Techniques for the

Seismic Rehabilitation

of Existing Buildings

FEMA 547/2006 Edition
Cover Photo Courtesy of Kelly Peterson, University of Utah 

Cover Photo: Unbonded Steel Braces are used as part of the seismic rehabilitation of the J. 

Willard Marriott Library on the University of Utah campus. FEMA provided partial funding of 

this project through the Pre-Disaster Mitigation Competitive (PDM-C) grant program.
Techniques for the Seismic Rehabilitation 

of Existing Buildings 

FEMA 547 – October 2006 

Prepared by: 

Rutherford & Chekene (R & C) Consulting Engineers (Subconsultant) under 

contract with National Institute of Standards and Technology (NIST). 

Project funding was provided by the Federal Emergency Management Agency 

(FEMA) through an Interagency Agreement - EMW-2002-IA-0098 with the 

National Institute of Standards and Technology (NIST). 

Additional funding was provided by the following agencies: 

General Services Administration (GSA) 

Air Force Civil Engineer Support Agency (AFCESA)

Naval Facilities Engineering Command (NAVFAC)

U.S. Bureau of Reclamation (USBR)

 
ii 

Notice 

Any opinions, findings, conclusions, or recommendations expressed in this publication do not 

necessarily reflect the views of the Federal Emergency Management Agency, the General 

Services Administration (GSA), the Air Force Civil Engineer Support Agency (AFCESA), the 

National Institute of Standards and Technology (NIST), the Naval Facilities Engineering 

Command (NAVFAC), and Rutherford & Chekene Consulting Engineers (R&C), and R&C’s 

subconsultants. Additionally, neither FEMA, R&C nor its subconsultants, AFCESA, FEMA, 

GSA, NIST, NAVFAC, USBR, or other ICSSC member agencies, nor any of their employees, 

makes any warranty, expressed or implied, nor assumes any legal liability or responsibility for 

the accuracy, completeness, or usefulness of any information, product, or process included in this 

publication. Users of information from this publication assume all liability arising from such 

use. 
iii 

Preface 

This seismic rehabilitation techniques document is part of the National Earthquake Hazards 

Reduction Program (NEHRP) family of publications addressing seismic rehabilitation of existing 

buildings. It describes common seismic rehabilitation techniques used for buildings represented 

in the set of standard building types in FEMA seismic publications. This document supersedes 

FEMA 172: NEHRP Handbook for Seismic Rehabilitation of Existing Buildings, which was 

published in 1992 by the Federal Emergency Management Agency (FEMA). Since then, many 

rehabilitation techniques have been developed and used for repair and rehabilitation of 

earthquake damaged and seismically deficient buildings. Extensive research work has also been 

carried out in support of new rehabilitation techniques in the United States, Japan, New Zealand, 

and other countries. Available information on rehabilitation techniques and relevant research 

results for commonly used rehabilitation techniques are incorporated in this document. 

The primary purpose of this document is to provide a selected compilation of seismic 

rehabilitation techniques that are practical and effective. The descriptions of techniques include 

detailing and constructability tips that might not be otherwise available to engineering offices or 

individual structural engineers who have limited experience in seismic rehabilitation of existing 

buildings. A secondary purpose is to provide guidance on which techniques are commonly used 

to mitigate specific seismic deficiencies in various model building types. 

FEMA sincerely thanks all of the federal agencies that contributed funds toward completing this 

report as well as the members of the Interagency Committee for Seismic Safety in Construction 

(ICSSC) Subcommittee 1, the Technical Update Team, and all of the federal and private sector 

partners for their efforts in development, review and completion of this publication. 
iv 

Acknowledgments 

Private sector consultants working with the Subcommittee 1, Standards for New and Existing 

Buildings, of the Interagency Committee on Seismic Safety in Construction (ICSSC) produced 

this document. The following persons contributed to this document. 

Contractors

Technical Update Team 

William T. Holmes, Team Leader, Rutherford & Chekene, San Francisco 

Bret Lizundia, Project Manager and Editor, Rutherford & Chekene, San Francisco 

James O. Malley, Degenkolb Engineers, San Francisco 

Kelly Cobeen, Cobeen & Associates Structural Engineering, Inc., Lafayette, California 

Contractor Staff 

Rutherford & Chekene: Afshar Jalalian, Gyimah Kasali, Mark Moore, Rich Niewiarowski, 

and Karl Telleen 

Degenkolb Engineers: Jack Hsueh 

Consultant on Foundation Rehabilitation Techniques 

Craig Comartin, CDComartin, Inc., Stockton, California 

Other 

Figures shown in Table 4-1 and repeated later in the text were adapted from those originally 

rendered by Anthony Alexander of Palo Alto, California for a separate FEMA-funded 

project. 

Project Review Panel

J. Daniel Dolan, Washington State University 

Terry Dooley, Consultant 

Kurt Gustafson, American Institute of Steel Construction 

Robert D. Hanson, FEMA Consultant 

Neil M. Hawkins, Consultant 

James R. Harris, J. R. Harris and Company 

Bela Palfalvi, General Services Administration (representing ICSSC) 

Daniel Shapiro, SOH and Associates 

FEMA Project Officer 

Cathleen Carlisle, DHS-FEMA, Washington, DC 
v 

ICSSC Subcommittee 1

H. S. Lew, Chair National Institute of Standards and Technology 

John Baals Department of Interior, U.S. Bureau of Reclamation 

Krishna Banga Department of Veteran Affairs 

Rosana Barkawi Department of Agriculture, U.S. Forest Service 

James Binkley U.S. Postal Service 

Larry Black Department of Army 

Cathleen Carlisle Federal Emergency Management Agency 

James A. Caulder Air Force Civil Engineering Support Agency 

Cathy Chan Department of Justice, Bureau of Prisons 

Harish Chander Department of Energy 

Anjana Chudgar U.S. Army Corps of Engineers 

Ronald Crawford Department of Housing and Urban Development 

Joseph Corliss Department of Health and Human Services 

Nathaniel Foster Tennessee Valley Authority 

Freda Gerard International Broadcasting Bureau 

Colleen Geraghty Department of Labor, Office of Safety and Health 

Asadour Hadjian Defense Nuclear Facility Safety Board 

Owen Hewitt Naval Facilities Engineering Command 

Howard Kass National Aeronautics and Space Administration 

Regina Larrabee Department of Commerce 

Catherine Lee General Services Administration 

S. C. Liu National Science Foundation 

Rita Martin Department of Transportation 

Thomas Myers Smithsonian Institution 

Thomas Nelson Lawrence Livermore National Laboratory 

Dai Oh Department of State, Office of Foreign Buildings 

Bela Palfalvi General Services Administration 

Erdal Safak Department of Interior, U.S. Geological Survey 

Subir Sen Department of Energy 

C.A. Stillions Architect of the Capital 

Lawrence Swanhorst Environmental Protection Agency 

Steven Sweeney Army Construction & Engineering Research Laboratory 

Doris Turner Federal Aviation Administration 

Terry Wong National Park Services 
vi 

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Techniques for the Seismic Rehabilitation of

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Table of Contents 

Notice.............................................................................................................................................. ii 

Preface...........................................................................................................................................iii 

Acknowledgments ........................................................................................................................ iv 

PART 1 - OVERVIEW 

Chapter 1 - Introduction ...........................................................................................................1-1 

1.1 Overview....................................................................................................................... 1-1 

1.2 Purpose and Goals......................................................................................................... 1-2 

1.3 Audience ....................................................................................................................... 1-2 

1.4 Scope............................................................................................................................. 1-2 

1.5 Other Resources............................................................................................................ 1-3 

1.6 Organization of the Document...................................................................................... 1-4 

1.6.1 Part 1 – Overview ......................................................................................... 1-6 

1.6.2 Part 2 – Rehabilitation Techniques for FEMA Model Buildings ................. 1-6 

1.6.3 Part 3 – Rehabilitation Techniques for Deficiencies 

Common to Multiple Building Types........................................................... 1-6 

1.7 Disclaimers ................................................................................................................... 1-7 

1.8 References..................................................................................................................... 1-7 

Chapter 2 - Seismic Vulnerability ............................................................................................2-1 

2.1 Introduction................................................................................................................... 2-1 

2.2 Seismic Evaluation........................................................................................................ 2-1 

2.2.1 Comparison with Requirements for New Buildings..................................... 2-2 

2.2.2 Prescriptive Standards .................................................................................. 2-2 

2.2.3 Performance-Based Evaluation Using Expected Nonlinear Response......... 2-3 

2.3 Categories of Seismic Deficiencies .............................................................................. 2-3 

2.3.1 Global Strength............................................................................................. 2-3 

2.3.2 Global Stiffness ............................................................................................ 2-4 

2.3.3 Configuration................................................................................................ 2-4 

2.3.4 Load Path...................................................................................................... 2-4 

2.3.5 Component Detailing.................................................................................... 2-5 

2.3.6 Diaphragms................................................................................................... 2-5 

2.3.7 Foundations................................................................................................... 2-6 

2.3.8 Other Deficiencies ........................................................................................ 2-6 

2.4 References..................................................................................................................... 2-8 
Techniques for the Seismic Rehabilitation of

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viii 

Chapter 3 - Seismic Rehabilitation...........................................................................................3-1 

3.1 Introduction................................................................................................................... 3-1 

3.2 Rehabilitation Standards ............................................................................................... 3-1 

3.2.1 Mitigation of Evaluation Deficiencies.......................................................... 3-1 

3.2.2 Rehabilitation Design Based on Nonlinear Response .................................. 3-2 

3.3 Classes of Rehabilitation Measures .............................................................................. 3-2 

3.3.1 Add Elements................................................................................................ 3-3 

3.3.2 Enhance Performance of Existing Elements................................................. 3-3 

3.3.3 Improve Connections Between Components................................................ 3-4 

3.3.4 Reduce Demand............................................................................................ 3-4 

3.3.5 Remove Selected Components ..................................................................... 3-4 

3.4 Strategies to Develop Rehabilitation Schemes ............................................................. 3-5 

3.4.1 Technical Considerations.............................................................................. 3-5 

3.4.2 Nontechnical Considerations........................................................................ 3-5 

3.4.3 Cost............................................................................................................... 3-6 

3.4.4 Seismic Performance .................................................................................... 3-6 

3.4.5 Short-term Disruption of Occupants............................................................. 3-6 

3.4.6 Long-term Functionality of Building............................................................ 3-7 

3.4.7 Aesthetics...................................................................................................... 3-7 

3.5 Other Common Issues Associated with Seismic Rehabilitation................................... 3-7 

3.5.1 Constructability............................................................................................. 3-7 

3.5.2 Materials Testing .......................................................................................... 3-7 

3.5.3 Disruption to Building Systems and Replacement of Finishes .................... 3-7 

3.5.4 Concealed Conditions................................................................................... 3-8 

3.5.5 Quality Assurance......................................................................................... 3-8 

3.5.6 Detailing for New Elements ......................................................................... 3-8 

3.5.7 Vulnerability During Construction............................................................... 3-8 

3.5.8 Determination of Component Capacity by Testing...................................... 3-8 

3.5.9 Incremental Rehabilitation............................................................................ 3-9 

3.6 Issues with New Techniques or Products ..................................................................... 3-9 

3.7 References................................................................................................................... 3-10

PART 2 - REHABILITATION TECHNIQUES FOR FEMA MODEL BUILDING TYPES 

Chapter 4 - FEMA Model Building Types ..............................................................................4-1 

4.1 Introduction................................................................................................................... 4-1 

4.2 History of Development................................................................................................ 4-1 

4.3 Model Building Type Refinements in this Document .................................................. 4-1 

4.4 Description.................................................................................................................... 4-2 

4.5 References..................................................................................................................... 4-8 
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ix 

Chapter 5 - Building Type W1: Wood Light Frames.............................................................5-1 

5.1 Description of the Model Building Type...................................................................... 5-1 

5.2 Seismic Response Characteristics................................................................................. 5-3 

5.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques................. 5-5 

5.4 Detailed Description of Techniques Primarily Associated with This Building Type .. 5-9 

5.4.1 Add New or Enhance Existing Shear Wall................................................... 5-9 

5.4.2 Add Collector at Open Front ...................................................................... 5-28 

5.4.3 Add or Enhance Anchorage to Foundation ................................................ 5-31 

5.4.4 Enhance Cripple Wall................................................................................. 5-40 

5.4.5 Rehabilitate Hillside Home......................................................................... 5-43 

5.4.6 Rehabilitate Chimney ................................................................................. 5-48 

5.5 References................................................................................................................... 5-52 

Chapter 6 - Building Type W1A: Multistory, Multi-Unit Residential Woodframes...........6-1 

6.1 Description of the Model Building Type...................................................................... 6-1 

6.2 Seismic Response Characteristics................................................................................. 6-5 

6.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques................. 6-5 

6.4 Detailed Description of Techniques Primarily Associated with This Building Type .. 6-9 

6.4.1 Add Steel Moment Frame............................................................................. 6-9 

6.4.2 Add New or Enhance Existing Wood Shear Wall...................................... 6-16 

6.4.3 Enhance Framing Supporting Shear Wall .................................................. 6-25 

6.4.4 Enhance Overturning Detailing in Existing Wood Shear Wall .................. 6-28 

6.4.5 Enhance Shear Transfer Detailing.............................................................. 6-35 

6.5 References................................................................................................................... 6-40 

Chapter 7 - Building Type W2: Woodframes, Commercial and Industrial......................... 7-1 

7.1 Description of the Model Building Type...................................................................... 7-1 

7.2 Seismic Response Characteristics................................................................................. 7-2 

7.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques................. 7-3 

7.4 Detailed Description of Techniques Primarily Associated with This Building Type .. 7-6 

7.4.1 Add Steel Braced Frame (Connected to Wood Diaphragm) ........................ 7-6 

7.4.2 Provide Collector in a Wood Diaphragm ..................................................... 7-9 

7.5 References................................................................................................................... 7-11 

Chapter 8 - Building Types S1/S1A: Steel Moment Frames..................................................8-1 

8.1 Description of the Model Building Type...................................................................... 8-1 

8.2 Seismic Response Characteristics................................................................................. 8-2 

8.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques................. 8-3 
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8.4 Detailed Description of Techniques Primarily Associated with This Building Type .. 8-8 

8.4.1 Add Steel Braced Frame (Connected to an Existing Steel Frame)............... 8-8 

8.4.2 Add Concrete or Masonry Shear Wall (Connected to An Existing Steel 

Frame)......................................................................................................... 8-14 

8.4.3 Add Steel Cover Plates or Box Existing Steel Member ............................. 8-22 

8.4.4 Provide Collector in a Concrete Fill on Metal Deck Diaphragm ............... 8-25 

8.4.5 Enhance Connection of Steel Column to Foundation................................. 8-26 

8.4.6 Enhance Beam-Column Moment Connection............................................ 8-31 

8.4.7 Enhance Column Splice.............................................................................. 8-38 

8.4.8 Add Steel Plate Shear Wall (Connected to An Existing Steel Frame) ....... 8-42 

8.4.9 Convert an Existing Steel Gravity Frame to a Moment Frame .................. 8-45 

8.5 References................................................................................................................... 8-47 

Chapter 9 - Building Types S2/S2A: Steel Braced Frames.................................................... 9-1 

9.1 Description of the Model Building Type...................................................................... 9-1 

9.2 Seismic Performance Characteristics............................................................................ 9-2 

9.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques................. 9-3 

9.4 Detailed Description of Techniques Primarily Associated with This Building Type .. 9-8 

9.4.1 Enhance Braced Frame Connection.............................................................. 9-8 

9.4.2 Enhance Strength and Ductility of Braced Frame Member........................ 9-10 

9.5 References................................................................................................................... 9-13 

Chapter 10 - Building Type S4: Steel Frames with Concrete Shear Walls ........................ 10-1 

10.1 Description of the Model Building Type.................................................................... 10-1 

10.2 Seismic Performance Characteristics.......................................................................... 10-2 

10.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques............... 10-2 

10.4 Detailed Description of Techniques Primarily Associated with This Building Type 10-2 

10.5 References................................................................................................................... 10-2 

Chapter 11 - Building Types S5/S5A: Steel Frames with Infill Masonry Shear Walls .....11-1 

11.1 Description of the Model Building Type.................................................................... 11-1 

11.2 Seismic Response Characteristics............................................................................... 11-3 

11.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques............... 11-4 

11.4 Detailed Description of Techniques Primarily Associated with This Building Type 11-9 

11.5 References................................................................................................................... 11-9 

Chapter 12 - Building Type C1: Concrete Moment Frames................................................12-1 

12.1 Description of the Model Building Type.................................................................... 12-1 

12.2 Seismic Response Characteristics............................................................................... 12-2 

12.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques............... 12-3 

12.4 Detailed Description of Techniques Primarily Associated with This Building Type 12-7 

12.4.1 Add Steel Braced Frame (Connected to a Concrete Diaphragm)............... 12-7 
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xi 

12.4.2 Add Concrete or Masonry Shear Wall (Connected to 

a Concrete Diaphragm)............................................................................. 12-17 

12.4.3 Provide Collector in a Concrete Diaphragm............................................. 12-25 

12.4.4 Enhance Column with Fiber-Reinforced Polymer Composite Overlay ... 12-32 

12.4.5 Enhance Concrete Column with Concrete or Steel Overlay .................... 12-36 

12.4.6 Enhance Concrete Moment Frame ........................................................... 12-39 

12.5 References................................................................................................................. 12-41 

Chapter 13 - Building Type C2b: Concrete Shear Walls (Bearing Wall Systems)............13-1 

13.1 Description of the Model Building Type.................................................................... 13-1 

13.2 Seismic Response Characteristics............................................................................... 13-3 

13.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques............... 13-4 

13.4 Detailed Description of Techniques Primarily Associated with This Building Type 13-8 

13.4.1 Enhance Shear Wall with Fiber-Reinforced Polymer Composite Overlay 13-8 

13.4.2 Enhance Deficient Coupling Beam or Slab.............................................. 13-14 

13.4.3 Enhance Connection Between Slab and Walls......................................... 13-18 

13.4.4 Reduce Flexural Capacity of Shear Walls to Reduce Shear Demand ...... 13-21 

13.5 References................................................................................................................. 13-23 

Chapter 14 - Building Type C2f: Concrete Shear Walls (Gravity Frame Systems)..........14-1 

14.1 Description of the Model Building Type.................................................................... 14-1 

14.2 Seismic Response Characteristics............................................................................... 14-2 

14.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques............... 14-4 

14.4 Detailed Description of Techniques Primarily Associated with This Building Type 14-8 

14.5 References................................................................................................................... 14-8 

Chapter 15 - Building Types C3/C3A: Concrete Frames 

 with Infill Masonry Shear Walls ......................................................................................15-1 

15.1 Description of the Model Building Type.................................................................... 15-1 

15.2 Seismic Response Characteristics............................................................................... 15-3 

15.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques............... 15-3 

15.4 Detailed Description of Techniques Primarily Associated with This Building Type 15-8 

Chapter 16 - Building Type PC1: Tilt-up Concrete Shear Walls........................................16-1 

16.1 Description of the Model Building Type.................................................................... 16-1 

16.2 Seismic Response Characteristics............................................................................... 16-4 

16.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques............... 16-5 

16.4 Detailed Description of Techniques Primarily Associated with This Building Type16-12 

16.4.1 Enhance Wall-to-Diaphragm Anchorage ................................................. 16-12 

16.4.2 Enhance Beam and Girder Connection to Supporting Elements.............. 16-21 

16.4.3 Enhance Anchorage at Base of Tilt-Up Panels......................................... 16-23 

16.5 References................................................................................................................. 16-25 
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Chapter 17 - Building Type PC2: Precast Concrete Frames with Shear Walls.................17-1 

17.1 Description of the Model Building Type.................................................................... 17-1 

17.2 Seismic Response Characteristics............................................................................... 17-4 

17.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques............... 17-5 

17.4 Detailed Description of Techniques Primarily Associated 

with This Building Type ........................................................................................... 17-11 

17.4.1 Add or Enhance Collector or Chord in Existing Precast Diaphragm ....... 17-11 

17.4.2 Enhance Connections Between Existing Precast Diaphragm, 

Shear Wall and Foundation Elements....................................................... 17-18 

17.4.3 Mitigate Configurations Creating Short Columns.................................... 17-22 

17.5 References................................................................................................................. 17-24 

Chapter 18 - Building Type RM1t: Reinforced Masonry Bearing Walls (Similar to 

Tilt-up Concrete Shear Walls)................................................................................................18-1 

18.1 Description of the Model Building Type.................................................................... 18-1 

18.2 Seismic Response Characteristics............................................................................... 18-3 

18.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques............... 18-3 

18.4 Detailed Description of Techniques Primarily Associated with This Building Type 18-7 

18.5 References................................................................................................................... 18-7 

Chapter 19 - Building Type RM1u: Reinforced Masonry Bearing Walls (Similar to 

Unreinforced Masonry Bearing Walls)..................................................................................19-1 

19.1 Description of the Model Building Type.................................................................... 19-1 

19.2 Seismic Response Characteristics............................................................................... 19-2 

19.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques............... 19-2 

19.4 Detailed Description of Techniques Primarily Associated with This Building Type 19-5 

19.5 References................................................................................................................... 19-6 

Chapter 20 - Building Type RM2: Reinforced Masonry Bearing Walls (Similar to 

Concrete Shear Walls with Bearing Walls)...........................................................................20-1 

20.1 Description of the Model Building Type.................................................................... 20-1 

20.2 Seismic Response Characteristics............................................................................... 20-2 

20.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques............... 20-3 

20.4 Detailed Description of Techniques Primarily Associated with This Building Type 20-7 

20.5 References................................................................................................................... 20-7 

Chapter 21 - Building Type URM: Unreinforced Masonry Bearing Walls .......................21-1 

21.1 Description of the Model Building Type.................................................................... 21-1 

21.2 Seismic Response Characteristics............................................................................... 21-3 

21.3 Common Seismic Deficiencies and Applicable Rehabilitation Techniques............... 21-4 

21.4 Detailed Description of Techniques Primarily Associated with This Building Type 21-9 
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21.4.1 Brace or Remove URM Parapet ................................................................. 21-9 

21.4.2 Add Wall-to-Diaphragm Ties................................................................... 21-12 

21.4.3 Add Out-of-Plane Bracing for URM Walls.............................................. 21-25 

21.4.4 Add Reinforced Cores to URM Walls...................................................... 21-31 

21.4.5 Add Concrete Overlay to Masonry Wall.................................................. 21-35 

21.4.6 Add Fiber-Reinforced Polymer Overlay to Masonry Wall ...................... 21-42 

21.4.7 Infill Opening in a URM Wall.................................................................. 21-45 

21.4.8 Add Concrete or Masonry Shear Wall (Connected to 

a Wood Diaphragm) ................................................................................. 21-47 

21.4.9 Add Steel Moment Frame (Connected to a Wood Diaphragm) ............... 21-50 

21.4.10 Add or Enhance Crosswalls...................................................................... 21-53 

21.4.11 Add Supplemental Vertical Support for Truss or Girder.......................... 21-56 

21.4.12 Add Veneer Ties in a URM Wall ............................................................. 21-59 

21.5 References................................................................................................................. 21-61 

PART 3 - REHABILITATION TECHNIQUES FOR DEFICIENCIES COMMON TO 

MULTIPLE BUILDING TYPES 

Chapter 22 - Diaphragm Rehabilitation Techniques ...........................................................22-1 

22.1 Overview..................................................................................................................... 22-1 

22.2 Detailed Description of Diaphragm Rehabilitation Techniques................................. 22-1 

22.2.1 Wood Diaphragm Strengthening ................................................................ 22-1 

22.2.2 Add or Enhance Chord in Existing Wood Diaphragm ............................. 22-11 

22.2.3 Add or Enhance Diaphragm Cross-ties for Out-of-Plane 

Wall-to-Diaphragm Loads in Flexible Wood and Steel Diaphragms....... 22-16 

22.2.4 Infill Opening in a Concrete Diaphragm .................................................. 22-26 

22.2.5 Add Fiber-Reinforced Polymer Composite Overlay 

to a Concrete Diaphragm.......................................................................... 22-28 

22.2.6 Infill Opening in a Concrete Fill On Metal Deck Diaphragm.................. 22-30 

22.2.7 Increase Shear Capacity of Unfilled Metal Deck Diaphragm .................. 22-32 

22.2.8 Enhance Masonry Flat Arch Diaphragm .................................................. 22-33 

22.2.9 Add Horizontal Braced Frame as a Diaphragm........................................ 22-38 

22.2.10 Improve Tension Rod Horizontal Steel Bracing ...................................... 22-41 

22.2.11 Improve Shear Transfer in Precast Concrete Diaphragm ......................... 22-42 

22.3 References................................................................................................................. 22-45 

Chapter 23 - Foundation Rehabilitation Techniques ...........................................................23-1 

23.1 Overview..................................................................................................................... 23-1 

23.2 General Goals for Seismic Rehabilitation of Foundations ......................................... 23-2 

23.3 Construction Issues ..................................................................................................... 23-2 

23.4 Analytical Issues ......................................................................................................... 23-3 
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23.5 Increasing Estimates of Capacity by In-Situ Testing.................................................. 23-5 

23.5.1 Plate Bearing Tests ..................................................................................... 23-6 

23.5.2 Static Tests on New Deep Foundations...................................................... 23-8 

23.5.3 Static Load Tests on Existing Deep Foundations..................................... 23-11 

23.6 New Foundations ...................................................................................................... 23-11

23.6.1 Types of New Foundations Commonly Used in Seismic Rehabilitation . 23-11 

23.6.2 Add Shallow Foundation Next to Existing Shallow Foundation.............. 23-11 

23.6.3 Add Shallow Foundation Next to Existing Deep Foundation .................. 23-16 

23.6.4 Add Deep Foundation Next to Existing Shallow Foundation .................. 23-16 

23.6.5 Add Deep Foundation Next to Existing Deep Foundation....................... 23-16 

23.7 Structural Rehabilitation for Existing Shallow Foundations .................................... 23-16 

23.7.1 Goals......................................................................................................... 23-16 

23.7.2 Add Micropiles Adjacent to an Existing Strip Footing ............................ 23-18 

23.7.3 Enlarge or Replace an Existing Spread Footing....................................... 23-22 

23.8 Structural Rehabilitation for Existing Deep Foundations......................................... 23-25 

23.8.1 Goals......................................................................................................... 23-25 

23.8.2 Add a Mat Foundation, Extended Pile Cap or Grade Beam..................... 23-25 

23.8.3 Add Drilled Piers to an Existing Drilled Pier Foundation........................ 23-27 

23.8.4 Add Micropiles to an Existing Drilled Pier Foundation........................... 23-29 

23.8.5 Add Top Bars to an Existing Pile Cap...................................................... 23-31 

23.9 Ground Improvement for Existing Shallow and Deep Foundations......................... 23-33 

23.9.1 Goals......................................................................................................... 23-33 

23.9.2 Compaction Grouting ............................................................................... 23-33 

23.9.3 Permeation Grouting Under Existing Shallow and Deep Foundations .... 23-36 

23.10 Mitigating the Impacts of Other Ground Hazards on Existing Foundations ............ 23-41 

23.10.1 Issues to be Addressed.............................................................................. 23-41 

23.10.2 Fault Rupture ............................................................................................ 23-41 

23.10.3 Lateral Spreading...................................................................................... 23-42 

23.10.4 Seismic-Induced Landslide....................................................................... 23-43 

23.11 References................................................................................................................. 23-47 

Chapter 24 - Reducing Seismic Demand ...............................................................................24-1 

24.1 Overview..................................................................................................................... 24-1 

24.2 Reduction of Seismic Weight ..................................................................................... 24-1 

24.3 Seismic Isolation......................................................................................................... 24-5 

24.4 Energy Dissipation...................................................................................................... 24-6 

24.5 References................................................................................................................... 24-9 

Glossary .....................................................................................................................................G-1 

Abbreviations ............................................................................................................................ A-1 

Workshop Participants..........................................................................................................WP-1 
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List of Figures 

Figure 1.6-1: Organization of Chapters and Parts......................................................................................1-5

Figure 5.1-1: Building Type W1: Wood Light Frames, One- and Two-Family Detached Dwelling.......5-1

Figure 5.4.1-1: Shear Wall Design Methods............................................................................................5-11

Figure 5.4.1-2: Preferred and Less Preferred Shear Wall Locations .......................................................5-12

Figure 5.4.1-3: Overturning Support Conditions for Upper Story Shear Walls ......................................5-13

Figure 5.4.1-4: New Continuous Foundation Cast Along Side Existing to Provide Capacity 

for Tie-Down Anchor......................................................................................................................5-15

Figure 5.4.1-5: Load Path from Roof Diaphragm to Top of Shear Wall .................................................5-18

Figure 5.4.1-6: Attachment of Blocking to Existing Sheathing...............................................................5-19

Figure 5.4.1-7: Load Path From Floor Diaphragm to Lower Story Shear Wall ......................................5-22

Figure 5.4.1-8: Load Path From Floor Diaphragm to Lower Story Shear Wall – Balloon Framing .......5-23

Figure 5.4.1-9: Load Path from Upper Story Wall-To-Floor Diaphragm................................................5-24

Figure 5.4.1-10: Load Path at Disruption in Shear Wall Sheathing.........................................................5-25

Figure 5.4.1-11: Load Path for Overturning (Tension and Compression) at Upper Story Shear Wall....5-26

Figure 5.4.1-12: Load Path for Overturning (Tension and Compression) at Foundation........................5-27

Figure 5.4.1-13: Reduction of Slender Shear Wall Effective Height.......................................................5-27

Figure 5.4.2-1: Collector from Garage Open Front to Adjacent Dwelling..............................................5-29

Figure 5.4.2-2: Collector from Garage Open Front to Adjacent Dwelling..............................................5-30

Figure 5.4.3-1: Added Anchor Bolt at Existing Concrete Foundation.....................................................5-33

Figure 5.4.3-2: Added Anchor Bolt at Existing Partially Grouted Concrete Masonry Foundation.........5-34

Figure 5.4.3-3: Anchorage to Existing Foundation Using Proprietary Connectors.................................5-35

Figure 5.4.3-4: Pier and Curtain Wall Foundation System with Inadequate Load Path 

Between Shear Walls and Foundation.............................................................................................5-36

Figure 5.4.4-1: Cripple Wall Enhancement .............................................................................................5-41

Figure 5.4.5-2: Anchorage of Floor Diaphragm Framing to Uphill Foundation in a Hillside Dwelling .5-46

Figure 5.4.5-3: Connections for Anchorage to Uphill Foundation ..........................................................5-47

Figure 5.4.6-1: Infill and Bracing of Masonry Chimney .........................................................................5-49

Figure 5.4.6-2A: Bracing of Masonry Chimney ......................................................................................5-51

Figure 5.4.6-2B: Bracing of Masonry Chimney ......................................................................................5-52

Figure 6.1-1: Building Type W1A: Multistory Multi-Unit Residential Woodframes ...............................6-1

Figure 6.4.1-1A: Elevation of Steel Moment Frame in W1A Building...................................................6-10

Figure 6.4.1-1B: Shear Transfer Between Moment Frame Beam and Diaphragm..................................6-11

Figure 6.4.1-1C: Shear Transfer from Moment Frame Beam to Collector..............................................6-12

Figure 6.4.1-1D: Shear Transfer from Moment Frame Beam to Collector..............................................6-13

Figure 6.4.2-1: Added Shear Wall Supported on Existing Foundation and Slab.....................................6-17

Figure 6.4.2-2: Added Shear Wall Supported by New and Existing Footings ........................................6-19

Figure 6.4.2-3: Added Shear Wall Supported on a New Footing ............................................................6-20

Figure 6.4.2-4: Enhanced Shear Wall Sheathing Fastening.....................................................................6-21

Figure 6.4.2-5: Enhanced Shear Wall With Sheathing Overlay ..............................................................6-22

Figure 6.4.3-1: Enhanced Overturning Support for Upper Story Shear Wall..........................................6-26

Figure 6.4.3-2: Enhanced Beam Supporting Discontinued Shear Wall...................................................6-27

Figure 6.4.4-1A: Shear Wall Elevation with Enhanced Overturning Detailing.......................................6-29

Figure 6.4.4-1B: Framing Fastening for Overturning Load Path.............................................................6-30

Figure 6.4.4-1C: Tie-down Details at Alternate Base Conditions ...........................................................6-31

Figure 6.4.5-1: Collector Details..............................................................................................................6-36

Figure 6.4.5-2: Collector Using Existing Framing Parallel to the Shear Wall.........................................6-38
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Figure 6.4.5-3: Collector Using Added Blocking at Framing Perpendicular to Shear Wall....................6-39

Figure 7.1-1: Building Type W2: Woodframes, Commercial and Industrial ............................................7-1

Figure 7.4.1-1A: Steel Braced Frame Added in W2 Building...................................................................7-7

Figure 7.4.1-1B: Shear Transfer and Collector for Steel Braced Frame....................................................7-7

Figure 7.4.1-1C: Shear Transfer and Collector for Steel Braced Frame....................................................7-8

Figure 7.4.1-2: Two-Story Steel Braced Frame in a W2 Building ............................................................7-8

Figure 8.1-1: Building Type S1: Steel Moment Frames............................................................................8-1

Figure 8.4.1-1: HSS Brace at Existing Beam-Column Connection in SCBF ..........................................8-10

Figure 8.4.1-2: HSS Brace at Existing Beam-Column Connection in Ordinary CBF .............................8-11

Figure 8.4.1-3: HSS Braced at Existing Beam-Column Connection .......................................................8-12

Figure 8.4.2-1: Cast-in-Place Concrete Wall at Existing Beam...............................................................8-16

Figure 8.4.2-2: Cast-in-Place Concrete Wall Encasing Existing Column ...............................................8-17

Figure 8.4.2-3: Discontinuous Wall at Existing Beam ............................................................................8-18

Figure 8.4.2-4: Wall at Existing Column.................................................................................................8-19

Figure 8.4.2-5: Combined Shotcrete and Cast-In-Place Construction.....................................................8-21

Figure 8.4.3-1: Cover Plate at Existing Beam .........................................................................................8-24

Figure 8.4.3-2: Box Section at Existing Column .....................................................................................8-24

Figure 8.4.4-1: Plate Collectors at Existing Beam...................................................................................8-26

Figure 8.4.5-1: Modified Base Plate to Increase Uplift Capacity ............................................................8-28

Figure 8.4.5-2: Concrete Pedestal at Existing Column............................................................................8-29

Figure 8.4.6-1: Reduced Beam Section at Bottom Flange of Existing Beam..........................................8-35

Figure 8.4.6-2: Welded Haunch at Bottom Flange of Existing Beam .....................................................8-36

Figure 8.4.7-1: Welded Splice Upgrade at Existing Column...................................................................8-40

Figure 8.4.7-2: Bolted Splice Upgrade at Existing Column ....................................................................8-41

Figure 8.4.8-1A: Unstiffened Steel Plate Shear Wall ..............................................................................8-43

Figure 8.4.8-1B: Fin Plate Connection Options.......................................................................................8-44

Figure 9.1-1: Building Type S2: Steel Braced Frames ..............................................................................9-1

Figure 10.1-1: Building Type S4: Steel Frames with Concrete Shear Walls...........................................10-1

Figure 11.1-1: Building Type S5: Steel Frames with Infill Masonry Shear Walls..................................11-1

Figure 12.1-1: Building Type C1: Concrete Moment Frames .................................................................12-1

Figure 12.4.1-1: Typical Braced Frame Configurations ..........................................................................12-8

Figure 12.4.1-2A: Typical Connection to Concrete Diaphragm..............................................................12-9

Figure 12.4.1-2B: Typical Connection to Existing Concrete Beam ......................................................12-10

Figure 12.4.1-2C: Typical Connection to Existing Concrete Column...................................................12-11

Figure 12.4.1-3: Test Specimen Connection Detail for Braced Frame..................................................12-12

Figure 12.4.1-4: Braced Frame to Concrete Column Connection .........................................................12-15

Figure 12.4.1-5: Braced Frame to Slab Connection...............................................................................12-16

Figure 12.4.2-1A: Concrete Wall Connection to Concrete Slab............................................................12-19

Figure 12.4.2-1B: Concrete Wall Connection to Concrete Slab – Partial Elevation View....................12-20

Figure 12.4.2-1C: Connection of Concrete Wall to Concrete Joists or Waffle Slab .............................12-22

Figure 12.4.2-1D: Concrete Wall Connection to Waffle Slab – Partial Elevation View.......................12-23

Figure 12.4.3-1: Concrete Collector at Concrete Slab ...........................................................................12-27

Figure 12.4.3-2: Concrete Collector at Waffle Slab ..............................................................................12-28

Figure 12.4.3-3: Concrete Collector at Existing Beam..........................................................................12-29

Figure 12.4.3-4: Steel Plate Collector....................................................................................................12-30

Figure 12.4.4-1A: Seismic Retrofit of Columns Using FRP Composites..............................................12-34

Figure 12.4.4-1B: Seismic Retrofit of Columns Using FRP Composites..............................................12-35

Figure 12.4.5-1: Concrete and Steel Overlays for Concrete Columns...................................................12-37

Figure 13.1-1: Building Type C2b: Concrete Shear Walls (Bearing Wall Systems)..............................13-1

Figure 13.4.1-1A: Shear Strengthening of Concrete Shear Walls Using FRP Composite ....................13-12

Figure 13.4.1-1B: Fiber Anchor Details ................................................................................................13-13
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Figure 13.4.2-1: Typical Corbel at Linking Slab ...................................................................................13-16

Figure 13.4.2-2: Typical Strengthening of Shallow Coupling Beam.....................................................13-17

Figure 13.4.2-3: Strengthening of Deep Coupling Beam ......................................................................13-19

Figure 13.4.3-1: Added Support and Shear Strength at Slab-Wall Joint ...............................................13-21

Figure 13.4.3-2: Added Shear Capacity at Slab-Wall Joint...................................................................13-22

Figure 14.1-1: Building Type C2f: Concrete Shear Walls (Gravity Frame Systems) .............................14-1

Figure 15.1-1: Building Type C3: Concrete Frames with Infill Masonry Shear Walls ...........................15-1

Figure 16.1-1: Building Type PC1: Tilt-Up Concrete Shear Walls.........................................................16-1

Figure 16.3-1: Plan of PC1 Building with Re-Entrant Corner at In-Set Panels.......................................16-9

Figure 16.3-2: Plan and Detail of Large PC1 Building Constructed in Three Sections ........................16-10

Figure 16.4.1-1A: Roof Plan with Wall Out-of-Plane Anchorage for Flexible Wood Diaphragm .......16-14

Figure 16.4.1-1B: Wall Out-of-Plane Anchorage for Flexible Wood Diaphragm at Subpurlins – 

Roofing Not Removed ..................................................................................................................16-15

Figure 16.4.1-1C: Wall Out-of-Plane Anchorage for Flexible Wood Diaphragm at Subpurlins – 

Roofing Removed .........................................................................................................................16-16

Figure 16.4.1-1D: Wall Out-of-Plane Anchorage for Flexible Wood Diaphragm at Purlins ...............16-17

Figure 16.4.1-2A: Roof Plan with Wall Out-of-Plane Anchorage for Flexible Steel Diaphragm.........16-18

Figure 16.4.1-2B: Wall Out-of-Plane Anchorage for Flexible Steel Diaphragm – 

to Decking for Load Parallel to Flutes ..........................................................................................16-19

Figure 16.4.1-2C: Steel Open Web Joist Anchorage to Exterior Wall ..................................................16-20

Figure 16.4.2-1: Enhanced Girder Connection – Collar at Pilaster .......................................................16-22

Figure 16.4.2-2: Enhanced Girder Connection at U-hanger ..................................................................16-23

Figure 16.4.2-3: Enhanced Girder Connection at Pilaster .....................................................................16-24

Figure 16.4.3-1: Enhancement of Tilt-up Panel Base Connection.........................................................16-25

Figure 17.1-1: Building Type PC2: Precast Concrete Gravity Frames with Shear Walls .......................17-1

Figure 17.2-1: Plan of Common Parking Structure Configuration ..........................................................17-4

Figure 17.4.1-1: Added or Enhanced Chord or Collector at Floor Perimeter 

with Cast-In-Place Topping Slab ..................................................................................................17-12

Figure 17.4.1-2: Added or Enhanced Chord or Collector at Floor Interior 

with Cast-In-Place Topping Slab ..................................................................................................17-12

Figure 17.4.1-3: Added or Enhanced Chord or Collector at Floor Interior with 

Cast-In-Place Topping Slab...........................................................................................................17-13

Figure 17.4.1-4: Added Collector Anchorage to Shear Wall with 

Cast-In-Place Topping Slab...........................................................................................................17-14

Figure 17.4.1-5: Rebar Embedment Considerations in Collector Anchorage 

to Existing Shear Wall...................................................................................................................17-15

Figure 17.4.1-6: Steel Chord or Collector at Floor Perimeter 

without Cast-In-Place Topping .....................................................................................................17-16

Figure 17.4.1-7: Steel Chord or Collector at Floor Perimeter without Cast-In-Place Topping .............17-16

Figure 17.4.2-1: Added or Enhanced Precast Wall-to-Foundation Connection ....................................17-19

Figure 17.4.2-2: Precast Wall Connection.............................................................................................17-20

