Technical Report Documentation Poge

1. Report No. 2. Government Accession No. 3. Recipient’s Catalog No.

-HWA-RD-75-129

4, Title and Subtitle

LATERAL SUPPORT SYSTEMS AND UNDERPINNING

Volume II. Design Fundamentals 6. Performing Organization Code

7. Author(s) Donald T. Goldberg, Walter E. Jaworski,

and M. Daniel Gordon

9. Performing Organization Name and Address

8. Performing Organization Report No.

1363

10. Work Unit No. (TRAIs)

Goldberg-Zoino & Associates, Inc.

30 Tower Road

FCP 35 B 104-s

11. Contract or Grant No.

Newton Upper Falls, Massachusetts 02164 DOT-FH-11-8499

12. Sponsoring Agency Name and Address

Offices of Research and Development I

13. Type of Report and Period Covered

FHWA

U. S. Department of Transportation

Washington, D. C. 20590

15. supplomentary Note.

Final report

‘14. Sponsoring Agency C$OQ 5 1 4

Contract manager for this study: J. R. Sallberg (HRS-11)

16. Abstract

This report provides current information and design guidelines on cut-and-cover

tunneling for practicing engineers. The main emphasis is on the geotechnical

aspects of engineering. Included in this volume is a state-of-the-art summary of

displacements and lateral pressure. Other topics are basic concepts of soil

mechanics, ground water in open cut, passive resistance, design aspects of

lateral earth pressure, stability analysis of sheeted excavations, bearing

capacity of deep foundations, and construction monitoring. Detailed explanations

of design methods and literature citations are inc:luded.

Other reports prepared as part of the study are FHWA-RD-75-128, Volume I,

Design and Construction; FHWA-RD-75-130, Volume III, Construction Methods;

and FHWA-RD-75-131, Concepts for Improved Lateral Support Systems.

r

f

19. Security Classif. (of this report)

I

20. Security Clarsif. (of this page\ 21. No. of Pager 22. Price 1

Unclassified Unclassified I 249 I

Form DOT F 1700.7 (E-72) Reproduction of completed page authorized
PREFACE

Volumes II and III of this three volume set present

the current state-of-the-art on the engineering aspects of the design

and construction of ground support walls and the closely related

techniques of underpinning, ground freezing, and grouting. So that

the reader will understand the rationale behind the subject matter,

the text contains detailed discussions, especially in areas of controversial or technically new issues. On the other hand Volume I,

a summary of Volumes II and III, is free from the detailed discussions

embodied in the latter two. Its purpose is to provide a ready reference manual.

Overall, the primary intent is to provide information

and guidelines to practicing engineers, in particular those engineers

with an advanced background in the disciplines of Soil Mechanics and

Foundation Engineering.

Volume II incorporates design fundamentals, primarily

those of a geotechnical nature. It places considerable emphasis upon

displacements of adjacent ground and adjacent structures and considers

those parameters which are primary contributors to excessive displacements.

Volume III is directed toward the essential design and

construction criteria associated with each of the following techniques:

(a) Support Walls - soldier pile walls, sheet pile walls, concrete

diaphragm walls; (b) Support Methods - internal bracing and tieback

anchorages; (c) Underpinning; (d) Grouting; (e) Ground Freezing.

Also, it presents an overview of these construction methods with

regard to selection, performance, and relative cost. Throughout,

an attempt has been made to provide a balance between the practical engineering considerations of construction and appropriate

corresponding considerations of engineering fundamentals.

These publications are produced under the sponsorship

of the Department of Transportation research program, a long range

plan to advance the technology of bored and cut-and-cover tunnels,

in particular those constructed in the urban environment.

ii
Part of this program involves a synthesis and evaluation

of existing knowledge and part involves a Research and Development

effort. These volumes fall under the category of the former, “State

of the Art”, aspect of the program from which it is hoped that progress through development of bold innovative approaches will emanate.

. . .

111
ACKNOWLEDGEMENTS

The authors acknowledge with gratitude: The assistance

of Melvin Febesh of Urban Foundation Company, New York, who

contributed so heavily to the underpinning chapter and whose insight

and construction experience assured proper direction to the effort.

The contributions of Dr. James P. Gould, of Mueser,

Rutledge, Wentworth and Johnston, New York, through his review of

the chapters in Volume I dealing with displacements, earth pressure,

and design aspects of land cofferdams. Thanks also to:

To Mr. John Shuster of Terrafreeze Corporation,

Bethesda, Maryland for the review and contributions for the section

on the techniques of ground freezing.

To Mr. John Jones of Schnabel Foundation, Co., Washington, D. C. for his review and comments on Volume III, “Construction

Methods”.

To Mr. D. Maishman of Foraky, Ltd., Great Britain

for allowing the authors the opportunity to review his yet unpublished

manuscript on ground freezing.

To those many engineers, manufacturers, and contractors who graciously ,provided comments, data, and photographs.

To Mr. John Dunnicliff of Soil & Rock Instrumentation,

Inc., of Newton Upper Falls, Massachusetts for the chapter on

Construction Monitoring.

To those individuals associated with Goldberg-Zoino

& Associates, Inc. who contributed greatly to the various chapters

in these manuals: Dr. Stephen A. Alsup on Ground Freezing;

Mr. Peter Riordan on Bearing Capacity; Mr. Jackson Ho for the

outstanding effort on computer work; the efforts of Ms. Andrea Wizer

and Mrs. Joan Jennings for their patience and endurance in the

typing of these manuscripts.

To Mr. Gardner L. Hayward, Jr. for a job well done

tin the drafting of the figures.

And finally, to Mr. J. R. Sallberg, contract manager

for the Department of Transportation and Federal Highway Administration, for his patience, support, and welcome criticisms throughout

the program.

iv
TABLE OF CONTENTS

INTRODUCTORY SECTION

TITLE PAGE

PREFACE

ACKNOWLEDGEMENTS

TABLE OF CONTENTS

CONVERSION FACTORS

SYMBOLS

TEXT

CHAPTER 1 INTRODUCTION 1

1.10 Purpose and Scope 1

1.20 Organization and Usage 1

CHAPTER 2 DISPLACEMENTS

2.10 General

2.20 Characteristics of Wall Deformation

2.30 Magnitude of Displacements

2.40 Parametric Studies

2.50 Distribution of Deformations

2.60 Lateral Deformations in Adjacent Soil

Mass

2.70 Effect of Construction Procedures

2. 80 Estimating Settlements

2.90 Summary

Appendix A

CHAPTER 3 BASIC CONCEPTS OF SOIL MECHANICS 51

3.10 General 51

3.20 Effective Stress 51

3.30 Soil Unit Weights 5 2

3.40 Shear Strength of Soils 5 3

CHAPTER 4 GROUND WATER IN OPEN CUTS 62

4.10 Ground Water Cutoff 62

4.20 Seepage Pattern to Excavation Face 64

4.30 Ground Water Recharge 66

Page Number

i

ii

iv

V

. . .

Vlll

X

Page Number

5

5

6

12

27

3 0

3 9

42

43

43

45

-v-
TABLE OF CONTENTS (conIt)

CHAPTER 5 LATERAL EARTH PRESSURE 67

5.10 Basic Considerations 67

5.20 Internally Braced Cofferdams 7 2

5.30 Tiebacks 8 3

5.40 Surcharge Loading 8 7

CHAPTER 6 PASSIVE RESISTANCE BELOW BASE OF

EXCAVATION

6.10 General

6.20 Soil Parameters

6.30 AnaLysis of Passive Resistance

6.40 Over cut Design Details

6.50 Berms

CHAPTER 7 DESIGN ASPECTS OF LATERAL PRESSURE

7.10 Load on Support Lievels

7.20 Analysis of Wales and Support Walls

7.30 Bracing and Tiebacks

7.40 Depth of Penetration Below Cut

7.50 Example Problems

7.60 Finite Element Analysis of Braced

Excavations

CHAPTER 8 STABILITY ANALYSIS OF SHEETED

EXCAVATIONS

8. 10 General

8.20 Bottom Heave

8.30 Local Failure

8.>40 Deep Seated Failures

Pape Number

96 ’

96

96

99

102

102

103

103

103

110

110

112

115

133

133

133

136

138

CHAPTER 9 BEARING PRESSURE OF DEEP FOUNDATIONS 145

9.10 General 145

9.20 Presumptive Beari.ng Value 145

9.30 Bearing Values Based on Shear Failure 145

9.40 Bearing Values Based on Settlement 150

-vi-
TABLE OF CONTENTS (can’t)

Page Number

CHAPTER 10 CONSTRUCTION MONITORING 168

10.10 Introduction 168

10.20 Planned Construction Monitoring 169

Programs

10.30 Parameters to be Measured 169

10.40 Example s 174

10.50 Recommended Methods of Measurement 177

lo. 60 Contracting for Instrumentation 180

10.70 The Key to Successful Construction 182

Monitoring

BIBLIOGRAPHY 183

-vii-
LIST OF CONVERSIONS

The list of conversions is designed to aid in converting from

British units of measure to metric units. This section has been divided

into two parts; general notation and arithmetic conversion.

General Notation

BTU

CXY-L

2

Cl-l-l

3

cm , cc

cfs

ft

ft2

ft3

fps,

gal

gPm

Ei? gr

h r

in

ln2.

ln3.

k

kg

m

2

m

3

m

min

British Thermal Unit

centimeter

square centimeter

cubic centimeter

cubic feet per second

feet

square feet

cubic feet

feet per second

gaLLon

gallons per minute

grams

hour

inches

square inches

cubic inches

kilo (thousand)

kilogram

meters

square meters

cubic meters

minute

..*

VLLL
rt-run

2

m m

3

m m

ml,

N

lbs

Pcf

Pl f

Ps f

psi

set

British Units

1 BTU

1 in

1 in2

1 in3

1 ft

1 ft2

1 ft3

1 pcf (lbs/ft3)

1 psf (lbs/ft’)

1 ksf (kips/ft’)

1 psi (lbs/in’)

1 l b

Lin- lb

millimeter 8

square millimeter 8

cubic millimeters

mil.liliter s

Newton

pounds

pounds per cubic foot

pounds per lineal foot

pounds per square foot

pounds per square inch

second

Conversions

Metric Units

0.2520 kg - calories

107.5 kg - meters

2.540 cm = 25.4 mm

6,452 cm2

16.103 cm3

30.48 cm = 0.3048 m

929 cm2 = 0.0929 m2

28,317 cm3 = 0.0283 m3

16.02 kg/m3 = 0.01602 g/cm3

4, 883 kg/m2 = 47.9 N/m2

4.45 N

0.1127 N-m

ix
List Of Symbois

The following .list of symbols has been prepared to aid the interpretation of symbol use in the text. This list identifies only the major

symbols used in the text and their general meaning. Each symbol (with

subscripts) is defined in the text for its particular usage. This list is

not a complete list of all symbols or all symbol usage in the text but

is a summary of major symbols and their usage.

0.1 symb

A

B, b

C

C

D, d

E

f

F. S.

H

K

KO

Ka

K

KP

L, 1

N

OCR

Represents Reference

general symbol for area

general symbols for width

cohesion intercept

heat capacity

general symbols for distance

and diameter

Volume I, Chapter 16

Volume III, Chapter 9

general symbol for modulus

general symbol for stress

factor of safety

depth of excavation: also

general symbol for height

general symbol for coefficient

of lateral earth pressure

coefficient of lateral earth pressure

at rest

coefficient of active earth pressure

coefficient of passive earth pressure

thermal conductivity Volume I, Chapter 16

Volume III, Chapter 9

general symbols for length

or distance

general, symboi for stability

number or standard penetration

resistance

over consolidation ratio

X
Svmbol

P

P

PH

R, r

s, s

S

U

U

w

s

v (max)

Jh (max)

E

x

*d

IIm

r

sub

Represents Reference

general symbol for load or force

general symbol for pressure

ne gative logarithm of effective

hydrogen ion concentration

general symbols for radius

general symbols for shear resistance

or shear strength

undrained shear strength

pore pressure

general symbol for weight

genera.l symbol for water content

general symbol for displace.ment

or movement; also angle of walL

fr i ction

vertical displacement (maximum)

horizontal displacement (maximum)

general symbol for strain

general symbol for unit weight;

tota. unit weight of soil unless

other wise specified

dry unit weight of soil

total unit weight of soil

bouyant unit weight of soil

unit weight of water

Poisson’s Ratio

Poisson’s Ratio

general symbol for friction

angle of soil

xi
Symbol Represents Reference

general symbol for settlement

general symbol for stress

total vertical stress

(effective vertica.L stress)

total horizontal stress

(effective horizontal stress)

iif v m

maximum past vertical

consolidation pr e s sure

(effective stress)

general symbol for shear stress

or shear resistance

Note: Line over symbols indicates effective stress parameters are to

be used. (e. g. cv = vertical effective stress).

xii
CHAPTER 1 - INTRODUCTION

1.10 PURPOSE AND SCOPE

This report, VoIume II, discusses the general design aspects

of cut-and-cover tunne,ling, in particular those factors affecting the

design of the ground support wall, such as earth pressure, Lateral

resistance , ground waters and bearing capacity. In addition, considerable emphasis is placed on displacements of adjacent ground and

adjacent structures.

The intent of Volume II is to provide the basic theoretica. elements and design framework with which to approach the engineering

of deep excavations and underpinning. Because emphasis is upon

geotechni cal consider ations, comments on structural factors are

included only when closely related. Legal and contractural re.lationships are not discussed.

1.20 ORGANIZATION AND USAGE

Including this introductory chapter, there are 10 chapters in

this volume, With the exceptions of Chapters 3 and 10, each of the

remaining chapters discusses analytical procedures or a particular

aspect of design. Chapter 3 presents basic soil behavior concepts

for reference, and Chapter 10 is an overview of construction monitoring as related to deep excavations and tunnels. Each of the

remaining chapters is discussed briefly below.

1.21 Displacements - Chapter 2

The disp.lacements occurring in the structures and soil

mass adjacent to an excavation will have an effect on the choice of

wall and the remedial or preventive measures required to protect the

s tru ctur e s . The greater the amount of movement, the greater will

be the protective measures required. An analysis of the performance data re car ded at approximately 60 excavations revealed

several important relationships between wall type, support type,

soil type, and movements .

In sands and gravels and very stiff to hard clays, the

wall type did not markedly affect performance. However, in softer

cohesive soils the stiffer support systems (concrete diaphragm walls)

limited movements to a much greater extent than did the more

-L-
flexible soldier pile or steel sheet pile walls. Some preliminary

conclusions can be drawn from the displacement data analyzed

which may be used as an aid in predicting movements behind a

wall.

By improving the displacement prediction techniques

for movements adjacent to an excavation, the engineer will be in

a better position to evaluate the many factors involved in the

decision to underpin structures. If displacements can be reliably

predicted, particularly with distance behind excavations, the

effect that the movements will have on structures can be evaluated,

and the costs of repair can be compared to the costs of alternative

procedures.

1.22 Basic Soil Parameters - Chapter 3

This chapter is a brief review of the basic soil

parameters and soil behavior that affect lateral support wall

design. Of particular interest are the differences in strength

behavior between cohesionless and cohesive soils. In cohesive

soils, it is important to identify the situations where undrained..

strength or draiaed strength is the critical controlling parameter.

The chapter is not intended to be a complete review of soil properties and soil behavior; rather, it is intended to provide a general

overview of factors affecting wall design.

1.23 Ground Water - Chapter 4

In many excavations, g round water control is often

the most difficult aspect of wall construction. Lowered ground

water levels may cause consolidation settlements in soil profiles

where compressible soils are present. Ground water flow into the

excavation may result in running soils and creation of voids behind

a wall. This chapter summarizes the basic concerns in ground

water as it relates to cut-and-cover tunneling; however, it does not

provide specific details on design of dewatering systems.

1.24 Lateral Earth Pressure - Chapter 5

Lateral earth, water, and surcharge pressures on a

wall represent the driving forces that must be resisted by the

passive resistance of the soil below the base of the excavation and

by the support members (internal bracing or tiebacks). This chapter reviews the established state-of-the-art, advances some new

-2-
concepts, and presents recommended design earth pressure distributions for internally braced and tied-back walls. Lateral pressures

caused by surcharge are also discussed.

1.25 Passive Resistance - Chapter 6

The passive resistance mobilized in the soil below the

cut may play a role in maintaining the stability of the wall and in

controlling the amount of lateral deflection. Design parameters for

passive pressures are presented, and specific design criteria are

discussed,

1.26 Design Aspects of Lateral Pressure - Chapter 7

This chapter discusses the techniques used to evaluate

strut loads, wale loads, and presents some typical design problems

incorporating the principles advanced in the previous chapters.

Specific design recommendations are made for allowable stresses

in steel members, determination of passive resistance, ‘and methods

of analyzing loads in structural members. Recommendations are

made relative to overcut, depth of embedment, and temporary

overstressing.

1.27 Stability - Chapter 8

The base stability of excavations must always be

examined, particularly in softer soil profiles. Methods of evaluating

overall stability as well as shear strains inherent in localized zones

of excessive shear, are also discussed.

1.28 Bearing Capacity of Deep Foundations - Chapter 9

This chapter is most useful in the design of underpinning members. However, this section may also be applicable to

tied-back walls where the vertical load induced in the wall by the

vertical component of tieback load can be significant. Design criteria

and design charts are presented for analysis of failure load and

settlement under various loading conditions.

1.29 Construction Monitoring - Chapter 10

Chapter 10 describes the basic considerations and

reasons for monitoring the performance of a lateral support wall.

-3-
The reasons may be to verify design assumptions, assure structure

stability, observe performance for possible legal litigations, or to

advance the state-of-the-art. The general procedure to be followed

in planning a monitoring scheme is outlined.

-4-
CHAPTER 2 - DISPLACEMENTS

2.10 GENERAL

2. 11 Purpose and Scope

The purpose of this section is to provide insight into displacements occurring adjacent to deep excavations - specifically, into

those factors influencing displacements and into the manner in which

displacements occur.

This section describes the basic performance of excavations

in terms of the magnitude and pattern of soil and wall movements,

Empirical plots are derived from the measured performance and presented. In addition, finite element analyses have been used to help

assess qualitatively the relative influence of the aforementioned

parameters. Together, the empirical studies of performance data and

accompanying computer analyses have provided new insight into the

understanding and prediction of displacement.

Several other empirical analyses of the performance of

laterally supported cuts have been performed. Peck (1969) and D’Appolonia (1971) are perhaps the most widely known, and their work has

been most valuable in the preparation of this section. This present

investigation incorporates data from the more recent cases, including

many with tiebacks and concrete diaphragm walls.

2. 12 Significance of Displacements

While the magnitude of settlement is a useful indicator of

potential damage to structures, the amount of settlement change with

horizontal distance (settlement profile) is actually of greater significance.

This fundamental concept is related to the concept of differential settlement, as opposed to gross settlement.

Horizontal displacements have proven to be a source of

severe damage (Gould, 1975). Therefore,, attention to the threat

of settlement should not cause us to overlook what may even be a

greater source of damage, Indeed, horizontal displacements are often

of greater concern than are vertical displacements in the presence of

underpinned structures (Febesh, 1975).

-5-
2. 13 Relationship to Underpinning

Historically, the decision whether or not to underpin has

been a subjective judgement based upon experience - experience which

reflects local soil conditions, contractors’ practices, attitudes of

engineers, and jurisdictional authorities. Rarely, if ever,

have engineers attempted to base a decision concerning underpinning on a quantitative evaluation of displacements. Rather,

structures within certain preestablished influence zones would be

underpinned. Alternatively, if the cost of underpinning was disprop0 r tionate in relation to the value of the structure and there was

no danger of collapse, one might accept the inherent risk of not

underpinning and make necessary repairs afterwards.

Fundamentally, the amount and distribution of the movements in a soil mass adjacent to an excavation is governed primarily

by soil type, stiffness of support wall, and construction procedures. A

better understanding of how these parameters control displacement

will lead to a more rational assessment of effects on adjacent buildings

and to the development of improved techniques to minimize displacements. Ultimately, these efforts will contribute to the decisions concerning .methods of support and underpinning of structures.

2.20 CHARACTERISTICS OF WALL DEFORMATION

2. 21 General Mode of Deformations

Figure 1 shows the possible range of deformations

for perfectly rigid walls and for walls displaying flexure. Basically

the range of behavior includes translation and either rotation about the

base or rotation about the top. In addition, wall deformation will

include some bulging as a result of flexure -- the amount of bulging

depending upon the stiffness of the wall support system.

2. 22 Internally Braced Walls

The upper portion of internally braced walls is

restrained from undergoing large horizontal movement especially

when braces are prestressed and are installed at or close to the

surface. This produces the typical deformation mode as shown in

Figure 2a. The degree of rotation will depend upon the toe restraint

below the bottom of the excavation.

-6-
(a) INFINITELY RIGID WALLS

I

I

/

I

I

-7-j

I

\

I

\

\

I

I

\

I

-Jm

TRANSLATION ROTATION

I

I

I

Y

I

I

I

I

I

ROTATION

ABOUT BOTTOM ABOUT TOP

(b) WALLS DISPLAYING FLEXURE

/

I

I

i

I

\

FF

FIXED

/ ’

i

\

\

TRANSLATION ROTATION

ABOUT BOTTOM

ROTATION

ABOUT TOP

FIXED

Figure 1. General deformation modes.
(0) TYPICAL FOR INTERNAL BRACING

/

I

4

i

. II

“BULGE” 5

ROTATION

ABOUT TOP

FIXED OR SLIGHT

TRANSLATION

(b) TYPICAL FOR TIEBACKS

FIXED OR SLIGHT

TRANSLATION

ROTATION

ABOUT BOTTOM

Figure 2. Typical deformation of tied- back

and internally braced walls.
2. 23 Tied-Back Walls

If the top of the tied-back wall remains fixed, then the deformation mode is similar to that of an internally braced wall (see Figure

2b, left panel). On the other hand, settlement of the wall, partial

yielding of the ties, gross movement of the soil mass, or shear deformation of the soil mass may result in inward movement of the top and

rotation about the bottom as shown in Figu:re 2b, right panel.

Nendza and Klein (1974) attributed the deformation mode of

Figure 2b, right panel, to a combination of shear deformation,

which contributed to inward movement of the top, and flexure, which

contributed to the bulging effect.

If the soil mass embodied by the tiebacks deforms somewhat

as a unit, the pattern would be similar to that shown in Figure 3.

Here, the top moves inward toward the excavation and the earth mass

mobilizes internal shear. Such a concept was originally proposed by

Terzaghi (1945) in connection with earth-filled cellular or doublewalled cofferdams. Such a deformation mode is not true for all situations, but is very likely in cases of an unyielding base and with the

bottom of the wall restrained from outward movement.

Overall, the deformation mode otf a tied-back wall is complex in that various factors develop in different ways. For example:

1. High prestressing pulls the tiop of the wall into the soil,

thus leading to a deformation mode of outward rotation about the bo-ttom.

In sensitive clays, this condition could induce consolidation (McRostie,

et al, 1972).

2 . Lateral translation of the ent.ire soil mass occurs from

shear strains within a weak underlying cohesive layer or from general

lateral strain following relief of large residual horizontal stress in

highly overconsolidated clay or soft shales. Observations show this

may continue even after reaching full depth. (Burland, 1974; St. John,

1974; Breth and Romberg, 1972; Romberg, 1973).

3 . Very stiff walls, such as diaphragm walls, will display

less bulging from flexure. Therefore, horizontal movement at the top

due to movement of the soil monolith will be large compared to the

effect of flexure and therefore will assume relatively more importance.

-9-
K

WALL

Figure 3. Internal shear development,

horizontal shift at top relative to bottom,
.

2. 24 Comparison of Braced Walls with Tied-Back Walls

Overall, there are insufficient data to quantitatively compare

deformations of internally braced walls with tied-back walls. Performance is highly dependent on construction methods and variables are

many. In competent soils (e. g. granular deposits, dense cohesive

sands, very stiff or hard clays, etc. ) performance data suggest that

tied-back walls display about the same deformation as internally

braced walls.

Although the observational data demonstrate little

difference between displacements with bracing or tiebacks,

from a purely qualitative aspect, a number of factors suggest a

superior performance should be attained with tiebacks in competent soils:

1. In granular soils in which soil modulus increases with

stress level, the prestressed soil mass engaged by the tiebacks is

made more rigid and therefore less deformable.

2 . Tiebacks are typically prestressed to about 125 percent

of the design load and then locked-off between 75 percent and 100

percent of the design load. Frestressing in this manner prestrains

and stiffens the soil monolith. Further, the process pulls the wall

back toward the soil to remove any “slack” in the contact zone.

3 . Internal bracing, if prestressed, is usually prestressed

to about 50 percent of the design load. Typically, the bracing gaitis

in load as the excavation deepens. In contrast, tiebacks, being lockedoff at higher loads, typically maintain the load or experience a slight

10s s of load with time. In the case of internal bracing elastic shortening of the strut continues after installation of the member.

4. Temperature strains are more important with bracing

than with tiebacks because the latter are insulated in the ground. Temperature drop may cause a drop in load and/or contraction of the member. If load remains constant, a 35-degree Fahrenheit temperature

drop from time of installation would cause a 75 foot member to shorten

by about .O. 2 inches.

5 . Frequently, internal bracing is removed then rebraced

to facilitate construction, whereas tiebacks do not have to be removed.
Obviously, the flexure occurring with strut re.moval is affected by wall

s tiffne s 6, span di stance, and concurrent backfill and compaction, Past

experience has shown strut removal to contribute significantly to

settlement of adjacent ground. The settlement is the result of lateral

wall deformation during the process of removing the support.

6. Contractors commonly overexcavate below bracing

levels to facilitate removal of materials. This induces greater movements, especially in weak soils.’ Wi&.tiebacks the contractor maintains

the excavation at or slightly above the tieback level.

2.30 MAGNITUDE OF DISPLACEMENTS

2. 3 1 Reported Horizontal and Vertical Displacements

Displacement of the soil retained by and adjacent to an excavation is a function of several factors including wall stiffness, constructtion technique, etc. Because of the inherent complexities of an actual

installation, it is difficult to isolate all variables and analyze each

separately on the basis of empirical data. However, some indication

of the effect of some variables can be obtained by simplifying the

primary characteristics of a cofferdam (soil type wall and bracing

type) and summarizing and comparing them with the results of field

measurements.

Figures 4 and 5 are an extension of a similar

plot presented by D’Appolonia (1971). The figures show normalized

vertical and horizontal displacements (ratio of the maximum displacements to the height of the cut) versus three general categories of soil

type and support type. References for this data are summarized in

Table 1. Diaphragm walls are distinguished from the relatively

more flexible soldier pile or sheet pile walls by symbol.

Vertical and horizontal displacements in the ground outside

the excavation arise from:

1. Horizontal and vertical displacement of the wall -- in

general, these are rotation, translation, and flexure,

2 . Movement of soil -- for example, loss of soil through

lagging, overcutting and improperly backpacking of lagging, spalling of

-L2-
0.5

a\”.

:: I

1.0

EX

&

:3

2.0

2.5

VERY STIFFTOHARD SOFT TO STIFF

SAND AND GRAVEL CLAY CLAY OTHER SOIL

S& 2ooopsf su ‘2000 psf CONDITIONS

TIEBACKS PRESTR. BRACING PRESTR. BRACING PRESTR. BRACING PRESTR

BRACING (STAND) TIEBACKS(STAND.) TIE BACKS (STAND) TIEBACKS BRACING

BRACING BRACING BRACING BTAND)

025 031 OS

*30 ---- I I I I---- I I I II I I I I l 23b+

I I

1 /////.ë/I////

- LIMITS OF ZONE1 (PECK, 1969)

I

-LIMITS OF ZDNEII (PECK,l969)

I

75% OF CASES EXPERlENCED LESS

MAXIMUM MOVEMENT. I

------ 100% OF CA!%% EXPERIENCED LESS

MAXIMUM MOVEMENT. I

+ ATYPICAL OR UNUSUAL CASES. REFER I

TO THE TEXT ArD TABLE 1.

I I

NOTE: NUMBERS REFER TO REFEdENCES

IN TABLE 1.

l Ir 5 o/o - DISPLACEMENT OFF SCALE,

MAGNITUDE EQUAL TO 5O/&

1 I I I I

,,, ,,,I ,,,,,,,

I I

I I

I I

I I

I

I I

I

(ORGANIC I

‘.p, 3.1 F SOlLS)

SOFT CLAY

AND PEAT)

Figure 4. Normalized vertical movements,

0 = DIAPHRAGM WALL

0 = SOLDIER PILE OR

STEEL SHEETING
0.5

P

8

1.5

2.0

2.5

VERY STIFF TO HARD SOFT TO STIFF

CLAY OTHER SOIL

SAND AND GRAVEL CLAY

su-,2000 psf Su’2000 psf CONDITIONS

TIEBACKS

PRESTR. BRACING BRACING PRESTR.

BRACING (STAND.)

TIEBACKS PRESTR. SRACINS TIEBACKS PRESTR. BRACINS

BRACING (STAN D-1 BRACING (sTAW TIE BACKS SRACCINB ISlANa)

047 04s

l 20

~32

I

l * 03

I

l 25

‘3QQS2

~2&‘-‘:-oM- a%-- 051 I I

041

OS 042

06 058

040 1 1

I I

.I5 044 Qaa

t

lJY;;!D,AwR,

---es

04 S24 I I

WALLS

031

oshfl I

I I I I I

I 1

I

034 ’

I

I

I 1 I

--m-m

I I

2:~~ 1

l 28b

* i i’r2, i i

l IP

1

OII* (Ot?Ih

SOILS)

I

I

I I 1 l II 1 I I

I I I I

+

085 I 022

I 014 I I

-4-e75% OF CASES EXPERIENCED LESS I L coRgfTy

MAXIMUM MOVEMENT. STEEL (SOFTCLP

WALLS 061

- -- -- 100°~ OF CASES EXPERIENCED LESS ,

MAXIMUM MOVEMENT. I

I -

- - .I7 a

a87

,,

/ ~AN~~IZ*

+ ATYPICAL OR UNUSUAL CASES. REFER STEEL

TO THE TEXT AND TABLE t. WALL I I

NOTE: NUMBERS REFER TO REFERENCES

IN TABLE I. I I *se

I 1

0 4 5 % - DISPLACEMENT OFF SCiLE, MAGNITUDE

EQUAL TO 5% I I *ai* l *

1

4: I I ----- I I I I I 1 I I 1

I

Figure 5. Normalized horizontal movements.

O= DIAPHRAGM WALLS

. = SOLDIER PILES OR

STEEL SHEETING

;. -_--_ .,.
Table 1. Summary of references on displacement,

Ruf.k Author(sl

Wail Brac,nC:

lype ‘Type Soil Type

Depth d,,max d,

of Cut “lax Commc”ts

-

I Lambe, Wolfskili, SSP strut, Fill, Organic 5X’ 7” 9” Consolidation settlements s,gnificant.

Ri Wang (19701 (Prcstrcsscd) Silt, till, rock (17. 7m) (17. 8cm) (22.9~~1) Settlements of 3” (7. 6cm) up to 70’

(21.3m) from excavation.

2 O’Rourke and SP St*UtS Dense Sand and 60’ 1.5” 9!’ Removal of struts increased settleCordmg (1974) (Prcstressed) gravel, Stiff (18.3111) (3, 8cm) (2.3cm) “lent from 0.9”(2.3cm) to 1. 5”

(G St. Excavatmn) day (3.8cm).

3 O’Rourke and SP struts Dense Sand and 82’ 1.5” I. 25” Some tune-dependent consolidation

Cording (1974) (Prcstresscd) gravel, Stiff (254 (3.8cm) (3.2cm) settlements.

(7th !a G Streets) clay

4 O’Rourke and SP Tiebacks Dense Sand and 40’ , 7” 2” Street settlements small while

Cording (1974) and gravel, Stiff (12.2m) (1.8cm) (5. Icm) soldier piles settled due to downWare, Mirsky, day 2” drag from tiebacks. Soldier piles

and Leuniz (1973) (5. lcm) settled 2”/5. lcml maximum.

5 Lambe, Wolfskill D W struts Fill, hard to 50’ 1 ” 1. 2” Minor consolidation settlements.

and Jaworski (Prestressed) medium clay, (15.2m) (2. km) (3. ocm) School located 5’(1. 5m) from wall.

(1972) till

6 Burland (1974) and D W Struts (Slabs Gravel and 52’ .6” 1”

St. John (1974) poured as very stiff (15.9m) (1.5cm) (2.5cm)

(New Palace Car excavation clay

Park) proceeded)

7 Burland (1974) and D W Tiebacks Very Stiff clay 26’ 1.1” 2. 2” Much of the wall movement was pure

St. John (1974) (7. 9m) (2.8cm) (5. 6cm) translation and continued with time.

(Neasden Under- Extremely small vertical settlements

8 OlRourke and SP struts Dense Sand and -- 2 % _- Did not report depth of excavation or

Cording (1974) (Prestressed) gravel and stiff amount of settlement.

I I 1th & G Streets1 Cl?l”

9 Burland (1974) and D W Slabs and Gravel and 52’ . 5” . 6”

St. John (1974) Tiebacks very stiff (15.9m) (1.3cm) (1.5cm)

(London YMCA) CL%” I

10 N.G.I. (1962) SSP struts Soft to medium 19. 5’ 3” -_ Consolidation settlements due to

(Oslo Technical clay (5.9m) (7.6cm) lowering of head in underlying

School) sand.

1, N. G. I. (1962) SSP struts Soft to medium 36’ 8.9” 5.1” Nearby underpinned structure

(Vaterland #2) (Prestressed) clay (llm) (22.6cm) (13cm) settled significantly.

12 McRostie, Burn SSP Tiebacks Medium ta 40’ 4.5” -4” Excessive tieback prestressing

and Mitchell stiff Clay (12.2m) (11.4cm)(-10.2cm) pulled wall away from excavation.

(1972) Sensitive clay consolidated due to

shearing stresses.

13 DiBiagio and D W Floor slabs Medium clay 62’ 1.6” 1-l. 2” Structure ( 2’(0. 6cm) from wall.

Roti (1972) used to (18.9~1) (4. lcm)(2.5-3. Ocm) All settlement appeared to be due

support wall to lateral wall deflection.

See Sheer 5 for notes. Sheet 1 of 5
Table 1. Summary of references on displacement. (Continued. )

Ref.# Author(s)

Wall

Type

Bracing

TY!X Soil Type

Depth d

of cut <““lax rnax C”nune”ts

I

14 N. G. I. (1962) SSP struts Soft to medium 37, 7” 6. 3” Part of excavation performed

(Grdnland #2) Cl.XV (11.3ml (17.8cml llb.Ocml under water. I

15 Shannon and SP Tiebacks Very stiff clay 78’ 3” 3” Maximum settlement measured at

strazer (1970) and sand (23.814 (7.6~“) (7.6Cm) wall. Settlement may be due to

downwa’rd force exerted by tiebacks.

- -

16 Swatek, Asrow, SSP struts Soft to stiff 701 9” 2. 3” Large settlement attributed to

and Seitz (1972) (Prestressed) clay (21.4m) (7.2.9cm) (5.8~1) localized heavy truck traffic.

Typically settlements < 5”( 12. 7cmI.

17 Rodriquez and SSP struts Soft to medium 37’ _- 7. 9” Staged construction to minimize

Flamand ( 1969) (Prestressed) clay (11.3m) (20. Icm) movsments. Dewatered to prevent

18 Scott, Wilson and SSP struts Dense fine SO’ _- 8” Poor performance attributed to

Bauer (1972) sands (15.3m) (20.3cm) poor construction techniques and

dewaterina problems. Nearby

19 Chapman, Cording SP Struts and Running soil encountered in one

and Schnabel (1972) Rakers ;:“,i;“d

(Prestressed) stiff cla

( p,: ;rn,<I

20 Boutsma and SSP struts Soft clay and 33’ 14” 6” Some settlement due to extensive

Horvat (1969) soft peat (10. lm) (35.6~1) (15. 2cm) dewatering for long time period.

Affected structures 600’ from I exca”atl0”. Liquefaction of backfill during extraction.

21 Insley (1972) SP Rakers Soft to medium 25’ -_ 2.5” One sectron tcstcd to fallurc.

Clay (7.6m) (6.4cm)

22 Tait and Taylor SSP Struts and Soft to 45’ 6” 7.5’, Larger movements attributed to

(1974) Rakers medium (13.8m) (15.2cm) (19. Icm) lack of firm bottom for wall.

(Prestressed) clay Utility lines damaged; no ma~ol

damage to ad,acont structures.

23a Hansbo, Hofman, SSP Rakers Soft clay 23’ 13.8” 11.8” Poor sheet pile interlocking. Lotis

and Mosesso” (7. Or”) (35. Icm) (29.9~1) time between excavation of center

(1973) portion and bracing. Disturbance

durmg pile driving for foundation.

23b Hansbo, Hafman,

and Mosesso”

(19731

SSP Tiebacks and Soft clay 2” 2”

Rakers (7% (5. Ian) (5. Icm)

Improved construction techniques.

24 Prasad, Freeman, SP Tiebacks Very stiff

(1: 84

- - -2” Top of wall moved away from

and Kla~nerma” Clay (-5. Icm) excavation. Maximum movcmcnt

(1972) at top.

25 Mansur and SP Tiebacks Very stiff to 45’ . 5” . 5”

Alizadeh (1970) hard clay (13.8m) (1.3~111) (1.3cm)

26 Sandqvist (1972) SSP Tiebacks Sand and silt 19.5’ 7.9” 2” Scttlemcnt in organics due to

with organic (5.9m) (20. lcm) (5. ICI”) lowered ground water level. f’ilc

soils driving also caused settlement.

See Sheet 5 for notes. Sheet 2 of 5

-16-
Table 1. Summary of references on displacement. (Continued. )

Ref. # Author(s)

Wall

TYPO

Bracing

TYPO Soil Type

Depth (/, dh

of cut max max Comments

27 Sigourney (1971) SP Tiebacks Clayey sand 20-26’

and hard clay (6.1-7~ 9m) --

. 5”

(1.3cm)

28 Goettle, Flaig, SP Tiebacks Dense sand 23’ . 25” . 25” Structure with footings only 2’

Miller, and and gravel (7. Om) (0.64cm) (0.64cm) (0.62~1) from wall was undamaged.

Schaefer (19741

29 Sigourney (1971) SP Tiebacks Very dense 35.43’ -- . 1”

silty sand (10.7- (0. 25cm)

and eravel 13. Zml I

30 Clough, Weber, and SP Tiebacks Vary stiff 64l I 25”t Top of wall moved away from

Lamont (1972) clay (19.‘6m) (3.. 2c$ (2.:‘& excavation.

31 Nelson (1973) SP Tiebacks Sandy over- 90: 1 ” 4” Cracking in’street indicated potenburden, hard (27.5m) (2.5~~11 (10.2cm) tial stabiliti failure (b’hax[ 15. Zcm])

clay shales Maljian & Van Beveren (1974).

32 Liu and Dugan SP Tiebacks

(1972)

Dense sand and 55,’ 8”‘~ 1”i Tops of soldier piles pulled away

gravel, very (16.8m) (2. Ocm) (2.5Cm) from excavation during prestressstiff clay ing.

33 Larson, Willette, SP Tiebacks Dense sand 50’ 1” 1 ”

Hall, and (15. 2m) (L.5cm) (2.5cm)

Gnaedinger (1972)

34 Dietrich, Chase, SP T,ebacks Silty sand 23.54’ 2.5” 1.8” Lateral movements measured at

and Teul (1971) (7.16.5m)(h. 3cm) (4. 6cm) top of wall.

35 Cunningham and D W Tiebacks Medium clay 23’ _. 4” Ticbacks anchored to deadman.

Fernandez (1972) under dense (7. Om) (10.2cm)

sand

36 Cole and Burland D W Rakers Very stiff 60’ 1.5” 2.5” Most movements occurred while earth

(1972) Clay (18.41~1) (3.8cm) (6. 3cm) berm supported wall. Excavation in

heavilv overconsolldated clav.

37 Tait and Taylor D W Tiebacks, Medium and _ _ 9 ” Minor settlements of nearby

(1974) prestressed soft clay (2.3cm) structures

struts and

rakers

I

38 Armento (1973) D W struts Sand and soft 70’ I. 7” I” Some settlement may have been

(Prestressed) to medium (21.4m) (4.3cm) (2.5cml caused bv other excavations in I

Underpinning of nearby footings

required after 5. 5”(13. 9cm) of

settlement. 50.70% oi movement

during caisson construction.

