in situations where there is a desire to cut off ground water or to

reduce seepage gradients at the bottom of the excavation.

8.33.2 Operational Considerations

Steel sheet pile walls are relatively flexible with

normal wale spacing, and they are frequently subject to relatively

large displacements when installed in weak cohesive soils.

Tearing of interlocks under hard driving conditions may cause ground Loss because of ground water infiltration through

the torn inter Locks .

While interlocked steel sheet piling effectively

intercepts ground water flow within pervious layers, the piezometric

level outside the excavation will often be depressed in impervious soil

strata. The presence of the interlocked steel sheet pile wall does

not prevent a seepage pattern to the face of the excavation. Such a

seepage pattern is accompanied by a drop in piezometric levels which

may induce consolidation of compressible soils. Removal. of steel

sheet piling from cohesive soils may also remove soils with it and lead

to settlement of adjacent ground.

8.40 SUPPORT METHOD

8.41 Tiebacks

8. 41. 1 ApplicabiLity

Tiebacks are most applicable in very stiff to hard

cohesive soils or in granular soils. In Lower shear strength, cohesive soils

the regroutable tieback has been used successfully while other anchor types

often experience relatively Large movements.

-lOl-
.I

8.41.2 OperationaL Considerations

A number of operational contingencies are listed

and commented on below.

Vertical Wall Movement

The vertical components of Load may cause settlement of soldier pile walls, and this may Lead to horizontal wall displacement.

Excessive Prestressing

With a relatively flexible wall, excessive prestressing of the upper Levels may cause inward movement of the top

and outward bowing below. The magnitude of the bowing increases

in response to excavation as the restraining force is removed on the

inside of the wall. The problem is accentuated in a soil sequence

of Loose - hard - Loose from the top.

Water Flow and Ground Loss into Drill Holes

Water flow through the drilled anchorage can

result in ground Loss, particularly inLoose fine sand. The magnitude

of the ground Loss is affected by the hydrostatic head, dril.Ling procedur es, and soil conditions. Water flow alone may Lead to a drop i-n piesometric Level and consolidation of compressibles.

Lateral Creep

Lateral movement several times greater than

settlement and extending relatively Large distances behind the face of

the excavation, has been reported in highly over consolidated cLays and

soft shares. The movement is believed to be associated with Lateral

expansion following stress relief from the excavation.

Another potential source of Later al creep is in

the presence of a weak Layer of cohesive soil be.Low the excavation.

8.42 Internal Bracing

8.42. 1 Applicability

Internal bracing is most applicable to situations

where a reasonably economical section can be used without need of

-102-
intermediate support, As the distance between the sides of the excavation increases, internal bracing becomes less efficient, and therefore

tiebacks become more attractive. In some cases inclined rakers are

economical alternatives.

8.42.2 0perationa.l Considerations

The most important contingency is believed to

be the connection details, especially alignment of members and welding.

Displacements may occur from slack in the support system (consisting of axial compression of the member, deformations in connections, bearing between wale and waLL and the adjoining

ground) . However, this can be Largely eliminated by preloading.

Brace removal is another source of displacement.

However, this can be controlled by a combination of well planned

restrutting and effective compaction of backfill between the wall and

the structure,

Preloading to about 50 percent of the design load

is common practice in areas where displacements are of concern.

Extreme temperature variations affect Load.

Reasonable precautions to prevent overstressing can be taken by covering steel members or by painting them with reflective silver paint.

8.50 UNDERPINNING

8. 51 Applicability

Underpinning elements transfer the load from an existing

foundation to a new foundation bearing below the zone of influence of

the adjacent excavation. The decision to underpinn a structure is

based on several factors including the cost of underpinning, the cost of

alternatives, expediency, and risk.

8. 52 Operational Considerations

A thorough study of the structure to be underpinned

should be made to determine load and Load distribution. Temporary

conditions that occur during underpinning will also require evaluation.

Because the elements pass through a zone undergoing vertical

and horizontal displacement, underpinning elements may be subject

to downdrag forces, lateral forces, and/or movement. Lateral movements have been a source of great damage.

-103-
A number of factors may cause ground Loss. Lagged

underpinning pits for construction of piers have many of the same

contingencies mentioned previously for soldier pile walls. The potential

for ground loss also exists when “blow conditions” develop in open

shafts or open-ended piles below gound water table,

8.60 STABILIZATION METHODS

8. 61 Scope

This section is a brief overview of grouting and freezing.

These methods are used to control ground water or to solidify a soil

mass. Applications may be to create an “arch*’ over a tunnel or

around a shaft or to solidify potentially unstable soils and badly jointed

rock encountered within the excavation.

Both methods are an “art” performed by specialty sub -

contractors, often with proprietary equipment or material. Details

of techniques are not highly publicized, although successful results

of applications are.

Performance type specifications are believed to be the

appropriate contracting procedure for both grouting and freezing.

8. 62 Grouting

Basic soil classification, particularly grain size ctlaracteristics, is essential for selecting the type of grout and planning ttle

grouting program. The 15 percent size of soil to be grouted if. commonly

used as a criterion for grout selection,

Less expensive grouts (cement and bentonite) are used

in coarse sands and gravels. Silicates may be used in fine to medium

sands. The most expensive grouts are the organic grouts, which are

used for fine sands and coarse silts. IIn stratified deposits multistage grouting consists of grouting with cement or bentonite to reduce

the permeability of relatively coarse soils followed by successive

stages of finer grouts and/or less viscous chemical grouts to penetrate

more fine - gr ained soils.

8. 63 Ground Freezing

Ground freezing methods have been used primarily in

conjunction with shafts and small diameter tunnels. Frequently, it has

-104-
been used in difficult ground water situations where more conventional methods have failed or are inadequate. However, the use of

ground freezing as a primary construction method is increasing.

Creep characteristics of the frozen soil are of interest

in deep shafts or tunnels. Creep is related to the stability of the ice

structure and displacements outside the frozen zone.

8.70 SOIL AND GROUND WATER CONDITIONS

The following is a brief check list of soil conditions that

may contribute to additional displacement. Some of these were mentioned

above.

1. Drawdown of ground water table: Ground settlement will

occur if compressible soils are present.

2. Soft shale and highly over consolidated clay: These

may display Lateral creep in tieback installations or may contribute

toward load buildup in braced excavations.

3. Rock within cut: A number of potential probiems exist:

a. Undermining of support wall from rock falls;

b. Over -blasting below and behind wall;

c. Difficulty in controlling flow at rock-soil

contact or through joints;

d. Inadequate toe restraint for soldier piles;

e. Inability to completely dewater overlying soils

to the top of the rock:

f. Ground water flow through highly jointed zones

in the rock: This may depress the ground

water table and/or carry fines.

(For further discussion see White, 1974a).

4. Pervious soils underlain by impervious soil within

This the dpthwill of the excavation: make it difficult to completely

dewater to the bottom of pervious formations. This concern is most

relevant to soldier pile walls.

5. Soft clay below excavation: Deformation character -

istics of soil (“elastic” range) will cause flexure of the wall below the

-105-
bottom of the excavation at intermediate stages and at final depths.

These uncontrolled displacements represent about 60 percent of the

total.

In deep excavations, the imbalance created by Load removal causes excessive shear strains!in the “plastic” range of stresses.

stresses.

6. Seenage: Seepage at the toe will weaken passive

restraint and/or cause ground flow into the excavation.

8. 80 COSTS

8. 81 Purpose and Scope

This section is intended to provide some general guidelines to enable engineers to make a “first pass” approximation of

costs or to compare alternate schemes. ObviousLy, these cost guidelines

are not precise, and they will vary by geographic area and job conditions.

Costs have been developed on the basis of 1975 prices

and Labor conditions prevailing in the urban northeast.

8. 82 Walls

Cost per Sq. Ft. (Typical Conditions)

Exposed Exposed with*

Wall only ALLowance for Toe

Soldier Piles and

Wood Lagging $4 to $7

Steel Sheet Piling

PZ-27

PZ-38

$6 to $7 $8 to $9

$8 to $9 $10 to $11

Concrete Diaphragm

Tangent Pile

(single row)

Cast-in-place Slurry1

WaLL (30” f thick)

$15 to $18 $19 to $23

$20 to $35 $31 to $44

*c When applied to the exposed portion of the wall, this includes carrying

the toe penetration to about 25 percent of exposed wall height below the

bottom of the excavation.

(1) Price variation is relatively insensitive to variations in wall

thickness in the range of 2 to 3 feet thick. Difficult excavation in

hard materials (till, boulders, weathered rock) will raise costs

to from $40 to $,60 per sq. ft. (Tamaro, 1975).
8. 83 Supported Walls

The following discussion presents costs of walls supported

with tiebacks or bracing. The upper and lower limits of each do not

represent corresponding situations and therefore do not represent the

cost differential between the two support methods. In general, tie -

backs are slightly more costly; however, many situations exist where

tiebacks are Less costly. Two examples are: rock within the excavation and a wide excavation, such as at a station.

8.83.1 Tiebacks

Typical tieback costs of small diameter (4 - 6 inches

usually percussion drilled) and large diameter anchors (12 - 18

inches f, usually installed with auger equipment) do not vary greatly.

The applicability of one type or the other will generally depend upon

soil conditions.

Total cost of tiebacks, including installation

and prestressing, is summarized below.

Easy job conditions $15 to $20 per lineal foot

Average job conditions $20 to $25 per lineal foot

Difficult job conditions $25 to $30 per lineal foot

Assuming average tieback lengths of about 50

feet long at $20 to $25 per foot, this represents a cost of $1000 to

$1250 each.

Costs for installed walls, supported by tiebacks

including the wale and connections, are as follows:

Cost per Square Foot1

Depth Soldier Piles and2 titer locked 2

(feet) Wood Lagging Sheet Piles

30 - 40 $17 to $22 $20 to $27

40 - 50 $21 to $26 $25 to $32

50 - 60 $24 to $30 $30 to $40

60 - 70 $30 to $40 $35 to $45

1 When applied to the exposed portion of the wall, this includes toe penetration to about 25 percent of the exposed wall height below the bottom of

the excavation.

2Water pressure is assumed to act on the sheeting, but is absent from

the soldier piles.

-107-
8,83,2, Internal Bracing

Costs for internally braced walls, including wale

and connections, are as follows:

Cost per

Soldier Piles and2

Square Foot1

Depth Interlocked’

(feet) Wood Lagging Sheet Piles

30 - 40 $15 to $20 $18 to $23

40 - 50 $20 to $25 $23 to $28

50 - 60 $25 to $30 $28 to $35

160 - 70 $30 to $40 $35 to $45

1 When applied to the exposed portion of the waL1, this includes toe

penetration to about 25 percent of the exposed wa.11 height below the

bottom of the excavation

2

Water pressure is assumed to act on the sheeting, but is absent from

the soldier piles.

8. 84 Underpinning

General guidelines are as follows:

a . Concrete Pit Underpinning

Installed cost is $275 to $350 per cubic yard of concrete.

b. Jacked Pile Underpinning

Installation cost includes cleaning out of piles.

Soft material $125 - $175 per lineal foot

Hard Materials150 - $250 per Lineal foot

c. Pali Radice

For piles 4 to 6 inches in diameter

Easy job conditions $20 to $25 per lineal. foot

Average job conditions $25 to $35 per lineal foot

Difficult job conditions $35 to $60 per lineal foot

For piles 8 to 10 inches in diameter, add about 25 percent.

-108-
8. 85 Ground Freezing

The main factors affecting costs are:

1. Geometry of excavation.

2. Earth and water pressures to be supported.

3. Amount of time available for completion of the

support system,

4, Duration of time for which the excavation is to be

held open after completion.

5. Union or non-union work rules. (Union work rules,

which demand r ound- the - clock manning of completely

automated electrically powered equipment, frequently

substantially increase the cost of ground freezing).

Installation of a cut-and-cover frozen excavation support

and ground water control system might typicaLLy range from $8 to $16

per square foot of exposed wall. Maintenance of the system during

subsequent excavation and subsurface construction might cost between

$. 20 and $. 80 per square foot of exposed wall per week. Underpinning

and tunneling costs vary too widely to allow any generalization. A s

a rule, circular, elliptical, or arch structures in which compression

rather than shear or tension stresses govern are least expensive to

construct.

8. 86 Grouting

The specialized nature of grouting work prevents an accurate

estimate of grouting costs. The cost data presented herein was obtained

from Halliburton Services (1975).

The cost of the grout materials can be accurately estimated

(cement grouts: $0.50 - $1. 301ft3; chemical grouts: $1.50 - $7. OO/ft3);

however, the installation costs are not as well known because of the

variables (time to grout, cost of equipment, etc. ). Only the grouting

contractor has an accurate idea of these costs, which will also vary

depending upon the amount of competition. Halliburton (1975) also

reports ranges in costs for final volumes of grouted soil (cement grouts:

$13.50 - $35. OO/yd3 of grouted soil; chemical grouts: $40 - $190/yd3 of

grouted soil).

-109-
CHAPTER 9 - SOLDIER PILE WALLS

9.10 INTRODUCTION

Soldier pile walls have two basic components, soldier piles

usually set at 6 to 10 foot spacings and lagging which spans the distance between the soldier piles. The soldier piles must carry the

full earth pressure load while the lagging must resist relatively

minor earth pressure loads.

Soldier piles are either installed with pile driving equipment

or are set in pre-excavated holes and then concreted in place. The

most common soldier piles are rolled steel sections, normally wide

flange or bearing pile. However, soldier piles can be almost any

structural member- -pipe section, cast-in-place concrete, or precast

concrete,

Figure 50 shows various types of steel soldier piles.

9.20 TYPES OF SOLDIER PILE WALLS

9.21 Lagging

Lagging is most commonly wood, but may also consist

of light steel sheeting, corrugated guard rail sections, or precast

concrete. Wood lagging is most commonly installed behind or in front

of the flange next to the excavation (front flange). As noted in Figure

5da, the lagging can either bear directly against the soil side (back

side) of the front flange or it can be wedged to make more intimate

contact with the soil and thus reduce associated lateral displacement.

Figure 51 shows various methods of attaching lagging

to the excavation side (front side) of the front flange. The cases

shown employ either a bolt or a T-section welded to the soldier pile

or a proprietary method known as “Contact Sheeting”. * In all cases

the vertical plate which holds the lagging can extend up over several

lagging boards so that the number of special attachments can be

minimized. One distinguishing feature of attaching lagging boards

to the front face is that the boards can run continuouslyacross

several soldier piles. This is not possible when lagging is installed behind the front flange.

Spacers between the lagging boards (called “louvres”)

allow the introduction of material for backpacking boards and filter-

*contact Sheeting, Inc. , Nyack, New York-

-llO-
(a) WF SECTION OR H-PILE SECTION

I.... . . . .

. - ., ’

. ,. .

.

-. - . -

. *. . ‘.*\’

’

^ *. a* ‘. \\\\\x\\9 .~ ..‘.: ,

.

. . ‘.‘. .

LWEDGED BEHIND FRONT FLANGE

(b) CHANNEL SECTION

LAGGING CAN

ATTACHED TO

FLANGE

ALSO BE

FRONT

LAGGING CAN ALSO BE

,,,A

ATTACHED TO FRONT

FLANGE, ADAPTABLE FOR

TIEBACKS

WEDGE

‘LANGE

Y-6. TIEBACK

e.g. WOSSERBDARRAGH

(1970)

(c) PIPE SECTION

LAGGING

TO FRONT

- WELDED BOLT. OR

ST SECTION

LAGGING

TIEBACK

e.g. DONOLO (1971)

Figure 50. Steel soldier piles.

-lll-
(a> CONTACT SHEETING

CONTACT SHEETING INCORPORATE0

( NYACK , N .Y.)

(td BOLT

BOLT PASSES BETWEEN AND PLATE HOLDS

THE TWO LEVELS OF LAGGING BOARDS.

t///./ THREADED BOLT ATTACHED BY

NELSON STUD OR RAM SET,

TOP AND BOTTOM LAGGING.

NEL SECTION HOLDS EL SECTION HOLDS

(c) SPLIT T- SECTION

SPLIT”T” WELDED TO FACE

Figure 5L. Wood lagging to front flange,

-112-
ing soil to protect against ground loss from seepage. In slow

draining ground the louvres are filled with salt hay. mis material

permits water to bleed through but also acts as a filter which prevents loss of ground (see Figure 52).

9.22 Concrete Wall

Examples of shotcrete or poured concrete walls constructed in conjunction with steel soldier piles are shown in Figure 53.

An application with precast concrete soldier piles is shown in Figure

54. In general, the typical construction procedure is to expose about

a 5-foot high section and to construct the wall by proceeding sequentially to the bottom of the excavation. In all cases the soil would have

to have sufficient cohesion to stand up while the section of the wall

is completed.

Figure55 shows precast soldier piles shaped to receive

either wood lagging or precast concrete lagging.

9.23 Soldier Pile Alone

Lagging may not be necessary in hard clays, soft shales,

or other cohesive or cemented soils if the soldier piles are spaced

sufficiently close together and adequate steps are taken to protect

against erosion and spalling of the face. Examples of this were described by Shannon and Strazer (1970) and by Clough, et al (1972) for

cases in cohesive soil in Seattle, Washington. In both cases, soldier

piles were set 3 feet on center.

Erosion or ravelling caused by drying of the exposed soil

can be inhibited by spraying the exposed soil face. Shannon and

Strazer, for example, reported use of Aerospray 52 Binder. In other

cases tarpaulins may be draped over soil to maintain moisture.

Workmen can be protected by welding wire fencing

or wire mesh to the soldier piles to prevent material from falling

into the excavation.

9.30 DESIGN CONSIDERATIONS

9.31 Soldier Piles

In addition to their function as support for lagging,

soldier piles must also develop vertical flexural strength, lateral

resistance below the level of the last strut or tieback level, and in

the case of inclined tiebacks bearing to support the vertical component

of tieback force.

-113-
PLAN

I 2-2’9 L ” LAGGING

BOARD

I

WOOD CLEAT

LOUVRE” OR

SPACE

FRONT ELEVATION

1

SOLDIER PILE

Figure 52. Louvre effect for waLL Lagging.

-114-
(a) CAST IN PLACE

L L- CONCRETE WA

STEEL W SECTION

(b) SHOTCRETE

.

:*:* *. . . . . .

., - . * * ’ :

. . . . *. ** l .í . .***... . . . .** * . - . .

Figure 53. Concrete infill between soldier piles.

-115-.
REINFORCING

PREFABRIC

ELEMENT

GROUT

EXPbSE HORIZONTAL \

SET REBARS AND FORMS

STEEL AS EXCAVATION PROCEEDS

DOWNWARD.

t

CAST-IN-PLACE

SECTION

F igure 54. Parisienne wall, precast soldier piles with

formed cast-in-place wall, (after Fenoux, 1974;

Xanthakos, 1974; and D’Appolonia, et al, 1974).

-116-
PREFABFiCATED

ELEMENT

CoNOCRRETE

Figure 55. Berlin wall, precast soldier piles with

wood or precast concrete lagging (after Fenoux, 1974).

-117-
9.32 Wood Lagging

9.32.1 Wood Materials

The most common wood used for lagging in the

United States is construction grade lumber, usually rough-cut. Structural stress-graded lumber may be specified though seldom used.

Preferred woods are Douglas Fir or Southern Yellow Pine, both of

which provide a desirable balance between flexural strength and deformation modulus.

Table 3 lists the properties of some woods that

may be used for wood lagging. The allowable flexural stress stated

in the table is for normal or repetitive use construction,

9.32.2 Arching

hrperience has shown that lagging installed in

the conventional manner in most reasonably competent soils does

not receive the total earth pressure acting on the wall. The lateral

pressure concentrates on the relatively stiff soldier piles; less pressure is applied to the more flexible lagging between the soldier piles,

This redistribution of pressure, known as arching, is inherently related to the usual manner of construction. The

lagging is supported on the front flange; a slight overcut is made

behind the lagging to facilitate placement of the boards; and the inter -

vening space behind the boards is filled with soil.

A related phenomenon is that the pressure on

lagging is relatively unaffected by depth. It therefore follows that

the greater forces associated with deeper excavations must be transmitted through soldier piles.

9.32.3 General Practice Concerninp Lagging Thickness

Lagging thickness design is based primarily

upon experience and/or empirical rules. One procedure is to vary the

amplitude of the pressure diagram with maximum pressure at the soldier pile and minimum pressure midway between the soldier pile (see

Lacroix and Jackson, 1972). Another procedure is to reduce the basic

pressure diagram used in the design of bracing and/or tiebacks by

applying a reduction factor. For example, Armento (1972), in designing lagging for the BARTD system, applied a 50 percent reduction

factor to the basic trapezoidal earth pressure diagram used for strut

design. The New York Transit Authority uses the basic pressure

-118-
Table 3. Strength properties for

typical grades of timber.

Wood Type and Grade

Allowable

Flexural Stress

fb, psi

I

Douglas Fir - Larch, surfaced

dry or surfaced green used at

max. 19% M. C.

Construction

Select Structural

1200

2050

Douglas Fir - South, surfaced

dry or surfaced green used at

max. 19% M. C.

Construction I

I

1150

Select Structural 1950

Northern Pine, surfaced at 15%

moisture content, used at 15%

max. 19% M. C.

Construction

Select Structural

1050

1750

Southern Pine, surfaced at 15%

moisture content K. D., used at

15% max. M. C.

Construction

Select Structural

1300

2250

Southern Pine, surfaced dry,

used at max. 19% M. C.

Construction 1200

I

Select Structural 2050

Modulus of

1,500,000

1,800,OOO

1,100,000

1,400,000

1,200,000

1,500,000

i

1,500,000

1,900,000

1,400,000

1,800,OOO

t

American Institute of Timber Construction, “Timber Construction

Manual”, 2nd Edition, Wiley, 1974.

-119-
diagram but allows a 50 percent increase in the allowable flexural

stress of stress graded lumber.

9.32.4 Recommended Lagging Thickness

A table of recommended thicknesses has been

developed and is presented as Table 4. Since the table has been developed on the basis of construction grade lumber, adjustments are required

for stress-graded lumber.

The so-called “competent soilsl’ shown herein

are typically either granular with relatively high angles of internal

friction or stiff to very stiff clays. Medium clays included in the table

are those with a ratio of overburden stress to undrained strength of less

than 5.

The category of “difficult soilsl’ includes loose,

granular soils with low angles of internal friction and soils having a

tendency to run when saturated, Heavily overconsolidated fissured

clays are also included because they have a tendency to expand laterally,

especially in deep excavations.

9.33 Displacements and Loss of Ground

9.33.1 General

Important factors contributing to ground loss

are the soil in zones immediately behind the lagging and the flexure

of the lagging board itself. The following discussion concerns ground

loss caused by the inherent characteristics of soldier pile walls, in

particular the techniques used in construction, The discussion does

not deal with overall deformations of the retained earth mass.

9. 33.2 Deflection of Lagging

The lagging board thicknesses recommended

in Table 4 will generally maintain deflection to less than about 1 inch.

9. 33.3 Gvercut

Movements caused by overcut are best controlled

by effective packing of soil behind lagging. The most effective way of

backpacking is to ram the soil into the space from the upperside of

each lagging board. If there is difficulty in obtaining sufficient cohesion in the material rammed in this manner and/or there is concern

with future washout from ground water action, the soil can be mixed

-120-
Table 4. Recommended thicknesses of wood lagging.

Recommended Thicknesses of

Unified Lagging (roughcut) for Clear Spans of:

Soil Description Classification Depth 5’ 6’ 7’ 8’ 9’ IO’

jilts or fine sand and silt M L

above water table SM-ML

Sands and gravels (medium GW, GP, GM, 0' to 25' 2" 3" 3" 3" 4" 4"

iense to dense). CC, SW, SP, SM -_

Clays (stiff to very stiff): CL, CH 25' to 60' 3" 3" 3" 4" 4" 5"

non-fissured.

slays, medium consis- CL, CH

tency and e < 5

SU

Sands and silty sands, SW, SP, SM

(loose).

Clayey sands (medium SC 0' to 25' 3" 3" 3" 4" 4" 5"

dense to dense) below

water table.

Clays, heavily over- CL, CH 25' to 60' 3" 3" 4" 4" 5" 5"

consolidated fissured.

Cohesionless silt or fine ML: SM-ML J

sand and silt below water

table.

Soft clays e> 5.

Slightly plastic silts

below water table.