Figure 17.4.2-3: Modification of Demand on Anchors Through Use of Panel-to-Panel Connections..17-21

Figure 17.4.2-4: Panel Setup for Testing of FRP Panel-to-Panel Connections .....................................17-22

Figure 17.4.3-1: Accidental Reduction of Column Height Due to Guardrail........................................17-23

Figure 17.4.3-2: Accidental Increase in Column Fixity Due to Embedment into Grade.......................17-23

Figure 17.4.3-3: Accidental Reduction of Column Height at Ramp......................................................17-24

Figure 18.1-1: Building Type RM1t: Reinforced Masonry Bearing Walls .............................................18-1

Figure 19.1-1: RM1u Building Type: Reinforced Masonry Bearing Wall.............................................19-1

Figure 20.1-1: RM2 Building Type: Reinforced Masonry Bearing Walls ..............................................20-1

Figure 21.1-1: Building Type URM: Unreinforced Masonry Bearing Walls..........................................21-1

Figure 21.4.1-1: Parapet Bracing ...........................................................................................................21-10
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Figure 21.4.1-2: Parapet Removal and Concrete Cap Beam .................................................................21-11

Figure 21.4.2-1: Drilled Dowels ............................................................................................................21-13

Figure 21.4.2-2: Through Bolt Anchor ..................................................................................................21-14

Figure 21.4.2-3: Tension Anchors Installed from Below the Floor.......................................................21-17

Figure 21.4.2-4: Tension Anchors Installed from Above the Floor.......................................................21-18

Figure 21.4.2-5: Floor-to-Wall Shear Anchors......................................................................................21-19

Figure 21.4.2-6: Bond Beam at a Sloping Roof.....................................................................................21-20

Figure 21.4.2-7: Bond Beam at a Sloping Roof with Limited Clearance ..............................................21-20

Figure 21.4.2-8: Tension Tie Connection Issues ...................................................................................21-21

Figure 21.4.2-9: Wall-to-Floor Tension Tie in Hollow Masonry ..........................................................21-24

Figure 21.4.2-10: Wall-to-Floor Tension Tie in Hollow Masonry Alternate ........................................21-25

Figure 21.4.3-1: Exposed Out-of-Plane Wall Bracing...........................................................................21-26

Figure 21.4.3-2: Vertical Bracing Alternatives......................................................................................21-27

Figure 21.4.3-3: Concrete Ribs in Hollow Masonry..............................................................................21-29

Figure 21.4.3-4: Connection of Strongback to Hollow Masonry...........................................................21-30

Figure 21.4.4-1A: Reinforced Cores......................................................................................................21-32

Figure 21.4.4-1B: Plan Detail of Reinforced Core in Masonry Wall ....................................................21-33

Figure 21.4.5-1: Concrete or Shotcrete Wall Overlay ...........................................................................21-36

Figure 21.4.5-2A: Potential Damage at Ends of Narrow Concrete Overlay..........................................21-38

Figure 21.4.5-2B: Alternatives to Distribute Overturning Loads in Concrete Overlay.........................21-39

Figure 21.4.5-2C: Alternatives to Distribute Overturning Loads in Concrete Overlay.........................21-40

Figure 21.4.6-1: Fiber Composite Wall Overlay on URM Wall............................................................21-44

Figure 21.4.7-1: Infilling an Opening in a URM Wall ..........................................................................21-46

Figure 21.4.8-1: Connecting a New Concrete Wall to an Existing Wood Floor ...................................21-48

Figure 21.4.8-2: Connecting a Concrete Wall to an Existing Wood Roof at Interior............................21-49

Figure 21.4.9-1: New Perimeter Steel Moment Frame to an Existing Wood Floor ..............................21-51

Figure 21.4.9-2: New Interior Steel Moment Frame to an Existing Wood Floor..................................21-52

Figure 21.4.10-1: Strengthen Existing Crosswalls.................................................................................21-54

Figure 21.4.10-2: Add New Crosswalls.................................................................................................21-55

Figure 21.4.10-3: Add New Moment Frame as Crosswall ....................................................................21-56

Figure 21.4.11-1: Supplemental Vertical Support .................................................................................21-58

Figure 21.4.12-1: Veneer Ties in Brick Masonry ..................................................................................21-59

Figure 21.4.12-2: Veneer Ties for Stone Masonry Facing.....................................................................21-60

Figure 22.2.1-1: Remove and Replace Existing Wood Sheathing with Wood Structural 

Panel at a Roof ................................................................................................................................22-2

Figure 22.2.1-2: Wood Panel Overlay with Blocking Over Existing Sheathing .....................................22-3

Figure 22.2.1-3: Wood Panel Overlay without Blocking Over Existing Sheathing ...............................22-4

Figure 22.2.1-4: Wood Panel Overlay without Blocking Over Existing Sheathing 

When the Bottom of the Existing Sheathing is Visible...................................................................22-5

Figure 22.2.1-5: Shear Transfer in New Overlay at Existing Partitions ..................................................22-6

Figure 22.2.2-1: Enhanced Chord Member Fastening at Wood Diaphragm .........................................22-12

Figure 22.2.2-2: Enhanced Chord Member and Fastening at Wood Diaphragm...................................22-13

Figure 22.2.2-3: Elevation of Wall Panels with Incomplete Chord Due to 

Vertical Offset in Roof Diaphragm ...............................................................................................22-14

Figure 22.2.3-1A: Roof Plan with Diaphragm Cross-Tie System Using Subdiaphragms, 

Shown for Wood Diaphragm ........................................................................................................22-17

Figure 22.2.3-1B: Subdiaphragm for Flexible Wood Diaphragm – Roofing Not Removed .................22-18

Figure 22.2.3-1C: Subdiaphragm for Flexible Wood Diaphragm – Roofing Removed ........................22-19

Figure 22.2.3-1D: Enhanced Wood Subdiaphragm with Added Wood Structural Panel Soffit............22-20

Figure 22.2.3-1E: Subdiaphragm for Flexible Wood Diaphragm at Purlins .........................................22-21

Figure 22.2.3-1F: Cross-Tie for Flexible Wood Diaphragm at Glulam Beams.....................................22-22
Techniques for the Seismic Rehabilitation of

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xix 

Figure 22.2.3-1G: Cross-Tie for Flexible Wood Diaphragm at Purlins.................................................22-23

Figure 22.2.3-2A: Roof Plan with Diaphragm Cross-Tie System Using Direct Ties, 

Shown for Steel Diaphragm ..........................................................................................................22-24

Figure 22.2.3-2B: Cross-Tie for Flexible Steel Diaphragm...................................................................22-25

Figure 22.2.3-2C: Steel Open Web Joist Connection for Diaphragm Cross-Ties .................................22-25

Figure 22.2.3-2D: Steel Open Web Joist Connection for Diaphragm Cross-Ties .................................22-26

Figure 22.2.4-1: Typical Infill Opening in a Concrete Diaphragm........................................................22-27

Figure 22.2.5-1: Shear Strengthening of Concrete Diaphragm Using FRP Composite.........................22-30

Figure 22.2.8-1: Failure Scenarios for Masonry Flat Arch Floors.........................................................22-34

Figure 22.2.8-2: Add Wall-to-Diaphragm Ties and Chord for Masonry Flat Arch Floor - 

Access from Below the Floor........................................................................................................22-35

Figure 22.2.8-3: Add Wall-to-Diaphragm Ties and Chord for Masonry Flat Arch Floor - 

Access from Above the Floor........................................................................................................22-36

Figure 22.2.8-4: Masonry Flat Arch Floor Strengthening .....................................................................22-37

Figure 22.2.9-1: Diaphragm Strengthening using Horizontal Braced Frame ........................................22-40

Figure 22.2.9-2: Horizontal Braced Frame Connection.........................................................................22-41

Figure 22.2.10-1: Tension Rod Connection at Wall ..............................................................................22-42

Figure 23.4-1: Modeling Approaches for Buildings with Basements......................................................23-4

Figure 23.5.1-1: Plate Bearing Tests for In-Situ Bearing Capacity Determination .................................23-7

Figure 23.5.2-1: Static Pile Load Test in Compression ...........................................................................23-9

Figure 23.5.2-2: Static Pile Load Test in Tension .................................................................................23-10

Figure 23.6.1-1: Types of New Foundations Commonly Used in Seismic Rehabilitation....................23-12

Figure 23.6.2-1: New Concrete Strip Footing Next to Existing Strip Footing ......................................23-13

Figure 23.6.2-2: New Shallow Footing is Deeper than Existing Shallow Footing................................23-14

Figure 23.6.2-3: New Shallow Footing Does Not Need to be as Deep as Existing Shallow Footing ...23-15

Figure 23.6.4-1: New Drilled Pier Next to Existing Strip Foundation ..................................................23-17

Figure 23.7.2.-1: Micropile Enhancement to Existing Strip Footing.....................................................23-19

Figure 23.7.2.-2: Micropile Details........................................................................................................23-21

Figure 23.7.3-1: Enlarge Existing Spread Footing.................................................................................23-23

Figure 23.7.3-2: Replace Existing Spread Footing ................................................................................23-24

Figure 23.8.2-1: New Mat Foundation Between Existing Drilled Piers ................................................23-26

Figure 23.8.3.-1: Adding Drilled Piers to an Existing Drilled Pier Foundation ....................................23-28

Figure 23.8.4.-1: Micropile Enhancement of an Existing Drilled Pier Footing.....................................23-30

Figure 23.8.5-1: Adding Top Bars to an Existing Pile Cap ...................................................................23-32

Figure 23.9.2-1: Compaction Grouting Under Existing Shallow Foundations – Pilot Test Program....23-35

Figure 23.9.2-2: Compaction Grouting Under Existing Deep Foundations – Pilot Test Program ........23-37

Figure 23.9.3-1: Permeation Grouting of Loose Sand Under Existing Shallow Foundation.................23-38

Figure 23.9.3-2: Permeation Grouting of Liquefiable Layer Under Existing Shallow Foundation.......23-39

Figure 23.9.3-3: Permeation Grouting of Liquefiable Layer Around Existing Deep Foundation .........23-40

Figure 23.10.2-1: Active Fault Traversing the Footprint of Existing Building .....................................23-42

Figure 23.10.3-1: Lateral Spreading of Soil Layer Underlying Existing Building................................23-43

Figure 23.10.3-2: Mitigation Measure for Lateral Spreading – Stone/Gravel Column .........................23-44

Figure 23.10.4-1: Landslide Hazards.....................................................................................................23-45

Figure 23.10.4-2: Mitigation of Landslide Hazards...............................................................................23-46

Figure 24.2-1: Generic Building Example of Decreasing Base Shear with Decreasing Height..............24-2

Figure 24.2-2: Generic Building Example of Increasing Base Shear with Decreasing Height ...............24-3

Figure 24.2-3: Real Building Example of Increasing Base Shear with Decreasing Height.....................24-4

Figure 24.4-1: Damper Alternatives for Rehabilitating an Existing Moment Frame ..............................24-7

Figure 24.4-2: Additional Damper Alternatives for Rehabilitating an Existing Moment Frame ............24-8
Techniques for the Seismic Rehabilitation of

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List of Tables 

Table 4-1: Model Building Types..............................................................................................................4-2

Table 5.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for W1 Buildings................5-6

Table 6.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for W1A Buildings.............6-7

Table 7.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for W2 Buildings................7-4

Table 8.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S1/S1A Buildings.........8-4

Table 9.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S2/S2A Buildings.........9-4

Table 10.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S4 Buildings .............10-3

Table 11.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S5/S5A Buildings.....11-5

Table 12.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C1 Buildings.............12-4

Table 13.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C2b Buildings...........13-5

Table 14.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C2f Buildings ...........14-5

Table 15.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C3/C3A Buildings ....15-4

Table 16.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for PC1 Buildings...........16-7

Table 17.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for PC2 Buildings...........17-6

Table 18.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for RM1t Buildings ........18-4

Table 19.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for RM1u Buildings .......19-3

Table 20.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for RM2 Buildings .........20-4

Table 21.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for URM Bearing Wall 

Buildings .........................................................................................................................................21-5
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 1 – Introduction

1-1 

Chapter 1 - Introduction 

1.1 Overview 

A considerable number of buildings in the existing building stock of the United States present a 

risk of poor performance in earthquakes because there was no seismic design code available or 

required when they were constructed, because the seismic design code used was immature and 

had flaws, or because original construction quality or environmental deterioration has 

compromised the original design. 

The practice of improving the seismic performance of existing buildings—known variously as 

seismic rehabilitation, seismic retrofitting, or seismic strengthening—began in the U.S. in 

California in the 1940s following the Garrison Act in 1939. This Act required seismic 

evaluations for pre-1933 school buildings. Substandard buildings were required to be retrofit or 

abandoned by 1975. Many school buildings were improved by strengthening, particularly in the 

late 1960s and early 1970s as the deadline approached. Local efforts to mitigate the risks from 

unreinforced masonry buildings (URMs) also began in this time period. In 1984, the Federal 

Emergency Management Agency (FEMA) began its program to encourage the reduction of 

seismic hazards posed by existing older buildings throughout the country. This program has 

included development of many resources to assist engineers and other stakeholders to reduce this 

risk; guidance on evaluation, costs and priorities; and ultimately, a comprehensive, performancebased, rehabilitation design guideline, FEMA 273, NEHRP Guidelines for the Seismic 

Rehabilitation of Buildings (FEMA, 1997a)—which was converted to FEMA 356 (FEMA, 

2000a) as an American Society of Civil Engineers (ASCE) prestandard. At this writing, ASCE 

is developing a standard entitled ASCE 41, Seismic Rehabilitation of Existing Buildings, using 

FEMA 356 as a basis. 

Recognizing that building rehabilitation design is far more constrained than new building design 

and that special techniques are needed to insert new lateral elements, tie them to the existing 

structure, and generally develop complete seismic load paths, a document was published for this 

purpose in 1992. FEMA 172, NEHRP Handbook of Techniques for the Seismic Rehabilitation of 

Existing Buildings (FEMA, 1992b), was intended to identify and describe generally accepted 

rehabilitation techniques. The art and science of seismic rehabilitation has grown tremendously 

since that time with federal, state, and local government programs to upgrade public buildings, 

with local ordinances that mandate rehabilitation of certain building types, and with a growing 

concern among private owners about the seismic performance of their buildings. In addition, 

following the demand for better understanding of performance of older buildings and the need 

for more efficient and less disruptive methods to upgrade, laboratory research on the subject has 

exploded worldwide, particularly since the nonlinear methods proposed for FEMA 273 became 

developed. 

The large volume of rehabilitation work and research now completed has resulted in 

considerable refinement of early techniques and development of many new techniques, some 

confined to the research lab and some widely used in industry. Like FEMA 172, this document 

describes the techniques currently judged to be most commonly used or potentially to be most 

useful. Furthermore, it has been formatted to take advantage of the ongoing use of typical 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 1 – Introduction

1-2 

building types in FEMA documents concerning existing buildings, and to facilitate the addition 

of techniques in the future. 

1.2 Purpose and Goals 

The primary purpose of this document is to provide a selected compilation of seismic 

rehabilitation techniques that are practical and effective. The descriptions of techniques include 

detailing and constructability tips that might not be otherwise available to engineering offices or 

individual structural engineers who have limited experience in seismic rehabilitation of existing 

buildings. A secondary purpose is to provide guidance on which techniques are commonly used 

to mitigate specific seismic deficiencies in various model building types. 

The goals of the document are to: 

Describe rehabilitation techniques commonly used for various model building types 

Incorporate relevant research results 

Discuss associated details and construction issues 

Provide suggestions to engineers on the use of new products and techniques 

1.3 Audience 

This document was written primarily for engineers who are inexperienced in seismic 

rehabilitation, or who provide these services infrequently. Secondarily, the material will be 

useful for architects and project managers coordinating rehabilitation projects or programs to 

better appreciate the potential scope and construction needs of such work. 

1.4 Scope 

This document is intended to describe the most common seismic rehabilitation techniques used 

for each type of building represented in the set of standard building types often used in FEMA 

seismic publications (see Chapter 4). The basics of seismic building engineering are not 

included herein nor are methods and procedures to seismically evaluate buildings. 

It is presumed that the user has a completed seismic evaluation of the building-of-interest, has 

concluded that some level of retrofit is appropriate, and has identified the seismic deficiencies to 

be corrected to achieve the desired performance objective. 

In this document, technique is used to describe a local action consisting of insertion of a new 

lateral force-resisting component or enhancement of the seismic resistance of an in-situ 

component in an existing building. A complete seismic rehabilitation scheme may consist of the 

use of several techniques. Detailed guidance on the strategies to develop such overall schemes is 

not included in this document, although a general discussion of the topic is given in Chapter 3. 

The overall organization of the document is intended to lead the user toward selection of 

realistic, practical, and cost-effective techniques to mitigate a given deficiency. 

The building types making up the FEMA set are described in Chapter 4. The building 

descriptions, performance characteristics, and potential mitigation techniques included are aimed 

at a broad, but not all-inclusive, range of buildings that fit into each category. The information 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 1 – Introduction

1-3 

may not apply to all buildings in the category, particularly those with configuration 

characteristics such as unusual story height or number of stories, or extreme irregularities. There 

are also buildings that do not fit neatly into one of the standard building types, but are 

combinations of standard types. Useful guidance can be obtained for such buildings by 

reviewing the recommendations for each type that is partially represented in Part 2. 

Certain important rehabilitation techniques, such as seismic isolation or the addition of damping 

devices, are complex, far reaching, and on a different scale than the common techniques included 

here for each building type. Although these techniques are described briefly in Chapter 24, they 

are not described in the same level of detail as more standard techniques. Users are encouraged 

to consider such techniques and seek more complete guidance from text books, conference and 

seminar proceedings, or from specialty consultants. 

A large number of research projects have been completed or are ongoing to develop new 

products or techniques for seismic rehabilitation in the United States and around the world. This 

document has included the most commonly used techniques at the time of this writing. For the 

rehabilitation of any specific building, products or techniques not included herein may be the 

most appropriate and economical. 

Guidance for selection of the most appropriate technique or combination of techniques is 

covered in general in Chapter 3. Overlapping and sometimes conflicting characteristics of each 

rehabilitation project—such as performance objectives, cost, disruption to occupants, and 

aesthetics—most often control development of the structural rehabilitation scheme and cannot be 

differentiated by building type in the context of this document. 

Seismic rehabilitation of nonstructural components is not included in this document. This broad 

category would include space-enclosing elements such as cladding, partitions, and ceilings; 

building service systems such as mechanical, electrical, and plumbing elements; and contents 

such as medical or laboratory equipment, storage shelves or racks, and furniture. 

1.5 Other Resources 

Technical design standards and analysis techniques can be obtained in documents such as: 

Standard for the Seismic Evaluation of Buildings, ASCE 31-03 (ASCE, 2003) 

Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 356 

(FEMA, 2000a) 

NEHRP Commentary on the Guidelines for Seismic Rehabilitation of Buildings, FEMA 

274 (FEMA, 1997b) 

Seismic Evaluation and Retrofit of Concrete Buildings, ATC 40 (ATC, 1996) 

Recommended Seismic Evaluation and Upgrade Criteria for Existing Welded Steel 

Moment-Frame Buildings, FEMA 351 (FEMA, 2000b) 

Improvement of Nonlinear Static Seismic Procedures, FEMA 440 (FEMA, 2005) 

Evaluation of Earthquake Damaged Concrete and Masonry Wall Buildings: Basic 

Procedures Manual, FEMA 306, (FEMA, 1999) 

International Existing Building Code, 2003 Edition (ICC, 2003) 

Uniform Code for Building Conservation, 1997 Edition (ICBO, 1997) 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 1 – Introduction

1-4 

The benefits to building owners of performance-based design, methods of managing seismic risk, 

and cost-benefit of seismic rehabilitation are discussed in: 

Primer for Design Professionals—Communicating with Owners and Managers of New 

Buildings on Earthquake Risk, FEMA 389 (FEMA, 2004) 

Planning for Seismic Rehabilitation: Societal Issues, FEMA 275 (FEMA, 1997c) 

Financial Management of Earthquake Risk, (EERI, 2000) 

Typical Costs for Seismic Rehabilitation of Existing Buildings, (FEMA, 1994 and 1995) 

A series on incremental seismic strengthening of selected occupancy types includes the 

following documents: 

Incremental Seismic Rehabilitation of School Buildings (K-12), FEMA 395 (FEMA 

2003a) 

Incremental Seismic Rehabilitation of Hospital Buildings, FEMA 396 (FEMA, 2003b) 

Incremental Seismic Rehabilitation of Office Buildings, FEMA 397 (FEMA, 2003c) 

Incremental Seismic Rehabilitation of of Multifamily Apartment Buildings, FEMA 398 

(FEMA, 2004a) 

Incremental Seismic Rehabilitation of Retail Buildings, FEMA 399 (FEMA 2004b) 

Many of these publications can be found on the FEMA-National Earthquake Hazard Reduction 

Program (NEHRP) website: http://www.fema.gov/plan/prevent/earthquake/nehrp.shtm. 

1.6 Organization of the Document 

As shown in Figure 1.6-1, the document is divided into three parts: 

Part 1 (Chapters 1-3) provides background on seismic evaluation, categories of seismic 

deficiencies, classes of rehabilitation techniques, and general strategies to develop 

rehabilitation schemes. 

Part 2 (Chapters 4-21) contains detailed descriptions of seismic deficiencies that are 

characteristic of each FEMA model building type and techniques commonly used to 

mitigate them. 

Part 3 (Chapters 22-24) contains chapters on seismic rehabilitation techniques common to 

multiple building types such as those related to diaphragms and foundations. A chapter is 

also included in Part 3 describing significant global techniques that could be applied to 

any building, such as seismic isolation or the addition of damping. 

An important aspect of the organization is to provide for flexible expansion of the material with 

future stand-alone printed documents, digital media, or with complete republication. Examples 

of such expansions include a chapter on nonstructural risk mitigation and descriptions of 

additional techniques not included in this edition or developed from future research results. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 1 – Introduction

1-5 

Figure 1.6-1: Organization of Chapters and Parts 

Part 1: Overview 

1. Introduction 

2. Seismic Vulnerability 

3. Seismic Rehabilitation 

Part 3: Rehabilitation Techniques Common to 

Multiple Model Building Types 

22. Diaphragm Rehabilitation Techniques 

23. Foundation Rehabilitation Techniques 

24. Reducing Seismic Demand 

Part 2: Rehabilitation Techniques Associated with 

Individual FEMA Model Building Types 

4. FEMA Model Building Types 

5. W1 

6. W1A 

7. W2 

8. S1/S1A 

9. S2/S2A 

10. S4 

11. S5/S5A 

12. C1 

13. C2b 

14. C2f 

15. C3/C3A 

16. PC1 

17. PC2 

18. RM1t 

19. RM1u 

20. RM2 

21. URM 

Chapter Organization 

12.1 Description of the Model Building Type 

12.2 Seismic Response Characteristics 

12.3 Common Seismic Deficiencies and Applicable 

Rehabilitation Techniques 

12.4 Detailed Description of Techniques 

12.4.1 Add Steel Braced Frame 

12.4.2 Add Concrete or Masonry Shear Wall 

12.4.3 Provide a Collector 

12.4.4 FRP Overlay of a Concrete Column 

12.4.5 Concrete/Steel Overlay of Column 

12.4.6 Enhance Concrete Moment Frame 

12.5 References 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 1 – Introduction

1-6 

1.6.1 Part 1 – Overview 

Chapter 2 gives a brief overview of evaluation methods and how seismic deficiencies, in general, 

can be placed into categories. A set of categories of seismic deficiencies is defined, both because 

such categories are useful to describe appropriate retrofit measures and also because the 

categories are useful as an organization of the chapters covering building types. 

Chapter 3 briefly summarizes various codes, standards, and guidelines that are normally used to 

define design procedures for seismic rehabilitation. These documents provide the numerical 

parameters for design but seldom describe the techniques for strengthening existing components 

or for adding new lateral force-resisting elements to an existing building. 

To relate various seismic rehabilitation techniques to seismic deficiencies, classes of techniques 

are established and described. Similar to the categories of seismic deficiencies defined in 

Chapter 2, these classes of techniques provide a consistent organization for the chapters covering 

building types. 

Finally, Chapter 3 includes a description of socio-economic characteristics that are common to 

most seismic rehabilitation projects and often control the selection of the rehabilitation scheme. 

1.6.2 Part 2 – Rehabilitation Techniques for FEMA Model Buildings 

This document is primarily organized around the FEMA model building types, first categorized 

in ATC 14 (ATC, 1987) in the late 1980s and then carried forward into FEMA 178 (FEMA, 

1992a) and almost all succeeding FEMA publications on existing buildings. It is expected that 

most users of this document will be interested in finding information on a particular building or 

building type, which suggested this organization. Each building type is therefore assigned a 

chapter. Common seismic deficiencies for each building type are identified and mitigation 

techniques suggested, although it is recognized that most buildings will have multiple 

deficiencies and may require a combination of mitigating actions. The rehabilitation techniques 

commonly used for each building type are identified in each chapter and, if closely associated 

with the building type, described in detail in that chapter. References are given to other chapters 

for other applicable techniques. 

To direct the user to appropriate chapters, the model buildings are briefly described in Chapter 4 

at the beginning of Part 2. 

1.6.3 Part 3 – Rehabilitation Techniques for Deficiencies Common to Multiple 

Building Types 

Although certain diaphragm and foundation deficiencies will be found more often in one 

building type than another, the issues and mitigation techniques are cross-cutting and therefore 

grouped together in Part 3 in Chapters 22 and 23. 

Two important rehabilitation techniques, seismic isolation and added damping, can be applied to 

any building type, are global in nature, and cannot be described as a local technique in the 

context of Part 2. These techniques are therefore described independently in Chapter 24. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 1 – Introduction

1-7 

1.7 Disclaimers 

The seismic rehabilitation techniques and details in this document are intended to provide 

guidance to qualified design professionals. Development of schemes that employ one or more 

techniques in this document is the sole responsibility of the engineer of record for the project. 

The details are not to be used in an actual rehabilitation project without review for technical and 

geometric applicability. In all cases, the details must be completed with additional project 

specific information. 

Some techniques included in the document have been developed using laboratory research. 

Conclusions from selected research and resulting product characteristics have been included in 

the document as a starting point for the design engineer. The adequacy of research methods and 

conclusions has not been verified as part of the development of this document. The search for 

applicable research and evaluation of results was not exhaustive, particularly for research outside 

the United States. Inclusion of research or products does not represent endorsement, and 

exclusion does not necessarily represent lack of confidence. 

1.8 References 

ASCE, 2003, Standard for the Seismic Evaluation of Buildings, ASCE 31-03, Structural 

Engineering Institute of the American Society of Structural Engineers, Reston, VA. 

ATC, 1987, Evaluating the Seismic Resistance of Existing Buildings, ATC-14, Applied 

Technology Council, Redwood City, CA. 

ATC, 1996, The Seismic Evaluation and Retrofit of Concrete Buildings, ATC-40, Applied 

Technology Council, Redwood City, CA. 

EERI, 2000, Financial Management of Earthquake Risk, Earthquake Engineering Research 

Institute, Oakland, CA. 

FEMA, 1992a, NEHRP Handbook for the Seismic Evaluation of Existing Buildings, FEMA 178, 

Federal Emergency Management Agency, Washington, D.C. 

FEMA, 1992b, NEHRP Handbook for the Seismic Rehabilitation of Existing Buildings, FEMA 

172, Federal Emergency Management Agency, Washington, D.C. 

FEMA, 1994, Typical Costs for the Seismic Rehabilitation of Existing Buildings, Volume 1, 

Summary, FEMA 156, Second Edition, Federal Emergency Management Agency, Washington, 

D.C. 

FEMA, 1995, Typical Costs for the Seismic Rehabilitation of Existing Buildings, Volume 2, 

Supporting Documentation, FEMA 157, Second Edition, Federal Emergency Management 

Agency, Washington, D.C. 

FEMA, 1997a, NEHRP Guidelines for the Seismic Rehabilitation of Buildings, FEMA 273. 

Federal Emergency Management Agency, Washington, D.C. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 1 – Introduction

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FEMA, 1997b, NEHRP Commentary on the Guidelines for Seismic Rehabilitation of Buildings, 

FEMA 274, Federal Emergency Management Agency, Washington, D.C. 

FEMA, 1997c, Planning for Seismic Rehabilitation: Societal Issues, FEMA 275, Federal 

Emergency Management Agency, Washington, D.C. 

FEMA, 1999, Evaluation of Earthquake Damaged Concrete and Masonry Wall Buildings: Basic 

Procedures Manual, FEMA 306, Federal Emergency Management Agency, Washington, D.C. 

FEMA, 2000a, Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 

356, Federal Emergency Management Agency, Washington, D.C. 

FEMA, 2000b, Recommended Seismic Evaluation and Upgrade Criteria for Existing Welded 

Steel Moment-Frame Buildings, Federal Emergency Management Agency, Washington, D.C. 

FEMA 2003a, Incremental Seismic Rehabilitation of School Buildings (K-12), FEMA 395, 

Federal Emergency Management Agency, Washington, D.C. 

FEMA 2003b, Incremental Seismic Rehabilitation of Hospital Buildings, FEMA 396, Federal 

Emergency Management Agency, Washington, D.C. 

FEMA 2003c, Incremental Seismic Rehabilitation of Office Buildings, FEMA 397, Federal 

Emergency Management Agency, Washington, D.C. 

FEMA 2004a, Incremental Seismic Rehabilitation of of Multifamily Apartment Buildings, FEMA 

398, Federal Emergency Management Agency, Washington, D.C. 

FEMA 2004b, Incremental Seismic Rehabilitation of Retail Buildings, FEMA 399, Federal 

Emergency Management Agency, Washington, D.C. 

FEMA, 2004, Primer for Design Professionals—Communicating with Owners and Managers of 

New Buildings on Earthquake Risk, FEMA 389, Federal Emergency Management Agency, 

Washington, D.C. 

FEMA, 2005, Improvement of Nonlinear Static Seismic Procedures, FEMA 440, Federal 

Emergency Management Agency, Washington, D.C. 

ICBO, 1997, Uniform Code for Building Conservation, 1997 Edition, International Conference 

of Building Officials, Whittier, California. 

ICC, 2003, International Existing Building Code, 2003 Edition, International Conference of 

Building Officials, Country Club Hills, IL. 
Techniques for the Seismic Rehabilitation of

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2-1 

Chapter 2 - Seismic Vulnerability 

2.1 Introduction 

In this document, a seismic deficiency is defined as a condition that will prevent a building from 

meeting the designated seismic performance objective. The performance objective for a building 

may be established by the choice of a prescriptive evaluation standard, or when using 

performance-based standards or guidelines, may be selected from a range of defined 

performance levels. A building evaluated against standards intended to minimize damage and to 

allow occupancy soon after the event may have significantly more deficiencies than the same 

building evaluated only to prevent collapse. Typically, techniques useful to mitigate a particular 

type of deficiency remain the same regardless of the performance objective, but the extent of the 

mitigating measure required may differ. 

The seismic protection systems for nonstructural components in a building have a profound 

effect on building seismic performance, particularly for higher performance levels and 

particularly in the weeks immediately following an event. However, the techniques for seismic 

retrofit of nonstructural components are relatively straightforward, and this document is devoted 

to structural issues. 

The most important issue when beginning to evaluate the seismic capabilities of an existing 

building is the availability and reliability of structural drawings. Detailed evaluation is 

impossible without framing and foundation plans, layouts of primary lateral force elements, 

reinforcing for concrete structures, and connection detailing for steel structures. Developing asbuilts from field information is extremely difficult, particularly for reinforced concrete, 

reinforced masonry, or structural steel buildings. In most cases, such structures must be 

seismically rehabilitated by placement of a new lateral force-resisting system, with enough 

physical testing performed to determine overall deformation capacity of the existing structure. 

This chapter and this entire document assume that sufficient information is available to perform a 

seismic evaluation that will identify all significant deficiencies. 

There are many different procedures and standards for seismic evaluation available to engineers, 

ranging from highly prescriptive sets of rules developed for a single building type to 

determination of probable performance considering nonlinear cyclic response to earthquake time 

histories. These methods are not delineated or described in detail here, nor are the basic 

principles of building seismic design. Instead, it is assumed that the user has already 

appropriately completed a seismic evaluation of some sort and has thus identified seismic 

deficiencies targeted for mitigation. 

This chapter describes the evaluation process in general terms and introduces categories of 

seismic deficiencies used throughout the document. 

2.2 Seismic Evaluation 

Seismic evaluation of older buildings may be commissioned as part of a municipal, regional, 

state, or federal risk reduction program that includes mandatory evaluation and rehabilitation of 

certain buildings. In these cases, the buildings may be identified by type of structural framing 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 2 – Seismic Vulnerability

2-2 

system, by age, by location, or by a combination of these risk factors. Seismic evaluations may 

also be required 1) by local building officials when alterations are made to a building such as a 

change in occupancy, addition, or revision to the structural system; or 2) as part of an owner’s 

voluntary seismic risk analysis. Lastly, building owners simply may be concerned about their 

economic investment or about post-earthquake use of the buildings. Evaluations that are 

mandated by the governing jurisdiction normally specify a minimum standard to be met. 

Evaluations performed voluntarily by owners are often performance-based—the seismic 

performance of the building is estimated by the engineer, rather than the building characteristics 

being compared to a set of prescriptive rules. 

Some types of evaluation techniques are briefly described below. 

2.2.1 Comparison with Requirements for New Buildings 

Until FEMA began an initiative to reduce the seismic risk from existing buildings in the mid1980s, there were very few standards or guidelines applicable to existing buildings. California 

engineers had developed rules for evaluation and retrofit of unreinforced masonry bearing walls 

buildings, but there was little else. Therefore, seismic adequacy was often determined by 

comparing the older building to the requirements for new buildings. This comparison is often 

difficult or impossible because the older building may include structural materials or systems 

prohibited in the code for new buildings, and it is often impractical to completely remove 

materials or change structural systems. Commonly, a completely new complying seismic system 

was introduced, often at great disruption and cost. Some jurisdictions still use this standard, 

particularly in cases of complete building renovation, but some form of performance-based 

equivalent is preferred.

2.2.2 Prescriptive Standards 

The most notable document available for seismic evaluation of existing buildings is ASCE 31-

03: Seismic Evaluation of Existing Buildings (ASCE, 2003), originally developed by FEMA as 

FEMA 310: Handbook for the Seismic Evaluation of Existing Buildings – A Prestandard 

(FEMA, 1998). FEMA 310 was converted to ASCE 31 as part of the Amercian Society of Civil 

Engineers standardization process. ASCE 31-03 is intended for use on older building and 

recognizes that older and out-moded structural systems may be incorporated in these buildings. 

The seismic life safety provided by a building is judged adequate if the requirements are met and 

many jurisdictions accept this level of performance for their community. 

The federal government has also developed Standards of Seismic Safety for Existing Federally 

Owned and Leased Buildings (NIST, 2002) that includes policy in addition to evaluation 

standards. 

Other prescriptive standards have also been developed, primarily for specific building types, 

such as unreinforced masonry bearing walls, timber residential construction, and tilt-up concrete 

buildings (ICBO, 1997; ICC, 2003). Local jurisdictions also may have a particular interest in a 

narrowly described building type within their region that is common and/or hazardous and may 

develop an appropriate minimum standard. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 2 – Seismic Vulnerability

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2.2.3 Performance-Based Evaluation Using Expected Nonlinear Response 

The most sophisticated and complex seismic evaluation is performed using analytical techniques 

that explicitly consider the expected nonlinear response of the structure in strong shaking. Such 

analysis can be performed for selected past ground motions or using slightly simplified 

techniques such a pushover analysis, as described in ATC 40 (ATC, 1996), FEMA 356 (FEMA, 

2000) , and FEMA 440 (FEMA, 2005). The results of such an analysis must be compared to 

responses associated with certain performance levels such as Immediate Occupancy, Life Safety, 

or Collapse Prevention. In order to use these techniques for evaluation, the governing 

jurisdiction or the owner must select the minimum acceptable performance for the building. 

2.3 Categories of Seismic Deficiencies 

Regardless of the evaluation method used, failure to meet the stipulated criteria will identify 

certain seismic deficiencies. It is convenient for the purposes of discussion and for developing 

strategies for seismic rehabilitation to place these deficiencies into categories. It is recognized 

that many building characteristics identified as a deficiency by a seismic evaluation could be 

identified in more than one category. For example, a shear wall structure with inadequate length 

of walls will probably have a deficiency in both global strength and stiffness. Similarly, a onestory tilt-up building with an inadequate diaphragm could be listed with inadequate global 

strength, inadequate global stiffness, or a diaphragm deficiency. Fortunately, these distinctions 

are not of great importance, because the options of mitigation techniques for a given deficient 

building characteristic are generally the same regardless of the category in which it is placed. As 

indicated above, the categories of seismic deficiencies, coupled with somewhat parallel classes 

of rehabilitation techniques described in Chapter 3, are incorporated to provide a convenient 

organizational format for Part 2. 

The categories of deficiencies used in this document are described below. In Part 2, the 

categories of deficiencies present in an individual building will lead a user to consider certain 

techniques for rehabilitation. Therefore, efficient use of this document is dependent on the user 

understanding the nature of the seismic deficiencies of the building targeted for rehabilitation. 

More building-specific seismic deficiencies that may be characteristic of each building type are 

described in each chapter of Part 2. 

2.3.1 Global Strength 

A deficiency in global strength is common in older buildings either due to a complete lack of 

seismic design or a design to an early code with inadequate strength requirements. However, it 

is seldom the only deficiency and the results of the evaluation must be studied to identify 

deficiencies that may not be mitigated solely by adding strength. 

Global strength typically refers to the lateral strength of the vertically oriented lateral forceresisting system at the effective global yield point, (as defined in documents that use simplified 

nonlinear static procedures based on “pushover” curves), but these concepts will not be described 

in detail here. Refer to FEMA 356 (FEMA, 2000) for details. For degrading structural systems 

characterized by a negative post-yield slope on the pushover curve, a minimum strength 

requirement may also apply as indicated in FEMA 440 (FEMA, 2005). In certain cases, the 

strength will also affect the total expected inelastic displacement and added strength may reduce 

nonlinear demands into acceptable ranges. 
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2-4 

If prescriptive equivalent lateral force methods or linear static procedures have been used for 

evaluation or preliminary rehabilitation analysis, inadequate strength will directly relate to 

unacceptable demand-to-capacity ratios within elements of the lateral force-resisting system. 

2.3.2 Global Stiffness 

Although strength and stiffness are often controlled by the same existing elements or the same 

retrofit techniques, the two deficiencies are typically considered separately. Failure to meet 

evaluation standards is often the result of a building placing excessive drift demands on existing 

poorly detailed components. 

Global stiffness refers to the stiffness of the entire lateral force-resisting system although the lack 

of stiffness may not be critical at all levels. For example, in buildings with narrow walls, critical 

drift levels occur in the upper floors. Conversely, critical drifts most often occur in the lowest 

levels in frame buildings. Stiffness must be added in such a way that drifts are efficiently 

reduced in the critical levels. 

Given an adequate minimum strength level, global nonlinear displacements and thus demands on 

most components in the building are more effectively reduced by increased initial stiffness than 

by increased global strength. 

2.3.3 Configuration 

This deficiency category covers configuration irregularities that adversely affect performance. In 

codes for new buildings, these configuration features are often divided into plan irregularities 

and vertical irregularities. Plan irregularities are features that may place extraordinary demands 

on elements due to torsional response or the shape of the diaphragm. Vertical irregularities are 

created by uneven vertical distribution of mass or stiffness between floors that may result in 

concentration of force or displacement at certain levels. In older existing buildings, such 

irregularities were seldom taken into consideration in the original design and therefore normally 

require retrofit measures to mitigate. 

In prescriptive evaluation methods, features that qualify as irregularities are defined by rules, 

similar to the rules used for new buildings. Evaluation methods that explicitly consider 

nonlinear behavior will normally identify concentrations of force or displacement due to 

configuration and the components affected by these concentrations will be shown to have 

inadequate capacity. 

2.3.4 Load Path 

Although all of the deficiencies described have significant effects on seismic performance, a 

break in the load path, or inadequate strength in the load path, may be considered overarching 

because this deficiency will prevent the positive attributes of the seismic system from being 

effective. The load path is typically considered to extend from each mass in the building to the 

supporting soil. For example, for a panel of cladding, this path would include its connection to 

the supporting floor or floors, the diaphragm and collectors that deliver the load to components 

of the primary lateral force-resisting system (walls, braces, frames, etc.), continuity of these 

components to the foundation, and finally the transfer of loads between foundation and soil. If 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 2 – Seismic Vulnerability

2-5 

the connection of the cladding panel or exterior wall fails and the element falls away from the 

building, the adequacy of the balance of the load path is moot. Similarly, if a new shear wall 

element is added to the exterior of a building as a retrofit measure, its strength and stiffness will 

have no effect if it is not connected adequately to the floor diaphragms. 