Settlement estimated on basis of

substantial damage to structure

See Sheet 5 for notes. Sheet 3 of 5

-17-
Table 1. Summary of references on displacement. (Continued. )

Ref. # Author(s)

Wall

TYPO

Bracing

TYPO

Depth d

Soil Type

” dh

of Cut max max Comments

41 Breth and D W struts Hard clay and 60’ _ _ . 4”

Wanoscheck (1969) limestone (18.4m) (I. Ocm) #

42 Huder (1969) D W Basement Slightly plastic 65’ _. 1.4”

slabs as silt and clay (19.9m) (3.6cm)

support

43 Thor, and Harlan D W struts Soft to medium 78’ 1” 1.2”

(1971) (Prestressed) clay (23. Em) (2.5cm) (3. Ocm)

44 Barla and Mascardi S W Tie backs Stiff clay 85’ 2.6” Cracking in nearby structures.

(19741 (25.9m) -- (6.6cmI

I

45 Heeb, Schurr, Bone, SP struts Sand

Hake, and Mu&r (142Y2m,

_- . 8”

(2. Ocm)

46 Breth and Romberg SP Tiebacks Stiff clay _ _ 5.9” Lateral movement of entire

(19721, Romberg and sand (2: Bm, (14.9cm) soil block.

(1973)

47 Schwarz (1972) and S W Tie backs Clayey marl 97.5’ ,211 . 6” Many levels of tiebacks at very

Andra, Kunzl, and (stiff clay) (29. Em) (0.5lcm) (1.5cm) close spacing.

Rojek (1973)

48 Corbett, Davies, D W Rakers Very stiff clay; -- _- . 2” Construction delayed after

and Langford upper sand (0.5lcm) hole opened.

(1974) and gravel

49 Hodgson (1974) D W Tiebacks Fill, gravel 26’ _- . 12” Special construction procedure

and struts very stiff clay (7.9m) (0.3cm) used.

50 Corbett and SP Tiebacks Fill, sand

Stroud (1974) and marl (15516,) -- (Z..cL,

Heave observed 18m from wall.

51 Littlejohn and D W Tiebacks Gravel and . 8”

MacFarlane (1974) very stiff clay (5185,) - - (2. Ocm)

52 Littlejohn and D W Tiebacks Gravel and

MacFarlane (1974) very stiff clay (P4:4m) (,.‘,‘Lm) (2::::)

53 Saxena (1974) D W Tiebacks Organic Silt _ _ 2.7” Tops of some wall sections moved

and sand (I’d: Em) (6.9cm) toward sail by fame amount.

54 Ware (1974) D W struts Sand and

(1:” 9m)

_ _ 1. 25”

Personal (Prestressed) gravel and (3.2cm)

communication stiff clay

55 Goldberg-Zoino SP Tiebacks Fill, organic 45’ 1.5” 1” Vertical settlements due to lagging

& Assoc. Files sand, stiff (13. 8x11) (3. 8cm) (2. 5cm) installation. Most horizontal

clay, till movement away from excavation.

56 Burland (1974) D W Cantilever Very stiff Small .5”

and St. John Wall clay (7% settle- (1.3cm)

(1974) merits

See Sheet 5 for notes. Sheet 4 of 5

-18-
Table 1. Summary of references on displacement. (Conthued. )

Ref.# Author(s)

Wall Bracing Depth ,$

TYPO: TYPO Soil Type of Cut “max d

hmax CO~>,,>CYJlS I

57 N.G.I. (1962) SSP struts Medium .and 26’ 3.9” 5.5” Significant movements aflu)

Telecommunications (Prestressed) soft clay (7.9m) (9.9cm) (13.9~1111 strut removal.

Center

58 N.G.I. (1962) bSP struts Medium and 26’ 4.2” 2,” Lateral deflections probably

Enerhaugen South (Prestressed) soft clay (7.9m) (10.7cm) (5.1cm) more than shown.

59 N.G.I. (1962) SSP struts Medium and 36’ 7. 9” 9”

Vaterland #I soft clay (11.0m) (20. lcm) (22.9cm)

60 N.G.I. (1962) SSP Slabs as Medium to 371 7.5” -- Air pressure and upside dorm

GrBnland # 1 support soft clay (11.3rnl (19.0cm) construction used

61 N.G.I. (1962) SSP struts Medium and 30’ 3.9” 5.9”

Vaterland fi 3 soft clay (9.2m) (9.9cml (14.9cm)

62 Mallian and SP Tiebacks Stiff to very 110’ 3” 2” Maximum vertical scttlemunt

Van Beveren stiff clay and (33.6111) (7.6cm) (5.Icm) atypical for the site--usually

(1974) cohesive sand lateral rnovemrnt greater than

and silt vertical.

h3 Jennings (cases Tiebacks Firm 48’ _ _ 3” Damage to utilities in street

reported by (14.7m) (7.6cm) and building across street.

Littlejohn and --- ------ --- - -

MacFarland [ 19741) Fissured -_ 1.5” Acceptable movements

South Africa (1: 7m) (3.8cm)

--- ----

Clay 74’ _ _ 1.5”

(22.6~1) (3.8cm)

---__------

vary st,ff 48’ -_ . i5”

fissured clay (14.7m) (1.9cm)

--- ---- ----

Soft jointed 59’ ._ 1”

rock (18. Om) (2.5Lrn)

Notes:

Sheet 5 of 5

I. SSP = Steel sheet piling

SP = Soldier pile \vall

DW = Diaphragm wall

SW = Secant wall

2. 6, and d

Y

are maximum horizontal and vertical displacements.

3. R’eference # represents references lIsted by author in Bibliography.

-19-
slurry trench walls, voids created from pulling of sheeting, etc. (See

Volume IlI, Construction-Methods, for a more detailed discussion

of various construction techniques. )

3 . Consolidation of soil -- for example, densification of

loose granular soils from vibration, or consolidation of soft cohesive

soils from lowering of ground water outside the excavation.

4. Base instability or near instability -- excessive shear

strains causeo by the imbalance created by removal of load contribute

to base heave and/or plastic conditions in soil.

5 . Stress relief from excavation -- this reduces vertical

stress below the base and relieves the K. horizontal stress (earth

pressure at rest). In turn, the possible displacement modes are

base heave, shear strains, and lateral strains.

The tabulated performance data indicates the following:

1 . Sand and Gravel; Very Stiff to Hard Clay

Seventy-five percent of the excavations in this material

experienced horizontal movements less than 0. 35 percent of the excavation depth. On the average the clays experienced approximately 30

percent greater movement’(0.35 percent vs. 0. 25 percent of H) than

the sands and gravels. Generally, the performance is not significantly

affected by support method or by wall type. Sheet piling, however, is

uncommon to these soil types due to difficulty in installation. One

probable reason for little apparent difference between wall type and

support method is the fact that the measured displacements are small

(typically less than 0. 10 feet for a 50-foot excavation). Many construction factors can contribute to displacements of similar magnitude and

therefore would mask the variation in displacement caused by wall

support type.

Two anomalous cases (no. 7 and no. 46, Table 1)

reveal a potential source of extraordinary lateral movement of a

tied-back wall retaining a predominately very stiff or hard clay (see

previous discussion in Section 2. 23 on tied-back walls). The mechanism causing this movement still is not clearly defined. However, the

practical implications are to approach similar cases with caution.

Ward (1972) cites horizontal strains as two to three times as large

as vertical strains in overconsolidated London clay.

-2o-
2. Soft to Stiff Clay

Wide variations for both horizontal and vertical displacements are evident. Sixty-five percent of the cases experienced

horizontal displacements which exceed 1 percent for steel sheet pile or

soldier pile walls, whether prestressed or n.ot. The data suggest

prestressing of these walls makes only minor difference.

The largest benefit is derived from concrete diaphragm

walls with prestressed bracing. Indeed, both horizontal and vertical

displacements are no different from those typical for sands and very

stiff to hard clay, being about 0.25 percent or less.

A unique case is included in Table 1. In this case the

wall was tied-back in a stratum stronger than the clay. The case reports

on the performance of tiebacks (McRostie, et al - no. 12) anchored in

underlying bedrock. The soil was a sensitive clay which experienced

significant consolidation settlements due to excessive prestressing

(McRostie, et al, 1972).

Another major cause of settlements in cohesive soils is due to

lowering of the ground water table. These settlements can often be quite

severe (Lambe, Woifski.L.L, and Wong, 1970; I3outsma and Horvat, 1969;

NGI, 1962; Sandqvist, 1972).

2.32 Effect of Wall Stiffness on Lateral Displacements in Clay

Wall stiffness refers not only to the structural elements

comprising the wall but includes the vertical spacing between the

support members. The measure of wall stiffness is defined as the

El inverse of Rowe’s flexibility number for walls -

L4

where:

E = modulus of elasticity of wall

I = moment of inertia. per foot of wall

L = vertical distance ‘between support levels

or between support level and excavation

base

-21-
Concrete walls generally have a much higher EI value than

soldier pile or sheet pile walls, and with comparable wale spacing,.~are

much stiffer. On the other hand, soldier piles or sheet pile walls with

closely spaced support levels may be stiffer than concrete walls with

widely spaced support Leveis .

Figures 4 and 5 indicate that diaphragm walls reduce

the magnitude of the movements in soft to stiff clay significantly

below the magnitude of the movements for the more flexible sheet

pile or soldier pile walls. In an attempt to further refine the

effect of wall stiffness on displacements in cohesive soils, a plot of

observed displacements versus corresponding stability number

(N = ti”

T

) and stiffness factor (4E1) is d

stabi % y number, k

eveloped on Figure 6. The

which consi ers both overburden stress (yH) and

the undrained shear strength (S,), is a me,asure of the relative

strength or deformability of the soil.

1. The maximum lateral displacement rather than the

vertical has been plotted, since consolidation settlements would

introduce a secondary variable.

2.‘ The maximum value of N was calculated on the basis

of the available strength data.’ H was taken as the depth of excavation

to the lower Limit of an intermediate clay stratum where it intersected

the wa.11.

3, The wall-support stiffness was based on the span distance, L, occurring where the “stability number, N, is a maximum. If

N was a maximum at an intermediate excavation depth, L was calculated as the wale spacing plus 2 feet (for overcut). If N was a maximum

at the excavation base, the span L was calculated from the lowest strut

to the excavation base.

The data plotted in Figure 6 demonstrate what is intuitively obvious -- namely, deformations are functions of soil strength

and wall stiffness. The contour lines of maximum lateral wall movement show this trend clearly. These data allow one to qualitatively

a’Ratio of overburden stress to undrained shear strength.

-22-
c I I I i

=-

E

” I =

=w ’ N ‘y

2

6 >=.q =’

X

i=

+v-- =t-u *y I cl
examine the relative change in anticipated lateral displacement for

a given change in wall stiffness and/or stability number of the soil.

As an examp.le, consider a soil with stability number of 6.

Assume we are evaluating PZ-38 sheeting versus a 30-inch thick concrete diaphragm wall, both with 8-foot spans between wale .leveis, For

an intermediate construction condition 2 feet is added to the span distance for overexcavation yielding a length, L=lO feet.

The stiffness factor of the steel sheeting (PZ-38) is:

EI (30 x 10% (281)

3 = (120)4 = 40. 7 psi = 5. 86 ksf

Co r re spondingly, the stiffness factor of a 30-inch concrete wall is:

EI

3 =

(3 x 106 ) x l/12 (12 x 303 )

(12014

= 391 psi = 56.3 ksf

The data in the figure show that the expected maximum

; -.- lateral displacement for the PZ-38 is appr0ximate.L.y 3 inches, whereas that for the stiffer diaphragm wall is approximately 1. 5 inches.

2.33 Wall Movement Versus Settlement

2.33. 1 Comparison for all Cases

Figure 7 compares observed maximum horizontal and vertical displacements for all types of soils, support systems,

and wall types. The absolute magnitude is shown in panel (a) and the

frequency distribution of the ratio of the movements in panel (b). The

figure shows that practically all the vertical displacements fall within

a range of l/2 to l-l /2 times the horizontal displacements, with most

of them lying in the range of 2/3 to l-l /3 times the horizontal movement.

2.33.2 Soft to Medium Clay

Figure 8 compares displacements for soft to

medium clays. The average curve shows that the vertical displacements are generally well in excess of the horizontal displacements

and that the disparity increases with the magnitude of the displacements.

-24-
IO

0 SOFT TO MEDIUM CLAYlSOME STIFF CLAY:

~-~~~~~~~T

019

2 4 6 8 IO 12

sh max. inches

(0)

MAXIMUM VERTICAL AND HORIZONTAL

DEFORMATIONS

g IO

H

6

i?