Clayey sands (loose),

below water table.

CL, CH 0’ to 15’ 3" 3" 4" 5" _- __

M L -- 15' to 25' a" 4" 5" 6" __ __

SC _ 25' to 35’ 4" 5" 6” __ __ __

Note:

*

In the category of “potentially dangerous soils”,

use of lagging is questionable.

-121-
with cement and dry packed. Louvres also aid in backpacking. Louvres

also provide an opportunity to take remedial measures to improve

filtering or to correct for ground loss behind previously installed lagging.

V. 33.4 Inherent Soil Properties

Soft clays and loose soils of low plasticity below

the water table are of particular concern. The physical act of exposing

a face in these soils below the last placed lagging board may in itself provide the opportunity for deformation.

An example of a rather dramatic failure after

exposure of soft sensitive clay was reported by Broms and Bjerke (1973).

Examples of a German procedure for dealing with soft unstable soils

are shown in Figure 56.

The extent of stress relief caused by arching

that occurs with very soft soils and soils subject to plastic creep is

in que stion. It is recommended that the pressure used for lagging

design be determined directly from the basic pressure diagram used

for design of struts and vertical members.

A procedure for constructing walls in silts and

other soils that are difficult to drain, is to dewater in advance of exc avation. An alternative procedure would be to maintain continuously

a sloped berm from the inside face of the soldier piles and to pump

from open sumps installed at the lowest portion of the excavation.

Dry cohesionless soil may also cause difficulties, especially in hot, arid areas. Under these circumstances, one

remedial technique is to mojsten the face by spraying while placing

the lagging board. Another technique is to use a board such as plywood to hold the soil temporarily in place while setting lagging.

9.33.5 Pre-excavation for Soldier Piles

There are several potential causes of material

loss during pre-excavation:

One cause is from the suction effect that occurs

during withdrawal of the auger. One way to prevent this is to provide ports within the auger which will prohibit the suction from developing below the auger. Another is to apply pressure to the inner hole

of a hollow stem auger as it is withdrawn.

A second cause of potential ground loss is from

-122-
(a) CURVED SOIL FACE AND DOUBLE WEDGING.

(b) CONCRETE

(cl VERTICAL STEEL SHEETS IN SOFT SOIL

/- OETAlL

STEELSHEET

OETAIL

ER PILE

STEEL

SHEET

EDGE

SOLDIER PILE

’ ;’ ; ;.,’ -WALE

I

‘1 iI/, I

I

I SFCTIC)N w-v . .-..A-A . . . \

Figure 56. German techniques to prevent deformations

(after Weissenbach, 1972).

-123-
collapse of soil into the augered hole. This can be prevented by using

casing or by using a bentonite slurry suspension to stablilize the hole.

A third possible cause of ground loss is from

improperly filling the pre-excavated hole following insertion of the

soldier pile. Normally, the filling is done with lean concrete or grout.

Cases have been observed in which ground water or surface water

concentrated along improperly filled holes, flowed downward alongside

the hole, emerged from the space between lagging boards, and carried

out a significant quantity of soil.

9.33.6 Surface Water and Ground Water

In any water-bearing cohesionless formation it

is absolutely essential that the ground water be drained piror to exposing the face. The depth of the cut below the water table, the porosity and permeability of soil, and the presence of underlying or interbedded impervious layers must all be considered in devising a dewatering s theme,

In soils which drain very slowly, the excavation

face can only be advanced about one foot at a time. The bottom of the

cut is sloped in a V-shaped fashion to allow for surface drainage and

to aid in depressing the phreatic surface at the side of the excavation.

When impermeable layers are interbedded with

more pervious layers, ground water is more difficult to control. The

ground water tends to flow for a relatively long period of time just

above the impervious layer (or layers).

Protection against ground water erosion through

lagging is commonly done by a combination of effective backpacking

and placement of salt marsh hay in the open space between the lagging

boards to filter out the soil. Another way of preventing erosion is

us.ing porous concrete as a filter behind the lagging. Such a procedure was reported by Mansur and Alizadeh (1970).

9*40 CONSTRUCTION CONSIDERATIONS

9.41 Soldier Piles

9.41. 1 Driven Soldier Piles

Conventional pile drivers may be used to drive

soldier piles. Bearing pile sections are the most desirable sections

-124-
for driving. In hard ground, bearing piles may be equipped with a

driving point to help penetrate boulders and/or to get sufficient depth

for adequate lateral resistance or bearing capacity.

9.41. 2 Soldier Piles Set in Pre-excavated Holes

Pre-excavated holes may be used for one or

more of the following reasons:

a. To reduce noise and vibrations.

b. To penetrate a hard layer.

c. To set a long soldier pile in the ground

so that it can conveninetly fit in the leads of a pile driving rig for further

driving.

d, To set the soldier pile at a precise location.

e. To install certain types of soldier piles

such as deep-web, torsionally flexible, wide flange sections, which

may be difficult to drive.

f, To minimize vibrations which could have an

adverse effect on loose unconsolidated sediments and nearby structures,

g* TO penetrate sufficiently far below the bottom

of the excavation to ensure lateral toe resistance and vertical bearing.

Such considerations may necessitate percussion or rotary drilling to

penetrate rock or boulders.

Pre-excavation is usually done with augers.

Equipment used for augering may be bucket type augers at the end of

a kelly bar or continuous hollow stem augers, In hard ground augers

may not be practical. Instead, percussion drilling or rotary drilling

may be necessary.

Pre-excavated holes facilitate setting the soldier

piles to a very close tolerance, both vertically and in plan. Where

alignment is critical, the soldier pile is set within the pre-excavated

hole with a centering spider,

It is common practice to use structural concrete

below the level of the excavation to assure vertical bearing and lateral

resistance against kick out and to use lean concrete for the rest of the

hole. It is believed that properly placed lean concrete can also be used

below the excavation level.

-125-
9,42 Installation of Lagging

TO rnmimiz;e overcut, hand tools should be used to shape

the soil and to fit the lagging board in place, If necessary, wedges

can be used to close the space between the lagging board and its bearing

area,

The depth of exposure below the last placed lagging

board may be as little as 1 foot, as in the case of saturated silt, or

as much as 4 or 5 feet in cohesive hardpan. The restriction in depth

of unsupported cut is the height of cut that is stable. If the unsupported

cut below the last board is unstable, excessive loss of ground may occur.

In circumstances of adverse soil conditions, proper:

cutting of the soil bank, backpacking of soil behind the lagging, and

filling the vertical space between lagging boards with a proper filtering and drainage material are all important details. Louvered lagging

is considered good practice,

9.43 Removal

There is a divergence of opinion among practitioners

with regard to whether or not untreated wood can be left in place

permanently above the ground water table. In this present state of

diverse opinion, the preferred options are to remove lagging that would

be permanently above the ground water table or to treat the wood

with chemicals for the purpose of preventing future deterioration.

When lagging is removed, the process should be in

stages of a few feet at a time. Concurrently, backfill should be compacted.

Soldier piles may be removed if it is practical to do so and provided

voids are not created.

Treatment standards are shown in Table 5.

-126-
Table 5. AWPA minimum retention standards

for sawn timber below ground.

Creosote, creosote solutions, and

oil-borne chemicals.

Creosote

Creosote-coal tar solution

Pentachlorophenol

Water -borne inorganic salts

(oxide basis).

( 1) Amoniacal copper

arsenite (ACA)

(2) Chromated copper

arsenate (CCA) type A

(3) Chromated copper

arsenate (CCA) type B

C hroma ted copper

arsenate (CCA) type C

Trade” Names:

(1) Chemonite

(2) Erdalith, Green salt

(3) Boliden K - 33

Osmose K - 33

lbs/cu. ft.

Retention

12

1 2

0. 6

0. 6

0. 6

0.6

0. 6

Note: This table presents minimum retention by assay in lbs. per

cu. ft. for Southern Pine, Douglas Fir, or Western Hemlock.

*Data from AITC; “Timber Construction Manual”, 2nd Edition, Wiley, 1974.

-127-
CHAPTER 10 - STEEL SHEET PILING

10.10 INTRODUCTION

This section concerns rolled Z-shaped or arch-shaped interlocked steel sheet piling. Because of

bending, Z-shaped sections are more

than are the arch-shaped sections.

their greater resistance in

common in American practice

10.20 DESIGN CONSIDERATIONS

10.2 1 General Applications

Steel sheet piling is typically used in soils that do not

permit easy placement of lagging, such as soft clays, saturated silts,

or loose silty or clayey sand. These soils are potentially unstable

when they are exposed during excavation.

Interlocked steel sheet piling is very effective in

cutting off concentrated flow through pervious layers within or below

the excavation and protecting against the possibility of a “blow” condition or other source of ground loss. On the other hand, the steel

sheet pile wall does not necessarily prevent lowering of the piezometric level and accompanying consolidation when the excavation is

made in relatively impervious soils. In these cases the steel sheet

pile wall has approximately the same permeability as the soil in which

it is driven (clayey sands and clays would fall into this category of

soil types).

10.22 Available Sections

Figure 57 schematically shows typical American steel

sheet pile sections used for relatively deep excavations. Table 6 gives

information concerning the properties of various steel sheet pile

sections (see Figure, 58and Table 7 for foreign sections,). Heavier

sections are available in foreign steel sheet piling than in domestic

piling.

Note that the PDA section and PMA section interlock on

the midline of the wall, whereas the “Z” sections interlock on the inside and the outside line of the wall. For the deep arch and medium

arch sections, it is conventionally assumed in American practice that

shear cannot develop along the interlocks and therefore the two sheet

piles which combine for the full wall depth cannot be considered effective in bending. European practice assumes interlock friction and

-128-
PZ 38 PZ 32 8 PZ 27

PMA 22

.I

--<’ ---L

PDA 27

;c_ -

.- -

Figure 57, Domestic sheet pile sections.

-1297
Section

PMA 22

Dimension (in)

D, depth L, length

3-l /2 x 2 = 7 (1) 19.6

PDA 27 5x2=10 16

PZ 27

PZ 32

PZ 38

1 2 1 8 27. 0

11.5 21 32. 0

12. 0 1 8 38. 0

Weight

lb/ sf

22. 0

Table 6. Domestic steel sheet pile sections.

27. 0

16 5.4

40 10.7

183 30.2

220 38. 3

281 46.8

(1) Single pile is 3-l/2” deep.

As driven, wall is 7” deep.

-130-
FRODINGHAM IBXN,2N,3N,4N

HOESCH Nos. 95,116,134,155,175,215

I--------\ I/------- .

//

I‘

//’ ‘#

‘/ ‘/

BELVAL BZ-25O,BZ-350, R-450, Bit-550

,----------e--.

,,,--------‘-w-e,,

1’1

---g

‘,s.

-L=19.68” A

Figure 58. Foreign sheet pile sections.

-131-
Table ‘7. Foreign steel sheet pile sections.

(1) Data fro& L. B. Foster Company, Pittsburgh, Pa.

(2) Data from Skyline Industries, Port Kearny, N. J.

-132-
therefore takes advantage of the full section modulus of both piles

(Tschebotarioff, 1974).

10.23 Allowable Stresses

The conventional ASTM grade used for sheet piling is

A 328, which has a minimum yield point of 38,500 psi. Some companies

produce steel sheet piling in higher strength steel using ASTM grade

A 572 in three types: 45,000; 50, 000; and 55, 000 psi yield point steel

(see Table. 8).

AISC allowable stresses may be used for the steel sheet

pile wall at full depth. T.emporary, intermediate conditions which exist

during the course of excavation may be analyzed using a 20 percent

overstress above the normal AISC allowable stress.

10.30 CONSTRUCTION CONSIDERATIONS

10. 31 Installation of Sheet Piling

The general installation technique is to drive the steel

sheet piling in waves, always maintaining the tips of adjoining steel

sheet piles no more than about 5 to 6 feet apart. The ball end (male

end) should always lead to prevent plugging of the socket end (female

end) with soil. This measure protects the interlocks from tearing.

Pile drivers may be impact type, single or double

acting hammers , or vibratory drivers. The vibratory drivers are

run by hydraulic or electric motors which power eccentric shafts

(Foster, 1971).

Silent pile drivers have been developed by Stabilator

AB of Stockholm, Sweden, and by Taylor Woodrow Construction,Ltd.,

of Great Britain. The former operates by compressed air, the latter

by hydraulic rams.

Perhaps the greatest cause of ground water leakage

and/or loss of ground is the ripping of sheeting out of the interlocks

as the result of poor alignment or hard driving conditions. Obviously,

the potential for this rises with the density of the soil and with the

frequency of boulders and obstructions below the surface.

10. 32 Removal of Sheet Piling

Conventional extractors can be used. Loose granular soils

-133-
Table 8. Steel types used for sheet piles.

‘ASTM Grade

fy, psi

Yield Point

fb, psi

‘AISC:: Design

Flexural Stress

A 328 38,500 25,400

A 572

Grade 45 45,000 29,700

Grade 5 0 50,000 33,000

Grade 5 5 55,000 36, 000

:::

fb = .66 f

Y

-134-
may, of course, consolidate as a result of vibrations during driving

or extraction. It is believed, however, that the influence of such

vibrations in loose granular soil will be confined to within about 10

to 15 feet of the sheet pile wall.

In cohesive soils the possibility exists that the clay

may adhere to the sheeting, especially at the sharp angular bend in

the corners of the PZ section. *This wouLd lead to displacements

in the adjoining ground.

Steps that can be taken to reduce the adhesion of clay

include prior application of bituminous material to the steel

and the application of direct electric current.

-135-
CHAPTER 11 - CONCRETE DIAPHRAGM WALLS

11.10 INTRODUCTION

The term concrete diaphragm wall refers to a continuous

concrete wall built from the ground surface, The walls may consist

of precast or cast-in-place concrete panels or contiguous bored

cone rete pile s ,

The most common wall type is a tremie concrete, diaphragm

wall cast within a slurry stabilized trench. The trenches are usually

about 24 to 36 inches wide and are excavated in 10 to 20 foot lengths.

After the individual panels are excavated, end stops and reinforcing

are placed. Concrete is poured, and the end stops are removed.

Once the concrete has set, the neighboring panel can be excavated.

The system using precast concrete panels lowered into a slurry

stabilized trench is quite popular in Europe. The use of this system

and concrete diaphragm walls in general is expected to increase in

the United States.

Figures 59, 60, 61 and 62 show various aspects of diaphragm

wall construction.

11.20 PROPERTIES OF BENTONITE SLURRY

The most commonly used bentonite contains the clay mine’taal,

sodium montmorillonite. When mixed with water, bentonite forms a

colloidal suspension (slurry). Following agitation, a bentonite slurry

will gel and develop shear resistance. Bentonite also displays plastic

viscosity, which means additional shear resistance develops depending upon the rate of shear application.

Fluids which develop both gel shear strength and additional

shear resistance from fluid viscosity are known as Bingham fluids.

Fluids which have no shear resistance in the static conditions but do

have viscosity characteristics are known as Newtonian fluids. Figure

63 iLLustrates the viscous character of Bingham and Newtonian fluids.

Bentonite fluid is thixotropic; that is, it will lose strength when

disturbed but will gain strength and gel when left undisturbed. In

diaphragm wall construction it is desirable to maintain a fluid slurry

which requires that the slurry be circulated and agitated.

-136-
.,i

. .

\.

i

:

-137-
_-

-.

/ REINFORCING STEEL CAGE BEING LOWERED

INTO SLURRY- FILLED TRENCH.

I,

_’ ._-

. . . . -: I. ,,/

__. -:

:

- -

_:.. .I,

*.’ _I... . .

. . . ‘,’ I

i :,,

,-;

. ;_... .

Y/END PIPE - -‘in...> :, ,r” 4 ‘. ..L’,.: ‘- -c _. _ ; .“.) -~~, __ “. ” *.~‘.

. .

Figure 60. Preparations for concrete place.ment.

(Courtesy of ICOS Corporation).
-139-
-140-
SHEAR APPLICATION RATE, F(v)

FLUID

Gel shear strength

Yield shear strength

Viscosity of Newtonian fluid

Apparent viscosity of Bingham fluid

Plastic viscosity of Bingham fluid

A parameter which is for particular geometry.

A function of shear application rate.

Figure 63. Viscous behavior of Newtonian and Bingham fluids.

-141-
11.30 SLURRY TRENCH STABILITY

11.31 General

Primarily, it is the fluid pressure of the slurry in combination with arching in the ground that maintains trench stability

in cohesionless soil. In addition, some local penetration into the

pervious, soil will impart cohesion to the soil and will prevent spalling.

The bentonite slurry in the trench is maintained at a

higher elevation than the surrounding ground water table. By a combination of hydrostatic pressure, osmotic pressure, and eLectroLytic

properties of the colloid, a membrane or “mudcake” forms against

the walls of the trench. The effect of this “mudcake” is to prevent

fluid loss and to maintain the fluid pressure against the trench wall.

11.32 Mudcake

The extent of fluid penetration into the soil voids depends

upon the permeability of the soil and the properties of the colloid.’

With very pervious soils such as sands and gravels, having permeabilities greater than 10 Am/set, there could be free penetration of the

slurry into the soil without the formation of a “mudcake”. With soils

having permeabilities between low2 and lb -1 cm/set, there may be some

time lag associated with the development of an impervious “mudcake”.

With soils having permeabilities less than 1V2cm/sec, the depth of

penetration for formation of an impervious “mudcake” is minimal and

there is essentially no time lag (Hutchinson, 1974).

With impervious soils, such as clay, the bentonite need

not form a “mudcake” because the clay itself is essentially impermeable.

In these cases the bentonite protects against fluid loss through pervious

seams that may be interbedded within the parent clay formation.

11. 33 Pressure of Slurry Fluid

11.33.1 General

It is common practice to maintain the water

level in the trench at least 4 feet above the ground water level.

This excess head in combination with the greater specific gravity of

the slurry creates an unbalanced pressure on the trench walls which

provides a force resisting a trench stability failure. Typically,

the bentonite concentration is about 4 to 6 percent by weight yielding

specific gravities of approximately 1.023 to 1.034.

-142-
11. 33.2 Stability Analysis

A number of simplified stabi.lity analyses are

presented in detail in Volume III (Construction Methods). All of these

cases examine trench stability on the basis of fluid pressure. None

of the cases consider arching which is a significant stabilizing factor

for panel excavations on the order of 15 feet or less.

Since panel lengths and fluid levels required

to maintenance trench stability are established large.ly on the basis of

experience and/or trial panel sections, stability analyses are not

generally used to evaluate trench stability. Their main value is helping

to assess the relative importance of fluid pressure and the depth of

the trench on overall stability.

11.34 Arching

In order to understand arching, the redistribution

of stresses away from plane strain conditions, two conditions must be

examined:

a . The strain conditions at great depth below the

surface.

b. The strain conditions near the surface.

At great depth, strain is essentially a two dimensional

condition acting in the horizontal plane outside the influence of Local

conditions. Horizontal strain is less near the ends of the panel than

near the center of the panel. As a result, load concentrates at the

ends of the excavated panel, thus re.lieving the stress condition near the

center and improving stability.

The very top of the trench is restrained by a guidewall

which is used to align the excavation and to introduce recirculated slurry.

The guidewall is essentially rigid and therefore restrains lateral .movement

so that arching develops in the vertical plane. Arching also occurs in - the horizontal plane.

Experience has shown that a rigidly placed guidewalt

is an extremely important element in maintaining the stability of the

top part of the trench.

-143-
11.35 Factors Contributing to Trench Stability

Fluid pressure and arching are the primary factors

maintaining trench stability. Other factors that contribute to trench

stability are the “mudcake” (formed by the electro-osmotic phenomenon)

and slurry penetration into the voids of cohesionless soil.

11.35.1 Electra-osmotic Phenomenon

Electra-osmosis contributes to the formation

of the impermeable “mudcake” which prevents fluid loss. The electrical

potentiaL at the slurry - soil interface causes a migration of colloidal

particles to the trench wall even in the absence of fluid flow under

hydraulic head.

11. 35.2 Penetration of Slurry into Cohesionless Soil

With slurry penetration of a few inches, an impermeable membrane effect is created; seepage pressures exist only

in the membrane; and the soil within the membrane is easily held by

the shear strength of the slurry in the soil voids. In this case, the

weight of particles tending to fall away is small compared to the shear

resistance of the soil. The seepage forces and the slurry shear strength

combine to prevent spalling.

As the zone of penetration increases, a larger

volume of soil is influenced. In this case the weight of the soil mass

within the zone of penetration is large compared to the shear resistance

of the soil, and the condition becomes less stable. A deeper penetration

lowers the gradient , lowers the seepage force per unit volume of soil

within the zone of penetration. A limiting case would be free penetration

in open gravel. Spalling of the trenches is typically the result of this

phenomenon.

It has been observed that trench wall collapses

(spalling) are more common near the top of the excavation than the

bottom. Miiller - Kirchenbauer (1972) points out that the slurry contains

few suspended soil particles when excavation first begins and only the

bentonite resists slurry penetration. As the trench depth increases, the

suspended soil particle concentration also increases. The suspended

particles aid in forming a more effective mudcake by plugging soil pores.

For this reason, in pervious soils it is advisable to maintain a specified

percentage of fine sand in the slurry to aid mudcake formation (Hutchinson,

et al, 1974). Soil arching is also less effective in preventing spalling at

the top of the excavation.

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11.40 PRACTICAL ASPECTS OF SLURRY STABILIZED EXCAVATIONS

11.41 Water Level

It is conxnon practice to maintain the trench fluid at

least 4 feet above the ground water level. In soft clays, loose silts,

and sands, cases have been reported where the level was maintained

8 feet or more above the ground water in order to assure stability.

Under certain conditions, this may necessitate the construction of

dikes paralleling the trench to maintain the slurry level at the desired

elevation or alternatively, pumping to draw down ground water.

The site investigation must carefully identify highly

pervious strata through which slurry loss may occur and also identify

the potential for artesian conditions in confined layers.

11.42 Control of Bentonite Slurry

11.42.1 General

The quality of the bentonite slurry must be

checked to assure that the design slurry properties are being maintained. Quality control includes a check on the freshly hydrated

bentonite slurry prior to insertion in the trench and checks on the

re-circulated slurry to assure that the slurry is not being contaminated.

Il. 42.2 Source of Contamination

a. Detritus Contamination. The clay, silt, and

sand particles that build up in the slurry are the contaminants. These

particles increase the slurry density with the effect most pronounced

at the bottom of the trench. The increased slurry density impairs

circulation and adversely affects concrete placement. The concrete can

not displace the denser slurry as easily as a thinner slurry.

b. Calcium Contamination. Cement, in contact

with the slurry, is the major source of calcium contamination. Fine

soils or artificial fill containing concrete demolition debris may also

be a source of calcium contamination. Calcium contamination causes

flocculation of bentonite particles, making the slurry more difficult

to circulate. An excessively thick mudcake may form which is more

difficult to displace during concreting then the thinner sodium bentonite

mudcake. The calcuim bentonite mudcake is also more permeable

which increases the chances for fluid loss in permeable soils. In some

soil conditions the use of a calcium bentonite slurry may minimize

the problems of calcium contamination.

-145-
c. Salt Contamination. Excessive salinity

changes the electrolytic properties and may cause the clay particles

to flocculate and settle. This makes it more difficult for the slurry

to form an effective mudcake and may lead to a loss of fluid and stability. The problem would be especially acute in pervious granular soils.

11.42.3Slurry Mix

The slurry must form an effective mudcake

on the sides of the trench and be dense enough to provide adequate

stability while still being thin enough to allow circulation and concreting. Agents may have to be added to the slurry to counteract chemical

contamination, to decrease slurry viscosity, or to aid mudcake development. The measures that can be taken to preserve slurry properties are described below.

a. Viscosity. Flocculation of the bentonite

particles will result in increased slurry viscosity. Mud thinners

(dispersants) retard flocculation and help maintain the desired viscosity.

Rogers (1963) classifies chemical mudthinners

in the following groups: molecularly dehydrated phosphates and

polyphosphates, plant tannins, lignosulfonate wood by-products, and

mineral lignins. “Dextrid” a trade name polysaccharide made by

Baroid, and ferro chrome lignosulfonate are two chemical mud

thinners mentioned by Puller (1974).

Use of mud thinners requires experience and

laboratory test verification of their effect. As a minimum, such

tests would include pH tests, viscosity tests, and standard API fluid

loss tests in order to diagnose the problem and to determine approprate

treatment.

b. Cement Contamination. A common approach

is to introduce sodium ions to retard ion exchange with calcium.