Many load path deficiencies are difficult to categorize because the strength deficiency may be 

considered to be part of another element. For example, an inadequate construction joint in a 

shear wall could be considered a load path deficiency or a shear wall deficiency in the category 

of global strength. As previously mentioned, the categorization does not make too much 

difference as long as the deficiency is recognized and mitigated. In this document, local 

connections of panels and walls to the diaphragm, and collectors or other connections to the 

lateral force-resisting elements are considered load path issues. Inadequacies within a lateral 

element such as a shear wall, braced frame, or moment frame are generally associated with the 

element and not considered a load path issue. Inadequacies at the foundation level are generally 

considered foundation deficiencies. 

2.3.5 Component Detailing 

Detailing, in this context, refers to design decisions that affect a component’s or system’s 

behavior beyond the strength determined by nominal demand, often in the nonlinear range. 

Perhaps the most common example of a detailing deficiency is poor confinement in concrete 

gravity columns. Often in older concrete buildings, the expected drifts from the design event 

will exceed the deformation capacity of such columns, potentially leading to degradation and 

collapse. Although the primary gravity load design is adequate, the post-elastic behavior is not, 

most often due to inadequate configuration and spacing of ties. 

Another common example is a shear wall that has adequate length and thickness to resist the 

design shear and moment, but that has been reinforced such that its primary post-elastic behavior 

will be degrading shear failure rather than more ductile flexural yielding. Examples in structural 

steel include braced frames with brittle and weak connections that are unable to develop the 

diagonal brace, or brittle beam-column connections in moment frames that are unable to develop 

the capacity of the frame elements. 

Identification of detailing deficiencies is significant in selection of mitigation strategies because 

acceptable performance often may be achieved by local adjustment of detailing rather than by 

adding new lateral force-resisting elements. In the case of gravity concrete columns, acceptable 

performance often can be more efficiently achieved by enhancing deformation capacity (e.g. by 

adding confinement) than by reducing global deformation demand (e.g. by adding lateral forceresisting elements). 

2.3.6 Diaphragms 

The primary purpose of diaphragms in the overall seismic system is to act as a horizontal beam 

spanning between lateral force-resisting elements. In this document, deficiencies affecting this 

primary purpose, such as inadequate shear or bending strength, stiffness, or reinforcing around 

openings or re-entrant corners, are placed in this category. Inadequate local shear transfer to 

lateral force-resisting elements or missing or inadequate collectors are categorized as load path 

deficiencies. 
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2-6 

Since the purpose, configuration, typical deficiencies, and retrofit of diaphragms are essentially 

independent of specific building types, techniques for rehabilitation are in Part 3, Chapter 22. 

2.3.7 Foundations 

Foundation deficiencies can occur within the foundation element itself, or due to inadequate 

transfer mechanisms between foundation and soil. Element deficiencies include inadequate 

bending or shear strength of spread foundations and grade beams; inadequate axial capacity or 

detailing of piles and piers; and weak and degrading connections between piles, piers, and caps. 

Transfer deficiencies include excessive settlement or bearing failure, excessive rotation, 

inadequate tension capacity of deep foundations, or loss of bearing capacity due to liquefaction. 

Analysis and identification of transfer deficiencies is problematic due to recognition that 

structural movement within the soil may be beneficial, or at least not detrimental, depending on 

the performance objective. Mitigation of apparent transfer deficiencies is often expensive and 

disruptive, adding incentive to more carefully consider their effects. Explicit modeling of soil 

resistance to foundation movement therefore is becoming more common and can affect the 

overall dynamic characteristics of the structure as well as base fixity of rigid elements. 

Similarly, the potential for liquefaction at the site is only a deficiency if the projected surface 

settlement is expected to compromise the performance objective for the building. 

This document assumes that apparent deficiencies in structure-soil transfer mechanisms have 

been confirmed by analysis to warrant mitigation. 

Similar to diaphragms, the issues surrounding foundation retrofit are generally independent of 

specific building types, and have been placed in Part 3, Chapter 23. 

2.3.8 Other Deficiencies 

Deficiencies that do not fit into one of the categories described above can be identified but are 

highly variable and unique. In some cases, such as certain geologic hazards or interaction with 

adjacent buildings, the hazard is created off the building site and may be out of the control of the 

building owner. Standard mitigation techniques cannot be identified for such conditions and are 

not included in this document. The significance of these deficiencies with regard to the 

designated performance objective must be discussed with the owner and if appropriate and 

feasible, mitigation actions developed. In rare cases, replacement of the building, abandonment 

of the site, or creation of a redundant facility may be indicated. 

Some of these potential deficiencies are briefly discussed below. 

Geologic Hazards 

On-site liquefaction can be categorized as a foundation deficiency and mitigated if deemed 

necessary. However, the liquefaction and/or lateral spread of adjacent off-site soils can disrupt 

utility service to the site or even cause lateral movement of the building. 
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Up-slope, offsite landslides or upstream dam failure and flooding can also be identified in 

geologic hazard studies. Similarly, potential slide planes may pass under the site but extend 

beyond the site in such a way that mitigation within the site is impractical. 

Although a rare condition, active fault traces can pass through the site or through the building 

footprint. 

Most of these hazards will not be identified unless a detailed geological hazard study is 

performed, which may not be justified unless exceptionally high performance is needed, or if 

required by the local jurisdiction. If identified, the risk of receiving unacceptable damage must 

be weighed against the cost of local mitigation or alternate means of meeting the owner’s 

requirements. 

The potential effects of these hazards on building foundations and possible mitigating actions are 

discussed in Section 23.10. 

Adjacent Buildings 

When the gap between buildings is insufficient to accommodate the combined seismic 

deformations of the buildings, both may be vulnerable to structural damage from the "pounding" 

action that results when the two collide. This condition is particularly severe when the floor 

levels of the two buildings do not match and the stiff floor framing of one building impacts on 

the more fragile walls or columns of the adjacent building. 

For conditions created by expansion joints that are commonly found in buildings, the slabs 

usually align, and the pounding damage is normally assumed to be a local problem. However, if 

the lateral systems on either side of the joint are of considerably different stiffness or strength, an 

independent analysis of both portions may be inappropriate as loads can be transferred from one 

portion to the other. 

For conditions along property lines or involving party walls, the two buildings likely have 

different ownership, and practical and legal issues may be more significant that technical ones. 

Without a high level of cooperation, performance to the satisfaction of both owners may not be 

possible. 

When one owner owns both adjacent buildings, these legal issues no longer apply, and tying the 

buildings together can focus on the technical issues. Like expansion joints in large buildings, if 

expansion and contraction movements between the structures are expected to be minimal, these 

joints can be structurally closed, eliminating the pounding problem and often increasing the 

options for the location of new seismic elements. 

Deterioration of Structural Materials 

Structural materials that are damaged or seriously deteriorated may have an adverse effect on the 

seismic performance of an existing building during a severe earthquake. Methods and techniques 

for repair of poor workmanship, deterioration, fire, settlement, or earthquake damage are not 

covered in this manual. If significant damage is suspected in a building, a condition assessment 

should be developed and carried out prior to development of a final seismic strengthening 
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scheme. The significance of the damage or deterioration must be evaluated with respect to both 

the existing condition and the proposed seismic strengthening of the building. Structural 

condition assessment is not covered in this document, but appropriate procedures and measures 

are well documented (Ratay, 2005). 

Timber: Common problems with timber members that require rehabilitation include termite 

attack, fungus ("dry rot" or "damp rot"), warping, splitting, checking due to shrinkage, strength 

degradation of fire-retardant wood structural panel in areas where high temperatures exist, or 

other causes. 

Unreinforced masonry: The weakest element in older masonry usually is the mortar joint, 

particularly if significant amounts of lime were used in the mortar and the lime was subsequently 

leached out by exposure to the weather. Thus, cracks in masonry walls caused by differential 

settlement of the foundations or other causes generally will occur in the joints; however, wellbonded masonry occasionally will crack through the masonry unit. 

Unreinforced concrete: Unreinforced concrete may be subject to cracking, spalling, and 

disintegration. Cracking may be due to excessive drying shrinkage during the curing of the 

concrete or differential settlement of the foundations. Spalling can be caused by exposure to 

extreme temperatures or the reactive aggregates used in some western states. Disintegration or 

raveling of the concrete is usually caused by dirty or contaminated aggregates, old or defective 

cement, or contaminated water (e.g., water with a high salt or mineral content). 

Reinforced concrete or masonry: Reinforced concrete and masonry are subject to the same types 

of deterioration and damage as unreinforced concrete and masonry. In addition, poor or cracked 

concrete or masonry may allow moisture and oxygen to penetrate to the steel reinforcement and 

initiate corrosion. The expansive nature of the corrosion byproducts can fracture the concrete or 

masonry and extend and accelerate the corrosion process. 

Structural steel: Poorly configured structural steel members may trap moisture from rainfall or 

condensation under conditions that promote corrosion and subsequent loss of section for the steel 

member. Even well-configured steel members exposed to a moist environment require periodic 

maintenance (i.e., painting or other corrosion protection) to maintain their effective load-bearing 

capacity. Older structural steel buildings often have little or no vapor barrier, particularly at the 

perimeter where failures in the weatherproofing of the cladding can lead directly to exposure to 

moisture. Light structural steel members (e.g., small columns or bracing members) in some 

installations may be subject to damage from heavy equipment or vehicles. While such damage 

may have no apparent detrimental effect on the vertical-load-resisting capacity of the steel 

member, its reserve capacity for resisting seismic forces may be seriously impaired. 

2.4 References 

ASCE, 2003, Standard for the Seismic Evaluation of Buildings, ASCE 31-03, Structural 

Engineering Institute of the American Society of Structural Engineers, Reston, VA. 

ATC, 1996, The Seismic Evaluation and Retrofit of Concrete Buildings, ATC-40, Applied 

Technology Council, Redwood City, CA. 
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FEMA, 1998, Handbook for the Seismic Evaluation of Buildings – A Prestandard, FEMA 310, 

Federal Emergency Management Agency, Washington, D.C. 

FEMA, 2000, Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 

356, Federal Emergency Management Agency, Washington, D.C. 

FEMA, 2005, Improvement of Nonlinear Static Seismic Procedures, FEMA 440, Federal 

Emergency Management Agency, Washington, D.C. 

ICBO, 1997, Uniform Code for Building Conservation, 1997 Edition, International Conference 

of Building Officials, Whittier, CA. 

ICC, 2003, International Existing Building Code, 2003 Edition, International Conference of 

Building Officials, Country Club Hills, IL. 

NIST, 2002, Standards of Seismic Safety for Existing Federally Owned and Leased Buildings, 

ICSSC RP 6, National Institute of Standards and Technology, Gaithersburg, MD. 

Ratay, Robert T., 2005, Structural Condition Assessment, John Wiley & Sons, Inc, Hoboken, NJ. 
Techniques for the Seismic Rehabilitation of

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Chapter 3 - Seismic Rehabilitation 

3.1 Introduction 

This document is primarily intended to provide descriptions of individual construction 

techniques used in seismic rehabilitation rather than to give complete guidance on the far more 

subtle process of developing and designing complete rehabilitation schemes. Although the latter 

may be useful to engineers inexperienced in seismic retrofit or seismic design in general, the 

schematic design process for seismic rehabilitation is complex and, not unlike other civil 

engineering design, often involves more art than science. 

Classes of rehabilitation methods are given in this chapter that address one or more of the 

potential categories of deficiencies described in Chapter 2. As previously mentioned, these 

categories and classes are somewhat arbitrary and sometimes overlap. However, they are 

intended to form a framework and logic for development of alternate overall rehabilitation 

schemes. This chapter describes the classes of rehabilitation methods and issues that commonly 

must be considered when developing overall schemes. 

3.2 Rehabilitation Standards 

Seismic rehabilitation guidelines and standards have developed parallel with, but somewhat 

behind, seismic evaluation documents. Often, however, they are the same. For example, 

minimum standards for URM buildings, developed in California, specified sets of configuration, 

maximum stress, and minimum inter-tie rules that were required. When used in an evaluation 

mode, the evaluator noted what was missing or deficient. When used in the rehabilitation mode, 

the engineer provided what was missing or added strength to eliminate deficiencies. 

However, it may not always be true that the evaluation standard and the rehabilitation standard 

are the same. Some engineers and policy-makers believe that the evaluation threshold should be 

set at a very minimum acceptable level because of the cost and disruption of rehabilitation, but 

that once rehabilitation is required, a higher, more reliable standard should be used. This is 

currently the case with the most commonly used documents, ASCE 31-03 (ASCE, 2003) for 

evaluation and FEMA 356 (FEMA, 2000) for rehabilitation. Slightly different methods are used 

which can lead to slightly different levels of deficiency and the general level of expected 

performance has also been set lower in ASCE 31-03. 

The types of common standards and guidelines used to seismically rehabilitate buildings are 

described below. 

3.2.1 Mitigation of Evaluation Deficiencies 

Most commonly, the scope of rehabilitation is determined by directly addressing the deficiencies 

determined by evaluation. This is certainly the case when using building type-specific codes 

such as the IEBC (ICC, 2003) and local ordinances, because the evaluation and retrofit standards 

are one and the same. Similarly, the Simplified Rehabilitation method contained in FEMA 356 

is based on use of the evaluation standard, ASCE 31-03 as a basis for design of rehabilitation 

measures. In the rare case where the code for new buildings is used as a standard for existing 
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buildings, the rehabilitation would also be determined by directly addressing deficiencies from 

an evaluation. 

3.2.2 Rehabilitation Design Based on Nonlinear Response 

Few, if any, evaluation methods fully consider nonlinear response (unless FEMA 356 itself is 

used to evaluate), so if rehabilitation designs are determined in this manner, the extent of retrofit, 

and in some cases, the entire strategy of retrofit, may differ from merely eliminating the 

evaluation deficiencies. Nonlinear techniques are intended to more reliably predict performance, 

so when this is desirable—rather than meeting an arbitrary standard—these methods of analysis 

and design of rehabilitation measures are indicated. 

3.3 Classes of Rehabilitation Measures 

In most cases, the primary focus for determining a viable retrofit scheme is on vertically oriented 

components (e.g. column, walls, braces, etc.) because of their significance in providing either 

lateral stability or gravity load resistance. Deficiencies in vertical elements are caused by 

excessive inter-story deformations that either create unacceptable force or deformation demands. 

However, depending on the building type, the walls and columns may be adequate for seismic 

and gravity loads, while the building is inadequately tied together, forming a threat for partial or 

complete collapse in an earthquake. In order to design an efficient retrofit scheme, it is 

imperative to have a thorough understanding of the expected seismic response of the existing 

building and all of its deficiencies. 

In the traditional sense of improving the performance of the existing structure, there are three 

basic classes of measures taken to retrofit a building: 

Add elements, usually to increase strength or stiffness 

Enhance performance of existing elements, increasing strength or deformation capacity 

Improve connections between components, assuring that individual elements do not 

become detached and fall, a complete load path exists, and that the force distributions 

assumed by the designer can occur 

The types of retrofit measures often balance one another in that employing more of one will 

mean less of another is needed. It is obvious that providing added global stiffness will require 

less deformation capacity for local elements (e.g. individual columns), but it is often less obvious 

that careful placement of new lateral elements may minimize a connectivity issue such as a 

diaphragm deficiency. Important connectivity issues such as wall-to-floor ties, however, are 

often independent and must be adequately supplied. 

In addition to improving the strength or ductility of the existing structural elements, there are less 

traditional methods of improving the performance of the overall structure. These methods can be 

categorized as follows: 

Seismic demand can be reduced by removing upper floors or other mass from the 

structure, adding damping devices to reduce displacement, or seismically isolating all or 

part of the structure. 
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Selected elements can be removed or weakened to prevent damaging interaction between 

different systems, to eliminate damage to the element or to minimize a vertical or 

horizontal irregularity. 

This document uses these five classes of retrofit measures, in conjunction with the categories of 

seismic deficiencies described in Section 2.3 as a framework to present specific retrofit 

techniques. The matrices in each chapter of Part 2 list rehabilitation techniques according to 

these classes of retrofit measures and the deficiency that they mitigate. Retrofit methods that are 

relatively independent of the model building being considered are described in Part 3. 

The classes of retrofit measures are discussed in more detail below. 

3.3.1 Add Elements 

This is the most obvious and most general class of retrofit measures. In many cases, new shear 

walls, braced frames, or moment frames are added to an existing building to mitigate deficiencies 

in global strength, global stiffness, configuration, or to reduce the span of diaphragms as 

described in Sections 2.3.1, 2.3.2, 2.3.3, or 2.3.6 respectively. New elements can also be added 

as collectors to mitigate deficiencies in load path as described in Section 2.3.4. 

Retrofit schemes are developed with a balance of additional elements and enhanced existing 

elements (see Section 3.3.2) that best fit the socio-economic demands described in Section 3.4.2. 

Either adding new elements or enhancing the strength of existing elements could create a load 

path issue. The designer must assure that the new loads attracted to these elements can be 

delivered by other existing components. Therefore, eliminating a deficiency in Global Strength

or Global Stiffness may create a deficiency in Load Path that did not exist initially. 

3.3.2 Enhance Performance of Existing Elements 

Rather than providing retrofit measures that affect the entire structure, deficiencies can also be 

eliminated at the local, component level. This can be done by enhancing the existing shear or 

moment strength of an element, or simply by altering the element in a way that allows additional 

deformation without compromising vertical load-carrying capacity. 

Given that certain components of the structure will yield when subjected to strong ground 

motion, it is important to recognize that some yielding sequences are almost always preferred: 

beams yielding before columns, bracing members yielding before connections, flexural yielding 

before shear failure in columns and walls. These relationships can be determined by analysis and 

controlled by local retrofit in a variety of ways. For example: 

Columns in frames and connections in braces can be strengthened, and the shear capacity 

of columns and walls can be enhanced to be stronger than the shear that can be delivered 

by the flexural strength. 

Concrete columns can be wrapped with steel, concrete, or other materials to provide 

confinement and shear strength. Composites of glass or carbon fibers and epoxy are 

becoming popular to enhance shear strength and confinement in columns. 

Concrete and masonry walls can be layered with reinforced concrete, plate steel, and 

other materials such as fiber composites. 
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An indirect method of mitigating an unreasonably small drift capacity of a gravity element or 

system is to provide a supplemental gravity support system. In some situations, the cost of 

adding sufficient new global strength and stiffness or of increasing deformation capacity of 

certain gravity elements is excessive. For seismic performance primarily aimed at life safety, 

adding supplemental gravity supports might provide efficient mitigation. A common example of 

this practice is the supplemental support required for concentrated wall-supported loads in 

unreinforced masonry bearing wall buildings contained in most standards for retrofit. 

Supplemental support techniques have also been used in several cases for parts or all of concrete 

gravity systems. 

Although enhancement of performance of existing elements can provide strength and stiffness 

for deficiencies similar to adding elements, these measures are most commonly used to mitigate 

inadequate component detailing as described in Section 2.3.5. 

3.3.3 Improve Connections Between Components 

The class of rehabilitation technique is almost exclusively targeted at mitigation of load path 

deficiencies as described in Section 2.3.4. With the exception of collectors, a deficiency in the 

load path is most often created by a weak connection, rather than by a completely missing link. 

However, some poor connections, particularly between beam and supporting column, are not 

directly in the primary seismic load path but still require strengthening to assure reliable gravity 

load support during strong shaking. 

3.3.4 Reduce Demand 

For buildings that contain a complete but relatively weak lateral system and that also have excess 

space or a site where supplementary space can be constructed, removal of several top floors may 

prove to be an economical and practical method of providing acceptable performance. However, 

like schemes that require strengthening, the noise and disruption or removing floors must be 

considered, particularly if the remaining floors are to remain occupied. In many cases, little 

or no retrofit work may be required on the lower floors, although due to a shortened period, the 

acceleration response of the base may be increased. This issue is discussed further in 

Chapter 24. 

Techniques to reduce demand on the seismic system by modification of dynamic response of a 

structure are also included in this class. Perhaps the most notable example is seismic isolation, 

although this procedure is relatively expensive compared to alternate techniques and is normally 

employed in existing buildings for historic preservation or for occupancies that cannot be 

disturbed. A technique to modify response that is often economically competitive with 

traditional rehabilitation is the addition of damping in a structure. The added damping may 

reduce deformations sufficiently to prevent unacceptable damage in the existing system. 

Systems that actively control dynamic response have also been the subject of research, but have 

not made their way into common use. Further descriptions of response modification techniques 

are given in Part 3. 

3.3.5 Remove Selected Components 

Lastly, deformation capacity can be enhanced locally by uncoupling brittle elements from the 

deforming structure, or by removing them completely. Examples of this procedure include 
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placement of vertical sawcuts in unreinforced masonry walls to change their behavior from shear 

failure to a more acceptable rocking mode and to create slots between spandrel beams and 

columns to prevent the column from being a “short column” prone to shear failure. 

3.4 Strategies to Develop Rehabilitation Schemes 

3.4.1 Technical Considerations 

The first overview by a retrofit designer should be studying the deficiencies identified by the 

evaluation. Typical deficiencies are categorized by model building type in Part 2 and a table for 

each is give that relates the deficiencies to common mitigation techniques. 

Some common seismic deficiencies are very localized and can be efficiently mitigated by 

narrowly targeting the retrofit activity. For example, for some one-story and two-story masonry 

or concrete wall buildings, the only deficiency may be out-of-plane wall ties to the diaphragm. 

Similarly, adequate resistance to overturning for a discontinuous shear wall may be made 

available by no more than providing confinement to the supporting column. Load path issues 

should be completely identified because there are often few choices for mitigation. 

Next, the appropriate deficiency table in Part 2 should be studied to identify if a potential 

mitigation technique is effective for more than one deficiency present in the building. Adding 

strength or stiffness is very common, and a few new elements may solve strength, drift, and 

configuration problems. 

When adding new lateral force-resisting elements such as shear walls, moment frames, or braced 

frames, several issues should be considered: Is the deformation compatible with the existing 

lateral force-resistsing or gravity load-carrying system? Will the new system sufficiently relieve 

the existing structure of load or deformation at all levels? Is the new system adding significant 

mass to the structure? Will this mass invalidate the previous evaluation? Will extensive new 

foundations be needed for the new system? 

For any early trial scheme, review that the altered structure will: 

Have a complete load path 

Have sufficient strength and stiffness to meet the design standard 

Be compatible with and will adequately protect the existing lateral and gravity system 

Have an adequate foundation to assume a fixed base building, or have appropriately 

considered foundation flexibility in the design 

3.4.2 Nontechnical Considerations 

The solution chosen for retrofit is almost always dictated by building-user oriented issues rather 

than by merely satisfying technical demands. There are five basic issues that are of concern to 

building owners or users: 

Construction cost 

Seismic performance 

Short-term disruption of occupants 
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Long-term functionality of building 

Aesthetics, including consideration of historic preservation 

All of these characteristics are always considered, but an importance will eventually be put on 

each of them, either consciously or subconsciously, and a combination of weighting factors will 

determine the scheme chosen. 

3.4.3 Cost 

Construction cost is always important and is balanced against one or more other considerations 

deemed significant. However, sometimes other economic considerations, such as the cost of 

disruption to building users or the value of contents to be seismically protected, can be orders-ofmagnitude larger than construction costs, thus lessening its importance. 

3.4.4 Seismic Performance 

If the governing jurisdiction is requiring seismic strengthening, either due to extensive 

remodeling or structural alteration, a design standard and resulting seismic performance 

expectation will normally be specified. When seismic rehabilitation is voluntary, the benefit-cost 

relationship of various performance levels may be considered explicitly, but in any case, the 

seismic performance factor will become important in the development of the scheme. 

Typically, in either situation, perceived qualitative differences between the probable performance 

of different schemes were often used to assist in choosing a scheme. Now that performancebased design is integral to most rehabilitation, specific performance objectives are often set prior 

to beginning development of schemes. Objectives that require a limited amount of damage or 

"continued occupancy" will severely limit the retrofit methods that can be used and may control 

the other four issues. 

3.4.5 Short-term Disruption of Occupants 

Often retrofits are done at the time of major building remodels and this issue is minimized. 

However, in cases where the building is partially or completely occupied, this parameter 

commonly becomes dominant and controls the design. 

To minimize disruption, schemes are often explored that place strengthening elements outside 

the building the building envelope. Concrete shear walls, pier-spandrel frames, and steel braced 

frames placed adjacent to or within the plane of exterior walls have been used in this way. Shear 

connection of the diaphragms to these new elements must be carefully considered. External 

elements that can also provide new strength and stiffness perpendicular to the exterior wall have 

also been used. In this case, a collector normally must be run into the building to connect the 

new element to the floor diaphragms. Installation of this collector may disrupt the internal 

systems, finishes, and occupants of the building to the extent that nullifies the exterior location of 

the new lateral element. Although there are many examples of exterior solutions that have been 

installed with continuous occupancy of the building, acceptability of the noise, dust, and 

vibration associated with the construction, as well as the potential disruption of access and 

egress, must be carefully considered during planning and design. 
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3.4.6 Long-term Functionality of Building 

The addition of shear walls or braced frame in the interior of a building will always change the 

functional use plan. If the seismic work is being done as part of a general renovation, new 

functional spaces can often be planned around the new elements. However, such permanent 

structural elements will always reduce the flexibility of future replanning of the space. This 

characteristic is often judged less important than the other four and is therefore sacrificed to 

satisfy other goals. Often the planning flexibility is only subtlety changed. However, it can be 

significant in building occupancies that need open spaces such as retail spaces and parking 

garages. 

3.4.7 Aesthetics 

In historic buildings, considerations for preservation of historic fabric usually control the design. 

In many cases, even performance objectives are controlled by limitations imposed by 

preservation. In non-historic buildings, aesthetics is commonly stated as a criterion, but, in the 

end, is often sacrificed, particularly in favor of minimizing cost and disruption to tenants. 

3.5 Other Common Issues Associated with Seismic Rehabilitation 

3.5.1 Constructability 

The options to obtain adequate access to the location of construction within the building as well 

as a sufficient local construction space are far more limited in a seismic rehabilitation project 

than in new construction. In addition, there may be issues related to undercutting existing 

footings, providing temporary shoring of gravity elements, or providing temporary lateral 

support for certain elements of the structure, certain floors, or even for the whole building. The 

design engineer must consider these issues when conceiving a rehabilitation scheme; the reality 

of field conditions may render a scheme physically or economically infeasible. 

To control their liability for site construction safety, engineers have generally avoided 

specification of “means and methods” of construction as part of the construction documents. 

This concern is no less true for rehabilitation projects, but, in cases where significant structural 

alteration is required, it is often difficult to develop a realistic scheme without a thorough 

understanding of probable construction methods. 

3.5.2 Materials Testing 

Destructive testing of existing material can be disruptive and expensive. Care should be taken in 

designing a program that suits the building-specific conditions. If basic information is available 

on structural materials, it is often prudent to delay testing until preliminary evaluation is 

completed to identify critical existing components, or on the other hand, to determine that the 

material strengths are relatively unimportant and material testing can be minimized. 

3.5.3 Disruption to Building Systems and Replacement of Finishes 

The significance of conflicts with mechanical, electrical, or plumbing distribution systems or 

equipment should be considered during development of rehabilitation schemes. Temporary 

disruptions of services may be acceptable if the building is not to be occupied during 

construction, but may need to be limited if the building is occupied. High costs may be 
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associated with permanent changes in routing or relocation of equipment due to the seismic 

work. 

Similarly, the cost and disruption of removal and replacement of finishes or cladding to gain 

access to the structure must be considered. In addition to certain finishes being unique and 

expensive or historic, the construction associated with gaining this access normally requires 

evacuation and closing off of the local area. 

3.5.4 Concealed Conditions 

Even when original construction drawings are available and certain material tests have been 

performed to gain confidence in the knowledge of existing conditions, different conditions may 

be exposed during construction. In addition to attempting to minimize the importance of such 

possibilities by field exposures and design, the design professional of record should also be 

engaged during construction, in order to properly assess such discoveries and to enable design of 

mitigating measures consistent with the overall scheme. 

3.5.5 Quality Assurance 

Quality assurance programs are probably more important in rehabilitation projects than with new 

construction. Given no control of existing conditions, the margin for error is often small. In 

addition, as indicated in Section 3.5.4, conditions in the field are often different than assumed 

and effective revisions often need to be developed. 

3.5.6 Detailing for New Elements 

In almost all codes, new elements installed into existing buildings as part of a seismic 

rehabilitation must meet the detailing requirements for new construction. For example, 

minimum reinforcing of concrete walls or columns, slenderness ratios of braces and connection 

details must be in accordance with new code requirements. With designs that utilized nonlinear 

analysis, deformation capacities will have been set using an assumed detailing pattern from the 

code from new buildings, and that level of detailing must then be provided. 

3.5.7 Vulnerability During Construction 

Installation of new seismic elements within an existing building often requires demolition of 

parts of the gravity load system as well as the effective lateral load system. Although safety 

during construction is the contractor’s responsibility under the “means and methods” principle, 

the engineer designing the seismic rehabilitation may be in a position to identify global 

weaknesses in the gravity or lateral load system that could develop during construction. Such 

conditions should be pointed out in the contract documents, although temporary strengthening 

measures that might be needed during construction should be designed by the contractor’s 

engineer. 

3.5.8 Determination of Component Capacity by Testing 

There are many unique components in existing buildings for which no data are available to 

define strength and/or deformation capacity. If certain components or connections occur in 

multiple locations and will potentially require extensive and costly retrofit, in-situ or laboratory 

testing may be justified. The cost of such testing and the possibility of acceptable performance 
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must be judged against potential savings in the cost of rehabilitation. Experts in material 

behavior and testing should be consulted to assist with such evaluations. 

3.5.9 Incremental Rehabilitation 

Disruption to occupants can be minimized if seismic rehabilitation is combined with other 

maintenance or renovation work. This may lead to phased or incremented construction. The 

potential to implement this type of seismic improvements is documented in a series of FEMA 

documents, FEMA 395 to FEMA 400 and FEMA 420. 

3.6 Issues with New Techniques or Products 

Part 2 discusses many commonly employed seismic rehabilitation techniques. However, it not 

possible to include all currently available techniques in the document; and there will always be 

new techniques, products, research, and approaches developed in the future. There is no 

substitute for engineering judgment. When considering a rehabilitation technique or product, the 

design engineer should consider the following issues. 

Prior Use 

Has the approach been used successfully before? 

How long have previous installations been in place? 

Have the installed rehabilitation measures been through actual seismic events? 

Testing 

General quality of testing. 

General quality of documentation. 

Was the testing performed by the manufacturer or by an independent entity? 

Relevance of test to actual elements. 

Type of testing: monotonic, cyclic quasistatic, dynamic. 

Number of specimens. 

How far into the nonlinear range did the testing go? 

Why was the test stopped? 

Were test results placed in performance-based design limit states? 

Construction 

Is the technique limited to only certain specialized subcontractors? 

Can the technique be documented sufficiently to be bid? 

Does installation involve noise, dust, vibration, harmful vapors, and/or danger? 

Are special tools or set-ups needed? 

Long-Term Stability of Mitigation Materials 

Do the materials creep, crack, shrink, lose strength, debond, rust/corrode, etc. over time? 

Can they be placed in exterior environments? 

Are there fire safety requirements or concerns? 

Are there temperature range limitations? 

Are coefficients of thermal expansion compatible with adjacent materials? 
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Do they react with other materials, such as galvanic corrosion from dissimilar metals, 

efflorescence in masonry, or breakdowns from ultraviolet light? 

Are moisture issues appropriately addressed or can they be mitigated sufficiently when 

the technique is used? 

Aesthetic and Historic Preservation 

Is the technique suitable for sensitive structures? 

Is it reversible? 

Code Considerations 

Are building code procedures and design methodologies available or applicable? 

Does the product have approvals? 

Quality Assurance 

Can an adequate field quality assurance program be developed to verify that in-situ 

properties meet design assumptions? 

Can a typical testing lab perform the inspection or testing or is special expertise needed? 

Cost 

Is adequate information available on pricing to make decisions during design? 

Is the work or product best procured lump-sum or by unit price? 

Is the cost worth the benefit? 

3.7 References 

ASCE, 2003, Standard for the Seismic Evaluation of Buildings, ASCE 31-03, Structural 

Engineering Institute of the American Society of Structural Engineers, Reston, VA. 

FEMA, 2000, Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 

356, Federal Emergency Management Agency, Washington, D.C. 

ICC, 2003, International Existing Building Code, 2003 Edition, International Conference of 

Building Officials, Country Club Hills, IL. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 4 – FEMA Model Building Types

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Chapter 4 - FEMA Model Building Types 

4.1 Introduction 

This document is primarily organized around the FEMA model building types. It is expected 

that most users of this document will be interested in finding information on a particular building 

or building type, which suggested this organization. Each building type is therefore assigned a 

chapter. Common seismic deficiencies for each building type are identified and mitigation 

techniques suggested, although it is recognized that most buildings will have multiple 

deficiencies and may require a combination of mitigating actions. The rehabilitation techniques 

commonly used for each building type are identified in each chapter and, if closely associated 

with the building type, described in detail in that chapter. References are given to other chapters 

for other applicable techniques. 

4.2 History of Development 

Several sets of standard structural types have been created to describe the building inventory of 

the U.S. Initially, these model building types were developed for the purposes of assigning 

fragility relationships to inventories of buildings for loss estimation in ATC 13, (ATC, 1985). 

Studies of buildings for development of ATC 14, Evaluating the Seismic Resistance of Existing 

Buildings (ATC, 1987), indicated a large number of types in existence, but identified 15 primary 

types around which evaluation considerations could be grouped. 

ATC 14 was later adapted for use in the FEMA series as FEMA 178, NEHRP Handbook for 

Seismic Evaluation of Buildings (FEMA, 1992a). This set of building types has subsequently 

been used extensively in other FEMA documents related to existing buildings, including FEMA 

154 (FEMA, 1988), FEMA 227 (FEMA, 1992b), and FEMA 156 (FEMA, 1995). 

When FEMA 178 was converted to a prestandard for input to the ASCE standards adoption 

process (FEMA 310 [FEMA, 1998] and ASCE 31-03 [ASCE, 2003]), the distinction between 

similar building types with flexible and rigid diaphragms was included by adding the suffix “A” 

to the alpha-numeric designation. For example, the definition of Building Type S1, Steel 

Moment Frames, was refined to designate steel moment frames with rigid diaphragms, and 

Building Type S1A was designated as steel moment frames with flexible diaphragms. 

However, this new designation was not assigned consistently. For example, W1A was defined 

to represent a W1 of larger size, rather than one with a flexible diaphragm; the designations 

RM1 and RM2 were used to differentiate flexible and rigid diaphragms in reinforced masonry 

buildings; finally, for the URM building type, the suffix A indicates a rigid diaphragm rather 

than a flexible diaphragm. 

4.3 Model Building Type Refinements in this Document 

Rather than causing additional inconsistency between documents, this document uses the preestablished model buildings types and designations described above. However, for the purposes 

of relating retrofit techniques to building types, additional minor refinements to the building type 

designations are convenient and clarifying. Specifically, concrete shear wall buildings (Building 
Techniques for the Seismic Rehabilitation of

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4-2 

Type C2) have been split into two groups, those with essentially complete gravity frames 

(Building Type C2f) and those primarily using bearing walls (Building Type C2b). Similarly, 

reinforced masonry buildings (Building Type RM1) have been split into two groups, those that 

are very similar to concrete tilts ups (Building Type RM1t) and those that are very similar to 

older, unreinforced masonry buildings (Building Type RM1u). Using these refinements, 

building performance characteristics, common seismic deficiencies, and applicable mitigation 

techniques can be more clearly described. 

Finally, building types that are less common or that seldom require retrofit have not been 

included or have been de-emphasized in this document, although techniques suggested for a 

similar building will generally be applicable. The excluded building types include Building 

Type S3, Steel Light Frames, and the following sub-types designated by the “A” suffix: C2A, 

PC1A, PC2A, and URMA. 

4.4 Description 

The model building types are summarized below. Detailed descriptions can be found in each 

dedicated chapter. Many real buildings have characteristics from more than one model building 

type. Useful information can still be obtained by referring to chapters for similar building types. 

Table 4-1: Model Building Types 

W1: Wood Light Frames 

Building Type W1 consists of one- and twofamily detached dwellings of one or more 

stories. Floor and roof framing are most 

commonly woodframe joists and rafters 

supported on wood stud walls. The first floor 

may be slab-on-grade or framed. Lateral forces 

in W1 buildings are resisted by woodframe 

diaphragms and shear walls. 

W1A: Multistory, Multi-Unit 

Residential Woodframes 

Building Type W1A is similar to Building Type

W1 in use of light-frame wall, floor and roof 

construction, but includes large multi-family, 

multistory buildings. In W1A buildings, 

second and higher stories are almost exclusively 

residential use, while the first story can include 

any combination of parking, common areas, 

storage, and residential units. Post and beam 

framing often replaces bearing walls in nonresidential areas. Multi-family residential 

buildings with commercial space at the first 

story are included in Building Type W1A due 

to similar building characteristics. Lateral 

forces in W1A buildings are primarily resisted 

by woodframe diaphragms and shear walls. 
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4-3 

Table 4-1: Model Building Types (continued) 

W2: Woodframes, Commercial and Industrial 

Building Type W2 consists of commercial, 

institutional, and smaller industrial buildings 

constructed primarily of wood framing. The 

first floor is most commonly slab-on-grade, but 

may be framed. Floor and roof framing may 

include wood joists, wood or steel trusses, and 

glulam or steel beams, with wood posts or steel 

columns. Post and beam framing is common at 

storefronts or garage openings. Lateral forces in 

W2 buildings are primarily resisted by 

woodframe diaphragms and shear walls, 

sometimes in combination with isolated 

concrete or masonry shear walls, steel braced 

frames, or steel moment frames. Diaphragm 

spans may be significantly larger than in W1

and W1A buildings. 

S1/S1A: Steel Moment Frames

Building Type S1 consists of an essentially 

complete frame assembly of steel beams and 

columns. Lateral forces are resisted by moment 

frames that develop stiffness through rigid 

connections of the beam and column created by 

angles, plates, and bolts, and/or by welding. 

Floors are cast-in-place concrete slabs or metal 

decks infilled with concrete. Building Type

S1A is similar but has floors and roofs that act 

as flexible diaphragms such as wood or 

untopped metal deck. 

S2/S2A: Steel Braced Frames 

Building Type S2 consists of a frame assembly 

of steel beams and columns. Lateral forces are 

resisted by diagonal steel members placed in 

selected bays. Floors are cast-in-place concrete 

slabs or metal decks infilled with concrete. 

Building Type S2A is similar but has floors and 

roofs that act as flexible diaphragms such as 

wood or untopped metal deck. 
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4-4 

Table 4-1: Model Building Types (continued)

S4: Steel Frames with Concrete Shear Walls 

Building Type S4 consists of an essentially 

complete frame assembly of steel beams and 

columns. The floors are concrete slabs or 

concrete fill over metal deck. These buildings 

feature a significant number of concrete walls 

effectively acting as shear walls, either as 

vertical transportation cores, isolated in 

selected bays, or as a perimeter wall system. 

The steel column and beam system may act 

only to carry gravity loads or may have rigid 

connections to act as a moment frame to form 

a dual system. 

S5/S5A: Steel Frames 

with Infill Masonry Shear Walls 

Building Type S5 is normally an older 

building that consists of an essentially 

complete gravity frame assembly of steel 

floor beams or trusses and steel columns. The 

floor consists of masonry flat arches, concrete 

slabs or metal deck and concrete fill. Exterior 

walls, and possibly some interior walls, are 

constructed of unreinforced masonry, tightly 

infilling the space between columns and 

between beams and the floor such that the 

infill interacts with the frame to form a lateral 

force-resisting element. 

C1: Concrete Moment Frames 

Type C1 buildings consist of concrete 

framing, either a complete system of beams 

and columns or columns supporting slabs 

without gravity beams. Lateral forces are 

resisted by cast-in-place moment frames that 

develop stiffness through rigid connections of 

the column and beams. 
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Table 4-1: Model Building Types (continued)

C2b: Concrete Shear Walls 

(Bearing Wall Systems) 

C2f: Concrete Shear Walls 

(Gravity Frame Systems) 

Building Type C2 covers buildings with 

concrete walls. For this document, the type is 

split into C2b and C2f. 

Building Type C2b is usually all concrete 

with flat slab or precast plank floors and 

concrete bearing walls. Little, if any, of the 

gravity loads are resisted by beams and 

columns. 

Building Type C2f has a column and beam or 

column and slab system that essentially carries 

all gravity load. Lateral loads are resisted by 

concrete shear walls surrounding shafts, at the 

building perimeter, or isolated walls placed 

specifically for lateral resistance. 

C3/C3A: Concrete Frames with 

Infill Masonry Shear Walls

Building Type C3 is normally an older 

building with an essentially complete gravity 

frame assembly of concrete columns and floor 

systems. The floors can consist of a variety of 

concrete systems including flat plates, twoway slabs, and beam and slab. Exterior walls, 

and possibly some interior walls, are 

constructed of unreinforced masonry, tightly 

infilling the space between columns 

horizontally and between floor structural 

elements vertically, such that the infill 

interacts with the frame to form a lateral 

force-resisting element. 
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Table 4-1: Model Building Types (continued)

PC1: Tilt-Up Concrete Shear Walls 

Building Type PC1 is constructed with 

concrete walls, cast on site and tilted up to 

form the exterior of the building. PC1

buildings are used for many occupancy types 

including warehouse, light industrial, 

wholesale and retail stores, and office. The 

majority of these buildings are one story; 

however, tilt-up buildings of up to three and 

four stories are common, and a limited 

number with more stories exist. For many 

years, tilt-up buildings have been primarily 

large box-type buildings with the tilt-up walls 

at the building perimeter; however, in recent 

years, tilt-up construction has been used in 

more complex and varied commercial 

building configurations. Lateral forces in PC1

buildings are resisted by flexible wood or 

steel roof diaphragms and tilt-up concrete 

shear walls. Floor diaphragms are most 

commonly composite steel decking. 