5

1I 1 r-ue-2

---- 1_-__ -1 0 ---- l -

0-5

O-5

----b

~~~~~~~~

0.4 0.6 1.2 1.6 2.0

&pax

bvmax

(b)

FREQUENCY DISTRIBUTION OF RATIO

OF MOVEMENTS

Figure 7. Comparison of maximum

vertical and horizontal displacements.

NOTE: NUMBERS REFER TD CASE STUDIES

LISTED IN TABLE I.
14

12

IO

:

5 8

.c

-

a

E

Lz

6

+ SOFT TO MEDllJM CLAYS

FOR OTHER SYMBOLS, SEE FIGURE 7

6 8 IO 12

sh mo”. inches

NOTE: NUWJERS REFER TO CASE

STUDIES LISTED IN TABLE I.

Figure 8. Comparison of vertical a.nd horizontal

displacements for soft to medium clays,

-26 -
This difference is believed to be directly attributable to consolidation settlements which are usually the result of

changes in water levels adjacent to the excavation. This situation

becomes more acute where deep deposits of soft clay underlie the

excavation.

2.33.3 Very Stiff to Hard Clays

Figure 9 compares the displacements of these

soils l As mentioned in a previous discussion (Section 2.31), comparatively large lateral displacements have been reported in severa. tieback projects. Notable among-these are cases no, 7 (Burland and St.

John), no. 31 (Nelson), and no.’ 46 (Breth and Romberg),

2.40 PARAMETRIC STUDIES

Finite element studies are useful in providing qualitative information on the behavior of cofferdams. Several studies of this type have

been undertaken for e’valuating the primary parameters affecting bracing

loads and deformations (Wong, 1971; Palmer and Kenney, 1972; Jaworski,

1973; Clough and Tsui, 1974).

The results of a finite element study for evaluating the effect of

wall stiffness on reducing deformations in various soil conditions is

shown in Figure 10. Also shown for comparison are the lines defining

deformation limits from Figure 6.

The finite element computer program used to develop these data

considered only cohesive soils and internally braced excavations. A

brief description of the program, its capabilities, and the soil properties used in the analysis is appended to this section.

Briefly, the conditions assumed in developing Figure 10 are:

1. The excavation was 60 feet deep, with a wall penetration of

30 feet. The wall was supported by five strut levels with

approximately a 10 foot vertical span between each level.

2 . Three wall stiffnesses were analyzed. The E+ ratios were

1. 00, 5. 8, and 230.

-27-
14

12

IO

ia

.-

.

::E

ci,

6

+ VERY STIFF

FOR OTHER

I

/’

/’

TO HARD CLAYS

SYMBOLS, SEE FIGURE 7

2 4 6 a IO Ii

Sh mar. inches

NOTE: NUMBERS REFER TO CASE

STUDIES LISTED IN TABLE I.

Figure 9. Comparison of vertical and horizontal

displacements for very stiff to hard clays.

-28-

3 . A uniform soil profile was assumed in which the shear

strengths were varied to obtain different stability numbers

(N). The value of N used to develop this plot was based on

the shear strength of the soil at the base of the 60 foot

deep excavation,

4 . The finite element analysis models ideal conditions. It does

not account for variations in construction procedure (such as

overexcavation for a strut level) or anticipated construction

events which are reflected in Figure 6.

Figure 10 shows the predicted lateral displacements are less

than the observe,d values for a given condition. This difference is

related to the inherent movements which are a function of the

construction process. Nonetheless, the theoretical results show a

trend similar to that described by the field observations; that is,

the stiffer walls result in lower movements for a given soil condition.

The results of the finite e.Lement analyses should not be taken in the

quantitative sense, The intent is that such analysis should be used

as a guide in the design and in the consideration of various options

for a bracing system.

2. 50 DISTRIBUTION OF DEFORMATIONS

In addition to knowing what the maximum lateral and vertical

displacements will be for a cut, it is also important to know the

influence zone of these deformations adjacent to an excavation.

Primarily, this is related to the distortion the deformations may

impart to a structure, for if it is anticipated that differential

displacements will result in structural distress, then alternate

procedures should be considered.

Currently many engineers rely on judgement and experience in

predicting defor.mation patterns adjacent to sheeted excavations. It

is the intent of this section to provide some information to aid the

engineer in evaluating what deformation patterns might be expected

adjacent to a cofferdam.

2. 51 Vertical Deformations

Peck (1969) suggested envelopes for the zone of influence of

settlements behind an excavation based on field measurements. The

envelope showed that significant vertical movements may occur up to a

-3o-
distance of twice the excavation height from the excavation face

depending on workmanship and soil profile, A refinement of this plot

was undertaken to provide more information on settlement patterns

adjacent to cofferdams. This was accomplished through a series of

normalized plots of vertical deformations versus distance from the

excavation face for three general soil classifications.

Figure 11 illustrates how the observed maximum settlement

patterns behind a wall varied with the soil conditions. The pattern of

movements indicates that maximum movements occur immediately

adjacent to the excavation. Also, one might expect significant movements a distance from the cut equal to twice the depth of the cut. A t

present, there are insufficient data to define any significant difference

in settlement pattern based on soil type or support wall.

Comparing the settlement patterns of sand versus cohesive soil,

the sands show essentially no settlement beyond twice the depth of the

excavation whereas the cohesive soils do. This is most likely caused

by the consolidation in the more compressible soils from lowering

of the ground water table. A factor to consider when viewing these

results is that when settlements are small consolidation can be a

large percentage of the total settlement; hence, in dimensionless

plots the total settlement may appear to extend over a greater

zone than is attributable to lateral movement or shear strain alone.

This is evident in the data presented by Lambe, Wolfskill, and Jaworski

(1972). The measured settlements due to drops in ground water table

appear significant even though the settlements were small (less than

1 inch). On the other hand, when the maximum settlements are large

(NGI, 1962 Oslo Technical School), consolidation settlements do not

appear as significant. The data is further influenced by variations

in the surface elevation caused by ground water fluctuations, freezethaw cycles, and other factors which will also be a greater percentage

of the maximum observed movement in those cases where 8 is

small. v max

Reviewing Figure 11, it appears that both soft clays and the

granular soils experience a significant angular distortion outside a

distance equal to the excavation depth (D/H = 1). The average

lines of settlement ratio versus normalized distance, shown as

dashed lines on the figure, may be used as a basis of comparison of this

distortion. On the other hand, the stiffer clays(Su) 2000 psf) seemto

experience more gentle distortion slope, even though the zone of influence

-3L-
0 OiS I*,0 Ii5 2LO 2;5

I t

AWERAGE ANGULAR

DISTORTION -v-v;IO

-A-A- I4

II

ENVELOPE ------- 5

CLAY -- -13

St,< 2000 DSf --- ---I2

D/H

I.5 +I 2.0 2.5

D/H

LEGEND

6” : SETTLEMENT AT DISTANCE D

b

“lllOX

= MAXIMUM SETTLEMENT ADJACENT

TO THE CUT.

D = DISTANCE FROM EXCAVATION !WIE.

H = EXCAVATION DEPTH

-v-o- 3(J

-o-o- 15

-q--J- 7

-A-A- 6

---- 9

7 I

,: / p/

I

AVERAGE ANGULAR

DlSTORTlON y

-Xx%- 283

-o-o- 8

-n-cl-4

NUMBERS REFER TO PROJECTS

SUMMARIZED IN TABLE I. /

Figure 11. Normalized settlements adjacent to a wall.
extends farther back from the excavation face. Of course these figures do

not show the absolute value of angular distortion. This is determined

by estimating 6 v from Figure 4, based on excavation depth and

max

method of wall support. Then the angular distortion is related to the

differential settlement by:

d=ocn x6

V

max

where:

4 = differential settlement over distance D

. Ocn = normalized differential settlement from Figure 11

6V

= maximum settlement from Figure 4.

max

2. 52 Parametric Study on Zone of Influence

Finite element studies were performed on several of the deformation modes shown in Figure 1. The se analyses were aimed at

obtaining some qualitative information on the settlement profile one

might expect adjacent to the excavation. The analytical approach used

was to apoly incrementally a specified mode of horizontal wall displacement to a soil profile and, using the finite element program

BRACE II, determine the induced settlement profile. The results of

this analysis portray only those constant volume settlements associated

with the wall displacement and ignore settlements associated with

consolidation of the soil.

Figures 12 through 14 show the wall deformations assumed

and the corresponding settlement profi.les predicted by the finite

element pr 0 gram. The results are reported in dimensionless

form. The basic patterns of deformation used were: tilting about

the base, rotation about the top, symmetrical bulging, a combination

of bulging with tilting, rotation and lateral translations.

The se patterns, although ideal, typify the more common modes

of deformations experienced in braced and tied-back walls. In those

cases where bulging of the wall was used to represent the flexural

deformation of the wall, the bulging was assumed to be symmetrical

around a depth equal to two-thirds of the excavation depth.

-33-
6v (x3

6 Vmox

6 v 05

fi,

. . 6v,, = 1.9” I

\J 0) TILTING

a

O/H

I;0 I;3 2;0, 2.5

VlEILL DEFDRMAfKm

MTA SEYWD REFLECTED

ELASTIC STRAINS AND I3

NOT REPRESENTATIVE

OF FIELD BEHAVIOR.

45 I.,0 I,5 &O 1.5

1 /’

.GESTIMATED FELD PERFORMANCE

Figure 12. Finite element prediction of surface settlement

profiles for normally consolidated clays,

-34-
D/H

0 0.5 1.0 1.5 2.0 2.5

01 y GATA BEYDND REFLECTS

I ELASTIC STRAINS AN0 IS / NOT REPRESENTATIVE

/ OF FIELD BEHAVIOR. OF FIELD BEHAVIOR. I

I 6

I

vmax = 5.4” WALL DEFORMATION

a5t - 0

V

;I!

-0

” & h=4in

Ii 2 ’

1 /

/

“rnax 1

4

j

6h,,‘8in

I.01 \

a) LATERAL LATERAL TRANSLATION PLUS FLEXURE FLEXURE \ I

\ I

D/H

O0

0.5 1.0 I.,5 &.O 2,5

/

1

/

WALL DEFORMATION

l----+0

b) TILTING PLUS FLEXURE

0 D’” 09.5 I.,0 I.,5 2:o 2,s 1

--- -Y.GTIMATE~ Flu0 PERFORMANCE

.-- I

WALL DEFORMATION

I I -0

i

6 VmX= 5.6”

C) ROTATION PLUS FLEXURE

Figure 13. Finite element prediction of surface settlement

profiles for normally consolidated clays,

-35-
D/H

o” o-5’ 1.0 I,5 2,x, qs

. WALL DEFORMATION

b,

&J = I.S”

max

a

6, 0.:

6 ?TUlX

1.0

0

sv 0.5

d

vmax

1-a

D/H

I.0 Ii5 q.0 2.5

V .

\

MLL DEFORMAllO)(

I-----CD

SV = 3.0”

mox

b) ROTATION

($5 I.&O I,5 2$ 2.5

. 1 I

WALL DEFORMATION

c---c-D

Figure 14. Finite element prediction of surface settlement

profiles in elastic medium,

-36 -
Two soil conditions were analyzed:

(a) Normally consolidated clay with both the soil strength

and soil modulus increasing with depth as a function of

the effective overburden stress.

(b) Elastic medium where the soil was assigned a constant

modulus with depth. The elastic cases were analyzed

primarily to provide background information to evaluate

the extent to which the elastic strains influence the

results from the yielding soil profiles of normally

consolidated clay.

In addition, the elastic cases provide some insight into the

settlement patterns for cohesive soil profiles where the stability

number is less than 4. The parameters for the cases studied are

summarized in Table 3 in the appendix to this section,

Figure 12 illustrates settlement profiles for ideal tilting,

rotation, and bulging. The first two conditions may be considered

representative of rigid wall behavior, whereas the bulging cases

represent deformations associated with a flexible wa.l.1.

The results indicate that for tilting and flexure the settlements

are concentrated within a distance one -half the excavation depth. Settlements beyond this distance are elastic and probably not representative

of actual field conditions. For both cases, the severe angular distortion

occurs within this zone as expressed by the slope of the settlement

profile. On the other hand, when rotation is the predominant mode of

deformation, significant deformations may occur at distances up to

1. 5 the excavation depth from the excavation face, This should be expected since the maximum lateral deformation is deep. The overstressed zone of soil is also deep, hence leading to a greater zone of

influence. However, the settlement profile for this latter case

suggests a less severe angular distortion along the ground surface.

Figure 13 shows the settlement profiles for wall deformations

which are a combination of rigid wall displacement plus flexural deformations (example, tilting plus Qexure shown in Figure 13b). The

results show the zone of influence is greatly affected by the nature and

volume encompassed by the horizontal wall movement. This behavior

agrees with the measured field performance of braced excavations

(Flaate 1966). The data also indicate that as the zone of influence

-37-
increases from the excavation face, the settlementprofile will have a

gentler slope. Hence, even though the area affected increases, the

actual danger to a structure may not be as severe since the angular

distortion is less.

Figure 14 shows the settlement pattern which results when

the soil is assumed to behave elastically. For all three wall deformations the vertical displacements become essentially horizontal

beyond the distance twice the excavation depth, Comparing

Figures 12 and 13, the same trend of constant displacement occurs , on the latter two figures beyond this same distance, thus suggesting

these deformation,s are related to elastic strains. The elastic

strains are transmitted to this outer zone by tensile stresses

which are not capable of developing in soils. This idiosyncrasy of

finite element analyses leads to behavioral trends which are not in

keeping with field observations. Consequently, the settlements shown

on Figures 12 and 13 beyond a distance of twice the excavation

depth should be discounted as not representative of field cond;tions.

A comparison of Figure 13 with the field measurements

summarized on Figure 11 shows some interesting trends. First,

the finite element predictions give zones of influence and distribution

of settlements similar to those recorded in the field. Conside ring

the deformations beyond a distance 2. 0 D/H from the excavation as

primarily elastic in origin, the finite element analysis gave a zone

of influence which ranged between 0.5H and 2. OH from the excavation

face, Correspondingly, the field measurements show a greater zone

since their results do contain some consolidation settlements not

accounted for in a finite element analysis, Nonetheless, it appears

that the finite element programs can be used to give qualitative

information on settlement profiles.

The effect wall movement has on the zone of influence is another

significant trend. Figure 12 and Figure 13 both show the importance

of minimizing movement below the excavation base which is often

associated with base instability or local overstressing adjacent to the

wall. (Chapter 8 discusses both types of instability in detail).

For all cases it appears that the deeper the maximum movement is

seated, the greater the zone of influence will extend from the excavation face.

-38-
2.60 LATERAL DEFORMATIONS IN ADJACENT SOIL MASS

Lateral movements of a structure have been observed to be

more damaging than vertical movements. Therefore, one should

attempt to evauluate the extent of lateral movement which may occur

as the result of constructing a temporary retaining structure. These

movements are most prevalent in heavily overconsolidated clays which

have large residual horizontal stresses which are released as a result

of the excavation.

Intuitively, one would think that the lateral movements would be

a maximum at the face of the wall and decrease with distance from the

wall. Also, the deflected wall shape would have some bearing on the

distribution of lateral displacements in the soil mass. Unfortunately,

very few measurements are available showing the distribution of

lateral deflection behind a wall.

A few normalized contour plots of horizontal deformation are

presented in Figures 15 and 16. The measurements were made

in heavily overconsolidated clays where tied-back walls were used to

support the excavation. The observations show that the lateral movements were time dependent. -h another similar case, (Burland, 1974 and

St. John, 1974) where only the Lateral movements at the surface were

monitored, the measured horizontal movements were 20 per cent of the max.

imum movement of 0.5 inches at a distance 1.5H from the excavation.

The aforementioned field data suggest two trends. First, the

pattern of the lateral movement follows closely with the deflected

shape of the sheeting. Second, the lateral movements can extend a

substantial distance from the excavation face as illustrated by the data

from Burland (1974) and St. John (1974).

Another factor to consider with respect to tied-back walls in

heavily overconsolidated clays is that the entire soil mass embodied

within the tiebacks may move laterally (Burland, 1974; St. John, 1974;

Breth and Romberg, 1972; and Romberg, 1973). Hence, in these types

of soils the tied-back excavation may not be as successful in limiting

wall movements as they would be in other soil types.

-39-
,L DEFLECTION

0.5

5

1.0

\-

I

I

I

I

\

\

\

\

\

\

\

I

I

O/H - NORMALIZED DISTANCE FROM CUT.

z/H - NORMALIZED DEPTH OF CUT.

1.0

‘?+

1.5

FROM BURLAND (1974)

AND ST. JOHN (1974)

Figure 15. Normalized lateral movement for

tied-back excavation in heavily overconsolidated clays.

-40.
WH II

+I

0 0.5 0.75 1.0 I.5

01 I I n

$0

I -i-, II/ I -

SEE FtGURE 13b FOR

SURF&E SETTLEMENT

‘1 v

I

1.5

c

WALL DEFLECTKIN

D/H

SEE FIGURE 13~ FOR

SURF= SETTLEMENT

Figure 16. Normalized Lateral movements from finite element

analysis for normally consolidated clays.

-4l-
There are .littLe data available regarding the distribution of

horizontal displacements for excavations in a normally consolidated

clay for comparison with the observed data for the heavily overconsolidated clays. Therefore, the results of the finite element studies

used to develop Figure 13 were reduced to provide some insight

as to the distribution which might be expected for ideal conditions.

These results are shown in Figure 16. In contrast to the data from

Burland (1974) and St. John (1974) for heavily over consolidated clays,

the finite element analysis indicates that in this normally consolidated

soil the z&e of significant movement is confined to an area described ’

by a 1 on 1 slope from the base of the sheeting. As expected, it is

within the theoretical yield zone. The movements are largely controlled

by the sheeting displacement.’ The zone of significant movements

increases with depth in the same pattern as the sheeting movements.

2.70 EFFECT OF CONSTRUCTION FROCEDURES

It is well known that construction procedures can have a significant effect on the performance of excavations.

Lowering of the ground water level either by pumping or by

seepage into the excavation can result in significant settlements.

These settlements could be associated with consolidation of the soil

or, in the case of granular soils, the piping of soil into the excavation.

Poor installation techniques for tiebacks or struts can lead to

surface settlements. Tiebacks should be carefully drilled to minimize the soil removed from holes. Also, any voids remaining after

the tieback is installed should be filled with grout. Struts, rakers,

and wales should be tightly wedged and preloaded to prevent movement

of the wall. In addition, hard wood or steel wedges should be used

for shimming to reduce movements caused by crushing. Earth beams

when used to provide temporary support before installing a strut

have been observed to be of little value in preventing wall .movement.

Cole and Burland (1972) and Hansbo, Hofmann, and Mosesson (1973)

report cases where earth berms did little to restrict wall movement.

Even though the entire support system may be in place, the sides

of the excavation may continue to creep inward with time. This problem

appears to be particularly acute in tied-back walls in very stiff to hard

-42-
clays. There is also some evidence to indicate that lagging in soldier

pile walls tends to pick up more load with time in all soils. Excessive

bulging or even failure of some lagging has been observed,

2.80 ESTIMATING SETTLEMENTS

The data presented in the section may be used to obtain rough

estimates of the ground movements which might occur adjacent to a

support wall. The reason for making this estimate is to provide some

additional input to aid in the decision of whether or not to underpin

adjacent structures or utilities.

Settlements may be estimated using both Figure 4 and Figure

11. Once the soil type and excavation geometry are defined, an

estimate of the maximum settlement may be made from Figure 4.

Figure 11 provides a means of estimating the angular distortion

and zone of influence of the ground movements. In the case of cohesive

soils, Figure 5 may be used to estimate the wall stiffness necessary

to limit the settlements.

2.90 SUMMARY

A review of available field measurements shows that wall and soil

movement at the site of a temporary cut are influenced by the soil

conditions, wall stiffness, vertical support spacing, prestressing, and

construction procedures. For any given wall any one of these may be

the most important factor. However, for situations were good construction procedures and typical wall types are used, Figures 4 and 5

indicate that,the magnitude oS.maximu~_veytical_a~d horizontal defle ction is

dependent on wall stiffness and method of support in soft to stiff clays,

but independent of these factors for walls in sands and gravels or

in very stiff clays.

In clays with an undrained shear strength of less than 2000 psf,

flexible walls (soldier piles and lagging or steel sheeting) commonly

experience vertical settlements in excess of 1.0 percent of the excavation depth. The magnitude of the se settlements can be reduced to less

than 0. 75 percent of the excavation depth by strictly controlled

construction procedures (Hansbo, Hofmann, and Mosesson, 1973).

Where stiff support wall systems (such as diaphragm walls) are used

in these soils, the settlements were less than 0.3 percent of the

excavation depths.

-43.
The maximum settlements for all wall types in sands and gravels

is typically less than 0. 25 percent of the excavation depth. For cuts in very

stiff clays, the maximum settlements may be slightly larger although

most of the maximum movements are still less than 0.25 percent of the

height of the cut.

Concerning lateral wall deformations, the maximum for walls in

sands and gravels is typically less than 0. 2 percent of the cut height.

However, the lateral wall movements in very stiff clays are somewhat

larger, reflecting the tendency for these soils to creep laterally with

time. This behavior is most prevalent in situations where tiebacks were

used in very stiff clays. For the very stiff clays, the maximum lateral

movements for internally braced cuts are about 0.2 percent of cut height

while maximum lateral movements for tied-back walls are generally

less than 0.4 percent of cut height.

In general, Figure 4 shows that the stiffness of the wall-support

system aids in controlling movements in virtually all soil types, although

the effect is much less marked in sands, gravels, and very stiff clays.

Figure 11 shows how the vertical surface deformations vary

with distance behind a cut. This data indicates that maximum settlements

can occur at distances equal to the excavation depth from the support

wall. To date, settlement profiles versus depth have not been measured.

Also, it is not known to what extent adjacent structures affect the

observed settlement profiles. In two cases (Lambe, Wolfskill, and

Jaworski, 1972; DiBiagio and Roti, 19721, the settlement profiles were

determined from the settlement of structures on shallow foundations.

These settlement patterns are very similar to those for other cases.

-44-
APPENDIX A to CHAPTER 2

A. 10 INTRODUCTION

Finite element analyses of complex engineering problems are becoming more common each day in the engineering world. These analyses

are often conducted to gain insight into the parameters which control

the performance of a structure. As part of the development of this

manual, computer analyses of braced excavations were made using the

finite element program known as BRACE II (Jaworski, 1973). A brief

description of this program’s capabilities along with the details of the

computer runs used to develop the data in this manual are presented in

this appendix.

A. 20 BRACE 11 - FINITE ELEMENT PROGRAM

A. 21 General

The computer pro gram, BRACE II, was developed expressly

for the purpose of analyzing internally braced excavations. The program

simulates the construction process for a braced excavation in an isotropic, bilinearly-elastic material by performing a total stress analysis.

It models sequentially the events of excavation and installation of struts.

It can also consider such effects as prestressing of struts or additional

movements at the strut levels after installation. Additional capabilities

of the program include the facilities for handling anisotropic, bilinearlyelastic materials and the overstressing of the sheeting piling.

A, 22 Program Description

The use of the finite element technique in analyzing complex

engineering problems is well described in the literature (Zienkiwiez,

1967). Briefly, this technique models a problem by an assemblage of

discrete triangular to quadralateral elements. The forces (8) at the

nodes of these elements is related to the node displacements (U) and the

global stiffness (i4f) of the element assemblage. A system of lgear

equations results which can be described by the equation:

This system of equations is solved to obtain nodal displacements. These

displacements are then used to determine individual element strains and

stresses.

-45-
Tt;lis technique was used in the development of a computer

program, BRACE II, for analyzing the behavior of braced excavations.

This current program is a second generation of the original program,

BRACE, developed by Wong (1971). The program models soil by discrete

elements with bilinearly-elastic stress/strain properties. Within each

element the strain is assumed constant. The retaining wall for the

excavation is simulated by one-dimensional linearly elastic bar elements.

The program simulates a specified excavation and bracing-construction

sequence by applying the load relief due to an excavation stage or by

applying a force at a node as a strut is prestressed. The loads from ’

a particular construction operation are applied incrementally. Fbr

each load increment a specified modulus is used until the yield strength

of the soil is attained. Thereafter, a reduced modulus is used for

each additional load increment.

Additional capabilities of BRACE II are: The retaining wall

is allowed to develop a plastic hinge when a specified yield moment is

exceeded; the soil behind the wall is allowed to slip unrestrained relative

to the sheeting. Both capabilities are important in the analysis of

braced excavations. In braced excavations in soft clay with large strut

spacings, the sheeting may become overstressed and large movements

will result. In cases where a concrete slurry wall is installed, a bentonite clay cake remains between the concrete and the soil. Since this

clay has essentially no shear strength, restraint of slippage between

the soil and concrete wall may be considered non-existant.

If, during a given excavation stage, the yield moment of the

sheeting is exceeded at a bar element node, that node is made a plastic

hinge for all additional incremental loads. Unrestrained siippage between

the soil and the sheeting is modelled by setting the axial stiffness of the

sheeting to zero.

The anisotropic variation of modulus of the soil is simulated

based on the methods proposed by Christian (1971).’ He recommends

the following appr oxirnate relation to account for the effect of stress

reorientation on the undrained .modulus:

E = E

6

uh- (Euh - Euv) cos4 0

where Eu h and E are the undrained modulus for tests with

the major principle s%\ss a

zontal directions P

plied to the soil in the vertical and horirespective y; 0 is the orientation of the major priciple stress from the verticl plane; and E

0

is the modulus used to

compute the deformations of the soil.

-46-
To account for anisotropic shear strength properties the

following yield criterion recommended by Davis and Christian (1971)

was used:

,cT,- S -s 2u v

2

uh ) + zxy2 &= a2

b2

\ 2

where:

a

u x 9 =y* Zxy

Su45

’UV’ ‘uh

conventional total stress components in the

x, y plane

shear strength of a soil sample oriented at

45 degrees from the vertical

shear strength for compression in the

vertical and horizontal directions

A. 23 Summary of Parametric Studies

The details of the parametric studies used to generate the

data in Chapter 2 and Chapter 7 are summarized in Tables 2 and 3.

The tables show the sheeting stiffness and corresponding soil

parameters used in each analysis. The soil parameters were selected

based on synthesized data presented by Ladd, et al (1971) and Ladd

and Vallaroy (1965 ). The soil parameters used in the analysis are

not exactly those reported by Ladd, et al (1971), rather they reflect

values which give stability conditions which would yield a broad

spectrum of excavation performance.

The finite element grid used to model this excavation is

shown in Figure 17. The soil mass was restrained against both vertical

and horizontal deformations along the bedrock base (I50 foot base) and

at a distance of 250 feet from the excavation face. Excavation stages

consisted of excavating approximately 3 feet below strut level. The

struts were instaLled without prestressing; but, once the strut was

installed, no further lateral movement wa’s permitted at the strut level,

-47-
Table 2. Summary of case studies for analysis of strut loads and sheeting deformations.

f

00

8

Fase

Depth Pene- EI Jg SOILS

Soil Wall of tration iTp

Profile Tme @ base

6,

cut (ft) (W @ base Type K, SW Suh %v %I

w-38 Soft

Uniform Steel 6 0 2 9 5. a3 6 . a 2 Uniform 0.5 0.2=, 0.15& 2OOZ’, 12%”

Pz-38 Medium

Uniform Steel 6 0 2 9 5 . a 3 4. 78 Uniform 0 . 6 0.406” 0.250, 2905” i aoq

26.5” Soft

Uniform Wall Diaphragm 6 0 2 9 38.40 6 . a 2 Uniform 0.5 o.za*, 0.156

v

2ooa; 1205”

26.5” Medium

Uniform Diaphragm 6 0 2 9 38.40 4. 78 Uniform 0. 6 0.406

Wall v

0.25q 290% 186”

48” Soft

Uniform Diaphragm 6 0 2 9 228.96 6 . a 2 Uniform 0.5

Wall

0.2aS-~ 0.15C” 2ooCv 1205”

48” Medium

Uniform Diaphragm 6 0 2 9 2 2 8.96 4. 78 Uniform 0. 6 0.405

Wall v

0.25e

”

29O?Q iaoCv

Fz-38 Stiff

Uniform Steel 6 0

Psf 1 0 Psf 5. a3 2. 25 Uniform Psf

1.0

Psf

3,200 3.200 3,000,000 3,ooo.ooo

Soft Fz-38 Soft

Over Steel 6 0 2 9 5. a3 2.25 0.5 800 540 800,000 500.000

Stiff Soft Fz-38 Stiff 1.0 3,200 3,200 3,000,000 2.500.000

Soft

Over Steel 6 0 1 0 5. a3 2.25 0.5 800 540 800,000 500,000

Stiff Stiff Soft 1.0 3,200 48” 3,200 3,000,000

Soft

2.500.000

Over Diaphragm 6 0 2 9 228.96 2.25 0.5 800 540 800,000 500,000

Stiff Stiff Wall PZ-39 Stiff 1.0 3,200 3,200 3,ooo.ooo

Stiff

2,500,000

, Over Steel 6 0 2 9 5. a3 6. 37 1.0 3.200 3.200 3,000,000 2.500.000

Soft Soft 0.5

Stiff

0.3(s; 0.25” 3608~

48”

216&J

Stiff

Over Diaphragm 6 0 2 9 228.96 6.37 1.0 3.200 3,200 3,000,000 2.500.000

Soft Wall Soft 0.5

Stiff

0. 3sv

Fz-38

0.2% 36@” 216%

Stiff

Over Steel 32. 5 17.5 1. 52 4. 30 1.0 2,000 2,000 2,400,000 2,400,000

Soft Soft 0.5 0.38” 0.2v

Stiff 48”

3607& 216q.

Stiff

Over Diaphragm 32. 5 17.5 59.60 4.30 1.0 2,000 2.000 2.400,000 2.400,000

Soft Wall Soft 0.5 0. 3Zv 0.2% 360& 21@v

_.

Notes: For all cases Ilsat = 120 pcf

GWT = 5.0 ft. below GL

Vertical strut spacing - 10’
Table 3. Summary of case studies for analysis

of zone of influence of deformations.

S

u h

Type

o f

Movement

Top tilt

yatxiri;m

a

Movement

6”

Rotation

about top

6”

Flexure

Top tilt

and flexure

6 ”

Rotation

and bulge

6”

Lateral shift

and flexu re

8”

Top tilt 6”

Rotation 6”

Flexure 4”

elastic elastic

1. 0 elastic elastic

1. 0 elastic elastic 3, 000,00(

Notes: For all cases:

lP= 0.01 psf

gwt = 150’ below G. L.

depth of cut = 60’

depth of penetration = 29’

Poisson’s ratio = )+I -3 hh- vh = 0.499

-49-
tiiiiiii i i

I

I I

I I 1 I

k

-5o-
CHAPTER 3 - BASIC CONCEPTS OF SOIL MECHANICS

3.10 GENERAL

This chapter briefly summarizes typical properties and stress

behavior of various soil types. Emphasis is placed upon basic concepts controlling behavior, particularly as related to the strength

proper ties . This discussion should not be construed as a substitute

for more comprehensive treatment given in soil mechanics texts, nor

should the data concerning typical soil parameters obviate the need

for soil testing. Rather, the data are given only as guidelines.

Soils may be classified under two broad categories, cohesive

and cohesionless soils. In the cohesive category are clays which have

low permeability and hence drain slowly. Most sands and gravels are

classified as cohesionless. Silts and cohesive sands are an inter -

mediate type whose engineering properties, especially strength, are

largely control.Led by their rate of drainage. Because of the general

relationship between permeabi,lity and plasticity, the latter is a useful

index for classification of these intermediate soils.

3.20 EFFECTIVE STRESS

Soil behavior is controlled largely by the effective stress in

the soil. Effective stress is defined as fo.llows:

is=&u

where: _

O-= effective stress

c= total stress

u = pore water pressure

Where @ is the vertical effective stress and ‘i is the horizontal

effective stresi, the in situ horizontal and vertical e4fective stresses

are related by the at-rest coefficient of stress, K , defined as the ratio

of the .initial horizontal to the vertical effective st%esses.

-51-
F

h

Ko= o

6 0

Soil strength parameters are governed by effective stresses

which make the evaluation of pore pressure an important consideration in any engineering analysis.

Changes in pore water pressure generated by shear strain

will dissipate immediately in cohesionless soil. Thus, the pore

water pressure is contro.lled simply by the depth below the hydrostatic water level.

With cohesive soil, pore water pressure dissipates very

slowly, in perhaps months or years. Moreover, one cannot accurately

predict changes in pore water pressure caused by shear strain. Asa corollary one cannot, with confidence, predict effective stress

conditions in cohesive soil before the start of construction. On the

other hand, pore water pressure measurements during construction

do provide a basis for computing effective stress.

3.30 SOIL UNIT WEIGHTS

The effective stresses in a soil mass depend upon the tota.

unit weight of the soil (&J, stress due to surcharge, and the pore

pressures within the soil mass (often controlled by the water table).

The unit weight of soil is a function of its specific gravity, void

ratio, and water content.

The fo.Llowing table gives a range of unit weights for various

soils for use in analysis.

Typical Values of Unit Weight

Moist Unit Weight Saturated Unit Wei&t*

Soil Type

(above water table)

Y: P

(below water table)

cf ksat, pcf

Poorly graded sands 105 - 115 115 - 125

Clean well graded sand 115 - 125 120 - 130

Silty or clayey sands 120 - 130 125 - 135

-52-
Soil Type

Moist Unit Weight Saturated Unit We#t*

(above water table) (below water table)

, pcf r sat, pcf I

Silty or clayey sands and

gravels

Soft to .medium clay

125 - 135 130 - 145

100 - 115 100 - 115

Stiff to very stiff clay 110 - 125 110 - 125

Organic silt or clay 90 - 100 90 - 100

“Submerged unit weights = Y sat- Y w = Y sub

3.40 SHEAR STRENGTH OF SOILS

3.41 General

Coulomb has defined soil shear strength in the general

form of the effective strength parameters r and 7, where 2 is the

cohesion intercept and? is the angle of internal friction as determined

in laboratory tests. The shear strength of a soil is:

Zff =E-k? ff tan. 3

where:

-c ff = shear stress on the failure plane at

failure

rff = effective norma. stress on the

failure plane at fai.lure

Figure 18 itlustrates the failure envelope.

3.42 Cohesionles s Soils

In cohesionless soils (sands and gravels) the cohesion

intercept, c, is zero. Therefore, the shear strength can be expressed

as:

r

ff = CT ff tan j?

Figure 19 illustrates the failure envelope for a typical.

cohesionless soil.

-53-
Figure 18. Failure envelope for soil,

general Mohr-Coulomb failure criterion.
P

ss3kLl.s w3l-6

-55-
As ,mentioned previously, sands and gravel.s are free draining

and therefore, pore pressure changes generated by loading or shear

strains dissipate quickly. Therefore, the shear strength of the soil

is related to the loading condition and the 2 value. For all practical

purposes, the pore pressure remains unchanged during the loading.

The value of the angle of internal friction is determined from

laboratory tests such as drained triaxial tests, direct shear tests, or

from empirical correlations. A crude guide for making an initial

estimate of 7 is the standard penetration test during sampling in the

bore hole. Peck, Hanson, and Thornburn (1974), for example, relate

ii values for sands and gravels to the standard penetration resistance

(N), as shown in the following table:

JJ vs N for Cohesion.less Soil

Relative Density 7

N:::

(blows /ft)

Loose 28O - 3o” 5 - 10

Medium 3o” - 36’ 10 - 30

Dense 36’ - 41° 30 - 50

Very dense 41° - 44Ot >50

>::

N, standard penetration resistance, is dependent on soil

density, gradation, drilling procedures, sampling procedures,

and depth below s-urface.

3.43 Cohesive Soils

Unlike cohesionless soils, which drain quickly, the

strength of cohesive soils is related to the changes in pore pressure

that occur during shear. The general expression for soil strength,

zff =Cks ff tan j?, is applicable to these soils; however, the

eva.luation of eff = 6

ff - u can only be made with knowledge of the

pore pressure, u. As stated previously, the current state-of-the-art

does not provide a sufficiently accurate means to predict pore

pressure, but it can be measured in situ.

Figure 20 illustrates the differences in the strength envelopes

for normally consolidated and over consolidated clay. Note that the

cohesion intercept (E) is observed in overconsolidated clays, but not

in normally consolidated clays.

-56-
‘At113 031v0170s

-NO3 AllVUWON UOJ

3d013hN3 HI9N3US XV13

lOSN03U3AO tlOj

031VO llOSN03

‘SU

B

‘3’0

-7!!
3.43.1 Undrained Strength

Cohesive soils are not normally permeable enough

to allow significant drainage during shear, and excess pore pressures

do not dissipate quickly. Therefore, to be on the safe side during

initial application of Load, the shear strength should be based on

undrained strength of the soil at natural water content. For most

loading conditions, normally consolidated soils will experience a pore

pressure increase (positive). Over consol.idated soils experience either

a smaller pore pressure increase or may even experience a pore

pressure decrease depending on the degree of overconsolidation.

When the change in pore pressure is positive, the

effective stress in the soil is lowered; therefore, the shear strength

is less than that which would be obtained using effective strength

parameters based on static water tab.le conditions. With ti.me,as pore

pressure dissipates, the effective stress, and therefore shear strength;

increases.

When pore pressure decreases as a result of shear

strain, as is characteristic of overconsolidated soils, the effective

stress increases. Therefore, the shear strength is greater than that

which would be computed using effective strength parameters based on

the static water table conditions. With tirne, the pore pressure

becomes increasingly positive (approaching the static water table

condition), the effective stress decreases, and the strength becomes

less.

Figure 21 shows two strength relationships. One

is a drained strength envelope in terms of effective stress. The other

is a plot of consolidation pressure vs. undrained strength. The plots

in the figure show cases where the undrained strength is greater than

the drained strength and, conversely, where it is Less.

Cohesive soils at their natural water content

exhibit a unique shear strength for al,L undrained loadings (and unloadings) regardless of the tota.L confining pressures. This value of shear

strength is referred to as the ,undrained shear strength, S , of the

soil. It is approximately equal to (and usually taken as) o&z half the

81 . $

compressive strength, 2 as determined in a compression test.

The undrained strength at natukal water content can also be determined

by a vane shear device, either in situ or in the laboratory.

-58-
I S3

9” SHEAR I STRENGTC

CURVE I

8

= EFFECTIVE NORMAL STRESS ON FAILURE PLAM w SHEAR STRENGTH

CURVE 2 = CON9oLlDATtON PRESSURE v s UNDRAINED SHEAR STRENOTH

FOR OVERCON~LIDATED CLAYS:

I

brolnodc %I-~**drained > Sundmkrsd

* ec = coNsocIDAfloN PRESSURE

L

c EFFECRT E NORMAL STRESS ON EWLURE PLANE (CURVE

& CDNSOLIDATION PRESSURE PRIOR TO SHEAR

Figure 21. Comparison of drained and undrained strengths.
3.43.2 Drained Strength

Ultimately the excess pore pressures generated by

shear strain in a cohesive soil will. drain, thus resulting in zero excess

pore pressures and a return to hydrostatic conditions. Drained and

undrained strengths present the limiting strength conditions for a

cohesive soil. Because one does not necessarily know what the pore

pressure conditions are, it is important to recognize which is the more

conservative for a particular case. At low confining pressures (with

respect to the degree of over-consolidation) the drained strengths control. On the other hand, the undrained strengths are more critical at

higher confining pressures. A comparison of undrained and drained

strengths is made in Figure 21.

3.43.3 Intermediate Cases

During drainage the strength lies between the

drained and undrained strengths and is a function of the pore pressure

dis sipation, i. e. it is related to the effective stresses at any given

instant.

3.43.4 Cohesive Soil Strength During Excavation

Consider the case of an element of cohesive soil

located at a depth inside an excavation. InitiaUy, the shear strength

of the soil wi1.l be equal to the in situ undrained shear strength, Su ,

Assuming an immediate unloading of the soil element (due to

excavation) , the soil element will experience a reduction in pore pr es -

sure, maintain a constant effective stress, and therefore, a constant

strength value (except for minor changes due to loading conditions).

However, as these negative excess pore pressures dissipate, the

strength of the soil will decrease until the ultimate drained conditions

are achieved.

This factor is particularly important for overconsoaidated cohesive soils with high undrained strengths. Accordingly,

upon initial excavation, the undrained shear strength at natural water

content should be assumed. But if the excavation is open sufficiently

long for excess pore pressure to dissipate, the drained strength should

be the basis for an analysis of this long term case. See Chapter 6 for

further discussion on the relationship to passive pressure.

-6o-
3.43.5 Consistency

Cohesive soils are described in terms of their

consistency, which in turn is directly related to their undrained strength,

Standard penetration resistance may also be used as a rough measure

of consistency for insensitive cohesive soils. See following table:

Consistency of N*

Cohesive Soil (blows /ft)

Undrained shear strength

psf

Very soft

Soft

Medium

Stiff

Very stiff

Hard

(2 < 250

2 - 4 250 - 500

4 - 8 500 - 1000

8 - 15 1000 - 2000

1 5 - 30 2000 - 4000

>30 ) 4000

N* - Standard penetration resistance

3.43.6 Heavily Over consolidated CLay

For heavily overconsolidated clays with fissures

and subject to strength deterioration with time, c should be set equal

to zero in the calculation of passive resistance.

B is also possible that in heavily over consolidated