Agents are: sodium ferro chrome lignosulfonate (FCL) (0.1 to 0.3 percent by weight), sodium bicarbonate, and other thinners.

c. Salt Contamination. A simple precaution

to counteract salt contamination is to mix the slurry with fresh water

and to be sure that it is fully hydrated before introduction into the

trench. Sodium ferro chrome lignosulphonate (FCL) is remarkably

effective in resisting excessive salinity (Xanthakos, 1974).

-146-
d. Fluid Loss in Highly Pervious Soils.

Merely increasing bentonite concentration in soils having permeabilities

greater than about 10’1 to 10” cm/set will not be effective (Sliwinski

and Fleming, 1974). Hutchinson, et al (1974) propose the.addition

of about 1 percent fine sand as a means to penetrate and block the pores

of pervious soils having permeability greater than 10’lcm/sec. Other

additives include a range of inert plugging substances such as; nut

shells, plant fibres, rayon, cellophane flakes, mica, ground rubber

tires, etc.

11.42.4 Control Testing

Appendix A to Chapter 4 of Volume III (Construction Methods) describes Standard API procedures, equipment, and

specifications for control testing.

Viscosity and Gel Strength

Fann Viscometer.

To determine plastic viscosity and yield shear

stress, the rotor is turned at 600 rpm and then at 300 rpm. The

plastic viscosity in centipoises is the difference between the 600 rpm

reading and the 300 rpm reading. The yield shear stress is the 300

rpm reading minus the plastic viscosity. (See Figure 64. )

With the viscometer, the gel strength is defined

by API as the maximum reading obtained at 3 rpm. Alternatively,

the rotor may be turned very slowly by hand. The tests are for an

essentially static condition, conventionally obtained after 10 minutes

of gel time.

Shearometer. The shearometer can be used

to obtain gel strength. Because of differences between the equipment

and procedures, the gel strength values from the shearometer are not

the same as these from the Fann viscometer. Specifications must therefore identify the procedure to be used as well as the control values.

Hutchinson, et al (1974) recommend gel strength

values of O.O5g/cm2 to 0. 20g/cm2 using the viscometer. The FPS

specification (1973) requires 0.014 to 0. log/cm2 using the shearometer.

Marsh Cone. The Marsh cone is a simple method

for obtaining an index of viscosity, especially useful as a quick field

method. The standard size cone is filled with slurry and the time for

-147-
i

/.

I

I _ /--------------------- $1

I t------i,

I I YIELD STRESS&

I

If

300 600 ROTOR SPEED (rpm)

PLASTIC VISCOSITY = z,600 - 7-300

YIELD STRESS

= ~300&00 -&oo>

Figure 64. Data from Fann viscometer.
the funnel to drain is reported as Marsh funnel viscosity. Obviously

the more viscous the fluid, the longer the drain time. The FPS

specification requires that the Marsh cone drain time be between

30 and 60 seconds.

Density. Density is a simple measurement

of a known volume of slurry using a Mud Density Balance.

Cement contamination, which adversely affects

the slurry by causing flocculation, increased viscosity, and more

permeable mudcake, also raises the pH. The FPS specification requires that the pH lie between 9.5 and 12. The pH can be determined

with litmus paper or with a pH meter.

Filtering Performance

The recommended device is the standard API

fluid loss device. Slurry (600 cc) is placed over filter paper; 100

psi pressure is applied; and fluid loss is measured in a 30 minute

time period.

Excessive sand content may unfavorably raise

the density of the slurry so that it is difficult to displace during concreting, and it may result in sand pockets in cast-in-place concrete

walls. On the other hand, fine sand may be added to the slurry being

circulated in the trench to control fluid loss in permeable soils.

11.42.5 Cleaning the Slurry

Depending upon the soil conditions and the method

of excavation used, the procedure for cleaning the slurry of suspended

detritus (gravel, sand, silt, etc. ) may include sedimentation tanks,

mechanical screening, and centrifugal separation using hydrocyclones.

11.43 Some Potentially Difficult Soils

Diaphragm walls have been constructed in virtually

al1 soil types. It is usually technically possible to install diaphragm

walls, but the costs may be greater than other methods or combinations

of methods. In severe cases treatment of the soils may be required

prior to wall construction.

-149-
11.43.1 Hiphly Pervious Soils

Loss of ground water through highly pervious

strata represents an obvious threat to the stability of the trench.

Bentonite concentrations between 4 and 6 perc_ept are satisfacLo2ry

for soils with permeability less than about 10 cm/set to 10 cm/set:.

Beyond merely increasing bentonite concentration, more permeable

soils may require a variety of measures such as the addition of fine

sand or various plugging agents to control seepage loss. A more

detailed discussion of pLugging agents is presented in Chapter 4 of

Volume III (Diaphragm Walls).

11.43.2 Saline Soils

In general, salinity not a severe problem

provided the bentonite is hydrated with fresh water. Even in coastal

sites where the land had been filled hydraulically with sand, the salt

concentration was not sufficient to cause adverse effects (Fuchsberger,

1974).

11.43. 3 Soft Clays

Soft clays with a shear strength of less than

500 psf must be approached with caution. Panel lengths and construction procedures must be verified by experimental test sections in the

early stages of construction. Moreover, such test excavations must

be accompanied by careful monitoring of displacements of the adjacent ground.

11.43.4 Calcium Laden Soils

Calcium contamination comes from lime soils,

gypsum, or anhydrite in the ground (Sliwinski and Fleming, 1974).

It may lead to flocculation and an ineffective mudcake on the trench

wall.

11.43.5 Organic Soils

Peat may overbreak and lead to an irregular

wall. Also, it may float free into the slurry and become embodied

within the concrete. Organic soils may also adversely affect the pH.

11.43.6 Residual Soils

Experience in residual soils in Brazil has shown

-150-
severe pH contamination due to presence of iron oxides. The slurry

became so thick and viscous that it was necessary to totally replace

before concreting.

11.43.7 Stiff Fissured Clays

Severe overbreaks and local collapses have been

experienced in highly fissured overconsolidated London Clay. This was

attributable in part to an unfavorable joint pattern in the clay (Puller,

1974),

11.43.8 Soft Silts

Local liquefaction may occur in non-plastic soft

silts, perhaps initiated by disturbance from excavation equipment.

11.44 Precautionary Measures

The site investigation must obtain sufficient data on

ground water chemistry, soil strength, and pervious strata to permit

an evaluation of slurry wall feasibility. Records of water loss during

drilling operations are essential as are in situ permeability tests in

very permeable strata.

During construction, trial panels can be excavated and

the lengths of panels can be varied to determine the optimum length

and to minimize the deformations and potential danger to adjacent

ground.

In cases where the potential for fluid loss is great,

stockpiling of backfill material should be considered in order to fill

panels in an emergency arising from a sudden loss of fluid. Acceptable

filling materials would be granular soils, gravelly soils, or crushed

stone.

Where the source of leakage is near the surface, the

excavation can be carried out in two steps. The first step is to dig

an over sized trench and refill with lean concrete; the second step

is to make the slurry trench and form the diaphragm wall in the conventional way.

11.50 STRUCTURAL ASPECTS OF CAST-IN-PLACE WALLS

11.51 Load Bearing

Provided that the slurry quality is adequately controlled,

-151-
the tremie concrete will satisfactorily displace the bentonite mudcake and develop effective bond against the soil. British practice

with cast-in-situ piling formed in slurry stabilized holes bears out

the successful development of soil adhesion, especially in cohesive

soils. In more pervious granular soils the mudcake is more difficult

to displace and may reduce side friction by about 10 to 30 percent

(Sliwinski and Fleming, 1974).

It is common in Europe to use load bearing diaphragm

wall elements (also referred to as slot caissons). Load is transferred

by side friction and end bearing.

11. 52 Concrete

11.52.1 Mix

The concrete must be a free flowing mix which

will displace the bentonite and bond t’o the reinforcing.

The FPS Specification (1973) for diaphragm

walls is appended to Chapter 4 of Volume III (Construction Methods).

Briefly, the requirements are as follows:

Slump - Minimum slump 150mm (61’) :

desirable slump 175 mm to 200mm

(7” to 8”)

Water Cement Ratio - Less than 0.6

Aggregate - Naturally rounded gravel and sand

(if available)

Sand Content - 35 to 40 percent of total weight

of aggregate

Cement Content - At least 400 kg/cubic meter

for tremie concrete

11.52.2 Placement

Concrete placement is performed simultaneously

through one or more tremie pipes in each panel. Pipe diameters are

normally 6 to 10 inches. General practice is to limit horizontal

travel distance of concrete to less than 8 to 10 feet to prevent significant segregation.

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11.53 Steel

11. 53.1 General Applications

The reinforcing can be a cage of rebars, a

combination of horizontal rebars and vertical wide flange sections,

or wide flange sections alone. The use of soldier pile reinforcing

alone is more common in the U.S. than in Europe.

11.53.2 Bond

Opinion varies with respect to the reduction

effect, if any, that bentonite slurry has on bond of concrete to steel.

The FPS Specification (1973) allows normal

bond stress on plain bars but allows only 10 percent more bond on deformed bars.

11.53.3 Cover

The FPS specification (1973) recommends

the following: Concrete cover over steel reinforcement should be

at least 75 mm (3 inches). Minimum clear spacing between main bars

should be at Least 100 mm (4 inches).

11.54 Panels and Joints

The most common type of joint used in cast-in-place

diaphragm wall construction is formed with a stop-end tube, a round

pipe placed in the end of the panel prior to concreting. Figure 65

is a schematic illustration of the joint configuration formed by means

of the stop-end tube.

Another procedure for joint formation is to use a

steel wide flange beam or precast I-beam to serve the dual purpose of

providing a joint for both shear transfer and vertical steel reinforcement. Figure 66 illustrates this joint.

11.60 EXCAVATION OF SLURRY TRENCHES

11.61 Guide Walls

A well-constructed guide wall is essential to prevent

caving of the trench wall in the uppermost part of the excavation. The

guide wall serves additional functions: a) to align the trench, b) to

contain the slurry, c) to suspend precast elements and d) to suspend

-153-
ROUND JOINT, USING STOP- END TUBE

CONDARY PAN L

f

PRIMARY PANEL

CONCRETED

-m--v

----e-w

STOP- END REMOVED

PRIMARY PANEL AND

SECONDARY PANEL

CONCRETED

Figure 65. Panel joint with stop-end tube.

-154-
TWO-STEP EXCAVATION

AUGER PRIMARY

HOLES.

SET SOLDIER PILES;

FILL WITH LEAN

CONCRETE OR GROUT,

TREMIE PRIMARY

PANEL.

TREMIE SECONDARY

PANEL.

ONE -STEP EXCAVATION

TREMIE PRIMARY

PANEL.

TREMIE SECONDARY

PANEL.

Figure 66. Panel joints with I-beams.

-155-
reinforcing steel in cast-in-place walls. Figure 67 shows guide wall,

sections.

11.62 Trenching

11,62.1 General

Procedures are:

a. Excavation Buckets These bring the

material directly to the surface, discharge load, and then are introduced back into the trench.

b. Direct or Reverse Circulation. These

methods break up the material into smaller particles so that it may

be suspended in the bentonite slurry and circulated to the surface to

the screening-desanding operation. Care must be taken to avoid

clogging of lines by boulders.

With cast-in-place walls, alternate panels

are excavated and concreted between stop-ends. Then the remaining

panels are completed. Another ,procedure is to proceed continuously

by excavating and concreting one panel at a time and always setting

a stop-end at the leading edge. In this case, the work proceeds at

two or more locations, so that the excavation equipment is busy during

concreting.

Typically, with precast panels, the trench

excavation proceeds continuously. However, the walls can also be

constructed in alternate panels.

11.62.2 Excavation Methods

ELSE Trenching Machine

An early technique was the ELSE trenching

machine which was introduced in Italy in 1958. This trenching shovel

operates like a power shovel. The ELSE trenching shovel is a specially

designed device which operates from a vertical mast that is advanced

into the trench with the excavation. With each bite the shovel is brought

to the surface to discharge its Load.

This device is still conventionally used in Japan

as reported by Ikuta (1974) but is rarely used in the United States. A detailed description of the operation of this device is provided by Xanthakos

(1974).

-156-
(a) COMPACT COHESIVE SOIL

BACKFILL

(b) LOOSE COHESIONLESS SOIL

CONCRETE WALL

COMPACTED BACKFILL

. :*

.

l I

I

. .

.

.

.

I

I - .

.

.

’ I .

.

I v

;. l

NOTE: ADD CEMENT TO

BACKFILL TO

PREVENT UNDERMINING AND TO

INCREASE STABILITY

Figure 67. Guide walls.

-157-
Clam Shell

The most common types of excavation equipment are specially designed clam shell buckets, conventionally referred to as grabbing tools or grabs.

Vertical and horizontal alignment of the bucket

is assisted by a guiding skirt (perhaps 15 or more feet high, 6 feet

or more long, and slightly less wide than the grab bucket). The bucket

extends just below the guide skirt.

The jaws of the grab may operate mechanically

or hydraulically. In the mechanical operation, the equipment weight

may not be fully effective and therefore is less effective in hard ground.

HydrauLic devices vary -- they may work from a single central piston

or from pistons on each side to close the jaws of the grab.

11.62.3 Direct and Reverse Circulation Methods

a. Soletanche, A Soletanche device, which

operates on rails that are set along the trench, is a reverse circulation device. The cutting tool benches back and forth between the

ends of the panels and cuttings are brought to the surface by surface

by suction and/or air lift through the tool itself. This device can

employ either percussion or rotary drilling techniques.

b. The BW Drill. The BW drill is marketed

through the Japanese firm, Mitsubishi International. Like the Soletanche

device, it operates on rails. It is a self-contained excavation tool

with four rotary cutter heads at its base (rotation about vertical axis).

Slurry cuttings are circulated through the device in suction lines,

desanded, and then reintroduced to the trench.

c. TBW Excavator. This device uses cutter

heads rotating about the horizontal axis. It is a product of the Japanese

firm, Takanaka, and was reported on by Ikuta (1974).

11.62.4 Hard Ground

Obstructions are broken up by heavy chisels

or chopping devices (material removed by grab buckets), by percussion tools, or by rotary tools. In genera.1, grab buckets or rotary

devices are used in soils of normal density or consistency. Percussion methods are necessary in cemented soils, hard boulders, clays,

and till.

-15%
Sliwinski and Fleming (1974) report a method

to penetrate soft rock by first boring 30 inch diameter holes at regular

spacing and then removing the material between the bored holes with

a hydraulically operated grab tool. Tamaro (1974) reports a similar

procedure used by ICOS to penetrate bouldery formations.

11.70 DIAPHRAGM WALLS OTHER THAN CONTINUOUS

CAST-IN-PLACE CONCRETE

11.71 General

This discussion covers the following:

a. Diaphragm walls constructed of precast elements

set within slurry stabilized trenches.

b. Hybrid technique using pre-set steel or concrete

soldier piles in combination with intervening cast -in-place concrete

panels.

lines.

c. A wall composed of bored piles set in one or more

11.72 Precast Concrete Methods

11.72. 1 General

Precast concrete elements are normally set

within a continuously excavated slurry stabilized trench. Figures 68

and 69 are schematic illustrations of the Soletanche and Bachy methods.

Franki uses a similar method.

Precast elements are carefully aligned and

suspended from the guide wall into the grout slurry (or cast-in-place

concrete) . The elements can not be set until the grout slurry has gained

sufficient strength to provide vertical support. The elements can be used

alone or in combination with an underlying conventional’ cast-in-place

diaphragm wall.

Grout fills the space between the back side

of the precast element and the soil, thus forming tight contact and

an impervious membrane. Grout adhering to the inside face is removed

during excavation.

The size of the precast concrete elements

is controlled by the load capacity of the crane. In urban areas the

crane size may be controlled by city ordinances thereby limiting

-159-
(a) TONGUE &GROOVE

r

GROUT REMAINING IN PLACE

v- TIE BACK, IF NECESSARY

EXCAVATION. SIDE

\GROUT RE[MOVED DURING EXCAVATION

(b) T BEAMS 81 SLABS

GROUT REMAINING IN PLACE

SOIL SIDE

EXCAVATION SIDE

OUT REMOVED DURING EXCAVATION

Figure 68. Panosol walls (Soletanchc, France).

-160-

panel size. Depending upon wall thickness the depth limitation is

normally in the range of 30 to 50 feet.

The T-beam/slab combination (Figure 68b)

offers flexibility with regard to depth. In this case the T-beam can

be carried to a lower elevation to engage a bearing stratum or to develop additional passive resistance. Slab panels need only extend to

the depths required for the permanent wall except where the wall must

also act as a ground water cutoff.

11.72.2 Grout and Slurry

The Soletanche method uses a special grout

mix which serves the dual purpose of stabilizing the trench and then

hardening in place. The base mix is cement and bentonite with additives to control setting time, viscosity, and strength,

Other companies employ conventional bentonite

mud slurries for trench stabilization during excavation but then introduce a cement-bentonite sealing grout (about 4 percent bentonite

and 14 percent cement) into the bottom of the panel prior to placing

the precast element. The panel then displaces the mud slurry so that

only the cement-bentonite mix remains. Such a method was described

by E. Colas Des Francs (1974) in the Bachy method.

11.73 Soldier Pile Combination Walls

11.73.1 General

The techniques described in this section all

use soldier piles at regular spacing along the wall in combination with

poured concrete between the soldier piles.

One of the features of first setting the soldier

pile in an augered hole and then concreting the panel is that the soldier

pile can be carried to a lower elevation than the wall panel for the

purpose of obtaining vertical bearing and/or increased lateral resistance.

Another approach is to eliminate the extra

step of augering and setting soldier piles separately. Following excavation of the panel, the soldier piles are positioned together with

the reinforcing cage , and the panel is concreted.

-162-
11. 73.2 Two Step Excavation: First for Piles:

Second for Panel

Two techniques are shown in Figures 70 and 71.

Figure 70 shows a wall formed using precast soldier piles while

Figure 71 shows a wall with steel, wide flange, soldier piles.

11.73.3 One Step Excavation

In this method (Figure 72) the soldier piles

and reinforcing are placed concurrently in the excavated panel.

11. 73.4 ,Discussion

Cost considerations aside, preset soldier piles

offer inherent advantages concerning protection of adjacent structures,

especially in unstable or weak soils and/or in the presence of heavily

loaded foundations. Risk exposure during setting of the soldier pile

is minimal; subsequently during excavation of the intervening panel,

the length between the soldier piles is relatively short - in the case

of BARTD, only about 6 feet. Thus, protection against movement,

or collapse, is always maintained. Also when soldier piles are installed separately, they can be extended to whatever depth is required

to develop bearing and/or toe restraint.

11. 74 Bored Pile Walls

11.74. P General

These walls are built by forming grouted or

cast-in-place concrete piles continuously along the line of the excavation. For purposes of this report, the methods have been classified

as “small-diameter piles”, conventionally formed by grouting using

hollow stem augering equipment, and “large-diameter piles” formed

by excavation with a solid auger and then filling with concrete after

withdrawal of the auger. In both cases the piles are reinforced.

Figure 73 illustrates these bored pile walls.

11.74.2 Small-Diameter Piles

Piles are formed using hollow stem augering

equipment with outside diameters ranging typically from 12 to 16 inches.

The procedure is to install alternate piles (primary piles) then after

the grout has set, to install the remaining piles (secondary piles).

The piles may be augered in one or more lines as necessary to achieve

the desired watertightness and/or structural strength (see Figure 73a).

-1630
I. SET SOLDIER PILE IN PRE-EXCAVATED HOLE

2.EXCAVATE AND CONCRETE PANEL.

CAST ;N PLACE RE&ORCING PREFABRICATED

CONCRETE STEEL JOINT

Figure 70. Two step excavation in slurry trench

using precast soldier piles and tremie concrete.

-164-
SET SOLDIER PILE 1N PRE- EXCAVATED HOLE.

STEEL WIDE FLANGE

SECTION DRIVEN TO

BEARING STRATUM LEAN

IF REQUIRED. CONCRETE

EXCAVATE AND CONCRETE PANEL(REINFORC\NG IF REQUIRED BY

REBARS OR I- SECTION)

. .

\

CAST-IN- PLACE ) LREBAR

CONCRETE REINFORCEMENT

I- SECTION REINFORCEMENT (AFTER

THON AND HARLON, 1971)

.

I‘r

(2b) I’ ‘1 \

/

\

. .

CAST-IN-PLACE

CONCRETE

Figure 71. Two step excavation in slurry trench

using steel wide flange soldier piles and tremie concrete.

-165-
0I

02

03

04

s

REBARS

cIz=LL~

SET SOLDIER PILES AND REBARS

IN ALTERNATE PANELS.

CONCRETE ALTERNATE PANELS.

EXCAVATE INTERVENING PANELS

AND SET REBAR CAGE.

CONCRETE

Figure 72. One step excavation with soldier piles

(after Tamaro, 1974).

-166-
(a) SMALL DIAMETER (COMMONLY @@TANGENT” PlLES)

\

d. 12” TO 16”

---+-I~‘~

ROW 2 83 etc.

AS REQUIRED

GROUT FlLLEO WITH HOLLOW STEM AUGER

(b) LARGE DIAMETER ( CoMMotyLY w C0tuTima~~” 0~4 SECANT’

PILES 1

COUCRETE FlCLEO

Figure 73. Reinforced bored pile walls,

-167-
The grout is a mixture of Portland cement,

fluidifier, sand, and water. Sometimes a mineral filler may be added

as well. The grout is injected under pressure through the central hole

as the auger is withdrawn, and soil cuttings are removed from the

auger flights as they emerge from the ground. Immediately following

grouting, a cage of reinforcing steel or a wide flange beam section

is inserted into the wet mortar.

11.74.3 Large-Diameter Piles

Shaft diameters range from about 2-l/2 feet

to 4 feet. Depending upon the nature of the soil and ground water conditions, the excavation can be made with or without casing, either in

the dry or in a slurry stabilized hole. As is the case for small

diameter pile 6, alternate piles are installed first, followed by installation of inte rmediate piles.

Reinforcing is positioned following excavation,

then the hole is filled with concrete. Contiguous piles are used where

there is not great concern over watertightness. Overlapping piles

(secant piles) can be used to provide additional assurance of watertightness (see Figure 73b).

11. 74.4 Discussion

A bored pile waL1 has several advantages over

walls cast in slurry trenches.’ Because of the mini.mum exposure of

excavated soil prior to concreting, additional protection is provided

for heavily loaded foundations and/or in weak or unstable soils. Also,

selected piles may be carried to a lower elevation for bearing or toe

restraint.

-168-
CHAPTER 12 - INTERNAL BRACING

12.10 INTRODUCTION

In general, internal bracing is most often used in relatively

narrow cuts, where cross-lot bracing can be used without intermediate

s uppo rt , or in wide excavations where suitable anchorage strata are

not available for tiebacks. In cut-and-cover tunnel work, braces typically do not require intermediate vertical support. A continuous

horizontal wale is typically used to transfer loads from the ground

support wall to the brace. Wale levels are normally set about

10 to 15 feet apart vertically and brace positions are set about 15 to

20 feet apart longitudinally along the cut. Recent excavation work in

Washington used discontinuous wales to aid installation.

12.20 DESIGN CONSIDERATIONS

12.21 Types of Bracing

The most common braces are pipe or wide flange sections. Projects have been reported in Europe in which a concrete

slab, poured on the ground, later serves as the roof of the structure

or a floor level within the structure. The excavation is carried out

by mining beneath the slab. The technique is often called “under the

roof” construction.

12.22 Allowable Stresses

Controlling design criterion is the column-action of

combined axial and bending stress. In that regard, a pipe section is

more efficient than a wide flange section.

AISC Code design stresses are recommended for the

completed cofferdam at its maximum depth. Temporary conditions

arising from intermediate situations during the course of excavation

will justify a 20 percent overstress above the AISC Code values.

12.23 Connections

Connections and details are of critical importance

in an internally braced excavation. Improper connections between

strut and wale or between the wale and the support wall are perhaps

the most common causes of difficulties in braced excavations. They

can lead to twisting, buckling, and rotation of members. Figures 74,

75, and 76 present typical connection details.