PC2: Precast Concrete Frames with Shear Walls 

Buildings designated as PC2 include wide 

ranging combinations of precast and cast-inplace concrete elements. Precast members 

may be limited to a floor system of hollow 

core or T-beam construction, or may include 

all elements of the gravity and lateral load 

systems. For this document, Building Type

PC2 includes concrete wall or frame 

buildings in which any of the horizontal or 

vertical elements of the lateral load system are 

of precast concrete, except for flexible 

diaphragm buildings which are addressed as 

Building Type PC1. 
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Table 4-1: Model Building Types (continued)

RM1t: Reinforced Masonry Bearing Walls 

(Similar to Tilt-Up Concrete Shear Walls) 

RM1u: Reinforced Masonry Bearing Walls 

(Similar to Unreinforced Masonry Bearing Walls) 

Building Type RM1 is constructed with 

reinforced masonry (brick cavity wall or 

concrete masonry unit) perimeter walls with a 

wood or metal deck flexible diaphragm. 

For this document, Building Type RM1 is 

separated into two categories, RM1u, which 

is multistory, and typically has interior CMU 

walls and shorter diaphragm spans, and 

RM1t, a large, typically one-story buildings 

similar to concrete tilt-ups. 

RM2: Reinforced Masonry Bearing Walls 

(Similar to Concrete Shear Walls w/ Bearing Walls) 

Building Type RM2 consists of reinforced 

masonry walls and concrete slab floors that 

may be either cast-in-place or precast. This 

building type is often used for hotel and 

motels and is similar to the concrete bearing 

wall type C2.
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Table 4-1: Model Building Types (continued)

URM: Unreinforced Masonry Bearing Walls 

Building Type URM consists of unreinforced 

masonry bearing walls, usually at the 

perimeter and usually brick masonry. The 

floors are typically of wood joists and wood 

sheathing supported on the walls and on 

interior post and beam construction. 

4.5 References 

ASCE, 2003, Seismic Evaluation of Buildings, ASCE 31-03, Structural Engineering Institute of 

the American Society of Civil Engineers, Reston, VA. 

ATC, 1985, Earthquake Damage Evaluation Data for California, ATC 13, Applied Technology 

Council, Redwood City, CA. 

ATC, 1987, Evaluating the Seismic Resistance of Existing Buildings, ATC 14, Applied 

Technology Council, Redwood City, CA 

FEMA, 1988, Rapid Visual Screening of Buildings for Potential Seismic Hazards, ATC 154, 

Federal Emergency Management Agency, Washington, D.C. 

FEMA, 1992a, NEHRP Handbook for the Seismic Rehabilitation of Existing Buildings, FEMA 

172, Federal Emergency Management Agency, Washington, D.C. 

FEMA, 1992b, A Benefit-Cost Model for the Seismic Rehabilitation of Buildings, FEMA 227, 

Federal Emergency Management Agency, Washington, D.C. 

FEMA,1995, Typical Costs for Seismic Rehabilitation of Existing Buildings, FEMA 156, Second 

Edition, Federal Emergency Management Agency, Washington, D.C. 

FEMA, 1998, Handbook for the Seismic Evaluation of Buildings—A Prestandard, FEMA 310, 

Federal Emergency Management Agency, Washington, D.C. 
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Existing Buildings: FEMA 547 Chapter 5 – Type W1: Wood Light Frames

5-1 

Chapter 5 - Building Type W1: Wood Light Frames 

5.1 Description of the Model Building Type 

Building Type W1 consists of one- and two-family detached dwellings of one or more stories. 

Floor and roof framing are most commonly wood joists and rafters supported on wood stud walls 

(called woodframe or wood light-frame). The first floor may be slab-on-grade or a raised framed 

floor. Lateral forces in W1 buildings are resisted by woodframe diaphragms and shear walls. 

Chimneys, where present, consist of solid brick masonry, masonry veneer, or woodframe with 

internal metal flues. Although materials for detached one- and two-family dwellings vary beyond 

woodframe, this chapter will focus on this most common type of construction. Figure 5.1-1 

provides one illustration of this building type. 

Figure 5.1-1: Building Type W1: Wood Light Frames 

One- and Two-Family Detached Dwelling 

Design Practice 

W1 buildings recently constructed near population centers may have a partial or complete 

engineered design; however, most W1 buildings will have been designed using prescriptive 

provisions (conventional construction). Where prescriptive design has been used, it can generally 

be expected that no numerical check of sheathing, fastening, wall overturning, or other load-path 

connections has been performed, and that no fastening or connections beyond basic fastening 

schedules have been used. In engineered design, the extent of analysis and detailing can vary 
Techniques for the Seismic Rehabilitation of

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5-2 

from a check of in-plane shear capacity of shear walls, to exhaustive design and detailing. 

Minimum fastening and connection needs to be assumed unless more is known to exist. 

Walls 

Wall bracing materials and detailing vary depending on dwelling age and location. Except for 

recently constructed or rehabilitated W1 buildings, it is most common for the finish material to 

also serve as the shear wall bracing material. Common interior finish and bracing materials 

include plaster over wood lath, plaster over gypsum lath (button board) and gypsum wallboard. 

Common exterior finish and bracing materials include board siding, shingles, panel siding, and 

stucco. Finish materials such as vinyl siding and EIFS are not included with these bracing 

materials due to low stiffness and negligible bracing capacity. Wall sheathing is sometimes 

present in addition to finish materials. In older W1 buildings, lumber sheathing--applied 

horizontally, vertically or diagonally--was often used. In newer buildings, wood structural panel 

(plywood or oriented strand board) sheathing is most often used. Because interior and exterior 

finish materials often also serve as bracing materials in W1 buildings, it is difficult to 

differentiate between structural and nonstructural materials. 

Early W1 building construction used post and beam wall framing systems in lieu of closely 

spaced studs. Most construction shifted to stud systems between the mid 1800s and early 1900s; 

however, some post and beam construction is still built. Except where braced frames or kneebraces provide alternate lateral force-resisting systems, post and beam wall systems still rely on 

wall finishes or sheathing to resist in-plane lateral loads. Stud systems were first constructed 

using balloon-framed walls, in which individual stud members extended from the foundation to 

the very top of the framed wall. This height often included cripple walls plus two stories. When 

walls are balloon framed, floor framing is hung off of the interior face of the studs. In the early 

1900s, most framing changed from balloon framed to platform framed, in which the wall framing 

stops at the underside of each floor, and the floor framing sits on top of wall framing rather than 

hanging off the face. These two wall framing systems have important differences for detailing 

load transfer, chords, and collectors for shear walls and diaphragms. 

Floor and Roof Diaphragms 

Floor and roof diaphragm materials and detailing vary depending on building age and location. 

In older buildings, solid lumber sheathing is most often applied straight or diagonally under 

built-up and membrane roofs, and spaced lumber sheathing is found under shingle and tile roofs. 

In older buildings, floor sheathing is often solid lumber sheathing applied horizontally or 

diagonally. In some cases, hardwood floors form both the sheathing and the floor finish. In 

newer or rehabilitated buildings, wood structural panel (plywood or oriented strand board) floor 

and roof sheathing is most common. The strength of wood structural panel diaphragms varies 

depending on whether they are blocked (interior sheathing panel edges supported on and edge 

nailed to blocking) or unblocked (interior edges not supported or nailed), sheathing panel layout, 

and sheathing nail size and spacing. The presence or absence of diaphragm chords and collectors 

also affects the diaphragm strength and stiffness. As is true with shear walls, the level of design 

detailing for diaphragms can vary significantly. 

Plank and beam framing became popular in the mid-1900s and is still in use today. This system 

uses 2x or thicker straight lumber plank sheathing for floors and roofs, supported on beams 
Techniques for the Seismic Rehabilitation of

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5-3 

spaced between four and eight feet on center. The planking is often left exposed on the 

underside for the ceiling below. Publications such as Plank and Beam Framing for Residential 

Buildings (AF&PA, 2003) describe this construction type. In the western U.S., wood structural 

panel overlays are often applied over the lumber sheathing to provide diaphragms for engineered 

designs. Shear walls are used as vertical elements to resist lateral loads. 

Distribution of seismic forces to the vertical elements of the lateral force-resisting system is 

influenced in part by the diaphragm stiffness. The selection of a flexible or rigid diaphragm 

model for purposes of force distribution is controversial at the time of this update, and details of 

building analysis are beyond the scope of this document. It is recommended that the reader refer 

to applicable building codes, local jurisdiction requirements, and the local standard of practice. 

System Between Lowest Framed Floor and Grade 

Where the lowest occupied floor in a W1 building is woodframed, there are a large number of 

structural systems that can occur between the framed floor and grade. Figure 5.1-2 illustrates 

some of the common systems for level building sites. The type of system can vary based on 

region, building age, soils, type of site, exposure to environmental hazard such as flood, etc. 

These may be foundation systems, or may include superstructure sitting on top of the foundation. 

Common weaknesses in these systems include 1) lack of a load path for lateral loads (Figure 5.1-

2A), 2) limited lateral load resistance, and 3) lack of adequate connection to transfer lateral loads 

to the foundation. 

Cripple walls (Figure 5.1-2C) are one common system between the framed floor and grade. 

Cripple walls are wood stud framed walls that extend from the top of a foundation to the 

underside of the first framed floor. Cripple walls often enclose an uninhabited crawl space, but 

may also sit on top of partial height concrete or masonry walls in a basement. In past 

earthquakes, dwelling drift and damage has often been concentrated in cripple walls. 

W1 buildings are often supported on continuous perimeter foundations or foundation walls 

(Figure 5.1-2D) in combination with continuous or isolated interior pier footings. Alternate 

foundation types may be used locally or regionally. Materials for continuous perimeter 

foundations or foundation walls vary depending on age and location. Many older dwelling 

foundations use unreinforced concrete, brick masonry, or stone masonry. Today, use of lightly 

reinforced concrete continuous foundations is most common in the western states, and 

unreinforced concrete masonry and brick masonry (pier and curtain wall) are common in other 

regions. In the 1970s and 1980s, use of post-tensioned slab-on-grade foundations became 

common in some areas with highly expansive soils; these present additional issues for anchorage 

that are discussed in Section 5.4.4. 

5.2 Seismic Response Characteristics 

The dynamic response of W1 buildings is very short period due to the stiffness of wall bracing 

and finish materials. Deflection and inelastic behavior occur primarily in the walls, while the 

floor and roof diaphragms remain close to elastic. Likewise, damage is mostly seen in the walls 

rather than the floor or roof systems. 
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Figure 5.1-2: Systems Between First Framed Floor and Grade – Level Lot Sites 
Techniques for the Seismic Rehabilitation of

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5.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

Life-safety performance of W1 buildings has generally been very good. A limited number of 

vulnerable configurations, however, have repeatedly resulted in significant damage, and in a few 

instances loss of life. In W1 buildings, damage to wall finish materials has contributed notably to 

repair costs. Wood chapters in two recently published earthquake engineering handbooks provide 

overviews of earthquake performance for woodframe buildings and extensive lists of references 

describing extent and details of damage (Dolan, 2003; Cobeen, 2004). Of the many discussions 

of performance, of particular note due to extent and detail is the Northridge earthquake 

reconnaissance report (EERI, 1996). 

It is not the objective of this document to address rehabilitation of buildings for wind loads; 

however, many of the rehabilitation measures that increase the strength and stiffness of the 

primary lateral-force-resisting system for seismic loads will also provide increased resistance to 

wind loads. Included is the addition of strength and stiffness in diaphragms, shear walls, and 

their connections. For load path connections, locations of greatest vulnerability and therefore 

priority items for seismic rehabilitation tend to be located at the base of the structure where 

seismic demand is greatest, such as anchorage to the foundation. In contrast, for wind 

rehabilitation, load path connections of greatest vulnerability and highest priority tend to be at 

the top of the structure, including roofing attachment to roof sheathing, roof sheathing 

attachment to rafters, rafter attachment to walls, etc. 

See below for general discussion and Table 5.3-1 for a detailed compilation of common seismic 

deficiencies and rehabilitation techniques for Building Type W1. 

Global Strength 

Inadequate strength, particularly in lower stories of multistory W1 buildings, has caused 

extensive damage to bracing and finish materials but has not generally resulted in hazard to life. 

Inadequate strength is most directly addressed by enhancing existing shear walls or adding new 

vertical elements. In one- and two-family dwellings this most often involves addition of wood 

structural panel sheathing and associated load path connections to an existing framed wall. While 

not commonly used in W1 buildings, steel moment and braced frames may be added to address 

global strength. 

Global Stiffness 

Global stiffness can occasionally be an issue in W1 buildings, particularly where archaic 

materials such as horizontal or vertical straight lumber sheathing are used for bracing and finish 

materials. In dwellings this is most likely to occur in unfinished garage, crawlspace or basement 

areas. This is a common condition for garage side walls in dense urban areas such as San 

Francisco. Where these types of sheathing are used, strength is usually an issue as well as 

stiffness. As with global strength, typical rehabilitation measures include enhancing existing 

shear walls or adding new vertical elements. Applicable rehabilitation measures are discussed in 

the Global Strength section. 
Techniques for the Seismic Rehabilitation of

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Table 5.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for W1 Build

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce Dem

Global Strength 

Insufficient inplane wall strength 

Wood structural panel 

shear wall [5.4.1] 

Steel moment frame 

[6.4.1] 

Steel braced frame 

[7.4.1] 

Enhance woodframe 

shear wall [5.4.1] 

Shear wall uplift 

anchorage and compression posts [6.4.4] 

Replace heav

roof finish wlight finish 

Global Stiffness 

Insufficient inplane wall stiffness 

Wood structural panel 

shear wall [5.4.1] 

Steel moment frame 

[6.4.1] 

Steel braced frame 

[7.4.1] 

Enhance woodframe 

shear wall [5.4.1] 

Shear wall uplift 

anchorage and compression posts [6.4.4] 

Missing or inadequate cripple wall bracing 

Add woodframe 

cripple wall 

Add continuous 

foundation and foundation wall 

Enhance woodframe 

cripple wall [5.4.4] 

Open front Wood structural panel 

shear wall [5.4.1] 

Collector [5.4.2] Moment frame [6.4.1] 

Enhance woodframe 

walls perpendicular to open front [5.4.1] 

Detailing of narrow 

woodframe shear wall piers 

Configuration 

Hillside Wood structural panel 

shear wall [5.4.5] 

Enhance woodframe 

shear wall [5.4.5] 

Anchor base level 

diaphragm to uphill foundation [5.4.5] 

Load Path Inadequate shear 

anchorage to foundation 

Anchorage to 

foundation [5.4.3] 
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Table 5.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for W1 Build

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce Dem

Inadequate shear wall overturning load path 

Supplement framing supporting woodframe shear wall [6.4.3] 

Shear wall uplift 

anchorage and compression posts [6.4.4] 

Inadequate shear transfer in wood framing 

Enhance load path for 

shear [5.4.1] 

Load Path (continued) 

Inadequate collectors to shear walls 

Enhanced existing collector 

Add collector [6.4.5], 

[7.4.2] 

Component Detailing 

Unreinforced & unbraced chimney 

Infill chimney [5.4.6] Brace chimney 

[5.4.6] 

Reduce unsupported chimney heig[5.4.6] 

Inadequate inplane strength and/or stiffness 

Enhance diaphragm [22.2.1] Diaphragm overlay [22.2.1] 

Replace heav

roof finish wlighter finish 

Inadequate chord capacity 

Enhance chord members and connections [22.2.2] 

Excessive stresses at openings and irregularities 

Enhance diaphragm detailing 

Diaphragms 

Re-entrant corner Enhance diaphragm 

detailing 

Foundations See Chapter 23 [ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 5 – Type W1: Wood Light Frames

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Configuration 

Several W1 building configurations have been observed to be vulnerable to damage, in some 

cases resulting in full or partial collapse. Vulnerable configurations include buildings with 

inadequate bracing systems between the lowest framed floor and grade, open front building 

portions, and split-level buildings. Primary rehabilitation measures are specific to each of these 

configurations. 

For level site buildings, inadequate bracing systems between the lowest framed floor and grade 

commonly include inadequately braced cripple walls and perimeter post and pier systems that do 

not provide a path for seismic forces. Inadequately braced cripple walls are commonly enhanced 

with sheathing and anchorage to the foundation. Where post and pier systems occur at the 

building perimeter, it is generally necessary to add a continuous footing and either a foundation 

stem wall or braced cripple wall. See related discussion of anchorage to the foundation in the 

Load Path section. 

Hillside buildings can be vulnerable when large variations occur in the stiffness of the system 

between the lowest framed floor and grade. Generally, the bracing for lateral loads at the uphill 

side will be significantly stiffer that the downhill side, attracting a much higher force. Flexible 

downhill systems permit significant deflection and diaphragm rotation. Hillside buildings can be 

improved by anchoring floor diaphragms to the uphill foundation, and by enhancing strength and 

stiffness of downhill bracing systems. 

Open front building portions occur when an exterior wall contains little or no bracing at any 

story level; common occurrences include garage fronts and window walls. Open front building 

portions can be rehabilitated by the addition or enhancement of shear walls or the addition of 

collectors, transferring seismic loads to portions of the building that have adequate shear walls. 

Split-level buildings have vertical offsets in the top of floor framing in adjacent portions of the 

building (i.e. sunken living room). Where floor framing with varying top elevations frames into a 

common wall, earthquake loading may cause one level of framing to separate from the wall, 

potentially causing local floor collapse. This behavior was seen in the San Fernando earthquake 

(ATC, 1976). Vulnerable split-level buildings are commonly rehabilitated by improving 

connections between framing on either side of the floor offset. 

Load Path 

The highest priority and most cost effective rehabilitation measure for W1 buildings is ensuring 

that the home is adequately anchored to the supporting foundation. Anchorage may use anchor 

bolts or proprietary retrofit anchors, and it may be done alone or in combination with cripple 

wall enhancement. In addition, a systematic evaluation of the seismic force-resisting system will 

often result in the need to rehabilitate load path connections. Load path improvements include 

shear anchorage to the foundation, uplift anchorage to the foundation, shear transfer load path in 

the wood framing, uplift load path in the wood framing, and collectors to shear walls. 

Rehabilitation measures primarily involve the addition of fasteners and connector hardware. 
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Component Detailing 

Many W1 buildings contain unreinforced, unbraced masonry chimneys, for which rehabilitation 

measures include removal, partial removal, infill, and bracing. Appendages such as exit stairs, 

porches and decks, and their roofs are commonly rehabilitated by improving seismic attachment 

to the main building structure. Inadequately anchored stone or masonry veneer in W1 buildings, 

if addressed, is most commonly removed, or removed and replaced with properly anchored 

veneer. 

Diaphragm Deficiencies 

A systematic evaluation may identify deficiencies in the diaphragm systems, including 

inadequate diaphragm strength and/or stiffness, inadequate shear transfer to walls, and 

inadequate detailing at large diaphragm openings and re-entrant corners. Diaphragm deficiencies 

have not stood out as a source of damage to W1 buildings. The removal and replacement of 

existing roofing, as part of regular building upkeep, often provides an opportunity for existing 

straight lumber sheathed diaphragms or spaced sheathing to be overlain with wood structural 

panel sheathing. Even though the roof diaphragm is seldom a top priority for W1 building 

rehabilitation, this can provide an opportunity to tie the roof together and achieve more 

monolithic behavior at a nominal cost. Rehabilitation measures can be found in Chapter 22. 

Foundation Deficiencies 

Common seismic deficiencies in foundations undergoing systematic rehabilitation include 

inadequate strength for shear wall overturning forces. Rehabilitation measures for foundation 

deficiencies are discussed in Chapter 23. Other deficiencies such as deteriorated foundations, 

sliding on unreinforced cold joints and settlement in cut and fill sites are not addressed by this 

document. 

5.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

5.4.1 Add New or Enhance Existing Shear Wall 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses insufficient global strength and/or stiffness, as well as 

local areas of insufficient strength and/or stiffness such as at open front conditions. Discussion is 

applicable to W1, W1A, and W2 buildings. In W1 buildings, it is most common for insufficient 

strength or stiffness to be local rather than global. 

Description of the Rehabilitation Technique 

This rehabilitation technique involves adding a new shear wall or enhancing an existing shear 

wall. The primary focus of the discussion is addition of wood structural panel (plywood or OSB) 

sheathing, fastening and connections to an already existing framed wall, as this is most common 

in W1 buildings. Additions or alterations may lend themselves to adding a new shear wall. 

Addition of a completely new shear wall and other options for enhancement of existing walls are 

discussed in Section 6.4.2. 
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As a fundamental element of shear wall addition or enhancement, this section includes 

discussion of load path for transfer of forces into and out of the shear wall. This discussion is 

applicable to building types W1, W1A, and W2, as well as other buildings types with wood floor 

and roof diaphragms. 

This section also discusses a few general topics relating to rehabilitation of existing woodframe 

buildings, including wood shrinkage, pre-drilling for fasteners, and wood species. The issues are 

applicable to building types W1, W1A, and W2, as well as other buildings types with wood floor 

and roof diaphragms. 

Design Considerations 

Research basis: Research that specifically discusses addition of shear walls in one- and twofamily detached dwellings has not been identified; however, a significant amount of testing and 

analysis on new shear walls and shear wall buildings can be considered applicable to this use. 

Primary references for shear wall testing are APA (1999a, 1999b), City of Los Angeles & 

SEAOSC (1996), Salenikovich (2000), Gatto and Uang (2002), and Pardoen et al. (2003). 

Testing of slender walls can be found in ATC R-1 (ATC, 1995). Testing of perforated shear 

walls can be found in Heine (1997). Testing of walls designed for continuity around openings 

can be found in Kolba (2000). Research addressing specific connections within the shear wall 

load path is referenced in the following discussions. 

Shear wall design method: New shear walls are primarily designed in accordance with 

provisions of the IBC (ICC, 2003a) or the Wind and Seismic Supplement (AF&PA, 2005b). 

Requirements for new shear walls should be used for design of new shear walls in existing 

residences in addition to the considerations addressed in this section. The IBC and the Wind and 

Seismic Supplement recognize three methods of analyzing wood structural panel shear walls: 

segmented, designed for continuity around openings, and perforated shear walls (Figure 5.4.1-1). 

All of these methods are acceptable. Design for continuity around openings will allow for use of 

slender wall piers, where necessary. The third method – perforated shear wall--was developed 

particularly for residential construction in order to minimize the required overturning restraint 

hardware. 

Shear wall location: Analytical studies have shown that one- and two-family dwellings will tend 

to have a concentration of deformation demand in the first story (lowest framed story) (Isoda, 

Folz, and Filiatrault, 2002) and (Cobeen, Russell, and Dolan, 2004). Therefore, under most 

circumstances, shear walls added in the lowest story are likely to have a larger impact on 

building performance than those added in upper stories, and lower stories should generally be 

given higher priority. 

In order for shear walls to function as part of the structural system, it is necessary to design for 

transfer of in-plane load from the diaphragm being supported into the wall top and transfer of inplane and overturning loads out at the wall base. In addition, the size and aspect ratio need to be 

adequate to meet demand, and significant disruptions over the height of the wall should be 

minimized. These considerations guide preferable locations for shear walls. 
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Figure 5.4.1-1: Shear Wall Design Methods 

Preferred shear wall locations: 

Exterior walls generally have inherent continuity of load path framing at the wall top and 

to a bearing foundation at the base (Figure 5.4.1-2). Conditions that can make exterior 

walls less effective include wall locations that are detached from the floor diaphragm 

(along stair opening or back of light-frame fireplace), walls that are balloon framed (wall 

studs continuous past floor framing), and walls that have interruptions over their height 
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(low roof framed into side of wall). Most of these conditions can be addressed with 

additional load path detailing, however, at greater cost and disruption. 

Interior bearing walls, like exterior walls, generally have inherent continuity of load path 

at wall top and bottom. 

Figure 5.4.1-2: Preferred and Less Preferred Shear Wall Locations 

Less preferable shear wall locations: 

Interior partition walls can be problematic due to lack of load path continuity at both the 

top and bottom of the wall. Inadequate support for overturning forces is generally the 

most difficult problem to solve. It is easiest to use second floor walls that are 

continuously supported on framed first story walls (Figure 5.4.1-3A) allowing 

transmission of uplift and downward loads to the foundation. As a second choice, it may 

be possible to use a section of an upper story wall that can be vertically supported by 

posts at each end, again allowing transmission of uplift and downward loads to the 
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foundation (Figure 5.4.1-3B). In both these support cases, the overturning stiffness of the 

shear wall should not be significantly different than shear walls located at the building 

exterior. As a last choice, floor framing systems can be enhanced to support interior shear 

walls (Figure 5.4.1-3C). 

Interior partition walls in residences with truss roof and/or floor systems (Figure 5.4.1-2) 

require special attention to wall location, and analysis and detailing in order to avoid 

damaging the trusses. 

Bathroom and kitchen plumbing walls can be problematic for use as bracing walls 

because of penetrations through the wall sheathing and because piping often results in 

breaks in the top and bottom plates serving as chords and collectors. 

Walls oriented at an angle to the primary framing direction can pose particularly difficult 

detailing issues. 

Figure 5.4.1-3: Overturning Support Conditions for Upper Story Shear Walls 

The addition of shear walls is often most needed in portions of residences where existing walls 

are too slender to provide effective bracing. Use of properly detailed wood structural panel shear 

walls assists in making slender shear walls effective in providing resistance. It may become 

necessary, however, to reduce window openings in order to provide adequate lengths of shear 

wall. 

All of the above limitations are only in response to the physical configuration of the residence. 

Other considerations in choice of wall locations include the level of disruption that is acceptable 

to the occupant and other planned work that may provide access for rehabilitation. 

Adequacy of foundation: Addition or modification to existing foundations can often be the most 

expensive portion of adding shear walls in existing residences. Shear walls produce concentrated 

uplift and downward loads at each end. Engineered shear walls are seldom added without 

addition of uplift anchorage. Where the shear wall is long enough and the overturning forces 
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low, the forces on the foundation can be modeling as two separate vertical forces, one up and one 

down. The downward load must be transferred to the supporting soils. Where the required 

bearing length does not exceed twice the depth of the foundation, the foundation capacity is not 

critical to footing resistance. Where a greater length is required, foundation shear and flexure 

capacity come into play. For the uplift anchorage it is necessary to have the foundation span far 

enough to mobilize dead load to resist uplift. Where slender walls are used, concentrated 

moments are introduced into the foundation by the closely spaced uplift and downward forces. 

This is particularly true of slender proprietary walls. 

Construction and capacity of the foundation will significantly impact the ability to withstand 

these concentrated forces. Continuous concrete foundations or foundation walls with reinforcing 

are preferred. Anchorage, shear capacity and flexure capacity can be particularly problematic 

with existing unreinforced brick masonry foundations, unreinforced concrete masonry 

foundations, partially grouted concrete masonry foundations, and isolated foundations of any 

material. Addition of new foundations is often required. New foundations cast along side and 

tied into existing foundations can have the advantage of mobilizing the resisting weight of the 

existing foundation, as can new foundations that run between and dowel into existing 

foundations. 

Figure 5.4.1-4 shows a new continuous footing cast alongside an existing footing. Adhesive 

anchors are drilled into the existing footing at a regular spacing so that if the new footing uplifts, 

it will also pick up the existing footing. The adhesive anchor can be a bolt, as shown, or 

reinforcing steel designed for shear friction. The bolt or reinforcing is designed to transfer the 

required vertical resisting load. Design must consider concrete anchor capacity including edge 

distance effects. Reinforcing steel should be anchored on both sides of the interface to develop 

the bar yield. Preparation of the existing concrete surface would normally involve cleaning only; 

intentional roughening is generally not practical. 

Detailing Considerations 

General: A few topics deserve general consideration before getting into the specifics of shear 

wall detailing, including shrinkage, predrilling, wood species, corrosion issues, and condition 

assessment of existing buildings. Shrinkage of wood framing members is an issue that must be 

considered in design of both new and existing wood buildings. Shrinkage of wood framing as it 

drops to equilibrium moisture content is accommodated in new construction every day. Whether 

in new construction or rehabilitation, the primary concern is differential shrinkage where 

members subject to shrinkage might act in a system with members subject to lesser shrinkage, no 

shrinkage, or possibly even slight swelling. In rehabilitation, new framing members subject to 

shrinkage may need to be added in parallel to members that are already at equilibrium moisture 

content. Shrinkage in the length of framing members is negligible. The primary shrinkage of 

concern is in the width of members. With a combination of radial and tangential directions, 

shrinkage on the order of 6% or 3⁄4” in 12 inches is reasonably possible. This could mean a gap of 

3/4 inches developing between blocking and the diaphragm above in a shear transfer or similar 

connection, greatly reducing the resistance provided. 
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Figure 5.4.1-4: New Continuous Foundation Cast Along Side Existing 

to Provide Capacity for Tie-Down Anchor 

Effects of shrinkage are best mitigated by use of dry framing members and detailing to minimize 

reliance on configurations susceptible to shrinkage problems. Equilibrium moisture content for 

enclosed buildings is most often in the range of 7 to 12 percent. The closer new framing is to this 

range at time of installation, the less the potential shrinkage problems. This can be accomplished 

by setting aside framing (purchased green, at MC19, or at MC15) in a protected location to dry. 

In a dry season, the moisture content can drop significantly in the range of several weeks to 

several months. Another approach is to use engineered wood members such as glulams, which 

are manufactured at low moisture content. Laminated veneer lumber (LVL) and similar 

engineered wood products can also be used; however, the manufacturers restrict the size and 

spacing of nails into the top and bottom faces of these members due to concerns of splitting 

along lamination lines; this limits these members to use for low to moderate shear transfer loads. 

Splitting of wood framing due to new fastening during rehabilitation is of significant concern. 

Nails that can easily be driven into new framing can be very difficult in existing framing, and 

splitting can occur. The current building code approach to splitting of members is primarily a 

performance approach. If members are split, the fasteners are not considered to provide capacity. 

This approach is of little help once splitting of critical structural members has already occurred. 

Repair and replacement of existing members can be very difficult. Predrilling for nails and other 

fasteners prior to installation will substantially reduce the risk of splitting framing members. 

Details of predrilling requirements are given in the NDS (AF&PA, 2005a). 
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Wood species is another item of general concern for detailing. The design values of wood 

fasteners and shear walls are a function of the framing density and, therefore, the wood species 

being fastened. The species of framing used may have varied over time. Older buildings may be 

framed with species that are no longer commonly used. Fastener, shear wall and diaphragm 

values need to be adjusted for the framing used. In very occasional cases, it might be desirable to 

determine the density of existing framing in order to identify the best choice of fastener values. 

Corrosion of fasteners and connectors due to pressure preservative treatments is currently a 

concern for new construction due to recent changes in treatment formulation. This concern and 

related cautions regarding use of corrosion resistant fasteners and connectors is equally 

applicable where preservative treated wood is added in rehabilitation. 

In woodframe buildings, deterioration of the structure can particularly impact seismic 

performance and the ability to implement seismic rehabilitation measures. For this reason it is 

important that condition assessment of critical elements of the existing woodframe structures be 

considered. See Section 2.3.8 of this document for additional discussion. 

Sheathing and fastening: Added sheathing will generally be wood structural panel sheathing 

(plywood or OSB). In very unusual circumstances, addition of diagonal lumber sheathing might 

occur. The choice of extent and unit shear for sheathing and fastening is a balance between cost 

and performance. In general, providing more sheathing at a lower shear capacity results in less 

building deformation and better building performance. As with any system, well-distributed 

resistance is always better than heavy concentrations of resistance in local areas. In addition, 

when sheathing fastening is being added to existing dry wood members, close fastener spacing 

increases the possibility of member splitting. This is particularly true in members on which 

sheathing panel edges abut. Under the IBC and Wind and Seismic Supplement, use of close nail 

spacing on shear walls will trigger a requirement for minimum 3x studs at adjoining panel edges. 

Since 3x framing will seldom already occur in an existing wall, two options generally result. 

First 3x or 4x members can be added, and wood structural panel sheets lain out to fall on these 

members, or a new 2x stud can be added along side an existing stud, and the two “stitch-nailed” 

to provide adequate interconnection. Shear walls with stitch-nailed 2x’s at abutting panel edges 

were tested recently by APA and found to provide acceptable behavior (APA, 2003). A 

provision permitting “stitch-nailing” has been incorporated into the 2004 supplement to the IBC 

(ICC, 2003a). 

The IBC requires the use of 3x foundation sill plates for shear wall unit shears over 350 plf, 

while the NEHRP Provisions (FEMA, 2003) permit 2x plates in combination with steel plate 

washers on anchor bolts. In rehabilitation work, it is seldom practical to replace or modify the 

existing foundation sill, so practice is to retain the existing sill. The IBC and predecessor UBC 

(ICBO, 1997a) requirements for 3x sills primarily address cross-grain splitting of foundation sill 

plates, observed in the Northridge earthquake and laboratory testing (SEAOC, 1999). In recent 

testing of shear wall anchorage to foundations (Mahaney & Kehoe, 2002), as discussed in 

Cobeen, Russell, and Dolan (2004), the best performance of foundation anchorage was seen with 

3x foundation sill plates; however, significant numbers of loading cycles were resisted by 2x 

plates with steel plate washers, supporting continued use of 2x plates in rehabilitation. Where a 
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performance objective more stringent than one such as the FEMA 356 Basic Safety Objective is 

being used, however, replacement with 3x sills should be considered. 

It is recommended that existing finishes be removed, allowing new structural sheathing to be 

installed directly over framing whenever possible. This permits an opportunity to observe the 

condition and fastening of existing framing, to install shear and overturning connections, and to 

add boundary member framing if required. The IBC permits wood structural panel sheathing to 

be installed over gypsum wallboard for fire-rating purposes; increased nail size is required. 

Increasing the distance from the center of sheathing nails to the edge of sheathing panels from 

3/8-inch to 3⁄4-inch has been seen to reduce fastener failure due to tear-out at the panel edge and 

greatly toughen the shear wall (Cobeen, Russell and Dolan, 2004). This is easily accomplished at 

top plates, bottom plates and end posts where only one row of edge nailing needs to be provided. 

It requires use of wider framing at interior wood structural panel joints where two panels abut 

and are edge nailed to a single framing member. 

Buildings that have exterior wood structural panel siding present a unique opportunity to 

improve sheathing fastening without opening up finishes. In many cases only one of the two 

edges at abutting panels will be properly nailed. Providing full edge nailing on both panels can 

improve shear capacity. Nailing may be exposed on the siding exterior, or may be under trim 

boards which can be removed and replaced. Corrosion resistant fasteners are needed for siding 

nailing. 

Sheathing to framing fastening with staples and use of wood structural panel overlays are 

discussed in Section 6.4.2. 

Shear transfer criteria (when using FEMA 356): FEMA 356 (FEMA, 2000) identifies fasteners 

used to transfer forces from wood to wood or wood to metal as being deformation-controlled 

actions. When coupled with several relatively high m-factors for static procedure acceptance 

criteria, this can result in less fastening being required by FEMA 356 than the current building 

codes. At the same time, the shear wall sheathing fastening is identified as the desired location of 

inelastic behavior, which suggests that fastening for shear transfer into the shear wall should be 

force-controlled and more fastening provided. Because shear transfer nailing has only rarely 

been seen as a critical weak link in earthquake performance to date, it is recommended that 

current building code requirements be used for a basic safety objective. For a higher performance 

objective where inelastic behavior of the shear wall is anticipated, the proportioning of fastening 

relative to anticipated shear wall demand should be considered. 

Shear transfer into top of wall: The addition of sheathing and fastening is not of value unless 

shear forces can be transferred into the top of the wall. Where sheathing is added to an existing 

wall line, the wall top plates will most often serve as the collector. Where top plates are not 

present, or are not continuous for a reasonable distance, a supplemental collector should be 

provided. 

Figure 5.4.1-5 shows a series of top of shear wall details where the shear is being transferred 

from a roof diaphragm into the top of the wall. Since most diaphragms in residential construction 

are not blocked, the unit capacity of the new shear wall is likely to be higher than the unit 
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Figure 5.4.1-5: Load Path from Roof Diaphragm to Top of Shear Wall 

capacity of the diaphragm above. For lightly loaded shear walls, the minimum length of the 

diaphragm to be connected into the shear wall can be calculated, and a collector provided to tie 

the diaphragm into the top of wall. For highly loaded vertical elements, it is recommended that 

the collector extend for the full diaphragm length, as discussed in Section 7.4.2. 

In new construction, attachment of floor or roof sheathing to shear walls below typically requires 

nailing through the sheathing into framing below, as shown in Figure 5.4.1-6A. While this 

attachment remains the preferred approach, installation of nailing is not possible where roof or 

floor finishes cannot be removed. Figures 5.4.1-6B, 5.4.1-6C and 5.4.1-6D show alternative 

attachments of roof or floor sheathing. Significant cautions are applicable when using either of 

the alternative approaches, as detailed below. 

Limited testing of the connections shown in Figures 5.4.1-6B and 5.4.1-6D occurred in the 

CUREE-Caltech Woodframe Project (Mosalam et al., 2002). The purpose of the testing was to 
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find the best method of attaching new steel moment frames to existing wood buildings. The 

specimens used 12-inch deep joists and blocks in two 16-inch bays and tested angle clip 

connections monotonically and cyclically and adhesive connections monotonically. Due to the 

geometry of the test specimen, overturning behavior was significant. Both methods of attachment 

increased the load capacity beyond that for minimum framing nailing. The attachment of the 

blocking to the sheathing was not a controlling factor in any of the tests. 

Figure 5.4.1-6: Attachment of Blocking to Existing Sheathing 

Where unit shears are low and a nailed sheathing to framing connection is not possible, 

connection of sheathing to framing using steel clip angles provides a possible alternative 

(Figures 5.4.1-6B and 5.4.1-6C). The clip angle is generally attached to the framing with nails 

and to the sheathing with wood screws. NDS requires a minimum penetration of six times the 

wood screw diameter into the sheathing (note that the length of the screw point is included when 
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calculating the 6 diameter penetration). This minimum penetration requirement results in use of 

very small screws, with very small capacities, making this connection type practical only when 

unit shears are low. If Number 4 screws are used (the smallest size generally available) the 

penetration into 1x sheathing with an actual dimension of 5/8 inches will be just short of meeting 

this penetration requirement. The Number 6 screws used in 3⁄4-inch plywood in testing also fell 

just short. Use of increased penetration is encouraged whenever possible. 

Along with caution due to the low capacity of the screws, two other significant cautions should 

be considered. First, during installation of the wood screws into the sheathing it is very easy to 

overdrive the screw, stripping out attachment to the sheathing. This is particularly easy when 

installation is with a screw gun, and it is even more so when the wood screw is connecting a steel 

clip, because the drawing of the screw head against the sheathing is not visible to the installer. 

Second, if the screw used is too long, it will penetrate the top surface of the sheathing. Care must 

be taken to not penetrate where the top surface is roofing or a sensitive finish. The thickness of 

the clip angle and protrusion of the fastener head generally reduce the screw penetration by 1/16 

to 1/8 inch, which should be considered in specifying screw length. Considerable attention to 

quality control and quality assurance is recommended if this detail is to be used. 

Where use of a nailed connection is not possible, adhesive connection from sheathing to framing 

provides a second alternative (Figure 5.4.1-6C). Adhesive attachment of sheathing to framing is 

discouraged in diaphragm assemblies in which inelastic behavior is anticipated, such as longspan and high load diaphragm systems. This is because adhesive connections do not allow slip 

between the sheathing and framing and do not permit energy dissipation, which generally occurs 

through nail bending. As a result, a glued diaphragm would be anticipated to behave nearly 

elastically up to a failure load and then fail in a brittle manner. In addition, adhesive sheathing to 

framing connections will be significantly stiffer than nailed connections, attracting higher loads 

to the adhesive where both types of attachment are used in combination. For these same reasons, 

use of adhesive in shear walls resisting seismic forces is not recommended, although the NEHRP 

Provisions do permit use in Seismic Design Category A, B or C, using and R-factor of 1.5. 

In most W1 and W1A buildings, however, it is anticipated that inelastic behavior will be 

concentrated in shear walls and other vertical elements, making use of adhesives in diaphragm 

connections an alternative. It is recommended that, when used, adhesive connections be designed 

for maximum expected forces (either overstrength forces, or using a very small R-factor or mfactor). 

Adhesives used in recent testing have included cartridge types, applied using caulking guns, and 

spray-on self-expanding foam adhesives. Foam adhesives are also being used for attachment of 

roof sheathing to framing in high-wind regions. In this case the adhesive improves both wind and 

seismic resistance. Cautions when using adhesive sheathing to framing connections include the 

following: first, great care must be taken in ensuring that adhesives do not harden before 

blocking placement, as adhesives can have limited pot lives. Second, adhesives should be used in 

connections that minimize overturning rotation (continuous joists or shallow blocks) so that 

tension on the glue joint is minimized. Again, significant attention to quality control and quality 

assurance are recommended when using this connection alternative. 
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When attaching to the roof, required roof cross-ventilation needs to be maintained. This can 

influence details both at the roof perimeter and interior, as shown in Figure 5.4.1-5. 