clays, where 5 > r a passive failure may occur in the soil due to

the release of vhertic&‘soil pressure. These passive failures have

been observed as base heave in excavations in heavily overconsolidated clays. In these soils, a detailed analysis of the effective

stresses in the soil with excavation depth is required. Careful

analysis of the in situ stress system is required for all clays.

Further dis cussion relative to strengths of

highly over consolidated clays is made in Chapter 8 (Stability Analysis).

-61-
CHAPTER 4 - GROUND WATER IN OPEN CUTS

4.10 GROUND WATER CUTOFF

4.11 General

Cutoff walls are used for the following purposes:

1. To avoid or to minimize dewatering of the excavation.

2. To lessen or to prevent lowering of ground water

level outside of the excavation because of possible settlement and

damage to adjacent structures.

3. Because it is otherwise impractical to place lagging

in soils that are extremely difficult to dewater in advance of excava-

( such as silts and/or dilatant clayey sands).’

4. To cut off pervious water bearing strata within or just

below the bottom of the excavation: thus, protecting against the possibility of a “blow” condition or other source of ground loss.

4. 12 Soldier File Wall

A soldier pile and lagging wall is not watertight. In order

to control ground water; dewatering, grouting, or freezing must be

performed. While excavating in “running” soils it is essential to

maintain the ground water level below the working face to prevent inflow

and subsequent ground loss.

4. 13 Interlocked Sheeting

The permeability of interlocked steel sheeting has been

studied both experimentally and by observations of actual installations

The results of these studies are quite variable, but as an approximation the seepage through intact interlocks may be assumed to be in

the order of 0.01 gallons per minute per square foot of wall under a

1 foot differential head (Fruco & Associates, 1966), Sherard, et al

(1963) suggest that flow through intact steel sheet piling is equivalent

to about 30 to 40 feet of soil (probably relatively pervious granular

soils), under the same hydraulic head.
Clearly, the effectiveness of an interlocked steel sheet

pile cutoff wall depends upon the permeability of the soil in which the

sheeting is driven. If the steel sheet pile wall remains intact and

penetrates into an underlying impervious stratum, the effectiveness

will be very significant in pervious sands and gravel.. On the other

hand, in granular soils of low permeability, such as silty or clayey

sands, the mterlocked sheeting will have little effect on the relatively

low flow into the excavation. In all cases, however, sheeting effectively cuts off Bow in pervious layers that are interbedded within a

parent stratum of impervious soil.

With regard to maintaining ground water level outside of

the excavation, interlocked sheeting is efflective in pervious granular

soils. For relatively impervious soils (such as clayey sands, silts,

and clays) the sheet piling is essentially elquivalent to the permeability

of the soil, and therefore will have little olr no effect on the seepage

pattern toward the excavation or on loweri.ng of piezometric levels.

The above discussion applies only to intact sheeting. The

presence of boulders, difficult driving conditions, or obstructions

can lead to ripping of the sheeting and/or ,jumping out of interlocks

which will seriously impair if not destroy the effectiveness of the cutoff wall.

Another common problem is when the effectiveness of a

cutoff in pervious soil depends upon achieving a tight seal on rock.

This situation may be especially acute when rock occurs within the

depth of excavation. Settlement may resu’lt from ground loss due to

water inflow or from lowering of ground water levels and thus induce

consolidation of compressible soils.

In a ca.se in Boston (Lambe, et al, 1970)) the lowering of

piezometric head in a sand and gravel deposit below organic silt contributed to consolidation of the organic silt contributed to consolidation of the organics and settlement of the adjacent ground. Interlocked

sheet piling could not completely cut off water in a deposit of pervious

sand and gravel over bedrock near the bottom of the excavation. A

similar case was reported in Oslo (Hutchinson, 1964).

-63-
4.14 Concrete Diaphragm Walls

For all practical purposes, a well-constructed concrete

diaphragm wall is essentially impermeable, It will effectively cut off

flow and prevent ground water lowering outside the excavation provided that there is penetration into an underlying impervious formation.

Nevertheless, lowering of ground water may be caused by several poor

construction procedures, such as: (a) leaky joints,(b) water loss

through drill holes made for tieback installations, (c) inadequate seal

on bedrock, especially within the excavation.

4.20 SEEPAGE PATTERN TO EXCAVATION FACE

As mentioned previously, interlocked steel sheeting has relatively little influence on the seepage pattern in impervious soils. As

a result, when cuts are made below ground water there will be flow

to the face of the excavation. In clays, such a flow will be so small

that it may not even be noticeable.

An example of a flow net for this type of situation is shown in

Figure 22. During the initial process of excavation, deformation

in the soil will generate shear strains and cause pore pressure

changes. Eventually, these pore pressures will be dissipated, and a

steady state seepage pattern will develop as shown in the figure.

The equipotential lines shown in the figure demonstrate the

changes in hydrostatic stress. Such changes in hydrostatic stress

lead to a time-dependent equivalent change in effective stress and

consolidation of the soil, In precompressed cohesive soils, the

amount of consolidation will be negligible; however, in soft normally

consolidated clays or organic soils the associated amount of consolidation can be significant and will contribute to displacements behind

the exe avation .

The foregoing case is important because it illustrates that steel

sheeting may not be effective in preventing consolidation of normally

consolidated soils within depth of the cut. Soil compressibility and rate

of consolidation must be considered.

-64-
INTERLOCKED STEEL SHEETING

EQUI POTENTIAL

ROCK (IMPERVIOUS)

- INDICATES CHANGE IN TOTAL HEAD

FROM INITIAL CONDITION TO STEADY STATE

SEEPAGE CONDITION.

Figure 22. Change in pressure head for cut in impervious

soil.

-65-
4‘30 GROUND WATER RECHARGE

Lowering of ground water is accompanied by a corresponding

decrease in hydrostatic pressure and an increase in effective stress.

Such an increase in effective stress may possibly be accompanied by

consolidation of compressible soils and settlement of surrounding

buildings .

Compressibility of cohesive soils is time-dependent; compressibility of loose sand or non-plastic silt is immediate. Recharging of

ground water can be accomplished through trenches, pits, or wells,

in communication with pervious strata adjacent to the excavation.

Most commonly, recharge wells are used in conjunction with excavations made ,in urban situations. Examples of recharge wells to maintain the ground water level outside of the excavation and to prevent ,.

settlement have been reported by Parsons (1959) ,and by Ball (1962).

One of the most difficult technical considerations in developing

recharge wells is the problem of avoiding plugging of the well screen

and surrounding soil. Contamination may develop from suspended

particles, from corrosion, or by microorganisms which grow on the

well screen. In addition, the recirculated water contains dissolved

air which expands and plugs the soil pores after diffusion back into

the soils* thus reducing the permeability of the soil.

Measures taken to counteract such contamination are to filter

and chlorinate the water or to use cathodic protection to prevent

corrosion.

A final consideration is to prevent buildup of excessive hydrostatic pressure by diffusion near the recharge well. Such a condition,

especially in loose granular soils9 can lead to loss of effective stress

and settlement of adjacent foundations.

-66-
CHAPTER 5 - LATERAL EARTH PRESSURE

5.10 BASIC CONSIDERATIONS

The following discussion presents fundamentals concerning

magnitude and distribution of lateral earth pressure when influenced

solely by conditions within the depth of excavation. The influence

of surcharge loadings are covered in Section 5.40.

In the case of excavations underlain by weak strata, the lateral

pressure may be greatly increased as a result of shear deformations

generated by marginal safety against base instability. These situations

are addressed empirically in Section 5.22.

5.11 Earth Pressure at Rest

The ratio of the geostatic horizontal stress to vertical

stress of a natural soil formation is defined as:

“6

K =-$-

0

V

where:

K =

0

coefficient of earth pressure at rest

Qh = horizontal effective stress

av = vertical effective stress

For granular soils Terzaghi and Peck (1968) suggest K,

values of 0.5 for loose deposits and 0,4 for dense soils, Generally K,

can be estimated for normally loaded soil deposits as:

-67-
K = 1 - sin 6

where: 0

K. = coefficient of earth pressure at rest

3 = angle of internal friction in terms of

effective stress

For cohesive soils, K is primarily dependent on

the over consolidation ratio (OCR =povm &) as shown on Figure

23. Normally consolidated clays typlcatly have K values of 0.5

to 0.6; lightly over consolidated clays (OCRL 4) ha?e K values up

to 1; for heavily over consolidated clays (OCR-16) K. gay range

up to a value of 2.

5.12 Active Earth Pressures

5.12.1 Mobilization

Lateral displacement (as shown in Figure

24) transforms the state of stress in the ground from the at-rest

condition to the active condition. The mechanics of this process

are the mobilization of full shear-resistance within the soil massa state of stress referred to as “plastic equilibrium”.

In their state-of-the-art report given at

Madrid in 1972, Bjerrum, Frimann-Claussen, and Duncan summarized previous work concerning the amount of lateral displacement sufficient for mobilization of active earth pressure. They

reported that lateral displacement of 0.1 percent of the wall height

was sufficient to mobilize fully the active earth pressure of sands;

whereas full active earth pressure develops in soft clays with displacements of 0. L percent to 0.2 percent of the wall height.

5. 12, 2 Distribution

Figure 24 shows the active earth pressure

distribution associated with displacement modes. The fully

active state stems from lateral translation, by rotation about the

bottom, or a combination of both. The earth pressure distribution

is triangular and the resultant occurs at the third height of the

wall.

-68-
3 6

28

2.6

2.4

2.2

2.0

1.6

1.6

‘6

’ 1.4

‘6

:

y 1.2

1.0

0.8

0.6

0.4

SYMBOL RI.%

I O

---- 20

- - 30

6 0

. I I

2 4

Figure 23. K o versus OCR for soils of varying plasticity,

from Ladd (1968).

-69-
(a) FULLY ACTIVE

FULLY ACTIVE

.

\ 2

\

\

\

\

(b) ARCHING ACTIVE

Figure 24. Earth pressure distributions for active and arching

active conditions.

-7o-
For a rigid retaining wall, the arching

active case occurs by rotation about the top (Taylor, 1948). The

resulting earth pressure distribution is parabolic in shape: it

exceeds active near the top but is Less than active near the bottom

of the wall. On a more flexible cofferdam wall the ‘pressure distribution may be highly irregular due to the sequence of excavation

and bracing or variation in the tightness of braces, both of which

affect Load concentration.

5.12. 3 Coefficients

The direction and magnitude of active

pressure depends upon whether or not there is wall friction. The

particular case of horizontal surface and zero wall friction is the

Rankine fully active condition, shown in Figure 24a. For this

case, the active stress acts horizontally on a vertical wall. The

R.ankine coefficient of active earth pressure, K.a, is the ratio of the

effective stresses.

ch @a ifc = ii, = Ka

For sands, Ka = tan2 ( 45’ - J/2)

For cohesive soils,

General case ( 3, E):

Ka = tan2 (45O- j/2) - $-j tan(45’-J/2)

Special case (# = 0, c = Su):

Ka = 2sUIH

where:

K

a

= coefficient of active pressure

3, “c = friction angle and cohesion

intercept

-71-
vertical effective stress

r

h

= horizontal effective stress

$a = active earth pressure (horizontal)

S

U

= undrained shear strength (6 = 0 case)

According to the Rankine expression, the pressure

distribution for cohesive soils is theoretically in tension in the

upper part of the wall as shown on Figure 25a. Frequently, adhesion simply does not (or cannot) develop and therefore tension

cannot 0 ccur . However, the net total lateral force on the wall is

equivalent to that described by subtracting the “negative” pressure

at the top from the positive pressure at the bottom, Assuming

that this net force increases Linearly with depth of wall, it can be

represented by a net pressure diagram with a triangular distribution of the same force magnitude as shown on Figure 25b. The

ordinate at the base of the wall is:

za = XH-4S

U

This procedure was described by Terzaghi and Peck

(L968) as a means of comparing measured lateral forces with

computed forces acting on braced cofferdam walls. White the

method is reasonable for short term conditions, it is probably unrealistic to assume that undrained strength is mobilized over Long

periods of time. Clearly, such an approach is unconservative

with very stiff or hard clays. . In such cases and given the time for

dissipation of pore pressure generated by shear strain, one should

examine pressures based upon the effective strength angle, 3.

5. 20 INTERNALLY BR.ACED COFFERDAMS

5.21 General

Initially the internal bracing is set near or at the

top, thus restraining inward displacement. With each stage of

excavation and bracing there will be progressive inward displacement beLow previously placed braces. The net displacement profile typically takes the form shown in Figure 26 (after Bjerrum,

et al, 1972).

-72-
l-H

1

(a) RANKINE ACTIVE PRESSURE: DISTRIBUTION

IN COHESIVE SOILS

. ,

PA= ‘I$ ‘d H * - 2%~ H

*.

I-2s, j,,..dH-*Sud

I

4I-2$,1

<-\bH ‘bH_I

f b) TRIANGULAR PRESSURE DISTRIBUTION

_EQUIVALENT TO NET RANK:INE FORCE

(N =

pn

= 1/2 H*(%I-45,)

= $ xH*- 2Sui4

= 4 ‘11H*(lp+4S, -+

=rH(I-4Su)=8H(I-+)

3T

Figure 25. Earth pressure distribution for cohesive soil (9 = 0).

-73-
i

1’

/

,I

. -Y+ ‘, I

\

--iA\

I

,II

\

DEI

STR

DENSE

STRATUM

Figure 26. Mode of deformation of internally braced cofferdam

(after Bjerrum, et al, 1972).
Characteristically, there will always be some inward

rotation about the top, at least in the upper portion of the cut. The

degree of bulging and displacement beLow the cut depends upon

several factors - the distance between braces, the stiffness of the

wall, and the stiffness of soils near the base of the wall. In general,

the resulting deformation pattern most closely resembies the arching

active condition. Therefore,- a parabolic, rather than triangular,

pressure distribution is most Likely to act on the wall.

The Load concentrated on individual levels of lateral

support is greatly affected by the construction procedure itself,

envir onmenta L than ge s , and design considerations.

Several factors that can affect the load in a strut are:

For bracing:

a. The tightness and consistency of the contact between

struts, wales, and wall.

b. Whether or not prestressing was employed.

C . Temperature of braces during and following their instaLLation. (Bracing loads may be significantly affected by

temperature changes; especially in excavations which are

not decked over. )

d. Excavation distance between lavels of support. (The

relative consistency of Loads is directly affected by the

variation in vertical distance between Levels of support. For

example, if the upper 2 wale levels were spaced 10 feet

apart and the third wale level were 15 to 20 feet below the

second, the load on level 2 would-be relatively higher.)

For bracing or tiebacks:

e. Weak soils below depth of excavation. (For example,

soft clay underlying the bottom of a deep cut would cause

relatively high loads in the lower strut levels because of

Lack of passive resistance below the excavation base. This

occurs even though soils within the depth of cut are highly

competent. Another example-of increased load on bottom

struts would be an upward seepage gradient causing a loss of

-75-
passive resistance in front of the cofferdam wall. )

f. A stratum of weak cohesive soils within the depth of

excavation. (Such a condition may selectively increase

Loading on a particular wale Level unfavorably positioned

in relation to the bracing sequence and the weak stratum. )

1 thrust on wall.

g. Concentrated construction surcharge.

h. Frost action causing additional Latera

i. Erratic ground wat er conditions - per

perched zones combined with seepage.

haps Locally

Because of the n,umber of variables affecting the distribution

of Load on ground support walls, design procedures are Largely dependent upon empirical studies and correlations. The design procedures summarized in the two state-of-the-art reports by Peck

(1969) and Bjerrum, Frimann..Clausen, and Duncan (1972) are

based upon data primarily from internally braced, relatively flexible

walls.

Figure 27 shows the conventional procedure for analyzing

empirical load data. The approach has been to develop an apparent

earth pressure diagram by distributing the maximum measured

strut Loads during construction over an area described as midway

between the upper and Lower adjacent spans of the particular strut

Load measurement. The resulting apparent earth pressure diagrams

are used to develop an envelope encompassing the maximum distributed pressures. This design envelope then represents the maximum

strut Load that can be anticipated at any stage of construction.

5.22 Design Earth Pressure Diapra,m

Apparent earth pressure diagrams suggested by Terzaghi

and Peck (1968) for design of braced walls are shown on Figure 28.

Strut Loads for a given Level are determined by reversing the procedure

used for development of the diagram. A strut is designed to support

a load described by the area between the midpoints of the adjacent

upper and Lower support Levels.

The following discussion does not incLude the effect

of surcharge (see Section 5.40).
RA

RB

RC

RD

f

A

LI

8

L2

C

‘3

D Lb- L4

. . .--*-

, .

. .

I .

REACTIONS

%*%FE l %

L24II*-

a

APPARENT EARTH

PRESSURES

P =-RA * etc.

A LA

Figure 27. Conventional procedure for development of

earth pressure diagram,

-77-
PRESSURE DISTRIBUTION TOTALFORCE

a) Sands

\

F!if----

KA = tan2(45 - $/2)

\

H

Rankine Active

\,

t\

-0.65 KAyH * ’

/- KA yH 4

pt = Trapezoid

PA= Rankine

Pt = . 65 KAyH2

PA= . 50 KAyH2

Pt

PA

= 1.30

f

b) Soft to Medium Clays m= 1.0

0.25H

c

(N> 6)

For clays, base the selection

pt = . 875yH2(1 - i)

\

\

on N = yH

%

PA= . 50yH2(1 - i)

0. 75H \

\

/Equivalent Rankine Active

KA = 1 -m&=1 4

Pt

- = 1.75

- -

YH N pA

\ v

--c l.OKAyH- m= l.Oexcept

where cut is underlain by deep

soft normally consolidated clay

c) Stiff-Clays 2

Pt = . 15yH to . 30yH2

For NK 4 pA-=4, p =o

0. 50H For 44 N1. 6, use the larger N A

of diagrams (b) and (c). NK 4, PAL 0

NOTE: Equivalent

Rankine Active = 0

I--

0.4yHFigure 28. Design earth pressure diagram for internally braced

flexible walls (sands, soft to medium clays, stiff

fissured clays), from Terzaghi and Peck (1968).

-78”
5.22.1 Sands

This diagram, which was developed from dewatered

sites applies to cohesionless soils. If the soils outside the excavations

remain submerged, then the earth pressure should be computed

using the bouyant unit weight of the soil. Hydrostatic pressures are

treated separately and added to the effect of the earth pressure.

5.22.2 Soft to Medium Clays

The recommended earth pressure diagram for these

soils is shown in Figure 28b. The seLection of an appropriate

design diagram is dependent upon the stability number, (N= lH/S

The earth pressure computation for clays is based upon the totalU

).

weight of soil, assuming undrained behavior. This follows from the

fact that the data were empirically developed on the basis of total

unit weights and the soils’ initial shear ‘strength.

Where the stability number (N) exceeds 5 or 6, shear

deformation becomes significant. Note, by inspection of the empirical diagram for sands and for soft clays, that the latter is

significantly greater than the equivalent Rankine pressure which is

shown for comparison. The value of ‘ml used in the determination, of

the ordinate for earth pressure applies to situations where the cut

is underlain by a deep deposit of soft clay. Its value can only be

determined by empirical means from measurements and performance

of an actual excavation. Experience thus far, reported by Peck

(1969) from cases in Mexico City and Oslo, Norway, lead to the

conclusion that the value of lrnl is in the order of 0.4 for sensitive

clays. For insensitive clays the value of ‘m’ may be taken as 1.0.

5.22. 3 Stiff Clays

The recommended apparent earth pressure diagram

for stiff clays is used when the stability number, N, is less than

4. This empirical diagram is independent of the value of shear strength,

rather the lateral earth pressure is a function of the gravity forces

only. Strains associated with excavations for cut and cover tunneling

in these relatively strong soils are small, and the shear strength

of the soil is only partially mobilized. However, the movement is

sufficient to drop the lateral earth pressure below the K. values

(Gould, 1970).

-79-
5.22.4 Heavily Overconsolidated Very Stiff Fissured Clays

Several cases have been reported which suggest

that stress relief from excavation leads to lateral deformation of these

soils toward the excavation. The mechanism probably includes

both elastic strain and volumetric expansion. The elastic deformations occur during the excavation process whereas the swelling

is time dependent and is a result of the development of negative

pore pressures caused by stress relief. Soil behavior would suggest that the deformations should increase with increasing overconsolidation ratio, increasing plasticity of clay, depth below

the water table, and intensity of fissuring in the soi.1. For

strutted excavations, this condition may lead to build up of strut

load with time.

Pending the reporting of more field experience,

design criteria for cases involving potentially, laterally expansive soils

are as yet undeveloped. Therefore, a laboratory test program

(possibly stress-path triaxial) should be undertaken to aid in

evaluating the magnitude of the problem. Prototype test sections with

construction monitoring are also recommended.

Recent papers concerning measurements of loads

in dense cohesive sands and sandy clays have been reported by

Armento (1972), Liu and Dugan (1972), O’Rourke and Cording (1974)

and by Chap.man, et al (1972). Several cases involved cohesionless

soils (either fill or natural deposits of sand) within the upper portions

of the excavations. Others included interbedded strata of stiff

clay. In all cases, the soils near or below the bottom of the excavation were extremely dense and highly over consolidated.

The cases reported that the following factors affect

the load distribution:

a . Cohesionless soils within the upper portion of

the cut.

b. The construction procedure.

C . The depth of the excavation and bracing sequence.

-8O-
Chapman, et al (1972) report that for a number of cases

studied in Washington, the ordinate of the apparent pressure diagram

increased from 0.15gH for 30-foot cuts to 0.23 8 H for 60-foot deep

cuts. The attribute the low pressures at shallow depth to the relative

importance of soil cohesion. O”Rourke and Cording (1974) in their report

to the Washington Metro also noted an increase in the ordinate of the

apparent pressure diagram with depth.

R.ecommended design diagrams for dense cohesive sands and

very stiff sandy clays are shown in Figure 29. The minimum pressure

line is associated with cuts having reasonab1.y consistent spacing between

wale levels, relatively uniform soil conditions, and depths less than

about 30 or 40 feet. The maximum pr e s sure line is recommended to

cover uncertainty regarding the effect of weak strata within the depth of

cuts, the contingencies arising from construction (for example, overexcavation below support level or ineffective toe berms), and cuts in

excess of 60 or 70 feet deep.

Cohesive soils near the top of the cut will justify pressure

reduction as shown in Figure 29a. Absence of cohesive soils near the

top of cut will require the higher pressures associated with Figure 29b.

5.22.6 Stratified Soils

The aforementioned cases are for readily idealized soil

profiles. Actual soil conditions may have a stratigraphy which does

not conveniently match these simplified cases. Moreover , an irregular

ground surface or surcharge may complicate the analysis.

Under such cir cumstances, one approach is to determine the

Lateral thrust either on the basis of classic active earth pressure or on

the basis of trial planar sliding surfaces and wedge stability analysis.

In this latter case the most critical wedge is used to determine the

Lateral thrust (see Chapter 8). In such cases, hydrostatic forces are

treated separately.

Once the lateral thrust is determined,, it should be increased

by the most appropriate value of P,/P (ratio of force from the empirical diagram to the force determined rom the analysis from active F

earth pressure or wedge equilibrium). The designer must choose the

most appropriate ratio based upon a comparison of the actual case to

one of the simplified cases presented in this section.

The final question is one of pressure Idistribution. Again, at

least initially, this is a question of the designer’s jtldgement by com-

-81-
T

H

1

r

H

i

(a) RFI ,ATlVE;Lv UNIFORM CONDITIONS

P? =0.112\6.H2 TO 0.188&H2

0.80H 10.20t-l

(b) UPPER THIRD OF CUT DOMINATED BY

COHESIONLESS SOIL

*T Pt 00.135 #H2 TO 0.225x H2

Figure 29. Proposed predsure diagram for internally braced

flexible walls (dense cohesive sands, very stiff sandy clays).

-82-
parison with the simplified cases. Serious questions may need field

measurements to provide data input during construction.

5. 30 TIEBACKS

5. 31 Background

Many practitioners have successfully applied the empirical rules developed for internally braced walls to tiebacks; others

make variations for tied-back installations. In any event, at the

present time there are no empirical methods for tied-back walls that

have been accepted as universally as Pe:ck’s rules for internally

braced flexible walls.

In a series of model tests, Hanna and MataLLana(l970)

studied the effect of prestressing a wall to different design pressures

and distributions. They observed with excavation that triangular

distributions tended to redistribute Load to an apparent trapezoidal

distribution. In addition, when ties were prestressed to loads corresponding to a trapezoida L distribution, there was Less Load redistribution and the movements were less than for the cases with triangular

distributions. One problem with Hanna and Matallana’s work was the

Location of the ties. They were connected to a rigid wall of the experimental setup rather than embedded in the soil mass.

Apparently the tieback prestress has a significant effect

upon the pressure distribution. Clough (1972) found, after studying

sever al tied-back cases, that the pressure distribution suggested a

parabolic shape; moreover, this was borne o,ut by finite element

analyses.

5. 32 Comparison Between Bracing and Tiebacks

Tied-back installations differ from internal bracing in

their deformation mode, in the mechanics of stress conditions in the

soil, and in various construction aspects.

a. Deformations associated with tiebacks and bracing are

discussed in Chapter 2. Internally braced walls are restrained at the

top and tend to move inward with depth by rotation about the top, whereas

tied-back walls are more free to move inward at the top. Thus, the deformation mode often develops as inward rotation about the bottom. This

latter mode is theoretically compatible with the ltfulLyt’ active state and

linear pressure increase with depth.

-83-
b. Temperature: Bracing loads may be significantly

affected by temperature increase, especially by direct sunlight - to

the extent that some projects require remedial measures to reduce

thermal effects. Tiebacks are not subject to severe temperature changes

since they are insulated in the ground. Thermal effects are more pronounced in prestressed struts when the strut is essentially between two

unyielding supports.

c. Preload: Tiebacks are typicaLLy Locked off at 75

percent or more of the design Load. Observations suggest that tiebacks

will either maintain their load or experience a slight load Loss with time. On

the other hand, struts are generally preloaded to 50 percent or Less

of the design Load and will gain Load as the excavation proceeds. (A

greater preload in struts may risk excessive Load, especially from temperature rise. ) These observations suggest that prescribed preloads

for tiebacks are greater than the earth pressure wishes to impose. In

effect, the tieback Lockoff Load predetermines earth pressure rather

than vice versa. I

d. Mechanics: Tiebacks do not act by themselves, but in

consort with the earth mass within which they are embedded. This

behavior tends to dampen out Local variations in a given soil stratum

and thus Leads to more uniform loading on the wall.

e. Load Variation: Overall, Load variation with tiebacks

is Less than wTth bracing. Production testing of each tie above design

Load, locking-off at 75 percent or greater of design Load, insulating from

temperature effects, and engaging of an earth mass between the wall --

all tend to Lessen the variation in Load between individual tiebacks.

5. 33 Recommendations for Tiebacks

The following discussion does not include the effect of

surcharge (see Section 5.40)

Because of the reasons cited above, the Load variation on

individual tiebacks is believed to be Less than that on internal bracing.

Thus, it follows that the design pressure envelope for tiebacks need

not be as conservative as that for internal bracing. This does not mean

that the resultant horizontal force on a section is Less with tiebacks.

The actual resultant Lateral force must be differentiated from the

empirical design envelope which is greater ‘because it assures that no

one tieback level is overstressed. Paradoxically, if one were to compare the actual forces on a given cross- section, the force on conventionally

-84-
installed tiebacks would probably exceed the force on a braced wall.

This is because of the .prestressing of tiebacks.

Only Limited documentation is available to quantify conclusions concerning the relative magnitudes of appropriate pressure

envelopes for tiebacks and bracing. Accordingly, there is no present

justification for significantly changing the pressure diagrams for tied-back

walls from those used for internally braced walls. The following recommendations for tied-back walls yield similar total forces, but the pressures

are distributed somewhat differently than for internally braced walls.

The soil classifications are the same as for internaLly

braced walls shown in Figures 28 and 29 namely; sands, soft to

medium clays, stiff clays, and dense cohesive sands or very

stiff sandy clays. A triangular pressure distribution, increasing

Linearly with depth, is recommended for soft to medium clay; a

uniform pressure distribution is recommended for all other cases.

a . Sands: Where deformations are critical, and it is intended to prestress to LOO percent of design Load, compute force using

K*- For dense sands K-= 0.4; for Loose sands K = 0.5. Thus, the

ur?iform ordinate will vayy from: 0

Uniform Pressure, p q

Force, Pt= 0.20 $H2 to

Where &formations are not critical,

0.20 $H to 0.25 $H

0.25 g-l2

useK =

K. +Ka

2Vg 2

, that is

a coefficient midway between active and at rest. A similar procedure

was used by Hanna and MataLLana (1970).

Typical range is:

Loose sand:

Ka = 0.33; K. = 0.50

K = 0.42

avg

Force, Pt = L/2 x 0.42 XH2 = 0.21 1(H2

Uniform Pressure, p = 0.21 #H

-85-
Dense sand:

Ka

= 0.24; Ko = 0.40

K = 0.32

avg

Force, Pt = L/2 x 0.32 YH2 = 0.16 YH2

Uniform Pressure, p = 0.16 1(H

b. Stiff to Very Stiff Clays: Use a uniform pressure

ordinate of 0.15x G to 0.30aH to produce the same force magnitude

as that for braced excavations. The higher value is associated with a

stability number of about 4. The lower nu,mber is associated with very

stiff clays where the stability number is less than 4. The force varies

as follows:

Stiff Clays, Pt = 0. 30 $H2

Very stiff clays, Pt = 0.15 yH2

c. Cohesive Sand, Very Stiff Sandy Clays: Compute the total

force associated with the diagram for braced excavations (Figure 28) and

distribute the force uniformly with depth. For relatively uniform conditions

use:

Force, Pt = 0.112 $H2 to 0.188 $H2

Uniform Pressure, p = 0.112 YH to 0. L88 yH

Where the upper third of the cut is dominated by cohesionless soi .L use:

Force, Pt = 0.135 rH2 to 0.225 NH2

Uniform Pressure, p = 0.135 XH to 0.225 ‘$H

d. Soft CLays: It is unlikely that tiebacks would be used

unless they could be embedded in an underlying denser stratum of soil

or in rock. The walls should be designed with a trianguLar earth pressure

diagram assuming at-rest conditions and a K, value between 0.5 and 0.6.

Force, Pt = 0.25 yH2 to 0.30 YH2

In normally consolidated sensitive clays, excessive prestressing should

be avoided because of the potential for induced consolidation (see

McRostie, et al, 1972).

-86 -
e. Stratified Soils: As with braced excavations an approach

based upon active earth pressure or wedge equilibrium should be investigated. Section 5.22.6 generally describes a procedure for increasing

the computed force in the same proportion as that of the most closely related simplified soil profile that exceeds active earth pressure. Use

the force distribution most closely related to the simplified case.

5. 34 Effect of Wall Stiffness on Load Distribution

Theoretical analyses of the effects of wall stiffness on tieback loads (Egger, 1972; Clough and Tsui, 1974) and model tests (James

and Jack, 1974) indicate that wall stiffness does affect anchor and wall

load distribution.

Finite element analyses have shown that a more uniform load

distribution occurs for a’ stiff wall than for a flexible wall. In the

more flexible walls the pressure distribution concentrates at the wale level

due to arching. The difference between the load distribution for stiff

versus flexible walls is greater with increased spacing between the wale

levels. Clough and Tsui (1974) suggest that for typical spacing of

tiebacks there is a relatively minor load distribution difference for

the different wall types, Therefore, there appears to be no present

justkfication for drawing a distinction in pressur.e distribution on the

basis of wall stiffness.

5.40 SURCHARGE LOADING

5.4L General Background

Surcharge near excavations may be the result of many different types of loading conditions including footing6 structures, storage

of construction materials, or traffic. The Lateral pressure caused by

a surcharge Load on a retaining wall has been investigated for a variety

of different loading and soil conditions (Spangler, 1940; Newmark, 1942;

Ter zaghi., 1954b). This pressure is in addition to the normal earth and

water pressure.

5. 42 Theoretical Cansiderations

The four basic Loading conditions for which soLutions,.df the

lateral stresses in an elastic medium are readily available are:

I. Point loading

2. Uniform line Loading

-87-
3. Irregular area Loading

4. Uniform area Loading

Typically, the stresses within a soil mass due to surcharge

loadings are computed on the basis of elastic half-space theory;’ When

the wall is represented as a rigid boundary, it is necessary,to double

all stresses obtained from half-space theory at the face of the wall in

order to maintain compatible boundary conditions.

When lateral strains of the same magnitude as those in ~

an elastic half-space do occur , it is not necessary to double the stress.

In general, the true value of lateral pressure due to surcharges will

be somewhere between these two cases. Since the assumption of an unyielding rigid boundary is conservative , uniform application of this rule

should be questioned, and judgements made as to the appropriateness of

the assumption for a given job condition (Gould, 1975).

5.43 Practical Considerations

With regard to surcharge loading from construction operations, it is common to take a distributed surface surcharge on the order

of 300 psf to cover storage of construction materials and general equipmerit. Usually, this surcharge should be considered within a rather

limited work area on the order of 20 feet to 30 feet from the cofferdam

wall.

A second major consideration is the question of concentrated

loads from heavy equipment (concrete trucks, cranes, etc. ). Lateral

thrust from such equipment would be easily covered within the 300 psf

surcharge, provided that the equipment were more than approximately

20 feet from the wall. On the other hand, such equipment within a few

feet of the wall may create a concentrated surcharge loading which would

be of far greater significance than a uniform surcharge Loading. This

must be accounted for separately. It may necessitate the designation of

specific areas rather than designing the entire cofferdam for such Loading.

5.44 Point Load

While it is impossible to have a perfect point loading situation,

the computed stress for an area Load or a point load is essentially

the same when the distance to the wall is large compared to the size

loaded area. The difference is small if a point load is assumed when

the distance is greater than twice the average dimension of the Loaded

area. There are several practicaL cases for which point loadings may

-88-
be as sumed. An isolated footing for a struciture or a heavy object

resting on a small base may be cases that can best be analyzed as

point Loadings a

In the work of Spangler and Gerber, summarized by

Terzaghi (1954b), it was shown that there is little change in the magnitude or distribution of Lateral stress from that determined by elastic

theory untiL the point Load is located at a distance x Less than 0.4 H from

the wall. This Leads to the following equations for’evaluating the effects

of a point Load on a rigid wall.

For ,m<O. 4:

0.28 n2 Q

crh =

P

(0. 16 t nL) ’ H2

where:

(rh = horizontal stress at a depth, z = nH

Qp = magnitude of point Load

H = height of cut

Z

n =- =

depth to point on wall

H depth of cut

m =X =

distance of point Load from waLL

For ti0.4:

H depth of cut

m n

1.17 rn’ n’ Q

ah = P

(m2 t n2)3 H2

Figure 30a presents solutions to these equations for selected values of

m. The equation for m<O.4 has been derived from measured lateral

pressures and does not correspond to results from elastic theory. For

m>O.4, the equation gives values twice that of elastic theory to account

for the wall as a non-yielding reflective boundary.

Figure 3Oc shows how the Lateral stress for a point Load varies along

the Length of the wall. The calculation of uh gives the horizontal stress

on a vertical plane lying perpendicular to the wall and through the point

Load. The horizontal stresses along the wall vary asbh’ = bh cos 2 (1. L0).

-89-
0.2

I

0.4

\

hi

II

t

0.6

w

3

s

0.8

VALUE OF O-h (8)

Q P

(a)

o;, co& 1.18)

X=mH

k) SECTION a-a

FOR m I 0.4:

(b)

FOR rnm.4;

H2

uh(q”

l.77m2n2

(rn+ d3

NOTE:

mm-&

n=+-

Figure 30. Lateral stresses on the face Iof an unyielding

wall from a point loading (NA.VFAC, 1971

and Terzaghi, 1954b).

-9o-
5.45 Line Load

Terzaghi (1954b) also synthesized previous works on the

effect of line load on rigid walls. Figure 31 presents simplified

equations for,evaluating horizontal stresses for a line load which is

parallel to the direction of the wall and Located a distance, x, from

the wall. In practice, such a situation may arise if a continuous strip

footing runs parallel to an excavation. A.gain, these equations are for

a non-yielding wall.

As for the case of the point loading, where m<O. 4, the

horizontal pressures predicted by elastic theory are too high. Hence,

the equations given have been modified to correspond to measured

lateral pressures. Where the load is located a distance less than

0.4 H from the wall, there is little change in the magnitude and distribution of lateral pressures from that computed at a distance x equal.

to 0.4 H (see Figure 31). The variation in the location of the resultant,

ph’ is also small until x>O. 4H. The equations presented in this section have been adjusted to represent the boundary conditions of a rigid

wall.

5.46 Irregular Area Loading

In some instances it may be unrealistic to assume a surcharge loading of infinite extent behind a wall. Theoretical solutions

for area loading of limited (and irregular) dimensions have been developed for elastic half-spaces. Newmark (1942) presents an influence

chart for use in determining the horizontal stress on a vertical plane.

Although the chart was developed on the basis of Poisson’s ratio

Y = 0.5, Newmark (1942) does give a method of converting these

values to soils with other values of Poisson’s ratio. The values of

horizontal stress derived from Newmark’s (1942) chart are for an

elastic half-space. If the wall is assumed to be rigid, the values from

the chart should be doubled.

Figure 32 shows an influence chart for evaluating the

lateral stresses acting on a rigid wall from a rectangular loading

(Sandhu, 1974). These charts assume a Poisson’s ratio of 0. 5 for

the soil mass. Using the influence charts for point loadings, the

lateral stress due to an irregular surcharge loading can be more

easily calculated.

-91-
0

0.2

II

c 0.6

5

a

> 0.8

0.4I

\

.

h

0.6

VALUE OFch (#

LINE LOAD QL (a) -L

FOR m 5 0.4:

I

z= nH

.43+

P

t-l .r

ph = 0.55 QL

FOR In> 0.4:

Figure 31. Lateral stresses acting on an unyielding

wall from a uniform line loading

(NAVFAC, 1971 and Terzaghi, 1954b).

-92-
m=-B

z ’

n:+-

, P’q x Ip

dq r0.S

q =SUF?cHAf?GE

L = LENGTH PARALLEL TO WALL

-

B = LgNgFL PERPENDICUL

0.4[ - . . - , . . . . , . . . .

0.3

0.2

0.1

0

,AR

Figure 32. Lateral stresses on an unyielding wall due

to irregular surface loading (Sandhu, 1974).

-93-
5.47 Uniform Area Loading

Intensity and distribution of Loading was discussed in

Section 5.43. One approach is to treat the surcharge as a stress in

an elastic medium. The solution for lateral stresses on a rigid wall

are presented in Figure 32. An example of the stress effect with depth

is shown in Figure 33. Note that the stress influence below a depth of 1

about 1. 5B is negligible.

A second approach is to apply an earth pressure coefficient,

K, to the surcharge loading and to consider the surcharge effective

within some portion of the cut. The magnitude of this coefficient will

range from Ka (active earth pressure) to K. (earth pressure at rest).

In evaluating which of the above approaches to use, one

should first establish whether or not there are significant design implications between the various methods. If there are then one must apply

judgement concerning the relative rigidity of the wal.l (see Section 5.42).

Moreover if the surcharge exists during the excavation process, then

the appropriate coefficient is closer to K,, If the surcharge is applied

after excavation and bracing against a relatively unyielding wall, then

one should use K. or Figure 32.

-94-
0.08.

0.5B

rtJ

- I.OB

E

%

1.5E

INTENSITY OF LATERAL STRESS

vRIGID WALL

Figure 33. Lateral stress on rigid wall from surcharge of

width B arid infinitely long (solution from Sandhu, 1974).

-95-
CHAPTER 6 - PASSIVE RESISTANCE BE:LOW BASE OF EXCAVATION

6.10 GENERAL

The design process frequently requires that the soils below the

base of an excavation provide passive resistance for force equilibrium

or to limit movement. The performance of the wall will depend upon

the spacing of the support Levels since the greater the spacing, the

greater the required passive resistance (and movement) below the

Lowermost support level. Figure 34 illustrates the case of a wall in

which the passive resistance of the soil is insufficient to’limit excessive wall movements.

This section will describe the selection of soil parameters and