-169-
SUFFICIENT WELD

LACE> WALE BRACE 2 -

TO HOLD IN PI

(WEB 

m---P-P

HORIZOTAL)

- -

-

7

m-------

STIFFENERS

(TACK WELD

IN PLACE)

Figure 74. Typical detail for horizontal brace

with brace web horizontal.

-170-
SUFFlClENT WELD I

TO t4DLD IN PLACE

,

BRACE 3

t

(WEB VERTICAL)

--- ---

a) BRACE DEPTH SMALLER THAN STIFFENERS

WALE FLANGE WIDTH, (TACK WELO IN PLACE)

PLATE

b) BRACE DEPTH GREATER THAN STIFFENERS

WALE FLANGE WIDTH. (TACK WELD IN PLACE) I

NOTE: DETAILS SIMILAR

FOR HORIZONTAL

BRACE.

Figure 75, Typical detail for horizontal brace with

brace web vertical,

-171-
INCLINED ’ KICKER” OR

‘SPUR BRACE” \ 1

PLATE IF

STIFF TB B

AT BRACE

(6 THICK)

WELDA=WELD BbC*

VERTICAL COMPONENT HORIZONTAL WOOD

OF BRACE LOAD SHEETING \

(3” THICK)

9LDIER

BEAM

)

Figure 76. Typical connection for inclined brace

and horizontal wale.

-172-
12.30 INSTALLATION

12.31 General

Typically, the first step is to attach brackets to the wall

to support the wale. Measurements are taken to cut the bracing members to proper length. The brace is cut to leave a few inches of clear

distance to facilitate placement. This extra space is taken up by plates

and wedges when final connections are made.

12.32 Installation Without Preloading

Cross-lot members are welded at one end and blocked

and shimmed at the opposite end. After the members are fitted in

place, steel wedges and plates are tack welded to hold everything

in place. For inclined braces (rakers) the member is welded at one

end (usually at the wale), and the reaction end cast into the concrete

slab. An alternate procedure would be to weld at the wale end and use

steel plates and wedges to make sure that the member is tight at the

reaction end.

In cases where wall displacements must be held to a

minimum, raker reactions against invert slabs are preferred to

reactions against concrete deadmen. If deadmen are used, they should

be preloaded to remove slack and to assure that the load can be accepted

without excessive movement.

12. 33 Installation With Preloading

The procedure is to jack to the desired load, to make the

connection, and then to remove the hydraulic jack. One procedure is

to jack to the desired load, and then to drive wedges between the member

and the wale until the jack load is essentially zero. A second procedure is to weld the connection tight while maintaining the jack load

and then to drop the pressure in the hydraulic jack, thus transferring the load through the connection to the wale. The second procedure may result in some wall movement as the load is transferred

although the magnitude of movement is generally small.

12.34 Preloading

Figures 77 and 78 show prestressing details for bracing.

Preloading is accomplished by loading hydraulic jacks to the desired

loads followed by securing the member with steel blocking, steel wedges, and welding. In the case of pipe struts, the connection can be

-173-
01 WEDGING

STIFFENERS

/ WALE

IPE 0.0. APPROXIMATELY

QUAL To*l.D. OF BRACE

PLIT SECTION OF PIPE WELDED IN PLACE

AFTER PRESTRESSING, LOAD STILL IN JACK.

b) TELESCOPING PIPE

Figure 77. Preetreeeing details for braces.

-174-
‘WALE

JACKING BRACKET

JvELD

.ATE

WEDGES

Figure 78. P restressing of pipe brace at

corners using brackets as reaction,

-175-
made using a telescoping strut or a split pipe which fits over the pipe

brace.

High preloads may cause overstressing of struts because of unforeseen job conditions or temperature effects. Accordingly,

the general practice is to preload bracing members to about 50

percent of their design load. This preload removes the slack from

the support system and at the same time reduces the risk of overstressing.

12.40 TEMPERATURE EFFECTS

12.41 General Background

Since temperature variations in strutted excavations

may easily be as great as 50oF, and even more if unprotected, the

changes in load accompanying temperature variation can be large.

A limiting case would be to assume a perfectly restrained strut (i. e.

no movement). The increase in load would therefore be equal to:

AP = AsEs (w x A0 F)

where:

AS

= area of strut

ES

= modulus of strut (30,000 ksi)

gC = thermal coefficient of expansion (6. 5 x 10 - 6

in/in/OF for steel)

AoF = change in temperature in degrees Fahrenheit

Since the soil behind the wall yields under the increased

load, the actual stress increase will be less than that indicated by

the limiting case condition.

12.42 Some Case Studies

The following table summarizes the load changes caused

by temperature variations on four projects.

-176-
Case Decked or Open

1. Chapman, et al (1972) Open

2. O’Rourke & Cording (1974) Covered

3. Jaworski (1973) Open

4. Armento (1972) Covered

12.43 Desipn and Construction Criteria

Load Variation

1.5 kip/OF

0. 5 kip/OF

20% f of measured

average

10% f of measured

average

Since the Peck (1969) diagrams have been developed

from measured maximum strut loads, the Peck diagrams implicitly

consider temperature variations. In critical cases where large temperature variations are expected in an unprotected excavation, strut

loads may be monitored to assure that overloading does not occur.

Although rarely done, struts may be painted with special reflective

paint or sprayed with water to prevent heat buildup in struts. Ideally,

strut installation (and preloading) should be at about the mean

temperature anticipated during the course of the job.

12.50 STRUT REMOVAL AND REBRACING

Strut removal (and rebracing) is an additional source of displacement. Factors controlling the amount of displacement are the

wall stiffness, the properties of the retained soil, the span distance

between remaining braces, and the quality and the compaction of the

backfill between the structure and the ground support wall.

-177-
CHAPTER 13 - TIEBACKS

13.10 INTRODUCTION

During the last 20 years the use of soil and rock anchors to

support side walls of excavations has increased significantly. Tiebacks (or anchors) have been used to support both temporary and

permanent excavations.

A tieback consists of 3 major components (see Figure 79):

1. An anchor zone which acts as a reaction to resist the

la.teral earth and/or water pressures.

2. A support member which transfers load from the wall

reaction to the anchor zone.

3. A wall reaction or point of support.

Since the wall reaction is the only part of the tieback in the

excavation, a tied-back system provides an open work area. At present the design of tied-back walls in the United States is based largely

on empirical relationships obtained from successful tied-back installations. The purpose of this chapter is to present the techniques

used in tied-back wall design and construction.

13.20 DESIGN AND. THEORETICAL CONSIDERATIONS

13.2 1 General

The design of tied-back systems involves several

major design considerations. The earth and water pressures acting on

the wall must be evaluated. There must be a suitable anchorage stratum. The soil-wall system must blCA stable with respect to overall

sliding stability.

must be evaluated.

The expected verti,cal and horizontal deformations

13.22 Deformations

In response to excavation, the unexcavated soil mass

will displace toward the excavation. Temporary support walls limit

but do not prevent movements. The consequences of movement and

the factors affecting soil and wall movements are discussed in greater

-178-
I

WALL q

REACTION

SUPPORT MEMBER (TIE)

ANCHOR

ZONE

Figure 79. Major tieback components.

-179-
detail in Volumes II (Design Fundamentals) and III (Construction

Methods) of this research.

The techniques used to predict movements are crude. Movements can be vastly influenced by specific conditions and construction

techniques which makes prediction of movements difficult. However,

certain factors in tied-back walls can be identified as affecting movements.

Some of the factors affecting performance of the wall and supported soil mass are listed below:

a. Vertical wall movement

b. Wall stiffness

c. Tieback prestress

d. Internal deformation of the soil block

e. Movement of the soil block

f. Ground loss associated with construction method

g* Volumetric strain

13.23 Overall Stability of Soil Mass

13.23.1 Circular Arc Analysis

A circular arc analysis of the stability of the

soils behind the anchors and below the wall should be performed. The

analysis should be performed using accepted circular arc methods of

analysis as described in basic soil mechanics texts.

13.23.2 Overturning Analysis

This method of analysis in combination with a

circular arc analysis is used to evaluate the stability of tied-back walls

throughout Europe. The basic method was proposed by Ranke and

Ostermayer (1968) who expanded on the work done by Kranz (1953).

Figure 80 schematically illustrates a failure by overturning.

-180-
Figure 80. Sketch of tied-back wall failing by

ove rtqrning.

,181-
Free Bodv Diagram and Forces

Figure 81 illustrates the free body diagram

and the forces acting on the free body. The free body is defined by

the points ACDE. Since the wall is not part of the free body, the forces

pA and Tnet

act in the directions shown.

The location of the free body diagram is predetermined in this method of analysis. Points A and E are located immediately above C and D. Point C is chosen as the point at which the shear

forces in the wall are equal to zero. In other words, point C represents

the point at which P -

Ah

T = P

d-h Ph’

Point D is uniquely defined as

the midpoint of the grouted anchor length. Therefore; in Figure 81

L I would be equal to Lz. In this method of analysis the entire anchor

load is assumed to be transmitted between points D and F.

below:

The forces acting on the soil mass are described

a. P represents the active pressure driving

force on the face DE from &e soil pressure. The force can be assumed

to be either inclined or horizontal.

b. W represents the weight of the soil mass

within the free body.

along the face AC.

c. PA represents the tot& active force acting

This ,resultant is inclined at the friction angle between

the soil and the wall.

resistance.

d. Sd represents the frictional component of soil

Full soil strength is assumed to be mobilized.

e. SC represents the full component of soil resistance from cohesive soil strength.

f. T represents the tieback force. The free body

cuts the tieback at points B and D. The force, Tnet, (Figure 81;) represents

the vector sum of the tieback forces at point B and point D. Since the

force at B must exceed the force at D, the force acts in the direction shown.

Safety in Terms of Tieback Force

The stability of the free body is evaluated in terms of

-182-
Figure 81. Free body diagram for a failure surface

in single anchor tieback system (internal free body).

-183-
the ratio of the maximum possible tieback force to the design tieback

T

force; F. S. = F. The maximum tieback force is determined from

des

the appropriate vector diagram.

a. Single Anchor. Figure 82 illustrates a

single anchor tied-back wall and the force diagram used to evaluate

the stability of the system. The vector diagram in Figure 82

defines the maximum tieback force consistent with the stability of

the earth mass. The design tieback force must be less than this

value, T

max.

A minimum factor of safety of 1.5 is recommended

in design.

The method described to this point has been

applicable to soil conditions where no water is present. If water is

present, the free body must consider the forces due to pore pressure,

and the analysis should be consistent with other basic methods of

stability analysis as described in many soil mechanics texts.

b. Multiple Anchor Systems. Figures 83, 84,

85, 86, 87, and 8 8 illustrate the free body and vector diagrams for

multiple anchor systems. The basic techniques are the same as for

a single anchor with the factors of safety defined as shown on the figures.

This method of stability analysis has several

apparent disadvantages. Among these is the rigid definition of the

failure plane. However, because of the method’s wide usage in Europe

with satisfactory results, it is believed that the method can be used to

evaluate wall stability against overturning. The method should be used

in combination with other methods evaluating sliding stability.

13.23. 3 Sliding Wedge Analysis

Free Body Diagram and Forces

A generalized free body diagram is presented in

Figure 89. The wall is included as part of the free body, and therefore,

the wall forces, H, and V, are included. The passive soil resistance is

included as is the net tieback force, T

net*

The sliding wedge analysis does not specify the

location of the failure surface as did the previous overturning analysis.

Several failure surfaces can be analyzed for a given anchor geometry.

-184-
A E

ti=$ on failure pl .ane

T

F.S. == > 1,s

Tdes

After Kranz (1953)and

Ranke & Ostermayer (1968)

(b)

Figure 82. Single anchor free body diagram with

appropriate vector diagram

(safety in terms of the tieback force).

-185-
d l $ ON FAILURE PLANE

Figure 83. Free bodies and forces for two completely

independent anchors (after Ranke and Ostermayer, 1968).

-186-
de8 a. Upper Tieback

T1

F. S. = T max 2 1.5

‘dce

Note: only the directions of Sq and

T are known.

max

d = 9 on failure pI8ne.

B. Lower Tieback

T,

L

F. S. = -2T 1.5 w2

2

des

P

a2

i

1-2

2

-7

es

max

IS

92

Figure 84. Vector diagram for case of two

completely independent anchors (safety in terms

of anchor force) (after Ranke and Ostermayer, 1968).

-187-
a( = B ON FAILURE PLANE

Pa - 2

Figure 85. Free body diagram with forces acting on the

bodies for the case of one independent anchor

(safety in terms of the tieback force)

(after Ranke and Ostermayer, 1968).

-188-
T

1-2

F.S. = max

T1 +T1 de8 des

T

1

F. S. = T max 2 1.5

1

des

d= q on failure plane.

1

des

w

2

1. 5

Figure 86. Vector diagrams used to evaluate the

stability of case with one independent anchor

(safety in terms of tieback force)

(after Ranke and Ostermayer, 1968).
d= jd ON FAILURE PLANE

Figure 87.. Free body diagram for anchor system with a

complex failure surface (safety in terms of the tieback fQrce).

-190-
-

J

S

‘Pa.2 Cl

-Be--

T2

F.S. = max

T2

1.5

des

F. S. =

L

 l-2

max

Tl tT

des

2

des

1.5

o(= 9 on failure plane

Figure 88. Vector diagram for a complex failure surface

(safety factor in terms of tieback force)

(after Kranz, 1953 and Ranke and Ostermayer, 1968).
Figure 89. Free body diagram for a failure surface

in a single anchor tieback system

(free body outside of wall).
Safety Factor in Terms of Soil Strength

Traditional studies of the stability of soil masses

express the factor of safety of the soil mass in terms of the available

S

and mobilized soil strengths; F. S. =

avail

s l mob

a. Single Anchor. In order to evaluate the force,

T = T (see Figure 89), it is assumed that the anchor load is distr%&ted &enly along the length of the anchor. Therefore, the magnitude of the forces, Ti and To,will depend upon the location of the

failure surface with respect to the achor zone.

Figure 90 shows the vector diagrams used to

analyze a single anchor system. For cohesive soil, the factor of

safety can be defined as the ratio of the undrained shear strength to

the mobilized shear strength along the failure surface.

For cohesionless soils, the factor of safety becomes: F. S. = G tan B L tan e6 .

R tan& tan&

b. Multiple Anchor Levels. Figure 91 illustrates

how the stability of a three anchor level system would be calculated using

this method. For simplicity, the example is for a cohesionless soil.

Using this method of analysis several trial failure surfaces can be analyzed

rapidly. The recommended factor of safety for this method of analysis

is 1.5.

13.23.4 Discussion

The evaluation of the stability of a tied-back

earth mass is a trial and error process involving the use of several

analytical techniques. The stability methods presented in this section

are those that have been commonly used in practice. It is recommended

that several techniques be used to evaluate tied-back wall stability.

13.24 Tieback Anchorage Design Considerations

13.24. 1 Suitable Anchorage Strata

Experience has shown that virtually all rock types

can be used as anchorage zones; however, not all soil deposits are suit-

-193-
S c avail -32 xL Shear strength available N tana

F. S. -s

c mob -Su mobXL

F. S. =Shear strength mobilized = N tano(

(a) f-b)

Cohesive Soil Cohesionless Soil

Note: Consideration of Horizontal and Vertical Wall

Forces Increases Factor of Safety

(cl

Soil With Both Cohesive and

Cohesionless Properties

Figure 90. Vector diagrams used in analysis with

factor of safety defined in terms of soil strength.
I

9’ ul

I

(a)

Free Bbdy (Cohesionless Soil)

F.S. = Ntan# = tan 6

N tan& tan&

(b)

Vector Diagram

Figure 91. Analysis of a multiple level anchor system

(safety factor in terms of soil strength).
able. The following list summarizes the appropriateness of various

soil and rock types for location of anchors.

1. Soft to medium clays are generally not suitable anchorage strata.

2. Stiff clays may or may not be suitable for

anchorages depending upon the project particulars (allowable movements and loads).

3. Loose cohesionless soils have provided

successful anchorages in some cases: however, other cases indicate

that these soils are not satisfactory.

4. Very stiff to hard clays and medium to very

dense granular soils are preferred anchorage strata.

anchorages.

5. Virtually all rock types provide suitable

13.24.2 Location of Anchors

In U. S. practice, anchors are generally located

beyond a line extending at a 30° to 45O slope from the vertical from the

base of the excavation to the ground surface (see Figure 92). Recent

cases indicate a more common use of 350 to 400 as an angle of inlination for the slip surfaces in granular soil deposits. In cohesive soil

deposits, anchors are often founded well behind 450 slip lines. For

normal anchor lengths this procedure results in a stable soil mass.

13.24.3 Soil Anchors

Load Transfer Mechanisms

The anchor transfers the tieback load to the soil

through two basic mechanisms: 1) frictional resistance at the anchorsoil interface and 2) end bearing where anchors have a larger diameter

than the initial drilled shaft diameter. The actual load transfer

mechanism (s) varies with anchor and soil type.

Table 9 summarizes the basic anchor-types

with respect to the soil types in which they are used. The remaining

sections discuss the specific methods used to estimate the load

carrying capacity of each of the anchor types.

-196-
” ,~:,i/,”

/ ,/ ,,(I f‘/ ’ , ,’
Table 9. Summary of tieback types and applicable soil types.

Method

. LOW PRESSURE

Diameter (inches) Grout

Shaft Bell Gravity Prsssors Suitable Soils Load Transfer

Type Type Concrete (psi) (1) for Anchorage Mechanism

Straight Shaft Friction 12-24” NA A NA Very stiff to hard clays Friction

(Solid stem auger) (30 - Dense cohesive sands

60cm)

Straight Shaft Friction 6-18” NA NA 30 - 150 Very stiff to hard clays Friction

(Hollow stem auger) (15 - (200 - Dense cohesive sands

45cml 1035kN/m2) Loose to dense sands

(12-14”

most

common)

Underreamed Single

Bell at Bottom

12-18” 30-42” A

(30 - (75 -

45cn-I) 105cm)

N A Very stiff to hard co- Friction and

hesive soils bearing

Dense cohesive sands

Soft rock

Underreamed Multibell4-8” 8-24” A

(10 - (20 -

20cm) 60~1x1)

NA Very stiff to hard co- Friction and

hesive soils bearing

Dense cohesive sands

Soft rock

2. HIGH PRESSURESMALL DIAMETER

Non-regroutable (2) 3-8” N A NA 150 Hard clays Frrction or friction

(7.5 - (1035kNNI Sands :md bearing in

ZOcml m21 Sand-gravel formations permeable soils

Glacial till or hardpan

Regroutablc (3) 3-8” N A N A 200-500 Same soils as for non- Friction and

(7.5 - (1380 - regroutable anchors bearing

20cml 3450kN/m2) plus:

a) stiff to very stiff

clay

b) varied and difficult

soils

(1) Grout pressures are typical

(2) Friction from compacted zone having locked in stress.

Mass penetration of grout in highly pervious sand/gravel forms “bulb anchor”.

(3) Local penetration of grout will form bulbs which act in bearing or increase effective diameter.

A - applicable

NA - not applicable

-198-
Large Diameter Anchors

Large diameter anchors can be either straightshafted, single-belled, or multi-belled. These anchors are most

commonly used in stiff to hard cohesive soils that are capable of

remaining open when unsupported; however, hollow flight augers can

be used to install straight-shafted anchors in less co,mpetent soils.

Figure 93 schematically illustrates several large diameter anchors.

The methods used to estimate the ultimate ,puLLout

capacity of large diameter anchors are largely based on the observed

performance of anchors and are, therefore, empirical in nature. The

following equations can be used to estimate anchor load capacity; field

testing of anchors is required to determine true anchor capacity.

a. Straight- shafted Anchor

pU

=wS 7fdsL

U S

where:

dS

= diameter of anchor shaft

LS

= length of anchor shaft

s = undrained shear strength of soil

-tl

DC= Reduction factor in Su due to disturbance,

etc. d = 0.3 - 0.5 (Hanna, 1973a; Broms,

1968; Littlejohn, 1970a; Neely and

Montague-Jones, 1974).

b. Belled Anchor

pU

=WSuVds Ls + n/4 (D2 - ds2) NcSu (Littlejohn, 1970a)

where:

d

s’ L

s, Su and d. are as before

D = diameter of anchor bell

NC= bearing capacity factor = 9

-199-
.-I

/ ESTIMATED LOAD FOR

ANCHORS IN COHESIVE SOIL

Pu= &SuL fl d

S

S

d = 0.3-O. 5

(a)

Friction Anchor

(b)

Belled Anchor

Pu= &Su Ls r 4+ n’ (D2-d,2) NcSu

4

d = 0.3-O. 5

N = 9

C

Pu= d SuL s n’d,+ --$ (D2-d;) NC

t 6 S, ‘l-l- DLu

Multi-Belled Anchor

Schematic representation of large diameter anchors .
c. Multi-belled Anchor

P

U

= OZS WdsLs t n/4 (D2 -

U

ds2) NcSu ‘,&u r DL

-cl

where:

dsI Ls’ sus oc tN and D are as before c

LU

= length of underreamed portion of anchor

P

= reduction factor in Su for soil between

underream tips = 0.75 - 1. 0 (Littlejohn,

1970a; Bassett, 1970; Neely and MontagueJones, 1974)

In order for failure to occur between the underream

tips, the tips must be spaced at 1.5 - 2. 0 times the belled diameter with

the bell diameter equal to 2.0 to 3.0 times the shaft diameter.

Small Diameter Anchors

Smatl diameter anchors are generally installed in

granular soils with grouting taking place under high pressures (usually

greater than 150 psi (1035 kN/m2). The anchor capacity will depend upon

the soil type, grouting pressure, anchor length, and anchor diameter.

The way in which these factors combine to determine anchor load is not

clear; therefore, the load predicting techniques are often quite crude.

The theoretical relationships in combination with the empirical data can

be used to estimate ultimate anchor load. Figure 94 schematically

illustrates several small diameter anchors.

a. Theoretical Relationshins

1. No grout penetration in anchor zones

pU

= pi 7/ds Ls tan (5

e

(Littlejohn, 1970a;

Broms, 1968)

where:

d = diameter of anchor shaft

S

LS

= length of anchor shaft

‘e = effective friction angle between soil and

grout

P .1

= grout pressure

or P

U

= Lsnl tan de (Littlejohn, 1970a)

where:

nl = 8.7 - 11.1 k/ft (127 - 162 kN/m)

-201-
(4

Friction Anchor

(No Grout Penetration)

(b)

Bulb Anchor

(Grout Penetration)

Grout Pipe (high pressure

Tie post grouting)

(cl

Regroutable Anchor

(Local Grout Penetration)

\ Primary Grout Zone

(low pressure)

Figure 94. Schematic representation of small diameter anchors.

^ ^-
2. Grout penetration in anchor zone (very pervious

soils)

pu = A 6v nDLs tan de t B -Qv@ end7//4 (D2 - ds2)

(Littlejohn, 1970a)

where:

ds, D, Ls and de are as before

C

= average vertical effective stress at anchor

entire anchor length

@v@ end = vertical effective stress at anchor end

closest to wall

A = Contact pressure at anchor soil interface

effective vertical stress ( 3,)

Littlejohn reports typical values of A ranging

between 1 and 2.

B = bearing capacity factor similar to N but

smaller in magnitude. A value of q

B zN

i!hT

is recommended provided

h’& 25;; where h is the depth to anchor.

Since the values of D, A, and B are difficult to

predict, Littlejohn (1970a) also suggests:

pU

= Ls n2 tan de

where:

n2

= 26 - 40 kips/ft (380 - 580 kN/m)

L =3

S

- 12 ft (0.9 - 3.7m), D= 15 - 24 inches

(400 - 610 mm).

depth to anchor = 40 - 50 ft (12.2 - 15. lm).

b. Empirical R.elationships

Figure 95 presents an empirical plot of the load

capacity of anchors founded in cohesionless soils. This figure was

developed by Ostermayer (1974) and represents the range of anchor

capacities that may develop in soils of varying densities and gradations.

-203 -
Iery Dense

Iense

EXe

vfed. Dense

hnre

lense

-Medium to

-Coarse Sand

,led. Dense

5 1 0 1 5 2 0 21 3

Length of Anchor, Ft.