Load transfer from a roof diaphragm through a roof truss system into the top of a shear wall can 

be very complicated at both bearing and nonbearing partition walls. The complication comes 

from two sources. First, the shear wall must be extended through the roof truss system. Where 

the shear wall is parallel to truss members, this may simply mean placing the shear wall off the 

roof truss line and extending it to the roof sheathing. Where the shear wall is perpendicular to the 

roof framing, infill panels between the roof trusses are added to act as shear wall extensions 

(similar to Figure 5.4.2-2). Second, because existing nonbearing walls will often be attached with 

clips that permit vertical movement of the truss, the addition of a shear wall can create an 

unintended reaction, changing truss forces, and if between truss panel points, potentially leading 

to fracture of a truss chord. Connections are best made to existing trusses at truss panel points 

and should never be made without evaluating the potential change in truss forces. 

Figures 5.4.1-7 and 5.4.1-8 show a series of details where shear is transferred into the top of a 

shear wall at a framed floor level. Note that Figure 5.4.1-8A shows an existing balloon framed 

condition prior to rehabilitation. Figures 5.4.1-8B through 5.4.1-8D show rehabilitation 

alternatives. 

Shear transfer out of wall: Second story or higher shear walls will generally be supported on 

wood floor framing. Figure 5.4.1-9 illustrates common details for shear transfer at the wall base. 

See also the following discussion of overturning forces. 

First story shear walls may be supported directly on foundations, or on framed floor systems 

supported on foundations or foundation walls. A detailed discussion of shear transfer anchorage 

to existing foundations can be found in Section 5.4.3. See also the following discussion of 

overturning forces. 

Disruption over height of wall: Where shear wall sheathing cannot be placed in a continuous 

plane over the full height of a shear wall, additional detailing for continuity is needed. Disruption 

of the shear wall sheathing occurs most often where a floor or roof frames into the wall between 

floor levels, such as at a stair side or landing, a one-story roof hitting the side of a two-story 

section, a split-level floor, or a deck ledger. Testing done in the CUREE-Caltech project showed 

that shear wall studs that lose support from the sheathing can fail in weak axis bending. The 

same vulnerability could potentially occur where shear wall sheathing stops below an obstruction 

and then starts again above. Figure 5.4.1-10 shows methods for maintaining shear wall continuity 

across this type of disruption. 

Overturning at wall base: Figure 5.4.1-3 illustrates a series of overturning support conditions 

that may occur at the base of second story shear walls. Continuity for the uplift and downward 

loads are required at each end of the upper story wall. Figure 5.4.1-11 shows common detailing 

for the overturning load path. Plumbing, electrical and mechanical utilities often run through the 

floor framing, greatly complicating addition of new floor framing members under second story 

shear walls. 
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Figure 5.4.1-7: Load Path From Floor Diaphragm to Lower Story Shear Wall 

Figure 5.4.1-12 illustrates common overturning support conditions at the base of first story shear 

walls. See also the earlier discussion of foundation design issues. See Section 6.4.4 for 

discussion of anchorage to concrete issues under recent ACI 318 (ACI, 2005) provisions. 
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Figure 5.4.1-8: Load Path From Floor Diaphragm to Lower Story Shear Wall 

– Balloon Framing 
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Figure 5.4.1-9: Load Path from Upper Story Wall-To-Floor Diaphragm 
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Figure 5.4.1-10: Load Path at Disruption in Shear Wall Sheathing 

Reduction of slender shear wall height: Shear walls at the sides of garage doors and other large 

openings are often very slender and, therefore, develop significant overturning forces. One easy 

and relatively inexpensive approach of modestly reducing overturning forces and increasing wall 

stiffness is shown in Figure 5.4.1-13. A steel collector strap is run across the full length of the 

wall near the bottom of the door header. This strap will effectively reduce the shear wall height 

to the height of the door opening; in addition, limited moment fixity may develop at the wall top. 

The strap is nailed to the header and to blocking added in line with the bottom of the header. 

The strap is best placed over the wood structural panel sheathing, so that strap nailing provides 

shear transfer to the sheathing. Alternately the strap can be placed on the opposite face of the 

framing, however fastening of the sheathing to the blocking and header is also required. This 

approach can be used alone, or in combination with rehabilitation of anchorage and sheathing. 

Cost, Disruption, and Construction Considerations 

Addition of sheathing and fastening to woodframe shear walls can often occur while the dwelling 

is occupied. Work will generally progress faster, however, without occupants. Where feasible, 

work on the outside face of exterior walls often provides not only the best access, but also the 

best load-transfer detailing options. Added sheathing that increases the thickness of a shear wall 

will require adjustment of trim at openings and reworking of water barrier detailing at windows 

and doors. Completely sheathing an exterior wall, including areas above and below windows and 

doors gives not only improved structural performance, but also the best surface for correctly 

installing windows and water barriers. 
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Figure 5.4.1-11: Load Path for Overturning (Tension and Compression) 

at Upper Story Shear Wall 
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Figure 5.4.1-12: Load Path for Overturning (Tension and Compression) at Foundation 

Figure 5.4.1-13: Reduction of Slender Shear Wall Effective Height 
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As in new construction, it can be a challenge to assure that rehabilitation measures are 

constructed with the fastener (nail, staple, screw, etc.) type and size that has been assumed in 

design and construction documents. Use of improper type and size often results in reduced 

rehabilitation measure capacity. Most nails are placed with nail guns. Most gun nails are ordered 

by diameter and length. Indications of type and pennyweight continue to be misleading. The only 

way to verify that required fasteners are being used is to measure them with calipers or a similar 

device. Fasteners connecting sheathing to framing should not be overdriven (not break the face 

of the sheathing). Where overdriving occurs, fastener capacity may be reduced up to 40%. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique other than the use of 

proprietary connectors and adhesives as part of the assemblage. 

5.4.2 Add Collector at Open Front 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses configuration deficiencies created by an open front 

condition such as at a garage or window wall. 

Description of the Rehabilitation Technique 

Often in W1 buildings, an open front will occur at a portion or wing of the building, while 

adequate shear walls are provided in an adjacent portion. A common example of this is a lack of 

shear wall at the front of a garage, while sufficient bracing exists in the adjacent portion, as 

illustrated in Figures 5.4.2-1 and 5.4.2-2. Where this type of condition occurs, a collector can be 

used to transfer seismic forces generated in the open front portion to the adjacent portion with 

adequate shear walls. In woodframe construction, most shear walls are capped by double top 

plates that can be used as collectors. Figure 5.4.2-1 shows the collector connecting from the roof 

diaphragm at the garage to top plate collectors at the front of the house. Figure 5.4.2-2 shows the 

collector connecting from the second floor diaphragm above the garage to double top plates at 

the front of the house. 

Design and Detailing Considerations 

Research basis: No research applicable to this rehabilitation measure has been identified. 

General design: Collector connections like the ones illustrated in Figures 5.4.2-1 and 5.4.2-2 are 

often complex, and they can include both vertical and horizontal offsets between bracing lines. 

Rehabilitation is seldom inexpensive, and alternatives such as added shear walls should always 

be considered. If used, however, collectors should help to mitigate differential movement 

between the one-story and two-story portions of the building and to reduce resulting damage. 

The collector will most often but not always need to resist both tension and compression. 

Figure 5.4.2-1 shows one of several possible methods of providing a collector. In the illustrated 

approach, a steel strap ties the top plates from the garage open-front to wood structural panel 

sheathed infill panels between the roof trusses in the one-story portion of the building. The infill 

panels transfer the load from the truss bottom chord up to the roof diaphragm, where loads can 

be carried to the shear walls. In Figure 5.4.2-1, a vertical eccentricity exists between the collector 
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Figure 5.4.2-1: Collector from Garage Open Front to Adjacent Dwelling 
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Figure 5.4.2-2: Collector from Garage Open Front to Adjacent Dwelling 
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level and the roof diaphragm in the one-story portion. This eccentricity is resolved by continuing 

the sheathing infill panels for the full width of the one-story building so that the vertical reaction 

can be resisted at the exterior walls. An alternate approach would be to install a wood structural 

panel ceiling diaphragm on the underside of the roof trusses, in which case no vertical 

eccentricity would exist. 

Figure 5.4.2-2 shows a steel strap from the underside of floor joist blocking above the garage to 

top plates in the adjacent framed wall. The floor blocking transfers load from the strap to the 

floor diaphragm above. The depth of the floor blocking creates a small vertical eccentricity, 

causing the blocks to overturn. End nails or toenails at each end of the blocking generally resist 

this overturning. See Figure 5.4.1-6 and related discussion for connection to the floor and roof 

diaphragm sheathing. 

Deformation of collector: The collector will only be able to protect the open front against 

excessive drift if the deformation in the collector system is kept to a minimum. Elongation of the 

steel collector strap and nail slip are likely to be the primary contributors to deformation. Loads 

in the strap and nails should be kept moderate. 

Other parts of the load path: When the double top plate serves as a portion of the collector, 

breaks in the double top plates may require steel straps in order to provide adequate capacity. In 

order to complete the load path, diaphragm capacity, roof diaphragm connections to the top 

plates, and splices in the top plates should all be checked. 

Cost, Disruption and Construction Considerations 

Installing the collector connection shown involves opening up ceiling finishes in both portions of 

the building and extensive work infilling between the roof trusses. Other solutions to bracing of 

the open front should be explored. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique, other than the use of 

proprietary connectors as part of the assemblage. 

5.4.3 Add or Enhance Anchorage to Foundation 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses insufficient shear connection between woodframe 

dwellings and their foundations. The highest priority and most cost effective rehabilitation 

measure for W1 buildings is ensuring that the home is adequately anchored to the supporting 

foundation. This technique is equally applicable to W1A and W2 buildings. Enhanced anchorage 

may be provided from the foundation to first story walls, to floor framing, or to cripple walls. 

Enhanced anchorage is often used in combination with cripple wall enhancement as discussed in 

Section 5.4.4. 

Description of the Rehabilitation Technique 

Foundation anchorage can often simply involve anchor bolts connecting a foundation sill plate to 

the supporting continuous foundation or foundation wall. The intent is to transfer the earthquake 
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horizontal base shear from the foundation sill plate into the foundation; nominal uplift capacity is 

often also provided by the anchorage. The primary objective is to keep the foundation sill and 

framed building above from sliding relative to the foundation under earthquake loading. Shear 

transfer to isolated footings or short foundation piers is not recommended without evaluation of 

the footing and transfer to the supporting soils. Where configuration and access prohibit 

installation of anchor bolts, proprietary anchors are used to transfer horizontal shear to the 

foundation. 

Figure 5.4.3-1 illustrates common anchorage details using anchor bolts to existing concrete 

foundations. Figure 5.4.3-2 illustrates an anchor bolt connection to an existing masonry 

foundation. Where possible, anchor bolts remain the preferred method of anchorage to 

foundations. Where the existing foundation is concrete masonry (Figure 5.4.3-2), grout may not 

exist in all masonry cells. The existence of grout at the added anchor should be confirmed. 

Where anchorage into grouted cells is not possible, cutting out face shells and pouring grout 

around an added anchor bolt is a preferred alternative. As a second alternative, adhesive anchors 

intended for connection to hollow bases can be used; however, capacities are very low. A 

combination of anchorage to grouted and ungrouted cells is not recommended. 

Steel plate washers need to be provided at each added anchor bolt between the foundation sill 

plate and the nut. Current codes require that the steel plate washer be a minimum of 1⁄4”x3x3”, 

and they allow a slotted hole to accommodate bolt location tolerances. Shear wall anchorage to 

foundations has been tested by Mahaney and Kehoe (2002). 

Installation of anchor bolts in first story shear walls involves the removal of finish materials. 

Where shear wall rehabilitation per Section 5.4.1 is already being provided, finishes will 

generally be removed in order to access framing. Where finishes or structural sheathing are not 

otherwise going to be removed, it is possible to create access for anchor bolt installation by 

removing finishes over the bottom two to three feet of the wall (Figure 5.4.3-1A). Where 

structural sheathing is removed for access, blocking needs to be provided at all sheathing panel 

edges so that edge nailing can be provided when the sheathing is replaced. 

In the configuration shown in Detail 5.4.3-1D, the existing foundation sill is wider than the 

existing studs. 2x4 blocking is added between the studs and nailed down to the foundation sill 

plate. In prescriptive provisions this is most often with four 10d common nails. Cripple wall 

retrofits using this base detail were tested by Chai, Hutchinson and Vukazich (2002) and 

performed well in testing. House inspectors, however, have reported seeing splitting of the 2x4 

block in homes that have been retrofitted using this approach. Alternative fastening approaches 

include using nails with pre-drilling, using staples, and using wood screws between the block and 

the foundation sill plate. Another approach is to cut the foundation sill plate flush with the studs 

above so that blocking is not required. No testing is available to judge the relative performance 

of these approaches. 

Addition of anchor bolts is often not possible with a crawlspace configuration due to inadequate 

vertical clearance for a rotary-hammer to drill down into the top of the foundation. Figure 5.4.3-3 

illustrates some of the alternate proprietary anchors that can be used for these configurations. 

Although shown with stud walls above the foundation sill plate, these connections work equally 
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Figure 5.4.3-1: Added Anchor Bolt at Existing Concrete Foundation 
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well when floor framing sits directly on the foundation plate. The steel angle connection in 

Figure 5.4.3-3B is generally not recommended as an alternate to anchor bolts for in-plane shear 

due to flexibility and potentially causing cross-grain splitting of the joists; other depicted anchor 

types resist in-plane shear much more effectively. ICC Evaluation Service reports should be 

consulted for anchors to the foundation and alternate proprietary anchors. 

Pier-and-curtain wall foundations (Figure 5.4.3-4) are used in some areas of the southern United 

States. As-built anchorage for shear transfer between the wood framing and foundation is 

generally minimal to non-existent. Rehabilitation of anchorage to this type of foundation is not 

known to have been undertaken to date. One possible approach is a continuous steel angle from 

the underside of the floor framing to the inside face of the single-wythe curtain wall, anchored to 

the curtain wall with veneer anchors and to the wood with nails or screws. Care would need to be 

taken in drilling for veneer anchors. An alternate approach would be new concrete or masonry 

foundations from pier to pier, allowing use of cast-in anchor bolts to the foundation and nailed or 

screwed connections to the wood framing. 

Figure 5.4.3-2: Added Anchor Bolt at Existing Partially 

Grouted Concrete Masonry Foundation 
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Figure 5.4.3-3: Anchorage to Existing Foundation Using Proprietary Connectors 
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Figure 5.4.3-4: Pier and Curtain Wall Foundation System with Inadequate 

Load Path Between Shear Walls and Foundation 

Design and Detailing Considerations 

Research basis: Testing of shear wall to foundation anchorage has been conducted by Mahaney 

& Kehoe (2002). Testing of prescriptive cripple walls anchored to foundations has been 

conducted by Chai, Hutchinson & Vukazich (2002). 

Anchor type and installation: A variety of proprietary anchors are available for anchorage to 

existing concrete and masonry foundations. Both manufacturer literature and ICC Evaluation 

Service reports should be consulted for information on conditions of use, allowable loads, and 

installation and inspection requirements. It is important to make sure that the anchor type is 

appropriate for the material being connected to, is approved for seismic loads, and is appropriate 

for weather and temperature exposure. Either adhesive or expansion anchors to the existing 

foundation are commonly used; however, because expansion anchors create splitting tensile 
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forces, the proximity to the foundation edge and strength of existing foundation material may 

make use of adhesive anchors a better choice. In addition, some concerns have been raised 

regarding potential relaxation of expansion anchors under seismic loading. Use of powder-driven 

fasteners for anchorage to concrete or masonry is not recommended due to concerns regarding 

performance under cyclic loading (Mahaney & Kehoe, 2002). The diameter of drilled holes is 

specified in installation requirements for each anchor type; variation from this size often leads to 

inadequate anchor capacity. 

Most manufacturers have caulking gun-like devices that make field placement of adhesives fairly 

simple and automatically mix two-part adhesives. Generally, these types of adhesives provide 

more than adequate strength, and there is no need to use more complicated high-strength 

adhesive types. The cleaning of holes prior to placing adhesive anchors is paramount for anchor 

capacity. When not well cleaned, the anchors can pull out at a small fraction of the design load. 

It is common to pull-test a portion of the adhesive anchors to verify adequate installation. The 

pull test load is usually in the range of one to two times the tabulated allowable stress design 

tension load. The bridge used for testing generally makes a concrete pull-out failure unlikely. 

The test load should not be near yield load for bolts or adhesive pull-out (bond) failure loads. 

Use of nonshrink grout in lieu of adhesives for anchor bolt attachment is another possible 

installation alternative. This approach was commonly used prior to adhesives being readily 

available. If used, literature from the grout manufacturer should be consulted for installation 

requirements and anchorage design procedures. The hole drilled for anchor placement is often 

required to be 1/4–inch (or more) larger that the diameter of the anchor being placed. This size of 

hole may not be practical near the edge of a foundation and in weaker foundation materials. 

When using this approach, it is important that the anchorage design consider the implications of 

full expected seismic loads, rather than just code level loads. 

Anchors will very often need to be installed near the exterior edge of a foundation. Typical 

anchor bolt placement in nominal 4-inch walls results in a distance from center of bolt to edge of 

concrete of 1-3/4 inches. Due to this edge distance, reductions in anchor capacity will likely 

apply. In addition, it is recommended that a minimum clear cover distance be maintained 

between the face of the anchor and the exterior face of the foundation. Where the exterior face of 

the foundation in the vicinity of the anchor bolt has been formed, ACI 318 Appendix D would 

require a clear cover of 1-1/2 inches in new installation. This provides reasonable guidance for 

rehabilitation also. In addition the placement of the anchors will be limited somewhat by the 

dimensions of the steel plate washers. Where possible, moving an anchor away from the edge of 

the foundation will result in a stronger foundation anchorage, but may not affect the wood to 

steel capacity. When the anchorage is at the base of a sheathed shear wall or cripple wall, it is 

best to keep the anchor as close as practical to the sheathed face of the studs in order to minimize 

risk of sill plate cross-grain splitting. 

Configuration implications: Where foundation anchors are being installed in a crawl space, the 

design of anchorage to the existing foundation will be driven almost entirely by the configuration 

of the existing foundation, sill plate and framing configuration. A good look at existing 

conditions is needed before design is started. Limitations on access for materials and equipment 

will often limit anchorage methods. 
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Prescriptive and engineered anchorage: Prescriptive provisions for anchorage of foundation sill 

plates and cripple walls can be found in the International Existing Building Code – IEBC -- 

(ICC, 2003b). These were developed from similar or identical provisions in the GSREB (ICBO, 

2001) and the UCBC (ICBO, 1997b). An extensive commentary to the GSREB Chapter 3 

provisions has been developed by SEAOC Existing Buildings Committee (ICC, 2005). Some 

organizations have developed local adaptations of these provisions. The objective of these 

prescriptive provisions is reduction of earthquake hazard; they are intended to provide a 

reasonable level of improvement for the majority of buildings within their scoping limitations. 

W1 buildings with unusual configurations, site slopes greater than one vertical in ten horizontal, 

or higher performance objectives should be addressed with an engineered design. An engineered 

design is recommended for all W1A and W2 buildings because of higher loads and potential 

configuration issues. 

Engineered design for anchorage without specifically identified superstructure shear walls: In 

cases where the prescriptive provisions are not applicable, it may be desirable to provide an 

engineered design for foundation anchorage, with or without cripple wall bracing. An engineered 

design allows load distribution to the cripple walls to be addressed for the specific building 

configuration and allows specific design for non-standard framing and foundation conditions. 

Where rehabilitation will be limited to anchorage to the foundation, it is common to make 

simplifying assumptions regarding force distribution. For small buildings, forces generated at 

and above the lowest framed floor are distributed by tributary area to the perimeter foundations. 

For larger buildings, force may also be distributed to interior foundations based on tributary area. 

In addition to providing foundation anchorage at engineered cripple walls, it is desirable to 

provide a minimum level of anchorage for all foundation sill plates to avoid loss of vertical 

support should building movement occur. 

Engineered design for anchorage with specifically identified superstructure shear walls (see also 

Section 5.4.1): Where shear walls are being added or enhanced in the story above the crawlspace, 

the foundation anchorage design will need to specifically provide a load path for the shear wall 

reactions. 

Adequacy of foundation: Shear anchorage of a woodframe building to a foundation generally 

puts modest demands on the foundation. In order to perform adequately, the foundation needs to 

resist local demands from the anchor installation (such as drilling as splitting tensile stresses if 

installing expansion anchors), and it needs to have enough continuity to distribute the seismic 

shear forces without local failure. Installation of shear anchorage into existing reinforced 

concrete or masonry footings or foundation walls is commonly done without any specific 

evaluation of the foundation capacity. Likewise, shear anchorage to an unreinforced concrete 

foundation in good condition is commonly done without specific evaluation. Evaluation is 

needed when any foundation shows signs of deterioration due to differential movement, 

moisture, or other causes. Foundations that are moving differentially should be stabilized prior to 

installation of anchorage. If not stabilized, further movement of the foundation can telegraph into 

deformation and damage in the building above. 

Views on addition of shear anchorage between woodframe dwellings and unreinforced masonry 

foundations vary widely. In some regions, there is considerable concern that unreinforced brick 
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foundations are fragile due to moisture driven deterioration and lack of confining overburden. 

Approaches to shear connections taken in these regions include casting new foundations 

alongside existing foundations and cutting out blocks of existing foundations in order to place a 

concrete key around added anchor bolts. In other regions, it is more common to recommend 

bolting woodframe dwellings directly to unreinforced brick masonry foundations that are in good 

condition. IEBC Chapter A3 requires an engineering evaluation of unreinforced masonry 

foundations, but does not provide details of the required evaluation. This allows some flexibility 

for anchorage practice to be determined locally based on local concerns, experience, foundation 

materials and construction practice. Load testing of anchorages should be considered as a quality 

assurance measure, particularly when new combinations of foundation materials and anchorage 

methods are being used. Addition of overturning anchors or concentrated loads requires specific 

evaluation of foundation capacity.

Special attention is needed where a masonry foundation is constructed of large cut stones 

because use of typical connections is impractical. 

Prestressed foundations: Where foundations contain prestressing tendons, it is important to 

locate tendons prior to drilling for foundation anchorage. Tendons cut during drilling for 

anchorage may fail explosively, either along the length of the tendon or at the tendon anchorage, 

potentially causing injury and damage. Original design drawings identifying tendon locations 

and profiles are of great value in understanding placement. Alternately, post-tensioning experts 

can field locate tendon anchorages and profiles. 

Alternate anchorage configurations: In California, encouragement of anchor bolting at the state, 

county and local government level has led to a noticeable amount of retrofit for anchorage to 

foundations. The lack of mandatory standards has led to a great variety of anchorage types being 

used, some appropriate for shear force transfer between the foundation and framing and some 

not. Where anchorage details used for prescriptive designs are not coming from national 

standards such as IEBC (ICC, 2003a), or guidance developed by local authorities, it is necessary 

to ascertain whether 1) the connection appropriately addresses the primary objective of 

preventing movement between the foundation sill plate and foundation, and 2) the capacity is 

comparable to the capacity that would have been provided by a prescriptive connection. In 

making this evaluation, consideration should be given to earthquake loading in both horizontal 

directions and a complete load path, additionally the occurrence of cross-grain tension should not 

be allowed. 

Construction Considerations 

Addition of foundation anchorage in a crawl space with minimum required code vertical 

clearance is difficult due to very cramped conditions; work areas are often hard to get to, let 

alone getting tools and supplies and executing work. New temporary access openings and 

disconnection of HVAC ducting may occasionally be needed to provide access to work. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique, other than the use of 

proprietary connectors as part of the assemblage. 
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5.4.4 Enhance Cripple Wall 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses enhancement of existing cripple walls. After addition of 

anchor bolts, as discussed in Section 5.4.3, enhancement of cripple walls is the most effective 

rehabilitation measure for older one- and two-family detached dwellings. Past earthquakes have 

repeatedly shown cripple walls to be a significant weak link in the performance of W1 buildings. 

W1A and W2 buildings with this configuration are equally susceptible. This rehabilitation 

measure is almost always done in conjunction with providing anchorage to the existing 

foundation (Section 5.4.3). 

Description of the Rehabilitation Technique 

This rehabilitation measure involves addition of wood structural panel shear wall sheathing to 

existing cripple walls and development of a load path into and out of the walls. The objective is 

to eliminate in-plane shear failure of the cripple walls, often resulting in the building falling off 

of the cripple walls and foundation. 

Prescriptive provisions for rehabilitation of cripple walls can be found in the International 

Existing Building Code – IEBC -- (ICC, 2003b). These were developed from similar or identical 

provisions in the GSREB (ICBO, 2001) and the UCBC (ICBO, 1997b). An extensive 

commentary to the GSREB Chapter 3 provisions has been developed by SEAOC Existing 

Buildings Committee (ICC, 2005). Some organizations have developed local adaptations of these 

provisions (ABAG, 2005). The objective of these prescriptive provisions is reduction of 

earthquake hazard; they are intended to provide a reasonable level of improvement for the 

majority of buildings within their scoping limitations. W1 buildings with unusual configurations, 

site slopes greater than one vertical in ten horizontal, cripple walls taller than 4 feet, or higher 

performance objectives should be addressed with an engineered design. An engineered design is 

recommended for all W1A and W2 buildings because of higher loads and potential configuration 

issues. 

The prescriptive provisions address: 

Shear transfer between floor framing and the cripple wall top plate 

Shear wall sheathing and fastening 

Anchorage of the foundation sill plate to the foundation (Section 5.4.3) 

Figure 5.4.4-1 illustrates common cripple wall enhancement. The top of wall detail shows angle 

clips to a continuous rim joist or blocking. It is assumed that both the floor sheathing and sole 

plate above are nailed to the rim joist or blocking. If not, shear transfer per Figure 5.4.1-7 should 

be provided. 

Where cripple walls are 14 inches tall or less, wood structural panel sheathing may no longer 

provide reliable bracing of the studs, and splitting of the studs becomes a more significant 

concern. For this configuration, use of solid blocking between studs is recommended in lieu of 

sheathing. 
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Figure 5.4.4-1: Cripple Wall Enhancement 

An engineered design of cripple wall bracing would be anticipated to use very similar detailing, 

although additional fastening to further complete the load path may be desirable. 

Design and Detailing Considerations 

Research basis: Research into prescriptive methods for strengthening of cripple walls was 

conducted by Chai, Hutchinson & Vukazich (2002). 

Bracing material vulnerability: Cripple walls have been seen in analytical studies and past 

earthquakes to often be subjected to much higher drifts than the occupied stories above. Wood 

structural panel sheathing is the preferred bracing material for cripple walls in order to 

accommodate required drifts without significant loss of capacity. Although still permitted for 

shear walls in new construction, stucco has not consistently provided adequate bracing of cripple 

walls. Often fasteners between the stucco and framing have withdrawn, resulting in damage and 
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collapse. As a result, rehabilitation is encouraged for cripple walls not braced by either wood 

structural panel or diagonal lumber sheathing. 

Horizontal force distribution: Where rehabilitation will be limited to the cripple walls and 

anchorage to the foundation, it is common to make simplifying assumptions regarding force 

distribution to the cripple walls. For small buildings, forces generated at and above the lowest 

framed floor are distributed by tributary area to the perimeter foundations. For larger buildings, 

force may also be distributed to interior cripple walls based on tributary area. Where buildings 

have had additions, cripple wall bracing may be needed on the foundation separating original and 

addition construction. 

Where crawl spaces extend under framed decks and porches, it is necessary to provide cripple 

wall bracing at the perimeter of the enclosed building, as well as at the perimeter of the framed 

deck or porch. With this configuration it is sometimes necessary to alter the bracing approach to 

allow continued under-deck access. Other bracing approaches should have load-deflection 

behavior similar to the rest of the cripple walls, or the system should be evaluated considering 

the differences is behavior. Occasionally perimeter foundations are not complete between the 

enclosed dwelling and the deck or porch. The simplest solution is often to complete the 

foundation and add braced cripple walls. 

Overturning anchorage: Tie-down anchors are not required by the IEBC provisions. This is 

primarily because the low unit shears in the sheathing (controlled by 15/32 sheathing and 8d 

common at 4” nailing) and a maximum wall height of four feet limit the overturning forces that 

are generated. Testing by Chai, Hutchinson & Vukazich (2002) indicates that good cripple wall 

behavior (strength, stiffness and energy dissipation) can occur with this construction. If the 

bracing unit shear capacity is increased or if the height of the cripple walls are increased, 

overturning anchorage may be required. See Sections 5.4.1 and 6.4.4 for discussion of 

overturning anchorage. 

Ventilation and access: Existing access openings, ventilation openings and flood vents should 

not be reduced and, if possible, should be increased to meet code requirements during cripple 

wall bracing. 

Construction Considerations 

Moisture exposure: Elevated moisture can sometimes occur at cripple wall construction. Possible 

moisture sources include seasonal rain coming through cracks in the wall finish and high relative 

humidity at the building location. Decay in the existing cripple wall framing is a good indication 

that the rehabilitation work may also have a potential for decay. Where decay exists in existing 

framing, it should be repaired. Where no specific source of water can be identified and stopped, 

it is recommended that both replacement framing and new construction use preservative treated 

wood products and corrosion resistant fasteners and connectors. See Section 5.4.1 for further 

discussion. 

Ventilation of stud bays: Where cripple wall studs are being sheathed on the interior face, it is 

recommended that ventilation holes be provided near the top and bottom of each stud bay to 

allow air circulation. Ventilation holes of 1-1/2 to 2 inches in diameter with centerline no closer 
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than three inches to the panel edge will generally not reduce the effectiveness of the cripple wall 

bracing. 

Variations in existing framing details: It is common to find variations in the framing details at 

the top of the cripple walls. The variations come from initial construction, repairs, and additions. 

Modification to typical details is often needed to address these conditions. Care should be taken 

that these modifications address the basic objective of transferring in-plane forces into the top of 

the cripple wall and providing capacity approximately equal to the detail being replaced. 

Access: Access openings and under-floor clearance are likely to control the size of wood 

structural panel sheet that can practically be placed. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique, other than the use of 

proprietary connectors as part of the assemblage. 

5.4.5 Rehabilitate Hillside Home 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses seismic vulnerabilities associated with hillside buildings. 

Buildings constructed on sites sloping downward from street level will often have cripple walls 

or skirt walls of widely varying heights around the building perimeter between grade and the 

lowest framed floor. The variation in height leads to widely varying shear wall stiffness. Seismic 

forces away from the hill can lead to the floor diaphragm pulling away from the uphill 

foundation (Figure 5.4.5-1A). Seismic forces across the hill can result in torsion due to stiff 

support on the uphill side and flexible support on the downhill side, also pulling the floor away 

from the uphill foundation and damaging stepped or sloped side cripple walls (Figure 5.4.5-1B). 

Similar behavior can result when steel rod bracing rather than cripple walls provide bracing 

between floor and grade. Collapse of hillside homes in the Northridge earthquake was attributed 

to this behavior. Information on damage from the Northridge earthquake and hillside building 

behavior can be found in City of Los Angeles & SEAOSC (1996), EERI (1996), von Winterfeldt 

et al. (2000) and Cobeen, Russell and Dolan (2004). 

Description of the Rehabilitation Technique 

The primary objective of this rehabilitation technique is to address hillside buildings that are 

vulnerable due to inadequate or missing bracing between the lowest framed floor and grade. A 

primary resource for this technique is voluntary rehabilitation provisions developed by the City 

of Los Angeles and included in the City of Los Angeles Building Code (City of Los Angeles, 

2002). The objective of these provisions is to reduce the risk of death or injury. The provisions 

are indicated to be applicable to buildings constructed on a hillside slope in excess of one vertical 

to three horizontal. The rehabilitation measures described, however, may not be applicable to all 

W1 buildings constructed on this slope. 

The basic elements of the City of Los Angeles voluntary provisions include: 
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Figure 5.4.5-1: Hillside Home Response to Seismic Forces 

Adapted from Von Winterfeldt, Roselund and Kitsuse (2000) 
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“Primary anchors” (designed for tributary seismic load) tying the floor diaphragm to the 

uphill foundation in line with each foundation extending in the downhill direction 

“Primary anchors” where interior shear walls occur in contact with the base level 

diaphragm 

“Secondary anchors” to the uphill foundation at a spacing not exceeding four feet 

Foundation load path at primary anchors (or addition of tie-beam extending downhill 

from anchorage location) 

Drift limits for tall downhill walls 

Alternates to primary anchors include wood shear walls, steel braced frames, and rod 

bracing, all within specific limitations 

The primary focus of this rehabilitation technique is providing direct tension anchorage from 

floor diaphragms to uphill foundations or foundation walls, as shown in Figure 5.4.5-2. This 

anchorage prohibits separation of the floor diaphragm from the uphill foundation or foundation 

wall, whether from direct tension or rotation. In doing so, the lateral and vertical load paths at the 

uphill foundation are maintained. The provisions require engineering evaluation and design. 

Figure 5.4.5-2 illustrates a primary anchor at the exterior wall, in line with the stepped 

foundation, a primary anchor interior with a concrete tie-beam added in line, and a secondary 

anchor between the two, with no requirements for load path beyond anchorage to the uphill 

foundation. 

To date, these are the only published provisions for addressing vulnerable hillside buildings. 

Further work is needed to identify which of the many possible hillside building configurations 

are vulnerable. At this time, there are no provisions addressing hillside buildings on pole or pier 

foundations where connection to the uphill foundation is not possible. 

Damage observed following the Northridge earthquake also raised questions about the 

performance of stepped woodframe cripple walls, common on the sides of hillside buildings. It 

was suggested that seismic forces might be concentrating in the shortest uphill step of the 

woodframe walls, causing overstress and progressive failure. City of Los Angeles provisions 

require that the concentration of forces be considered in stepped cripple wall analysis. Testing of 

stepped cripple walls by the CUREE-Caltech Woodframe Project (Chai, Hutchinson and 

Vukazich, 2002) did not observe concentrations of seismic force, but instead saw well distributed 

forces and good performance. No explanations are currently available for the contrast between 

performance in testing and observed Northridge earthquake behavior. 

Design Considerations 

Research basis: Limited testing of the load-deflection behavior of tie-down devices used for 

diaphragm anchorage to uphill foundations has been conducted by Xiao and Xie (2002). See 

Cobeen, Russell and Dolan (2004) for discussion of the use and limitations of this information. 
Techniques for the Seismic Rehabilitation of

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Figure 5.4.5-2: Anchorage of Floor Diaphragm Framing 

to Uphill Foundation in a Hillside Dwelling 

The configuration tested is similar to Figure 5.4.5-3A. Figure 5.4.5-3B is another commonly 

used configuration. Care has to be taken to make the steel angle stiff enough to protect the 

framing connection to the uphill foundation. 

Alternate bracing approaches: Steel concentric braced frames have sometimes been used in lieu 

of primary anchors at exterior stepped foundation walls (the right hand end wall in Figure 5.4.5-

2). When this approach has been used, there is often only a single diagonal brace member at each 

foundation line, acting in tension for seismic loads towards the hill and in compression for 

seismic loads away from the hill. This does not conform to code requirements for braced frame 

design in which a balance of tension and compression resistance is required. If this approach is 

taken, conservatism in estimating brace and anchorage forces is recommended, to avoid 

premature failure and compensate for limited energy dissipation capacity. 
Techniques for the Seismic Rehabilitation of

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Figure 5.4.5-3: Connections for Anchorage to Uphill Foundation 

Detailing Considerations 

The objective of anchoring to the uphill foundation is to protect the ledger or foundation sill plate 

connection to foundation or foundation wall. These connections can experience brittle crossgrain tension failure at very small deflections. As a result, a very stiff primary or secondary 

anchor connection is needed to mitigate this failure. Stiff, direct axial connections should be 

favored over connections that allow movement; for example, the direct connection in Figure 

5.4.5-3A would provide better protection against damage, while Figure 5.4.5-3B might flex to 

result in damage to the foundation sill connection but still prevent collapse. Testing has not been 

performed to determine what level of deformation is acceptable for the varying details that can 

occur at the uphill foundation. 

Cost and Disruption Considerations 

Because the majority of the work is intended to be in the crawl-space area under the dwelling, 

little disruption is generally caused by this rehabilitation work. 

Construction Considerations 

Construction on steep hillsides can be very difficult. In the extreme case, chemical grouting to 

stabilize loose soils may be required to keep the hillside from deteriorating during construction. 

At the end of construction, care should be taken to remove all soil that is in contact with wood 

framing. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique, other than the use of 

proprietary connectors as part of the assemblage. 
Techniques for the Seismic Rehabilitation of

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5-48 

5.4.6 Rehabilitate Chimney 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses inadequate component detailing associated with 

unreinforced and unbraced masonry chimneys. Damage to masonry chimneys has occurred in 

virtually every moderate to major United States earthquake. A falling hazard can be created if 

portions of the chimney break free. 

Description of the Rehabilitation Technique 

Techniques for mitigating the hazards posed by unreinforced and unbraced chimneys include: 

Removal of the chimney and fireplace 

Removal of the chimney and replacement with light-framing 

Filling of the chimney 

Anchorage of the chimney to the building 

Complete removal of the masonry chimney and fireplace is the only method that will ensure 

elimination of the potential for damage or falling hazard. The chimney and fireplace can be 

removed without replacement or with replacement by well-anchored light-framing surrounding a 

factory-built fireplace and flue. All other rehabilitation measures mitigate rather than removing 

hazards. 

Recommendations for removal of the masonry chimney and replacement with light-framing are 

published by the City of Los Angeles (2000) and California OES and FEMA (OES and FEMA, 

2000). The transition to light-frame construction is shown to occur either at the top of the firebox 

or at a specified minimum dimension below the roof level. The farther down the chimney is 

removed, the more areas of potential damage are eliminated. A concrete bond beam is provided 

at the top of the remaining masonry. The bond beam is doweled into the existing masonry to 

remain and allows cast-in anchors for attachment of the light-framing above. Attention to 

maintaining required clearances to combustible materials is important at the transition and above. 

Anchorage of the flue is provided per manufacturer installation instructions. Anchorage of the 

light-frame enclosure to the building at floor, ceiling and roof levels is required. The OES and 

FEMA publication also illustrates replacement of an unreinforced masonry chimney with a codeconforming reinforced masonry chimney. The transition between existing and new construction 

should be carefully evaluated if this rehabilitation approach is chosen. 

Figure 5.4.6-1 illustrates a possible scheme for filling in a vulnerable chimney with reinforced 

concrete. Reinforcing is placed in the chimney down to the smoke chamber. Most fireplace 

geometries will make it impractical to extend the reinforcing down further. Ties and spacers are 

recommended to hold the reinforcing at adequate clearances off of the flue wall so that bond is 

adequate to develop the reinforcing. Wheel-type spacers, sometimes put on tie or spiral 

reinforcing in drilled-pier foundations, could help with placement. Figure 5.4.6-1 shows the 

concrete extending to the damper location. Where possible, reinforcing and filling the fire-box 

would improve the strength and continuity of the infill. Anchorage of the chimney to floor, roof 

and ceiling levels needs to be provided in conjunction with chimney infilling. Filling the 

chimney will reduce the falling hazard of an unreinforced chimney, by providing strength and 
Techniques for the Seismic Rehabilitation of

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5-49 

stiffness continuity at the commonly seen weak points (roof line and transitions in width). 

Filling the chimney may reduce, but is unlikely to eliminate damage. This rehabilitation measure 

is most often used for buildings of historical significance where there is a strong desire to 

maintain the current appearance. In some cases the height of very tall chimneys are reduced 

prior to filling with concrete. Use on chimneys already in poor condition due to deterioration or 

foundation movement is not recommended. Placement of grout between the flue liner and 

masonry is also recommended where this grout is completely missing or has significant gaps. 

Figure 5.4.6-1: Infill and Bracing of Masonry Chimney 
Techniques for the Seismic Rehabilitation of

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5-50 

Figure 5.4.6-2 illustrates anchorage of an exterior masonry chimney to floor, roof and ceiling 

framing. This detail is an adaptation of prescriptive information for new construction in the IBC 

(ICC, 2003a) and the Masonry Fireplace and Chimney Handbook (Amrhein, 1995). The steel 

strap is intended to keep the chimney from falling away from the building. In order to do this, the 

strap must be anchored into existing floor and roof framing with a capacity and load path 

adequate to resist forces from the chimney. Anchorage to wall studs or a single framing member 

will not accomplish this. It is often difficult and disruptive to anchor far enough into the building 

to develop required capacity. Figure 5.4.6-2 is intended for small to medium size chimneys 

common in single-family residences. Large and irregularly configured chimneys require 

additional consideration. 

Design Considerations 

Research basis: No research applicable to these rehabilitation techniques has been identified. 

Earthquake reconnaissance reports provide a limited record of earthquake performance of 

rehabilitation techniques. 