methods used to evaluate passive resistance. This section will not

consider the depths of penetration required to maintain overall stability of the earth mass or to limit displacements.

6.20 SOIL PARAMETERS

This section sum.marizes the soil properties relevant to the

calculation of the passive pressures. Chapter 3 of this volume

presented the basic concepts in more detail.

6.21 Granular Soil

Granular soils are free draining and cannot sustain

positive or negative pore pressures generated by strain or Load

changes for even a short period of time. Therefore, analyses of

the stability of granular soils is performed on the basis of drained

strength parameters and effective stresses in the ground referenced

to the static water level. The appropriate soil strength parameter

for the soil is the angle of internal friction, p, for the soil. For

design, granular soils are assumed to have no cohesive strength

component.

6.22 Cohesive Soil

Passive stress conditions occur with excavation below

the Last placed support Level. Because of the load decrease from

excavation, soils in the passive zone just below the excavation will

initially experience a pore pressure decrease. As a result, a

gradient is set up which causes water to flow into the voids of the

soil. This causes excess pore pressure to rise (i.e. become Less

-96-
Figure 34. Movement at wall base due to

insufficient passive resistance.

-97-
negative. ) This may be accompanied by heave caused by swelling

of the soil.

Limiting case strength parameters for passive

pressure computation are:

a. Immediate Condition: Pore pressures generated by

unloading and strain do not have time to dissipate. Moreover, pore

pressure cannot be reliably predicted. Use undrained strength of

soil at natural water content, S . Conventionally, this is determined’

from vane shear, unconfined cokpressiomf or unconsolidated

undrained compression tests.

b. Ultimate Condition: Pore pressures generated by

unloading and strain are dissipated by drainage. Effective stresses

can be computed on the basis of static water levels, Use strength

parameters from the effective stress envelope, E and j&

Specific cases obviously require soil testing and analysis in the light of soil properties, boundary conditions, and construction time. General recommendations for strength relationships

are to use undrained strength parameters for the “during excavation”

stage, that is, during the period of sequentially excavating and

installing braces or tiebacks. For the fixed depth conditions, pore

pressures will generally have sufficient time to dissipate, and

therefore, effective stress parameters will apply for this limiting

case condition. With in-situ pore pressure measurements during

construction, the passive force can be assessed in terms of effective

stress strength parameters based upon the computed effective stress

conditions.

As was pointed out in Chapter 3, the undrained strength

at natural water content may be greater than the drained strength of

over consolidated soils. Therefore, indis criminate use of. undrained

strength without regard for pore pressure dissipation may be on the

unsafe side.

Two factors that affect strength loss with pore pressure dissipation are the proximity of the soil element to the bottom

of the excavation and the amount of unloading. Typical.ly, the

drained strength of cohesive soils in the passive zone of deep cuts

will be the controlling strength.

-98-
6.30 ANALYSIS OF PASSIVE RESISTANCE

Several articles and texts address the problem of passive

pressures that can develop behind a continuous wall (Terzaghi, 1954b;

NAVFAC, 1971). In cohesionless soil wal.1 friction modifies both

the direction and magnitude of the passive resistance. Typica.lLy, the

resultant of the passive pressure acts at an angle $’ equa.l to l/2 to

Z/3 of the angle of internal friction. The following table (from

Terzaghi, 1954b) summarizes values of Kp for various values of

$ and 6 .

Values of Passive Earth Pressure Coefficient

as a Function of fi and&

6 g=o 4= $12 6 = 213 $

25O 2.46 3. 00 3.20

3o” 3.00 4.20 4. 80

35O 3.70 6.50

4o”

7. 30

4.60 9.20 11.00

The passive pressure for drained loading or in terms of

effective stress at depth, z, will be:

cP

= av tan2 (45’ - y/2) t 2C tan (45O t 3/2) Eq. 6.30.1

excess

be:

where :

d

P

= passive pressure (effective stress)

ev = vertica.1 effective stress = 82; - u

3 = angle of internal friction (effective stress envelope)

z’= cohesion intercept

For this drained condition, in which by definition there is no

pore pressure, the total lateral stress at any depth, z, will

0, = bp t x 2

W

Eq. 6. 30.2

where:

0

h

= lateral stress

8 w = unit weight of water

-99-
The passive resistance of cohesive soils in an undrained

condition should be evaluated on the basis of the undrained shear

strength, S

U

, and the in situ total vertical stress, 0

V'

For a continuous wall, the passive pressure at a given depth will equal:

tTp=crv + 2s

U

=v zt 2 s

U

Eq. 6.30.3

where:

6

= total vertical stress = YZ

Su = undrained shear strength of the soil

In this case, the water pressure is not added because pore

pressure effects are already accounted for in the determination of

undrained shear strength, Su. Therefore, the total lateral. stress

at any depth, z, will be:

oh= op= xzt2s

U Eq. 6.30.4

Soldier pile walls are not continuous walls, therefore the

passive earth pressure coefficients must be modified from those

used for continuous wa.Lls, Broms (1965) showed that the passive resistance of lateraLLy .Loaded piles based upon piLe width and on Kp va.Lues

for continuous walls was too conservative. His study showed that soil

ar thing and non-plane strain conditions increase the capacity of

individua.L pi.Le s . Indeed, the process is probably closely related to

lateral bearing capacity. Broms’ recommendations are given in the

charts shown in Figure 35. It should be noted that for cohesive soils

the lateral resistance of the soil should be neglected to a depth of

1.5 pile diameters. In cohesionless soils where the depth of penetration is greater than one pile diameter, soil arching causes an

effective increase of 3.0 in the value of K

P’

A factor of safety of 1.5 is recommended for use in passive

pressure calculations.

-LOO-
I

I --

/)

/

FILE

WIDTH’

b

T

E

l.Sb m-

*

L c

>

T

D

1

UNDRAMED STRENGTH

mmEswE)

38vbKp

-T-=b

,- -

0

1

EFFECTIVE STRESS

fARAMETERS

KDkESK#ULESS or CDHESIVE)

Figure 35. Passive pressure for soldier piles

(after Broms, 1965). (Modified. )

-lOI-
6.40 OVERCUT DESIGN DETAILS

Over-excavation below the required support level depth is

common either to obtain working room or to muck up the bottom,

During intermediate excavation phases, assume a minimum of two

feet of overcut before strut placement. At final depth, assume a

minimum of one foot of over cut.

6.50 BERMS

Lateral resistance of berms will, of course, be lower than

the case of a horizonta,l plate at the top elevation of the berm. One

method of analysis is by wedge or logarithmic spira.1 force equilibrium

of trial failure surfaces. Another procedure is to replace the berm

with an equivalent sloping plane and assign the appropriate passive

coefficient (Terzaghi and Peck, 1968; NAVFAC, 1971).

-102-
CHAPTER ‘7 - DESIGN ASPECTS OF LATERAL PRESSURE

The analysis of forces acting on a support wall, the related

sizing of members, and the determination of wall penetration below

the bottom of the excavation is related primarily to the wall stiffness

El

and the type of wall. The wall stiffness is related to the ratio 3.

For example, steel sheet piling and soldier pile walls with a typical

wale spacing of eight feet or more and generally greater

horizontal distance between support members are considered

to be “flexible” walls Design earth pressure diagrams should be

determined in accord&e with Chapter 5.

7. 10 LOAD ON SUPPORT LEVELS

Commonly,. wale loads are determined by area proportioning from

the pressure diagrams developed from field measurements. This

method for evaluating wale loads merely consists of reversing the

procedure for developing the apparent earth pressure diagrams shown in

Figure 28. This procedure is illustrated in Figure 36.

7. 20 ANALYSIS OF WALES AND SUPPORT WALLS

7. 21 General

Deflection of structural members supporting soil causes

arching of earth resulting in a reduction of pressure near the center of

spans and a concentration of pressure at the supports. Hence, the

actual bending moments in wall elements and wales is less than that

which would be computed assuming a uniform loading on these flexural

members.

Several approaches have been used to determine moments

in support members. Armento (19721, for example, used 80 percent of

the uniform apparent pressure and computed moments assuming

hinges at support levels. Peck, Hanson, and Thornburn (1974) propose using

using 2/3 of the apparent pressure and assuming continuity over

supports in computing moments.

The approach used herein, for moment computation in

wales and support walls, is to use 80 percent of the loading diagram.

For evaluation of loads in internal bracing and tiebacks, the full loading

-LO3-
* ; :‘;’

AL,/*

li I

12

,.: ‘, :.;.

. b

RA

R B

RC

RD

RE

‘k

1

2

L.3IL2

- --

bp = KifH ~-4

AREA A

AREA B

AREA C

AREA D

AREAE

STRUT LOAD PER LINEAL FOOT OF WALL IS EQUAL TO DESIGNATEDAREA

EXAMPLE: Rc=P ($‘++2 )

Figure 36, Load determination from apparent earth

pressure diagram.

-104-
diagram (100 percent) is used. This recommendation is linked to a

number of other associated design recommendations -- the pressure

diagram itself, methods for moment computation, preload practice,

allowable s tre s se s, etc.

Where rigid walls support the earth, such as diaphragm

EI

walls with r4 greater than 50 ksf/ft, arching will be minimal; therefore, structural design of the wall as well as other elements should

be based on the full pressure diagram,

7. 22 Continuous Members

The following expression should be used for computing moments over continuous members with uniformly applied loads:

M=Cw12

where:

M = moment

C = moment coefficient

w = distributed load on span

1 = span length

Hinged ends would have a coefficient, C = to. 125, with a

maximum positive moment in the center of the span, Fixity at each

support (no rotation) results in a maximum negative moment at the

support and a moment coefficient, C= 0. 087. Since construction

methods greatly influence the position of the elastic line of members

(especially vertical members), there is no practical way that the moment can be precisely analyzed. Therefore, a coefficient of C = 0. 10 is

recommended for continuous members supporting a uniform distributed

load.

7. 23 Discontinuous Wales

The moment in the wale will depend on the splice detail.

For splices which occur at a strut and tie the wale with a steel strap,

-105-
to transfer shear hut not moment, zero moment should be assumed

at that point.

Wales supporting uniform load with moment splices over

less than three spans, should not be considered continuous. Three

spans or more should be considered continuous using a moment coefficient, C = 0. 10.

The moment in wales supporting concentrated load (as

from soldier piles or tiebacks) should be calculated on the basis of

statics. Assume full continuity where moment splices are used;

assume zero moment in other splices.

7. 24 Member Connections

It is common to design splices for the full structural capacity of the member (both shear and moment). This is often done with

a combination of fully penetrating butt welds and cover plates.

Figures 37, 38, and 39 show some typical

details for splices and wale to strut connections. For splices that are

butt welded it is often assumed that the butt weld is only 50 percent

to 75 percent effective since the beveled edges at the splice are field

cut. Hence, the cover plates are designed to carry 25 percent to 50

percent of the member capacity. In designing a strut to wale connection,

stiffness must be provided to prevent web crippling. Also, if raked

struts are used, a knee brace is required at the strut to prevent buckling

of the wale from the vertical component of load.

7. 25 Lagging

Arching of soil to the soldier piles results in substantial

reduction of loads on lagging between soldier piles. This reduction

depends on soil type and construction procedure, and it is not possible

to predict by rational analysis. Therefore, the determination of lagging

size is largely based on the past experience of the construction industry.

The soldier pile section in Volume III (Construction Manual) summarizes

recommended lagging sizes versus soil type and excavation depth.

-106-
SOLDIER PILE

I

-I

GAS CUT ENDS OF SOLDIER PILE

FULLY PENETRATING BUTT

WELD

Figure 37. Typical splice with butt welding,
u

PLATF )r

+-@4

SPACER

fCONTACT

LAGGING

I

I

?

1

,

/

L

1

I

i

c

1

1

-- i

---_ --- --

$

/

/

/

/

/

/

Figure 38.

loCLEAR SPACE

IF NO BUTT

WELDING

arc------------- PLATE

SHEAR

- PLATE

/a- STIFFENER AT SOLDIER PILE

J

f PLATE IF STRUT ---m-L- ------e DEEPER THAN WALE

a_\

-STIFFENER

AT STRUT

PLan view of typical wale splice and strut connection.

-108-
SPACER

llccn “YLY WITU ..II”

:ONTACT LAGGING

ERECTION SUPPORT

AS REQUIRED

Figure 39. Typical strut -wale- soldier pile

connections (elevation view).

-109-
7.30 BRACING AND TIEBACKS

Bracing and tieback loads must be determined for the most

critical construction condition. This may be an intermediate depth

of cut or at full depth.

For bracing, the allowable axial loads are governed by the

,member’s slenderness ratio. Posting and lacing may be necessary 1

to cut down on unsupported length to provide economical bracing

members. Pipe sections may be utilized because of their efficiency

as column members. Wide flange sections with vertical webs are

also efficient, but this orientation may complicate wale connections.

For bracing:

a. At final depth, use allowable stresses by AISC Code.

b. For temporary conditions at intermediate depth of excavation use AISC t 20 percent.

For tiebacks, use the stress values stated in Volume III,

Chapter 6 (Tiebacks).

7. 40 DEPTH OF PENETRATION BELOW CUT

7. 41 Lateral Resistance

When design pressure diagrams are used, a reaction

at the base of the cut is assumed to ,exist which is equal to the

lowest area shown in Figure 36. This reaction is provided by the

passive resistance of the soil beneath the cut. The magnitude of the

passive resistance is analyzed for continuous walls using the modified

Coulomb earth pressure coefficients given in Section 6. 30. Passive

resistance for soldier piles reflects an added resistance due to soil

arching as explained in Chapter 6.

Figure 40 illustrates the method for determining the depth of

penetration in competent soils that are capable of developing adequate

passive resistance. Soils satisfying this condition are medium-dense

to dense granular soils and stiff to hard clays. The general method of

analysis is:

a. Compute the equivalent reaction at the base of the cut (hE).

-llO-
Pt (FROM DESIGN EARTH PRESSURE

FlGURE 28)

BOTTOM OFCUT I

\

\

V * SHEAR AT TIP OF SHEETING *O

Kg8 COULOMB EARTH PASSIVE cmEFFlCIENT

(SEE SECTION 6.30 1

1 . Compute RE = 0.5 pt Ld e

2. Compute depth x such that: P = RE + PA

P K

Use minimum F. S.

. = 1. 5 for passivedoofticient, K’ = fi

P l

3. Check Mmax 5 yield moment of sheeting

4. Drive to depth D = 1.2x

c

Figure 40. Procedure for determining depth of penetration

in relatively uniform competent soil conditions.

-lll-
b. Determine the depth required to satisfy force equilibrium on the horizontal plane.

C . Check the maximum moment at or below RD against

overstressing o.f the support wall.

d. Drive sheeting to a depth 20 percent greater than that

required for force equilibrium.

In cases where the soils below the base of the cut are very

loose to loose granular soils or soft to medium clays, the sheeting

should penetrate to only a minimal depth of approximately 20 percent

of the excavation depth and be designed as a cantilever below RD.

The reason is that in loose granular soils the sheeting must experience large lateral deformation befo,re building significant passive

resistance. It is probable the sheeting will be overstressed before

this deformation is attained. Hence, the sheeting will either act as

a cantilever, or if it is driven to great depths it will act as a simple

beam with a substantial span. In either case, the sheeting would

most likely be overstressed at the lowest strut level. If the cut is

underlain by soft to medium clays, the net pressure on the sheeting

often is in the active state, hence there is theoretically a net active

pressure. This situation can arise in deep cuts even when the base

of the cut is stable as illustrated in Figure 41.

7.42 Bearing Capacity Considerations

Load capacity must be evaluated when there is a downward

component of load, as is the case for inclined tiebacks. This may be

accomplished using pile driving formulas or by the’empirical

and semi-empirical methods outlined in Chapter 9.

7.50 EXAMPLE PROBLEMS

7. 51 Introduction

Three example problems are analyzed to illustrate methods

of evaluating the depth of penetration required for sheeting stability

and to show the effect soil stratification can have on the variation in

strut load. The se example s, shown at the end of this chapter, consider the following three conditions:

-1l.2-
SAND

-

BOTTOM OF

Yf!I”v”

su 8 + D 0.20ifH

RANKINE MZTIVE = 8 H -- 2( .20rH ) =

O.t3O~H AT L = H

RANKINE PASSIVE

AT t’H=O t 2(.201(H) = 0.4&d

1 . Theoretical passive resistance is not available

below bottom of cut to develop horizontal reaction.

In fact, the net force below cut is theoretically

toward excavation, based on active and passive

pressure.

2. Use nominal penetration of 0. 2 H or 5 feet whichever

is greater,or penetration to cut off pervious layers.

3. Check base stability (see Chapter 6).

4. Design for cantilever condition below E.

Figure 41. Method for analyzing sheeting with

weak under Lying layer.

-113-
Case I. Homogeneous soil profile.

-

Case II. “Soft” soil stratum to base of excavation underlain

by a dense stratum.

Case III. A soft layer underlying a more competent soil.

Above the base of the excavation an empirical design ’

pressure diagram is used as shown in Chapter 5. Below the base

level Rankine active and passive pressures are used.

It is recognized that when using Peck’s design envelopes

the largest strut load for any condition is taken into consideration,

However, a review of the development shows that in a few cases,

strut loads gave apparent pressures which were greater than the

de sign envelope s. The intent of this exercise is to illustrate a

means by which a designer may estimate the magnitude of a strut load

for unique conditions as well as providing a basis for judging whether or

not to increase the design load on a given strut over that predicted

by the design envelope. In addition, the analytical approaches will

aid in the evaluation and understanding of observed strut loads obtained from instrumentation programs.

i’. 52 Results of Analysis

Case I is the analysis of a homogeneous soil profile which provides abasis for comparison of required penetration depth and strut

load variations. It represents, most ideally, the conditions where

the design envelope is appropriate. The method for analyzing

soldier piles set in concrete-filled, pre-augered holes is also presented.

Case II analyzes the effect of a weak soil overlying a more

competent one. It illustrates how the load in the second lowest

strut can exceed that of the lowest strut.

As the excavation proceeds below level D to level E,

little passive resistance is provided to the retaining wall; hence,

the wall deflects inward. Effectively, the wall spans from level D

to the excavation base with full active pressure applied and no passive

resistance. The deformation of the sheeting is such that, during

this excavation stage, it resists essentially the same load over the

span D to F whether or not strut level E is installed. This would be

particularly true in the stiffer diaphragm walls. The effect of this

-114-
large unsupported length is twofold:

a, Since the sheeting has already assumed an elastic line

such that it resists the full active load, little load is transferred to

strut level E. Hence, strut level D effectively takes a disproportionate share of the load.

b. The moment in the sheeting is greatly increased by the

long unsupported length. As expected, the required depth of penetration for the sheeting will be significantly less than for Case I.

Case III depicts a method for evaluating the maximum strut

load when a relatively weak soil layer starts immediately beneath

the base of the excavation. Since little pa.ssive resistance can be

expected from the weak layer the sheeting acts as a cantilever

member; thus, a large load is developed in the lowest strut. For

these conditions, where the base is stable against bottom heave,

little is gained from driving the sheeting to any depth below the

bottom of the cut. Therefore, a minimum penetration is recommended

of five feet or 20 percent of the excavation depth, whichever is

greater. In situations where the base is unstable, consideration

may be given to deeper penetration and stjtffer sheeting to prevent

bottom heave.

7.60 FINITE ELEMENT ANALYSIS OF BRACED EXCAVATIONS

7.61 Introduction

In recent years several computer programs, based on the

finite element methods of analysis, have been developed to analyze

braced excavations (Wang, 1971; Palmer and Kenney, 1972; Jaworski,

1973; Clough and Tsui, 1974). Currently, the primary use of these

programs is to provide insight into the behavioral trends of braced

cuts. Using computer programs parametric studies can be conducted to evaluate, at least in a qualitative manner, the effect

of wale spacing, sheeting stiffne s s, and soil stratification

on strut loads and sheeting deformation. Further, these

studies may be used to provide guidelines for engineering judgement

and for obtaining some qualitative verifica’tion of design assumptions.

7.62 Case Studies

To illustrate how the finite element programs can be

-115-
used as an aid to the design engineer, the program BRACE II (see

appendix to Chapter 2 for description) was used to analyze three

different soil profile s. These profiles are similar in concept to

those described in Section7. 50, except that cohesive soils are

assumed due to program limitations.

As was the intent of the design examples of Section 7. 50,

the results of these analyses are used to give some insight into

the effect soil stratification and sheeting stiffness m.ay have on the

variation of strut load and sheeting forces.

Specifically, four soil conditions were analyzed:

Case la. Homogeneous soil profile of soft, normally consolidated clays.

Case lb. Homogeneous soil profile of medium-stiff clay.

Case 2. A soft soil stratum above the base of the excavation

underlain by a stiff stratum.

Case 3. A soft soil layer underlying a more competent

stiff soil.

For all cases, the ground water table was taken at a five

footdepth, andundrained soil parameters were assumed for both the

shear strength (S,) and deformation modulus (E,). For soft and

medium soils, these parameters increased linearly with depth

within a stratum as a function of the vertical effective stress

For the stiff soils, the strength and modulus were considered

(8v).

constant with depth. The soil parameters used in the analysis are

summarized as follows:

-116-
Sheark So il*<

Strength Modulus

mu) tEu)

0.28B I 2005

V V

0.4oq 290?v

3200 3 x 1Obpsf

* The parameters for the soft an.d medium-stiff cohesive

soils are from Ladd, et al(19711 These piarameters are normalized

against the vertical effective stress on the soil, i. e. they are directly

related to the effective overburden pressure zv. Henc e, the

strength and modulus increase linearly with a”v and therefore with

depth in a soil stratum.

For the purpose of providing a basis of comparison, the

cofferdam geometry was the same for all (cases:

Strut spacing (L) - 10’ c/c

Depth of excavation (H) - 60’

Width of Excavation ‘- 60’

Sheeting penetration - 30’

Also, two wall types were considered in order to provide some information on the effect of wall stiffness. The wall types were a

steel sheet pile wall (PZ-38) and a 4 foot thick concrete diaphragm

wall.

7. 63 Distribution of Earth Pressures

Figures 42 and 43 show normaliz,ed apparent earth pressure

diagrams predicted by the finite element a:nalysis for the four soil

conditions outlined in Section 7. 62. These apparent earth pressures

were obtained in the same manner as shown on Figure 27 for

apparent earth pressures from field measurements, The

pressures are normalized by taking the ratio of the predicted apparent pressure at a strut level (P) to the maximum apparent pressure

(Pm2x)I both computed from the finite elelment analysis. The diagrams were developed using the maximum strut load computed in

any strut level and during any stage of excavation.

-117-
E

I-J

‘b ymlJT LW

I

PmaX = 2osK

.

‘8

48”CONC. WALL

N16.8 AT BASE OF CUT

p’%ulx CASE la-HOMOGE NIOUS SOFpT/pCLAY

?--Y-i0

max

0

o:s”

PZ-38

43” CONC. WALL

N~4.8 ATBASEOFCUT

CASE lb- -ENlOW hEMUM-STIFF CLAY

Figure 42. Normalized qpparent earth pressure

diagrams predicted by finite element analysis.

-118-
7

Pnwl’~ K

PZ -36 SHEETING

E

+LJ

Pa srRU TwDl Pm’73K

0 Lla

48”coNc. WALL

N* 2.3 AT BASE OF CUT

CASE 2 - SOFT OVERLYING STIFF CLAY

Oo;5 94JLp /

/Ic\

A-+-P A-++

0-c+ 7b++ I

C-V C-t+

D+@ Dj--

I

E+ Ed

/jq

Pmx = VK = MAXIMUM STRUT LOAD

7fi

Pmox = 62K

PZ-38 48” CONC. WALL

N=6.4 ATSASEOFCUT

CASE 3 - STIFF OVERLYING SOFT CLAY

Figure 43. Normalized apparent earth pressure diagrams

predicted by finite element’ analysis,

- 1 1.9 -’
Comparing Case la with Case lb in Figure 42, the analysis

shows that walls in the soft clay should be expected to experience

relatively higher pressures near the base of the cut than the wall in the

mediu.m- s tiff clay, This trend is .more obvious for the stiffer concrete

walls. In addition, the predicted earth pressures in the soft soil may

be much higher, as shown by the maximum strut loads. The apparent

reason for this behavior is the lack of lateral support be.low the base

of the excavations; hence, an inward rotation around the lowest strut.

The stability number of N = 6. 8 in Case la results in a factor of safety

of less than 1. Hence, a bottom heave failure occurred which resulted

in the Loss of passive resistance below the excavation Level. On the

other hand, the stability number for Case lb is low (N = 4. 8), resulting

in a stable bottom. With increasing wall stiffness less curved deformation is expected, and the potential for unloading the second lowest

strut and overloading of the lowest strut is increased. As Case lb

shows, this behavior becomes less pronounced as the soil becomes

stiffer . One possible remedy for reducing this effect in soft soils

would be to prestress the second lowest strut and lock in a high

residual compressive force.

On Figure 43, Case 2 (soft clay overlying stiff clay) shows

opposite effect to that experienced in the homogeneous soil mass.

This stiffer layer provides an adequate reaction for the wall, restricting its inward deflection in the overlying soft clay. This leads

to a larger strut load in the second to last strut and a reduction in

the load received by the lowest strut. This results because the wall

has already deflected inward close to its maximum amount before

the last strut is installed and final excavation completed. Therefore,

this last excavation stage results in little load transfer to the lowest

strut.

For Case 3, where the soils within the depth of cut are

stiff, stability number N<4, and soft soils exist immediately below

the base of the excavation, the results show that the strut loads are

greatest in the second lowest strut. This occurs for the same

reasons given for Case la, that is, lack of support below the excavation base. For this soil profile, the pattern of pressure distribution appears independent of wall rigidity since both give essentially

the same normalized pressure diagram.

-L20-
7.64 Magnitude of Strut Loads

Figure 44 shows the magnitudes of the predicted loads

for Cases 2 and 3. In both cases, the diaphragm wall receives much

greater apparent pressures, on the order of 2 to 4 times that of the

more flexible PZ-38 steel sheeting,

The higher apparent pressures in the concrete wall are

attributed to smaller lateral deformations, hence, less mobilization of shear strength in the soil adjacent to the wall,. The stiffer

wall tends to retain a large portion of the initial stresses, conse -

quently its loading is more dependent on the K o value of the

supported stratum.

This behavior is particularly acute. in the heavily

overconsolidated soils such as those assumed for Case 3. Considering the pressure diagrams for this case, the steel sheeting

experienced a greater inward movement. Therefore, the high undrained shear strength of the soil was mobilized resulting in

relatively low apparent pressures compared to those for the concrete wall which experienced little inward movement.

There is scant field evidence to support this trend.

Observations of tied-back wa.LLs in heavily overconsolidated clays

show that the walls move Laterally with time. The movement may

be associated with the stress relief and subsequent .Lateral swelling

of the soil. This swelling is time dependent and could result in the

build up of strut Loads somewhere between the Rankine active stress

and the initial horizontal stresses in the clay. Inward bulging of

Lagging was observed in an internally braced cut made in overconsolidated soil in the Washington, D. C. area. The severity of the

bulging increased with time suggesting a. load build up on the lagging

and hence, an increase of load in the support system. In any case,

when over consolidated soils are present, one should be aware that

loads may build up on the support system with time causing over -

loading, especially if a relatively rigid waL.1 is used which restricts

the lateral swel.Ling of the soil.

7.65 Structural Behavior

Table 4 compares the predicted location of zero

moment and zero shear versus the sheeting stiffness for the three

cases described in Section 7.61. For all conditions analyzed

-L2L-
APPARENT LATERAL PRESSURE ( psf)

I , I 1 1 f

PZ-38 WALL

b I

C

0

E P

I

I

I

---I=--

48”CONC. WALL

AT BOTTOM OF CUT: N* 2.3

CASE-2 (SOFT OVERLYING STIFF CLAY)

r r 1 ’ 1 ’ I

0 2-0 4ooo ~~00

1 ’ I ’ I

4 a- w@J GDESIGN PRESSURE DIAGRAMS 2

A-) FROM FIGURE 28. A-+ ’

\

PZ-38 WALL

AT BOTTOM OF CUT: N = 6.4

NOTE: EELS FOR DIscwsKlN

CASE.3 (STIFF OVERLYING SOFT CLAY 1

(SEE SECTION 164 FOR DISCUSSlON)

Figure 44, Comparison of predicatd apparent

earth pressures from finite element analysis on

stratified soils.

-122-
Table 4. Summary of structural behavior for

braced walls from finite element analysis.

I

N’

w

I

4’ Diaphragm

Wall 4. 8 8.2 229. 0 )lO ft. 13 ft. 19 ft.

4’ Diaphragm I

Wall 6. 8 11.8 1 229. I 0 > 10 ft. >30 ft. )30 ft.

I

2 PZ-38 2. 3 9.0 5. 8 2ft. 230 ft. 2 ft.

Stiff

Soil 1PZ-3815) 2.3 9. 0 5. 8 3 ft. ) 10 ft. 4 ft.

Below

Base of 4’ Diaphragm

cut Wall 2. 3 9. 0 229. 0 ) 10 ft. 2 ft. 7 ft.

3 PZ-38 6.4 2.2 5.8 7 ft. 1 1

Soft Soil

ft. 19 ft.

Below

“y”, Of 4’ Wall Diaphragm 6.4 2.2 229.0 !

10 ft. 17 ft. 19 ft.

(11All excavations 60 ft. deep; wall penetration 30 ft. (3hased

below base of excavation.

on average S, within the depth of the excavation

(based

r u s acing 10 ft. c/c; lowest strut 10 ft. above

on S, at base of excavation. (4k.set 0P excavation .

(5) Assumed depth of penetration below base was 10 ft.
except one; the sheeting penetration was assumed to be 30 feet,

The exception was for Case 2 where, for one analysis, the PZ-38

sheeting penetration was assumed to be 10 feet to provide a comparison of the effect of sheeting penetration in stiff soils.

It is important to recognize that the finite element

method is for analysis of a soil condition; therefore, the 30 foot

depth of penetration was chosen to provide information on the

effective depth of sheeting.

7.66 Implications of Finite Element Analysis

A comparison of the behavioral trends predicted by

the finite element method (FEM) with the results of the simplified

analytical method (SAM) described in Section 7.50 shows the benefits of performing finite element studies. Even though such a comparison is not theoretically justified since two different soil types

are considered, the results from the FEM are indicative, in a general

sense at least, of the type of behavioral pattern one might expect

for the soil profiles considered,

For example, Figures 42 and 43 indicate that sheeting stiffness has little effect on the apparent normalized earth

pressure diagrams. Figure 44, however, suggests that sheeting

stiffness has a marked effect on the magnitude of the loads, The

stiffer sheeting gives higher apparent earth pressures. This trend

stems from the reduction in ground movement achieved by the

stiffer wall versus the movements associated with the more flexible

steel sheeting. This behavior is not inherently considered in the

SAM. For this latter method, one must rely on engineering judgement to account for the effects of sheeting stiffness on strut loads.

Regarding Case 2, the SAM approach predicts higher

strut loads on the second from the bottom strut and low strut loads

at the lowest strut. The finite element analysis indicates a similar

trend, i. e. , the predominance of load is in the upper strut levels.

However, with the FEM the greatest loads are higher. For example,

the maximum load is in strut B.

-124-
7. 67 The Use of the Finite Element Method in Design

The finite element method is a more realistic mathematical modeling of the complex soil profile and the soil-structure

system, thus making it a powerful tool in the analy,sis of supported

excavation s. However, this method should be used cautiously. The

computer programs are developed by making various assumptions

concerning soil behavior. Naturally, these assumptions have a

large effect on the accuracy of the results. The results are only as

good as the input, particularly the soil parameters assumed. Therefore, these programs should be used only by experienced individuals

who are aware of the assumptions used in the program development

and how the assumptions could possibly effect the results. Finally,

when making a finite element analysis, the results should be carefully

evaluated for consistency and behavioral trends. The results may

appear correct, but because of inappropriate input or misapplication

of the program, the analysis may be giving erroneous-results.

It is recommended that until substantially more experience is gained with the FEM as a design tool it be used primarily

as an aid to guide engineering judgement.

-125-
Case 'i ' case 2 &se '3'

-126- I
._. ---_- __.^.. -_-. ___-

-127-

x= /5: y=

0

= - /‘Q44
/

70)4*

I

\

-- -

Gbc/bl 77/l
-131-
SUMMARY Ah!! CG?MPAHSON Of 77fEE c%5ES

--.. 1

Jand

/O I / 1 ,

I

/

\

\

I

-132-
CHAPTER 8 - STABILITY ANALYSIS OF SHEETED EXCAVATIONS

8.10 GENERAL

There are three primary modes of instability of concern in

sheeted excavations in clay:

a. Bottom heave (Figure 45a).

b. Deep seated failures (Figures 45b and 45~)

c. Local failures immediately adjacent to the support wall.

Of these modes, b). and c). are related to the overall stability of

the excavations. They often will dictate the procedures to be used

in constructing a cofferdam. These condi.tions of instability may

also result in heavier strut loading than predicted by the method

described in Chapter 5. An example of th.e potential for increased

loading if bottom heave occurs is described in Section 7. 50”

Local failures (mode c) are of concern where it is necessary

to limit inward sheeting deformations. Failures of this type occur

below the excavation level immediately adjacent to the sheeting,

resulting in partial loss of lateral support.. This loss of support

creates a large unsupported length and can lead to excessive inward deflections of the sheeting.

8.20 BOTTOM HEAVE

Bottom heave is a problem primarily in soft to medium clays

where the strength of the soil is nearly constant with depth below

the base of the excavation. The failure is analogous to a bearing

capacity failure, should be analyzed (Bjerrum and Eide, 1956)

using the stability chart given in Figure 46. The factor of safety

against a bottom heave is determined as:

-133-
I

I

I

I

I

H

I

I T1)

I HEAVE

I --- /- ,I

T

I ------ / \ \

'ly ; \/ \

/l

’ ‘1 /.’ \

-A /

a) BOTTOM HEAVE FJULWE- -

I a

&n I

i

I

I

I

I

w--w--

‘s

I

t u

‘1

/

/

b) DEEP SEATED FAILURE

INTERNALLY BRACED

EXCAVATION

. ROTATION QcFNTER

C) DEEP SEATED FAILURE

TIED BACK

EXCAVATION

Figure 45. Potential failure surfaces.

1134-
L* LENGTH OF TRENCH -

0 I 2 3

H/B

Figure 46. Bearin.g capacity factors for

bottom stability analysis,

-13!j-
S N

F. S = Ncb (

YH+Uq

= c b

N

where:

N = stability number = XHt q

%

N cb= bearing capacity factor from Figure 46

S

U

= the undrained shear strength of the clay

# = total unit weight of the soil

cl = uniform surcharge loading on the area adjacent

to the excavation

Where the soil is stratified within the depth of excavation and

below, a weighted average of undrained strength should be used for

%I* This average should be taken over a zone described between B/Jr

below the excavation base and 2.5B above the base.

8.30 LOCAL FAILURE

When braced excavations reach a certain depth in clay soi.ls, the

lateral pressures on the retaining wall coupled with the stress relief

from the excavations can be of sufficient magnitude to cause local

yielding of the soil immediately adjacent to the inside of the sheeting.

This localized overstressing results in loss of passive resistance

which in turn leads to uncontrolled inward movements of the sheeting.

As the excavation proceeds, these inward movements become additive, resulting in large inward movements and a corresponding loss

of ground adjacent to the excavation. D’Appolonia (1971), Jaworski

(1973)) and O’Rourke and Cording (1974a) all show data which indicate

these uncontrolled movements can account for up to 50 percent of

the loss of ground.

Figures 47a and 47b can be used to estimate when local failure is

imminent in cohesive soils where flexible sheeting is used The failure

is related to the shear strength (Su) and to -the initial state of stress

in the ground. Figure 47a shows the factor of safety against bottom

heave necessary to prevent local yield as a function of excavation

geometry and the shear stress ratio. The shear stress ratio (f) is

a convenient dimensionless parameter which defines the initial state of

-136-
I

CuRwiS PNlEmNED Fnou

RESULTS OP PINITE ELEMENT

cwPuTERPRcxMM-m&E

-

( FROM D’AFI’OLONI A. 1971)

-1.0 -0.5 0 +o.s + 1.0

SHEAR STRESS RATIO, f =

I-K,

2s

=vQ

Figure 47a. Factor of safety required to prevent

local yield below bottom of excavation in clay.

- 1.0 -0.5 0 to.5 t 1.0 t1.3

SHEAR STRESS RATIO, f

Figure 47b. Shear stress ratio vs. over consolidation ratio.

-137-
stress in the ground and the strength of the soil. Figure 47b gives

the variation of the ratio (f) versus over consolidation ratio for Boston

Blue Clay.

The depth at which 1oca.L failures begin to develop is related to the

shear strength of the soil and the initial state of stress in the ground.

The potential for local yielding is most prominent in the over consolidated

soils. The reason for this is the high value of K,, ( ~ho/~vo), which

is close to or can exceed 1. The failure at the base of the excavation is

one of extension; that is, the shear stress (or deviator, stress) is ’

increased by a decrease in the vertical stress. It follows that the

higher the K, value the closer the soil is to a failure condition. Hence,

it takes less stress relief to cause overstressing of the soil.

Figure 47a shows that where the depth to width ratio of an

excavation is 1 and the OCR is 6, a F.S.Y 3 may be necessary to

prevent local yielding. On the other hand, for the same excavation in

the softer normally consolidated soils (OCR = l), a F. S. 5 1.5 is

sufficient to limit local yielding. This lower factor of safety is associated

with the low K, value (K,s 0.5) in normally consolidated soils. Thus,

they can experience much Larger stress release before failure.

Figure 48 shows when Local yielding starts in normally consolidated clays as a function of sheeting stiffness (K) and excavation

geometry (H/B). These data were developed using a finite element

program (see Appendix to Chapter 2). The results indicate that for a

given excavation geometry (up to H/B ti 1.0) stiffening the sheeting

reduces the factor of safety required to prevent local failure. This trend

is related to the ability of the stiffer sheeting to act as a cantilever

wall while minimizing inward movement.

8.40 DEEP SEATED FAILURES ---

8. 41 Internally Braced Excavations

8.41.1 Circular Arc Analysis

In situations where internally braced excavations

are either underlain by weak soils or the ground adjacent to the excavation slopes upward, the overall stability of the excavation should

be analyzed.

One approach to analyze the stability is by the

classical circular arc analysis as illustrated in Figure 49. It consists

of assuming a series of centers of rotation and failure surfaces to find

-138-

BRACED WITH EITHER

CROSS LOT OR RAKED e7m, ITCI

*COMPETENT

u

3

SOIL LAYER

Consider overall stability:

Moments around center of rotation

Forces to consider:

1) Weight of driving mass (WT)

2) Resisting strut loads (P,, Pz) (Horizontal component

of support load.)

3) Resisting shear capacity of wall (Hs) from combdtent soilkyer.

4) Shear strength of soil, frictional component (T). and

Cohesion, (c)

Note: If rakers used, kicker must be located outside failure

mass for PI and P2 to be considered in analysis.

=“R

Safety Factor = zh/l = w

(N tan $ t CL) R

D T aw - p111 - P212 - HsR

Figure 49. Stability of internally braced cut (circular arc method).

-140-
the minimum factor of safety against a rotational failure. The conditions shown are for the case of a homogeneous soil where the driving

forces are the total weight of the soil mxss plus any surcharge loads.

Resisting forces consist of the strut loads (PL, P,), the shear strength

along the failure arc, and the shear capacity of the sheeting below

the failure arc. If the soil is stratified,then the stability analysis should

be made using the classical “Method of Slices”.

The analysis will determine the hypothetical failure

surface bounding the failed soil mass. If rakers support the wall, care

must be taken to insure their kicker support is outside the failure zone.

Otherwise, the thrust forces from the rakers should not be considered in

the analysis.

The sum of the strut forces necessary to maintain

a stable excavation should be compared to those predicted from the

Lateral earth pressure diagram as outlined in Section 7. 50. The greater

of the two total Loads should be used to estabLish the ordinate of the

design earth pressure diagram.

In the cases where the retaining wall extends through

a weak Layer into a highly competent soil, the structuraL resistance of

the retaining wall (Hs) should be considered in the analysis. The shear

resistance should be taken equal to the passive force determined in

accordance with Chapter 6.

8.41.2 Wedge Stability Analysis

An alternate means of evaluating the overall stability

of an internally braced excavation is to make a wedge stability analysis.

It is often a simple method for analyzing a stratified soil deposit for

the maximum loads which might occur in a support system.

Figure 50 shows this method of analysis for an

internally braced excavation. The analysis illustrated is general, with

no assumption for either failure surface or direction of active and

passive Loads. However, this Leads to a tedious analysis. A simple

alternative for analyzing this condition is to assume Rankine conditions

for failure surfaces and direction of Load. ALthough this Latter approach

does not yield theoretically correct answers, the results will be adequate

for most problems.

-141-
L- ACTIVE WEDGES

Stratum I

(P T t CT,

v

T

1%

5itratum II

4 ‘11’ d 51 T T

P I

P2

WlV

Stratum III

“III9 %I’ YIII

For general solution varyd, 8, and&angles to obtain minimum value

for factor of safety.

Method of Analysis:

1. Assume o(, ,&, & angles.

2 . Sequentially analyze the active and passive segment for loads

P

III and PV’ Include water pressure.

3. Sum forces in horizontal di

i.e. F.S. =

p1 +p2

pIII - p v

“<I

w.u

--9

PI

v

P WI \

\

\ %+2 4 o*IT

\P

ij

\%

F

(% :?I

\

P

---

a ’

wI’

P

u=

I

P

wII

= horizontal water pres

uplift water force on Wedge II

rection for factor of safety

sure

Typical Analysis of Wedge

(Wedge II)

Figure 50. Wedge stability analysis for braced cut.

-1421
8. 42 Tied-Back Walls

Detailed procedures for analyzing the stability of tied-back

walls by a variety of methods employing trial planar surfaces and wedges

are presented in Volume III. By and Large, these methods placed emphasis

upon the failure surface passing through the zone of tiebacks. This

technique may be used as a design tool for establishing the appropriate

Length of tiebacks.

This section makes a simplified presentation of the circular

arc method below as a means to examine overall stability for a failure

surface passing beyond, or nominally through, the tieback zone. This

concept is particularly appropriate when weak soils occur near or below

the excavation base.

The analysis is quite similar to that used for internally

braced excavations. Figure 51 illustrates the general approach for

an assumed circular failure surface. The analysis must consider the

position of the anchor relative to the failure surface. The example

illustrated shows the surface cutting through the lowest anchor. The

resisting force contributed by this anchor is a function of the amount of

anchor outside the failed mass. If the failure surface passed before

the anchor zone, then the tension force may be assumed to be the full

design force in the tieback, T . For the surface shown, the failure

plane passes through the anch& zone, therefore, it is necessary to make

an assumption concerning tension force remaining in the tieback.’ With

ties having essentially uniform resistance in the anchor zone the load

variation will be linear. Thus, the value of T to be used in the analysis

may be taken as:

TzY T

XfY

C

where T c is the total force in the anchor. In cases where ties are

anchored in rock or belled anchors in highly competent soils are used,

the full tension force may be assumed since the failure surfaces will not

cut through these strata.

-143-
’

Take moments about center of rotation.

ZM = Wll tP 1

0 9 4

- U& + HB R1)

e”R = (W COB 0 tan $ + CL) Rl + TC cos 8 tan 9

Safety Factor:

=“R F. S. = 

CM

-

0

Figure 51. Stability of tied-back excavation,

-144-
CHAPTER 9 - BEARING PRESSURE OF DEEP FOUNDATIONS

9.10 GENERAL

This section is directed toward those basic considerations used

to establish bearing values for elements used in connection with cut and

cover operations. These principles would be applied for underpinning

units or for walls and soldier piles subjected to a vertical component

of load.

Typically, the bearing stratum is deep -- that is, it lies at great

depth relative to the width of the bearing area. Accordingly, those

design rules developed for “shallow foundations” such as are presented

in Terzaghi and Peck (1968) will be overly conservative.

Fundamentally, allowable bearing value must recognize two

governing criteria -- first, adequate safety against shear failure of

the foundation and second, a limitation of settlement. Usually, it is

shear which controls for clays and settlement which controls for sands.

The following discussion presents those basic tools required to assess

the above stated criteria.

9.20 PRESUMPTIVE BEARING VALUE

Table 5 presents a summary of the range of allowable bearing

values for building foundations resting on a variety of soil types. This

tabulation is not intended to represent a recommendation for design;

rather its purpose is to convey a means to assess the relative competency

of different materials and to provide a crude initial guide. Because the

values typically apply to shallow foundations, acceptable values for deep

foundations will be somewhat higher.

9.30 BEARING VALUES BASED ON SHEAR FAILURE

9.31 General

The following represents a summary of theoretical procedures for calculating net ultimate bearing capacity using shear strength

parameters, 3, o co f hesionless soil., and undrained shear strength, S,, of

cohesive soils. A factor of safety of 2 to 3 should be applied depending

upon risk and the confidence level in data.

-145-
Table 5. Abstract of presumptive bearing capacity, ksf.

A B C D E

Mass. State New York Atlanta National Board BOCA

Code (1974) City (1968) (1950) of Fire Under- (1970)

Glacial Till:: 20 - - - - - - - -

Hardpan:: - - 16 - 24 2 0 20 20

Gravel, wellgraded sand and

gravel::

Coarse sand+

10 8 - 20 8 -12l 8 - lzl 8 - 12l

6 6 - 12’ - - 6- 8l 6- 8l

Medium sand:: 4 - - - - - - 4(100 se)

Fine sand 2 -4 4- 83 - - 4- 6l - -

Hard clay 1 0 1 0 - - - - - -

Stiff clay - - - - 4 5 - -

Medium clay 2 4 - - 5 - -

*

Massachusetts and New York Code allow 5 percent increase in bearing value per foot of additional

embedment, but not more than twice tabulated value.

1 - Range reflects compactness, gradation, and/or silt content

2 - 0. 1 x N, but not less than 6 ksf nor more than 12 ksf (where N= no. of blows in SPT)

3 - 0. 1 x N, but not less than 4 ksf nor more than 8 ksf (where N= no. of blows in SPT)
9. 32 Sand

For deep piers, in sand, the end bearing load capacity is

generally expressed as:

%l

where:

N =

4

%l =

ultimate bearing capacity (load per unit area)

Values of N, vary depending upon assumptions made in the

derivation. Vesic (1965)jpresents ranges for the values as shown in

Figure 52. The lower curves represent modes of failure in which the

shear strength of the soil is developed below the footing. Higher values

of N will be obtained by assuming that the failure surface extends above

the p ane P of bearing, thus engaging shear resistance above that level,

As a practicaL matter, because of the high bearing capacity

of sand there is little penalty in adopting a conservative value. For

example, consider an extended underpinning pier bearing at a depth of

20 feet on a sand with d - 35O. Assuming a unit weight of 125 psf, the

effective stress would be:

= Nq ‘,

dimensionless bearing capa city factor that

is a function of the shear strength’parameter,

d, of the bearing meterial and shape of the

loaded area

effective stress in the soil at the bearing surface

%

= 20 x 125 = 2500 psf

N

Q

= 58 (Veeic)

= 75 (Berezantsev)

= 110 (Brinch Hansen, 1961)

- 147-
1 I I I I 10 250 300 350 400 450 so0

ANGLE OF INTERNAL FRICTION 4 -e

FROM VESIC (1965)

Figure 52. Bearing capacity factors for deep

circular foundations.

-148-
Thus, the ultimate bearing pressure will range from:

Vesic: x =

Berezantsev:

58 x 2500 = 145,000 psf

Brinch Hansen:

Obviously, all values are quite acceptable. Accordingly,

there is Little to be gained in most applications by debating the appropriate value of Nq.

ultimate values.

In general, a safety factor of 3 is applied to these

As stated above the settlement limitation usually

controls in granular soils.

9.33 Clay

In clays the undrained strength, Su, rather than drained

strength will control the bearing capacity of a foundation element.

Skempton (1951) presented bearing capacity factors N, for net ultimate

bearing capacity in clays. In this case, ‘1 net” means pressure in excess

of the effective overburden stress of the bearing level.

where:

9, net ultimate bearing capacity (load per

unit area)

N =

c

dimensionless bearing capacity factor

S

that is a function of the shape of the loaded area

u

undrained shear strength of soil

For deep foundations (at depth greater than 4 to 5 times

the breadth of the Loaded area) values of Nc are as follows:

Circle: N =9

c

Strip: N = 7.5

c

Rectangle: N

c

= 7.5 (1 t 0.2 B/L)

where: B = breadth and

L - length

Note that for clays, the net ultimate bearing pressure is

independent of depth (and therefore overburden stress). It is a function

only of the shape of the loaded area and undrained shear strength of the

soil.

In addition to the load bearing capacity at the base, the

side friction may be determined on the basis of the embedded area and

-149-
adhesion along the shaft. In soft clays, the adhesion is equal to or only

slightly less than the undrained shear strength. However, in stiff to

hard clays the adhesion is typically less than one-half the undrained

strength. Tomlinson (1969) presents a summary of data for adhesion

in both driven and bored piles.

Much of the data on shaft adhesion was developed on bored

piles in London clay. The practice is to apply a reduction factor, d ,

to the undrained strength to estimate adhesion. Thus:

S

eff =o(sU

where :’

oc = reduction factor

S

U

= undrained shear strength, psf

S

eff = adhesion along shaft, psf

Figure 53 (after Peck, et al, 1974) shows that o( decreases as the shear strength of clay increases. In general, shaft

adhesion is counted on for load support in very stiff to hard clay, In

this range, & varies from about 0. 3 to 0.5. For stiffer cLays, the

average developed adhesion, a( S,, shows Little variation with

increasing shear strength. It varies from only about 1 tsf to 1. 3 tsf.

Jn areas where there is Little prior data, Skempton

recommends a maximum adhesion of 1 tsf (Tomlinson, 1969) when

using the chart. The total capacity of the shaft is equal to:

Qshaft = cd. SUA

where:

A = shaft area

Again, a safety factor of between 2 and 3 should be used.

9.40 BEARING VALUES BASED ON SETTLEMENT

9.41 General

The following presents the recommended procedures for

estimating the settlements of deep foundations.

-150-
0.8

0.6

d

0.4

0.2

C

2

I.5

I

0.5

AVERAGE

0(x S”

tsf

FROM PECK, HANSON a THORNBURN

(1974)

Figure 53. R.eduction factor in S from observed

capacity of friction pile:.

-151-
9.42 Soils Having Constant Modulus of Deformation with Depth

9.42. 1 Surface Loading

Theoretical procedures for determination of

settlements have been developed by Fox (1948) for square and rectangular bearing areas and by Woodward, et al (1972) for round bearing

areas. These are based on integration of the Mindlin solution for

a point load within an elastic half space. At a depth equal to zero,

the Mindlin solution is identical to the familiar Boussinesq solution.

These solutions all have the general form.

9

= q $I? (1 -U2) Eq. 9.42. 1

where:

/p = settlement

q = distributed load

B = least dimension of foundation unit

E = modulus of deformation

9 = Poisson’s Ratio

I

P

= influence factor which depends on rigidity of

footing, shape of footing, and depth of footing

A simplified method for determining settlement

at the surface is based upon a coefficient of subgrade reaction defined as

follows:

k =

p”

Eq. 9.42. 2

And thus, the settlement is computed as follows:

p=+-.

Eq. 9. 42. 3

where:

P

and q are defined as above

k = coefficient of subgrade reaction in general units

of pressure per unit deflection

-152-
The value of the coefficient of subgrade reaction is

commonly determined by plate loading tests or by correlation with in

situ soil indices such as relative density and standard penetration resistance. By comparison of Eq. 9.42. 1 and 9.42.3, the coefficient of

subgrade reaction is related to the theoretical settlement as follows:

k = E

B(1 - 9’) Ip Eq. 9.42.4

For a constant footing shape and depth and constant

material properties, the coefficient of subgrade reaction for a footing of

size B is therefore related to a footing of size B’ by the following

relationship:

k I B= kg (5, Eq. 9.42.5

It is common to express the coefficient of subgrade

reaction in terms of the value for a 1 foot square plate (kl) as this is

the size for conventional plate loading tests. Therefore,

kl

kB = B

Typical values for kl are shown in Figure 54.

9.42.2 Rectangular Footings

Influence values for other than square footings can

be determined from elastic theory. These values however, become

very large for long narrow footings and in fact approach infinity

(k approaches zero) for an infinitely long footing. These results directly

follow from the fact that the Boussinesq solution does not approach

the actual plane strain conditions when integrated to infinite limits.

Therefore, the elastic solution is unrealistic for long footings. To

solve the problem, Terzaghi (1955) has proposed the following empirical

relationship for rectangular footings:

k

=k (L+.5B/L)

LxB B I. 5

Eq

. 9.42.6

where:

k

LxB=

coefficient of subgrade reaction for footings of

Length, L, and width, B

kg = coefficient of subgrade reaction for square

footing of dimension, B

-153-
350

300

250

200

150 ~

100

5 0

0

FINE GRAINED SOIL

1

STIFF I VERY STIFF I

I 2

I

UNCONFINED COMPRESSIVE STRENGTH 0, TS.E

300 k, =COEFFlClENT OF SUBGRADE

REACTION FOR I FOOT

SQUARE BEARING PLATE d

AT GROUND SURFACE.

GRAt NEd SOILS

IO 20 3% 40 5 0 6 0 40 Jo Jo loq

I VERY 1 LOOSE I MEDIUM DENSE

LOOSE 1 DENSE 1 o”EENRsyE 1

COARSE GRAINED SOIL

Figure 54. Coefficient of subgrade reaction vs.

in situ soil indices (NAVFAC, 1971).

-154-
This relation suggests that the subgrade modulus

for an infinitely Long footing approaches a value equal to 2/3 of that

for a square footing.

Both the elastic theory and Terzaghi’s empirical

relationship are plotted in Figure 55. The recommended procedure is

the Terzaghi relationship especially for Larger values of L/B.

9.42.3 Effect of Depth

For a footing with constant Loading, shape, and

mater ial pr opertie s , the subgrade modulus of that footing is inversely

proportional to the influence factor (see Eq. 9.42.4). Thus, when the

influence factor varies with depth, the ratio of subgrade ,modulus at the

surface to the subgrade modulus at depth may be computed as follows:

ksB = I;

- -

kD Is

B P

Eq. 9.42.7

where:

S

33 = coefficient of subgrade reaction for a footing (breadth B)

at the surface.

k; = coefficient of subgrade reaction for a footing ( breadth B)

at depth, D.

$ = influence factor for footing at depth D.

1: = influence factor for footing at surface.

Elastic theory demonstrates a variation of

influence factor, I with depth. On this basis, the depth factor FD has

been plotted in FiGre 56 for circular and rectangular footings. Note

that this figure is for the special case of constant modulus of deformation.

Also shown on this figure is a plot of depth

factor for a circular shaft which relies on 100 percent side friction

and no end bearing. It can be seen that the depth factor is Less for this

case than for the case of 100 percent end bearing for depth ratios.

( D ) greater than approximately 1.25. For all cases,

d LxB

. the effect of depth is to increase the subgrade modulus and thus to

reduce settlement.

-155-
1.0

0.9

0.8

0.7

0.2

0.1

0

LENGTH TO

BREADTH RATIO

v s

SHAPE FACTOR

FOR A CIRCLE OF

DIAMETER 8,G ‘0.89

TERZAGHI (1955)

ELASTIC THEORY

K

LX’

= COEFFICIENT OF SUBGRADE REACTION FOR

RECTANGULAR FOOTING OF WIDTH 8, LENGTH L.

KB = COEFflClENT OF SUBGRADE REACTION FOR

SQUARE FOOTING.

I I

0.1 1.0

L/B

10.0 100.0

LENGTH TD BREADTH RATIO

Figure 55. Shape factor for rectangular footings.
1.0

.95

.90

.05

K

co

80

gkloy .I5

XII

Lo .?O

x’ .65

.60

.55

.50

DEPTH RATID

v8

*

1

:LE DEPTH Fm

g0.5)

\

I

0.2 0.4 0.6 0.0 I.0 0.0 46 \ 0.4 42

l - K; COEF-fICIENT OF SUBGFMDE REACTION AT SURFACE.

. K= II ,I ”

” DEPTH, D

. LxB = FOOTING DIMENSIONS

0

RECTANSULAR AREAS FRDM FOX( l940)

CIRCULAR -SF-WOOD-0 ti al (lS72)

NOTE: THIS FlGURE BASED UPON THEORY WITH

CONSTANT MODULUS OF ELASTICITY

(DEFORMATIONJ

Figure 56. Influence of depth on coefficient of subgrade reaction (based on

modulus of deformation that is constant with depth).
9.43 Soils Having Modulus of Deformation that Increases with Depth

Surface 9.43.1

In the case where the modulus of deformation

increases linearly with depth, it can be seen that the coefficient of

subgrade reaction will vary in direct proportion to the increase in

modulus. This effect comes into play in two ways: first, for larger

size footings a larger area is loaded and consequently a greater depth

of stress influence is created; and second, for footings at depth, the ,

deformation modulus E, will not be the same as at the surface, thereby invalidating the relationships in the preceding section.

For the former case, Terzaghi (1955) has proposed

the following empirical relationship to convert the coefficient of

subgrade reaction for a 1 foot square area to an area B x B square.

Eq. 9.43. 1

9.43.2 Rectangular Footings

Once kgisdetermined, kL x B at the surface can

be obtained from Figure 55.

9.43.3 Depth Effects

The special case of constant modulus of

deformation was discussed in Section 9.42. 3. In addition, Taylor

(1948) has proposed an embedment correction to account for the increase

in modulus of deformation with depth as follows:

k; = k; (L t 2 D/B) Eq. 9.43.2

Where: D = depth of footing

B = least breadth of footing

In using this relationship, care must be taken to assure that

the value of kl used to determine kB does represent the material at the

surface. If the value of kl is determined from correlation with indices

such as standard penetration resistance or relative density at the bearing

level., the correction for increase in modulus would not be made.

-158-
A second method of evaluating the effect of increasing modulus

of deformation is to consider the coefficient of subgrade reaction to be

directly proportional to the initial tangent modulus Eit. Janbu (1963)

shows that Eit for granular soils is proportionaL to a power function of

stress Level. SpecificaLLy:

E*1t - (a,) n Eq. 9.43. 3

where:

cr3= Ita era1 stress (assumed to be effective Lateral stress)

n = 0.3 for gravels, 0.5 for sands

Accordingly, it follows that:

ksB =

FDG Eq. 9.43.4

where: FDC is defined as the depth factor for granular soil.

In normally consolidated deposits, 33 is proportionaL to the overburden stress and therefore to the depth. The following

equation results:

ksB

D

kB

n

= F

DG Eq. 9.43.5

expression is:

The recommended val.ue of ‘z’ to be used in this

z=D

F

t 0. 75B Eq. 9.43.6

where :

DF= depth of footing from average ground surface

The results of both methods for determining depth

effects in soils with varying modulus of deformation are presented in

Figure 57. Note that a Limitation of FDG = 0.5 has been set on the

Taylor expression. The relationships shown in the figure are typically

-159-
1.0

I

0.9

UJ P

YY

0.8

2 07.

%

J 0.5

z

$

(3 0.4

DEPTH RATIO L-v s

DEPTH FACTOR

BASED ON JANBU(l963)

JANBU(1963)

KS = COEFFICIENT OF SUBGRADE REACTION AT SURFACE.

KD = COEFFICIENT OF SUBGRADE REACTION AT DEPTH, D

1.0 too

Figure 57. Influence of depth on coefficient of subgrade reaction

for granular soils (based on modulus of deformation that increases with depth).
applicable to granular soils because their modulus of deformation

is a function of stress level, and therefore, depth. Such soils rarely

exhibit “over consolidated” behavior: however, they would only be

appLicabLe to cases where the determination of the basic value of subgrade modulus did not already include the effects of increasing modulus

with depth. For instance, if k were based on the average standard

penetration resistance in the zone of significant stress increase beneath

the proposed footing,no correction for increasing modulus with depth

would be made (F

=E

= 1). On the other hand, if k were based on a

plate loading test a the surface, the value of FDG as determined from

Figure 57 should be used.

9. 43. 4 Water Table Effects

The presence of ground water in granular soils will

effect the modulus of deformation by reducing the Lateral effective

stress. The effect of the ground water table can be esti,mated by

considering equation 9.43.4 and computing the lateral effective stress

at midpoint of the zone of significant stress increase. If the water

table is below a depth of 1.5B beneath the base of the footing, then no

water table correction would be necessary and Figure 57 could be

used directly. If the water table is at or above the base of the footing,

then FDG would be computed using effective stress values at the

average depth of significant stress increase substituted into

Eq. 9.43.4. Where the water table lies between these limits, a

pro-rated correction should be made.

The effects of the water table would only be considered where the water table effects have not been considered in

determining the basic value of the subgrade modulus. For instance,

if k is based on a plate loading test at the surface, with the water

table also at the surface, Figure 57 could be used directly since

water table effects would be accounted for in the value of k; however,

if the footing and water table were at some depth D greater than the

influence area of the loading test, F

D C would be computed as described above. Conversely, if the value of k were based on the average standard penetration resistance in the zone of significant stress

increase, no adjustment to k would be necessary, either for water

table effects or the increase in modulus with depth.

-161-
9.44 Recommended Procedure for Determination of

Settlements of Deep Foundations

9.44.1 Clays

Assume that modulus of deformation is constant

with depth. Compute settlement for Eq. 9.42.2.

P,= q/k

where:

4 = load in tsf

k = coefficient of subgrade reaction in tsf/ft

P= settlement in feet

Determine k by first determining kl at the

surface of the soil from Figure 54 or from plate load test.

Modify k, as follows, .to account for size,

shape and depth:

kl FS

k=y F

ClD

where:

FS = shape factor from Figure 55

FD = depth factor from Figure 56

B = least dimension of bearing area in feet

9.44.2 Sands

Assume modulus of deformation .increases

with depth. Compute settlement from Eq. 9.42.2 as above. Determine k by first determining kl as above.

account for size, shape and dept&

Modify kl as follows to

2

kl x FS

k =

FD x FDC

-162-
where:

FS, FD, and B as defined above

FDG = depth factor for granular soil from

Figure 57

9.45 Example Problems

9.45,l Problem Number 1

Determine the average settlement of a

continuous footing for underpinning; 4 feet wide at a depth of i0 feet

in sand. The water table is at a depth of 50 feet and.the average

load per lin3ear foot of footing is 20 kips. The subgrade modulus of

200 tons/ft was determined from a plate Ioading test at the surface

prior to excavation. The unit weight is 125pcf -

4 Make correction for footing size and shape:

(1) Find k4

(coefficient

-2

for 4’ x 4’ footing)

kB

4t1 2

k4 = 200 r 8 3 = 78 tons/ft3

(2) Pind shape factor for an itiinitely long

footing 4 feet wide. See Figure 55.

FS = 2/3 (Tereaghi curve)

b) Make correction for footing depth.

(1) Find depth factor, based upon elastic

theory and constant modulus. See

Figure 56.

FO

= 1.0 for L = 00 by depfW

(2) Find depth factor due to increasing

modulus with depth. See Figure 57.

-163-
D/ B = 30/4 = 7.5

from Figure 57

FDG = 0.5 (both Taylor and Janbu criteria)

cl Find corrected coefficient of subgrade

reaction.

G3

k = 78 x

FD x FDG

=78x$x

1

0.5 x 1.0

= 104 tons/ft3

d) Settlement Computation

P

= q/k

20k/ft

4=-4ft = 5 ksf

k = 104 tons /ft3 = 208 k/ft3

5

P = 208 = 0.024’ = 0.288”; say& 0. 3”

9.45.2 Problem Number 2

Determine the settlement of a 3 foot square footing for

an underpinning unit at a depth of 5’ in stiff clay. The unconfined

compressive strength of the clay was L. 5 tsf,and the footing load

is 45 k.

a) Determination of k

1

From Figure 54 for fine grained soils:

kl = 60 tons/ft3

b) Make correction for size

kg = kL (l/B) = 60/3 = 20 tons/ft3

-164-
c) Make correction for depth

1) Find depth factor based upon elastic theory and

constant modulus. See Figure 56.

JLfB, =5+ = 1.67

FD= 0.64

d) Find corrected coefficient of subgrade reaction

k=20x+- = o%4

D ’

31 tons /ft. 3

e) Settlement Computation

p= q/k

J

45k

q =-

9 ft.2

= 5 ksf

k = 31 tons/ft. 3= 62klft. 3

P

=-

5

=

6 2

. 0.08’ = 1.0”

9.45. 3 Problem Number 3

Same as problem number I except that the water

table is at 5 feet. The only difference will be the determination of FDG

.

a) Make corrections for footing size and shape

k4 = 78 tons/ft. 3and FS = 2/3 (from example I)

b) Find depth factor ) due to increasing modulus with

depth. In this case F

(FDC

must be determined by Eq. 9.43.4, since the

original estimate of su73Ggrade modulus did not include the effects

of the water table. Let K. = at rest earth pressure coefficient.

Under a 4 foot wide load at the surface

33 = 125 (0.75 x 4) K, = 375K,

-165-
Under a 4 foot wide load at 30 feet with the water

table at 10 feet.

a3 = b25 x 5) t (25 x 62.61 K. = 219OK

0

- S

03 375

-D = -=0.17

a3

2190

FDG

= (0.17) Oo3 = 0.59

c) Find depth factor based on elastic theory and constant

mo dulu s .

PD = 1.0 from Example 1,

d) Find corrected coefficient of subgrade reaction

F

k=78 FSxF

DG D

= 78x$x ’

0.57 x I.0

91 tons/ft. 3 = 182 k/ft. 3

e) Settlement computation

P

=q/k= s2 =0.027' = 0.324"; say - 0.3"

9.45.4 Problem Number 4

Determine the settlement of a 5 x 5 footing at a depth

of 10 feet in sand and gravel. The load is 80 to s

3

and the coefficient of

subgrade reaction is estimated to be LOO tons/ft for a L square foot

footing on the basis of the standard penetration resistance between 10 .feet

and 17. 5 feet. The water table is at 10 feet.

a) Correct for size effects

k5 =kl( 2B BA) 2 = LOO (;o)2 = 36 tons/ft. 3

where: k = coefficient of subgrade modulus for

5’ x 5 ’ fzoting at surface.

-166-
b) Shape effect - None required for square footing.

Fs = 1.0

c) Make corrections for depth effects

I) Consider elastic theory and constant modulus.

See Figure 56

D/B = 5/10 = 0.5

F

D

= 0. 85

2) Consider the effect of increasing modulus with

depth.

No correction for increasing modulus or water table

effects since k was based on data from the zone

of influence of p!roposed footing. FDG = I. 0

d) Find corrected coefficient of subgrade reaction

k=36x FS

FDx FDG

= 36x I. 0

0.85 x I. 0

k = 42 tons/ft. 3

e) Settlement computation

P

= q/k

80 tons 2

q= 5x5 = 3.2 tons/ft.

P

3.2

= - = 0.076’

42

= 1. I”

-167-
CHAPTER 10 - CONSTRUCTION MONITORING

10. 10 INTRODUCTION

Geotechnical engineering, by its nature, involves

contingencies from unforeseen conditions that are encountered

during construe tion. Bold innovative designs may justify experimental test sections and then continuing re-assessment and verification during actual construction.

The preceding chapters in this volume are addressed to

engineering during the design phase. Geotechnical engineering

must extend beyond design into construction, and therefore it is

essential that data be obtained for re-evaluation of design assumptions and implementation of appropriate modifications.

10. 11 General

This chapter presents an overview of the purpose of

construction monitoring, what is measured, and how the task is

planned and executed. Emphasis is placed on open cut deep excavations and adjacent structures. Construction monitoring case

histories were reviewed by Schmidt and Dunnicliff (19741, who also

describe construction monitoring of soft ground tunnels.

10. 12 Reasons for Construction Monitoring

If a construction monitoring program is performed for

the right reasons, planned properly, and executed by diligent engineers,

it can make a large contribution towards increasing safety,reducing

cost, and reducing the impacts of construction on the environs. Some

valid reasons for monitoring are:

Diagnostic: To verify adequacy of design

To verify suitability of construction

technique s

To diagnose the specific nature of an

adverse event

To verify continued satisfactory

performance

Predictive: To permit a prediction of behavior later

on at the same job

-168-
Legal:

Research:

To establish a bank of data for possible

use in litigation

To advance the state-of-the-art by

providing better future design data

10.20 PLANNING CONSTRUCTION MONITORING PROGRAMS

Many construction monitoring programs fail to achieve their

purpose because the engineer does not approach the program design in

a logical sequence. There is a tendency a,mong engineers to select

an instrument, make some measurements, and then wonder what to

do with the data (Peck, 1970). The essential elements required for

successful planning to a construction monitoring program are presented

in Table 6. This table is not a substitute for an experienced engineer,

but if used as a guide by such a man it will help to minimize the

the possibility of a monitoring program failing in its purpose.

10.30 PARAMETERS TO BE MEASURED

10.3 1 General

The most important parameters to be measured, irrespective of wall or support type, are load, pore water pressure (or

ground water level), and horizontal and vertical displacement.

Temperature measurements have special application to

excavations supported by internal bracing because of the influence that

temperature may have on bracing load. In general, direct earth pressure measurements have been unreliable except where backfill has

been placed carefully against an instrumented structure, which is

generally not possible with a deep excavation. It is preferred to

determine earth loading from load measurements in supports.

10. 32 Instruments

Types of instruments suitable for measuring the above

parameters together with advantages and limitations are given in Table

7 . Less suitable instruments, although available and occasionally

used, have not been included in this table. Schematic diagrams

illustrating instrument operation principles are given by Dunnicliff

(1970, 1971, 1972).

-169-
Table 6. Steps for planning construction monitoring programs.

1 . Define the Problem

Project type

Soil conditions

Ground water conditions

Status of nearby structures

2 . Define the Purpose of the Instrumentation

Diagnostic: To verify adequacy of design

To verify suitability of construction

techniques

To diagnose the specific nature of an

adverse event

To verify continued satisfactory performance

Predictive: To permit a prediction of behavior

later on at the same job

Legal: To establish a bapk of data for possible

use in litigation

To demonstrate a contractor’s compliance with contract requirements

Research: To advance the state-of-the-art by

providing better future design data

3. Select Monitoring Parameters

Load or stress

Pore water pressure

Earth pressure

Settlement or heave (surface or subsurface)

Horizontal movement (surface or subsurface)

Tilt

Temperature

4. Make Predictions of Behavior and Define Specific

Instrumentation Needs

Range

Accur’acy

Duration of readings

Frequency of readings

Data evaluation schedule

5. Decide Who Will Do What

Who will procure the instruments?

Who will install the instruments?

Who will monitor the instruments?

Who will maintain the instruments?

Who will process the data?

Who will analyse the data?

Who will decide on implementation?

Who will implement?

6. Select Instruments, Components, and System

General Criteria

Select each part of the system with equal care

Will it achieve objective?

Maximum simplicity

Maximum durability in installed environment

Minimum susceptibility to vandalism

Appropriate accuracy, range, longevity

Good past performance record

6. General Criteria (continued)

Minimum cost (to furnish, install, read, process)

Maximum environmental stability

Calibration can be verified after installation

Consistent with skills of available personnel

as in 5. above

Minimum interference to construction while

installing and reading

Minimum falsification of measured parameter

Selection

Refer to Table 7 and Section 10.40

7 . Determine What Factors May Influence Measured Data

(to permit an analysis of cause and effect)

IDetailed record of all construction particulars, progress

and other data

Incidence of any observed distress or unusual event

Environmental factors which may, in themselves, affect

monitored data, e.g. temperature, nearby construction

activities

8. Plan Procedures for Ensuring Reading Correctness

Consider necessary redundancy

Consider duplicate measuring system

Plan how instruments will be calibrated and corrected

for environmental effects

Consider possibility of feature to check-calibrate in place

9 . Determine a Numerical Value of Deviation from

Anticipated Performance at which the Engineer Should:

Be concerned

Press the panic button

10. Plan Instrument Layout

How many?

Where?

11. Write Instrument Procurement Specifications

12. Plan Installation

Write installation specifications

Prepare field data sheet for recording details of installatim

Examine every detail of the planned installation procedure

and think through alternative methods in the event problems

arise

Make detailed list of all materials and tools required

13. Plan Procedures Subsequent to Installation

Plan monitoring arrangements

Prepare field data sheets

Plan maintenance arrangements

Plan data processing arrangements

Plan analytical procedures

Plan remedial measures (in the event data indicates adverse

event) or other methods of implementation, and forewarn all

concerned parties

-170-
Table 7. Types of available instruments.

Parameter Instrument Advantages Limitations

Surface movement of Optical survey for settlement Simple and direct. Care required to prevent

buildings and using settlement reference pin disturbance. Requires

adjacent ground points. reliable benchmark.

(horizontal and

vertical). Optical survey for horizontal Simple and direct. R equires immovable

movement, using offsets from reference stations.

Crack opening using portable

mechanical gage.

Simple, inexpensive Care required to prevent

and direct. disturbance to reference

points.

Hose level for settlement

within buildings.

Tilt using “tidal quality

resolution” tiltmeters.

Precise. Can monitor Requires skill to read.

manv wine 8. Manpower reading costs high.

Very precise, hence Expensive and complex.

give useful data in

short monitoring

Subsurface settlemerit of adjacent

ground.

Single or multi-point rod

extensometer with mechanical readout.

Simple and reliable. Rods can hang up within

surrounding sleeves if

many anchors in one hole,

thereoy falsifying readings.

Requires manual access to

read (may create traffic

interference and danger to

reading personnel).

Magnet/reed switch

vertical pipe gage.

Anchors follow pattern Requires manual access to

of settlement without read (may create traffic interfalsification. Simple ference and danger to reading

and reliable. personnel).

Single or multi-point embedded Can be read remotely, Rods can hang up. More prone

rod extensometer with elec- without traffic inter- to malfunction, damage and

trial readout. ference. vandalism than mechanical

Subsurface horizontal Horizontal or inclined Only few required to Expensive. Must relate to datum

movement’ of adjacent extensometer. locate zone of no for absolute movements.

ground. displacement.

Inclinometer. No embedded electrical Expensive but usually cheaper

parts. Gives full than extensometers. Many

full depth profile of required to locate zone of no

movement. displacement. Must relate to

immovable reference station.

Movement of soldier Optical survey. Simple and direct. Requires reliable benchmark

piles, sheet piles, and immovable reference

walers and diaphragm stations. No readings

walls (horizontal possible at depth until

and vertical). member exposed.

Inclinometer installed on

pile or in wall.

Readings at all depths Not suitable for driven piles.

available immediately For soldier piles, must install

after pile or wall inclinometer casing inside pipe

installation. for protection during pile installation. Best to weld pipe and

install casing after pile is in

place.

-171-
Table 7. Types of available instruments. (Continued).

Parameters

Movement of tieback

anchors.

Instrument

During proof test: Dial gage,

mounted on survey tripod.

After proof test: Sleeved

unstressed telltale rod

attached to anchor.

Advantages

Simple and direct.

Simple and direct.

Limitations

Measures only relative movement

of soldier pile and anchor. Must

relate to datum for good understanding of anchor load test.

Bottom heave of

excavation.

Anchor embedded in borehole

below eventual bottom

Elevation read with probing

Simple.

Inexpensive.

Requires survey crew.

Risk of boring caving

during excavation.

Single or multi-point em- Precise. Cann connect Risk of electrical failure and

bedded rod extensometer with several sensors at damage during excavation.

electrical readout. different elevations to Bottom anchor must be deep

one anchor. Can enough to serve as benchmark.

become settlement Rage.

Load and stress in Mechanical strain gage. Inexpensive. Simple. Access problems. Many temperstruts, soldier piles, Easy to install. ature corrections required.

sheet piles, walers Minimum damage Limited accuracy. Readings are

and diaphragm walls. potential. subjective.

Vibrating wire strain gage. Remote readout. Read- Expensive. Sensitive to temperout can be automated. ature, construction damage.

Potential for accurracy Requires substa. tial skill to

and reliability. Freq- install. Risk of Nero drift. Risk

quency signal permits of corrosion if not hermetically

data transmissionwer sealed.

long distances. Gages

Electrical resistance

strain gage.

Inexpensive. Remote Sensitive to temperature,

readout. Readout can moisture, cable length

be automated. Poten- change in connections, contial for accuracy and struction damage. Requires

reliability. Most limi- substantial skill to install.

tations listed opposite Risk of zero drift.

can be overcome if

proper techniques are

Load in tieback

anchors.

Telltale load cell.

Mechanical load cell.

Inexpensive. Access problems. Cannot be

Simple. used with all proprietary

Calibrated in- place. anchor systems.

Direct reading. Expensive.

Accurate and reliable. Access problems.

Rugges and durable.

Electrical resistance strain

gage load cell.

Remote readout.

Readout can be

automated.

Expensive. Sensitive to temperature, moisture, cable length,

change in connections. Risk of

Vibrating wire strain gage

load cell.

Remote readout. Expensive. Sensitive to

Readout can be auto- temperature. Risk of zero drift.

mated. Frequency

signal permits data

transmission over

long distances.

Photelastic load cell. Inexpensive. Limited capacity. Access

problems. kequires skill to

read.

-172-
Table 7. Types of available instruments. (Continued).

Earth pressure on Hydraulic, pneumatic or

sheet piles and electrical interface stress

diaphragm walls. Cdl.

Direct method. Few successful case

records.

Backfiguring from strut load

measurements.

As for strain gages

above.

As for strain gages above.

Groundwater 01

piezometric level.

Standpipe piezometer or

wellpoint.

Simple. Reliable, Slow response time. Tubing

Long experience must be raised nearly vertical.

record. No elaborate May create traffic interference

terminal point needed. and danger to reading personnel.

Heavy liquid version Freezing problems.

available for reducing

response time and

overcoming freezing

Pneumatic piezometer. Level of terminal Must prevent humid air from

independent of tip level. entering tubing.

Rapid response.

Vibrating wire strain gage

or semi-conductor pressure

transducer piezometer. Suitable for automatic readout.

Level of terminal Expensive. Temperature corrindependent of tip ection may be required. Errors

level. Rapid response. due to zero drift could arise

High sensitivity. (although most manufacturers

have overcome major problems).

Temperature Thermistor Precise Delicate, hence susceptible to

damage. Sensitive to cable length.

Thermocouple Robust. Insensitive Less precise than thermistor,

to cable length. Avail- but premium grade can give

able in portable version *lo F.

as “surface pyrometer”

-173-
10. 33 Related Parameters

To permit a cause and effect analysis, a complete L

record of other relevant parameters should be monitored:

a. A record of depth of excavation versus time, at

close stations.

b. Time of installation of all wales, struts, and ties, ’ with preload records if any, and depth of excavation below strut or

tie at time of installation.

C . Incidences of extraordinary ground losses, ground

water behavior, observed distress, or any other unusual event.

d. Complete as-built construction plans and records,

including records of any pile driving.

e. Environmental factors which may, in themselves,

affect monitored data, e. g. temperature, nearby construction activities.

10.40 EXAMPLES

Two somewhat overly simplified situations are demonstrated

in the accompanying Figures 58 and 59.

Figure 58 is at a test section well removed from adjacent

structures that might be damaged by displacements caused by the excavation.

The objective of the test section is to determine the magnitude and

influence zone of displacements. Note that vertical and horizontal

displacements can be measured at and below the surface.

Figure 59 is an example of monitoring performed for a

building close to the excavation. Horizontal and vertical displacements

are measured at and below the surface; also, settlement points are

e s tabli s hed around the building. Concern over po s sible consolidation

requires monitoring of piezometric levels above and below the clay.

-174-
.4H I

--k @2

.6Hrt.6H

b A

0

SECTION A-A

LEGEND

1.q -

v

multi-point subsurface settlement system

(e. g. magnet/reed switch gage)

2.W (:

V

, (h - multi-point subsurface settlement and

horizontal movement system (e. g.

inclinometer with magnet/reed switch

gage )

3 : A,- .surface settlement point

4.+ & - heavepoint (use when underlain by

cohesive soil; otherwise optional)

Figure 58. Example of instrumentation for measurement of displacement at

test section.
LEGEND

1.e 6-

z;+( -

settlement point at surface

inclinometer (for measurement of

subsurface horizontal movements)

3*&fh,6v- inclinometer with multi-point

subsurface settlement system

TYPICAL ItiNoMEti TO 0’

BEIBW CUT IN ROCK

SECTION A-A

4.8 - pie zome te r ’

5.4 - tilt meter on building

6. Monitor horirontal.~vements at the face of

the diaphragm wall by optical surfkey. Monitor

settlement of diaphragm wall by optical survey.

Figure 59. Example of instrumentation adjacent to building

and diaphragm wall.
lo. 50 RECOMMENDED METHODS OF MEASUREMENT

10.51 General

The seiection of the measurement method depends upon case

specific factors, and no general rules can be made. However,

Table 7 provides basic information to assist in the selection process.

The most difficult task, and the task with the poorest

success record, is measurement of load in supporting members.

The following sections therefore provide more detailed guidelines

to assist in selecting a method for monitoring load in braced and

tied-back excavations.

10. 52 Strut Loads ,.

10. 52. 1 Instrument Type

For monitoring strut loads, strain gages are

preferred rather than load cells, primarily because inclusion of a

load cell will tend to create non-typical loading conditions and will

interfere with the contractor’s work. Strain gages permit measurement of bending stresses, whereas a load cell does not.

Since the most expensive feature of an instrumentation program is often the disruption to construction activities,

remote readout is a desirable feature, and vibrating wire and electrical

resistance strain gages are the preferred instruments. Selection

between the two gage types should be based on the experience and skills

of available personnel, rather than on any quality inherent to one or

the other type.

In general, a backup system should be established,

using mechanical strain gages, although of course, their use is limited

by access restrictions. Each gage is discussed below, and advantages

and limitations are sur-nmarized in Table 7. In general, an accuracy

of no better than f 10 percent of design load can be attained, and this

accuracy is usually adequate.

10, 52. 2 Vibrating Wire Strain Gages

O’Rourke and Cording (1974b) provide a detailed

and up-to-date technical quide in use of vibrating wire gages. Gages that

are perfectly temperature compensated (equal thermal coefficients for

gage and structural member) will provide optimum accuracy, and the

vibrating wire itself should be as close as possible to the surface of

the strut.

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The problem of thermaL response, in particular caused

by temperature differentiaL between gage and strut, and the problem

of zero drift remain the most severe. Estimates of long term zero