Diameter of Anchor 4”-6”

NOTE: Ift. = ,305 m Depth of Overburden L 13 feet

tin. = 2.54cm

I Wtt = 14.6 kN/m

Figure 95. Load capacity of anchors in cohesionless

soil showing effects of relative density,

gradation, uniformity, and anchor length

(after Ostermayer, 1974).

-204-
The following table summarizes the load capacity of

single injection small diameter anchors :

Ultimate Load

Soil kips /ft kN/m

Clean sand/gravel soils 10 - 20 145 - 290

Clean medium to coarse 7 - 15 100 - 220

sands

Silty sands 5 - 10 70 - 145

C . Re gr outable Anchor s

Regroutable anchors are small diameter anchors

that allow the Load carrying capacity of the anchor to be improved after

installation and testing. Figure 94 schematically illustrates a regroutable anchor.

Jor ge (19 69) reported an improvement of anchor

load capacity in both cohesionless and cohesive soils with a regroutable anchor. Figure 96 presents a summary of the results with data

on very stiff clay from Osterxnayer (1974).

A summary of data on cohesive soils for regroutable anchors is presented in Table 10 and Figure 97. These values

can be used to estimate regroutable anchor loads.

13.24.4 Rock Anchors

R.ock anchors may fail in any one of the following

modes:

1. Failure of the rock mass

2. Failure of the grout-rock bond

3. Failure of the grout-steel bond

4. Failure of the steel tendon

The last two modes of failure are true of all anchors

in Section 13.25.

Failure of the Rock Mass

and will be discussed

The criterion for failure in a rock mass is based on

the weight of the rock contained within a specified cone emanating from
20

I6

12

8

I .---- 1 I 1 I

I I

I

I ----- I

I

-+- -+- --Y- I

I I Sands I & Gravels ;

I‘

I

I

I

I

LJU I 400 600

Maximum Grouting Pressure, psi

NOTE: I k/ft * 14.6 kN/m

I psi = 6.90 k N/m2

I in = 2.54 cm

BRUSSELS

“**” Medium Sand

MAR COULE

--6 Alluvium & Marl

BERCY

-Cl Se& Alluvial

Sands

ST. JEAN-DE-LUZ

-l Clayey Sand and

Gravel

h ST. CLOUD BRIDGE - -

Soft Chalk

BAGNOLET

- - 4 Gypsum Marl

w-m--.Oetermayer (1974)

Very Stiff Clay

wI = 48 to 58

wP = 25 to 35

Figure 96. Ultimate anchor capacity as a function of grout pi-essure.
Table 10. High pressure small diameter tiebacks

in cohesive soil (after Ostermayer, 1974).

Typical Skin Friction

(per square foot of grouted zone)

Soil Type

Without With

Post-Grouting Post-Grouting

Marl Clay - medium plastic

(ti = 32 to 45; w = 1 4

1

to 25)

P

Stiff 2200 - 3500 -----m----

Very Stiff 3500 - 6500 ----------

Marl Sandy Silt - medium plastic

(w’1=45; wp=22)

Very stiff to hard 6500 - 8500 8500 - 10,500

~ Clay - medium to highly plastic

~ (w 2 45 - 59; w = 1 6 -

1 P

35)

Stiff 500 - 2000

Very Stiff 2000 - 3000 3000 - 5500

Note:

1. Tieb,acks 3-l/2” to 6” 0. D.

2. Values are for lengths in marl - 1 5 to 2 0 feet and

for lengths in clay - 25 to 30 feet

3. 1 psf = 0.48 kN/m2

1 in = 2.54 m

1 ft = 0.305 m

-207.
8

6

Anchor Load in Cohesive

Soil I

With

, Post\\\\

Grouting

v

\

I I .

Med. M d

Stieff ’

Stiff Stiff V. Stiff V. Stiff

V. Stiff to hard

(5) No. of Test

Anchors

Without Post

Grouting ////

Without

O------O EF%ting

With

0-a Post

Grouting

Clay Consistency

Ostermayer (1974)

11 Marl Clay wl= 32 to 45; w = 14 to25

12 Marl Sandy Silt w1=45; wp= 22

13 Clay-medium to

highly ‘plastic

w145 to 59; w;=16 to 35

Goldberg-Zoino &,Associates NOTE: I K/f?= 47.9 KN/m*

21 Medium Plastic Clay

Figure 97. Effect of post grouting on anchor capacity,

-208-
a point on the anchor and extending to the top of the rock. Figure 98

illustrates the geometry for this case. The criteria used to evaluate the

value of the angle, 8, and the location of the apex of the cone vary with

the type of rock, method of load transfer, and designer (Littlejohn, 1975).

Typically, the design value of 8 will vary from 60~

to 90° although in badly fissured or jointed rock the design criteria may

be significantly different. If the weight of the rock within the contained

cone is greater than the design anchor load, then the anchor is generally

believed to be safe since any cohesion or other rock strength properties

have been ignored. However, if desired, a factor of safety can also be

a,pplied to the weight of the rock mass and the anchor load. This measure

may be required if the rock is badly jointed.

Grout-Rock Bond

Most rock anchors are straight-shafted friction

anchors of 4 inches to 6 inches in diameter. In the past it has been

assumed that the load is transmitted uniformly along the grout-rock

interface, and most anchor design has been based upon this assumption. Littlejohn (1975) reports the results of studies performed by

several authors that indicate that this assumption may not be valid.

However, in the absence of more detailed information the established

methods should still be used. The designer should be aware of the

potential problems of local debonding, R.igid field testing should be

establish anchor adequacy.

is:

The equation used to estimate anchor capacity

pu = 7/ds Ls ‘skin

where:

ds = diameter of anchor shaft

LS

= length of anchor shaft

d

skin = grout-rock bond strength

The values of skin friction, gskin, for various

rock types are summarized in Table 11.

In soft rock, it is also possible to form belled or

multi-underreamed anchors. The equations governing the ultimate loads

in these rocks are given in previous equations (Section 13.24.3 ). In

these cases, the cohesive strength of the rock becomes the controlling

quantity.

-209-
e= 600 - 90'

I? = Apex of Cone

(Varies from Midpoint

to Base of Anchor)

From Littlejohn (I 975)

PU =x x VOLUME OF ROCK

IN CONE

Figure 98. Schematic drawing of design quantities for

failure in a rock mass.

-210-
Table 11. Typical values of bond stress for selected rock types.

Rock Type

(Sound, Non-Decayed)

Granite & Basalt

Limestone (competent)

Dolomitic Limestone

Soft Limestone

Slates and Hard Shales

Soft Shales

Sandstone

Chalk (variable properties)

Marl (stiff, friable, fissured)

Ultimate Bond Stresses Between

Rock and Anchor Plug (dskin)

7

250 - 800 psi

300 - 400 psi

200 - 300 psi

150 - 220 psi

120 - 200 psi

3 0 - 120 psi

120 - 250 psi

3 0 - 150 psi

2 5 - 3 6 psi

.-_. ._-.. ._--

1 psi = 6.90 kN/m2

Note: It is not generatly recommended that design bond stresses

exceed 200 psi even in the most competent rocks.

Data is summary of results presented in:

1. Inland-Ry,erson (1974 - AC1 Ad Hoc Committee)

2. Littlejohn (1970)

3. Littlejohn (1975)

-711-
13.24. 5 Safety Factor of Soil or Rock

Safety Factor With Respect to Shear

a. Soil Anchors. In cases where there has been

considerable experience with the soil and anchor type and where 5 percent or more of the anchors are to be proof-tested to 150 percent of

design load, the anchors should be designed with a minimum factor of

safety of 2. The design parameters should be based on previous pullout

tests or the results of pullout tests performed on the site.

In special cases where a comprehensive field

testing program is specified, the factor of safety may be reduced to

1.75. The general requirements for the reduction in the factor of

safety are extensive experience with the anchor in the soil type and a

minimum of five carefully monitored pullout tests (or to 175 percent

of design load) for the full anchor load. Production test monitoring

including creep measurements is also required.

b. Rock Anchors. The factor of safety that should

be applied against pullout of a rock anchor depends upon the rock type and

the type of failure. For failure in’the rock mass itself, a factor of safety

of 1. 1 applied to the weight of the rock mass inside the cone of rupture is

considered adequate because of the beneficial contributions of rock shear

strength. In heavily jointed rock the factor of safety may be increased.

The factor of safety applied to the grout-rock

bond should be a minimum of 2.0. This factor of safety is recommended

because of stress buildup and debonding.

Safety Factor With Respect to Creep

In some instances the design anchor load may

be based on the creep of the anchor. At a particular load, the anchor

may have an adequate factor of safety against pullout; however, the

anchor may creep causing a loss of load and result in wall movement.

The rate of creep of the anchor must be kept to acceptable values. T o

date, the criteria for determining acceptable creep rates are based upon

field observations. R.ecommended creep criteria appear in the section

on field testing of anchors (see Section 13.40).

13.24.6 Discussion

The theoretical and empirical methods for predicting

anchor load are used as a first step in anchor design. Final anchor capacity should be verified by testing of each anchor beyond its design load.
13.25 Tendon Design and Load Transfer

13.25.1 Anchor Zone and Bond Free Zone

The anchor zone is that part of the tieback which

is grouted in the soil and through which the tieback load is transferred

to the soil. The transfer of load to the grout zone can be done either

through bonding forces between the tie and the grout (tension anchor) or

by a device or plate at the base of the anchor rigidly attached to the

tie (compression anchor).

The bond free zone refers to that portion of the

anchor inside the theoretical or assumed slip line. Since anchor resistance will not be developed in this area when the wall reaches its full

depth, it is unconservative to test load the anchor if load can be transferred through this zone during testing. Therefore, the following methods

are used to insure that all load is carried in the anchor zone.

1. Prevent tendon load transfer.

a . Wrap the steel tie in a pla.stic sheath to

prevent bonding in this zone.

2. Prevent compressive force from developing.

a . Do not backfill; or wash out grout in bond

free zone.

b. Backfill the bond-free zone with sand or very

lean cement grout to within a foot of the back

face of the wall.

Although the technique of grouting to the back of the

wall has been used, the technique is not as an effective a debonding technique as the others mentioned. Figure 99 illustrates the recommended

treatment for bond free zones.

13.25.2 Steel Tie Member

The design of steel tie members depends on the

ultimate load that the member can carry in tension. The exceptions to

this rule would be where the bond between tie and grout is the controlling factor (rare) or where end connections cause a significant decrease

in steel tie area. Bonding is not a significant problem until large anchor

capacities are required. Bond strengths will typically be between 200

and 250 psi (1.38 - 1.75 N/mm2) for cement grouts and concrete.

-213-
No Backfill Within 1 ’ of

Wall,

Grout

. /

\ITic Member \ Sheathing ‘\

,

1 Tie Member

\\I-\ Sand or Lean ~~n,nn+

l.AA.l..llk

\’ ’ Grout (Backfill)

Figure 99. Recommended treatment for bond free zone.

-214-
High strength steel wire strands, cables, and

bars are most commonly used for tie members. Often the choice of

the type of tie is limited by the method of installation or convenience.

Table 12 Lists typical properties and dimensions of steel wires, strands,

and bars for tie members.

i3.25.3 Grout and Concrete

Resin Grouts

Resin grouts are used because of their quick

setting times of ten to twenty minutes (for 80 per cent to 90 percent ultimate

strength). This allows anchor testing shortly after installation. The

strength of the resin grouts is comparable to that of concrete or cement

grouts. The major disadvantage of resin grouts is their relatively high

cost.

Cement Grouts

Cement grouts are most commonly used in

small diameter anchors . Generally, high early strength cement is mixed

with water to form a neat cement grout. A low water to cement ratio

is used during the mixing process. The anchors are usually tested 24 -

72 hours after installation of the grout. While expansive additives have

been used in grouts, recent experience has shown that such additives are I

not necessary to the satisfactory performance of the grout or anchor.

Concrete

In large diameter anchors greater than ten

inch (25cm) diameter the anchor zone is generally grouted with a mixture

of high early strength cement, water, and sand or fine gravel. This

concrete mixture is used because the sand or gravel filler is cheaper

than cement and does not appreciably reduce the strength of the grout.

13.25.4 Factors of Safety

Table 13 presents recommendations for design steel

stresses in temporary ties. These values represent minimum factors

of safety and should be increased for permanent tieback installations or

critical temporary anchors. For permanent anchors, it is recommended

that a minimum factor of safety against ultimate load of 2.0 be applied

for the design loads in the steel members.

-215-
Table 12. Typical steel properties and dimensions for ties.

TYPe Diameters Ultimate Stress Yield Stress Ultimate Load Yield Load

of Tie (inches). fu (ksi) fy (% fu’ Wps) hips)

Wire (1) .25” 240 .80 11. 8 9.4

Cables or .25” 270 . 85 10.3 8. 8

Strands (2) .50” 270 .85 41.3 35.1

.60” 270 .85 58.6 49. 8

.50” 160 . 85 34.1 29.0

.625” 230 . -85 70.6 60.0

Bars or 1.00” 150 . 8 5 127.8 108.6

Rods (3) 1.00” 160 .85 136.3 115.9

1.25” 150 . 85 187.5 159.4

1.25” 160 .85 200.0 170.0

1.375” 150 .85 234.0 198.9

1 . 25” - 132 . 85 165.0 140.2

1

Wire Members: ASTM A-421 Note: 1 inch = 25.4 nxn - ,,

Cable or Strands: ASTM A-416 1 ksi = 6.898 N/mm&

Bars or Rods: ASTM A-322 1 kip = 4.45 kN

(1) Many wires are used in anchor to obtain load carrying capacity.

(2) Several cables or strands are used in an anchor.

(3) There are many bar or rod types and manufacturers. The data presented

here is typical and is not meant to indicate the only bar types available.
a

-21?-
13.25.5 Corrosion Protection

Corrosion protection for temporary earth or rock

anchors is generally minimal. In those cases where the anchors are

expected to be in use for two years or less, the only corrosion protection

consists of greasing and sheathing the ties in the bond free zone. Where

unusually corrosive soil and water conditions are encountered, additionaL

corrosion protection methods may be used. Specially treated grout, steel

members or extra steel may be used to ensure that the anchors will

perform adequately. Compression anchors have been used to prevent

radial era eking of the grouted portion of the anchor. These cracks

result in faster corrosion of the anchor parts.

13.30 CONSTRUCTION CONSIDER.ATIONS FOR. TIEBACKS

The purpose of this section is to describe the basic construction

procedures and techniques used to install tiebacks. Since the major

difference between tied-back wall construction and internally braced wall

construction is the installation and testing of the tiebacks, this section

is limited to a discussion of the construction considerations for each of

the different tieback types.

13. 31 Tied-Back Walls Versus Internally Braced Walls

The basic construction sequences and procedures are the

same for both wall types.

1. Install wall (soldier piles, steel sheeting, slurry wall,

etc. ).

2. Excavate to support level.

3. Install tieback, strut, or raker.

4. Repeat steps 2 and 3 until excavation is complete.

The differences between the wall construction methods are

very minor and primarily reflect ways of installing tiebacks through the

walls. For example, one common procedure is to place tiebacks between

back-to-back channels --set either vertically as soldier piles or horizontally as wales (see Figure 100).

13. 32 Construction Techniques Common to Tiebacks

Stated very simply, the construction sequence for the

installation of a tieback consists of the following steps:

-218-

1. Excavate a hole for the tieback

2. Install the tendon (tie)

3. Grout the anchor to the specified point (usually to the

“slip” line)

4. Backfill the bond free zone with a weak material to

prevent bonding.

5. Tension and test the tie

6. Make final anchorage at the wall.

The type of tie, the treatment of the bond free zone, the

method of tensioning the tie, and anchoring of the tie at the wall are all

virtually independent of the type of tieback.

13.32.1 Compression or Tension Anchors

Compression anchors are those where the entire

anchor load is transferred to the tie at the base of the anchor. The

tie is connected to a plate or a point which is embedded in the anchor

base. The plate or point transfers all of the anchor load to the tie with

no bond allowed to develop between the tie and the grouted zone except

at the very base of the anchor.

In a tension anchor the load transfer from the

an.chor to the tie is accomplished through the steel-grout bond acting

over the surface area of the tie. Generally, the anchor geometry is

such that no problems are’encountered in obtaining the desired load in

the tie through the steel-grout bond. However, when bonding problems

are anticipated, the wires or cables may be unraveled at the end to ensure that there is enough surface area for bonding. The tensile and

shearing forces in the concrete are larger for a tension anchor, and

hairline cracking in the anchor has been observed in these anchor types

(Ostermayer, 1974).

A partial compression anchor is one in which a

plate or point is fixed to the end of the tie to help transfer Load. However,

bonding of the tie to the grout is allowed so that such anchors have

characteristics of both compression and tension anchors.

13.32.2 Centering Ties

Spiders or other centering devices are required in

larger diameter holes. This is particularly true for wire or cables be-

-220-
cause of their flexibility. In small diameter holes steel bars or rods

often require centering while cables and wires generally will not because

of their irregular surface.

13.32.3 Tendons

The different tendon (tie) types and their material

properties have been described in an earlier section (see Section 13.25).

The choice of which tendon type to use (bar, strand, or wire) is virtually

independent of anchor type. Recentle, the use of high strength steel rods

has increased since these rods can be easily threaded into detachable

points in the base of the anchor and allow for easy connections at the wall.

13.32.4 Anchorage at Wall

Friction connections have ridges or teeth that

grip the tendon and cut into it slightly. The steel area is therefore

reduced which leads to increased stresses at that point in the tie. Figure

101 illustrates a typical friction connection.

Button head connections are generally preferred

over friction connections where substantial retesting of anchors is anticipated. The connection is less likely to slip or cause damage to the

tendons. Figure 102 illustrates a typical button head connection.

Threaded conections also allow much retesting

of anchors without damage to the tendon. The- design steel area for the

tendon is based on the interior area of the threads. Figure 103 illustrates

a threaded connection. In practice, threaded connections are more

commonly used than button head connections.

13.33 Construction Techniques and Procedures for Different

Anchor Types

Table 14 summarizes the equipment installation techniques

and preferred soil types for common anchor types. Figures 104 through

107 show several tieback installations and some of the installation

equipment.

13.40 FIELD TESTING

13.41 Reasons

The major reasons for field testing are:

-221-
Figure 101. Frlction * connection used to tie anchor to wall.

-222-
Figure 102. Button-head connection for wire ties.

-223 -
.

* l 0

a 0 *

l . 0 . 0 . 0 0 b 0 . . . . . .

. ..e..

. . . ...*

. . ..b.o

0 l 0 . 00 l

l e 0 0 0 i 0

.

0

9 * . 0 0 . 0 .L

l #..*....

~*b.*ea~*.

.* 6 * . 0 0 0 c 0

J

.

rr....r*cre

.oo*.....*

l . . . . ..~...

0 l b. 0 0 0 0 l e I.

, b . I e 0 0 . 0 . * , . :::*

.eCee..... . *

l *e***.*c

..+e~o.**.

l e . . . . . l . . . . . . . . .

9 b . . . . a I , /.::.

w,rs....

..*ì.~~

HO..... . .*ë,.

1 8 l 0 e 0 /..

r

.*..........

.,.........

.r..e......

l 0e0.0.0..

l . . . . . . ..b

0 . . . . . . .

. . . . 0 . 0 0 . ..o..

*e.oo.

- b 0 0 - .

0 l . 0 . .

4 0

e 0 . .

l 0 . l . l

0 .

. 9

De sign

Tic

Figure 103. Threaded connection for tying anchor to wall.

-224-
Table 14. Typical equipment for construction of tiebacks.

Range in

Diameter

(typical)

Angle of

Inclination

(to horizontal) Bond Free Zone

SpXe*S

Preferred Soil and

TYPO

Typical

Grout Type

Anchor

TYPO

Straight

ShaftLarge

Diameter

(1) Solid

stem

Augers

Equipment Lengths Plate

00 - 9o”

(better at

shallow

angles)

12”-24”

(30cm60cm)

50’-130’

(typical)

(15x-n -

40m)

Pumped concrete.

Spacers and

plate generally used.

Lean concrete or

sand backfill.

Plastic sheathing.

Competent cohesive soil which

can remain open

unsupported.

Truck-mounted

crawler-mounted

or crane-supported augers

guided by Kelly

Bars.

Truck-mounted

crawler-mounted

or crane-supported with

guides.

f,“- 18”

(15 cm45 cm)

12”- 14”

most

c CnnmOn

Reported

to 160’.

(50 In)

No spacer

*eCeSSZ&*y

since hollow

stem serves

as guide.

Points are

generally used

in anchor.

00 - 900

(proprietary

methods may

not be able to

achieve lower

angles)

Lean concrete or

sand backfill.

Plastic sheathing.

High strength

concrete

pumped water

p*l?SSU*e

through hollow stem

-150 psi or

less. (1035

kN/m2)

(2, Hollow stem

Augers

Preferred in

competent cohesive soils.

Often used in

sandy soils.

Pumped concrete.

Truck-m&ted

crawler-mounted

or crane-supported augers

with guides.

Belling equipment same as

used for caisson

work.

12”-24”

(30cm -

60cm)

Shaft

30”-42”

(75cm -

105cm)

Bell

Belled

Anchor

Spacers used

to center ties.

Plates or

washers usually aid load

transfer.

Lean concrete or

sand backfill.

Plastic sheathing.

Generally installed at

angle (300 -

60’)

Typical

length to

bell of

approximately

50’ (15m).

Lengths

up to loo’+

(30x-n) in

California,

Total

lengths in

excess of

50’ (15I-n).

Spacing

between

bells

approximately

1.5 - 2.0

x diameter

bell.

Competent cohesive soils which

can remain open

unsupported.

cement grout

or concrete.

[Concrete for

larger diameter anchors]

Spacers used

to center ties.

Plate used in

some methods

to transfer

entire load.

Generally installed at

angle (300 -

60°)

Lean concrete,

weak cement grout,

or sand. Entire

tie length except

for plate is unbonded in some

methods.

4”- 8”

(10cm -

20cn-l)

Shaft

*~~- 24”

(20cm -

60~1x1)

Underreams

I,,>- 1.5ds

MultiUnderreamed

Anchor

(MultiBell)

Competent cohesive soil or rock

that can remain

open unsupported

To date experience in United

Kingdom.
Anchor

TYPO

Table 14. Typical equipment for construction of tiebacks, (Continued).

Range in spacers Angle of

Preferred Soil Diameter Typical and Inclination

Type Equipment (typical) Lengths Grout Type Plate (to horizontal) ‘Bond Free Zone

SIIdl

Diameter

Anchors

(Not Regroutable)

(1) Driven Sands and gravels Crawl&-mounted 4”-8” Generally High early spacers may Generally in- Weak grout or

preferred but can percussivadriv- (10cm - lengths strengthczemnt be used if ties stalled at sand used to backbe installed in all ing equipment. 20cm) about 70’ grout. Grout are not 150 - 60° fill. In some

soils except Casing driven Shaft (2Omb. has high ce- attached to de- angle. cases. holes lift

those with ob- and then extract- merit to water tachable points open. Ties

structions. ed. ratio. High with threaded typically sheathed

prl?**tl*e rods more and greased.

grouting (>150 common.

psi) (1035&N/

m2)

:2) Drilled Sands and gravela Crawler-mounted 3”- 8” Generally High early Spacers may Generally Weak grout or

LSI Anchors Generally used drilling equip- (7.5cm- lengths strength ce- be required installed at sand recommended

in soils with ob- n-lent. Drill bit 20cm) less Ix-lent grout if fletible tie angle of 150 - or hole left open.

structions or precedes casing ahoft, if than 70’ with high ce- is used or no 60“. Ties sheathed or

where driving or inside crsine, soils are (20m). ment to water plate or point greared or both.

casing is diffi- perlYlC!*- ratio. Grout- is med.

cult. ble, bell ing pressure

may form. generally

>15Opsi

(103%iWm~)

legrouta- All soil types. Same equipment 4”- 8” As before Cement grout spacers not As before As before.

ale Usually used in as before for (10cm - for small (1) 1st grout generally ‘for small

inchors softer soils, drilling or driv- 2Ocm) diameter at 100 psi needed for diameter

variable con- ing casing Shaft anchors. (690 kN/m2) bars although anchors.

ditions, or (depends on soil (2) 2nd grout good forfledwhere obstruc- conditions). through indi- ble ties.

tions *re Grout pipe for vidual packers Points beencountered. each anchor. at pressures coming

up to 800 psi co-on.

(5520kN/m2)

(3) successive grouts

as needed.