Cautions: Some in the earthquake engineering community recommend against rehabilitation 

measures involving unreinforced masonry chimney anchorage to light-frame buildings on the 

basis that the anchorage is unlikely to eliminate earthquake damage. Indeed, damage and 

occasionally partial collapse of anchored chimneys have been seen in past earthquakes. The 

inherent difference in stiffness between masonry chimneys and fireplaces and light-frame 

construction is a likely contributor, along with widely varying adequacy of anchorage detailing 

and installation. The potential hazard posed by an unreinforced and/or unanchored chimney and 

the ability to reduce the hazard using one or more rehabilitation techniques need to be weighed 

for each building under consideration. Other practical measures to reduce life-safety threats due 

to unreinforced chimneys include limiting activities (interior as well as exterior) in the 

immediate vicinity of the chimney and fireplace and placing wood structural panel sheets on 

ceiling rafters alongside the chimney to slow down any portions falling to the interior 

(ABAG, 2005). 

Variations in existing chimney conditions: Either careful evaluation of the existing chimney 

construction or worst-case assumptions regarding construction are suggested. Even when 

chimneys would have been required by buildings codes to be grouted and reinforced, it is 

common to find chimneys ungrouted, poorly grouted and unreinforced. 

Foundations: In areas of poor soils, the weight of the chimney and firebox can result in higher 

settlement, and sometimes differential settlement, leading to leaning. Foundation problems need 

to be resolved before other rehabilitation measures are considered. 

Detailing Considerations 

Anchorage of a strap or other tie to an existing masonry chimney should be avoided where 

possible and otherwise approached with caution. Expansion anchors cause splitting tensile 

stresses that can result in cracking of the masonry. Adhesive anchors change properties under 

elevated temperatures that might be experienced during use of the chimney. 
Techniques for the Seismic Rehabilitation of

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Figure 5.4.6-2A: Bracing of Masonry Chimney 
Techniques for the Seismic Rehabilitation of

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Figure 5.4.6-2B: Bracing of Masonry Chimney 

Cost, Disruption and Construction and Construction Considerations 

Any penetrations of the building exterior walls or roof need to be properly detailed for water 

resistance. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique. 

5.5 References 

ABAG, 2005, Association of Bay Area Governments website: www.abag.org. 

ACI, 2005, Building Code Requirements for Reinforced Concrete and Commentary, ACI 318, 

American Concrete Institute, Farmington Hills, MI. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 5 – Type W1: Wood Light Frames

5-53 

AF&PA, 2003, Plank and Beam Framing for Residential Buildings, American Forest & Paper 

Association, Washington, D.C. 

AF&PA, 2005a, National Design Specification for Wood Construction, ASD/LRFD, American 

Forest & Paper Association, Washington, D.C. 

AF&PA, 2005b, Special Design Provisions for Wind and Seismic, ASD/LRFD, American Forest 

& Paper Association, Washington, D.C. 

Amrhein, J., 1995, Residential Masonry Fireplace and Chimney Handbook, Second Edition, 

Masonry Institute of America, Torrance, CA. 

APA, 1999a, Wood Structural Panel Shear Walls (Research Report 154), APA The Engineered 

Wood Association, Tacoma, WA. 

APA, 1999b, Research Report 158, Preliminary Testing of Wood Structural Panel Shear walls 

Under Cyclic (Reversed) Loading, APA The Engineered Wood Association, Tacoma, WA. 

APA, 2003, Shear Wall Lumber Framing: Double 2x’s vs. Single 3x’s at Adjoining Panel Edges

(Report T2003-22), APA The Engineered Wood Association, Tacoma, WA. 

ATC, 1976, A Methodology for Seismic Design and Construction of Single-Family Dwellings, 

ATC 4, Applied Technology Council, Redwood City, CA. 

ATC, 1995, Cyclic Testing of Narrow Plywood Shear Walls, ATC R-1, Applied Technology 

Council, Redwood City, CA. 

Chai, Y.H., T.C. Hutchinson and S. M. Vukazich, 2002, Seismic Behavior of Level and Stepped 

Cripple Walls, CUREE Publication No. W-17, Consortium of Universities for Research in 

Earthquake Engineering, Richmond, CA. 

City of Los Angeles, 2002, “Voluntary Earthquake Hazard Reduction in Existing Hillside 

Buildings (Chapter 94),” City of Los Angeles Building Code, Los Angeles, CA. 

City of Los Angeles, 2000, Reconstruction and Replacement of Earthquake Damage Chimneys

(P/BC 2001-70), City of Los Angeles Department of Building and Safety, Los Angeles, CA. 

City of Los Angeles & SEAOSC, 1996, Findings and Recommendations of the Hillside 

Buildings Subcommittee of the City of Los Angeles Department of Building and Safety & 

Structural Engineers Association of Southern California Task Force on Evaluating Damage 

From the Northridge Earthquake, Final Report, City of Los Angeles, CA. 

Cobeen, K., 2004, “Recent Developments in the Seismic Design and Construction of Woodframe 

Buildings,” Earthquake Engineering From Engineering Seismology to Performance Based 

Engineering, CRC Press, Boca Raton, FL. 
Techniques for the Seismic Rehabilitation of

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Cobeen, K., J. Russell and J.D. Dolan, 2004, Recommendations for Earthquake Resistance in the 

Design and Construction of Woodframe Buildings, CUREE Publication No. W-30b, Consortium 

of Universities for Research in Earthquake Engineering, Richmond, CA. 

Dolan, J.D., 2003, “Wood Structures,” Earthquake Engineering Handbook,” CRC Press, Boca 

Raton, FL. 

EERI, 1996, “Supplement C to Volume II - Northridge Earthquake of January 17, 1994 

Reconnaissance Report Volume 2,” Earthquake Spectra, Earthquake Engineering Research 

Institute, Oakland, CA. 

FEMA, 2000, Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 

356, Federal Emergency Management Agency, Washington, D.C. 

FEMA, 2003, NEHRP Provisions for New Buildings and Other Structures, FEMA 450, Federal 

Emergency Management Agency, Washington, D.C. 

Gatto, K. and C.M. Uang, 2002, Cyclic Response of Woodframe Shearwalls: Loading Protocol 

and Rate of Loading Rate Effects, CUREE Publication No. W-13, Consortium of Universities for 

Research in Earthquake Engineering, Richmond, CA. 

Heine, C., 1997, Effect of Overturning Restraint on the Performance of Fully Sheathed and 

Perforated Timber Framed Shear Walls, Virginia Polytechnic Institute and State University, 

Blacksburg, VA. 

ICBO, 1997a, Uniform Building Code, International Conference of Building Officials, Whittier, 

CA. 

ICBO, 1997b, Uniform Code for Building Conservation, International Conference of Building 

Officials, Whittier, CA. 

ICBO, 2001, Guidelines for the Seismic Retrofit of Existing Buildings, International Conference 

of Building Officials, Whittier, CA. 

ICC, 2003a, International Building Code, International Code Conference, Country Club Hills, 

IL. 

ICC, 2003b, International Existing Building Code, International Code Conference, Country Club 

Hills, IL. 

ICC, 2005, 2003 International Existing Building Code (IEBC) Commentary, International Code 

Conference, Country Club Hills, IL. 

Isoda, H., B. Folz, and A. Filiatrault, 2002, Seismic Modeling of Woodframe Index Buildings, 

CUREE Publication No.W-12, Consortium of Universities for Research in Earthquake 

Engineering, Richmond, CA. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 5 – Type W1: Wood Light Frames

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Kolba, A., 2000, The Behavior of Wood Shear Walls Designed Using the Diekmann’s Method 

and Subjected to Static In-plane Loading, Marquette University, Milwaukee, WI. 

Mahaney, J. A. and B. E. Kehoe, 2002, Anchorage of Woodframe Buildings, Laboratory Testing 

Report, CUREE Publication No. W-14, Consortium of Universities for Research in Earthquake 

Engineering, Richmond, CA. 

Mosalam, K., C. Machado, K-U. Gliniorz, C. Naito, E. Kunkel and S. Mahin, 2002, Seismic 

Evaluation of an Asymmetric Three-Story Woodframe Building, CUREE Publication No. W-19, 

Consortium of Universities for Research in Earthquake Engineering, Richmond, CA. 

OES and FEMA, 2000, Guidelines to Strengthen and Retrofit Your Home Before the Next 

Earthquake, State of California Governor’s Office of Emergency Services, Sacramento, CA and 

Federal Emergency Management Agency, Washington, D.C. 

Pardoen, G., A. Waltman, R. Kazanjy, E. Freund, and C. Hamilton, 2003, Testing and Analysis 

of One-Story and Two-Story Walls Under Cyclic Loading, CUREE Publication No. W-25, 

CUREE, Richmond, CA. 

Salenikovich, A., 2000, The Racking Performance of Light-Frame Shear Walls, Virginia 

Polytechnic Institute and State University, Department of Wood Science and Forest Products, 

Blacksburg, VA. 

SEAOC (Structural Engineers Association of California), 1999, Recommended Lateral Force 

Requirements and Commentary, Seventh Edition, Structural Engineers Association of California, 

Sacramento, CA. 

von Winterfeldt, D., N. Roselund, and A. Kitsuse, 2000, Framing Earthquake Retrofitting 

Decisions: The Case of Hillside Homes in Los Angeles, PEER Report 2000-03, Pacific 

Engineering Research Center, Richmond, CA. 

Xiao Y. and L. Xie, 2002, Seismic Behavior of Base-Level Diaphragm Anchorage of Hillside 

Buildings, CUREE Publication No. W-24, Consortium of Universities for Research in 

Earthquake Engineering, Richmond, CA. 
Techniques for the Seismic Rehabilitation of Chapter 6 - Type W1A: Multistory, 

Existing Buildings: FEMA 547 Multi-Unit Residential Woodframes

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Chapter 6 - Building Type W1A: Multistory, Multi-Unit 

Residential Woodframes 

6.1 Description of the Model Building Type 

Building Type W1A is similar to Building Type W1 in use of woodframe wall, floor and roof 

construction, but includes large multi-family, multistory buildings. In W1A buildings, second 

and higher stories are almost exclusively residential use, while the first story can include any 

combination of parking, common areas, storage, and residential units. Post and beam framing 

often replaces bearing walls in non-residential areas. Multi-family residential buildings with 

commercial space at the first story are included in building Type W1A due to similar building 

characteristics. Lateral forces in W1A buildings are primarily resisted by wood diaphragms and 

shear walls. Figure 6.1-1 provides an illustration of this building type. 

Figure 6.1-1: Building Type W1A: Multistory Multi-Unit Residential Woodframes 

This chapter addresses W1A buildings where the first story walls are of woodframe construction. 

This includes both multistory woodframe buildings supported at grade and the multistory 

woodframe portion of buildings with concrete or masonry walls at one or more lower stories. 

The stories with concrete or masonry walls represent building types other than W1A, and they 

are addressed by other chapters in this document. 
Techniques for the Seismic Rehabilitation of Chapter 6 - Type W1A: Multistory, 

Existing Buildings: FEMA 547 Multi-Unit Residential Woodframes

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Variations in the W1A building type can include a combination of multi-family residential use 

and the hillside building configuration discussed in Section 5.4.5. For this combination, 

rehabilitation measures from this chapter and Section 5.4.5 are applicable. 

Design Practice 

W1A buildings including apartment and condominium buildings, residential hotels, motels, and 

residential use over commercial space are very common in the current building stock, with some 

dating back to the early 1900s or earlier. While many W1A buildings constructed in the 1980s 

and later will have had a partially or fully engineered design, the majority of older W1A

buildings will not. Case studies of California tuckunder buildings constructed in the 1970s 

(Schierle, 2001) indicate that a check of first-story walls for in-plane shear capacity was 

common, shear wall overturning was not considered, and bracing of upper stories commonly 

relied on prescriptive construction provisions. Steinbrugge, Bush and Johnson (1996) chronicled 

changes in California design practice of multi-family residential buildings since the 1960s. In 

some regions, these buildings are currently constructed using prescriptive codes. 

Walls 

Wall bracing materials include the same range discussed for W1 buildings. Checks of first floor 

shear capacity in California tuckunder apartment buildings led to the use of wood structural 

panel sheathing without overturning anchorage in some first story walls in the 1960s and 1970s. 

Cripple walls, also discussed with the W1 building type, are common in W1A buildings up until 

the 1950s. 

Floor and Roof Diaphragms 

Floor and roof diaphragms include the same materials as the W1 building type, however plank 

and beam systems are rare in W1A buildings. 

Foundations 

Foundation types and issues for W1 buildings are also applicable to W1A buildings. Of note, the 

gravity dead and live loads in W1A buildings can be significantly higher than in W1 buildings. 

Identification and Performance of Vulnerable Buildings 

Several W1A building vulnerable configurations have become prominent in literature and 

discussion because of collapses or near collapses of lowest woodframe stories in the Loma Prieta 

and Northridge earthquakes. While these vulnerabilities are important for the W1A building 

type, they are not the only deficiencies that require consideration. See Section 6.3 for a 

systematic discussion of seismic deficiencies. 

The discussion of these prominent vulnerable configurations requires a common understanding 

of terminology. In addition, a brief review is provided of documents that discuss performance, 

identification and rehabilitation provisions for vulnerable stories in W1A buildings. 

W1A buildings, regardless of design approach, gain much of their strength and stiffness from 

bracing and finish materials on exterior walls and interior walls between and within residential 

units. This is true whether or not these walls are identified as shear walls. Where residential use 
Techniques for the Seismic Rehabilitation of Chapter 6 - Type W1A: Multistory, 

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occurs in multiple stories, it is common for residential unit layouts to be similar at each story, 

providing substantially uniform story strength and stiffness. Where the lowest story includes uses 

such as parking, common areas, commercial use, etc., the amount of exterior and interior wall is 

reduced, often resulting in significantly reduced story strength and stiffness. At the same time, 

the lowest story experiences the highest earthquake demands. 

The terms weak story and soft story are used for this condition in which a story has less strength 

or stiffness than the story above. Concentration of deformation demand is understood to occur in 

a soft story. Inadequate strength and story failure may occur in a weak story. These would be 

identified as global strength and stiffness deficiencies for purposes of this chapter. Exact 

definitions of what constitutes a soft or weak story vary, as do opinions as to when soft and weak 

stories become vulnerable enough to recommend rehabilitation. Little research is available to 

assist in identifying when these configurations pose a hazard to life. 

Where parking occurs in all or a portion of the lowest woodframe story, significant openings in 

the exterior walls are generally provided in order to allow access to the parking. Often there is 

little or no interior wall in the parking area. The term tuckunder parking (named due to the 

parking being tucked under the residential units) is used for this type of building configuration. 

Tuckunder parking buildings with woodframe walls at the parking story will often have a soft 

story and a weak story. Occasionally, parking only exists in a very small portion of the building 

plan area, and it does not significantly affect the story. 

An open front building occurs when at any story level there is little or no bracing in one or more 

exterior walls. The term open front is a misnomer in that the open exterior wall can occur at any 

side of the building. Woodframe buildings are generally considered to have flexible diaphragms, 

and as a result bracing elements are generally provided at or near each edge of the diaphragm, 

most often at exterior walls. When an open front occurs, the diaphragm is required to transmit 

forces to other wall lines by rotation, creating torsional building behavior. This behavior is 

particularly critical when an exterior wall is provided at upper stories but discontinued in the first 

woodframe story, as this creates a significant discontinuity in the load path at the lowest story. 

Open front buildings often have tuckunder parking, but can also have commercial and other uses. 

Open front buildings will often but not always also have soft and weak stories at the open front 

story. Addition of vertical elements at the open front is the most direct rehabilitation approach to 

open front buildings. In buildings studied to date, capacities in the direction perpendicular to the 

open front have also been significantly lower that required by current codes and may also require 

rehabilitation. 

What the terms soft story, weak story, tuckunder building and open front building all have in 

common is that they are identifying buildings that have potentially vulnerable stories due to 

deficient global or local strength or stiffness. In most cases, the vulnerable story is the lowest 

woodframed story. 

Appendix Chapter 4 of the International Existing Building Code (IEBC) (ICC, 2003b) and 

Chapter 4 of the Guidelines for Structural Rehabilitation of Existing Buildings (GSREB) (ICBO, 

2001) contain identical provisions for hazard reduction in W1A buildings. These provisions 

identify a broad range of multistory woodframe buildings as vulnerable based on: 
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Open front conditions (defined by IEBC as diaphragm cantilever in excess of that 

permitted by the applicable building code) 

A weak wall line (defined by IEBC as story strength less than 80% of the strength of the 

story above), or 

A soft wall line (defined by IEBC as not meeting story drift limits) 

The IEBC provisions require evaluation and retrofit, including resisting elements from the 

diaphragm above the soft, weak, or open front story to the foundation-soil interface. Design is to 

be in accordance with the current building code except use of 75% of the code base shear is 

permitted. Specific rehabilitation measures are not detailed (with the exception of a prescriptive 

rehabilitation for limited building configurations); however, additional requirements for shear 

wall rehabilitation are included. The IEBC evaluation provisions create the challenge of 

calculating strength and stiffness for a variety of current and archaic finish materials not 

generally considered to be part of the lateral force-resisting system. Some guidance on strength 

and stiffness can be found in FEMA 356 (FEMA, 2000) and the AF&PA Wind and Seismic 

Supplement (AF&PA, 2005). Focus on the vulnerable first story may be lost in the calculation 

process. The IEBC also creates the challenge of identifying a wide range of buildings as 

potentially vulnerable, going well beyond open front and tuckunder configurations observed to 

be vulnerable to date. No guidance is given in judging relative hazard. If using IEBC Appendix 

Chapter 4, a commentary to the GSREB (ICC, 2005a) and ICC proposed changes (ICC, 2005b) 

are important additions. 

The City of San Jose has developed several documents that assist in identification of vulnerable 

W1A buildings. The Apartment Owner’s Guide to Earthquake Safety (Vukazich, 1998) uses a 

procedure based on ATC–21 rapid screening provisions in a broad approach to identifying 

vulnerable buildings and suggests shear wall enhancement and addition of steel moment frames 

as primary rehabilitation measures. Practical Solutions for Improving the Seismic Performance 

of Buildings with Tuckunder Parking (Lizundia and Holmes, 2000) illustrates rehabilitation 

techniques for three model building types, primarily using shear wall enhancement and steel 

moment frames. Rehabilitation measures address both life-safety and limited down time 

objectives. The focus of life safety measures is the first woodframed story. Work for limited 

down time objectives extends into upper stories. 

A joint task force of the City of Los Angeles Department of Building Safety and the Structural 

Engineers Association of Southern California prepared the report Wood Frame Construction 

Report and Recommendations (City of Los Angeles & SEAOSC, 1994), which contains a series 

of observations and recommendations for multi-family residential construction based on 

performance in the Northridge earthquake. Issues include 

Poor performance of gypsum wallboard and stucco bracing, attributed in part to high 

values given to these materials in past Los Angeles codes, 

Poor performance of plywood shear walls, attributed to core gaps (gaps in the center ply 

of three-ply plywood) and slender walls, 

Poor performance of tie-downs, attributed to design and installation problems, and 
Techniques for the Seismic Rehabilitation of Chapter 6 - Type W1A: Multistory, 

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 6-5 

Excessive drift at steel columns and excessive building rotation, attributed to lack of drift 

checks on steel columns used as lateral-force-resisting elements. 

Details and photos of observed damage are provided. 

Finally, the CUREE-Caltech Woodframe Project included testing and analytical studies of openfront buildings and retrofits, summarized in Topical Discussion J (Cobeen, Russell and Dolan, 

2004). One observation of note is that walls perpendicular to the open front suffered the greatest 

damage and degradation in testing and analysis, due to combined direct and torsional loading. 

Simultaneous earthquake loading in both horizontal directions should be evaluated in open-front 

buildings. Rehabilitation measures studied and recommended for use include: 

Steel moment frames (designed as special moment frames per building code requirements 

or at R = 1) at the open front in combination with enhancement of other first story walls, 

and 

A longitudinal wall near the building center of mass designed to carry the entire building 

base shear. 

The CUREE research found that soft first stories are very common in woodframe construction 

and do not necessarily create a hazard. 

Among these documents, there is currently no widely accepted definition of the point at which 

soft, weak and open-front stories become vulnerable to damage or constitute a life-safety hazard. 

The first story is the primary focus of evaluation and rehabilitation in most W1A buildings, and 

it is generally acceptable to reduce hazard through rehabilitation of the first story without 

improvement to upper stories. Steel moment frames and added or enhanced shear walls are the 

primary rehabilitation measures recognized in these documents. 

6.2 Seismic Response Characteristics 

Like the W1 buildings, the dynamic response of W1A buildings is short period, and inelastic 

behavior is primarily concentrated in the vertical wall elements rather than the diaphragms. The 

first woodframed story will generally drift significantly more than upper stories and experience 

higher damage as a result. Configurations with open fronts have been seen to respond with 

significant torsional behavior as well as weak story behavior. 

6.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

While similar in construction to the W1 building type, damage to W1A buildings has been more 

significant in areas of strong ground motion. Notably, the damage to finish and bracing materials 

and residual drift have been significant enough that re-occupancy of numerous buildings has not 

been permitted. Full and partial collapse of open front or tuckunder parking W1A buildings has 

occurred in recent earthquakes and resulted in loss of life in one building complex in the 

Northridge earthquake. The first story of these buildings was partially or completely occupied by 

parking; fewer and shorter bracing walls combined with archaic or heavily loaded bracing 

materials and rotational or torsional response contributed to vulnerability. Significant structural 

damage also occurred in W1A buildings having only residential units at the lowest story, as seen 
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in Schierle (2001) Case Study 10, a three story residential building constructed in the early 1960s 

and braced with stucco and plaster over gypsum lath. See below for general discussion and Table 

6.3-1 for a detailed compilation of common seismic deficiencies and rehabilitation techniques for 

Building Type W1A. 

Global Strength and Stiffness 

Global strength and stiffness are of particular concern in the first story of W1A buildings and 

have contributed significantly to damage in past earthquakes, sometimes accentuated by open 

fronts. Rehabilitation measures for global strength and stiffness include adding new vertical 

elements and enhancing existing elements. Common added elements are steel moment frames 

and added or enhanced shear walls. Steel braced frames may be added, but are not common in 

W1A buildings, since the brace would restrict access for parking or other uses. 

Configuration 

Although most common in W1 buildings, some W1A buildings have missing or inadequately 

braced cripple walls. See Chapter 5 for rehabilitation techniques. Where W1A buildings are of 

large plan area, it may be necessary to add interior cripple walls and new interior foundations. It 

is common to enhance or add cripple walls in W1 buildings without specifically accounting for 

overturning behavior in the stories above. Caution should be exercised in taking this approach 

with W1A buildings due to the larger size and weight. In addition, where uplift anchorage is 

being provided in stories above, the load path must be carried through the cripple wall to the 

foundation. 

Torsional irregularities due to open fronts are prevalent and of significant concern in W1A

buildings. Open fronts are often in the first story, and they combine with weak and soft story 

behavior. Where open fronts occur in tuckunder buildings, continued use of the first story 

parking often dictates that this deficiency be mitigated by the addition of steel moment frames. 

Wood shear walls and steel braced frames are alternate measures. It is important that walls 

perpendicular to the open front also be evaluated and enhanced, as these can be significantly 

deficient also. 

Load Path 

Adequate load path connection is a concern for W1A buildings, particularly so in first stories, 

which are likely to experience the majority of force and deformation demands. Many W1A

buildings constructed in California in the 1960s and 1970s used wood structural panel sheathing 

in the first story, but did not have overturning detailing. Testing suggests that significant 

reductions in shear wall strength and stiffness can occur when overturning detailing is not 

provided. Likewise, many W1A and W2 buildings are braced with diagonal lumber sheathing 

without overturning anchorage. Addition of overturning anchorage to these buildings could 

potentially greatly improve performance. 

Related to the overturning load path, in W1A buildings where upper story shear walls are 

discontinued in lower stories, beams and posts providing vertical support at shear wall ends are 

potentially vulnerable. Instances of rehabilitation of members supporting shear walls in W1, 

W1A and W2 buildings are very limited to date. This is because W1A building retrofits have 

focused on first story vulnerability, because earthquake damage to date has not shown this to be 
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Table 6.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for W1A Buil

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Dem

Global Strength 

Insufficient in-plane wall strength 

Wood structural panel 

shear wall [6.4.2] 

Steel braced frame 

[7.4.1] 

Steel moment frame 

[6.4.1] 

Enhance 

woodframe shear wall [6.4.2] 

Shear wall uplift 

anchorage and compression posts [6.4.4] 

Replace hea

roof finish wlight finish 

Global Stiffness 

Insufficient in-plane wall stiffness 

Wood structural panel 

shear wall [6.4.2] 

Steel braced frame 

[7.4.1] 

Steel moment frame 

[6.4.1] 

Enhance 

woodframe shear wall [6.4.2] 

Shear wall uplift 

anchorage and compression posts [6.4.4] 

Weak story, missing or weak cripple wall 

 Add woodframe cripple 

wall 

Add continuous 

foundation and foundation wall 

Enhance 

woodframe cripple wall [5.4.4] 

Configuration 

Open front Wood structural panel 

shear wall [6.4.2] Proprietary wall Steel moment frame [6.4.1] 

Enhance 

woodframe shear walls perpendicular to open front [6.4.2] 

Inadequate shear anchorage to foundation 

Anchorage to 

foundation [5.4.3] 

Inadequate detailing for shear wall overturning 

Enhance framing 

supporting shear wall [6.4.3] 

Shear wall uplift 

anchors and compression posts [6.4.4] 

Load Path 

Inadequate shear transfer in wood framing 

Enhance load path 

for shear [5.4.1], [6.4.5] 
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Table 6.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for W1A Buil

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Dem

Load Path (continued) 

Inadequate collectors to shear walls 

Enhance existing collector 

Add collector 

[6.4.5], [7.4.2] 

Component Detailing 

Unreinforced & unbraced chimney 

Infill chimney 

[5.4.6] 

Brace chimney 

[5.4.6] 

Reduce 

unsupported chimney he[5.4.6] 

Inadequate in-plane strength and/or stiffness 

Enhanced diaphragm [22.2.1] 

Replace hea

roof finish wlight finish 

Inadequate chord capacity 

Enhance chord 

members and connections [22.2.2] 

Excessive stresses at openings and irregularities 

Enhance diaphragm 

detailing 

Diaphragms 

Re-entrant corners Enhance diaphragm 

detailing 

Foundations See Chapter 23 [ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 
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a critical weakness in woodframe construction and because rehabilitation of these supports can 

be difficult and expensive. 

Shear transfer into and out of shear walls and other vertical elements must be adequate in order 

for the vertical element to fully contribute to building performance. While systematic evaluation 

may identify insufficient shear transfer at any story, shear transfer in first story walls is of 

particular concern due to reductions in the amount of shear wall and increases in unit loads. As in 

the W1 building, adequate anchorage to the foundation is a high priority rehabilitation measure. 

Component Detailing 

Damage to unreinforced masonry chimneys has occurred in practically every earthquake to date. 

Approaches to rehabilitation include bracing, reducing height, infilling or removing. See 

Chapter 5. 

Diaphragm Deficiencies 

Although diaphragm deficiencies have not been seen as a significant contributor to damage to 

date, systematic evaluation can identify this as a deficiency. Rehabilitation measures include 

enhancing existing diaphragms through added fastening, blocking, and overlaying. Detailing can 

also be added at openings and re-entrant corners. See Chapter 22. 

6.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

6.4.1 Add Steel Moment Frame 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses insufficient global or local strength or stiffness through 

the addition of steel moment frames. This rehabilitation technique is particularly beneficial in 

buildings with open fronts due to tuckunder parking, because the use of moment frames permits 

continued use of parking stalls. It is similarly beneficial for other buildings where continued use 

does not allow the addition of shear walls. 

Description of the Rehabilitation Technique 

This rehabilitation technique most commonly involves the addition of steel moment frames 

immediately adjacent to existing beams and columns, at or near a first story open front. Moment 

frames are less commonly added in other locations and in stories above the first story. 

Figure 6.4.1-1A illustrates an elevation of a typical single-bent steel moment frame added 

immediately in front of existing beams and columns. Such frames might be added at every 

second or third framing bay across the building front. Moment frames can be brought to the job 

site in a complete beam plus two-column bent or in two L-shaped pieces with a field-bolted 

splice at beam mid-span. The use of two L-shaped pieces allows the critical beam to column 

connections to be welded in the fabrication shop with better access and quality control. The 

height required to tilt the frame into place is the factor most commonly governing whether 

frames are fabricated in one or two pieces. A new foundation will often be required to support 

the moment frame. This can either be an isolated footing at each end or a continuous footing. 
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Figure 6.4.1-1A: Elevation of Steel Moment Frame in W1A Building 

Footing placement will generally require the shoring of the upper stories and full or partial 

removal of existing footings. Transfer of earthquake load from the diaphragm above to the steel 

moment frame will commonly involve a collector that runs the full length of the open front and a 

series of connections from the collector to the steel moment frame. 

Figures 6.4.1-1B, 6.4.1-1C and 6.4.1-1D illustrate possible connections. See discussion of 

collectors and shear transfer in the Design Considerations section. A number of detailing 

considerations discussed in Section 5.4.1 are applicable to frame connection to the existing wood 

building. In particular, detailing must accommodate shrinkage and possible swelling of wood, 

and alternate fasteners to existing sheathing may be needed. 

This rehabilitation measure is not intended to address systems of steel columns cantilevered from 

the foundation without moment connections to a beam at the top. This cantilevered column 

system should be used with caution due to the difficulty of quantifying and limiting the many 

potential sources of rotation and deflection and to inadequate knowledge of post-elastic system 

behavior. 
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Figure 6.4.1-1B: Shear Transfer Between Moment Frame Beam and Diaphragm 

Design Considerations 

Research basis: Research specifically addressing steel moment frames in woodframe buildings 

includes: Seismic Evaluation of an Asymmetric Three-Story Woodframe Building (Mosalam et 

al., 2002) and Improving Loss Estimation for Woodframe Buildings (Porter et al., 2002). Results 

from these studies are also discussed in Cobeen, Russell, and Dolan (2004). 

Moment frame design criteria: Chapter 8 of this document addresses steel moment frame 

rehabilitation in buildings where steel moment frames are the primary lateral force-resisting 

system. In contrast, when used for rehabilitation of W1A buildings, steel moment frames will 

generally only be used in one story and along one building line. The response modification factor 

of the woodframe building above makes use of either an ordinary or intermediate moment frame 

a logical choice for the first story of a multistory W1A building. Limitations addressing use in 

light-frame buildings have been in a state of flux. The most current seismic design provisions, 

ASCE 7-05 (ASCE, 2005) and AISC Seismic (2005), permit: 

Single story ordinary moment frames (OMF) for new buildings in Seismic Design 

Category (SDC) D and E, to a height of 65 feet, provided dead load tributary to the roof 

does not exceed 20 psf and tributary wall dead load does not exceed 20 psf 
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Figure 6.4.1-1C: Shear Transfer from Moment Frame Beam to Collector 
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Figure 6.4.1-1D: Shear Transfer from Moment Frame Beam to Collector 
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OMFs for new buildings in SDC D and E in light frame construction up to a height of 

35 feet, with roof and floor dead load to tributary to the frame not exceeding 35 psf and 

wall dead load tributary to the frame not exceeding 20 psf 

Intermediate moment frames (IMFs) in SDC D up to a height of 35 feet 

IMFs in SDC E up to a height of 35 feet with tributary floor and roof dead load not 

exceeding 35 psf and tributary wall dead load not exceeding 20 psf 

A three-story W1A building will generally just meet the height and weight limits to allow use of 

an OMF. This allows the choice of OMF, IMF or special moment frame (SMF). While SMFs are 

always acceptable, the response modification factor must not be taken as greater than for the 

lateral force-resisting system above (typically wood shear wall), and use of pre-qualified welded 

joints may require use of steel beam and column sizes larger than acceptable. 

Because the limitations for use of moment frames in light-frame construction have been in a state 

of flux, a number of organizations and jurisdictions have developed local guidance for design 

and rehabilitation. Among these are: 

Provisions used by the City of Santa Monica with the response modification factor set as 

one (used in CUREE Woodframe Project research) 

Draft guidelines by the SEAOSC Steel Ad Hoc Committee (SEAOC, 2002) addressing 

up to two-story buildings and recommended reduced drift and quality assurance measures 

Draft procedures by the ICC Peninsula Chapter (2004) addressing design procedures and 

quality assurance measures 

The need for these guidelines in addition to the latest design standards requires review. One of 

the recommendations made in the guidelines is that moment frame drift be limited to less than 

required by code in recognition of the lesser ductility of the connections. 

Shear transfer and collector detailing: Provision of adequate strength and stiffness for shear 

transfer from the building wood framing into the steel moment frame is key to improved building 

performance. Where the shear transfer detail allows significant slip, undesirable building 

deflection will occur. This was observed to be a significant issue in the CUREE-Caltech 

Woodframe Project testing of moment frames (Mosalam et al., 2002) (Cobeen, Russell, and 

Dolan, 2004). Figure 6.4.1-1C3 illustrates the shear transfer detail used in a simplified moment 

frame. The shear transfer was designed using tributary seismic forces, but without consideration 

of overstrength or the force that could be developed by the system. The connection used wood 

filler pieces and through bolts, and it was intended to reflect common design practice. Excessive 

slip developed between the wood beam and the filler. At peak capacity, the slip accounted for 

40% of the total system drift, and the bolts cut long slots into the beam and fillers. 

Figure 6.4.1-1C1, based on a Rutherford & Chekene detail for the CUREE testing, shows a shear 

transfer detail used for the special moment frame tested in the shake table tuckunder building. 

Two significant differences occur in this detail. First, the shear transfer connection was designed 

to develop the capacity of the diaphragm above; and second, the wood-to-wood connection was 

replaced with a lower-slip wood-to-steel connection. Although the forces seen by the frame were 
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moderate, the connection resulted in less slip, suggesting that better control of building drift 

would result. 

Although the first (Figure 6.4.1-1C3) connection could be improved by design using 

overstrength forces, the second (Figure 6.4.1-1C1) connection approach is recommended. It is 

further recommended that the approach of using overstrength forces and limiting slip be applied 

to other shear force transfer connections, including those shown in Figures 6.4.1-1B and 6.4.1-

1D. 

Steel moment frame design and detailing: Design and detailing of steel moment frames used in 

rehabilitation should be in accordance with the most recent edition of IBC and AISC provisions. 

Moment frame column bases: Columns in the CUREE testing used base plate details that are 

commonly considered to provide pinned conditions. This was done to minimize the moment 

demand put on the foundation, keeping foundation rehabilitation to a minimum, and to keep 

inelastic behavior in places where performance could be more easily predicted. The column base 

behavior during testing corresponded well to the assumed near-pinned condition, with little or no 

deterioration of the base plate connection seen. The use of pinned column base detailing is 

recommended. 

Lateral bracing of columns: Bracing at the beam top and bottom flange elevations is required at 

the moment frame columns. For the CUREE testing, steel angle braces were provided between 

bottom flange continuity plates and wood floor joists. 

Lateral bracing of beam flanges: Continuous bracing of the moment frame beam top flange is 

generally easily accomplished by the addition of a bolted nailer and connection to new or added 

framing, as shown in Figure 6.4.1-1B. Provisions for SMFs may require the bracing of the beam 

bottom flange just beyond the plastic hinge zone if bracing was included in prequalification 

testing. Bracing forces that are easily accommodated in steel construction can be more of a 

significant detailing issue in woodframe rehabilitation, depending on how far the bracing force is 

developed into the wood framing system. As a minimum, the brace member and its connection at 

either end should develop required forces. 

Addition of moment frames in upper stories: Where moment frames are added in upper stories, 

provision for a load path to the foundation is required. The load path should be designed using 

the forces that can develop in the frame using overstrength, force-controlled action, or target 

displacement approaches. 

Detailing Considerations 

Accommodation of wood shrinkage: Figure 6.4.1-1 details use vertical slotted holes in the steel 

side plates to accommodate wood shrinkage (or expansion) and possible vertical movement due 

to deformation of the steel moment frame beam. This approach should be used at any location 

where steel side plates are placed against wood framing, provided the connection is only 

intended to transfer horizontal forces. See Section 5.4.1 for further discussion of wood shrinkage 

issues. 
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Cost and Disruption Considerations 

It is very unlikely for the addition of a steel moment frame to be the least expensive or quickest 

way to rehabilitate for global or local strength or stiffness. The steel moment frame requires the 

involvement of multiple building trades: fabrication in a steel fabrication shop and site assembly 

by steel workers, in addition to foundation and framing work at the job site. The addition or 

enhancement of shear walls will be less expensive. In buildings were the addition of shear walls 

is not acceptable, however, the addition of a steel moment frame does provide a reasonable and 

common rehabilitation approach. 

Construction Considerations 

Plumbing, HVAC or electrical lines may be running in the floor framing in the vicinity of steel 

moment frame locations. Either accommodation in the structural design or relocation of utilities 

may be necessary. Job site welding of steel members requires adequate access and special 

ventilation measures in enclosed buildings. Welding of steel members in the vicinity of 

woodframe construction can be a significant fire hazard and should only be undertaken by 

experienced welders and only when absolutely necessary. Smoldering droppings from on-site 

welding and cutting have repeatedly caused structure fires. Welding should always be done by 

certified welders using approved welding techniques in compliance with building code welding 

and special inspection requirements. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique. 

6.4.2 Add New or Enhance Existing Wood Shear Wall 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses insufficient global or local strength or stiffness though 

the addition of or enhancement of vertical elements of the lateral force-resisting system. In W1A

buildings, stories with inadequate global first story strength and first story open fronts have been 

vulnerable in past earthquakes. Rehabilitation of shear walls perpendicular to the open front is 

often necessary. 

Description of the Rehabilitation Technique 

This rehabilitation technique involves the addition of a shear wall (framing and sheathing) or 

enhancement of an existing shear wall by the addition of sheathing, the addition of sheathing 

fastening, or a wood structural panel overlay. 

Added shear walls: When new shear wall framing and sheathing are being added, the most 

difficult design issue is mobilizing dead load to resist uplift due to shear wall overturning. 

Design for transfer of overturning forces to the supporting soils requires an understanding of the 

existing foundation configuration. Added shear walls can then be located to specifically make 

use of or avoid existing foundations. 

Figure 6.4.2-1 shows a shear wall located so that it can use the dead load carried by a building 

column to resist uplift at the left hand side and an existing bearing wall foundation at the right 
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Figure 6.4.2-1: Added Shear Wall Supported on Existing Foundation and Slab 
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hand side. The existing foundations need to be checked for adequate dead load resistance and 

adequate capacity to resist both up and down forces within material and soil strengths. At the left 

hand side, the existing column connection to the foundation needs to be capable of picking up 

the footing and surrounding slab. Use of this detail is limited not only by the adequacy of the 

foundation for overturning forces, but also the adequacy of the slab for shear anchorage. The slab 

must be thick enough to allow the installation of expansion bolts or adhesive anchors for anchor 

bolts. This starts being possible at a slab thickness of about four inches and is best with a slab of 

five inches or greater. Use of powder-driven fasteners for shear transfer to the slab is not 

recommended. Testing has found that these anchors fail prematurely under cyclic loads 

(Mahaney and Kehoe, 2002; and Cobeen, Russell, and Dolan, 2004). 

Figure 6.4.2-2 shows a shear wall supported on a new strip footing. The new footing runs 

between and is doweled into existing footings at each end, allowing the dead load of the existing 

footing to resist overturning. The addition of a new footing allows new anchor bolts to be cast-in, 

greatly simplifying shear anchorage. It also allows the addition of a curb to help reduce decay 

exposure in areas like garages that might have water exposure. 

Figure 6.4.2-3 shows a shear wall added away from any existing footings. A large pad-type 

footing will be needed to provide enough dead load to resist overturning forces. 

Enhanced shear walls: Section 5.4.1 provides a detailed discussion of enhancing shear walls by 

the addition of structural sheathing to walls currently braced with finish materials. This 

discussion is equally applicable to W1A building, and it is also applicable when it is decided to 

remove existing wood structural panel sheathing and replace it with new sheathing of higher 

capacity. 

Other approaches to enhancing shear wall capacity include the overlaying of new wood 

structural panel sheathing over existing sheathing and addition of fastening (added nails or 

staples) to existing sheathing. Figure 6.4.2-4A illustrates the addition of nails to increase shear 

wall capacity. New nails do not need to be added between every existing nail pair. It is 

acceptable to space them out to every second or third nail pair, as long as the average over two to 

three feet meets the needed spacing. It is desirable to distribute added nails as evenly as possible 

over the height of the wall. Too many nails can reduce performance: additional detailing 

requirements may be triggered, the wall overstrength will be increased, and demand on 

anchorages will be increase. Additionally, if not symmetrically placed, added nails can reduce 

the capacity of the shear wall (Cobeen, Russell and Dolan, 2004). 

Figure 6.4.2-4B illustrates the addition of staples. Staples are placed with their long direction 

parallel to the stud longitudinal direction in order to maintain edge distance in the stud and 

sheathing. It has been noted that workers placing staples have very little feel for whether the 

staple penetrates the stud, or is off the stud and only penetrates the sheathing (called a “shiner”). 