drift are necessary to judge the reliability of measurement. Individdual gages set to different frequencies should be mounted on unloaded

sections of strut steel and monitored throughout the life of the project.

No-load gage readings should be taken after strut removal and then

compared with the initial no-load readings taken before strut installat ion. Temperatures should be measured at the time of each recording,

and readings should be appropriately corrected on the basis of temperature changes between the initial and subsequent values.

10. 52.3 Electrical Resistance Strain Gages

Electrical resistance (SR4) strain gages have

been used very successfully to monitor strain in laboratories, but their

use in field measurements has often yielded poor results, Largely because

of the irie%perience of personnel undertaking the monitoring

program. Since an electrical property of the wire rather than a mechanical property is being measured, it is important that these personnel

have experience in field electronics, and such personnel are rare

among geotechnical firms.

Dunnicliff (1975) elaborates on key factors:

gage selection and installation; sensor configuration; wiring; amplification

and readout equipment. The task requires attention to many minute

details, and wherever possible the gage installation work should be

performed in a controlled labor environment prior to installation of the

structural member.

When using electrical resistance gages zero

drift tests should be made, as described above for vibrating wire gages.

Weldable gages have not yet been used on a

widespread basis, perhaps because they are more costly than bonded

gages, but the ease of installation and hermetic insulation of the gages

are great advantages. Their sensitivity, although less than that of

bonded gages, will normally be adequate.

10. 52.4 Mechanical Strain Gages

Although less accurate than the gages discussed

above, use of mechanical gages provides valuable backup data provided

three rules are adhered to. First, the Demec rather than the Whittemore type should be used (Schmidt and Dunnicliff, 1974). Second,

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gage points must be rigidly attached,to the structural member, by

drilling into the member or by welding. Third, proper temperature

correction procedures must be used, taking into account thermal

coefficients of strut, gage, and gage reference bar.

10. 52. 5 Temperature Correction

Strain-gages are used to measure strain, which

then has to be converted to stress and load by using a value for modulus.

However, temperature change also causes strain, and any such strain

must be subtracted from measured strain before the conversion to

stress is made. In the absence of complete temperature compensation

(possible using resistance gages, approachable using certain vibrating

wire gages, impossible using mechanical gages) temperature must

be measured and thermal strain accounted for. Temperature variations

always contribute to inaccuracy, and any effort to *minimize temperature variation is worthwhile.

10.53 Tieback Loads

Load cells have been more commonly used to monitor

tieback loads than have strain gages. Strain gages are inapplicable for

use on stranded wire tendons since no convenient method is available

for attaching the gages. Furthermore, a single load cell with a central

hole can surround an entire group of tendons. It is possible to attach

strain gages to steel rods, although the rate of gage attrition is

generally high. Advantages and limitations of the five basic types of

load cell are given in Table 7. Portable ’ calibrated’ hydraulic

jacks have also been used, but measurement error may be up to 30

percent.

Selection of cell type depends on the factors given in

Table 6, on past personal experience of the engineers executing

the monitoring program, and load cell availability. Dunnicliff (1975)

describes an inexpensive home -made “telltale load cell” capable of

monitoring 150 kip loads with an accuracy of &5 kips, and in view of its

simplicity and economy, it seem-s logical to use this method wherever

feasible.

A backup system is desirable, although less necessary

than for strut load monitoring. This can be done in one of two ways,

although neither way is always practical. First, one or more Load I

cells, on a special test frame or rod, can be retained on the site in

-179-
a loaded condition. Readings can be examined for drift, and the cells

can be checked periodically by calibrating in the normal way. Second,

it may be possible to check and calibrate selected cells in place by

pulling a tie to unload the cell and then releasing the tie. For this test,

it is necessary to connect a calibrated load cell, in addition to the

stressing jack, in series with the cell under test.

10.60 CONTRACTING FOR INSTRUMENTATION

Table 8 presents the three basic contracting methods for

furnishing and installing instrumentation.

Sophisticated instrumentation should not be included as a bid

item in the prime contract, as the task requires professional skill and

dedication, usually unobtainable if the prime contractor shops between

’ specialist’ subcontractors. A separate contract between owner and a

specialist firm is suitable for sophisticated instrumentation provided

the specialist and prime contractor’s work areas do not overlap. I f

sophisticated instrumentation is to be installed within the prime contractor’s work area the only viable method is use of a cost plus item

in the prime contract. The essential elements of this procedure are:

a. Work which is within the capability of the average prime

contractor is bid in the normal way.

b. The prime contract specification defines the nature of

special instrumentation work. This work is included in the bid

schedule as an allowance item, with an estimate of cost, and the prime

contract bidder bids a markup, carrying forward the marked up total

to the amount column. The estimate is not an upset.

c. The owner selects an instrumentation specialist firm, using

normal professional procedures for engagement of engineering services,

and agrees on a basis of payment for the firm’s services.

d. The owner instructs the prime contractor to enter into a

subcontract with the specialist firm and to pay the firm in accordance

with the agreed basis. The contractor is reimbursed by the owner at

cost plus the bid markup.

This procedure requires a clear and thorough prime contract

specification. It also requires close coordination, cooperation, and trust

between owner and specialist to ensure that a.1.L expenditures are necessary,

-180-
Table 8. Contracting for instrumentation.

Type of Location of

Instrumentation Instrurrients Example

Contract for Furnish and Install

Separate cost Plus

, Bid Item- Contract Item in

in Prime With Prime

Contract Specialists Contract

Simple Outside or with- W ellpoint s Suitable Not Not

in contractor’s Optical survey necessary necessary

work area.

Sophisticated Outside contrac- Inclinometer with Not Suitable Suitable

tar’s work area top on sidewalk suitable

Within contrac -

tar’s work area

Strain gages on

struts

Load cells on

tiebacks

Not

suitable

Not

suitable

Suitable
thereby keeping costs to a minimum. If handled properly,

it results in cooperation, flexibility to accommodate changes as the

work proceeds, and a successful monitoring program at minimum cost

to the owner.

10.70 THE KEY TO SUCCESSFUL CONSTRUCTION MONITORING

The key to successful construction monitoring may be stated as:

Have a valid reason for monitoring, and perform complete

and logi.al planning (Table 6).

b. Select the most appropriate parameters and instruments

(Table 7).

c. Establish workable contractual arrangements with experienced

personnel who have a full understanding of the monitoring objective, and

who have the patience and desire to ensure the success of the program

(Table 8).

d. Achieve cooperation between all parties in the field. Cooperation can best be gained by explaining to the contractor’s personnel the

purpose of the program, gaining his respect by performing top quality

work, then throughout the program being responsive to the effects of

the.program on him/her and working with him/her to minimize any

adverse effects.

e. Observe and record all relevant construction data (Section

10.33).

f. Make use of the data in the way intended.

-182-
BIBLIOGRAPHY

Abbreviations:

ASCE American Society of Civil Engineers

ICSMFE International Conference on Soil Mechanics

and Foundation Engineering

ECSMFE European Conference on Soil Mechanics

and Foundation Engineering

JSMFD Journal Soil Mechanics and Foundation

Division

GTED Journa.1 Geotechnical Engineering Division

SGDMEP Symposium on Grouts and Drilling Muds

in Engineering Practice

-183-
BIBLIOGRAPHY

ADLER, F., Vereinfachte Lgngenberechnung F& Erkander (Simplified

Method for Length Calculations of Tie-Backs), Schweitzerische

Bauzeitung, Vol. 91, No. 5, pp. 103-105, 1973.

AMERICAN Institute of Timber Construction, Timber Construction

Manual, Sedond Edition, Wiley, 1974,

AMERICAN Petroleum Institute, Oil Well Drilling Fluid Materials,

API Specification 13A, Sixth Edition, January 1974.

AMERICAN Petro1eu.m Institute, Standard Procedure for Testing

Drilling Fluids, API RP 13B, Fifth Edition, February 1974. -

AMERICAN Society of Civil Engineers, JSMFD, Proceedings, Chemical

Grouting, Vol. 83, SM4, Paper No. 1426, pp. l-106, November 1957.

AMERICAN Society of ‘Civil Engineers, JSMFD, Proceedings,

Bibliography on Chemical Grouting, Vol. 92, SM6, pp. 39-66,

November 1966.

ANDERSEN, Paul, Earth Anchors and Tieback Installations, 18th Annual

Soil Mechanics and Foundation Engineering Conference, University of

Minnesota, March 26, 1970.

ANDRX, W., Kunzl, W. and Rojek, R.. , Bohrpfahlwand Fiir die

Allianz-Neubauten in Stuttgart - Entwurf und Messungen (Secant Wall

for the New Allianz Insurance Building in Stuttgart - Design and

Measurements), Bie Bautechnik, Vol. 50, No. 8, pp. 258-264, 1973.

ARMENTO, William, Design and Construction of Deep Retained

Excavations, ASCE /SEAONC, Continuing Education Seminars,

November 1970.

ARMENTO, W. J., Criteria for Lateral Pressures for Braced Cuts,

ASCE Specialty Conference on Performance of Earth and Earth

Supported Structures, Purdue University, pp. 1283-1302, June 1972,

ARMENTO, W. J., Coffer dam for BARTD Embarcadero Subway

Station, ASCE, JSMFD, Vol. 99, SMlO, pp. 727-744, 1973.

-184-
BACHY Enterprise, Brochure on Firm Capability,; France 1970.

BADAPPANAVAR, M. G., Chemical Grouting in Pervious and Rock

Foundations*, Journal of the Institute of Civil Engineers (India),

Vol. 54, Part C13, pp. 110-115, 1974.

BALL, Donald G., Prudential Center Foundations, Journal of Boston

Society of Civil Engineers, July 1962.

BARES, F. A., Use of Pali Radice in the Solution of Difficult

Foundation Problems, ASCE/ The Engineering Institue of Canada/

Roads & Transportation Association of Canada, Joint Transportation Meeting, Montreal, July 1974.

BARES, F. A., Use of Pali Radice (Root Piles), Open Cut Seminar,

Metropolitan Section, ASCE, New York, February 1975.

BARES, F. A., Personal Communication, 1975

BARLA, G., and Mascardi, C. , High Anchored Wall in Genoa,

Conference on Diaphragm Walls and Anchorages, Institution

of Civil Engineers, London, pp. 101-106, September 1974

BARTHEL, H., Verbesserung der Gebirgsfestigkeit durch Chemische

Injektionen (Improvement of Rock Mass Strength by Chemical

Injections), Bergbau-Wissenschaft, Vol. 17, No. 8, pp. 281, 285,

1970.

BASSETT, Discussion, Conference Organized by Institution of

Civil Engineers, London, 1970.

BASTIAN, S., and Grueno, M., Erweiterung des Bereiches der

Zementiverungsarbeiten auf Grund von Neuen Eigenschaften der

Zement Suspensionen (Increasing the Range of Soils Suitable for

Cement Injection Through New Properties of the Cement Suspension),

Bergbau-Wissenschaft, Vol. 17, No 8, p. 303, 1970.

BAUER, M., Personal Communication, 1974.

BAUER Schrobenhausen, Company Catalogue

BAUERNFEIND, P. and Hilmer, K., Neue Erkenntnisse aus

Sohldruck-und Erddruck-Messungen bei der U-Bahn Nurnberg

(New Information from Earth Pressure Measurements on the

Subway Tunnel in Nurenberg) Die Bautechnik, Vol. 51, No. 8,

pp. 523-259, 1974.

-185-
DEHRENDT, J; , Die Schtitzwandbauweise beim U-Bahnbau in Kiiln

(The Slurry Trench Walt Construction at the Subway in Cologne)

Der Bauingenieur, Vol. 45, No. 4, pp. 121-126, 1970.

BETHEL, R., Underpinning a Landmark,, Civil Engineering, ASCE,

Vol. 38, No. 8, pp. 67-68, August 1968.

BJERRUM, L., and Eide, O., Stability Iof Strutted Excavations in Clay,

Geotechnique, Vol. 6, pp. 32-47, 1956.

BJERRUM, L., Frimann-Clausen, C. J., and Duncan, J. M., Earth

Pressures on Flexible Structures - A State-of-the-Art Report,

General Report, 5th ECSMFE, Madrid, pp. 169-196, April 1972.

BJERRUM, L., and Johannessen, I. J., Reduction of Negative Skin

Friction on Steel Piles to Rock, 7th ICSMFE, Vol. 2, pp. 27-34,

1969.

BOGUTH, K., Vibrationsprobleme beim R.ammen .-

 und Ziehen (Vibration

Problems by Driving and Retrieval), Tiefbau, Vol. 15, No. 4,

pp. 345-347, 1973.

BONARES, R., Baugrubensicherung und Sohlplatte des Bfg-Hochhauses

in Frankfurt (Excavation Support and Base Stab for the “Bank of

Gemeinwirtschaft” in Frankfurt), Der Bauingenieur, Vol. 49, No. 6,

pp. 214-218, 1974.

BONZEL, J. and Dahms, J., Uber den Einfluss des Zements und der

Eigenschaften der Zementsuspensionen auf die Injizierbarkeit in

Lockergesteinsboden (The Influence of the Cement and the Suspension

Characteristics on the Groutability of Soils), Beton Herstellung und

Verwendung, vol. 22, No. 4, pp. 156-166, 1972.

BOOTH, W. S., Tiebacks in Soil for Unobstructed Deep Excavation,

Civil Engineering, ASCE, pp. 46-49, September 1966.

BOTTKE, H., Geologis the und Sedimentpetrographis the Grundlagen

der Stabilizierung und Verfestigung von Locker gesteinen (Geological -

and Sediment-Petrographical Principles for t,he Stabilization and

Consolidation of Loose Soils), Bergbau-Wissenschaft, Vol. 17,

No. 8, pp. 278-281, 1972.

-186-
BOUTSMA, K., and Horvat, E., Soil Mechanical Aspects of the Metro

Construction in Rotterdam, Proc., 7th ICSMFE, Mexico City,

Vol. 2, pp. 417-422, 1969.

BRAIN, M. G., Tunnel Interface Crossing Solved by Specialized

Consolidation, Ground Engineering, Vol. 5, No. 5, September 1972.

BRANDL, H., Bemessungsdiagramme fur einfach verankerte

Spundwsnde (Dimensioning Diagrams for Simply Anchored Sheet

Pile Walls), Oesterreichische Ingenieur-Zeitschuft, Vol.. 15,

No. 6, pp. 161-169, 1972.

BRAUN, B., Arbeiten in Hamburg, (Work in Hamburg), Unser Betrieb,

No. 3, Deilmann-Haniel, Dortmund-Kurl, 1969.

BRAUN, B., Die bisher groesste, allein durch Bodenvereisung

gesicherte Baugrube der Welt, (The Largest Present Day Safe

Excavation in the World Wholly Supported by Frozen Ground),

Unser Betrieb, No. 6; Deilmann-Hanie.L, Dortmund-Kurl, 1970.

BRAUN, B., Ground Freezing for Tunneling in Water-Bearing Soil

at Dortmund, Germany, Tunnels and Tunneling, Vol. 4, No. 1,

pp. 27-32, January 1972.

BRAUN, B. and Macchi A., Ground Freezing Techniques at Salerno,

Tunnels and Tunneling, vol. 6, No. 2, pp. 81-85, 87, 89,

March - April 1974.

BRAUN, W. M., The Art of Tunneling in Vienna, Tunnels and Tunneling,

Vol. 6, pp. 29-32, May, 1974.

BRETH, H. and Romberg, W., Messungen an einer verankerten Wand

(Measurements on a Tied-Back Wall), Mittei.lungen der Versuchsanstalt fir Bodenmech und Grundbau der Technische Hochschule

Damstadt, No. 11, 1972-73.

BRETH, H. and Wanoschek, H. R., Steifenkraftmes sungen in einer

durch’pfahlwande gesicherten Tiefbahngrube im Frankfurter Ton

(Strut Load Measurements in a Subway Excavation in Frankfurt

Clay Supported by Secant Walls), Der Bauingenieur, Vol. 44,

No. 7, pp. 240-243, 1969.

BRISKE, R., Anvendung von Druckumlagerungen bei Baugrubenumschliessungen (The Use of Pressure Redistribution in Supported

Excavations), Die Bautechnik, Vol. 35, No. 6 and 7, pp. 242-244

and 279-281, 1958.

-187-
BRISKE, R., Erddr uckumlagerungen bei abgesteiften Tragerbohtwanden

(Earth Pressure Redistribution at Strutted Soldier Pile Walls),

Die Bautechnik, Vol. 48, No. 8, pp. 254-259, 1971.

BRISKE, R. and Pirlet, E., Messungen iiber die Beanspruchung des

Baugrubenverbaues. der Kolner U-Bahn (Strain Measurements on

the Excavation Wall in the Cologne Subway), Die Bautecknik,

Vol. 45, No. 9, pp. 290-299, 1968.

BROMS, B. B., Design of Laterally Loaded Piles, ASCE, JSMFD,

Vol. 91, SM3, pp. 79-99, 1965.

BROMS, B., Swedish Tieback Systems for Sheet Pile Wafts, Swedish

Geotechnical Institute, Report No. 32, Reprinted from Proceedings

of Third Budapest Conference on Soil Mechanics and Foundation

Engineering, Budapest, Vol. 1, 13 pages, 1968.

BROMS, B. B., and Bennemark, H., Stability of CLay at Vertical

Openings , ASCE, JSMFD, Vol. 93, SM 1, pp. 71-94, 1967.

BROMS, B. B. and Bennemark, H., Stability of Cohesive Soils Behind

Vertical Ooenings in Sheet Pile Walls. Contributions to the 3rd

Budapest Conference on Soil Mechanics and Foundation Engineering,

Budapest, 1968.

BROMS, B. B., Bjerke, H., Extrusion of Soft Clay Through a

Retaining Wall., Canadian Geotechnical Journal, Vol. 10, No. 1,

pp. 103-109, 1973.

BRoOKER, E. W., and Ireland, H. O., Earth Pressures at Rest

Related to Stress History, Canadian Geotechnical Journal,

vol. 2, No. 1, pp. 1-15, 1965.

BROWN, D. R. and Warner J., Compaction Grouting, ASCE, JSMFD,

Vol. 99, SM8, pp. 589-601, August 1973.

BURLAND, J. B., Recent Research on Excavations and Diaphragm

Walls, Paper presented at BRS Symposium, 1974.

, Grouts and Drilling Muds. in Engineering

Practice, Butterworths, London, 1963.

BUTTNER, J. H., Herstellung von Injektionssohlen zur Abdichtung

von Baugruben (Construction of Horizontal Grouted Surfaces for

Waterproofing of Excavations), Tiefbau, Vol. 15, No. 5, pp. 450-

455, 1973.

-188-
BtiTTNER, J. H., Neue Technik zur Herstellung dtinner horizontaler

Injektionssohlen (New Method for P.Lacing Thin Horizontal Grouting

Shells), Die Bautechnik, Vol. 51, No. 2,-pp. 62-65, 1974.

-18%
CARSON, A. B., Foundation Construction, McGraw Hill, New York,

pp. 116-147, 1965.

CASHMAN, P. M. and Haws, E. T., Control of Groundwater by

Water Lowering, Proceedings of a Conference Organized by the

Institution of Civil Engineers in London, London, pp. 23-32, 1970.

CHAPMAN, K. R., Cording, E. J., and Schnabel, H. Jr., Perfor.mance

of a Braced Excavation in Granular and Cohesive Soils, ASCE

Specialty Conference on Performance of Earth and Earth-Supported

Structures, Purdue University, Vol. III, pp. 271-293, 1972.

CHARGE, J., The Raising of the Old Wellington Inn and S’inclairs

Oyster Bar, The Structural Engineer, Vol. 50, No. 12, pp. 483-494,

December 1972.

CHEN, W. F. and Rosenfarb, J. L., Limit Analysis Solutions of Earth

Pressure Problems, Soils and Foundations, Vol. 13, No. 4,

pp. 45-60, December 1973.

CHRISTIAN, J. T., Personal Communication, 1971.

CHRISTIAN, J. T. and Wong, I. H., Errors in Simulating Excavation

in Elastic Media by Finite Elements, Soils and Foundations, Vol. 13,

No. 1, pp. l-10, March 1973.

CLOUGH, G. W., Performance of Tie-Back Support Systems, ASCE

Specialty Conference on Performance of Earth and Earth-Supported

Structures, Purdue University, Vol. III, pp. 259-264, June 1972.

CLOUGH, G. W., Deep Excavations and Retaining Structures, Paper

to be Presented at Lehigh Conference on Foundations for Tall

Buildings, August 1975.

CLOUGH, G. W. and Tsui, Y., Performance of Tied-Back Walls in

Clay, ASCE, JGED, Vol. 100, GT 12, pp. 1259-1273, 1974.

CLOUGH, G. W. ,.Weber, P. R., and Lamont, J. Jr., Design and

Observation of a Tied-Back Wall, ASCE Specialty Conference on

Performance of Earth and Earth-Supported Structures, Purdue

University, Vol. I, Part 2, pp. 1367-1389, June 1972.

-190-
COLAS DES FR.ANCS, E., Prefasif Prefabricated Diaphragm Walls,

Conference on Diaphragm Walls and Anchorages, Institution of

Civil Engineers, London, pp. 63-70, September 1974.

COLE, K. W. and Burland, J. B., Observation of Retaining Wall

Movements Associated with a Large Excavation, 5th ECSMFE,

Madrid, Vol. I, pp. 445-453, April 1972.

CONSEDINE, R. L., Hi-Rise Costs Cut by Excavating After Subgrade

Walls in Place, Engineering Contract Record, Vol. 85, No. 9,

pp. 48-49, 1972.

CONSEDINE, R. L., Slurry Trench Method Used for Olympic Stadium

Retaining Walls, Engineering Contract Record, Vol. 86, No. 8,

pp. 72-73, 1973.

CONSTRUCTION Methods, Long, Buried Tiebacks Leave Cofferdam

Unobstructed, pp. 130-135, May 1966.

CONSULTING Engineer, The, Ground Anchors, Panel Presentation on

Ground Anchors, pp. l-39, May 1970.

CORBETT, B. O., Davies, R. V., and Langford, A. D., A Load

Bearing Diaphragm Wall at Kensington and Chelsea Town Hall,

London, Conference on Diaphragm Wa’lls and Anchorages,

Institution of Civil Engineers, London, pp. l-10, September 1974.

CORBETT, B. 0. and Stroud, M. A., Temporary Retaining Wall

Constructed by Berlinoise System at?entre Beaubourg, Paris,

Conference on Diaphragm Walls and Anchorages, Institution of

Civil Engineers, London, pp. 77-83, September 1974.

COYLE, H. M., Bartoskewitz, R. E., Milberger, L. J., and

Butler, H. D. , Fie.ld Measurement of Lateral Earth Pressures on

a Cantilever Retaining Wall, Paper prepared for presentation at

the 53rd Annual Meeting of the Highway Research Board, January

1974.

CUNNINGHAM, J. A. and Fernandez, J. I., Performance of Two

Slurry Wall Systems in Chicago, ASCE Specialty Conference on

Performance of Earth and Earth-Supported Structures, Purdue

University, Vol. I, Part 2, pp. 1425-1449, June 1972.

-191-
CZARNOTA-BOJARSKI, R. and Rymsza, B., Influence of Neighboring

Structures on Deex, Foundations, 8th ICSMFE, Moscow, Vol. 2.1,

pp. 61-66, August 1973.

-192-
D’APPOLONIA, D. J. , Effects of Foundation Construction on Nearby

Structures, Proceedings of the 4th Panamerican Conference on Soil

Mechanics and Foundation Engineering, Vol. I, pp. 189-236, 1971.

D’APPOLONIA, D. J., D’Appolonia, E., and Namy, E., Proceedings

of Workshop on Cut and Cover Tunneling: Precast and Cast-in

Place Diaphragm Walls Using Slurry Trench Techniques, Report

prepared for FHA Office of Research and Development, Report No.

FHWA-RD-74-57, Washington D. C., January 1974.

DARR.AGH, R.. D., Design and Construction of Deep Retained Excavations, ASCE/SEAONC Continuing Education Seminars, Lecture dated

November 23, 1970. I ,

DAVIS, E. H,, and Christian, J. T. ; Bearing Capacity of Anisotropic

Cohesive Soil; ASCE JSMFD, Vol. 97, No. SM5, pp. 753-769,197l.

DEILMANN-HANIEL, GmbH, Soil Stabilization by Freezing, Undated

Procedures, Dortmund-Assein, Germany.

DEMPSEY, J. A. and Moller K., Grouting in Ground Engineering,

Proceedings Conference on Ground Engineering - London, Paper 1,

pp. 3-10, 1970.

DEUTSCHE Normen, DIN 1054 (German Codes, DIN 1054), BeuthVertrieb, GmbH, Berlin 30, November 1969.

DEUTSCHE Normen, DIN 4014 Bohrpfahle (German Codes, DIN 4014,

Angered Piles) Beuth-Vertrieb, GmbH, Berlin 30, November 1969.

DEUTSCHE Normen, DIN 4017, Baugrund (German Codes, DIN 4017,

Foundations), Beuth-Vertrieb, GmbH, Berlin, March 1965.

DEUTSCHE, Normen, DIN 4026, Rammfahle (German Codes, DIN 4026,

Driven Piles) Beuth-Vertrieb GmbH, Berlin 30, July 1968.

DEUTSCHE Normen, DIN 4033 (German Codes, DIN 4033) BeuthVertrieb GmbH, Berlin 30.

DEUTSCHE Normen, DIN 4123, Gebaudesicherung im Bereich von

Ausschachtungen, Grundungen und Unterfangungen (German Codes

DIN 4123, Underpinning of Buildings Near Excavations and Foundation Pits), Beuth-Vertrieb GmbH, Berlin 30, May 1972.

DEUTSCHE Normen, DIN 4124, Baugruben und Graben (German Codes,

DIN 4124, Excavations and Pits), Beuth-Vertrieb GmbH, Berlin 30,

January 1972.

-193-
DEUTSCHE Normen, DIN 4125, Verpressanker fur vorubergehende

Zwecke in Lockergestein (German Codes, DIN ‘4125, Grouted

Temporary Soil Tie-Backs), Beuth-Vertrieb GmbH, Berlin 30,

1972.

DEUTSCHE , Stahlbau-Verband, Stahlbau. Vorsehriften, Normen

Profile. (Steel Construction Rules, Codes, Profiles). Deutscheq

Stahlbau-Verband, Cologne, 1974.

DIBIAGIO, E. and Myrvoll, F., Full Scale Field Tests of a Subway

Trench Excavation in Soft Clay, 5th ECSMFE, Madrid, Vol. I,

pp. 461-471, April 1972. I /

DIBIAGIO, E. and Roti, J. A., Earth Pressure Measurements on a

Braced Slurry Trench Wall in Soft CLay, 5th ECSMFE, Madrid,

Vol. I, pp. 473-483, 1972.

DIDLAUKIES, J., Vorschactebau mit Hilfe des Gefrierverfahrens am

Beispiel des Schactes Kurl 4, Unser Betrieb, No. 8, DeilmannHaniel, Dortmund-Kurl., West Germany, 1971.

DIETRICH, M., Chase, B., and Teul, W., Tieback System Permits

Uncluttered Excavation, Foundation Facts, Vol. VII, No. 1, pp. 3-7,

1971.

DONOLO, Louis, Mur de Soutenement Special, L’Ingenieur, Montreal,

November 1971.

DROSSEL, M. R., Drill Steel Tubes Act as Temporary Tieback Casing,

Construction Methods and Equipment, Vol. 55, No. 8, pp. 50-51,

August 1973.

DUMONT-VILLARES, A., The Underpinning of the 26-Story Companhia

Paulista De Seguras, Building, S&o Pau.Lo, Brazil, Geotechnique,

Vol. 6, pp. 1-14, March 1956.

DUNCAN, J. M., Use of the Finite Element Method in Analysis of

Soil Structure bteraction, ASCE Specialty Conference on Performance of Earth and Earth Supported Structures, Purdue University,

Vol. III, pp. 265-269, June 1972.

-194-
DUNNICLIFF, C. J. , Equipment and Methods for Measurement of

Displa-cements and Settlement, Lecture Notes, Met Section

AXE, Seminar on Field observations in Foundation Desig,n and

Construction, New York, April. (Reprinted in Highway Focus,

U. S. Department of Transportation, F.H. W.A., Vol. 4, No. 2,

June 1972, pp. 105-137), 1970.

DUNNICLIFF, C. J., Equipment for Field Deformation Measurements, Proc. 4th Panamerican Conf. on Soil Mechanics and

Foundation Engineering, San Juan, Vol. 2, pp. 319-332, 1971.

DUNNICLIFF, C. J., Schematic Arrangements of Various Tyypes of

Soil Mechanics and Rock Mechanics Measuring Instruments,

Highway Focus, U. S. Department of Transportation, F. H. W.A.,

Vol. 4, No. 2, June, pp. 139-143, 1972.

DUNNICLIFF, C. J., Instrumentation of Open Cut Construction,

Lecture Notes, Met. ‘Section ASCE, Seminar on Open Cut

Construction, New York, February, 1975.

DUNNICLIFF; C. J., and Schmidt, B., An Engineering Approach

to Monitoring the Performance of SoftGnd Tunnels Duris

Construction, Proc. Rapid Excavation and Tunneling Conference,

San Francisco, Vol.’ 1, pp. 377-396, 1974.

DUNS, C. S. and Butterfield, R., Flexible Buried Cylinders,

Part I, Static Response, Part II, Dynamic Response, and I

Gt III, Buckling Behavior,xternational Journal of Rock

Mechanics and Mining Sciences, Volume 8, No. 6, November,

1971.

-195-
EC1 Sotetanche, Brochure of Firm Capabilities, 1971.

EDELING, H.and Schulz, W., Die neue Osterreichische Tunnelbauweise (The New Austrian Tunnel Driving Method, Der

Bauingenieur, Vol. 47, No. 10, 11 pages, 1972.

EDER., P. and RGmmeli, H., Die erste vorfabrizierte Schlitzwand in

der Schweiz (The First Precast Slurry Trench Wall in Switzerland),

Schweitzerisch Bauzeitung, Vol. 91, No. 28, pp. 689-691, 1973.

EGGER, P., Die Grundung der Kraftwerkserweiterung BordeauxAmbes (Foundations for the Extension of the Bourdeaux - Ambes

Power Station), Gsterreichische Ingenieur - Zeitschrift, Vo.L 15,

No. 7, pp. 203-209, 1972 a.

EGGER, P. Influence of Wall Stiffness and Anchor Prestressing on

Earth Pressure Distribution, 5th ECSMFE, Madrid, Vol. I,

pp. 259-264, April 197213.

EIDE, O., Aas, G., and Josang, T., Special Application of Cast-inPlace Walls for Tunnels in Soft CLay in Oslo, 5th ECSMFE, Madrid,

Vol. 1, pp. 485-498; 1972.

ELLINGER, M., Die Untergrundbahnbau in Wien (The Vienna Subway),

Osterreichische Ingenieur--Zeitschrift, Vol. 15, No. 9, pp. 257-

262, 1972.

ELSON, W. K., An Experimental Investigation of Slurry Trenches,

Geotechnique, Vol. 18, No. 1, pp. 37-49, 1968.

ENDO, K., Artificial Soil Freezing Method for Subway Construction,

Civil Engineering in Japan, Vol. 8, pp. 103-117, 1969.

ENDO, K., Problem Factors and Exe cution of Under g’r ound Excavation

Methods for Bad and Soft Ground, Proceedings of Sixth National

Tunnel Symposium, Tokyo, Japan, September 14-16, 1970.

, Unbraced Walls Hold Up a Street , Engineering

News Record p. 43, November 8, 1962

Soil-Anchored Tiebacks Aid Deep Excavation.

Engineering News ‘Record, pp. 34-35, August 10, 1967.

-196-
, Foundation for Tallest Towers: Water Out,

Trains In, Engineering News Record, pp. 30-32, October 31, 1968.

Swedes Quick Freeze 39 Ft. Tunnel Section,

Engineering News Rk cord, P. 13, October 29, 1970.

Massive Underpinning Job Paves Way for

Construction of D. d. Subway, Engineering News Record, p. 37,

February 4, 1971a.

, Brussels Subway Work Runs Obstacle Course,

Engineering News Record, pp. 17-18, August 5, 1971b.

, Gr.out, Slurry Protect Neighboring Foundations,

Engineering News Record, March 9,. 1972a.

, Small-Diameter Piles Protect Tunnel, Builgings,

Engineering News Record, pp. 22-23, April 20, 1972b

Engineer Subway, ini

Europes Tallest Building Encases an Operating

News Record, pp. 22-23, July 15, 1972~.

, Deep Excavation in Bad Geology Secured by 60

Miles of Tiebacks, Engineering News Record, p. 31, January 11, 1973a.

, Special Formwork Casts Two-LeveL Arch

Station Under Street, Engineering News Record, pp. 20-21,

July 26, 1973b.

, Tiebacks Driven In Place Keep Excavation

Open, Engineering News Record, p. 25, August 9, 1973~.

, Rio Subway Builders Beat High Water Table,

Engineering News Record, Vol. 192, No. 20, p. 54, 1974a.

, New Underpinning System Solves Explosive

Problem, Engineering News Record, p. 21, July 21, 1974b.

ERLER, K., Keilformige Zupanker i.m Baugrund bei Senkr e cht -und

Schragzug(Wedge Shaped Tie-Backs in Soil. by Vertical and Oblique

Pull), Bauplanung - Baute chnik, Vol. 27, No. 12, pp. 586-591,

1973.

-197-
ETHERIDGE, D. C., Tie-Backs Unclutter Deep Fpundation, Construction Methods, pp. 66-71, September 1966.

-198-
FEBESH, M., Earth Tie Backs, Draft of Paper Presented at New

York Metropolitan Section, ASCE, undated.

FEBESH, M., Personal Communication, 1975.

FEDDERSEN, J., Verpressanker im Lockergestein (Pressure Grouted

Tie Backs in Soil), Der Bauingenieur, Vol. 49, No. 8, pp. 302-

310, 1974. -

FEDERATION of Piling Spe cia.Lis ts , Specifi,cation for Cast-in-PLace

Concrete Diaphragm Walling, Ground Engineering, Vol. 6, No. 4,

July 1973.

FENOUX, G. Y., LaMise en Precontrainte Des Tirants (Prestressing of Anchor Rods), Travaux, No. 449-450, pp. 33-43, AugustSepte.mber 1972.

FENOUX, G. Y., La Paroi Pr’Lfabriqude, Ses Applications (The

Precast Wall, Its Application), Annales de L’Institut Technique

du Batiment et des Travaux Publics, No. 313, pp. 198-219,

Januar y 19 74.

FENOUX, G. )I, and Milanise, S., hlications Re/cents de La Paroi

Prefabriquee Panasol (Recent Appliczions of the Precats “PanasoL”

Wall), Travaux, K24-31, December 1971.

FERNANDEZ-RENAU, L. F., Discussion of Slurry Trench Analysis,

5th ECSMFE, Madrid, Vol. 2 pp. 366-373, 1972.

FREI, R. and Graf, W., Schlitzwknde mit Vorfabrizierten Betonteilen

(Slurry Trench Walls with Precast Concrete Elements, Schweitzer is the Bauzeitung, vol. 91, No. 28, pp. 691-692, 1973.

FLAATE, K. S., Stresses and - -Movemen in Connection with Braced

Cuts in Sand and Clay, Ph. D. e- Thesis, University of Illinois,

Urbana, Illinois, (1966).

FLATAU, A. S. ,’ Brockett, R. W., and Brown, J.V., Grouts and

Grouting, A Survey of Materi_aLs and Practice_, Civil Engineering

and public Works R.eview, Vol. 68, No. 804, July 1974.

-199-
FLEMING, W. G. K. ; Fuchsberger, M. ; Kipps, 0. ; and Sliwinski, 2 * Diaphragm Wall Specificatis

l P

and Anchorages,

Conference on Diaphragm Walls

September 19 74.

Institution of Civil Engineers, London, pp. 57-61,

FOLLENFANT, H. G., The Victoria Line, Proceedings Institute of

Civil Engineers, Supplemental Paper 727OS, pp. 335-478, 1969.

FONDEDILE, The Pali R.adice for Underpinning, Company Brochure,

England.

FORAKY, Execution d’une Congelation de Sol Pour Creas’ement d’un

Collecteur Souterrain a Watterlos (Nord), Brussels, Be.lgium,

- -

Undated Company Brochure.

FOREST, K., Der Verbau von Leitungsgraben mit Verbauteilen aus

Stahl (Soil Support with Steel Profiles for Utility Excavations,

Schriftenreihe der Tiefbau-Berufsgenossenschaft, No. 6, 1970.

FOSTER Company, L. B., Vibrating Pile Driving/Extracting Information Booklet, Company Brochure, Pittsburgh, Pennsylvania, 1971.

FOX, E. N., The Mean Elastic Settlement of a Uniformly Loaded Area

at a Depth Below the Ground Surface, 2nd ICSMFE,Vol. 1,~. 129, 1948.

FRANKE, E., Zur Frage der Standsicherheit von Bentonit-Schlitzwanden (The Question of Stability in Bentonite-FilLed Slurry Trenches).

Die Bautechnik, Vo.1. 40, No. 12, pp. 408-411,1963.

FRUCO & Associates, Dewatering and Ground Water Control, for Deep

Excavations, Report prepared for Waterways Experiment Station,

U. S. Corps of Engineers, January 1966.

FUCHSBERGER, M., Some Practical Aspects of Diaphragm Wall

Construction, Conference on Diaphragm Walls and Anchorages,

Institution of Civil Engineers, London, pp. 57-62, September 1974.

FUJII, T., The Practical Application of Thermal and Freezing Methods

to Soils Stabilization, Proceedings of the First Australia-New Zealand

Conference on Geomechanics, Melbourne, pp. 337-343, 1971.

-2oo-
GAIL, C. P., Tunnel Driven Using Subsurface Freezing, Civil

Engineering, ASCE, Vol. 42, No. 5, pp. 37-40, May 1972.

GAIL, C. P., Subsurface Freezing As An Aid to Soft Ground Tunnel

Construction, Paper presented at University of Wisconsin -

Extension Professional Development Seminar on “Soft Ground

Tunneling”, April 24-25, 1973.

GALLER, S., Slurry Wall Technique Expedites Subway Construction,

Public Works, Vol. 104, No, 8, pp. 72-73, 1973.

GEDDES, W. G. N. ; Rock, G. ; and Scrimgeor, J. ; The Backwater

Dam, Proceedings of the Institution of Civil Engineers, London,

Vol. 51, pp. 433-464, March 1972.

GEFFRIAUD, J. P., Controle de la Tension des Tirants Precontraints

Definitifs D’ancrage de la Paroi Moulee du Musee Archeologique de

Fourviere a Lyon (Stress Checking of Prestressed Tie Rods of the

Diaphragm WalZ of the Fourviere Archeological Museum at Lyon),

Travaux, No. 449-450, pp. 44-54, August-September 1972.

GENERAL Under ground Structures, Inc. , Construction Freezing,

Undated Company Brochure, Wheatridge, Colorado, 21 pages.

GERBER, E., Untersuchungen uber die Druckverteilunp in ortlich belasteten

Sand, Zinich, Switzerland, 1929.

GIRNAU, G., Pr obleme und Entwicklungstendenzen beim unterirdis then

Bahnbau in der Bundesrepublik (Problems and Developments in

Subway Construction in Germany), Strasse Brucke Tunnel, Vol. 21,

No. 6, pp. 141-150, 1969.

GLOSSOP, R., The R.ise of Geotechnology and its Influence on Engineering Practice, Geote chnique, Vol. 18, No. 2, pp. 105-150, June 1968.

GOETTLE, R. ; Flaig, J. J. ; Miller, A. J. ; Schaefer, S. E. ; Deep

Excavation and Buildings Supported by Tie-Back System, Accompanying Inland Ryerson Catalog entitled “Prestressed Soil and Rock

Anchor Systems”, August 1974.

GOLDBER.G-Zoino & Associates, Inc. , Job Files

GOLDER, H. Q; Gould, J. P. ; Lambe, T. W. ; Tschebotauioff, G. P. ;

and Wilson, S. D. ; Predicted Performance of Braced Excavations,

ASCE, JSMFD, Vol.. 96, SM 3, pp. 801-815, 1970.

-201-
GOLDFINGER, H., Permanent Steel Used as Subway Shoring, Civil

Engineering, ASCE, Vol.. 30, No. 3, pp. 47-49, March 1960.

GOULD, J. P., Design of Temporary Retaining Structures for the

Washington Subway System, ASCE Specialty Conference on Earth

and Earth-Supported Structures, Purdue University, Vol. III,

pp. 251-253, 1972.

GOULD,’ J. P. , Personal Communication, 1975.

GOUVENOT, D., Essais en France et a L’etranger sur Le Frottement

Lateral en Fondation: Amelioration par Injection, (Fr,ench & Foreign

Works on Lateral Contact in Foundations: Improvement by Grouting),

Travaux, pp. 64-72, November 1973.

GRAF, E. D., Compaction Grouting Technique and Observations,

ASCE, JSMFD, Vol. 95, No. SM 5, pp. 1151-1158, September 1969.

GREENWOOD, D. A . , Baugrundverbesserung dur ch Tiefenverdichtung

(Improvement of Foundation Soil by Deep Compaction), Baumaschine

und Bautechnik, Vol. 19, No. 9, pp. 367-375, 1972.

GRINER@D, M., Boreslam pa Norsk Kontinentalsokkel (Drilling Mud on

the Norweigian Continental Shelf), Teknisk Ukeblad, Vol. 121,

No. 34, pp. 37-42, 1974.

GRUNDBA U- Tas chenbuch Band I (Foundation Engineering Handbook,

Volume I) 2nd Edition, Berlin Munich, Wilh. Ernst’ & Sohn, 1966.

-202-
HAFFEN, M., Unterirdischc Bauwerka in Kohasionsarmen, kornigen

Boden, Vorbehsndlung cur Abdichtung und Verfestigung mittels

Injectionen (Unbrground sructures in Low-Cohesion, Granular

Soils, Injection Treatment for Sealing and Compaction), Ber gbauWissenschaft, Vol. 17, No, 0, pp. 290-294, 1970.

HAFFEN, M., Der Stand der Schlitzwandtechnik (The State of Slurry

Trench Wall Techlogy), Ortcsrreichische Ingenieur - Zeitschrift,

Vol. 16, No. 10, pp. 321-330, 1973.

HAFFEN, M. and Janin, J., Grouting Cohesionless Water -Bearing

Soils in City Tunnqls, Proceedings, North American R.apid Excavation and Tunneling Conference, Vol. 2, June 5-7; 1972.

HAGERTY, D. J., Noise, Pile Driving and OSHA, Paper presented at

Piletalk Seminar o& Curreti Practices in Pile Design and Installation. 1974.

HALLIBURTON Services, Grouting in Soils, Vol. I, A

State-of-the-Art Report FHWA-RD-76-26; Vol. II,

Design and Operations Manual FHWA-RD-76-27, 1976. ’

HANKE, W. K., Die V-Balm - Tunnel unter dem Munchener Rathaus

(The Subway Tunnel Belaw the Munich City Hall), Der Bauingenieur,

Vol. 44, No. 2, pp. 46-51, 1969.

HANNA, T. H. , Design and Behavior of Tie-Back Retaining Walls,

Acta Technica Acadsmiae Scieatiarum Hungaricae Tomus, Vol. 63,

pp. 423-431, 1968a.

HANNA, T. H., Ground Anchors in Civil Engineering,

Engineering, Vol. 1, No, 5, pp. 38-41, 1968b.

HANNA, T. H., Ground Anchor o in Civil Engineering,

Ground

Institution of

Civil Engineers, Norther’n Ireland Association, 26 pages,

February 9, 1970.

HANNA, T. H., The Influence of Anchor Inclination on Pull-Out

Resistance of Clays, Canadian Geotechnical Journal, Vol. 10,

No. 4, pp. 664-669, 19’73a.

HANNA, T. H., Anchored Inclined Walls; A Study of Behavior, Ground

Engineering, Vol. 6, No, 6, pp. 24-33, November 1973 b.

-203-
HANNA, T. H., Foundation Instrumentation, Trans Tech PubLications, Cleveland, 1973~

HANNA, T. I-I., -Personal Communication, 1975. .-.-. -.__ --- , _ __~ _-._

HANNA , T. H. and Kur di, I. I. , Studies on Anchored Flexibte Retaining Walls in Sand, ASCE, JGED, GT 10, pp. 1107-l 122, October

1974.

HANNA, T. H. and Littlejohn, G. S., Design and Construction

Considerations Associated with Retaining Walls Supported by

Prestressed Tie-Backs, The Consulting Engineer, Part 1, May

1969, Part 2, June 1969.

HANNA , T. H. and Matallana, G. A. , The Behavior of Tied-Back

Retaining Walls, Canadian Geotechnical Journal, Vol. 7, No. 7,

pp. 372-396, 1970.

HANNA, T. H. and Seeton, J. E., Observations on a Tied-Back

Soldier Pile and Timber Lagging Wall, Ontario Hydro Research

Quarterly, Vol. 19, No. 2 pp. 22-28, 1967.

HANSBO, S., Hofmann, E. and Mosesson, J., bstra Nordstaden,

Gothenburg, Experiences Concerning a Difficult Foundation Problem

and Its Unorthodox Solution, 8th ICSMFE, Moscow, Vol. 22,

pp. 105-110, 1973.

HARMS, H. and Berg, R., Die Baugrubenaussteifung beim Bau von

Undergrundbahnen nach der Berliner Bauweise (The Bracing in

Open Cuts for Subway Constructions with the Berlin Method), Die

Bautechnik, Vol. 34, No. 1, pp. 16-20, 195’7.

HASHEMI, H. T. and Sliepcevich, C. M., Effect of Seepage Stream on

Artificial Soil Freezing, ASCE, JSMFD, Vol. 99, SM 3, pp. 267-

287, March 1973.

HEAT’ON, E., Grouting, Theory of Grouting, Mining in Canada,

February 1968.

HEATON, E., Grouting, Methods and Procedures, Mining in Canada,

March 1968.

-204-
HEATON, E., Grouting, Properties of Chemical Grouts and Pumping

Systems, Mining in Canada, June 1968.

HEATON, E. , Grouting, Pumping Equipment Specifications, Mining

in Canada, November 1968.

HEATON, E., Grouting, Estimating Chemical Quantities and Grouting

Systems, Mining in Canada, December 1968.

HEATON, E., Grouting, Examples of Chemical Grout Technology-,

Mining in Canada, March 1969.

HEEB, A., Schurr, E., Bonz, M., Henke, K. F., and MuLLer, Hi,

Erddruckmessungen am Baugrubenverbau fur Stuttgarter

Verkehrsbauwerke (Earth Pressure Measurements on Sheeted

Excavations during Construction of the Stuttgart Subway), ‘Die

Bautechnik, Vol. 43, No. 6, pp. 208-216, 1966.

HENKE, K. F. and Bonz, M., Steifenkraftmessungen beim U-BahnBau in Stuttgart (Strut Load Measurements during the Subway

Construction in Stuttgart), Strasse Briicke Tunnel, Vol. 24, No. 12,

pp. 318-321, 1972.

HENKEL, D. J., Geotechnical Considerations of Lateral Stresses,

ASCE Specialty Conference on Lateral Stresses in the Ground and

Design of Earth - Retaining Structures, Cornell University,

pp. l-49, June 1970.

HENNING, D., Warmetechnische Verfestigung bindiger Lockergesteine

(Thermic Compaction of Cohesive Loose Soils), BergbauWissenschaft, Vol. 17, No. 8, pp. 311-314, 1970.

Ruhr schnetlwegtunnel im Essen (The R.uhr

Highway Tunnel in Lessen), Hochtief-Nachrichten, 1970.

HODGSON, F. T.; Design and Construction of a Diaphram Wall at

Victoria Street, London, Conference on Diaphragm Walls and

Anchorages, Insititution of Civil Engineers, London, pp. 45-50,

September 1974.

-205 -
HOFFMANN, M., Bangr ubenums chliesqungen mit r uckyartiger

Verankerung (Excavation Walls with Tie-Backs), Der Bauingenieur,

I Vol. 46, No. 10, pp. 374-377, 1971.

HOLZMANN, A. G., Technische Berichta, UIBahsl-Bau (Subway

Construction), April 1967. 6

HOUK, I. I., Arch Dams, published in Section 3 of the second

edition of the Handbook of Applied Hydraulicar, C. Vi Davis,

Editor in Chie,f, McGraw Hill, 1952

I (

HUDER, J., Deep Braced Excavation with High Ground Water Level.,

7th ICSMFE, Mexico, Vol. 2, pp. 443-448, 1969.

HUDER, J., Stability of Bentonite SLurry Tre.n,ches; $ome Experience

in Swiss Practice, 5th ECSMFE, bIrdrid, Vol. 1, pp. 517-522,

1972.

HUNT, Hal W., Design and Installatian of Pile Founc$ationo, Associated Pile & Fitting Corp., Clifton, New Jersey, 1974,

HURTADO, J., Parois prkfabriquees scelleer dans Les sol, (Precast

Concrete Walls Sealed in the Soil), Annalea de L’Inrtitut Technique

du Batiment et des Travaux Publics, No. 299, pp. 34-48, November

1972.

HUTCHINSON, J. N., Settlements in Soft Cl,a_rAround a Pumped

Excavation in Oslo, Norwegian Geotechnical Institute, Oslo,

Publication No, 58, pp. l-8, 1964.

HUTCHINSON, M. T.; Daw, G. P.; Shotton, P. G.; and James, A.N.;

The Properties of Bentonite Slurries used in Diaphragm Walling

and Their Control, Conference on Diaphragm Walls and Anchorages,

Institution of Civil Engineers, pp. 27-34, September 1974.

-206-
ICOS Corporation, The ICOS Company in the Underground Works,

Company Brochure, 1968.

IKUTA, Y., Diaphragm Walling in Japan, Ground Engineering,

pp. 39-44, 1974.

INGLES, 0. G., and Metcalf, J. B., Soil Stabilization, Principles

and Practice, Wiley, New York, 1975.

INGLES, 0. G. and Metcalf, J. B., Soil Stabilization, Halsted Press,

1973.

INLAND - Ryer son, Catalog on Soil and Rock Anchors, March, 1974.

INSLEY, A., A Deep Excavation and a Raft Foundation in Soft Clay,

Canadian Geotechnical Journal, Vol. 9, No. 3, pp. 237-248, 1972.

‘, Contributions to Soil Anchors, Proceedings

of a Conference Organized by the Institution of Civil Engineers in

London, pp. 84-94, June 1970.

, Contributions to Lateral Support Systems,

Proceedings of a Conference Organized by the Institution of Civil

Engineers in London, London, pp. 9’7-101, June 1970.

ISCHY, E. and GLossop, R., An Introduction to Alluvial Grouting,

Proceedings Institution of Civil Engineers, Vol. 21, pp. 449-474,

March 1962.

-207-
JACKSON, R. H., Freeze Watts Key to Deep Excavation, Western

Construction (reprint), April 1969.

JACKSON, W. T.; Perez, J. V.; and Lacroix, Y.; Foundation

Construction and Performance for a 34-Story Building in St. Louis,

Geote chnique, Vol. 24, No. 1, pp. 63-90, 1974.

JAMES, E. L., and Phillips, S. H. E., Movement of a Tied Diaphragm

Retaining Wall During Excavation, Ground Engineering, July 1974.

JAMES, E. L., and Jack, B. J., A Design Study of Diaphragm Walls,

Conference on Diaphragm Walls and Anchorages, Institution of Civil

Engineers, London, pp. 35-44, September 1974. ”

JANBU, N., Soil Compressibility As Determined by Oedometer and

Triaxial Tests, ECSMFE, Weisbaden, Vol. 1, 1963.

JANIN, J. J. and LeScieLlour, G. F., Chemical. Grouting for Paris

Rapid Transit Tunnels, ASCE, Journal of the Construction Division,

Vol. 96, COl, pp. 61-74, June 1970.

JAWORSKI, W. E., An Evaluation of the Performance of a Braced

Excavation, Massachusetts Institute of Technology, June 1973.

JELINEK, R. , Constructive Problems and Case Histories, 5th ECSMFE,