Lock

Lnchors

1) Multi- Used in softer

lnder- competent rock.

‘earned

See Section on Underreamed

Soil Anchors.
Table 14. Typical equipment for construction of tiebacks, (Continued).

Range in SpaCerS

Anchor

Angle of

Preferred Soil Diameter Typical and

TYPO TYPO

Inclination

Equipment (typical) Lengtha Grout Type Plate (to horizontal) Bond Free Zone

(2) Drilled All competent As above for soil 3”- 8” Generally Cement grout Bolts or

Anchors

Rock anchors Weak cement or

rock types. anchors. Rotary (7.5cm- 4 30’ (9m) at high pres- washers in

drilling equipgenerally at

20cm)

aand backfill in

into rock. crure 150psi bottom with 450 angle.. roila above.

n-tent for rock shaft (1035kN/m2) spacers. Ties sheathed

drilling. Per- depend- or quick and greased.

cussive drills ing on setting resin.

also. rock and

load.

Gravel Competent cohe- Both augering 4”- 8” As before Cement grout

Packed

Casing serves As before As before.

aive soils that and driving (10cm - .at high presAnchors

as spacers.

will remain equipment is 2Ocm) sure >150 psi Points usuopen when not required. Driv- Shaft (1035kN/m2) ally used.

supported. ing equipment

for casing inserted after

gravel in hole.




1. Load - Theoretical bases for establishing design

load are given in Section 13.24. These are crude at best and should only

be used for preliminary estimate of safe load. Field testing of anchors is

the only method of assuring that the design anchor load can be carried by

the anchor.

2. Quality and Safety - Prooftesting of each production tie must meet general acce.ptance criteria to assure safety and to

develop uniformity of the anchors.

3. Creep - Creep rates, inferred from long term

tests, provide additional data for design and acceptance.

13.42 Current Practice

The specific criteria used by engineers/contractors

to field test anchors varies. However, it is generally recommended that

pullout tests be performed on anchors to determine the design characteristics and anchor capacities. These may involve actual failures or loadings

to some specified load greater than the design load (e.g. 200% of design).

During anchor installation several special anchors may be tested to a load

significantly in excess of design (e.g. 150 - 200%) to ensure that an adequate factor of safety against pullout is being maintained. Production

anchors should be tk’sted to a load in excess of the design load (e. g.

120 - 150%) to prove the adequacy of every anchor. It is also recommended that the rate of anchor creep be observed during all anchor

testing.

13:43 Recommendations

The following recommendations are made for installation

of temporary anchors to support excavations in the presence of nearby

structures.

13.43.1 Recommendations for Special Load Tests

Test Loads

Soil and Site Conditions Load Remarks

1. R.easonable experience with soil 150% 5% of production ties

and anchor. Nearby structures of should be tested in this

outside “zone of influence”. design manner.

,

-232-
2. Reasonable experience with soil 150% 5% of production ties

and anchor. Nearby structures of should be tested in this

within the “zone of influence”. de sign manner. In addition, 3

ties in each soil formation should be tested to

200% of design.’

3. Little experience or unsatis -

factory experience with soils and/

or anchor. Nearby structures

within “zone of influence”.

150% 10% of production anchors

of tested in this manner.

design In addition, 3 ties in each

soil formation should be

tested to failure or 250%

of de sign. 2

1 . For ties loaded to 200% of design, the ties should be loaded

to 150% of design and tested as other special test anchors.

If the anchor passes the special test criteria, the anchor

should then be loaded to 200% of design. If the anchors

satisfy the creep criteria for special test anchors at this

load, they may then be used as production anchors.

However, it is recommended that these anchors be tested

prior to actual construction to verify anchor design criteria

(length, diameter, grouting pressure).

2. These anchors should be loaded to 150% of design and tested

as special test anchors prior to increasing the load. If the

anchor passes the special test criteria, the anchor should

then be loaded to 250% of design or failure. The anchor

design should be modified if failure occurs at less than

2OO%.of design. It is recommended that these anchors be

installed and tested prior to actual construction. Anchors

tested prior to construction should be of varying lengths

and geometries to establish the appropriateness of the design

a s sumption s .

Duration of Special Test Load and Criteria for Creep

Cohesionless Soil Load duration of 1 - 2 hours depending upon prior

experience with soil and anchor. The creep rate

at a load of 150% of design should not exceed 2mm

(0.08”) per logarithmic cycle of time (see Figure

108).

Cohesive Soil Load duration of 24 hours for all cohesive soils.

Creep rate should not exceed 2mm (0.08”) per

logarithmic cycle of time (see Figure 108).

-233 -
Acceptable Creep Rate,

-------- km = 0.08” / log cycle -SC------x- * . -* Observed and Predicted

0

0.05

0.10

0.15

0.2a

0.25

Unacceptable Creep Rate,

kc > 0.08” 1 log cycle

Creep Rates for Production

Tieback Tested for 20 min.

(km = 0.06” /log cycle)

I \ .

.

I

\

\

\

\

I \

\

\

\

I

\

\

1 \

\

I

\

I \

I”= 25.4 mm I

\

\

\

I \ \

I \ \

1i-I *”3n 1no 1 nnn 1 nnt I” --- ---- ----

Time, minutes (log scale)

Figure 108. Example of recommended method of recording creep data.
Method of Load Application

1. Load anchor in increments of 25% of design load to

12 5% of the design load.

2. Unload to zero

3. Reload in increments of 25% of design load to the

desired load (or loads).

4. Maintain load for

5. Unload anchor to

prescribed period.

specified lock-off load.

13.43.2 Recommendations for Production Anchor Load Tests

The following recommendations are the minimum test

test criteria for anchors.

Test Load

Load the anchor to 125 percent of the design load. Care

must be taken to ensure that the recommended stresses in the ties are

not exceeded (see Section 13.25).

Duration of Test Load

The load should be maintained for a minimum of 20 minutes

or until a creep rate of less than 2 mm (0.08”) per logarithmic cycle of

time is achieved. This criterion for creep is applicable for both stiff

clays and granular soils.

Method of Load Application

1. Load to 125% of the design .load in increments of

25% of the design load.

2. Unload to zero.

3. Reload in increments of 25% of the design load to

125% of the design load.

-235-
13.43. 3 Evaluation of Anchor Test Loading

Anchor Capacity

The verification of anchor capacity is initially

obtained when the applied load reaches the appropriate test level

(125% - 150% of design load). However, this simple test is not

enough to ensure that the anchor capacity is sufficient. Any proof

loading of ties should include a plot of load versus tie enlogation.

Figure 109 shows a typical load vs. elongation plot

for a tie. A comparison of the observed elongation curve can be made

with theoretical elongation curves for several cases of “effective

length” in the grouted zone. The “effective length” can vary from zero

( leff= 0) to the length of the tie in the grouted zone ( 1,ff = lg). Zero

effective length indicates an anchor in which the entire tie load is

transferred at the end of the anchor nearest the bond free zone. In

this case the elongation wouldbe equal to the elongation of the tie in

the bond free zone (lb). The other limiting condition is where the

entire anchor load is transferred at the base of the anchor zone. Figure

110 schematically illustrates the load distribution in the tie for several

cases.

A comparison of this type provides some insight

into the manner of Load distribution in the anchor and in the soil. Since

the data can be recorded and plotted directly, it is a convenient method

for use in the field and during evaluation.

The ties will generally be acceptable if the

measured elongation is less than the maximum theoretical elongation

at 125% of design load (i. e. leff 5 1 g).

Creep Considerations

To assess the creep characteristics of an anchor,

a plot of anchor movement versus time should be prepared. Anchor

movement should be plotted to an arithmetic scale while the time

readings should be plotted to a logarithmic scale. Figure 108 illustrates

a plot and defines the creep coefficient, kc, which must be less than 2 mm

(0.08”) per logarithmic cycle of time.

If there is concern that even these creep movements may affect the performance of the tieback system, the testing

criteria may be more rigid. For instance, the maximum allowable

creep coefficient, kc, may be reduced or the test load may be increased.

-236 -
100

25

Theoretical Elongation

-mm Curve with l,ff=O

Theoretical Elongation

-- ---- Cume with leff=O. 5 lg

Theoretical Elongation

-a- Curve with l,rr=l. 0 1,

I

-

CXL 6

Observed Elongation for

Typical Tie During Testing

Elongation, inches

Figure 109. Typical plot of load vs. elongation during test loading.
rI

I

I E

t-

-.- /

I

I

I

I

I

I

I

-

a1.t UT a3.Io.g uo!sua&L = ;c ‘pea?

-238-
13.43.4 Recommended Lockoff Loads

Earth Pressure

Lock-off

$6 of design

For triangular active

distributions

For triangular at-rest

distributions

For trapezoidal and

rectangular distributions

50 - 80‘70

100%

Upper ties - 80%

Lower ties - 100%

13.43.5 Permanent Anchors

At least three full-scale pullout tests should be conducted for each soil type in which anchors are to be installed. Evaluation

of the creep rate at each stage of loading above the design load should be

made. This information can be used to determine, more accurately, what

the most appropriate value for use as the creep coefficient should be.

A conservative testing criterion for anchor failure

under creep would be to maintain a creep coefficient, kc, less than 1 mm

(0.04”) per logarithmic cycle of time at a test load of 150 percent of the

de sign load. As a matter of routine all permanent anchors should be

tested to a minimum of 150 percent of the design load as opposed to the

125 percent of design load recommended for testing of temporary production anchors.

It is also recommended that selected anchors (5

percent) from a permanent anchor installation be retested at later periods

after installation. The loads in these anchors should be checked to determine if the anchor’load is being maintained and if there is a possible

dangerous buildup of load in the anchors.

-239-
CHAPTER 14 - UNDERPINNING

14.10 INTRODUCTION

Underpinning is the insertion of a new foundation or support

below an existing foundation and the transfer of load from the old to

the new foundation. In some cases underpinning elements may be

installed on either side of the foundation, but in these cases cross

beams or some other method is used to support the old foundation

element.

A part of the underpinning process is to evaluate the existing

structure for total footing loads, existing bearing pressures, soil

conditions, ground water level, and possible excess foundation capacity. This investigation will determine the extent of the underpinning

operation, the suitable underpinning techniques, and the constraints

required to maintain structural integrity,

14.20 DESIGN AND THEORETICAL CONSIDERATIONS

14.21 Load Computation

14.21.1 Existing Structure

The load of the existing structure can be determined from building drawings. Failure to locate the plans for the

building (as is often the case in older structures) necessitates an

analysis of the structure to estimate the existing foundation loads.

14.21.2 Load Distribution

As the load is progressively transferred to the

new foundation, the distribution of the foundation load ‘changes. The

existing foundation should be analyzed for each of the intermediate

stages since the foundation could fail or settle excessively if allowable

loads are exceeded.

14.22 Deformations

14.22.1 Displacements Resulting from Adjacent

Construction

Even though a structure is successfully underpinned, it still may suffer damage from the adjacent excavation,

Lateral displacement leads to cracking when one portion of the structure

-240-
shifts relative to another portion of the structure. Tiebacks or braces

may be used to proide the resistance needed to withstand horizontal

forces. Vertical displacement below the bearing level obviously

cusses settlement. Vertical displacements above the bearing level

contribute to additional load on underpinning elements. This may

also cause settlement,

14.22.2 Settlement from the Underpinning Installation

General sources of settlement are noted below:

a. Structural Elements. Increased loads

in structural elements may cause elastic deformations. Non- elastic

deformations may be caused by creep and shrinkage of the concrete

used for underpinning,

b, Bearing Stratum, Settlements are caused

by strain in the bearing stratum.

c. Construction Procedures. The two main

sources of settlement during construction are loss of ground and the

strains associated with load transfer (slack in plates and wedges, jacks,

etc. ).

14.23 Design of Underpinning Elements

14.23.1 General

While the actual design of the underpinning

elements is relatively straighfforward, the choice of an underpinning

system and selection of a bearing stratum are more complex. Experience in working with the various types of underpinning systems is

absolutely e s sential.

14.23.2 Downdrag and Horizontal Forces

Underpinning elements are influenced by displacements occuring in the soil mass within the zone of influence of

adjacent excavations or tunneling. Underpinning elements may settle,

may shift laterally, and/or may receive additional load.

-241-
14.23.3 Group Action

Because of interaction between piles, a pile

group stresses soil to a greater depth than does a single pile. Thus,

for a given load per pile, the settlement of the group of piles will

be larger than the settlement of a single pile.

The significance of group action depends upon

a number of variables --the proximity of piles, the characteristics of

bearing stratum, and the sequence of preloading, Normally, group

action will not be important for piles spaced greater than 3 diameters

apart or piles bearing on very competent granular soils or rock.

14.30 CLASSICAL UNDERPINNING PROCEDURES

14.31 General Considerations

The objectives of underpinning are to transfer the foundation

load to a firm bearing stratum with a minimum of movement. The

underpinning operation must be coordinated with the overall construction

project, especially when the underpinning system is incorporated into

the lateral support system or the final new construction.

14.32 Pit or Pier Underpinning

14.32.1 General

Probably the most common method of underpinning is a concrete filled pit or pier which has been excavated using

horizontal wood sheeting to retain the earth. The construction procedures for this method have not changed significantly since the

technique was first used. The techniques used for access below the

foundation form the basis for other underpinning procedures.

14.32.2 Procedure

The basic procedure for installing a concrete

underpinning pier is to excavate an approach pit below the footing,

advance the underpinning pit to the bearing stratum, and fill pit with

concrete (see Figure 111).

14.32.3 Other Considerations

Load Tranfer

The space between the top of the pier and the

-242 -
OPEN PIT w

PLAN

(0)

ELEVATION

P AN

1cl

APPRWH PIT

ELEVATION

kJ)

Figure 11 1. Pit or pier underpinning.

ELEVATION

2x4 CLEA

PLAN

(b)

SHEETED PIT

UNDER FOOTING

ELEVATION

(d)

DRYPACK Cjff

AND WEDGES

CONCRETE

DERPINNING

PL ATES

-243 -
foundation is normally filled with dry-pack- -a mixture of cement and

moist sand that is rammed in place. Plates and wedges may also be

used.

Under certain circumstances the settlement

associated with load transfer may not be acceptable. In such cases

jacks may be inserted between the top of the concrete piers and the

underside of the footing and loads maintained on the jacks. This

permits the concrete pits to settle while maintaining the structure at

its original elevation.

Horizontal Wood Sheeting

The thickness of the sheeting (commonly 2

inches for wood sheeting) is essentially independent of depth as the

stresses in the soil are distributed by arching. The material used

for sheeting is commonly untreated wood. Occasionally, because of

concern over future deterioration, specifications require treated wood,

concrete planking, or steel sections.

Belled Piers

If required, underpinning pits can be enlarged

or belled at the bottom. There is a possibility for loss of ground if this

operation is not performed carefully, especially in non-cohesive

soils.

14.32.4 Sources of Settlement

Some causes of settlement are improper

backpacking of sheeting, excessive deflection of the sheeting, and loss

of ground. Pit excavations in weak soils or below the ground water

level may cause significant movement of the adjacent soil. In general,

fast excavation and concreting minimize movements of adjacent footings

or slabs.

Weak soils, such as saturated silt or soft clay,

tend to flow or squeeze into the pit excavations. Ground loss may

occur during exposure of the soil face before lagging placement or after

lagging placement by movement through open lagging or by movement

intoan overcut zone behind the lagging.

Pit underpinning is generally limited to use in

dry ground. If other underpinning techniques cannot be used, vertical

wood or steel sheeting maybe required to maintain the sides of the

pit. Settlement maybe caused by loss of ground behind sheeting, erosion of
soil through lagging, or an unbalanced hydrostatic head causing a “blow”

at the bottom of the pit.

14.32.5 Examples

unde rpinning .

Figures 112 and 113 illustrate examples of pit

14.3 3 Pile Underpinning

14.33.1 General

Generally, H-beams or steel pipe piles (both

open-and close-ended) are used in underpinning. Open-ended pipe is

usually preferred to c.lose’-ended pipe. Open-ended pipe permits- cleaning

out soil to reduce end resistance and side friction. Close-ended pipe

is used to penetrate through soft soils and/or where displacements

and vibrations from pile driving do not have a significant effect.

Piles can be installed either directly under or

alongside a footing. If the piles are alongside the footing, the load

can be transferred either to a beam connecting two piles or to a

bracket on a single pile.

14.33.2 Pile Installation

Jacked Piles

Typically, aluminum hydraulic jacks are used

because they are light and easy to handle in a confined pit, Since the

foundation is used as a reaction, the jacking load must be monitored

to prevent an excessive upward force.

Except in soft material, jacking is done with

open-ended pipe to permit removal of soil from within. in soft soils,

a plug of cinders, sand, or lean concrete is formed within the pipe

to permit advancement of the pipe by displacement.

The typical proc.edure is first to excavate an

approach pit and then to jack the pile downward in approximately 5

foot long sections. When the required depth is reached, the pile is

filled with concrete and test loaded to 150 percent of design load,

Load is transferred by inserting a wedging

beam (e. g. I-section), plates, and wedges while maintaining the full

-245 -


load on the jacks. As a final step, the wedging beam and plates are

encased in concrete (see Figure 114).

Driven Pile s

Conventional hammers or drop weights can be

used to drive piles. When using a conventional hammer, the energy

that can be developed by the hammer is often limited by the size of the

pit that can be excavated beneath the footing.

Piles are driven in sections with splices made

between successive lengths. Open-ended pipe may be cleaned out,

if required, to reduce resistance.

Where installed below foundations, driven piles

may be test loaded by jacking aginst the foundation, Load transfer is

done in a fashion similar to that used for jacked piles.

Advancing Open-Ended Pipe

Side friction or end resistance is reduced

during installation by periodically cleaning out the soil from within

open-ended pipe. Sections of pipe are connected by tight fitting sleeves,

generally fastened on the outside of the pipe to avoid interference during

cleaning out. The sleeves are not normally welded and are designed

to keep the sections of pile in alignment.

Piles can be cleaned using various tools such

as pancake augers, flight augers, orange-peel buckets, water jets,

air jets, or water/air jets. When using any of the jet cleaning methods,

care should be taken not to clean below the bottom of the pipe as this

may cause loss of ground and ultimately lead to settlement in the surrounding soils. While cleaning the piles and during driving, a positive

hydrostatic pressure must be maintained to prevent a “blow” at the

bottom.

14.33.3 Piles on Both Sides of Footing - Support

with Beams

Main steps are:

The basic procedure is shown in Figure 115.

a. Excavate to near bottom of footings and

‘install piles.

b. Sequentially install a series of beams below

the footing by excavating a sheeted trench.

-248-
I

STRUT WEDC

PREVENT RE

AFTER PRES

6’ J

BE PILE USUAL

PEN-ENDED.

Figure 114. Jacked pile installation.

-249-
COLUMN l-w I IWU

--------

I, ,

I

I I.‘ I

WALL DR COLUMN

0

FOOTING

DRYPACK Of? PLATES

AND WEDGES

PILE DRIVEN TO BEARING

Figure 115. Piles driven alongside footing, support by beam,

-250-
Transfer load with drypack, plates and

wedges, or jacks. This transfer of load

can be made at either the bottom of the

footing, the top of the pile, or a combination

of the two.

14.33.4 Piles on One Side of Footing - Bracket Pile

Underpinning

This method is normallyused for light structures,

especially for exterior walls or continuous footings. Bearing is developed

by a driven pile, usually an H-pile, or by a belled or straight-shaft

caisson.

Driven Piles

The piles are usually installed alongside the

footing and the load is carried by brackets welded to the pile, Plates

and wedges or drypack are used to transfer load.

Pre-excavated Vertical Piles and Caissons

A steel beam or a concrete shaft may be used.

a. Steel Beam: The hole is filled with lean

concrete and a bracket is welded on the

steel beam similar to driven bracket pile

underpinning (see Figure 116). An alternative procedure without brackets is shown in

Figure 117.

b. Concrete Shaft: A hole is excavated under

the footing, the necessary reinforcing steel

is placed, and the pile and bracket are

poured monolithically (see Figure 1181.

Pre-excavated Battered Piles

This method, shown on Figure 119, consists

of drilling a hole at a batter or a l’slantt’l starting adjacent to the existing

footing or as close as feasible to the footing and continuing the hole

to the bearing stratum. A vertical slot below the footing intersects

the slant pile, and reinforcing ties the slot and pile together.

-251-
.-

WEDGES

STEEL BEAM /

LAUGER HOLE

FILLED WITH

LEAN CONCRETE.

BRACKET

ELEVATION

NOTE: SIMILAR DETAILS IF

STEEL PILE IS DRIVEN

IN PLACE.

Figute 116. Steel pile with steel bracket.

-252-
ORYPAC

O R

PLATES

AND

WEDGES

STEEL

COLUMN

SLOT - I

I

s

! AUGER HOLE

I t

Ll- --I

fk

Figure 117. Auger hole with pile installed in slot.

-253 -
- -

t--------- WALL

DRYPACK

O R

PLATES

AND

WEDGES

REIN. STEEL

T

CONCRETE--J

BRACKET

CAISSON REIN- >

FORCING STEEL

(IF REQ’D)

d FOOTING .

1

-

BRACKET REINFORCING STEEL

:I: A\ I+ CAISS(

CAISSON REINFORCING STEEL

m

(IF REQ’D)

I I I

I PLAN

I

I -

\AUGERED

CAISSON

IN

BELL IF REQUIRED

Figure 118. Augered concrete caisson with concrete bucket.

-254 -
DRYPACK OR

PLATES 8

UNDERPlNNlNG

ELEMENT

I AU&R HOLE

I PROPERTY LINE MUST

I

INTtRSECT BELOW

SUBGRADE. L

Figure 119. Battered pile underpinning.

-255-
14.40 GROUTED PILES

14.41 Hollow Stem Auger

A continuous flight, hollow shaft auger is rotated into

the ground to the specified pile dc pth. As the auger is withdrawn,

high strength mortar is placed under pressure throug’l its center to

form a pile of regular length and diameter. A reinforcLg cage is

placed into the wet grout. Typical sizes range from 12 inch to 16

inch diameter.

Special low headroom equipment permits installation of

these piles inside buildings. These piles can be installed adjacent to

or through existing footings, and loads can be transferred from the

structure to the piles by beams or brackets or by making the piles

integral with the footing through bond.

14.42 Root Piles (Pali Radicel

14.42.1 General

This system is capable of providing vertical

and/or lateral support to foundations and excavations (Bares, 1974)

(see Figure 120). The piles range from 3-l/2 inches to 12 inches in

diameter and are usually reinforced.

14.42.2 Root Pile Underpinning

Installation

When used for underpinning, root piles are

normally installed through existing foundations. The drilling muck or

cuttings are brought up to the surface by direct circulation of the drilling

fluid (bentonite slurry or water). Installation in granular soils usually

requires a casing throughout its entire length to prevent collaspe of

the hole.

Concreting of the pile is accomplished by filling

from the bottom with mortar placed through a pipe. Compaction of the

mortar is achieved by blasts of compressed air (about 70 to 100 psi)

done in stages as the casing is withdrawn. This improves the contact

of mortar and soil and facilitates the withdrawal of casing.

Reinforcing consists of a cage or a single bar.

The smaller root piles (generally 4 to 5 inches nominal diameter) are

reinforced by a deformed high strength bar while the larger piles

-256 -
-257-
(generally 6 to 12 inches nominal diameter) are usually reinforced with

a spiral cage. The steel is placed after concreting in the smaller

pile and before concreting in the larger piles.

Design Considerations

The design of root piles should follow procedures

for friction piles and end bearing piles modified by experience. The

load carrying capacity is in the range of 10 to 15 tons for the smallest

diameter piles and 40 tons or more for the larger diameter piles.

Load is transferred to the soil through friction, end bearing, or a

combination of the two, depending upon soil conditions,

Table 15 summarizes the results of load tests

on root piles obtained from published and unpublished data. III general,

the tests were not carried to failure and therefore, the data do not

permit an evaluation of safety factors. However, since settlement data

were available it was possible to develop, at least in crude fashion,

a relationship between pile geometry, load, and settlement.

14.423 Reticulated Root Piles

The term “reticulated” is used by Fondedile to

describe an application where the piles resist lateral displacement of

the soil, as differentiated from the underpinning application where the

piles support vertical load. Jn these cases the underpinning piles

carry vertical and lateral loads and resist soil displacement (Bares,

1974, 1975, and Lizzi, 1970).