For this reason, careful attention to staple placement is required. This is only an acceptable 

approach when very modest increases in capacity are required, such that changes in detailing are 

not required (load path connections into and out of the shear wall, etc.). To date practice has been 

to waive the requirement for 3x framing at abutting panel joints when stapled shear walls are 

used. This is because the staples are thought to significantly reduce splitting of the wood 
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Figure 6.4.2-2: Added Shear Wall Supported by New and Existing Footings 
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Figure 6.4.2-3: Added Shear Wall Supported on a New Footing 

framing, greatly reducing the likelihood of stud failure. See further discussion in the Design 

Considerations section. 

Figure 6.4.2-5 illustrates use of shear wall wood structural panel overlay over existing wood 

structural panel sheathing. The figure illustrates the staggering of panel edges so that edge 

nailing of abutting panel edges on the inside and outside sheathing layers do not occur on the 

same framing member. Adequacy of overlay sheathing nail penetration into the framing member 

needs to be verified. This may be a problem where “short” sheathing nails are used, but not 

likely if full length common nails are used. At shear wall boundary members, both the inside and 

overlay sheathing need to be fastened to the boundary member. This may require the addition of 

a new boundary member at this location. One set of nails should not be relied on to fasten both 

sheathing layers. This approach has some potential issues, discussed in the Design 

Considerations section. 

Another possible use of an overlay is over existing lumber sheathing. See Design Considerations

section for discussion. 

Design Considerations 

Research basis: A significant amount of research for new shear walls can be considered 

applicable to this use. See Section 5.4.1. Testing of stapled shear walls has been conducted by 

APA (1999), Zacher and Gray (1985) and Pardoen (2003). Testing of sheathing-to-framing 

connections with staples, wood screws, and nails using two sheathing layers has been conducted 

by Fonseca et al., (2002). Limited testing of plywood overlays of plywood diaphragms has been 

conducted by APA (1999). 

Foundation design: The foundations, new or existing, have to be capable of resisting imposed 

forces. In Figure 6.4.2-1, the existing foundations need to be checked for both adequate dead 

load resistance and adequate capacity to resist up and down forces within material and soil 
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Figure 6.4.2-4: Enhanced Shear Wall Sheathing Fastening 
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Figure 6.4.2-5: Enhanced Shear Wall With Sheathing Overlay 

strengths. At the left hand side, the existing column connection to the foundation needs to be 

capable of picking up the footing and surrounding slab. In Figure 6.4.2-2, the new footing needs 

to be specifically designed for the loading; use of a typical footing section and reinforcing may 

not be adequate. The existing footings need to be checked for capacity to mobilize overturning 

resistance and to distribute downward reactions to the supporting soils. At the interface between 

the new and existing footings, vertical uplift and downward reactions are generally transferred 

through rebar doweling. Generally this is designed as a shear-friction connection, with the face 

of the existing footing cleaned; roughening the concrete surface to reduce the factor below 1.0 

is seldom practical, so a of 1 is generally used in design. In order to develop shear friction, the 

yield strength of the reinforcing needs to be developed on either side of the interface. 

Embedment depths to develop the reinforcing are generally available from the adhesive anchor 

manufacturer. If dowels are installed too close to the top or bottom of the footing, spalling can 

occur. Locating dowels near the center of the footing height reduces avoids spalling issues. 

Stapled shear walls: Use of stapled fastening of shear wall sheathing has been studied as a 

desirable approach to enhancement of existing shear walls for rehabilitation. Testing by Zacher 

and Gray (1985) found that use of staples avoided splitting of the framing members, making it 

possible to achieve higher capacities without adding in 3x studs at abutting panel edges. Stapled 

shear walls tested Pardoen, et al. (2003) show behavior indistinguishable from equivalent nailed 

shear walls. Testing of stapled connections by Fonseca et al., (2002) shows adequate load and 

deflection behavior, suggesting them to be equally acceptable. All of the staples tested eventually 

experienced fatigue failure, but this was after significantly more cycles than required by the 

loading protocol. When staples are being used to increase the capacity of existing shear walls, 

enough staples should be provided to carry the entire design shear. This is because the loaddeflection behavior of the staples can be expected to be different than existing nails due to the 
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very different fastener shank diameter. Stapled shear wall allowable design values are provided 

in the IBC (ICC, 2003a). 

Wood screw shear wall fastening: Wood screws are occasionally used for fastening of shear wall 

sheathing to wood framing. The very limited research available suggests that there are concerns 

with using this attachment type. In testing by Mahin (1980s), the brittle fatigue failure of cutthread wood screws was first noted. The screws failed at the transition from a full shank to a cut 

shank, this coincided with the framing to sheathing transition in the wall. This failure was 

repeated by Fonseca et al. (2002) when screw length was chosen to give minimum embedment. 

An increase in screw length to three inches significantly reduced but did not eliminate fatigue 

failure. Testing of rolled thread wood screws has not been identified. 

Shear wall overlay over wood structural panel sheathing: There are two primary reasons for 

using an overlay rather than removing existing sheathing and putting in new. One is to avoid the 

expense of removing material, the other is to make use of the capacity already provided and 

reduce thickness of added sheathing. The downside of using an overlay is that observation and 

modification of framing and framing connections is not possible. Overlay of wood structural 

panel sheathing has been used in past rehabilitation projects; however, concerns arise that 

deserve consideration. The deflection of shear walls under load involves the rotation of the 

sheathing panel as the wall framing racks. The primary energy dissipation method is through 

bending of sheathing nails due to the different deflection pattern of the sheathing and framing. 

The addition of an overlay with staggered edges will theoretically put significant deformation 

demands on nails being driven in two different deformation patterns (one by each sheathing 

layer). Available testing on fasteners in overlay conditions (Fonseca et al., 2002) showed a 

significant increase in fatigue failure of nails. APA (2000) investigated plywood overlays at the 

end of plywood diaphragms as a means of increasing shear capacity. Slow stepped loading 

without load reversals was used, and the overlay was found to successfully increase capacity. 

Because definitive information about performance of shear wall overlays is not available, caution 

in using this approach is recommended. 

Shear wall overlay over straight lumber sheathing: Straight lumber sheathing is generally 

flexible enough and of low enough capacity that when overlayed, the behavior of the wood 

structural panel sheathing can govern. This makes it acceptable to overlay straight sheathing; 

however, there is no benefit from the sheathing remaining, other than reduced work due to 

removal. Where removal of the straight sheathing is possible, it is preferred. Only the capacity of 

the wood structural panel sheathing should be relied upon. It is recommended that edge nailing 

of wood structural panel sheathing be through straight sheathing into framing in all cases, since 

reduced embedment could lead to reduced overstrength capacity due to nail withdrawal. Special 

attention needs to be paid to developing shear transfer to boundary members, since nailing must 

be through straight sheathing to the boundary member framing behind. 

Shear wall overlay over diagonal lumber sheathing: The load-deflection behavior and fastener 

deformation patterns of diagonal lumber sheathing and wood structural and sheathing are 

considerably different, raising questions about the behavior resulting from the combination of the 

two. Due to lack of information, use is not recommended without a detailed study of behavior. 
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Mixing of shear wall deformation capacities: Designers are particularly cautioned against using 

shear wall systems or enhancements with deformation capacities less than the balance of the 

story or building (i.e. less than the two percent of story height drift permitted by current codes 

for ordinary occupancy structures). Because the building or story deformation demand or target 

displacement will be largely determined by the rest of the vertical elements, introduction of a 

stiffer element with limited deformation capacity could result in premature failure. 

Detailing Considerations 

Shear walls separating parking areas from residential areas may be part of fire-rated assemblies. 

Any fire rating needs to be maintained in the rehabilitation work. When wood structural panel 

sheathing is applied over gypsum wallboard, increased nail sizes are required by the building 

code. Because cyclic testing has not been conducted for sheathing applied over gypsum 

wallboard, the implications for drift are not known. Testing of gypsum wallboard has shown 

crushing of the gypsum, with cycled loading resulting in slotting of the wallboard and significant 

slip. The same behavior may lead to increased deflection where wood structural panel sheathing 

is applied over gypsum wallboard sheathing. 

Cost and Disruption Considerations 

The primary cost of enhancing existing shear walls comes from the disruption of the occupants 

and the removal of finishes to gain access to the structural walls. The cost of materials and 

connections is generally minor is comparison. As a result, it is preferable to keep the variation in 

sheathing, nailing, and connections to a minimum, making execution of the work as simple as 

possible. Planning on removal and replacement of existing sheathing can facilitate project 

schedule by minimizing the need to address unexpected existing sheathing conditions while 

construction is in progress. Other design and detailing measures that can make execution of the 

work more predictable are encouraged. 

See Section 6.4.1 for discussion of field welding cautions. 

Construction Considerations 

As in new construction, it can be a challenge to assure that rehabilitation measures are 

constructed with the fastener (nail, staple, screw, etc.) type and size that has been assumed in 

design and construction documents. Use of improper type and size often results in reduced 

rehabilitation measure capacity. Most nails are placed with nail guns. Most gun nails are ordered 

by diameter and length. Indications of type and pennyweight continue to be misleading. The only 

way to verify that required fasteners are being used is to measure them with calipers or a similar 

device. Fasteners connecting sheathing to framing should not be overdriven (not break the face 

ply of the sheathing). Where overdriving occurs, fastener capacity may be reduced up to 40%. 

Often plumbing, HVAC or electrical lines will be running in the floor framing in the vicinity of 

shear walls. This is particularly problematic where they cross over the shear wall at critical 

locations for shear or overturning transfer. Some disruption in the transfer of shear into the top of 

a shear wall will generally need to be accommodated, typically this means that there are a 

number of joist bays in which blocking and clips can not be installed. Within residential units, 

relocation of utilities is often not an option. In other areas, relocation of utilities may be more 

practical. 
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Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique other than the use of 

adhesive anchors as part of the assemblage. 

6.4.3 Enhance Framing Supporting Shear Wall 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses inadequate beams, posts, and their interconnection 

supporting vertical overturning forces from ends of discontinued upper story shear walls. The 

primary focus is support of existing shear walls, but the discussion applies equally to support of 

enhanced shear walls. 

Description of the Rehabilitation Technique 

This rehabilitation measure involves the addition or supplementing of beams, posts, beam-topost connections and post-to-foundation connections to support discontinued upper story shear 

walls. 

Figure 6.4.3-1 illustrates the addition of new supports and connections where an upper story 

shear wall is added or enhanced. This figure shows a new beam, post and foundation system 

being added. Ideally, the posts would be added immediately under the shear wall ends; however, 

the layout of the first story will often dictate other support locations. The beam, post, beam-topost and post-to-foundation connections must be designed for overstrength or special seismic 

load combinations is using ASCE 7 or IBC, or as force-controlled members per FEMA 356. 

Either approach will amplify the demand on these members and connections. Overturning 

anchorage of the shear wall is addressed in Section 5.4.1 and Figure 5.4.1-11C. Shear transfer at 

the wall base is addressed in Figure 5.4.1.9. Where overturning forces from the wall are 

significant, wood beam sizes may prove too large to be practical, in which case a steel beam may 

be needed. Where a steel beam is used, use of steel columns may also be practical and provide 

stronger and stiffer beam-to-column connections. Where an existing beam exists but is not 

adequate, the addition of new steel channels on either side of the beam can provide a practical 

solution. See Figure 6.4.3-2. Attention is needed to adequate load transfer into and out of the 

channels, including end supports and uplift connections. 

Design Considerations 

Research basis: No research applicable to the rehabilitation measure has been identified. 

History: The failure of concrete columns supporting the Olive View Hospital during the 1971 

San Fernando earthquake dramatically demonstrated the significant demands placed on members 

supporting discontinued bracing systems; however, this was not commonly considered in design 

of woodframe buildings until the 1997 NEHRP Provisions (FEMA, 1998) and 1997 UBC 

(ICBO, 1997) when special requirements for supporting members were expanded from columns 

to beams, columns and connections, and explicit application to woodframe was noted. The 

requirement of design for expected forces for new construction is now included in ASCE 7 

(ASCE, 2005) and the IBC (ICC, 2003a), for regions of high seismic hazard, but not other 

regions. As a result, most buildings will not have been designed considering expected forces ( 0

overstrength or special seismic load combinations). A systematic evaluation in accordance with 
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Figure 6.4.3-1: Enhanced Overturning Support for Upper Story Shear Wall 
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Figure 6.4.3-2: Enhanced Beam Supporting Discontinued Shear Wall 

FEMA 356 requires that these supporting members be evaluated as force-controlled, with forces 

coming from 1.5 times the yield strength of the supported wall. This will have the same or a 

more critical effect than design per ASCE 7 and IBC requirements. As a result, support upper 

story shear walls will most likely be identified as a deficiency. As discussed in Section 6.3, 

however, rehabilitation for this deficiency has seldom occurred to date. 

Support in crawl spaces: Where vertical support is needed for interior first story walls above 

crawlspaces with post and pier floor systems and spread footings, the easiest and least expensive 

rehabilitation is the addition of new foundation to support the shear wall. This is best 

accomplished by addition of blocking under the shear wall, fastening of a pressure treated sill 

with pre-placed anchor bolts, and casting of the concrete footing to the underside of the 

foundation sill. Access and ventilation openings in the new foundation may be required. 
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Detailing Considerations 

See Section 5.4.1 for discussion of wood framing issues, applicable to floor blocking and added 

beams. Any time wood and steel members are connected to each other, the detailing needs to 

accommodate wood change in dimension with moisture content (either shrinkage or expansion). 

Figure 6.4.3-2 provides one example of where this must be considered. An existing wood beam 

inside of a conditioned building would be anticipated to have very little dimensional change, 

while a new beam or a beam with exposure to weather or humidity could have significantly 

more. In Figure 6.4.1-1D dimensional change was accommodated through the use of slotted 

holes. In Figure 6.4.3-2 it is important that the holes in the steel strap and channels not be 

slotted. Oversized holes in the wood beam could be used to accommodate dimensional change. 

Cost/Disruption 

This rehabilitation measure will require simultaneous access to the story with the shear wall and 

the story below. Significant areas of ceiling will need to be removed to access work. The ceiling 

in the garage of a W1A or W2 building may be plaster rather than gypsum wall board and may 

be part of a fire-rated assembly separating the garage area from the residential units. Any fire 

rating would have to be maintained in the rehabilitation work. 

Construction Considerations 

Often plumbing, HVAC or electrical lines will be running in the floor framing in the vicinity of 

shear walls. This is particularly problematic where they cross under the shear wall at critical 

locations for shear or overturning transfer. In some cases it is practical to accommodate these 

utilities in the structural design. In other cases relocation of utilities may be more practical. 

Welding of steel members requires adequate access and special ventilation measures in enclosed 

buildings. Welding of steel members in the vicinity of woodframe construction can be a 

significant fire hazard and should only be undertaken by experienced welders. Welding should 

always be done by certified welders using approved welding techniques in compliance with 

building code welding and special inspection requirements. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique. 

6.4.4 Enhance Overturning Detailing in Existing Wood Shear Wall 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses inadequate or missing load path detailing for uplift and 

downward forces at the ends of shear walls, between shear wall and foundation, or between 

upper story and lower story shear walls. The uplift load path may have inadequate or missing tiedown devices and detailing. The compression load path may have inadequate compression 

capacity in the wall framing or through the floor framing depth. 

Description of the Rehabilitation Technique 

Where there is a calculated net uplift force at the ends of shear walls, proprietary tie-down 

connectors are fastened to the wall framing and foundation to resist the uplift forces. The tiedown connectors may be fastened to existing framing or new framing. They may be used in 
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combination with existing shear wall sheathing (generally on the exterior face of exterior walls), 

new sheathing on the interior face, or new sheathing on the exterior face. Tie-down vertical bolts 

are generally fastened to existing foundations with adhesive anchors. 

Except for very lightly loaded walls, tie-downs are generally needed to develop the in-plane 

strength and stiffness of wood structural panel and diagonally sheathed shear walls, as discussed 

in the Design Considerations section. Tie-downs may potentially be used on stucco shear walls, 

but are seldom used on gypsum wallboard shear walls due to the low capacity. 

Figure 6.4.4-1A illustrates a shear wall elevation with commonly used tie-down connectors for a 

slab-on-grade condition. Figure 6.4.4-1B illustrates fastening to develop a load path between the 

tie-down connector and the shear wall sheathing edge nailing. For sheathing and framing 

conditions other than those shown, similar fastening must be provided to complete the load path. 

Figure 6.4.4-1A: Shear Wall Elevation with Enhanced Overturning Detailing 
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Figure 6.4.4-1B: Framing Fastening for Overturning Load Path 
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Figure 6.4.4-1C: Tie-down Details at Alternate Base Conditions 
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Tie-down connectors in first story walls above woodframed floors require detailing 

modifications; Figure 6.4.4-1C illustrates anchorage to the foundation in locations with a frame 

floor and a framed floor plus cripple wall. For both of these conditions, there is generally not 

enough height to install tie-down connectors in the floor framing or cripple wall space, so the tiedown is installed in the first story wall and the tie-down bolt is extended through the joist and 

cripple wall height to anchor into the foundation. Occasionally cable or rod tie-down systems 

running the full height of one or more stories will be used in lieu of the tie-down brackets shown 

in the figures. 

Design Considerations 

Research basis: Research results comparing in-plane strength and stiffness with and without tiedowns are summarized in Cobeen, Russell and Dolan (2004). Applicable research includes 

Mahaney and Kehoe (2002), Salenikovich (2000), Ni and Karacabeyli (2000), Salenikovich and 

Dolan (1999) and Fischer et al. (2001). The drop in shear wall capacity without tie-downs varies 

as a function of wall length and wall axial loading. Strength reductions up to approximately 80% 

(20% retained strength) were observed without tie-downs. Reductions in wall stiffness varied, 

but in general mirrored the drop in strength. Unless rehabilitation of uplift capacity is provided, 

the reduced strength and stiffness needs to be accounted for in building evaluation. 

Adequacy of tie-down post or studs: The stud or post that the tie-down connector is fastened to 

must be designed to carry required tension and compression forces. Calculations of tension 

capacity must consider any reduction in the post/stud net section, such as would occur at bolted 

tie-downs. Where multiple stories contribute tension or compression forces to a post/stud, the 

full accumulated force must be considered. Single 2x studs should be carefully evaluated before 

they are used as tie-down studs and should be limited to appropriate loads. Where existing 

framing members are not adequate, new tie-down posts can be added if fastening is provided to 

complete the load path (see Detailing Considerations section). Where multiple 2x studs are to 

form a built-up post, it is important that stitch nailing between studs be adequate to develop the 

wall shear capacity. 

In addition, tie-down connectors are believed to create flexure as well as tension in the post/stud 

being connected (Pryor, 2002). Where bare posts have been tested alone (no sheathing, wall 

framing), the flexure has been seen to cause both failure of the post and pull-through of bolts 

connecting the tie-down to the post (Nelson, 2005, and Nelson and Hamburger, 1999). The stud 

or post should be checked for combined tension and flexure. The type of tie-down chosen can 

reduce the flexure. Use of tie-downs fastened with nails or wood screws rather than bolts avoid 

net section reduction at the bolts and reduce possible slip. This type has been favored in 

California since the Northridge earthquake. Alternately, bolted tie-downs can be placed 

symmetrically on each side of a stud or post to minimize flexure. 

Tie-down bracket devices developed by manufacturers since the Northridge earthquake have also 

tended to be stiffer, minimizing deformation within the bracket device. The stiffer tie-down 

reduces the portion of wall drift generated by uplift at the tie-down. In addition, less uplift at the 

wall end should reduce the likelihood of foundation sill plate splitting because sill uplift is also 

restrained. Stiffer tie-downs are recommended to the extent practical, as reduced wall drift 
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should translate into less damage. Tie-downs that might have brittle failures at expected 

earthquake loads should be avoided. 

Tie-down design criteria: FEMA 356 (FEMA, 2000) identifies fasteners used to transfer forces 

from wood to wood or wood to metal as being deformation-controlled actions. When coupled 

with several relatively high m-factors for static procedures, this can result in less fastening being 

required by FEMA 356 than the current building codes. At the same time, the shear wall 

sheathing fastening is identified as the desired location of inelastic behavior, which suggests that 

shear wall overturning restraint should be force-controlled and more fastening provided. It is 

recommended that current building code requirements be used for FEMA 356’s Basic Safety 

Objective. For a higher performance objective, a capacity-based approach is suggested. 

Foundation anchor type and installation: Discussion of foundation anchor type and installation 

can be found in Section 5.4.3. Anchorage of tie-down tension bolts to existing foundations will 

almost exclusively use adhesive anchors, which have more compatible capacities and allow more 

convenient installation. To date, it has been common to install the adhesive anchor straight down 

into the footing, or at a very slight angle if required for access. The capacity based on adhesive 

bond can be taken from manufacturer information. In the past, concrete anchorage design 

methods in the UBC (ICBO, 1997) have allowed calculation of the concrete pull-out capacity 

based on an assumed failure surface. New provisions in ACI 318 Appendix D (ACI, 2005) will 

not allow tie-down anchorage using current configurations. With typical anchors centered at 1-

3/4 inch from the edge of concrete, required cover cannot be met, the seismic load requirement 

that steel rather than concrete control is difficult to meet, and side blow-out tends to restrict 

calculated capacities. Although this appendix chapter excludes adhesive anchors, it is difficult to 

consider rehabilitation anchorages acceptable that would not be acceptable for new cast-in-place 

connections. One possible alternative is to angle the tie-down rod in the concrete to get better 

cover and reduce calculated side blow-out. Although some proprietary cast-in anchors use this 

configuration, testing for rehabilitation use has not occurred. 

Adequacy of foundation: Tie-down connectors should be attached to substantial existing footings 

that have the shear and flexural capacity to mobilize required resistance. Alternately, new 

footings or footing reinforcement can be provided. Addition of tie-down connectors at isolated 

footings or unreinforced masonry footings should receive very careful design consideration. 

Detailing Considerations 

Vertical shear load path: It is important that a load path be provided between the tie-down 

connector and a stud or post that has adequate fastening to the structural sheathing. Where new 

shear wall sheathing is provided, the tie-down connector is installed on a post/stud that receives 

sheathing edge nailing over the entire wall height (Figures 6.4.4-1B2 and 6.4.4-1B3). Where 

existing panel sheathing is being used, it is necessary to install the tie-down at an existing 

post/stud with sheathing edge nailing (Figure 6.4.4-1B1). Additional nailing may be required to 

maintain a load path between the tie-down post/stud and the post/stud with sheathing edge 

nailing, as seen at the right hand side of Figure 6.4.4-1B1. The nail size and spacing will need to 

be calculated to match the shear wall capacity. Where this is not possible, the structural 

sheathing should be exposed at the tie-down locations in order to provide adequate nailing into 

the tie-down member. Use of adhesive attachment of the tie-down post/stud to the structural 
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sheathing should not be used as part of this load path because the stiffness of this sheathing to 

framing connection is not compatible with expected slippage between the sheathing and framing 

during shear wall racking. 

Vertical compression load path: When shear wall uplift is occurring at one end of a shear wall, a 

downward reaction is occurring at the other end. A load path to transmit this compression 

through the wood framing to the foundation is generally provided at the same location as the tiedown. Often, compression blocking is added in the floor framing depth to provide full bearing of 

the post/stud on the top and bottom plates, as shown in Figures 6.4.4-1A and 6.4.4-1C. It is 

important that dry framing be used; otherwise, shrinkage is likely to make the blocking 

ineffective. See the Section 5.4.1 discussion of shrinkage. 

Tie-down connectors: Tie-down connectors are almost exclusively proprietary. Connector types 

used for retrofit include brackets, straps, and occasionally full-height rod or cable systems. 

Where possible, it is preferred to not mix the connector types within a shear wall. All connectors 

should be installed in accordance with the manufacturer’s recommendations and applicable ICC 

Evaluation Services report recommendations. 

Where straps are used, the manufacturer specified capacity of the strap is dependent on the 

number of fasteners (nails or screws) installed at each end of the strap. It is important that the 

required fasteners are provided between the strap and the wall studs. Nails into the floor framing 

or top and bottom plates should not be counted toward the required amount. The length of the 

strap must be adjusted to allow installation of the proper number of fasteners into the studs. This 

should be clearly specified on the tie-down strap detail. 

Tie-down bolts: The vertical bolt between the tie-down bracket and the foundation, or between 

the bracket in a story above and below, is usually all-thread rod. 

Anchorage to the foundation: It is most common to use adhesive anchors for anchorage of the 

vertical tie-down bolt to the foundation. The calculation of the required anchorage depth must 

take into account the edge distance to the near face of the foundation and the foundation 

capacity. It is often desirable to lengthen the embedment into the foundation beyond that 

required by the adhesive anchor manufacturer, in order to better mobilize the foundation 

capacity. Adhesive anchors must be installed in accordance with the manufacturer’s 

recommendations and the applicable ICC Evaluation Services report recommendations. 

Cost/Disruption 

Rehabilitation of woodframe shear walls often occurs while the building is still being occupied. 

This generally involves phased construction and moving furniture from room to room ahead of 

the work. This slows down the work, but can be less expensive and disruptive for the occupants 

than relocating them. When the building will be occupied a choice is sometimes made to do all 

of the work from the building exterior, keeping the interior as functional as possible, or 

completely from the building interior, avoiding opening of the building finishes. This choice 

greatly affects design and detailing, so it should be made very early in the design process. 
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Where existing shear wall sheathing has adequate shear capacity, it may be possible to 

selectively open interior finishes to install tie-down connectors, greatly limiting the disruption to 

the occupants. If locations of sheathing edge nailing are well known, it may be possible to only 

open up a space one stud bay wide and several feet high at each wall end. More likely, however, 

it will be necessary to open up the stud bay for the full wall height to provide adequate 

interconnection of framing members. Often shear transfer connections will also need to be 

provided, requiring the opening of a strip of wall finish along the base of the wall and another 

strip of ceiling at the wall top. 

Construction Considerations 

It is not uncommon for significant variation to occur in the framing detailing of existing 

buildings. It is important that conditions be observed during construction of rehabilitation 

measures, and details be modified for as-built conditions. This is most effectively done by 

scheduling time between opening of finishes and start of installation for the engineer to observe 

conditions and provide needed guidance. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique other than the use of 

proprietary connectors and adhesives as part of the assemblage. 

6.4.5 Enhance Shear Transfer Detailing 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses detailing for transfer of shear into and out of shear walls. 

Description of the Rehabilitation Technique 

The addition or enhancement of shear walls is not of value unless shear forces can be transferred 

into and out of the wall. Section 5.4.1 addresses a wide variety of shear transfer details for the 

top and bottom of shear walls where the existing wall top plates will serve as collector elements. 

This will be applicable in most instances in W1 and W1A buildings. Where new shear walls are 

added, however, it is likely that new collector elements will be needed. In W1A buildings, shear 

walls are generally well distributed and resist moderate loads. This section discusses collectors 

and shear transfer for moderate loads in new shear walls. Sections 7.4.1 and 7.4.2 address 

addition of collectors for new vertical elements with significant strength and stiffness, including 

steel braced frames and concrete and masonry walls. These elements are most often added in W2

buildings. 

In a W1A building, the capacity of the roof of floor diaphragms are very likely to be less than the 

capacity of added or enhanced shear walls. The collector needs to extend well beyond the length 

of the shear wall, as a minimum engaging adequate diaphragm length to resist forces. Ideally a 

collector would extend for the entire length of the diaphragm being supported. 

Figure 6.4.5-1 illustrates collectors transferring load into the top of a new or enhanced shear 

wall. Figures 6.4.5-1A and 6.4.5-1B show new or existing framing parallel to the wall used as a 

collector. Detail A assumes that fastening of the diaphragm sheathing to framing exists or can be 

provided. Load transfer to the diaphragm can occur over the length of the new or existing 
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Figure 6.4.5-1: Collector Details 
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framing member without any splices being required. Where additional length of attachment to 

the diaphragm is required, splicing of the collector framing in accordance with Figure 6.4.5-2 is 

needed. Often framing in older buildings has a significant lap length over interior supports, 

sometimes making a direct nailed, screwed or bolted connection between existing members 

possible. Where bolts are used, detailed attention is needed to provide required bolt end and edge 

distances and spacing. Alternate splice approaches include steel straps and plates. 

Where collector member connection to the diaphragm above is not practical, a wood structural 

panel soffit, as shown in Detail B can be used to transfer load from the collector to the 

diaphragm. A minimum soffit width of four feet will generally ensure that at least one row of 

diaphragm edge nailing is engaged. For large unit shears, additional soffit width and length can 

distribute loads further. 

Figures 6.4.5-1C through 6.4.5-1E illustrate collector details where the existing framing is 

perpendicular to the wall. Because continuous framing is not available to act as a collector, steel 

straps or sections are used. Straps will generally be assumed to only carry tension loads. 

Blocking, already provided for shear transfer is assumed to carry compression loads. Blocking 

needs to have a tight fit in order to minimize deformation. Detailing is needed if splices will 

occur in the collector. Figure 6.4.5-3 illustrates an elevation of a collector where framing is 

perpendicular to the shear wall, corresponding to Figures 6.4.5-1C or 6.5.4-1D. 

Design and Detailing Considerations 

Research basis: Testing of shear transfer connections between wood structural panel diaphragms 

and shear walls below was conducted by Ficcadenti et al. (2004). No research applicable to steel 

straps and blocking for collectors has been identified.

Deformation in the collector: In order to be the most effective, the deformation of a collector 

should be as compatible as possible with the roof diaphragm it is attached to. Generally roof 

diaphragms in W1A buildings will be short span and quite stiff, suggesting that a stiff collector 

is preferable. This can be best achieved through use of existing framing members of as long 

lengths as possible, as illustrated in Figure 6.4.5-2. Splices in the collector members should also 

be reasonably stiff, as slip at the splice could result in tension in the diaphragm. 

Where framing runs perpendicular to the framing direction, there is sometimes little choice but to 

use steel straps for tension and blocking or framing members for compression, as shown in 

Figure 6.4.5-3. Unless the straps are reasonably stiff and blocking is installed tight, significant 

deformation could occur in the collector, resulting in limited efficiency for transferring loads. 

Although this type of collector is used commonly in new construction and rehabilitation, little is 

know about its effectiveness and resulting building performance. Conversely, however, 

significant distress in diaphragms in W1A buildings has only been seen at significant changes in 

geometry such as re-entrant corners (Schierle, 2002). If collectors are to be installed, it is 

recommended that they be made as stiff and tight-fitting as possible. Sizing of the collector 

member using overstrength forces or as force-controlled actions will help keep the collector stiff, 

therefore increasing likely performance. 
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Figure 6.4.5-2: Collector Using Existing Framing Parallel to the Shear Wall 
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Figure 6.4.5-3: Collector Using Added Blocking at Framing Perpendicular to Shear Wall 

Where the top of the existing diaphragm sheathing cannot be accessed for additional sheathing 

nailing, sheathing added at the ceiling soffit can help distribute forces into the existing 

diaphragm, as shown in Figure 6.4.5-1B. 

Cost and Disruption Considerations 

Removal of existing floor or roof finishes to nail diaphragm sheathing into new collector 

members can be both costly and disruptive. It is, however, going to provide the most predictable 

performance and is recommended for highly loaded walls and where a performance objective 

higher than life-safety is intended. Other fastening methods can be calculated and detailed, 

however not enough is known about their ability to perform adequately. 

Construction Considerations 

Tight fit of framing, blocking and straps is critical to limiting deformation and improving 

performance of the collectors and shear transfer connections. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique other than the use of 

proprietary connectors as part of the assemblage. 
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Existing Buildings: FEMA 547 Multi-Unit Residential Woodframes

6-40 

6.5 References 

ACI, 2005, Building Code Requirements for Reinforced Concrete and Commentary, ACI 318, 

American Concrete Institute, Farmington Hills, MI. 

AF&PA, 2005, Special Design Provisions for Wind and Seismic, ASD/LRFD, American Forest 

& Paper Association, Washington, D.C. 

AISC, 2005, Seismic Provisions for Steel Buildings, American Institute of Steel Construction, 

Chicago, IL. 

APA, Revised 2000, Research Report 138, Plywood Diaphragms, APA The Engineered Wood 

Association, Tacoma, WA. 

APA, Revised 1999, Research Report 158, Preliminary Testing of Wood Structural Panel Shear 

Walls Under Cyclic (Reversed) Loading, APA The Engineered Wood Association, Tacoma, WA. 

ASCE, 2005, Minimum Design Loads for Buildings and Other Structures (ASCE 7), American 

Society of Civil Engineers, Reston VA. 

City of Los Angeles and SEAOSC, 1994, Wood Frame Construction Report and 

Recommendations, City of Los Angeles and Structural Engineers Association of Southern 

California, Los Angeles, CA. 

Cobeen, K., J. Russell, and J.D. Dolan, 2004, Recommendations for Earthquake Resistance in 

the Design and Construction of Woodframe Buildings, CUREE Publication No. W-30, 

Consortium of Universities for Research in Earthquake Engineering, Richmond, CA. 

FEMA, 2000, Prestandard and Commentary for the Seismic Rehabilitation of Buildings (FEMA 

356), Federal Emergency Management Agency, Washington, D.C. 

FEMA, 1998, NEHRP Recommended Provisions for Seismic Regulations for New Buildings and 

Other Structures (FEMA 302), Federal Emergency Management Agency, Washington, D.C. 

Ficcadenti, S., E. Freund, G. Pardoen, and R. Kazanjy, 2004, Cyclic Response of Shear Transfer 

Connections Between Shear Walls and Diaphragms in Woodframe Construction, CUREE 

Publication No. W-28, Consortium of Universities for Research in Earthquake Engineering, 

Richmond, CA. 

Fischer, D., A. Filiatrault, B. Folz, C.M. Uang, and F. Seible, 2001, Shake Table Tests of a TwoStory Woodframe House, CUREE Publication No. W-06, Consortium of Universities for 

Research in Earthquake Engineering, Richmond, CA. 

Fonseca, F., S. Rose and S. Campbell, 2002, Nail, Screw and Staple Fastener Connections, 

CUREE Publication No. W-16, Consortium of Universities for Research in Earthquake 

Engineering, Richmond, CA. 
Techniques for the Seismic Rehabilitation of Chapter 6 - Type W1A: Multistory, 

Existing Buildings: FEMA 547 Multi-Unit Residential Woodframes

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ICBO, 1997, Uniform Building Code, International Conference of Building Officials, Whittier, 

CA. 

ICBO, 2001, Guidelines for Seismic Retrofit of Existing Buildings (GSREB), International 

Conference of Building Officials, Whittier, CA. 

ICC, 2003a, International Building Code, International Code Council, Country Club Hills, IL. 

ICC, 2003b, International Existing Building Code (IEBC), International Code Council, Country 

Club Hills, IL. 

ICC, 2005a, 2003 International Existing Building Code (IEBC) Commentary, International Code 

Conference, Country Club Hills, IL. 

ICC, 2005b, Proposed Changes to the 2003 IEBC, International Code Council, Country Club 

Hills, IL. 

ICC Peninsula Chapter, 2004, Simplified Design Procedure for Structural Steel Ordinary 

Moment-Resisting Frame (OMRF) 9/3/04 draft), ICC Peninsula Chapter, CA. 

Lizundia, B. and W. Holmes, 2000, Practical Solutions for Improving the Seismic Performance 

of Buildings with Tuckunder Parking, prepared for the City of San Jose Department of Housing 

and Office of Emergency Services, San Jose, CA. 

Mahaney, J. A. and B. E. Kehoe, 2002, Anchorage of Woodframe Buildings, Laboratory Testing 

Report, CUREE Publication No. W-14, Consortium of Universities for Research in Earthquake 

Engineering, Richmond, CA. 

Mahin, S., 1980s, personal communication. 

Mosalam, K., et al., 2002, Seismic Evaluation of an Asymmetric Three-Story Woodframe 

Building, CUREE Publication No. W-19, Consortium of Universities for Research in Earthquake 

Engineering, Richmond, CA. 

Nelson, R., 2005, personal communication. 

Nelson, R. and R. Hamburger, 1999, “Hold-down Eccentricity and the Capacity of the Vertical 

Wood Member,” Building Standards, November-December 1999, International Conference of 

Building Officials, Whittier, CA. 

Ni, C. and E. Karacabeyli, 2000, “Effect of Overturning Restraint on Performance of Shear 

Wall,” Proceedings of the World Conference on Timber Engineering, University of British 

Columbia, Vancouver, CA. 
Techniques for the Seismic Rehabilitation of Chapter 6 - Type W1A: Multistory, 

Existing Buildings: FEMA 547 Multi-Unit Residential Woodframes

6-42 

Pardoen, G., A. Waltman, R. Kazanjy, E. Freund, and C. Hamilton, 2003, Testing and Analysis 

of One-Story and Two-Story Walls Under Cyclic Loading, CUREE Publication No. W-25, 

Consortium of Universities for Research in Earthquake Engineering, Richmond, CA. 

Porter, K., J. Beck, H. Seligson, C. Scawthorn, L.T. Tobin, R. Young and T. Boyd, 2002, 

Improved Loss Estimation for Woodframe Buildings, CUREE Publication No. W-18, Consortium 

of Universities for Research in Earthquake Engineering, Richmond, CA. 

Pryor, S. 2002, The Effect of Eccentric Overturning Restraint in Complete Shear Wall 

Assemblies, Simpson Strong-Tie Company, Inc., Dublin, CA. 

Rutherford & Chekene, 2000, Seismic Rehabilitation of Three Model Buildings with Tuckunder 

Parking: Engineering Assumptions and Cost Information, prepared for the City of San Jose 

Department of Housing and Office of Emergency Services, May. 

Salenikovich, A., 2000, The Racking Performance of Light-Frame Shear Walls, Virginia 

Polytechnic Institute and State University, Blacksburg, CA. 

Salenikovich, A. and J. Dolan, 1999, “Effects of Aspect Ratio and Overturning Restraint on 

Performance of Light-Frame Shear Walls Under Monotonic and Cyclic Loading,” Proceedings 

of the Pacific Timber Engineering Conference, 1999, Volume 3, New Zealand Forest Research 

Institute, Rotorua, New Zealand. 

Schierle, G.G., 2002, Northridge Earthquake Field Investigations: Statistical Analysis of 

Woodframe Damage, CUREE Publication No. W-09, Consortium of Universities for Research in 

Earthquake Engineering, Richmond, CA. 

Schierle, G. G. editor, 2001, Woodframe Project Case Studies, CUREE Publication No. W-04, 

Consortium of Universities for Research in Earthquake Engineering, Richmond, CA. 

SEAOC (Structural Engineers Association of California), 2002, Recommended Guidelines for 

Ordinary Moment Frames (OMF) for Buildings, drafted by the SEAOSC Steel Ad-Hoc 

Committee for the SEAOC Seismology Committee (August 5, 2002 version), Structural 

Engineers Association of California, Sacramento, CA. 

Seismic Safety Commission, State of California, 1994, 1994 Northridge Earthquake Buildings 

Case Studies Project (Proposition 122; Product 3.2 SSC 94-06), Sacramento, CA. 

Steinbrugge, J., V. Bush, and J. Johnson, 1996, “Standard of Care in Structural Engineering 

Woodframe Multiple Housing,” Proceedings, 1996 Convention, Structural Engineers 

Association of California, Structural Engineers Association of California, Sacramento, CA. 

Vukazich, S., 1998, The Apartment Owner’s Guide to Earthquake Safety, City of San Jose, San 

Jose, CA. 
Techniques for the Seismic Rehabilitation of Chapter 6 - Type W1A: Multistory, 

Existing Buildings: FEMA 547 Multi-Unit Residential Woodframes

6-43 

Zacher, E. and R.G. Gray, 1985, “Dynamic Tests of Woodframed Shear Panels,” Proceedings, 

1985 Convention, Structural Engineers Association of California, Structural Engineers 

Association of California, Sacramento, CA. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 7 - Type W2: Woodframes, Commercial and Industrial

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Chapter 7 - Building Type W2: Woodframes, Commercial and 

Industrial 

7.1 Description of the Model Building Type 

Building Type W2 consists of commercial, institutional, and smaller industrial buildings 

constructed primarily of wood framing. Most W2 buildings have first floor slab-on-grade 

construction; however, woodframe floors supported on foundation walls or cripple walls occur. 

The upper floor and roof framing consist of wood joists and can include wood or steel trusses, 

beams, and columns. Post and beam framing is common at interior and at storefronts or garage 

openings. Lateral forces are resisted by woodframe diaphragms and shear walls. In older 

buildings, steel rod bracing systems may also be used in place of diaphragms. In newer 

buildings, wood shear walls are sometimes used in combination with isolated concrete or 

masonry shear walls or steel braced frames or moment frames. Figure 7.1-1 provides one 

illustration of this building type. 

Figure 7.1-1: Building Type W2: Woodframes, Commercial and Industrial 

Design Practice 

Design practice for W2 buildings can include no design, design per conventional constructions 

provisions, engineered gravity design and conventional construction bracing, and engineered 

gravity and lateral design. More W2 buildings are likely to have an engineered gravity design 

than W1 and W1A buildings because the framing systems often fall beyond conventional 

construction provisions. Lateral bracing of multistory W2 buildings in accordance with 

conventional construction provisions was permitted by the UBC (ICBO, 1994) through the 1994 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 7 - Type W2: Woodframes, Commercial and Industrial

 7-2 

edition. Construction of single-story W2 buildings using conventional construction provisions is 

still permitted under the IBC (ICC, 2003). 