hladrid, pp. 295 -310, April 1972.

JENNE, G., Wirtschaftliche Spundwandberechung (Economical

Designing of Steel Walls), Die Bautechnik, Vol. 49, No. 8, pp. 260-

264. 1972.

JESKE, T. and Przedecki, T., The Rankine Pressure Theory in the

Light of Kinematics of Granular Medium, 8th ICSMFE, Moscow,

pp. 149-154, August 1973.

JESSBERGER, H. L., and Nussbaumer, M., Einfluss von Zeit und

Temperatur auf die Bemessung von Konstruktionen aus geforenem

Lockergestein (,Influence of Time and Temperature on the Dimensioning of Frozen Soil Structures), Bergbau-Wissenschaft, Vol. 17,

No. 8, pp. 299-303, 1970.

-208-
JESSBERGER, H, L, and Nussbaumer, M., Anwendung des Gefrierverfahrens (Application of Artificial Soil Freezing,

Vol. 50, No. 12, pp. 414-420, 1973.

Die Bautechnik,

JOAS, H., Weber, J., and Heine, E., Die Sicherung vonGeb:uden

und sonstigen Anlagen beim MGnchener U-Bahn-Bau (The Underpinning of Buildings and Other Structures during the Subway

Construction in Munich), Die Bautechnik, Vol. 48, No. 2, pp. 37-

42, 1971.

JORGE, G. R., Le Tirant IRP Reinjectable Special pour Terrains

Meubles, Karstiques ou a Faibles Caracteristiques Gcotechniques,

7th IGSMFE, Mexico (Abstracted translation by EC1 Solelanche,

Pittsburgh, PA), 1969.

JUMIKIS, A. R. , Thermal Soil Mechanics, Rutgers University Press,

New Jersey, 267 pages, 1966.

-zoq-
KAMENSKII, R. M. Thermal Engineering Calculations of the Frozen

Soil Watertight Cuttoff Dams, Takinp into Account the Mutual

Influence of the Freezing Columns, published in Russian in Gidroteckhnicheskoi Stroitel’stvo, No. 4., pp. 38-k2, April, 1971.

KANY, M., Empfehlungen zur Berechnung und Konstruktion von

Tunnelbauten (Recommendation for Design and Construction of

Tunnels), Die Bautechnik, Vol. 5 1, No. 8, pp. 276-280, 1974.

KAPP, M. S. , Tall Towers Will Sit on Deep Foundations, Engineering

News Record, pp. 36-38, July 9, 1964.

KAPP, M. S., Slurry Wall Construction for Basement Will of World

Trade CenterTCivil Engineering, Vol. 39, No. 4, pp. 36-40,

April 1969.

KLRCHER, K., Erdwieder s tand bei Bohlwanden in bindigen B&den

(Earth Pressure on Soldier Pile Walls in Cohesive Soils), Die

Bautechnik, Vol. 49, No. 8, pp. 264-268, 1972.

KAROL, R. H., Chemical Grouting Technology, ASCE, JSMFD,

Vol. 94, SM 1, pp. 175-204, 1968.

KASTNER, R. and Lareal, P., Experimental Excavation of Length

50m Supported by Strutted Cast Diaphragm Walls: An Analysis of

Stress Distribution in the Struts, Conference on Diaphragm Walls

and Anchorages, Institution of Civil Engineers, London, pp. 17-20,

1974.

KEUTE, H., and Schurrer, V., Bemessungshilfen und Ankerlangenbestimmung fur Baugrubensicherungen mit Hilfe von Diagrammen

(Design Diagrams for Tie-Back Length and Excavation Support),

Die Bautechnik, Vol. 51, No. 5, pp. 1.55-160, 1974.

KERSTE’N, M. S. Final Report, Laboratory Research for the

Determination of the Thermal Properties of Soils, Engineering

Experiment Station, University of Minnesota, 1949.

KEZDI, A., Stability of Rigid Structures, General Report, 5th

ECSMFE, Madrid, pp. 105-130, April 1972.

-210-
KIENBER.GER, H., Diaphragm Walls .as Load Bearing Foundations,

Conference on Diaphragm Walls and Anchorages, Institution of

Civil Engineers, London, pp.’ 19-22, September, 1974.

KJAERNSLI, B. , Empirisk regel for bestemmelse av avstivingskrefter

i utgravninger i Oslo-leire (Empirical Rules for Determination of

Strut Loads in Excavations in Oslo Clay), N. G. I. - Report No. 83,

Oslo, pp. l-10, 1970.

KLEIN, K., Lateral Pressure of Clayey Soils on Structures,

Discussion in Specialty Session 5, Vol. 4. 3, 8th ICSMFE, p. 232,

Moscow, 1973. 1,

KLOPP, R., Verwendung vorgespannter Felsanker in geklGftetem

Gebirge aus ingenieurgeologischersicht (The Application of

Prestressed Anchors in Jointed Rock), Der Bauingenieur, Vol. 45,

No. 9, pp. 328-331, ,197O.

KNAUPE, W., Hydraulischer Grundbruch an Baugrubenumschliessungen

in Schichtweise gelagerten Baugrund (Hydraulic Ground Failure in

Excavations in Layered Soils), Bauplanung, Bautcchnik, Vol. 26,

No. 9, pp. 428-431, 1972.

KNAUPE, W., Aushub umschlossener Baugruben im Schichtweise

gelagerten Baugrund bei Wasserandrang (Excavations in Layered

Soils with High Ground Water Table), Bauplanung-Bautechnik,

Vol. 27, No. 10, pp. 500-504, 1973.

KOTULLA, B. , Konstruktive Proble,me bei der Sicherung von

Bauwerken im Bereich tiefer Baugruben (Constructional Problems

by the Support of Buildings Adjacent to Deep Exca,vations), Tiefbau,

Vol. 15, No. 9, pp. 784-788, 1973.

KRAMER, J., Wasserhaltung bei tiefen Baugruben im Ruhrgebiet

(Groundwater Control in Deep Excavations in the Ruhr Area),

Tiefbau, Vol. 15, No. 5, pp. 526-530, 1973.

KRANICH, W. and Musfeldt, T., Das Einpressen ganzer Bauwerke

unter Eisenbahngleisen (The Jacking of Complete Structures

Through Railway Embankments), Beton-und Stahlbetonbau, Vol. 67,

No. 12, pp. 265-271, 1972.

-211-
KRANZ, E., Uber die Verankerung von Spundwanden, 2 Auflage,

Mitteilungen aus dem Gebiete des Wasserbaues und der

Baugrundfor s chung, Heft 11, Berlin: ’ ‘Wilh, Ernst & Sohn, 1953.

KRIMMER, H., Technische Entwicklung der Baugrubenwande beim

Frankfurter U-Bahn-Bau (Technical Development of the Excavation

Support for the Subway Construction in Frankfurt), Baumaschine

und Bautechnik, Vol. 19, No. 4, pp. 136-144, 1972.

KRIZEK, R. J., and Corotis, R. B., Probabilistic Approach to Heave

of Soft Clay Around Sheetpilc Walls, 8th ICSMFE, Moscow,

Vol. 1.3, pp. 143-150, 1973. 2,

KUTZNER, C. , c ber di

- -

e mechanischen Eigenschaftcn der mit

Silikatgelen injizierten Erdstoffe (The Mechanical Properties of

Soils Injected with Silicate or Grout), Die Bautechnik, Vol. 45,

No. 3, 12 pages, 1968.

KUTZNER, C., and Ruppel, G., Chemische Bodenverfestigung zur

Unterfahrung des Hauptbahnhofs beim U-Bahnbau in Koln (Chemical

Soil Stabilization as Underpinning of the Main Railway Station in

Cologne for Subway Construction), Strasse Brucke Tunnel, Vol. 22,

No. 8, pp. 202-205, 1970.

-212-
LACKNER,’ E., Technis cher Jahresbericht 1971 des Arbeitsausschusses “Ufereinfassungen” (Annua.1 Report 1971 of the German

CommitteTfor Quay and Shore Walls), Die Bautechnik, Vol. 48,

No. 12, pp. 409-420, 1971.’

LACKNER, E., Tochnischer Jahresbericht 1972 des Arbeitsausschusses “Ufereinfassungen” (A nnual Report 1972 of the German

Committee for Quay and Shore Walls), Die Bautechnik, Vol. 49,

No. 12, pp. 405-420, 1972.

LACKNER, E.,Technischer Jahresbericht 1973 des Arbeitsausschusses “Ufereinfassungen” (Annual Report 1973 of the German

Committee for Quay and Shore Walls), Die Bautechnik, Vol. 50,

No. 12, pp. 397-414, 1973.

LA CROIX, Y. and Jackson, W., Design and Construction of

Supported Temporary Excavations in Urban Environment, 3rd

Ohio Soil Seminar, October 1972.

LADD, C. C., Strength Parameters and Stress-Strain Behavior of

Saturated Clays, Research Report R71-23, Soils Publication 278,

Massachusetts Institute of Technology, 1971.

LADD, C.C.;Bover, R.B.;Edgers, L.; andRixner, J. J.;

Consolidated Undrained Plane Strain Shear Tests on Boston Blue

Clay, R.esearch in Earth Physics, Phase Report 15, M.I. T.

Department of Civil Engineering, Research Report R71-13,

243 p., 1971.

LADD, C. C., and Varallyay, J; The Influence of Stress System

on the Behavior of Saturated Clays During Undrained Shear,

M. I.. T. Department of Civil Engineering, Research Report

R65-11, 263 p., 1965.

LAMB, B., Massive Concreting Follows Earth-Anchoring, Construction Methods and Equipment, Vol. 55, No. 2, pp. 64-67,

1973.

LAMBE, T. W., Soil Stabilization, Foundation Engineering, Chapter

4, G. A. Leonards, Ed., McGraw-Hill Book Co., pp. 351-437,

1962.

-213-
LAMBE, T. W., Predicting Performance of Braced Excavations,

ASCE Specialty Conference on Performance of Earth and EarthSupported Structures, Purdue University, Vol. III, pp. 255 -256,

1972. 2

LAMBE, T. W., Wolfskill, L. A., and Jaworski, W. E., The

Performance of a Subway Excavation, ASCE Specialty Conference

onthe Performance of Earth and Earth-Supported Structures, , Purdue University, Vol. I, Part 2, pp. 1403-1424, 1972.

LAMBE, T. W., ,Wolfskill, L. A., and Wong, I. H., Measured

Performance of Braced Excavation, ASCE, JSMFD, Vol. 96,

SM3, pp. 817-836, 1970.

LARNACH, W. J., The Behavior of Grouped Inclined Anchors in

Sand, Ground Engineering, Vol. 6, No. 6, pp. 34-41, November

1973.

LARSON, M. L. , Willette, W. R..; Hall, H. C. ; and Gnaedinger, J. P. ;

A Case Study of a Soil Anchor Tie-Back System, ASCE Specialty

Conference on Performance of Earth and Earth-Supported Structures,

Purdue University, Vol. I, Part 2, pp. 1341-1366, 1972.

LAUNINGER, K. and Song, F., Stutzmauern. Beispiele r-nit Anwendung

von Pressbetonbohrpfahlen und riickwartigen Verankerungen

(Retaining Walls. Examples of the Use of Bored Concrete Piles

with Tie-Backs), Beton-und Stahlbetonbau, Vol. 69, No. 4,

pp. 88-92, 1974.

LEISER, A. , and Schmaus, W. , A bfangung Nochbelas teter StGtzen

und Umbau der Griindung bei einer Stahlbetonskelettkonstruktion

(Underpinning of Heavy Loaded Columns and Reconstruction of the

Foundations for a Reinforced Concrete Skeleton Building), Betonund Stahlbetonbau, Vol. 64, No. 10, pp. 225-229, 1969.

LEONARD, M. W., and Moller, K., Grouting for Support with

Particular Reference to the Use of Some Chemical Grouts, Grouts

and Drilling Muds in Engineering Practice, Butterworths, London,

1963.

-214-
LEONARD, M. S. M., Precast Diaphragm Walls Used for the Al3

Motorway, Paris, Conference on Diaphragm WaLls and Anchorages,

Institution of ?%z Engineers, London, pp. ‘71-75, September 1974.

LEONARDS, G. A.;, Foundation Engineering, McGraw-Hill. Book

Company, Lnc. , 1962.

LINELL, K. A., Hennion, F.B., and Lobacz, E. F., Corps of

Engineers’ Pavement Design in Areas of Seasonal Fro&, H. R. B.

Research Record, No. 33, 1963.

LITTLEJOHN, G. S., Soil Anchors, Proceedings of a Conference

Organized by the Institution of Civil Engineers in London, pp.

33 -44, June 1970a.

LITTLEJOHN, G. S., Discussionof Soil Anchors, Proceedings of a

Conference Organized by the Institution of Civil Engineers in

London, pp. 70-71, 1970b.

LITTLEJOHN, G. S., Ground Anchors Today - A Foreword, Ground

Engineering, Vol. 6, No. 6, pp. 20-23, November, 1973.

LITTLEJOHN, G. S., sock Anchors - Some Design Considerations,

Draft of Report on Rock Anchors, 1974.

LITTLEJOHN, G. S. ; Acceptable Water Flows for Rock Anchor

Grouting, Geotechnics Research Group, Department of Engineer -

ing, University of Aberdeen, 1974.

LITTLEJOHN, G. S. and Bruce, D.A., Rock Anchors-State-of-theArt, Part I: Design, Ground Engineering,~ Vol. 8, No. 3, May,

pp. 25-33, 1975.

LITTLEJOHN, G. S. ; Jack, B. J. ; and Sluvinski, Z., Anchored

Diaphragm Walls in Sand, Reprinted articles from Ground

Engineering, September to November 1971, January 1972.

LITTLEJOHN, G. S. and MacFarlane, I. M., A Case History Stud2

of Multi-Tied Diaphragm Walls, Conference on Diaphragm Walls p-c---

and Anchorages, Institution of Civil Engineers, London, pp. 91-

100, September 1974.

LIU, T. K. and Dugan, J, P. Jr., An Instrumented Tied-Back Deep

Excavation, AS’CE SpeciaLty Conference on Earth and EarthSupported Structures, Purdue University, Vol. I, Part II,

pp. 1323-1339, June 1972.

-215-
LIVER, N. L., Tension Piles and Diagonal Tiebacks, ASCE, Journal

of the Sanitary Engineering Di-vision, pp. 107-120, July 1969.

LIZZI, I. F., Special Patented System of Underpinning and. More

Generally, Subsoil Strengthening by Means of “Pali Radice” (Root

Piles) with Snecial Reference to Problems Arising from the.

Construction of Subways in Built-Up Areas, Illustrated Text of

Lecture on Root Piles, Fondedile Corp, 1974.

LIZZI, F., Reticolo Di Pali Radice, Italian Geotechnical Association

Meeting BARI, October 1970.

LOCHTE, J., Mzgvlichkeiten zur Anwendung des Injectionsverfahrens

bei gering durchlassigen Lockergesteinen (Application Possibilities

of the Injection Method in Low-Permeability Soils), BerbauWissenschaft, Vol. 17, No. 8, pp. 295-299.

LOERS, G., Neue Erkenntnisse bei der Herstellung und Aus bildung

von Schlitzwanden (New Facts for the Construction and Design of

Slurry Walls), Tiefbau, Vol. 15, No. 5, pp. 445 -449, 1973.

LOGEAIS, L. and Graux, D., Les tirants d’ancrage, (Tie-Backs),

Annales de L’Institut Technique du Batiment et des Travaux

Publics, No. 3 12, pp. 126- 146, December 1973.

LORENZ, H., Utilization of a Thixotropic Fluid in Trench Cutting

and the Sinking of Caissons, SDGMEP, Butterworth, London,

pp. 202-205, 1963.

LOVENBURY, H. T., and Vaughan, P. R., Discussion to the

Proceedinps, 5th ECSMFE, Madrid, Vol. 2, p. 339, 1972.

-216-
MALJIAN, P. A., and Van Beveren, J. L., Tied-Back Deep Excavations in Los Angeles Area, ASCE, Journal of the Construction

Division, CO 3, pp* 337-356, September 1974.

MAIDL, B. and Nellessen, H. G., Bewehrungsst8sse im Tunnelbauunter Beriicksichtigung der neuen DIN 1045 (Reinforcement in

Tunnel Construction as a Result of the New DIN 1045), Die Bautechnik, Vol. 50, No. 11, pp. 361-370, 1973.

MAISHMAN, D., Ground Freezing, Methods of Treatment of Unstable l

Ground (in press), Butterworth and Co, 1975.

MAJTENYI, S. I., Personal Communication, 1975.

MANSUR, C. I., and Alizadeh, M., Tie-Backs in Clay to Support

Sheeted Excavation, ASCE, JSMFD, Vol. 96, SM 2, pp. 495-509,

1970.

MARGOLIN, V. M., and D’Yakonov, V. P., Stability of Vertical

Walls of Trenches Under Hydrostatic Heads, Soil Mechanic and

Foundation Engineering, Vol. 10, No. 1, pp. 60-62, 1973.

MARKER, H., Die Anwendung der Schlitzwandbauweise (The use of

Slurry Walls), Schweitzerische Bauzeitung, Vol. 88, No. 33,

pp. 735-740, 1970.

MARSAK, S. A., Herstellung eines Wassertunnels im Gefrierverfahren, Gidrotechniceskoe stroitel stvo (Wassertechnisches Bauwesen), Volume 35, No. 6.

MARTIN, A., Vorgespannte Schlitzw&de ffr die Zeltdachyerankerung

der Olympia - Sports&ten in Munchen (Prestressed Diaphragm

Walls for Anchoring the Tent-Roof over the Olympia Stadiums in

Munich), Baumaschine und Bautechnik, Vol. 18, No. 10, pp. 415-419,

1971.

Massachusetts Institute of Technology, “Grouting Applications in

Civil Engineering, Department of Civil Engineering, Prepared

for Oak Ridge National Laboratory, Contract: ORNL-Sub-3960-1,

December 1974.

-217”
Massachusetts Institute of Technology, If)roject Report, M. I. T.

Test Section Instrumentation - Massalchusetts Bay Transportation

Authority Haymarket - North Extension-2 Soil Mechanics Division,

Project Mass. MTD-2, 1972.

MAZURKIEWICZ, B. K., The Rupture Figure for a Double Wall

Cofferdam, ASCE Specialty Conference on Performance of Earth

and Earth-Supported Structure, Purdue University, pp. 1271-1281,

June 1972.

McROSTIE, G. C.: Burn, K. N. ; and Mitchell, R. J. ; The Performance of Tied-Back Sheet Piling in C1a.b Canadian Geotechnical

Journal, Vol. 9, No. 2, pp. 206-218, 1972.

MEAD, H. T., Complex Earth Holding System Prepares for Deep

Excavation in Congested Area, World Construction, Vol. 27, No. 5,

pp. 18-22, 1974.

MEISSNER, H., Zur Verankerung von Wanden, die Gelandespriinge

verformungsarm abstutzen sollen (Design of Tie-Backs to Obtain

Minimum Wall Deformation), Der Bauingenieur, Vol. 45, No, 9,

pp. 337-339, 1970.

MERGENTIME, C. E., Underpinning Massive Structures, Civil

Engineering, Vol. 42, No. 3, pp. 40-41, March 1972.

MEYERHOF, G. G., Stability of Slurry Trench Cuts in Saturated Clay(

ASCE Specialty Conference on Perfor:mance of Earth and EarthSupported Structures, Purdue University, Vol. I, Part 2,

pp. 1451 - 1466, 1972.

Milan Polytechnic, Report on the Measurements Taken on a Portion

of Reticulated “Pali Radice” Retaining Wall, on Behalf of the

Metropolitan Transit Authority of Milan, Institute of Geodesy,

Topography and Photogrammetry. -

MITCHELL, J. K., Fundamental Aspects of Thixotropy in Soils,

ASCE, JSMFD, Vol. 86, SM 3, pp. 19-53, 1960.

-218-
MITCHELL, J. K., In Place Treatment of Foundation Soils, Paper

presented at specialty Conference: Placement and Improvement of

Soil to Support Structures, MIT, Cambridge, Mass., August 26-28,

also ASCE JSVLFD, Vol. 96, SMl, 1970.

MITCHELL, J. M., Some Experiences with Ground Anchors in

London, Conference on Diaphragm Walls and Anchorages, Institution of Civil Engineers, London, pp. 1 - 11, September 1974.

MijBIUS, C. H. , and Giinther, T. H., Modellver suche zur Standfestigkeit flussigkeitgestutzer Erdwande (Model Tests to Determine Stability of Slurry Trenches), Die Bautechnik, Vol. 50,

No. 5, pp. 156-160, 1973.

MOLLER, K., Grouting Now, Consulting Engineer, London, Vol. 36,

No. 8, pp. 44 - 49, 1972.

MORGENSTERN, N. and Amir-Tahmasseb, A., The Stability of a

Slurry Trench in Cohesionless Soils, Geotechnique, Vol. 15,

No. 4, pp. 387-395, 1965.

MORONEY, N., Recharge Wells and Pit Underpinning Overcome

Settlement Caused by Dessication, Civil Engineering, AXE,

Vol. 31, No. 1, January 1961.

. .

MULLER, D. and Krense, P., Rammversuche mit Verschiedenen

Stahlspundbohlen in einem Ton-und Geschiebemergel (Driving

Tests With Various Steel Sheetings in Clayey Marl with Obstructions), Die Bautechnik, Vol. 49, No. 11, pp. 370-374, 1972.

MULLER-Haude, H. Ch., and von Scheibner, D., Neue Bodendruckmessungen an Baugruben und Tunnelbauten der Berliner U-Bahn

(New Earth Pressure Measurements on Support Walls and Tunnel

Structures at the Berlin Subway), Die Bautechnik, Vol. 42, No. 9

and 11, pp. 293-298 and 380-385, 1965.

MULLER-Kirchenbauer, H., E rddruckme s sungen an e inem ho hen

Brucken-wiedew (Earth Pressure Measurements on a High

Bridge Abutment ), Baumaschine und Bautechnik, Vol. 20, No. 5,

pp. 169-173, 1973.

-219-
MULLER-KIRCHENBAUER, H., Stability of S1urr.y Trenches, 5th

ECSh4FE, Madrid, Vol. 1, pp. 543-553, 1972.

MGLLER-KIRCHENBAUER, H., Einflcsse der Filtratbildung auf die

Stabilitgt von Schlitzwsnden (The Influence of Filter Cakes on the

Stability of Slurry Trenches), Vortrzge der Baugrundtagung 1972

in Stuttgart, pp. 843-860, Deutsche Gesellschaft fur Erd-und

Grundbau e.v. Essen 1973.

MURPHY, D. J. ; Clough, G. W.; and Woolworth, R,S. ; Temporary

Excavation in Varved Clay, ASCE, JGED, Vol. 101 GT 3, March,

pp. 279-295, 1975. /N

MYRVOLL, F., Feltforstik ved utfdrelse av tverrslissegrdft 41 ved

fellestunnelen i Studenterlunden (Field Test by the Construction of

the Transverse Slurry Trench 41 for the Studenterlund Tunnel),

NGI Internal Report 71075-1, September 25, 1973.

MYRVOLL, F., Malinger og observasjoner utfQrt under arbeidet

med tverslissegrdft 42 i Studenterlunden (Measurements and Observations Made During the Work with Transverse Slurry Trench

42 in Studenterlunden, Oslo), NC1 Internal Report No. 71075-2,

March 25, 1974.

MYRVOLL, F., and Karlsrud, K., Skiaerforsdk i full mklestokk &

blQt leire i Studenterlunden, Oslo (Full Scale Shear Test on Soft

Clay in Studenterlunden, Oslo), NGI Publication 102, pp. 31-38,

1974.

-220 -
NASH, 5. K. T. L. , and Jones, G. K. , The Support of Trenches

Using Fluid Mud, SMDMEP, Butterworths, London, pp. 177-180,

1963.

NAVARRO, H., Les tirants en terrain meuble, type T. M. (Anchoring

Rods of T. M. Type for Loose Soil), Travaux, No. 451, pp. 15-21,

October, 1972.

NAVFAC DM-7, Design Manual - Soil Mechanics, Foundations and

Earth Structures, Department of the Navy, Naval Facilities

Engineering Command, Washington, D. C. , 197 1.

NEELANDS, P. .J., and James, A. N., Formulation and Selection of

Chemical Grouts with Typical Examples of Field Use, Grouts and

Drilling Muds in Engineering Practice, Butterworths, London,

pp. 150-155, 1963.

NEELY, W. J., and Montague-Jones, M., Pullout Capacity of StraightShafted and Underreamed Ground Anchors, Die Siviele Ingenieur,

South Aftica, Vol. 16, pp. 131-134, April, 1974.

NELSON, J. C., Earth Tiebacks Support Excavation 112 Feet Deep,

Civil Engineering, Vol. 43, No. 11, pp. 40-44, November, 1973.

NENDZA, H., Sicherung tiefer Baugruben neben Bauwerken (Lateral

Support of Deep Excavations Adjacent to Buildings), Tiefbau, Vol.

15, No. 8, pp. 698-702, 1973.

NENDZA, H. , and Klein, K. , Bodenverformung beim Aushub’tiefer

BauPrugen (Soil Deformation by Deep Excavations), Strasse Brucke

Tunnel, Vol. 26, No. 9, pp. 23 l-239, 1974.

NEUMANN, H. , and Wilkins, L. S., Soil Solidification by Chemical

Injection, Civil Engineering and Public Works Review, pp. 635-636,

June, 1972.

NEWMARK, N. M., Influence Charts for Computation of Stresses in

Elastic Foundation, University of Illinois Bulletin, Vol. 40, No. 12,

1942.

NICHOLSON Anchorage Co., Nicholson Ground Anchors, Company

Catalogue, 1974.

-221-
NORTH-LEWIS, J. P. and Lyons, G. H. S., Continuous Bored Piles,

Conference on Diaphragm Walls and Anchorages, Institution of

Civil Engineers, London, pp. 43-48, 1974.

NORWEGIAN Geotechnical Institute, Measurements at a Strutted

Excavation, Report Nos. 2-9.

NORWEGIAN Geotechnical Institute, Measurements at a Strutbed 1

Excavation, Oslo Subway, Gronland 2, km 1, 692, Oslo, 111 pages.

-222-
OOSTERBAAN, M. D., and Gifford, D. G. p A Case Study of the Bauer

Earth Anchor, ASCE S pecialty Conference on Performance of Earth

and Earth-Supported Structures, Purdue University, Vol. 1, Part

2, pp. 1391-1401, 1972.

O’ROURKE, T. D., and Cording, E. J. , Observed Loads and Displacements for a Deep Subway Excavation, Proceedings Rapid Excavation

and Tunneling Conference, San Francisco, Vol. 2, p. 1305, 1974a.

O’ROURKE, T. D., and Cording, E. J., Measurement of Strut Loads

by Means of Vibrating Wire Gages, Proceedings Sy,mposium on

Performance Monitoring for Geotechnical Construction, A.STM,

STP 584, pp. 58-‘77, 1974b.

OSTERBY, N. P., The Lowry Hill Tunnel Freeze Wall, Paper presented

at the Highway Research Board meeting, January, 1971.

OSTERMAYER, H., Discussion on Ground Anchors, 5th ECSMFE,

Madrid, Vol. II, pp. 334-336, April, 1972.

OSTERMAYER, H., Construction, Carryinp Behavior and Creep

Characteristics of Ground Anchors, Conference on Diaphragm Walls

and Anchorages, Institution of Civil Engineers, London, pp. l-12,

September, 1974.

-223-
F’ALMER, J. H. ) and Kenney, T. C., &lalytical Study of a Braced

Excavation in Weak Clay, Canadian Geotechnical Journal, Vol. 9,

-_-_.._--

No. 2, pp. 145-164, 1972.

PARSONS Associates, Report on Underpinning with Root Piles, Nonpublished Technical Report prepared for Washington Metropolitan

Area Transit Authority,

PARSONS, J. D., Foundation Installation Requiring Recharging of

Ground Water, Journal cf the Construction Division, ASCE,

September, 1959

PASSARO, Aldo, St. Lucia Tunnel - Salerno Underpinning by Reticulated

Root Piles, Report to Italian State Railway System.

PATERSON, A. C. , Underpinning, Proceedings, Conference on Ground

Engineering organized by the Institution of Civil Engineers,

London, Paper No. 5, pp. 45-54, 1970.

PAVLOW, Fred, Design and Construction of Deep Retained Excavations,

ASCE/SEAONC Continuing Education Seminars, Lecture dated

November 23, 1970.

PECK, R. B. , Deep Excavations and Tunneling in Soft Ground, Stateof-the-Art Report, 7th ICSMFE, Mexico City, State-of -the -Art

Volume, pp. 225-290, 1969.

PECK, R. B., Observation and Instrumentation, Some Elementary

Considerations, Lecture Notes, Met. Section ASCE, Seminar on

Field Observations in Foundation Design and Construction, New

York, April, 1970. (Reprinted in Highway Focus, U. S. Dept.

of Transportation, F.H.W.A., Vol. 4, No. 2, pJ?* l-5, June, 1972.)

PECK, R. Be, Hanson, W. E., and Thornburn, T. H., Foundation

Engineerin& Second Edition, Wiley, New York, 1974.

PEER, G., Concrete Caissons Shore Highway Wall, Reprint from

Heavy Construction News, May 29, 1972.

PEER, G. A. , Montreal Job has Two Types of Shoring, Reprint from

Heavy Construction News, January 22,, 1973.

PETERSEN, G., and Schmidt, H., Untersuchungen uber die

Standsicherheit verankerter Baugrubenwande an Beispielen des

Hamburger Schnellbahntunnelbaues (Stability Analysis of Anchored

Walls for the Subway in Hamburg), Strasse Brucke Tunnel, Vol.

23, No. 9, pp. 225-233, 1971.

-224-
PETERSEN, G, , and Schmidt, H., Zur Berechnung von Baugrubenwande

nach dem Traglastverfahren (Design of Support’Walls by the Method -.

of Load Capacity), Die Bautechnik, Vol. 50, No. 3, pp. 85-91, 1973.

PETERSON, C.

Journal of the Construction Division, Vol. 

R., Force Reduction in Excavation Devices ASCE, )

99, No. COl, pp 21-33,

1973.

PIASKOWSKI, A. , and Kowalewski, Z. , Thixotropic Properties of

Suspensions of Soils with Different Grain Sizes and of Various -.

Mineralogical Types, 5th ICSMFE, Vol. 1, -pp. 293 -296, 1961.

PIASKOWSKI, A., and.Kowalewski, Z. , Application of Thixotropic

Clay Suspension for Stability of Vertical Sides of Deep Trenches

Without Strutting, 6th ICSMFE, Canada, Vol. II, pp. 526-529,

1965.

PILARSKI, L., Precast- Walls Drop Into Place in Six Easy Steps,

Construction Methods and Equipment, Vol. 56, No. 7, p* 71,

1974.

PLAISTED, A. C., Application of Chemical Grouts to Ground

Consolidation, Ground Engineering, Vol. 7, No. 4, pp. 42-44,

July, 1974.

POTEVIN, G., La Congelation des Terrains Dans les Travaux Publics,

Industrie Minerale, April, 1972.

PRASAD, G. D., Freeman, C. F., and Klajnerman, D., Deep

Excavations for the Manufacturers Life Centre, Toronto, Canadian

Geotechnical Journal, Vol. 9, No. 2, pp. 137-144, 1972.

PREGL, O., Ein allgemeiner Extremwertverfahren zur Bestimmunq

de s E rddrucke s und de r E rddruckve r te ilung auf Abs tli tzungen

(A General Maximum Value Method to Calculate Earth Pressure

and Pressure Distribution on Lateral Support Walls), Der

Bauingenieur, Vol. 45, No. 9, pp. 332 -336, 1970.

PRENTIS, E. A. , and White, L., Underpinning Columbia University

Press, 374 pp., 1950.

-225 -
PROCHAZKA, J.; Berechnung unveranketer Spundwande bei trapezformiger Verteilung des Erddruckes (Design of Unsupported Steel

Sheet Piling with Trapezoid Shaped Earth Pressure), Die Bautechnik,

‘Vol. 48, No. 5, pp. 153-160, 1971.

PULLER, M. J., Slurry Trench Stability Theoretical and Practical

Aspects, Ground Engineering, pp. 34-38, September, 1974,

-226-
RANK, D., and Nussbaumer, W., Bodenverfestigungsversuch bei der

Schwedenbr’ucke in Wien - eine ne.ue Anwendung von Radionukliden

in der Bauwirtschaft (Earth Stabilization Test by the Swedish Bridge

in Vienna, A New Way of Using Nuclear Dye in the Construction

Business), Oesterreichische Ingenieur-Zeitschrift, Vol. 16, No. 10,

pp. 331-338, 1973.

RANKE, A., and Ostermayer, H., Beitrag zur Stabilitats - Untersuchung

mehrfach verankerter Baugrubenumschliessungen (A contribution to

the Stability Calculations of Multiple Tied-Back Walls), Die

Bautechnik, Vol. 45, No. 10, pp. 341-349, 1968.

Tunnel), Holzmann-Nachrichten, 1970.

RICH A R DS, R., Jr., and AgrawaI, J. S.,

Circular Pipe by Transformed Section,

100, GT6, June, 1974.

Stresses on fhallog

ASCE, GTED, Volume

RODRIGUE Z, J. M. , and Flamand, C. L. , Strut Loads Recorded in a

Deep Excavation in Clay, Proceedings, 7th ICSMFE, Mexico City,

Vol. 2, pp. 459-467, 1969.

RAPPERT, D., Beeinflussung von Bauwerken durch Unterfahrungen

(The Influence of Tunnel Constructions on Buildings Above the

ROGERS, Walter F., Composition and Properties of Oil Well Drilling

Fluids, Gulf Publishing Co. , 1963.

ROTH, B., Ruckwartige Verankerung an Baugrubenwanden, Teil II,

(Tied-Back Excavation Walls, Part II), Vortragsveroffentlichungen,

Haus der Technik, Essen, No. 241, 1970.

ROMBERG, W., Messunpen an einer verankerten Baugrubenwand

(.Measurements on a Tied-Back Excavation Wall), Mitteilungen der

Versuchsanstalt fur Bodenmechanik und Grundbau der Technischen

Hochschule Darmstadt, No. 12, December, 1973.

RUFFANI, E. , Baugrubensicherung fur eine extrem tiefe Pressengrube

(Bracing of an Extremely Deep Structural Excavation), BauplanungBautechnik, Vol. 26, No. 9, pp. 418-421, 1973.

RUFFERT, G., Betonspritzverfahren zum Schutz unterirdischer Bauten

(Shotcrete for Protection of Subsurface Constructions), Tiefbau,

Vol. 15, No. 3, pp. 213-216, 1973.

RUPPEL, G. , Ausfurhrung von Injektionen in Lockergesteinen (Injection Technique in Cohesionless Soil), Bergbau-Wissenschaft, Vol.

17, No. 8, pp. 285-290, 1970.

-227-
SABZEVARI, A., and Ghahramani, A., Theoretical Investigations of

the Passive Progressive Failure in an Earth Pressure Problem,

Soils and Foundations, Vol. 13, No. 2, pp. 1-18, June 1973.

SADLIER, N, A., and Dominioni, G. G; , Underground Structural

Concrete Walls, SGDMEP, Butterworths, London, pp. 189-195,

1963.

SALAS, J. A. J., Ureil, S., and Serrano, A. A., Behavior Under

Horizontal Forces of Two Cast-in-Place Diaphragm Walls in

Loose Layered Soil, Proceedings 5th ECSMFE, Madrid, Supplement B. , pp. 79-89, 1972.

SANDHU, B.S. ,Earth Fressure on Walls Due to Surcharge, Civil

Engineering, ASCE, Vol. 44, No. 12, pp. 68-70, December, 1974.

SANDQVIST,G., Back-Tied Sheet Pile Wall in Friction Soil. Defor -

mations and Drag Forces Due to Pilinp and Freezing, 5th ECSMFE,

Madrid, Vol. I, pp. 293 -298, April, 1972.

SANGER, F. J., Ground Freezing in Construction, ASCE, JSMFD,

Vol. 94, SM 1, pp. 131-158, 1968.

SAXENA, S. K., Measured Performance of a Rigid Concrete Wall at

the World Trade Center, Conference on Diaphragm Walls and

Anchorages, Institution of Civil Engineers, London, pp. 85-90,

September, 1974.

SCHEUCH, G., Die Grundung des Fariser Turmhochhauses MaineMontparnasse (The Foundation of the Paris Skyscraper MaineMontparna sse ), Tiefbau, Vol. 14, No. 12, pp. 1151-1152, 1972.

SCHEUCH, G., U-Bahn Bau in Hannover (Subway Construction in

Hanover), Tiefau, Vol. 15, No. 5, pp* 540-542, 1973.

SCHMID, W, E., and Prieto, L. A., Ea,rth Anchors. Stress Distribution Analysis and Design, 5th ECSMFE, Madrid, Vol. 1, pp.

25-31, April, 1972.

SCHMIDT, B., and Dunnicliff, C. J. , Construction Monitoring of

Soft Ground Rapid Transit Tunnels. A Definition of Needs and

Potential Developments, Report on Contract DOT/TSC 661, Rail

Program Branch of the Urban Mass Transportation Administration,

Office of Research, Development and Demonstrations, 1974.

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SCHMIDT, H. , Zur Ermittlung der kritischen tiefen Gleitfuge von

mehrfach verankerten hohen Baugrubenwanden ‘(Construction of

The Critical Deep Failure Surface by High’ Support Walls with

Several Anchorages), Die Bautechnik, Vol. 51, No. 6, pp. 210-

212, 1974.

SCHMITT, G. P., Ein Beitrag zur K&rung des Tragverhaltens einfach

verankerter Baugrubenwande (A Contribution to the Understanding of

Soil-Structure Interaction by Single .Tied-Back Walls), Mitteilungsheft

14 der Versuchsanstalt fur Bodenmechanik und Grundbau der

Technischen Hochschule Darmstadt, 1974.

SCHMITZ, H., and Schmaus, W., Tieflegung der Bundesbahn in Bonn

(Subsurface Reconstruction of the Railway System in Bonn),

Strasse Brucke Tunnel, Vol. 25, No. 7, 9 pages, 1973.

SCHNABEL, H., Jr., Procedure for Testing Earth.Tiebacks, ASCE

National Structural Engineering Meeting, Cincinnati, Ohio, April

22-26, 1974.

SCHNEEBELI, P. G., La Stabilite Des Trenches Profonde s Forees

en Presence De Bouf (The Stability of Deep Trenches Excavated in

a Slurry), La Houille Blanche, No. 7, pp. 815-820, 1964.

, Durchpressungen (Horizontal Pile Jacking)

Schriftenreihe der Tiefbau-Berufsgenossenschaft, No. 8, 1972.

SCHULTZ, E. W., Die Anvendung der Spritzbetonbauweise beim Bau

der Frankfurter S-Bahn (The Use of Shotcrete in the Construction of

the Frankfurt Subway), Der Eisenbahningenieur, Vol. 24, No. 6,

pp. 177-181, 1973.

SCHWARZ, H.,Permanentverankerung einer 30 m hohen St’utzwand im

Stuttgarter Tonmergel durch korrosionsgeschutze Injectionsanker,

System Duplex (Permanent Anchoring of a 30 m High Secant Wall

in the Stuttgart Clayey Marl by Means of the Duplex System of

Corrosion Protected Injection Tiebacks), Die Bautechnik, Vol. 49,

No. 9, pp. 305-312, 1972.

SCOTT, J. D., Wilson, N. E., and Bauer, G. D., Analysis and

Prformance of a Braced Cut in Sand with Large Deformations,

Canadian Geotechnical Journal, Vol. 9, No. 4, pp. 384-406, 1972.

SEELING, R., Die Schlitzwandbauweise (Diaphragm Wall Construction),

Baumaschine und Bautechnik, Vol. 18, No. 2, pp. 55-63, 1971.

SEELING, R. , Ruckwartige Verankerungen fur Baugrubenumschliessungen

(Tieback Systems for Excavation Pits), Baumaschine und Bautechnik,

Vol. 20, No. 1, pp. 3-7, 1973.

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SEEWIG, K., and Schebetz, F., Horizontale Baugrundvereisung unter

Bundesbahngleisen, Der Eisenbahnigenieur, Vol. 20, No. ll., 19x9.

SENING, G., and Klotschke, G., Die Unterfahrung des Hauptbahnohofes

beim U-Bahnbau in K&Ln (The Subway Tunnel Construction Below

the Main Railway Station in Cologne), Strasse Brticke Tunnel., Vol. 22,

No. 8, pp. 197-202, 1970.

SHANNON, W. L., and Straier, R. J., Tied-Back Excavation Wall for

Seattte First National Bank, Civil Engineeing, ASCE Vol. 40, pp.

62-64, March 1970.

SHERARD, Woodward, Gizienski, and Clevenger, Earth and Earth-Rock

Dams, Wiley, 1963.

SHUSTER, J. A., Controlled Freezing for Temporary Ground Support,

Proceedings , North American Rapid Excavation and Tunneling

Conference, Vol. 2, Chicago, June 5-7, 1972.

SHUSTER, J. A. 9 Personal Communication, 1975.

SIGOURNEY, J. W., cInstallations ’

Conference on Design and Construction of Earth and Rock Tiebacks.

George Washington University, February 3, 1971.

SKEMPTON, A. W., The Bearing Capacity of Clays, Bldg. Research

Congress, England, 1951.

SKEMPTON, A. W., and Cattin P., A Full-Scale Alluvial Grouting

zest at the Site of the Mangla Dam, Grouts and Drilling Muds in

Engineering Practice, Butterworths, London, pp. 131-141, 1963.

SLIWINSKI, Z. J., Permeability of Concrete Diaphragm Walls, Ground

Engineering, Vol.7, No. 4, p. 21, July 1974.

SLIWINSKI, Z. , and FLeming, W. G. K., Practical Considerations

Affecting the Construction of Diaphragm Walls, Conference on

Diaphragm Walls and Anchorages, Institution of Civil Engineers

London, pp. l- l,O, September 1974,

SOOS, P. von, Anchors for Carrying Heavy Tensile Loads into the

Soil, 5th ECSMFE, Madrid, Vol. 1, pp. 555-563, 1972.

SPANGLER, M. G., Horizontal Pressures on Retaining Walls Due to

Concentrated Surface Loads, Bulletin No. 140, Iowa Engineering

Experiment Station, 1940.

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STAVROPOULOS, D., Ein Beitrag zur Frage der Verankerung von

Stiitswanden durch Injektionsanker (A Contribution to the Problem

of Anchoring Support Walls with Tie-Backs), Die Bautechnik, Vol.

49, No.’ 8, pp. 253-260, 1972.

ST. <JOHN, H. D., Per sonal Communication, 1974.

STOSS, K., Die Abteufen kleiner Gefrierschacte mit mobilen

Grosslochbohranlagen, Gluckauf, Vol. 197, No. 20 (reprint), 1971.

STROH, D. Berechnung verankerter Baugruben nach der Finite

Element Methode (Design of Anchored Walls with the Finite Element

Method), Mitteilungsheft 13 der Versuchsanstalt fur Bodenmechanik

und Grundbau der Technischen Hochschule Darmstadt, 1974.

SVERDRUP and Parcel and Associates, Cut-and-Cover Tunneling

Techniques, National’Technical Infor,mation Service, Report No.

FHWA -RD-73-40, Springfield, Virginia, 1973.

SWATEK, E.P., Asrow, S.P., and Seitz, A.M., Performance of

Bracing for Deep Chicago Excavation, ASCE Specialty Conference

on Performance of Earth and Earth-Supported Structures, Purdue

University, Vol. 1, Part 2, pp. 1303-1322, June 1972.

SZECHY, K., Some Practical Comments to the Slurry Trench Method,

5th ECSMFE, Madrid, Vol; 2, pp. 319-323, 1972.

SZECHY, K., Uber die -*. Stabilitat von Schlamms chlitzen ,( The Stability

of Slurry Trenches) A eta Technica A cademiae Scientiarum Hungaricae,

Tomus 75 (l-4), pp. 371-381, 1973.

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TAIT, R, G. , and Taylor, H. T. , _Uesign Construction and Performance

of Rigid and Flexible Bracing Systems for Deep Excavations in San

Francisco Bay Mud, ASCE National Meeting on Water .Resources

Engineering, Los Angeles, Preprint No. 2162, 25 pages, January 21-

25, 1974.

TAMARO, G.; Concrete and Slurry Walls Using Slurry Trench

Construction, Lecture Notes at 20th Annual Soil Mechanics and

Foundation Engineering Conference, University of Minnesota,

March 23, 1972.

TAMAR.0, G., Slurry Walls, Open Cut Construction Seminar,

Metropolitan Section, ASCE, New York, 1974.

TAN, S.B., Settlements Due to Deep Foundation Construction,

8th ICSMFE, Moscow, vol. 22, pp. 221-226, August 197c

TAYLOR, D. W., Fundamentals of Soil Mechanics, John Wiley and

Sons, New York, 1948.

TSCHEBOTARRIOFF, G. P., Foundations Retaining and Earth

Structures 2nd Ed., McGraw Hill, 1973.

TENG, W. C., Foundation Design, Prentiss-Hall, pp. 418-419,

1962.

TERZAGHI, K., Stability and Stiffness of C,ellular Cofferdams, TRANS.

ASCE,Vol. 110, pp. 1083-1202, 1945.

TERZAGHI, K., Theoretical Soil Mechanics, John Wiley & Sons,

New York, Seventh Printing, 510 pages, 1954a.

TERZAGHI, K., Anchored Bulkheads, Transactions, ASCE, Vol. 119

Paper 2720, pp. 1243-1324, 1954b.

TERZAGHI, K., Evaluation of Coefficient of Subgrade R.eaction,

Geo technique , ,,Vol. 4, pp. 279, London 1955.

TERZAGHI, K., and Peck, R. B., Soil Mechanics in Engineering

Practice, Wiley, 1968.

THON, J. G., and Harlan, R. C., Slurry Walls for BART Civic Center

Subway Station, ASCE, JSMFD, Vol. 97, SM 9, pp. 1317-1333,

1971.

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U-Bahn-Barr in Numberg (Subway Construction in

Number g), Tiefban, Vol. 15, No. 12, pp. 1051 - 1055.

TOMLINSON, M. J., Foundation Design and Construction, 2nd Ed.,

Wiley Inters cience, Division of Wiley, New York, 1969.

TOMLINSON, M. J., Lateral Support of Deep Excavations, Proc. of

Conference Organized by the Institution of Civil Engineers in London,

pp. 55-64, June 1970.

TOMLINSON, M. J., Discussion of Lateral Support Systems, Proc.

Conference Organized by Institution of Civil Engineers in London,

pp. 72, 74, June 1970.

TROW, W.A., Temoorarv and Permanent Earth Anchors: Three

Monitored InstaLlations, Canadian Geotechnical Journal, Vol. 11,

No. 2, pp. 257-268, 1974.

TSYTOVICH, N. A., (Translated from Russian by L. Diashevska),

Bases and Foundations on Frozen Soils, Highway Research Board

Special Report No. 58, National Academy of Sciences - National

Research Council Publication 804, Washington, D. C., 1960.

-233-
VIALOV, S. S. , The Strength and Creep of Frozen Soils and Calculations

for Ice Soil Retaining Structures, Translation 76, U. S. Army Corps

of Engineers, Cold Regions Research and Engineering Laboratory,

September 1965 ( original article published in 1962 in Russian),

VEDER, C., An Investigation on the Electrical Phenomena at the

Area of Contact Between Bentonite Mud and Cohesionless Material,

5th ICSMFE, Vol. 3, pp. 146-149, 1961.

VEDER, C., Excavation of Trenches in the Presence of Bentonite

Suspensions for the Construction of Impermeable Load-Bearing

Diaphragms, SGDMEP, Butterworth, London, pp. 181-188, 1963.

VEDER, C., Behavior of Concrete Poured Under Bentonite Suspensions,

7th ICSMFE, Mexico, Vol. 2, pp. 483-486, 1969.

VEDER, C., Oral Discussion Conference on Diaphragm Walls and

Anchorages, Institution of Civil Engineers, London, 1974.

VESIC, A. S. , Ultimate Loads and Settlements of Deep Foundations

in Sand, Bearing Capacity and Settlement of Foundations,

Proceedings of a Symposium held at Duke University, pp. 53-68,

April, 1965. (Published 1967).

-234-
WAKABAYASHI, J., Water and Air Inject Instant Slurry Wall, Construction Methods and Equipment, Vol. 56, No. 7, pp. 64-65, 1974.

WANOSCHEK,‘R., and Breth, H., Auswirkung von Hauslasten auf die

Belastung ausgesteifter Baugrubenwande (The Effect of Building

Loads on Lateral Earth Pressure in Braced Excavations), Strasse

Briicke Tunnel, Vol. 24, No. 8, pp. 197-200, 1972.

WARD, W.H., zerformance of Braced Excavations in London Clav.

ASCE Specialty Conference on Performance of Earth and EarthSupported Structures, Purdue University, Vol. III, pp. 257-258,

June 1972.

I I

WARE, K. R., Underpinning Methods and Related Movements, RETC

Conference - San Francisco, Vol. II, 1974.

WARE, K, R., Personal Communication, 1974.

WARE, K.R., Mirsky, M., and Leuniz, W.E., Tieback Wall Construction

Results and Controls, ASCE, JSMFD, Vol. 99, SM 12, pp. 1135-

1152, 1973.

WARNER, J., Strength Properties of ChemicaLly Solidified Soils,

ASCE, JSMFD, Vol. 98, SM 11, pp. 1163-1185, November 1972.

WARNER, J., and Brown, D. R.. , Planning and Performing Compaction

Grouting, ASCE, JGED, Vol. 100, CT 6, pp. 653-666, 1974.

WEBSTE.R, L. 0, How to Get a Clear Excavation by Using a New Shoring

Method, Engineering and Contract Record, Vol. 85, No. 2, pp. 62-65,

February 1972.

WEINHOLD, H., and Kleinlein, H., Berechnung and Ausftihrung einer

schragen Bohrpfahlwand (Design and Construction of an Inclined

Secant Wall), Der Bauingenieur, Vol. 44, No. 7, pp. 233-239, 1969.

WEISS, F. ,‘ Schlitzwand Bauweise (Slurry Trench Construction),

Schriftenreihe der Tiefhau-Berosgenossenschaft, No. 7, 1971.

WEISSENBACH, A., Berechnung von mehrfach gestiitzten Baugruben

spundwanden and Tragerbohlwanden nach dem Traglastverfahren

(Design of Multiple Supported Soldier Pile and Steel Sheet Piling

W3.11~ with the “Carrying Load Method”), Strasse Bri.icke Tunnel,

Vol. 2 1 Nos. 1, 2 , 3, and 4, pp. 17-23, 38-42, 67-74, 130-136,

1969.

-235-
WEISSENBACH, A., Empfehlungen des Arbeit:skreises”Baugruben” der

Deutschen Gesellschaft fur Erdund Grundbau e. V. (R.ecommendations

from the German Committee for Excavations), Die Bautechnik,

Vol. 47, No. 7, p. 223, 1970.

WEISSENBACH, A., Empfehlungen des Arbeitskreises “Baugruben”

der Deutschen Gesellschaft fur Erdund Grundbau e. V. (Recommendations

from the German Committee for Excavations), Die Bautechnik,

Vol. 49, No. 6 and 7, pp. 192-204 and 229-239, 1972.

WEISSENBA CH, A., Lastannahmen fur Bauten, Kenngrossen des

Bodens (Building Loads and Soil Parameters), Die Bautechnik,

Vol. 51, No. 6, pp. 2011206, 1974(a).

WEISSENBA CH, A., _Empfehlungen des Arbeitskreises “Baugruben”

der Deutschen Gesellschaft fur Erd-und Grundbau e. V. (Recommendations)

of the German Committee for Excavations), Die Bautechnik, Vol.

51, No. 7, pp. 228-232, 1974 (b).

WELZIEN, K., and Wyleciol, H., Unter suchungen zur Beurteilung der

Wasserdurchlassigkeit von Schlitzwanden (Investigations to

Determine the Permeability of Diaphragm Walls), Bauplanung-Bautechnik,

Vol. 27, No. 4, pp. 189-190, 1973.

WHITE, E. E., Underpinning, Foundation Engineering, ed. G. A.

Leonards, McGraw-Hill, pp. 826 - 893, 1962.

WHITE, E. E., Sheeting and Bracing Systems for Deep Excavations,

World Construction, Vol. 26, No. 1, pp. 20 L 24, January 1973.

WHITE, R. E., Caissons and Cofferdams, Article in Foundation Engineering, Leonards, McGraw- Hill, 1962.

WHITE, .R. E., Anchored Walls Adjacent to Vertical Rock Cuts,

Conference on Diaphragm Walls and Anchorages, .Institution of Civil

Engineers, London, pp. 35-42, 1974 (a).

WHITE, R.. E., Discussion of Tieback W?Ll Construction - Results and

Controls, Proceedings, ASCE JGED, No. 10, pp. 1167-1168, 1974(b).

-236”
WILDFENER, F., Konsolidierung bei nicht standfestem Boden durch

Injektion (Consolidation of Unstable Soil with Grouting), Der Mens ch

un die Technik, March 8, 1973.

WaHLBIER, H., Verbesserung der Gebirgsfestigkeit

Stability), Ber gbau-Wissenschaft, Vol. 17, No. 8,

WONG, I. H., Analysis of Braced Excavations, SC. D. T

Massachusetts Institute of Technology, 1971.

Improving Soil

pp. 276-277, 1970.

lesis,

WOODWA RD, R. J. , GARDNER, W. S. , and GREER, D. M. , Drilled

Pier Foundations, McGraw-Hill, 287 p. 1972.

WOODWAR.D-MOORHOUSE & ASSOC., INC., Sheetpile Interlock Seepage

Study for AISI, undated.

, Quieter Sheet Pile Driving, World

Construction Vol. 22, No. 4, p. 28, 1974.

WOSSER, T. D., and DARRAGH, R. D., Tiebacks for Bank of America

Building Excavation Wall, Civil Engineering, Vol. 40, No. 3, pp. 65-

67, 1970.

WROTH, C. P., General Theories of Earth Pressures and Deformations,

General Report, 5th ECSMFE, Madrid, Vol. II, pp. 33-52, April 1972.

XANTHAKOS, P. P., Underground Construction in Fluid Trenches,

University of Illinois at Chicago Circle, 1974.

YOKOYAMA, H., Excavation Work of a Large Scale in a Soft Layer at

the Tokyo Underground Station, Proc. of the 6th National Tunnel

Symposium, Tokyo, pp. 6-l --6-29, September 14-16, 1970.

ZELGER, C., Meesungen der Krgfte in Baugrubenaussteifungen der

Mi;nchener U-Bahn (Load Measurements on the Support Walls for

the Munich Subway), Die Bautechnik, Vol. 50, No, 1, pp. 2-7, 1973.

ZIENKIEWICZ, 0. C., and Cheung, Y.K., The Finite Element Method

in Structural and Continuum Methods, McGraw-Hill Publishing

Company, Ltd., London, 1967.

ZIMMERMANN, H., Krenzungsbauwerk Nord-Siid-Fahrt/Eigelstein in

Kijln (Crossing Construction for North-South Highway in Cologne),

Strasse Brucke Tunnel, Vol. 21, No. 5 and 6, pp. 113-120 and

151-156, 1969.

*U.S. GOVERNMENT PRINTING OFFICE: 1976-28t-173/663

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