The principle is to engage an earth mass by

installing a root pile network at close spacing and in a particular pattern of pile batter and orientation. See Figure 121 for an example.

14.50 TUNNELING BELOW STRUCTURES

14.51 General

This discussion concerns instances when tunnels pass

below structures. In such instances, it is likely that vertical underpinning elements cannot be used directly below the foundations.

14.52 Column Jacking

Figure 122 illustrates a common technique used to support

columns during below grade construction. The column is first isolated

-258-
Table 15. Reslrlts of load tests in Pali Radice.

Assumed Settlement Settlement

Nominal Effective Max. Test at

Case Diameter Length Max. Load

Modulus 2 Soil

Length Load (1) in-ft TYPO

NO. D, inches L, feet L’, feet P, tons p, inches k ‘ton (2) Location

1” 4 21 21 22 0.04 0.013 G School Building, Milan, Italy

i” 4 40 40 22 0.16 0.097 C Olympic Swimming Pool, Rome

3 ’ 12 90 90 50.6 0.32 0.570 G Bausan Pier, Naples

4** 4 49 20 19.8 0.08 0.0270 Si, G Italian State Rome Railrod,

5* 4 52 42 17.6 0.09 0.072 G Bank of Naples

6* 8.5 99 66 108 0.22 0.087 G Corps Engineers, Naples of

7** 5 65 24 50 0.32 0.062 G Washington, D.C., Subway

8* 9 19.5 10 45 0.45 0.075 G Queen Anne’s London Gate,

9 ’ 7 28 18 50 0.30 0.063 G Queen Anne’s Gate, London

10** 4 52.8 52.8 23. 1 0.236 0.1798 C-G Salerno-Mercatello Hospital,

Salerno-Mercatello

11** 8 82.5 43 108 0.472 0. 125 G Marinella Port of Wharf, Naples, Naples

12** 8 47.5 47.5 59.4 0.035 0.0187 G Main Genoa Switching Plant,

13** 8 73 73 62.5 0.065 0.0506 G Mobil Gil Italiana, Naples

(1)k =+D =+ (2) G = Granular; C = Clay; Si = Silt

*

Bares, F.A.. Personal Communication, September 1975.

* *

Bares, 6 A. (1974).
Table 15. Results of load tests in Pali Radice. (Continued).

Assumed Settlement ~t~ement

Nominal Effective Max. Test at

case Diameter Length Length Load Max. Load Soil

NO. D, inches L. feet L’. feet P. tons p, inches Typ”(2, Location

14** a 66 66 58.7 0.037

15** a 63 63 56.5 0.065

16** 8 60.5 60.5 56.5 0.028

17** a 73.5 73.5 27.5 0.252

0.0247

0.0483

0.0200

0.4490

la** 8 66 66 24.2 0.386 0.7010 C

19** a 66 66 48.5 0. 205 Q 1860 C

20** 8 99 110.2 0.213 0.0850 G Belt (Expressway) East-West, Naples

21** a 99 88.2 0. 127 0.0634 G Belt (Expressway) East-West, Naples

22** a 59.5 68.3 0.061 0.0354 G Swimming Pool - Scandone Pool, Naples

23** 4 33 21.5 0; 087 0.0445 G

24+* 8.5 82.5

66

66

59.5

33

82.5

82.5

69.7 0.148 0.1241

25** 8.5 69.7 0.150 0. 1258

G

G

Railway Terminal, Naples (Corso A. Lucci)

Plant (Brindisi)

Plant (Brindisi)

Special Foundations for Transmission

(Electrical Towers between GariglianoLatina)

Special Foundations for Transmission

(Electrical Towers between GariglianoLatina)

Special Foundations for Transmission

(Electrical Towers between Garigliano-.

Latina)

Casa Albergo in Viace Piave

Port of Naples

Port of haples

CU G = Granular; C= Clay; Si = SiH

c

Bares. F.A. (1974)
EXISTIN@

bUlLDING

II i r’ .I,

DILP ’

QLES

NRE

‘I

1

ATE

NO

Figure 12 1. Schematic showing principle of

reticulated root piles.

(Courtesy of Warren-Fondedile, Inc. ).

-261-
FOOTING Q CWJMN

CM MOVE MMPENOCNTLY

FooTammuwEL FnssEs

UUW)

(0)

BRACKETS INSTALLED, ANCHDR BOLTS

LOOSENED, JACKS INSTAlLED.

(b)

SHIMS INSTALLED, ANCHOR BOLTS

TIGHTENED.

Figure 122. Schematic of column jacking to prevent

structure settlement during tunnel construction.

-262-
from the footing and is maintained in place by jacks, The footing can

settle while the column remains in place. After construction the column

is reconnected to the footing.

14.53 Pipe Shield Technique

The procedure is to install a series of contiguous horizontal pipe tunnels, on the order of 3 to 4 feet in diameter, which

are later reinforced and concreted to provide a protective roof (or shield)

above the tunnel. Typically, the contiguous tunnels, called pipe shields,

are installed by jacking pipe from an open cut jacking pit or from the

side of a primary drift tunnel if this is not possible.

Figures 123a and 123b illustrate examples where jacking

pits were excavated from the surface.

14.54 Inclined Secant Piles

Inclined secant piles in lieu of underpinning are applicable

where there is a slight encroachment below utilities or structures

(see Figure 124).

14.55 Bridging

Figure 125 schematically isllustrates measures that can

be taken to bridge across the tunnel area.

14.60 LOAD TRANSFER

The transferring of the load from the old foundation or temporary

shoring to the new underpinning element is similar for all underpinning

methods. Sources of potential settlement are compression of the under -

pinning member, displacement of the bearing stratum, and compression.

of plates and wedges or dry pack.

14.61 Dry Pack Alone

The use of dry pack alone is generally limited to pit

underpinning. Preloading techniques may not be required because stresses

and deformations are relatively small. The dry pack is a dry mortar

mix, generally consisting of one part cement, one part sand, and sufficient

water to hold the mixture together. It is placed in the void between the

underpinning element and the existing footing by ramming with a 2” x

4” and maul.

-263-
(01 CONSTRUCTION

-

.,* . . - . - ~ :,... * -. . . . - -. * ‘. r

. . . . . .

,-

..,._. - *..-

. . - -. .

(b) AFTER COMPLETION

Sequence

1 .

2 .

3 .

4.

Underpin bridge with steel piles and jacks to rdjurt for settlement.

Construct jacking pits on each side of highway, jack 1. 2 m plpee

and cone rete pipes.

Construct 3 m wide x 2 m hig’h tunnels below pipe& Concrete em h

tunnel before building next one.

Construct walls of highway tunnbl.

Figure 123a. Pipe shield technique (after Zimmerman, 1969).

-264-
0) PLAN \

\

\

\ ~RRYING BEAM

\

v

TUNNEL -j

b) SECTION

4 WILDJNQ 4

RUBBISH

FILL

Figure 123b. Pipe shield technique (after Rappert, 1970).

-265-
ORAVEL

SAND

s&

r” STIFF FISSURED ua+s CLAY WITH SAND

* 8 SILT LENSES

“c:

Figure 124. Example of bored pile wall used to

pro tee t structure (after Braun, 1974).

-266 -
*. . *. .

’. % .

PILk *

. -.

.

1

.

. -

.

- .

. ’

.CO*UETEr)

. . r .

.’ . .

* * * * . .

a. .

.

;.01

TUNNEL ,:

. . *

. . .

. :

. . * ‘.

.

. . - - .

.

. . *

. . .

* . * .

. . .

. . - .

“UNO6RF+tNNIN@ LImEWENT

* :

c>

.

. * c

* .

Figure 125. Bridging.

-267-
14.62 Plates and Wedges

This method consists of using pairs of steel (or wooden)

wedges driven between steel plates in the void between the underpinning

element and the footing. As the wedges are driven, their combined

width increases. The footing then acts as a reaction, and the road

in the underpinning element increases. For a permanent installation,

dry pack may be used to fill voids. If the wedges are steel., they can

be we.lded together to prevent future deformation.

14.63 Jacking

Jacking is done with mechanical jacks, hydraulic ram jacks,

or with hydraulic flat jacks where the space is too restricted to accommodate conventional jacks. Where creep is minirna.1, the load can be

transferred immediately by a steel or conc:rete plug or dry packed. The

jacks are then removed. Where there is concern over settlement, the

Load can be maintained and periodically adjusted as needed.

14.70 TEMPORARY SUPPORT OR “SHORING”

14.71 Basic Considerations

The need for temporary support during underpinning is

controlled by the integrity of the structure being underpinned, the

effect of a temporary bearing pressure increase adjacent to the underpinning operation, and the degree of foundation undermining.

Usually it is very difficult and often impossible to

predict the loads which the shores will carry. Movements of the

shored element should be monitored throughout construction. The

shoring can be jacked or wedged to compensate for settlement, if and

when it occurs.

14.72 Needle Beams

Typical “needle beam” configurations are shown in Figure

126. ;More elaborate systems might consist of concrete pads and steel

needles with jacks at the support points to control the movement of the

structure.

-268-
r-k- WALL

DRYPACK & PLATES

OR WEDGES

CUT HOLE IN WALL

~-BLOCKING AND FOOTING

(a) NEEDLE BEAM THROUGH WALL

(SIDE VIEW)

-BLOCKING AND FOOTING

-COLUMN

7

F-FOOTING

I I

T I A

I--->

I

I I

1

1 rk 1

FOOTING

NEEDLE BEAM

(b) NEEDLE BEAMS SUPPORTING A COLUMN

(PLAN VIEW)

Figure 126. Needle beam detail.

-269-
14.73 Inclined Shoring

Some typical configurations are presented in Figure 127.

In a.ll cases, the lateral loads transmitted through the shores must be

accounted for. Some common details of shoring connections are shown

in Figure 128 and 129.

When cast iron columns are encountered, special

attention must be given to prevent damage to the column. Often it is

necessary to fill or encase the cast iron column with concrete. The

pin and clamp method is presented in Figure 130. The shoring of cast

iron columns might also be accomplished using a concrete collar

placed over either a roughened surface or using welded shear connections on the column. Regardless of the method, eccentric loadings

should be avoided.

Figure 131 illustrates a case where inclined shoring was

used to protect a structure.

14.80 PERFORMANCE

Underpinning is no guarantee that the structure will be totally

free from either settlement or lateral movement. About l/4 - l/2 inch

of settlement should be expected during the underpinning process- -even

under the best of conditions. Additional movements may be associated

with the subsequent adjacent excavation, including lateral displacements

occurring in the retained soil mass adjacent to the excavation. Ware

(1974) presents settlement and lateral movement data for underpinned

structures in the Washington, D. C. area.

-270-
l

. .

. t ë.. . “;i:,í

i.;

APPROAti PIT

CUTNOTCH IN FOOTING

DRYPACK, PLATES AND WEDGES

TEMPORARY SHORING

TEMPORAHY JACK IF

REOUIRED

TEMPORARY FOOTING (CONCRETE OR

FLOOR

FLOOR

!

FOUNDATlON

*

WALL I

-WOOD)

(01 INCLINED SHORING UNDER A FOOTING

~WHORE

” y APPROACH AND \PLATES AND if EDGES AS

UNDERPINNING PITS

(b) SHORING A WALL OR COLUMN

FLOOR 3

?

REQ’D

Figure 127. Inclined shoring details.

-271-
(a) WELD STEEL SHORES DIRECTLY TO

COCJJ IutN

-BRACKET -

*

(b) SHORE TO BRACKETS WELDED ON

COLUMN

Figure 128. Shoring details, steel column.

-272-
I-. CONCRETE COLUMN

ROUGHEN

EXISTING

L RElNFORClNG STEEL

V- ’ STEEL

SHORE

(a) LIGHTLY LOADED

COLUMN

(b)SHORE AGAINST

CONCRETE

COLLAR

CONCRETE

COLUMN

-STEEL [OR 1

//I’ 1 ‘\r STEELSHOREti/‘,’ ‘I’\

(cl CLAMP STEEL BEAMS OR CHANNELS TO COLUMN

AND SHORE AGAINST CLAMPED BEAMS.

Figure 129. Shoring detail, concrete colunm.

-273 -
CAST IRON COLUMN

&-- GROUT HOLE

BOLTS

CLAMP TYPES

Figure 130. Pin and clamp details for a cast iron column.

-274-

CHAPTER 15 - GROUTING

15.10 INTRODUCTION

Since 1802 when Charles Be’rigny initiated the practice of

grouting; the techniques, procedures, grouts, and applications of

grouting have increased and improved. Today, grouting is used to

stabilize soils, provide ground water cutoffs, and underpin structures.

Any one or all three of these features of grouting may be required

during cut -and- cover or soft ground tunneling in urban areas.

Injection grouting of a porous soil or rock mass may be done

with particulate or chemical grouts. The principal advantage of

injection grouting is that the engineering properities of a soil mass

can be substantially improved with little or no disturbance to existing

structures. Grouting can also be performed in locations where access

limitations inhibit the use of conventional construction techniques.

The design and construction inform.ation on grouting presented in

this report is provided to aid the engineer and/or contractor in deciding

whether grouting is a feasible construction technique for the project

being considered. Final design of a grouting system could not be performed

on the basis of the information presented herein. A detailed design

and construction manual on grout.ing is being prepared by HaUiburton

Services and will be avai.Lable through the National Technical Information

Service.

15.20 DESIGN AND THEORETICAL CONSIDERATIONS

15.21 Purpose

Grouting can be used to control ground water, to solidify

or stabilize a soil mass, or underpin an existing structure. For a given

project, grouting may be used to achieve one or all three of these

purposes.

Grouts injected into a soil mass reduce the permeability

of the deposit. A ground water cutoff, either vertical or horizontal

(Biittner, 1973), can be formed to replace or to supplement other

dewatering s themes. Selective grouting of specific strata may also be

performed. Figure 132 iLLustrates several situations in which grouting

techniques could be used to control ground %water.

Grouting can significantly improve the strength and

deformation characteristics of the soil mass. Strength grouting can

be used to prevent large deformation behind lateral support walls, to

-276-
a) PERVIOUS LAYERS WITHIN DEPTH OF CUT

(IMPERMEABLE)

---A--

------

(IMPERMEABLE)

GROUTED ZONE

b) DEEP PERVIOUS DEPOSITS

SAND 8 GRAVEL

(PERVIOUS)

DIAPHRAGM

WALL

f

f

Z

3 -

-

By.

*.,.1

’

- 1 ::

( NOTE: X 8&Z 8 w)

HARDPAN (IM

Figure 132. Grouting for ground water control.

-277-
c) BOULDERSORSHATTEREDROFK

y-[,STEEL SHEETING r

ROCK- BADLY

EETING

NETRATION

CERTAIN

I--------a----- J

Figure 132~. Grouting for ground water control.

-278-
prevent running of soils, or protect overiying structures during bored

tunnel construction. Figure 133 Uustrates several examples of soil

stabilization through grouting.

Grouting for underpinning is a special appLication of

grouting for soil so.lidification. This technique may be particularly

valuable if the grouted mass can also be used to control ground water

or act as a lateral support wa.11. Figure 134 illustrates a case

where grouting could be used to underpin a structure.

15.22 Soil ProfiLe and Soil Type

15.22. 1 FieLd Investigations

Fie.Ld investigations undertaken for a proposed

grouting scheme faLL into two phases. The first investigation phase

would involve obtaining an accurate definition of the soil profile. This

would incLude careful mapping of the depth and extent of strata.

The second phase includes obtaining more data

pertaining to the specific soil properties controlling groutability,

including field permeability tests and soil sampling or laboratory testing.

The in situ soil permeability can be determined from borehole permeability

tests or pumping tests. Pumping tests are preferred because they provide

more reliable values of permeability.

In rock, instances of water Loss during driLling

should be recorded, and rock core Logging should reflect jointing,

weathering, and RQD-- all of which bear a relationship to permeability.

15.22.2 Laboratory Investigations

Laboratory testing wiLL be limited primarily to

detailed Logging to map stratigraphy, grain size analyses, and Laboratory

permeability tests. The stratigraphy may determine the appropriate

grouting methods and procedures. Grouting uniform soil deposits may be

much different than grouting highly stratified deposits.

Grain size analyses may determine whether the

deposit can be grouted. Although soils with greater than 10 percent by

weight pas sing the No. 200 sieve can be grouted, it is generally very

expensive to do so. Coarse silt deposits can also be grouted but more

extensive analyses of the gradation and particie nature is required.

HaLLiburton (1976) describes the more rigorous grouting Limit criteria.

Laboratory determination of permeability is of

limited usefulness since the tests are generally performed on reconstituted

-279-
a) CUT AND COVER

SOLDIER PILE

\ WALL

STIFF CLAY

b) SOFT GROUND TUNNEL

. - -

. ~. . . . .: - - .

. . .

‘_ . . - * . .-- _-

* . .

-.

. .

- _.. $L\NO,’ ‘:‘ . . . . &- . I--.

I’ARCHll : . . --

. . _

STIFF CLAY

Figure 133. Grouting for soil solidification.

-280-
STRUCTURE

SHEET PILE

WALL 7

EXCAVATION FOR CUT

ANDCOVERTUNNEL

r i,. -;

_ I. ,. .

I. .’ , _ .

. .

1’ . -. 4

f

I ,I8. ,,

~

,

ROCK

, . ’ BENEATH.. _ . . FOOTING . .- .

‘. ‘a’ ., ,. *

r

‘_ .

.

25

,* .’ -

-*_ -.’ .P

.

._

Figure 134. Example of grouting used to

underpin an existing structure.

-281-
soil samples. Therefore, laboratory and field permeabilities may

differ considerably. An assessment of all parameters --grain size

distribution, stratigraphy, laboratory permeability tests--provides

a basis for judging whether a soil deposit can be successfully grouted

and what grouts are like.ly to be most efficient.

15.23 Grout T,ype

15.23. 1 Particulate Grouts

Particulate grouts are fluids that consist of a

suspension of solid particles --such as ce.ment, clay, a processed clay

Like bentonite, or a mixture of these elements. The groutability, or the

ability of a grout to penetrate, is limited by the size of the particle in

suspension and the size of the voids in thse material to be grouted.

Mitchell (1968) defines a groutability ratio for soils as the ratio of the

15% size of soil to the 85% size of the particulate grout. For successful

grouting the ratio should exceed 25.

Groutability ratio =

D15 (soil) > 25

D 85 kFOut)

In practice, normal cement based grouts are limited in use .to

coarse sands while a pure bentonitic grout might be injected into a medium

sand.

15.23.2 Chemical Grouts

Chemical grouts are frequently classified into

two major groups: silica or aluminum based solutions and polymers.

Metathetical precipitation processes (M. I. T., 1974) generally use

silicate solutions with sodium silicate being the best known although

aluminates are &so used. The basic process consists of adding acid

to a soluble silicate salt to form a silicate gel and salt. Chromelignosulfates

also fall into the general category of metathetical precipitation type

grouts.

Po.lymers are gen.era.lLy more fluid than the

metathetical precipitation grouts and use a process by which monomers

or partially polymerized polymers react to form macromolecules. The

reaction can be triggered by catalysts or by application of heat, pressure,

or radiation (M. I. T., 1974). Table 16 summarizes the basic grout

types and lists some of the common grouts according’to these general

groupings .

Unlike particulate grouts that are injected as

suspensions in a fluid, chemical grouts are injected as true so.lutions.

-282-
Table 16. Classification of common grout types

(from Massachusetts Institute of Technology, 1974).

Particulate Grouts

Chemical Grouts

P ret ip tat ion

Polymers

Cement

Clay

Bentonite

Silicate Chemicals

Aluminate Chemicals

Chromelignosulfates

AcryIamides (e. g. AM9)

Phenoplasts or Aminoplasts

(e. g. recorcineformol,

urea -fo rmol)

EPOXY

Polyester-resins

Suspensions

Injected in form

of monomers

Injected partially

polymerized

-283 -
Chemical grouts are therefore idealized to behave as Newtonian fluids

of characteristic viscosity. Viscosity of the grout, together with the

permeability of the soil and the injection pressure will control the

groutability. E. Maag in 1938 (Ischy and Glossop, 1962) developed a

simplified model of the behavior of a Newtonian fluid:

t

ocn 5

3khr o

(R3-ro3)

where:

r

0

n

k

oc

h

t

radius of grout at time (t)

radius of the injection pipe

porosity of the soil

permeability of the soil

ratio of g.rout viscosity to that of water

piezometric head in the grout pipe

time of grouting

Maag’s formula is based upon several simplifying assumptions--a uniform

homogeneous soil, spherical flow, radius of injection pipe small with

with respect to depth below water, and injection toccurring above

impermeable boundaries. In view of the many unknowns inherent in any

soil mass, however, a more precise theoretical solution to the problem

of rate of grout penetration is of questionable value. For a more precise

determination of the rate of grout penetration field injection tests are

required.

15.24 Design Factors

The final grouting design is performed by a grouting

specialist; however, the engineer and/or c:ontractor should be aware

of the features that influence grout selection and design. In some

cases the grouting procedure may involve several injections with grouts

of decreasing viscosity to achieve the desired product (most commonly

done in Europe).

Grout selection must include evaluation of required soil

strength and permeability as well as grout gel time, setting characteristics, volume of grout, and penetrat:ion. TabLes 17 and 18 and

-284 -
Table 17. Limits of grouting ability of some mixes.

Coarse Sands Medium to Silty or Clayey

Type of Soils and Gravels fine Sands Sands, Silts

aI

2 Grain diameter : dl O> 0. 5mm

Q),

Q02<d, 0<05mm dlO< 0. 02mm

[ *ecific surface s < 100 cm-l - 1 100cm-l<lCKXIan’. s > 1OOOcm

-I

2 Permeability k >10e3m/s 10m3> k> 10m5m/ s k ( 10e5m/s

series of Mix Bing ham Colloid Solutions Pure solutions

Suspensions ,(Gels) (Resins)

Double-shot

Consolidation Cement silica-gels Aminoplastic

Grouting (k > ld2m/s) (Joosten)

Phenoplastic

Aerated Mix Single -shot

silicate

mpermeability

Grouting

Aerated Mix Bentoni te Gel Acrylamide

Bentonite Gel Lignochromate Aminoplastic

Clay Gel Light Carongel Phenoplastic

Clay/Cement Soft Silicagel

Vulcanizable’Oils

Polyphenol

After Janin and Le Sciellour, 1970

-285-
Table 18. Grout types for ground stabilization.

Soil Type Particle Size Minimum Grout Type

Fissured rock to coarse 5mm Cement

sand PFA

Bentonite

Coarse sand to medium sand lmm Silicate

Medium sand to fine sand 0. lrnm Resins

Coarse silt 0. 011?run c Acrylamide

After Flatau, et al, 1973

-286 -
Figure 135 provide guidelines for grout selection.

Layout of grout injection pipes depends upon soil types,

grout viscosity, injection pressure, and gel time. Spacing will depend

upon grout penetration and the desired grouted soil properties. In

Europe, less expensive grouts (coarser grouts) are often used as a first

step in grouting to fill the largest voids and reduce the need for less

vis cous, but more expensive grouts. However, the ‘labor costs of

switching from more viscous to less viscous grouts may exceed the

savings in materials. Using less viscous grouts for all grouting may

be cheaper than using the sequential grouting system.

15.30 CONSTRUCTION CONSIDERATIONS

15.31 Materials

15.3.1.1 Particulate Grouts

Cement grouts are used primarily to increase

strength but also have the added benefit of l;ovxering permeability.

These grouts are the least expensive grout types and are often mixed

with natural clay or bentonite to prevent cement segregation in coarser

soil deposits.

Natural and processed (bentonite) clays can be

used as grouts primarily to reduce permeabkty. The properties of

natural clays must be carefully examined to: determine their suitability

for use. It is common to mix clays with cement to form the final grout.

15.31.2 Chemical Grouts

The basic divisions of chemical grouts are by

their respective chemical processes, inorganic (metathetical precipitation)

and organic (polymerization). Table 19 summarizes the basic types of

commercial grouts available and their relevant mechanical properties.

Inorganic grouts are silica or .aluminum based

grouts. A great variety of these grouts exist and range from high

strength, high viscosity grouts with little penetration to relatively low

viscosity grouts with lower strength and greater penetration.

Chemical grouts are generally combined or

activated using one of the following techniques:

a. A two-shot process in which two fluids are

injected separately into the same mass. The grout sets when the fluids

come into contact with each other. The classic Joosten process is an

example of this.

-287-
AASHO CLASSIFICATION

GRAVEL SAND

OARSE kW4 t fINE SILT

CE’MENT SOIL

I

Cl AY

I

I

SILICATE SOI ,UTiONS 1

TS)

GRAIN SIZE MILLIMETERS

GRAVEL 1 SAND

FINE kOARSd MEDIUM 1 FINE SILT I

UNIFIED’SOIL CLASSIFICATION

Figure 135. Range of usefulness of various grout

types (from Mitchell, 1968 1.

-288-
Table 19. Physical properties of chemical grouts (after Neelands and James, 1963).

Special Fields

Class Example

silica gel low Silicateconcentra- bicarbonate

tion

Viscosity Gel Time Spe cifi c Water - Consolidation

CP Range Gravity 8 topping

Min.

Fine Medium High

Soil Strength Strength

1. 5 0.1-300 1.02 X X

silica gel Silicate- 4-40 5.- 300 1.10 X

high con- formamide

centr ation

Zhrome

li gnin

TDM 2.5-4 5-120 1.10 X X

7inyl

polymer

AM-9 1.3 0.1-300 1.02 X X

tiethylol

bridge

polymer

U F 6 5-300 1.08 X

Xl-based un- Polythixon

saturated FRD

fatty acid

polymer 8

lo-80 25-360 0.99- X

1.05
b. A one-shot process where the gel strength

of a very low viscosity grout gradually gains strength with time and

eventually forms a stiff gel.

c. A one-shot process where the gel strength of.

a very low viscosity grout remains constant for a period of time (which

is controlled by the mix) and then gels almost instantaneously.

15.32 Procedures

15.32.1 Driven Lance

Probably the most widely used method for injection at shallow depths (10 - 12m) is the driven lance method (Dempsey

and Moller, 1970). The method consists of driving a lan-ce using a

pneumatic hammer and extracting the lance by jacking. The injection

can be through perforations at the end done either during driving or

withdrawal (or both in a two-shot process). Alternately, a loose point

may be used during driving; and, upon withidrawal, injection can be

made through the open end with the point remaining in place. A nonreturn valve may be installed to prevent influx of firm material when

driving. Figure 136 schematical.ly illustrates the driven lance method.

15.32.2 Sleeved Grout Tube

The sleeved grout tube or lltube-~-manchette”

method was introduced by Ischy and is the standard method of injecting

grouts in deep or intricate grouting operations (Ischy and Glossop, 1962).

The basic system consists of a tube, now ge:nerally of PVC, which is

installed in a borehole and surrounded by a clay cement, sleeve grout

to seal the *be into the ground. At short intervals (approximately 300 mm)

the tube is perforated and rubber sleeves are used to cover these

perforations. The grout is injected through a double packer arrangement

which isolates each perforated zone. Under grout pressure the rubber

sleeves are forced open, the sleeve grout ruptures, and the grout passes

into the soil.

The primary advantage of this system is that

multiple injections can be made from the same tube. This a.llows the

use of different grouts and better control of the grouted soil mass

properties. Figure 137 shows the basic ‘ltube-&-manchettell and

grouting procedures.

-290-
.,

\.

\

I

1

\

.: 1.

\ ‘1 *, :

.

‘.

I’ \ ;

\

’ ,

/ ’ * .

/ I’ /

- . -

. *

. .

* ,

.

/

.

,

* ,

/

.

’ t

:

I 1

I ’

I

/

f . .

/ .

. 9 .

r .

, ’

I

,

/

. I

7. T. . .

.

2:’ . . .

. .

. . -

. .

. *

-/-1

, .

t

/

‘I .

* 1

1

. -

.

. .

II

.:/ : . .

. /

/ *

/ .

.

1 .

.

(a)

DRIVE LANCE.

‘(b) (cl

SEPARATE POINT WITHDRAW LANCE

AND PERFORM WHILE INJECTING

FIRST INJECTION. GROUT IN ONE FOOT

INTERVALS.

Figure 136. Schematic of driven lance method.

-291-
BORING B

CASING

,OOUBLE PACKER

OF GROUT H6LE

- SEMI - PLASTIC SEALlNG

SHEATH

PIPE SEALEO INTO

‘RUBBER”MANCHE1

GROUTING ORIFICE

‘GROUTING PIPE

‘DOUBLE-PACKER

HOLE

‘TE”

(a)

iNSERTlNG THE SEALING- IN OF GNJECTION BY

TUBE b TUBE b MANCHETTES MEANS OF

MANCHETTES AND WITHDRAWL DOUBLE-PACKER

OF CASING

(b)

Figure 137. Sleeved grout tube

(after Ischy and Glossup, 1962).

-292 -
15.32.3 Injection Pressures

In general, injection pressures for normal

grouting operations are limited to 1 psi injection pressure for each

foot below ground surface. The purpose of limiting the injection

pressure is to prevent fracturing of the ground. In specific instances

where high confining pressures are known to exist (below heavy structures,

for example) the 1 psi per foot of depth limitation may be raised.

15.32.‘4 Special Techniques

Vibratory Lances

Biittner (1974) reported a cage in the Netherlands

in which a horizontal cutoff below an excavation was placed using

vibratory techniques to install the lances to the proper depths. A

detachable point with a plastic pipe attached was connected to the

vibrating lance. The point was detached and grout pumped through the

plastic pipe to form the horizontal cutoff. The primary advantage of this

method is that the grouts can be injected at depths of up to 23 m or

approximately twice the depth possible using driven lances.

Short Gel Times

Karol (1968) reports the use of AM-9, an

acrylamide grout, with a gel time Less than the pumping time. Pumping

of the grout continues after the initial grout has set; creating an everincreasing size grout bulb. The mechanism controlling this behavior

is still unknown, however, it has been found that it can be used to create

a grouted formation in the presence of flowing ground water.

15.40 FIELD TESTING AND QUALITY CONTROL

15.41 General

Since it is becoming increasingly important to know how

successful the grouting has been prior to the start of construction,

techniques for evaluating grouted soil performance have been developed;

but much improvement is needed. Some techniques are discussed in this

section. Halliburton (1975) discusses

15.42 Ground Water Control

15.42.1 Core Borings

field test&g in greater detail.

This technique consists of drilling core holes into

the grouted soil .mass and recovering grout-soil samples. These samples

-293 -
can then be tested in a laboratory to determine the permeability

characteristics of samples. Since the samples are difficult to obtain

and since there are no standarized procedures for testing grouted

soils, this method is of limited value.

15.42.2 Pu,mping Tests

Pumping tests, similar to those preceding the

grouting operation can be performed. Perhaps, the easiest test to

perform is the test using water and calculating the resulting permeability.

A slight variation of this test is to use a very low viscosity chemical

grout and calculate the permeability based on the known flow and viscosity

at the time of pumping. The grout will eventually gel and further reduce

the permeability (Halliburton, 1975).

15.42.3 Flow Tests

In certain instances it may be possible to judge

the effectiveness of a grouted soil mass by observing the flow through it.

Two methods could be used to evaluate the grout curtain. By pumping

on one side of the grout curtain and observing the loss of head on both

sides of the curtain the effectivenss of the ground water barrier could be

determined. Alternatively, dyes could be injected on the side of the

curtain away from the pump and the travel times observed.

15.43 Soil Stabilization

At present the methods of evaluating the effectiveness of

grouting to stabilize a soil mass are pr imi.tive. The only widely

accepted method of determining the in situ strength is to take core

borings and test the recovered samples in a laboratory. However, the

same problems apply in this type of testing as in permeability testing.

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CHAPTER 16 - GROUND FREEZING

16.10 INTRODUCTION

The first reported use of ground freezing as a stablization method

was for a mine shaft excavation in South Wales in 1862 (Maishman, 1975).

The process was patented in Germany by F.H. Poetsch in 1883,. The

basic’ mea&d of circulating cooled brine, through underground tubing

described in the patent, known as the “Poetsch Process”, remains

the basic process in use today.

The fundamental process in ground freezing is the removal of

heat from the ground to cause lowering of subsurface temperature below

the freezing point of moisture in the pore spaces. The frozen moisture

acts as a cementing agent to bind,the soil particles together and as a

structural support framework in the soil mass. Heat is removed by

circulating coolants through pipes installed from. the surface into the zone

to be frozen, and the heat removed is transferred into the atmosphere.

In practice, a designed pattern of freezing pipes or “probes” is

placed in the zone to be frozen. The probes are commonly two pipes of

different size, one within the other, so that the coolant can be pumped

into one and extracted or allowed to escape from the other. Freezing

in the soil progresses radially outward from the probes as a frozen

cylinder along the length of the probe. The cylinders eventually coalesce

between probes to form a wall or zone enclosing the area to be excavated

with a mechanically strong and impervious barrier within the soil mass.

Closed systems, where the coolant is continuously circulated,

cooled, and recirculated through the heat removal system, are the most

common techniques used. In open systems the cooling is accomplished

by sublimating a solid (typically CO2) or releasing pressurized liquefied

gas to evaporate in the zone where cooling is wanted.

16.20 DESIGN AND THEORETICAL CONSIDERATIONS

16.21 Design Parameters

Basic design parameters considered necessary for a ground

freezing program include the thermal, hydrological, and mechanical

properties of the soil ‘mass to be frozen.

16.21.1 Thermal Properties

a. Initial subsurface temperatures (T,)

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b. Volumetric heat (C) of both the fluids and

solids in the zone to be frozen, or the ratio

of the amount of heat required to change the

temperature of a unit mass of material one

degree to the amount of heat required to raise

the same mass of pore water one degree.

Frozen and unfrozen soils have different heat

capa citie 8. Mois’ture content (w) (weight of

water in percent of dry weight of soil) is the

major factor that must be considered in

calculating heat capacity. The approximate

volumetric heat capacity is:

C

U

= $, (0.2 ++) in BTU/ft3/OF (unfrozen)

cf = Iid (0.2 .++$q in BTU/ft3/OF (frozen)

where:

xd = dry unit weight of soil (in pounds

per cubic foot, pcf)

Typical values for dry unit weight and water content

of soils are given in th.e table below:

WATER CONTENT AND DRY UNIT WEIGHT

OF TYPICAL SOILS,

Soil Type

Silty or clayey well-graded sand

and gravel

Clean well-graded sand and

gravel

Well-graded sand

Poorly-graded sand

Inorganic silt or fine sand and

silt

Stiff to stiff clay very

Soft to medium clay

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Typical Values

if

(% drrwt. ) (PC:)

5 140

8 130

10 120

15 110

15 - 25 LlO - 85

20 - 3 0 9 5 - 80

30 - 40 8 0 - 70
C . After the temperature of water is just

lowered to 32 F, Latent heat of fusion (L)

of the pore water is the amount of heat

removal needed to convert the water to ice.

Because latent heat is large compared to all

other heat losses, it usually represents the

most important factor in the freezing process.

144 BTU are required to convert one pound of

water into ice (or approximately 80 cal/gm).

L = Id 0.8~ gm-Cal/cm3 ( 3, in gm/cm3)

o r

L = 1 d 1.44~ BTU/ft3

d. Therma.L conductivity (K) expresses the

quantity of heat transfer through a unit area

in unit time under a unit thermal gradient.

Typical values for soils are about 1.0

BTU/Hour -ft-OF and about 2.0 BTU/Hour-ft-OF

for frozen soils. Thermal diffusivity (or

temperature conductivity) is the quotient of

conductivity and volumetric heat capacity

(& =K/C). Kersten (1949) provides a summary

of thermal conductivities for typical frozen

and unfrozen soils.

16.21.2 Mechanical Properties

General

A frozen soil mass is a visco-plastic material

in that it will creep under stress application. Normally the creep rate,

rather than ultimate strength, will control the design. The latter,

however, is a useful index parameter in assessing creep. Tests may

be performed in the laboratory or in situ. Schuster (1975) uses in

situ pressuremeter tests for determination of short term deformation

characteristics.

Creep

The creep rate of frozen soil is dependent upon

stress Level and temperature. Typical behavior patterns are shown in

Figures 138 and 139. Figure 138 shows the effect of increasing

compressive stress on axial strain. Figure 139 shows strain increase

with both higher stress and higher temperature.

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3 0

2 5

2 0

15

IO

5

GRAY, SILTY CLAY

W=45%, sds 1180per cu m.

e= -2s”c, Cs 5 0.0 kq per sq cm.

t& = (0; - c3> = DEVIATOR STRESS

IO I5 2 0

Tf3

DURATION Of= LOADING (T), HOURS

(AFTER SHUSTER.1972)

Figure 138. Stra in versus time and load.i ng for a frozen soil,
I60 G-l-7

tf= TIME TO

CREEP FAILURE

TIME (t), HOURS

( A FTER SANGER 1968)

Figure 139. Creep curves for an organic silty clay with

temperature influences.

-299-
Stress is held constant for each of the three curves.

Point “F” in Figure 139, represents the line

at which the rate of strain becomes progressively greater. Sanger

(1968) refers to this as creep failure.

Creep tests, such as those shown in Figure 139

are carried out under constant stress and temperature while measuring

strain. In any given installation the designer must be assured that

actual stress levels are safely below values that would produce

excessive creep over the duration of the project.

Ultimate Strength

A summary of ultim.ate compressive strengths

of common soils as a function of temperature beLow the freezing point

of water is given in Figure 140. As may be noted, sandy soils have

greater strengths than clayey soils. As the clay content of the soil

increases, the shear strength decreases.

The strength of frozen granular soil at a given

temperature increases as the moisture content increases. Figure 141

shows the ultimate compressive strength increase of frozen sand. The

figure also shows that the strength of a clay does not increase with

moisture content.

16.21.3 Geometry and Capac:ity of the Freezing System

Cost and time factors for ground freezmg programs

are strongly influenced by both the ge0metri.c arrangement of the freezing

probes and the capacity of the refrigeration equipment. The ground

freezing process proceeds radially outward from each of the freezing

probes, and the rate of progess is a function of:

1.

2.

r

3.

4.

The capacity of the equipment relative to the

thermal load of all of the combined probes

and surface piping.

The thermal gratdient between the probe and

surrounding materials.

The rate of heat transfer between the probefrozen ground system and the unfrozen soil

mass.

Fringe losses at the freezing frqnt due to

ground water flo8w.

-3oo-
I

owc

I

ULTIMATE COMPRESSIVE STRENGTH (Q&,kg/cm*
2000

.a

%

c

T

a 1500

iz

E

cn

5 IOOC

Gi

%

E

5 5oc

v

(

SOIL HELD CONSTANT,

5 IO 15 20 23 53

WATER CONTENT (w), %

% 1200

.

E4 1000

gIs 800

y 600

z

wit 400

8 200

WATER CONTENT(w:,%,

(GENERAL UNDERGROUND

STRUCTURES INC. 1975)

Figure 141. Ultimate short term compressive strength

of ground vs. moisture content.

-302-
In the design process , increased freezing rates

can be obtained by decreasing freeze element spacing and/or increasing

the temperature differential by increasing the capacity of the cooling

equipment.

Fringe losses are :reduced as the radial freezing

fronts converge betieen probes since both the frontal areas between

frozen and unfrozen masses are reduced and thermal Losses due to

ground water movements through the freezing mass are effectively

blocked.

16.22 Approaches To Design

16.22.1 Thermal Considerations

The analysis must consider two basic phases of

operation including (1) reducing the temperature of the soil mass to a

level where the required frozen ground behavior wiLl be obtained, and

(2) maintaining all or some part of the frozen mass at a temperature

where the mass will behave in a satisfacto:ry and predictable way during

construction a ctivitie 8. All methods are fundamentally an exercise in

heat transfer from the ground to the atmosphere.

The basic approach to simplify the analysis is

to (1) identify the zone to be frozen, (2) establish existing temperatures

and temperatures after freezing, and (3) compute the amount of heat

loss required to transfer the volume of soil in the zone from existing

condition to frozen condition. This simplification implicity neglects

temperature drops (and therefore heat loss) at distances beyond the

ice front. However, for practical applications the heat loss within the

frozen zone is large compared to heat losses beyond the frozen zone.

frozen zone are:

The total heat losses that occur within the

J’= Q = Uheat flow from soil, solids, and pore

water required to drop temperature

from soil temperature,To,to the

freezing temperature, Tf .

Q, = Latent heat flow to transfer from water

to ice (occurs at constant temperature,

of Tf).

Q, = heat flow from soil, solids, and pore

water required to drop temperature from

freezing point, Tf, to the design subsurface

-303-
temperature, T2. Therefore, the total

heat loss from a unit volume of soil is:

*u =: Cu (T 0

- Tf)

*L

:: #, (1.44) w

*f

:: Cf tTf - T2)

where:

TO

2 Initial ground temperature

(usually mean annual temperature).

Tf = Freezing temperature.

TZ = Final temperature.

Cu and C

f

= are as previously defined, heat

required to drop temperature

one degree per unit volume.

Typically the latent heat is large compared to the

volumetric heat associated with temperature drop.

16.22.2 Mechanical Considerations

Consideration of creep is fundamental and indeed,

for some cases special measures will be needed to offset contingencies

associated with excessive creep which is especially important with frozen

arches or tunnels.

Open surface excavations with frozen walls are

normally designed as simple massive gravity structures or as cantilevered

beams. The latter requires Less thickness of frozen’earth.

16.22.3 Ground Movement Considerations

Knowledge gained from studies of frost action

below pavement indicates that clean, free draining soils have insufficient

fines to develop’capillarity and therefore do not develop ice segregation.

An old rule of thumb is that soils having more than 3 percent by weight

finer than the 0.02 mm size are frost susceptible.

-304-
Ground freezing below pavement differs from

ground freezing as a construction method in severa. ways: (1) it is

much slower; (2) the ice front is usually parallel to the stratigraphy;

and (3) it is typically in the capillary zone above the water table.

Where ground freezing is used as a construction

method, frost heave is not a problem with free draining, non-frost

suspectible soils. In poorly drained soils, heave is generally attributed

to two separate phenomena. The first is an approximate nine per cent

expansion of pore water upon freezing. The second is expansion from

pore water migration and ice segregation.

Rapid freezing can be used as a device tomitigate

ice segregation. However, after a period of time when the rate of ice

front advance slows down or stagnates, th.e threat of ice segregation and

associated expansion will increase. In su.ch cases, careful monitoring

is essential, especially where structures are adjacent to the excavation.

16.22.4 SeLection of Freezing System

Figure 142 shows the basic elements of some

freezing systems that have been used.

The most common. and Least expensive method

of soil freezing in use today is the Poetsch Process and is essentially

the same system used by Poetsch in 1883. The system consists of an

ammonia or freon primary refrigeration plant to chi.LL a secondary

brine coo.lant which is circulated into freeze pipes in the soil. Depending

on the brine, temperatures to -65Oc can be obtained. The most common

system uses ca.lcium chloride as the brine with a minimum temperature

of -4oOc.

Additional methods of freezing are now being

used which have as their principa.1 advantage a much lower operating

temperature at the soil interface and a resultant much quicker freezing

time. SpecificaLly, the alternatives to the Poetsch Process can be

broken down as follows:

a. On-Site Refrigeration Plant

The first alternative is an on-site refrigeration pLant with the primary

refrigerant pumped directly into the freezing pipes. This system has

been tried using ammonia. One disadvantage is that the system operates

under a vacuum making leaks undetectable. With carbon dioxide, the

system operates under high pressure to keep the CO2 liquid. Hence,

expensive high pressure pLumbing is required.

-305-
REFRlGERATlON

lJNlT

BRINE

CHILLER

CIRCULATING COOLANT RELIQUEFICATION LIQUID/GAS

SYSTEM SYSTEM

CHILLED

COOLANT

CONDENSOR

t

f

1

EVAPORATING

REFRIGERANT

.

c

l

I

t

I

I

EVAPORATING

REFRIGERANT

LIQUIFIED

GAS

EXPANDABLE GAS

SYSTEM

Figure 142. Basic refrigeration system elements for ground freezing.
b. Primary and Secondary Refrigerants

A second alternative is to use a thermally cascaded system employing

a primary refrigerant which can produce low temperature and a secondary

refrigerant capable of transmitting this low temperature. A system using

freon as the primary and CO2 as the secondary coolant seems the most

feasible and would be capable of temperatures of -2OOC to -55OC. The

problem with this system is that field cont:rol of the secondary refrigerant

is more expensive. Improved techno.logy in the field, primarily in the

direction of simple control units , will make this approach practical.

C . Expendable Refrigerants

A third technique is to use expendable refrigerants, such as commercially

available liquid nitrogen. A less efficient but cheaper alternative might

be solid or .liquid CO2. These materials are piped into the ground and

then vented to the atmosphere. Expendab1.e refrigerants are maintained

at a lower temperature than can be achieved in the brine by on-site

refrigeration units; and therefore, their rate of freezing will be more

rapid. Typically, they are used for a sho.rt term and/or emergency

situations. See Figure 143 for an examp.le.

The basic freezing; method consists of choosing

one of the freezing processes discussed above and drilling freeze holes

into which the freezing pipes are installed. A cylinder of frozen

material forms around the pipes and increases in size until the heat

gain at the perimeter is equal to the heat taken out in cooling. The

freeze pipes are installed such that the final frozen zones will overlap

and a continuous barrier will be formed.

In the freezing process, the greatest amount

of heat removal required is to actually change the water from liquid to

solid: i. e. the latent heat of fusion in the soil mass. Once the desired

size of the frozen zone has been reached, the energy requirement to

maintain the frozen condition in equilibrium is considerably less than,the

energy required for initial freezing. Therefore, the capacity of the

refrigeration plant can be reduced after initial freezing.

16.30 CONSTRUCTION PROCEDURE

16.31 General Approach

Freeze probes are emplaced vith spacing (“s”) and probe

size (ro) according to time requirements and required freeze wall thickness for strength. Strength requirements are based upon the type of

frozen structure (i. e. gravity wall); strength requirements determine the

average temperature of the frozen mass. A photograph showing piping

-307-

connections between a series of freezing pipes is shown in Figure 144.

Obtaining the desired ice waLl thickness is usually not

a problem unless goundwater flows in excess of about 6 feet per day

are encountered. Frequently, low temperature freezing techniques

are employed to overcome heat Losses to the moving water above this

range.

Special care must be taken when drilling the holes and

placing the freeze pipes to insure proper alignment. This is a very

critical part of the operation. If the freeze pipes are out of line,

closure of the freeze wall may not be adequate to prevent leakage of

ground water . In this regard, the interfaces between soil and bedrock

or between sands and underlying clays are critical. A closely

monitored freezing program is required to prevent any gap in the

freeze wall.

It is common practice to design the frozen structure so

that it either bottoms in an impervious stratum or a frozen bottom is

part of the design. When the former procedure is used, the freezing

probes are commonly inserted several feet into the impervious zone

to assure that watectight closure of the frozen structure is accomplished.

16.32 Protection of the System

During the construction process, care must be taken to

avoid mechanical damage to the distribution system that might cause

loss of refrigerants and leaks in the frozen wall.’ Maintenance of the

frozen mass of earth after it is formed depends on a constant removal

of heat to compensate for any heat gain at the fringes of the frozen zone.

Open excavations are coxnmonly covered with reflective thermal insulation

that provides protection against sun and rain. An aerial view of a

protected freeze waLl is shown in Figure 145.

16.33 Special Construction Problems

Special details are necessary to work in areas containing

existing utilities, especia.Lly steam, water, and sewage. Not only can

these conduits be frozen and flows interrupted, but if.not frozen, they

constitute a heat source and a potential leak in the freeze wall. One

possible solution is to temporarily reroute the utilities, or if freezing

must proceed through the utilities, the utilities can be insulated prior

to freezing so that the 3Z°F isotherm remains in the insulation.

-309-


16.34 Construction Monitoring

Monitoring subsurface and brine temperatures is a

requirement during construction of a froze:n ground structure. This is

usually accomplished by measuring the profile of subsurface

temperatures in small diameter observation pipes (1”O. D., or so)

distributed throught the frozen zone. Comznercially available

thermistors or thermocouples are widely used as the temperature

sensor, and relative.ly inexpensive readout devices are adequate for

the monitoring requirements. Whether a problem exists in the

refrigeration system, or in unexpected subsurface conditions, can

normaLly be detected with an accurate profile of subsurface temperatures

and routine coolant temperature data obtained during plant operation.

-312-