In California, woodframe school buildings constructed in the 1950s included engineered lateralforce-resisting systems with tie-down anchors and diagonal lumber or plywood sheathed shear 

walls and diaphragms (Jephcott and Hudson, 1974). Commercial construction in California 

would generally be anticipated to match school construction, with a time lag. It is anticipated that 

most W2 buildings constructed today will have engineered gravity and lateral designs. 

Walls and Other Vertical Elements 

While wall bracing materials can include the same range discussed for W1 buildings, the use of 

diagonal lumber sheathing or wood structural panel sheathing is much more likely in W2

buildings. The use of overturning anchorage would be varied between the 1950s and 1970s, but 

common in engineered buildings from the 1980s on. 

W2 buildings often have significantly fewer interior bracing walls than W1 and residential 

portions of W1A buildings. School buildings have moderate room sizes. Commercial buildings 

with simple geometries are often braced only at the building perimeter, creating large open 

rooms. Interior bracing may be added for more complex geometries. In newer commercial 

buildings, concrete or masonry shear walls, steel moment frames, or steel braced frames are 

sometimes used at the street front or as interior bracing walls in order to maximize the 

occupant’s or user’s ability to see across the retail or office space. These vertical elements are 

used specifically because needed bracing capacity can be provided by much shorter element 

lengths than with woodframe shear walls. Inclusion of these vertical elements requires additional 

attention to force distribution, potential torsional irregularities, and collectors adequate to 

transmit lateral loads to the elements. 

Cripple walls, also discussed with the W1 building type, sometimes occur in W2 buildings. See 

Chapter 5. 

Floor and Roof Diaphragms 

Floor and roof diaphragms include the same materials as the W1 building type; however, plank 

and beam systems are rare in W2 buildings. Significant in W2 buildings is the occurrence of 

longer diaphragm spans and more complicated roof diaphragm configurations. The longer spans 

should result in larger force and deformation demands in the diaphragms and more out-of-plane 

movement of walls following the diaphragm deflection. More complicated roof configurations 

require attention to boundary members at diaphragm edges and vertical offsets in chords and 

collectors. 

7.2 Seismic Response Characteristics 

Many W2 buildings, like Building Types W1 and W1A are short period with inelastic behavior 

concentrated in the vertical elements. Some W2 buildings (primarily single story), however, 

have long-span diaphragms, creating the possibility of high stresses, inelastic behavior, and high 

deformation in the diaphragm. 
Techniques for the Seismic Rehabilitation of

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7.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

Very little information has been published on the earthquake performance of W2 buildings. 

Earthquake reconnaissance report discussions of wood buildings have tended to focus on 

residential rather than commercial and light industrial uses. One exception to this is an 

exhaustive and detailed review school building performance in the 1971 San Fernando 

earthquake (Jephcott and Hudson, 1974). Only occasional and generally moderate damage is 

reported to have occurred in one-story and two-story woodframe school buildings. This is 

consistent with observations of schools made following the Northridge earthquake (EERI, 1996). 

For schools, however, nonstructural damage was reported to be significant. 

While reports of damage are scarce, the construction materials and demands are essentially the 

same as in W1 and W1A buildings, and many of the vulnerabilities and damage types should be 

expected to be similar. In fact, the generally larger building size and fewer interior walls should 

make W2 buildings more vulnerable than W1 or W1A. This was true in a case study of the 

Satellite Student Union Center at California State University, Northridge (Schierle, 2001), where 

significant damage to finish materials occurred. In addition, the W2 category includes buildings 

such as churches that often have very irregular building configurations, which should make them 

susceptible to damage. 

See below for general discussion and Table 7.3-1 for a detailed compilation of common seismic 

deficiencies and rehabilitation techniques for the Building Type W2. 

Global Strength and Stiffness 

Global strength and stiffness can be of concern in W2 buildings. This is particularly true where 

use of the first story results in few structural and nonstructural walls and open fronts. 

Rehabilitation is commonly addressed by the addition or enhancement of wood shear walls, or 

the addition of steel moment frames, steel braced frames or concrete or masonry shear walls. 

Where W2 buildings are large in plan area, it may become practical to introduce a steel braced 

frame to resist lateral loads. The braced frame can resist higher loads than wood shear walls and 

lighter moment frames, allowing concentration of lateral loads into fewer and shorter bracing 

elements. This increases the level of force in the collector and at the base resisting shear and 

overturning. Occasionally concrete or masonry shear walls are used for rehabilitation; this must 

be done with caution however, because the weight of the wall will increase seismic forces 

perpendicular to the wall and attention to wall anchorage is required. 

Configuration 

The open-front torsional irregularities and weak cripple wall configuration deficiencies 

introduced in W1 and W1A buildings are equally applicable to W2 buildings. In addition, 

mixing of lateral force systems in W2 buildings can lead to torsional irregularities. Where shear 

walls are mixed with other vertical bracing elements, care should be taken in evaluating the 

distribution of lateral forces and deformations. Torsional irregularity may have contributed to 

damage to the CSU Northridge Satellite Student Center (Schierle, 2001). Common measures for 

rehabilitation of torsional irregularities include the addition of steel moment frames, wood shear 

walls, steel braced frames and concrete or masonry shear walls. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 7 - Type W2: Woodframes, 

7-4 

Table 7.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for W2 Build

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce Dem

Global Strength 

Insufficient in-plane wall strength 

Wood structural panel shear wall [6.4.2] Steel braced frame [7.4.1] Steel moment frame [6.4.1] 

Enhance woodframe shear wall [6.4.2] 

Uplift anchorage and compression posts [6.4.4] 

Replace heavroof finish wlight finish 

Global Stiffness 

Insufficient in-plane wall stiffness 

Wood structural panel shear wall [6.4.2] Steel braced frame [7.4.1] Steel moment frame [6.4.1] 

Enhance woodframe shear wall [6.4.2] 

Uplift anchorage and compression posts [6.4.4] 

Weak story, missing or weak cripple wall 

Wood structural panel shear wall [5.4.3], [6.4.2] Add woodframe cripple wall Add continuous foundation and foundation wall 

Enhance woodframe shear wall [6.4.2] Enhance woodframe cripple wall [5.4.4] 

Configuration 

Torsional irregularity including open front 

Wood structural panel shear wall [6.4.2] Proprietary wall Steel moment frame [6.4.1] Concrete or masonry wall 

Enhance woodframe shear wall [6.4.2] 

Inadequate shear anchorage to foundation 

Anchorage to foundation [5.4.3] 

Load Path 

Inadequate overturning anchorage 

Uplift anchors and compression posts [6.4.4] 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 7 - Type W2: Woodframes, 

7-5 

Table 7.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for W2 Build

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce Dem

Inadequate shear transfer in wood framing 

Enhance load path for shear [5.4.1], [6.4.5] 

Load Path (continued) 

Inadequate collectors to vertical elements 

 Enhance existing 

collector [7.4.2] 

Add collectors [6.4.5], [7.4.2] 

Inadequate in-plane strength and/or stiffness 

Enhanced existing diaphragm [22.2.1] 

Replace heavroof finish wlight finish 

Inadequate chord capacity 

Enhance chord members and connections [22.2.2] 

Excessive stresses at openings and irregularities 

Enhance diaphragm detailing 

Diaphragms 

Re-entrant corners Enhance diaphragm 

detailing 

Foundations See Chapter 23 [ ] Numbers note in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 7 - Type W2: Woodframes, Commercial and Industrial

7-6 

W2 buildings with inadequate cripple wall bracing and foundation anchorage are just as 

vulnerable as similar W1 and W1A buildings. Rehabilitation of these deficiencies is 

recommended to be highest priority. For W2 buildings, use of an engineered rather than 

prescriptive design for cripple wall bracing and bolting is recommended. 

Load Path 

The load path deficiencies in W2 buildings are much the same as W1 and W1A buildings. 

Rehabilitation measures typically involve fasteners and connectors to resist shear and 

overturning, and addition of collectors. As in W1 buildings, anchorage to the foundation is a 

high priority for rehabilitation in W2 buildings. 

Diaphragm Deficiencies 

W2 buildings can have highly irregular diaphragms, with vertical offsets, folded plates, and sawtooth configurations. Rehabilitation of chords and collectors is key to adequate performance of 

irregular diaphragms. Rehabilitation enhancing the capacity of the diaphragm is discussed in 

Chapter 22. 

7.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

7.4.1 Add Steel Braced Frame (Connected to Wood Diaphragm) 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses inadequate global or local strength or stiffness through 

the addition of a new steel braced frame element. 

Description of the Rehabilitation Technique 

Figure 7.4.1-1A illustrates an elevation of a steel braced frame added in a W2 building. This type 

of element is generally introduced because it can provide needed bracing capacity in a short 

element length. The resulting highly loaded element will almost always require the addition of a 

significant collector to transfer load into the top and the addition of a significant new foundation 

to transmit forces to the supporting soils. Because existing foundations and collectors would not 

likely be adequate, it is common to add these elements in an area clear of existing foundations 

and beams. 

Figures 7.4.1-1B1 and 7.4.1-1B2 illustrate unit shear transfer over the length of the braced frame 

for framing perpendicular and parallel to the frame. 

Figures 7.4.1-1C1 and 7.4.1-1C2 illustrate collector members and their connection to framing 

perpendicular and parallel to the frame. These details are discussed further in Section 7.4.2. 

Figure 7.4.1-2 illustrates a two-story steel braced frame added in a W2 building. Significant in 

this detail is that the second floor is opened up allowing the braced frame to run continuous over 

the two-story height. Installing a separate frame at each story would lead to unmanageable 

connection details; the two-story configuration provides the strongest and stiffest solution. Shear 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 7 - Type W2: Woodframes, Commercial and Industrial

7-7 

Figure 7.4.1-1A: Steel Braced Frame Added in W2 Building 

Figure 7.4.1-1B: Shear Transfer and Collector for Steel Braced Frame 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 7 - Type W2: Woodframes, Commercial and Industrial

7-8 

Figure 7.4.1-1C: Shear Transfer and Collector for Steel Braced Frame 

Figure 7.4.1-2: Two-Story Steel Braced Frame in a W2 Building 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 7 - Type W2: Woodframes, Commercial and Industrial

7-9 

transfer, collectors, and frame member bracing need to be provided at the second floor as well as 

the roof. 

Design Considerations 

Research basis: See Chapter 9 of this document for detailed discussion of steel braced frames. 

No research applicable to steel braced frames in wood buildings has been identified. 

Foundation: The cost of the new braced frame foundation will be a significant part of the cost of 

this rehabilitation measure. Generally, the dead load available to resist foundation uplift will be 

minimal, so a large and heavy foundation is often needed. In some cases, it may be necessary to 

resort to drilled piers or helical anchors to provide uplift resistance. 

Detailing Considerations 

See Chapter 9 for discussion of detailing of the steel braced frame. See Section 5.4.1 for 

discussion of basic issues related to rehabilitation of woodframe structure, including wood 

shrinkage and splitting. These issues are pertinent to load path connections for attachment of 

steel frames and collectors into the existing woodframe structure. 

Steel Connections: Details for connections within the steel braced frame and from the frame to 

the collector will need to give careful consideration to access for field assembly and field 

welding. Out-of-plane bracing will be required at the top of the columns and as required by the 

AISC Seismic Provisions for Structural Steel Buildings (AISC, 2005) along the length of the 

beam. See Section 6.4.1 for discussion of similar connection at a steel moment frame. 

Cost and Disruption Considerations 

The addition of the steel frame will be disruptive to the area immediately surrounding the frame. 

Field welding is very difficult to avoid when adding steel braced frames to existing buildings, but 

should be minimized as much as possible. See Section 7.4.2 for discussion of interruption due to 

the steel collector. 

Construction Considerations 

Placement of the steel columns is one of the significant construction challenges, particularly in a 

multistory frame as shown in Figure 7.4.1-2. It may be desirable to place the steel columns and 

then cast the foundation and/or slab concrete around them. This allows the depth of the footing to 

be used for maneuvering the steel column. See Section 6.41 for discussion of field welding 

issues. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique. 

7.4.2 Provide Collector in a Wood Diaphragm 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses provision of collectors to new high capacity vertical 

elements in wood diaphragm buildings. This technique is primarily intended for use with steel 

braced frames, as discussed in Section 7.4.1, but would also be applicable to a collector for a 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 7 - Type W2: Woodframes, Commercial and Industrial

7-10 

new concrete or masonry shear wall in a W2 building. Sections 6.4.5 and 5.4.1 discuss shear 

transfer and collectors for woodframe vertical elements with low to moderate loads. 

Description of the Rehabilitation Technique 

Figures 7.4.1-1C1 and 7.4.1-1C2 illustrate a steel collector member added perpendicular or 

parallel to existing diaphragm framing. Fastening is provided for shear transfer between the 

collector and the diaphragm. As discussed in Section 7.4.1, it is assumed that this collector is 

added in a location away from existing continuous framing members. See Section 6.4.5 for 

collector alternatives where this is not the case. 

In Figure 7.4.1-1C1 new blocking is added, and nailing is provided between the diaphragm 

sheathing and blocking and between the blocking and a wood nailer. The nailer is attached to the 

collector with welded steel studs. A 3x framing member is shown as the collector because it 

gives better bolt values and because it provides enough depth to allow counter sinking of the 

washer if required. 

In Figure 7.4.1-1C2, a wood structural panel ceiling soffit is used to distribute collector forces 

into the existing diaphragm. 

Design Considerations 

Research basis: No research applicable to installation of steel braced frames in a W2 building 

have been identified; however Section 6.4.1 discussion of shear transfer and collector detailing 

for steel moment frames (Mosalam et al., 2002) (Cobeen, Russell, and Dolan, 2004) is similar to 

this technique. 

Design demand: The primary deformation in a wood diaphragm should occur as nail slip in the 

fasteners attaching the sheathing to the framing. The most effective collector will allow this slip 

to occur, while not adding other sources of significant deformation. This is why the collector is 

illustrated as a substantial steel section rather than a light steel strap. The collector and 

connections would require design for overstrength forces in accordance with current building 

codes. Similarly, use of force-controlled actions would be appropriate when using rehabilitation 

guidelines. Design to avoid yielding of the steel section is recommended. 

Nailing of the existing sheathing to the framing or blocking should not be increased beyond what 

is required for design level forces or deformation-controlled actions. Note, however, that design 

level forces may require two rows of diaphragm edge nailing at the collector member, as the sum 

of the unit shear from two sides may be up to twice the diaphragm unit shear capacity. It is 

highly recommended that roof sheathing or floor finish materials be removed to allow the 

sheathing edge nailing to be installed from the top of the existing sheathing (Figure 5.4.1-6A). 

The other fastening and connections between the steel collector and existing roof sheathing 

should be designed for amplified forces to the extent possible in order to limit deformation. 

Connection alternatives shown in Figures 5.4.1-6B, 5.4.1-6C and 5.4.1-6D are not recommended 

for this level of demand. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 7 - Type W2: Woodframes, Commercial and Industrial

7-11 

Detailing Considerations 

Extent of collector: It is recommended that the collector be extended for the full dimension of the 

diaphragm wherever practical. If the collector is stopped short of the end, the change in 

diaphragm shear and therefore, deformation may occur at the collector end. 

Detailing of splices: Collector member splice locations should be planned, and splice details 

should be developed. Collector interruptions at beams may make logical splice locations. In 

Detail C2, it may be possible to move the collector up into the area between joists, thus avoiding 

collector breaks at beams. Where this is the case, ideal splice locations may be a few feet away 

from beams. Collector compression forces should be considered in splice design. 

Tolerances in existing floor framing: It can generally be expected that there will be some 

unevenness in the underside of the existing floor framing in the areas where the steel beam and 

collectors are to be added. It is best to anticipate and include in detailing shimming or other 

approaches to dealing with this tolerance. Detailing should show locations where shimming is 

acceptable, set upper limits on acceptable shimming, and adjust fastener capacity or length to 

account for reduced fastener penetration when shimming is provided. 

Cost/Disruption 

The addition of collectors at the underside of roof or floor framing can be quite disruptive in 

buildings that are in use because of the extent of the work; however with adequate planning the 

work can generally be installed quickly. Where quick work is desirable, the ceiling should be 

removed for observation of existing conditions over the full extent of the collector prior to steel 

fabrication. In a one-story building, disruption of occupants can be reduced by installing the 

collector member on the roof top. This will require removal and replacement of roofing, and 

adjustment of roof drainage if drainage is altered by the added collector member. 

Construction Considerations 

Welding of steel studs: Threaded steel studs should be welded to the steel collector in a 

fabrication shop with periodic special inspection. Field welding of the studs is discouraged due 

to a lower level of control and fire hazard. Smaller fabrication shops may not have fusion 

welding equipment for attachment of the studs. A fillet weld around the stud perimeter is 

acceptable when used with wood nailers. Slight routing of the wood nailer may be required to 

accommodate the weld. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique. 

7.5 References 

AISC, 2005, Seismic Provisions for Steel Buildings, American Institute of Steel Construction, 

Chicago, IL. 

Cobeen, K., J. Russell, and J.D. Dolan, 2004, Recommendations for Earthquake Resistance in 

the Design and Construction of Woodframe Buildings, CUREE Publication No. W-30, 

Consortium of Universities for Research in Earthquake Engineering, Richmond, CA. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 7 - Type W2: Woodframes, Commercial and Industrial

7-12 

ICBO, 1994, Uniform Building Code, International Conference of Building Official, Whittier, 

CA. 

ICC, 2003, International Building Code, International Code Council, Country Club Hills, IL. 

Jephcott, D. K. and D. E. Hudson, September 1974, The Performance of Public School Plants 

During the San Fernando Earthquake, Earthquake Engineering Research Laboratory, California 

Institute of Technology, Pasadena, CA. 

Mosalam, K., et al., 2002, Seismic Evaluation of an Asymmetric Three-Story Woodframe 

Building, CUREE Publication No. W-19, Consortium of Universities for Research in Earthquake 

Engineering, Richmond, CA. 

Schierle, G. G. editor, 2001, Woodframe Project Case Studies, CUREE Publication No. W-04, 

Consortium of Universities for Research in Earthquake Engineering, Richmond, CA. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 8 - Types S1/S1A: Steel Moment Frames

8-1 

Chapter 8 - Building Types S1/S1A: Steel Moment Frames 

8.1 Description of the Model Building Type 

Building Type S1 consists of an essentially complete frame assembly of steel beams and 

columns. Lateral forces are resisted by moment frames that develop stiffness through rigid 

connections of the beam and column created by angles, plates, and bolts, and/or by welding. 

Moment frames may be developed on all framing lines or only in selected bays. It is significant 

that no structural walls or steel braces are provided. Floors are cast-in-place concrete slabs or 

concrete fill over metal deck. These buildings are used for a wide variety of occupancies such as 

offices, hospitals, laboratories, and academic and government buildings. Figure 8.1-1 shows an 

example of this building type. 

Building Type S1A is similar but has floors and roofs that act as flexible diaphragms such as 

wood or untopped metal deck. One family of these buildings is older warehouse or industrial 

buildings, while another more recent use is for small office or commercial buildings in which the 

fire rating of concrete floors is not needed. 

Figure 8.1-1: Building Type S1: Steel Moment Frames 
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Variations Within the Building Type 

The use of structural steel in building construction started in the last decades of the 19th century. 

Lateral load resistance was initially provided by masonry infill. Additional resistance may have 

been provided by the encasement of the steel members in concrete, though these encased 

members were not designed with composite considerations. In the 1920s, the use of riveted 

connections introduced steel moment frames. Beam flanges and webs were joined to the 

columns by structural shapes, most commonly T-sections. Rivets had low strength and ductility, 

which limited the overall capacity of the frames. In the 1960s, high strength bolts replaced the 

rivets. Bolts were faster to install and permitted larger clamping forces, increasing the rigidity of 

the frames. Connections became smaller when cover plates bolted to the beam flanges and 

welded to the columns replaced T-sections in the 1950s. By the 1960s, beam flanges and webs 

were welded directly to the columns to create fully restrained connections, initiating the welded 

steel moment frame (WSMF) construction era. Shear tabs bolted to the beam webs and welded 

to the columns later replaced welded beam webs. These welded-flange and bolted-web 

connections were used extensively in the 1970s and 1980s and are now known as pre-Northridge 

connections. After the 1994 Northridge earthquake, it was concluded that these connections did 

not provide enough plastic rotational capacity for most seismic applications (FEMA, 2000c). 

Several types of connections have been developed and tested to address the flaws in the preNorthridge connections. The newer connections are designed to develop the full moment 

capacities of the beams and provide large inelastic rotation capacity. 

Floor and Roof Diaphragms 

Diaphragms associated with this building type could be either rigid or flexible. The typical rigid 

diaphragm found in modern buildings consists of structural concrete fill on metal deck. 

Diaphragm forces transfer to the frames through shear studs welded to the beams. Older steel 

buildings that were constructed before metal decks were commonly used may have concrete 

slabs or masonry arches that span between the beams. Flexible diaphragms include bare metal 

deck or metal deck with nonstructural fill. These are frequently used on roofs that support light 

gravity loads. Decks could be connected to the steel members with shear studs, puddle welds, 

screws, or shot pins. The steel members also act as chords and collectors for the diaphragm. 

Foundations 

There is no typical foundation for this building type. Foundations can be of any type including 

spread footings, mat footings, and piles, depending on the characteristics of the building, the 

lateral forces, and the site soil. Spread footings are used when lateral forces are not very high 

and a firm soil exists. For larger forces and/or poor soil conditions, a mat footing below the 

entire structure is commonly used. Pile foundations are used when lateral forces are extremely 

large or poor soil is encountered. The piles can be either driven or cast-in-place. Vertical forces 

are distributed to the underlying soil through a combination of skin friction between the pile and 

soil and/or direct bearing at the end of the pile; lateral forces are resisted primarily through 

passive pressure on the vertical surfaces of the pile cap and piles. 

8.2 Seismic Response Characteristics 

Steel moment frame buildings are generally flexible, but subject to large interstory drifts. The 

ductility of these buildings is achieved through yielding and plastic hinging of beams and/or 
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shear yielding of column panel zones at beam-column connections. This inelastic behavior 

allows moment frames to sustain many cycles of loading and load reversals. Historically, it was 

believed that large plastic rotations could be developed without significant strength degradation. 

Up until the Northridge earthquake, the performance of steel moment frame buildings was also 

believed to far exceed that of masonry and concrete buildings based on observations from 

previous earthquakes. 

The Northridge earthquake exposed severe deficiencies in WSMF connections. A significant 

number of the frames inspected after the earthquake exhibited visible cracking in the beam 

flange-to-column flange welds. In a few rare cases, the flanges completely fractured and the 

damage extended into either the shear tab or the column panel zone. Newer buildings, which 

relied on deeper beams with thick flanges and less redundancy, were discovered to be even more 

susceptible to this type of damage. In retrospect, a review of data from past earthquakes 

indicates that WSMF buildings rarely received close inspection following the event, and often 

these buildings were overlooked due to the more obvious damage in other types of structures 

(FEMA, 2000d). This oversight also applies to older steel buildings with riveted or bolted 

connections. These older buildings, though designed as moment frames, may have suffered 

limited damage due to the interaction of the steel frames with infill walls and concrete cores. 

Lastly, if steel damage was discovered after an earthquake, engineers often attributed this to poor 

construction quality (FEMA, 2000d). 

8.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

See Table 8.3-1 for deficiencies and potential rehabilitation techniques particular to this system. 

Selected deficiencies are further discussed below by category. 

Global Strength 

The lack of global strength is caused by insufficient frame strength, resulting in excessive 

demands on the existing frames. Yielding or fracturing of the beams, columns, and/or 

connections could lead to excessive drifts. As a result, the building could be deemed irreparable 

after an eathquake. 

Global Stiffness 

Moment frames are much more flexible than other types of lateral force-resisting systems. Their 

flexibility could lead to excessive building drifts and interstory drifts. This is likely to cause 

structural damage to the connections and nonstructural damage to the partitions and cladding. 

Additional concerns include P- effects and pounding with adjacent buildings. Common 

rehabilitation measures include strengthening the existing frames or providing new vertical 

lateral force-resisting elements. 

Configuration 

Soft story conditions occur when stiffness from one floor to the next changes abruptly. This is 

common at ground floors of commercial and office buildings with tall first stories. It could also 

occur at mid-heights of five-story to fifteen-story story tall buildings that have not been designed 

for higher mode effects and near field motions. 
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Table 8.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S1/S1A Bui

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce De

Global Strength 

Insufficient frame strength 

Moment frame 

Braced frame [8.4.1] 

Concrete/masonry shear wall [8.4.2] 

Steel plate shear wall [8.4.8] 

Strengthen beams [8.4.3], columns [8.4.3], and/or connections [8.4.6], [8.4.9] 

Seismic is[24.3] 

Supplemendamping [

Global Stiffness 

Excessive drift Moment frame 

Braced frame [8.4.1] 

Concrete/masonry shear wall [8.4.2] 

Steel plate shear wall [8.4.8] 

Strengthen beams [8.4.3], columns [8.4.3], and/or connections [8.4.6], [8.4.9] 

Supplemendamping [

Configuration Soft story Moment frame 

Braced frame [8.4.1] 

Concrete/masonry shear wall [8.4.2] 

Steel plate shear wall [8.4.8] 

 Re-entrant 

corner 

Moment frame 

Braced frame [8.4.1] 

Concrete/masonry shear wall [8.4.2] 

Collector [8.4.4] 

Enhance detailing [8.4.3], [8.4.4] 

Missing collector 

Load Path Add collector [8.4.4] 

Inadequate shear, flexural, and uplift anchorage to foundation 

Embed column into a pedestal bonded to other existing foundation elements [8.4.5] 

Provide steel shear lugs or anchor bolts from base plate to foundation [8.4.5] 
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Table 8.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S1/S1A Bui

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce De

Load Path (continued) 

Inadequate outof-plane anchorage at walls connected to diaphragm 

Tension anchors 

[16.4.1] 

Component Detailing 

Inadequate capacity of beams, columns, and/or connections 

Enhance beam-column connection [8.4.6] 

Add cover plates or box members [8.4.3] 

Provide gusset plates or knee braces [9.4.1] 

Encase columns in concrete 

 Inadequate 

capacity of panel zone 

Provide welded continuity plates [8.4.6] 

Provide welded stiffener or doubler plates [8.4.6] 

 Inadequate 

capacity of horizontal steel bracing 

Provide additional secondary bracing [9.4.2] 

Strengthen bracing elements 

Reduce unbraced lengths 

Strengthen connections 

Diaphragms Inadequate 

in-plane strength and/or stiffness 

Collectors to distribute forces [8.4.3], [8.4.4] 

Moment frame 

Braced frame [8.4.1] 

Concrete/masonry shear wall [8.4.2] 

Steel plate shear wall [8.4.8] 

Concrete topping slab overlay 

Wood structural panel overlay at flexible diaphragms [22.2.1] 

Strengthen chords [8.4.3], [8.4.4], and [22.2.2] 

Add nails at flexible diaphragms [22.2.1] 

 Inadequate 

shear transfer to frames 

Provide additional shear studs, anchors, or welds [22.2.7] 
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Table 8.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S1/S1A Bui

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce De

Diaphragms (continued) 

Inadequate chord capacity 

Add steel members or reinforcement [8.4.3], [8.4.4] 

 Excessive 

stresses at openings and irregularities 

Add reinforcement [8.4.3] 

Provide drags into surrounding diaphragm [8.4.4] 

Foundations See Chapter 23 [ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 
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Load Path 

Load path deficiencies in steel moment frame buildings include inadequate collectors and frame 

anchorage to foundations. In Type S1 buildings, seismic forces transfer from the diaphragm to 

the frame through shear studs welded to collectors or directly to the frame beams. The collectors 

or the connections to the frame may be too weak and insufficient to transfer these forces. 

Connections from columns to a base plate or pile cap have to resist shear, flexural and potentially 

uplift forces. Connections that cannot develop these frame forces do not allow the frame to 

develop its full capacity. 

Component Detailing 

The most common detailing deficiencies in steel moment frames are related to beam-column 

connections. Pre-Northridge welded moment connections consist of complete penetration flange 

welds and bolted or welded shear tabs. These connections were previously thought to be ductile 

but were found to fracture at small plastic rotations or even under elastic loads. Rehabilitation 

techniques primarily focus on forcing the yielding and plastic hinging to occur away from the 

joint to reduce stresses on the welds. This can be achieved with the use of reduced beam 

sections (RBS) or strengthening the section of a beam adjacent to the beam-column joint. These 

techniques are presented in Section 8.4.6 and discussed in detail in AISC Design Guide 12 (Gross 

et al., 1999) and FEMA 351 (FEMA, 2000b). Before welded moment connections became 

common, connections were either riveted or bolted. These types of connections are prone to net 

section fracture. 

Panel zones were previously thought to provide excellent ductility and strain hardening and were 

given increased predicted shear strength in the 1988 Uniform Building Code (FEMA, 2000c). 

The increased strength also meant larger inelastic demands in the panel zones. As a result, panel 

zones are subjected to yielding or buckling before adjacent members fully develop their 

capacities. Studies performed after the Northridge earthquake found that large panel zone 

deformations could increase the potential for connection failures (FEMA, 2000c). 

Welded column splices are vulnerable to fracture when subjected to large tensile loads (FEMA, 

2000b), since they generally use partial penetration groove welds and thus, are not designed for 

the full capacity of the smaller column. 

Diaphragm Deficiencies 

Common diaphragm deficiencies include insufficient in-plane shear strengths, inadequate 

chords, and excessive stresses at openings. Causes for these deficiencies could be due to lack of 

slab or fill thickness, lack of reinforcing steel in the slabs, insufficient connections to chord 

elements, and poor detailing at openings. See Chapter 22 for common rehabilitation techniques. 

Foundation Deficiencies 

Foundations that are inadequate do not develop the full capacity of the lateral force-resisting 

system. Their deficiencies result from insufficient strengths and sizes of footings, grade beams, 

pile caps, and piles. See Chapter 23 for common rehabilitation techniques. 
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8.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

8.4.1 Add Steel Braced Frame (Connected to an Existing Steel Frame) 

Deficiency Addressed by Rehabilitation Technique 

Moment frame buildings that are insufficient to resist lateral forces or too flexible to control 

building drifts can be converted into braced frame buildings. 

Description of the Rehabilitation Technique 

The seismic performance of a building may be improved by adding braces to existing welded or 

riveted steel moment frames. Braces can be added without substantially increasing the mass of 

the building. Various concentrically braced frame (CBF) configurations should be considered, 

though some tend to perform much better than others in earthquakes. In addition, systems that 

meet the provisions for special concentrically braced frames (SCBF) are expected to exhibit 

stable and ductile behavior in large earthquakes. See Chapter 9 for a discussion of different CBF 

configurations and their behaviors. Moment frames are not commonly converted to eccentrically 

braced frames (EBF) due to complicated design and detailing issues that would be encountered. 

Different brace types can be used, including W-shapes, hollow structural sections (HSS), steel 

pipes, double angles, double channels, double HSS, and buckling-restrained braces. 

Design Considerations 

Adding braced frames to a moment frame building increases its stiffness considerably. The 

upgraded structure should be evaluated for higher lateral and overturning forces accordingly. 

The primary system performance issues are those associated with CBF systems, in which the 

most vulnerable elements are the braces and their end connections. Thorough knowledge of the 

existing material behaviors and strengths are necessary to ensure that new and existing elements 

to interact in the desired manner. Other design issues include the following: 

Research basis: No references directly addressing the addition of steel braced frames to moment 

frame buildings have been identified.

Brace locations: Preference of brace locations should be given to existing moment frame bays to 

utilize the strengths of existing members, connections, and foundations. If this is not possible 

and braces are added at other locations, the existing moment frames should be considered when 

forces are distributed to the lateral force-resisting system. If the building drifts are large enough, 

deficiencies in the moment frames may still require mitigation whether the frames are included 

in the new lateral force-resisting system or not. 

Brace selection: Use compact and non-slender sections whenever possible to avoid premature 

fracturing or buckling of the braces during post-yield behavior. Issues related to conventional 

structural shapes are discussed in Chapter 9. Two particular brace types not common in older 

braced frame buildings are double HSS section and buckling-restrained braces. Double HSS 

sections can be used in configurations similar to double angles or channels (Lee and Goel, 1990). 

They provide reduced fit-up issues and smaller width-to-thickness ratios compared to a single 

HSS, resulting in increased energy dissipation capacity. The other type of brace, used in a 
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buckling-restrained braced frame (BRBF), is typically used in new buildings but has also been 

used successfully in new BRBF systems in existing buildings. One example of a brace used in a 

BRBF consists of a steel core inside a casing, which consists of a hollow structural section (HSS) 

infilled with concrete grout. Proprietary materials separate the steel core and concrete to prohibit 

bonding between the two materials. There are other buckling-restrained braces that do not use 

grout or additional separating agents between the steel and grout. The main advantage of these 

braces is the ability of the casing to restrain the buckling of the steel core without providing any 

additional axial force resistance beyond the capacity of the steel core. Provisions for new 

building BRBF design are included in the NEHRP Recommended Provisions (FEMA, 2003) and 

the AISC Seismic Provisions for Structural Steel Buildings (AISC, 2005b). 

Nonstructural issues: The addition of braces to an existing structure changes the architectural 

character of the building. Braces in exterior frames will be visible in buildings with clear glazing. 

At interior bays, braces have to be configured to avoid obstruction of existing corridors, 

doorways, and other building systems. Braces are also commonly exposed and incorporated into 

the interior architecture. For this particular option, note that gusset plates designed in accordance 

with the AISC Seismic Provisions for SCBF can be fairly large and should be discussed with the 

architect and tenants. If the braced frames are hidden in partition walls, the architect should be 

aware that these walls will be thicker than typical walls. Beams that are increased in size and 

new collectors may affect nonstructural components by reducing clear floor heights. These 

components typically include suspended ceilings, pipes, conduits, and ducts. Coordination with 

the architect and other trades should not be overlooked or underestimated.

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 

available, and conducting comprehensive field surveys, the following issues should be noted: 

Connections: Gusset plates provide greater tolerances for the installation of braces than directly 

attaching the braces to the frame members. It may be impossible to completely eliminate 

welding at a braced frame connection in a seismic rehabilitation. Continuity and doubler plates 

on columns, stiffener plates on beams, brace-to-gusset plate connections, and most gusset plateto-frame member connections all require welding. A typical fully welded connection appropriate 

for use in SCBF is shown in Figure 8.4.1-1. If the brace capacity is governed by out-of-plane 

buckling, stable post-buckling behavior can be achieved by allowing the gusset plate to develop 

restraint-free plastic rotations. AISC recommends an offset of two times the plate thickness 

along the brace centerline, measured from the end of the brace to a line perpendicular to the 

nearest point on the gusset plate constrained from out-of-plane rotation. If gaps are provided 

between the gusset plate and concrete fill, then only the beam or column can act as constraints, 

as is the case in Figure 8.4.1-1. On the other hand, if concrete is placed directly against the 

gusset plate, then the slab can also act as a constraint. 

For low to moderate seismic applications, Figure 8.4.1-2 shows a more compact connection. An 

example of a W-shape used as a brace and welded directly to the beam and column is shown in 

Figure 8.4.1-3. The two latter connections do not allow for restraint-free plastic rotations out-ofplane and should be used primarily in situations where in-plane buckling of the braces govern or 
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Figure 8.4.1-1: HSS Brace at Existing Beam-Column Connection in SCBF 
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Figure 8.4.1-2: HSS Brace at Existing Beam-Column Connection in Ordinary CBF 

ductility demands are low. In addition, the connection shown in Figure 8.4.1-3 requires 

extensive field welding and could present fit-up issues if field dimensions are not verified on a 

case-by-case basis. 

Buckling-restrained braces are typically bolted to gusset plates. Their bolt configurations at 

these connections allow for very small tolerances. Double angle and double channel braces can 

be easily bolted to a gusset plate but space restrictions may limit the number of bolts and 

subsequently, the strength of the connection. 
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Figure 8.4.1-3: HSS Braced at Existing Beam-Column Connection 

It may not always be practical or possible to locate a work point at the intersection of the 

centerlines of the beam and column, e.g. deep beams. This may deemed to be acceptable if the 

eccentricity is included in the connection design. 

Built-up brace members: While double angles, double channels, and double HSS offer 

advantages for installation, more stringent criteria apply to these members when used in SCBF. 

These include stitch spacing, member compactness, and strength of the stitches (AISC, 2005b). 

Reinforcing cover plates: HSS and pipe braces are subject to net section fracture at the gusset 

plate slots (Uriz and Mahin, 2004, and Yang and Mahin, 2005). This brittle failure mode can be 

eliminated by adding reinforcing cover plates to the sides of the HSS without the slots, as shown 

in Figure 8.4.1-1 and Figure 8.4.1-2. For pipes, the reinforcing plates can be oriented at right 

angles to the pipe and appear like stiffeners. 
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Cost/Disruption 

This system could be cost-effective when compared to other alternatives for upgrading steel 

moment frame buildings. Costs will be less when existing moment frames are converted into 

braced frames to take advantage of the existing strength and stiffness of the frame members, 

connections, and foundations. Designs that are simple and details that are not overly 

complicated will also minimize costs. 

Costs can also be reduced if disruption is minimal during construction. Installing braces at the 

perimeter frames reduces logistical issues associated with working in confined spaces and 

temporary removal of the nonstructural elements. Noise associated with this type of work is loud 

and disturbing to the tenants if the building is occupied while the work is being performed. 

Construction Considerations 

The engineer’s involvement during the construction phase is critical during a seismic 

rehabilitation. The design of the retrofit scheme must not neglect the construction phase and 

should consider these issues at a minimum: 

Welding issues: A work environment in which the welder can perform quality welds is critical. 

This includes an environment with adequate space to properly operate welding equipment, 

adequate lighting, and a stable work platform for overhead welds, all while ensuring worker 

health and safety. Additional considerations include venting of welding fumes and fire 

protection. 

Removal of existing nonstructural elements: Exterior cladding and interior partitions must often 

be removed to deliver and install the braces to their final locations. Connection modifications at 

the roof level may warrant the removal of roofing and waterproofing. Installation of the 

connection to the underside of beams, column continuity and doubler plates, and beam stiffeners, 

will affect ceilings, lights, and other mechanical/electrical/pluming components. Almost all steel 

buildings will have some form of fireproofing around the members and connections. With older 

buildings, asbestos may be present in the fireproofing, which could require costly abatements to 

expose the members. Older buildings with items of historical significance may require 

additional coordination and effort to ensure that these items are removed and restored properly. 

Buildings constructed in the early 20th century commonly had steel members encased in 

concrete. 

Removal of existing structural elements: Slabs and metal decks must be chipped and cut away to 

install connections. Slabs oriented perpendicular to the beams require temporary shoring. 

Temporary openings in the slab not only require shoring but also consideration for how the 

openings will be closed. Beams are rarely removed but if unavoidable, the slabs supported by 

the beams need shoring and the columns supporting the beams may require bracing. 

Construction loads: Loads from construction activities vary and may be either temporary or 

permanent. Temporary loads could include the weight of construction equipment and patterned 

loading on perimeter framing when heavy cladding is temporarily removed. Permanent loads 

could be induced if connection stiffness is modified, such as converting a simple shear tab 
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support into a fully rigid moment connection, and when members are temporarily removed, 

causing forces to redistribute. 

Proprietary Concerns 

Braces used in BRBF are proprietary. There are a limited number of manufacturers of the braces 

used in BRBF. 

8.4.2 Add Concrete or Masonry Shear Wall (Connected to An Existing Steel 

Frame) 

Deficiency Addressed by Rehabilitation Technique 

Moment frames buildings that are insufficient to resist lateral forces or too flexible to control 

building drifts can be strengthened and stiffened by adding shear walls. The shear walls may be 

used alone as the new lateral force-resisting system or in conjunction with the moment frames. 

Description of the Rehabilitation Technique 

Shear walls can add considerable strength and stiffness to a structure. These walls could be 

considered as an alternative to adding braced frames if the existing beams and columns are 

incapable of resisting forces in the braced frame system. Concrete shear walls can be placed 

using conventional formwork or shotcrete can be used instead if skilled operators for placing 

shotcrete walls are available. Masonry shear walls, though typically weaker than their concrete 

counterparts, have the advantage of not requiring formwork when filling the cell cavities with 

grout. 

Design Considerations 

The addition of a shear wall to an existing steel frame forms a composite shear wall system. The 

horizontal shear forces are resisted by the wall elements and the vertical overturning forces are 

primarily resisted by the steel columns that become boundary elements. This system is almost 

always controlled by shear due to the substantial flexural strength provided by the steel column 

boundary elements. Though shear cracking and yielding of a wall is not as ductile as flexural 

hinging at the base of the wall, limiting the wall to small drifts prevents early loss of strength and 

stiffness degradation. Other design issues are as follows: 

Research basis: No references directly addressing the addition of shear walls to moment frame 

buildings have been identified.

Design forces: The forces on the structure could increase significantly when this mitigation 

technique is employed due to the increased stiffness and mass of the walls. The entire structure 

should be reanalyzed, including all components that were previously determined to have 

sufficient capacity to resist the forces on the more flexible structure. In-plane forces due to the 

self-weight of the walls remain in the walls and do not increase demands on other elements. The 

out-of-plane forces, on the other hand, must be transferred through the diaphragms and collectors 

to other walls or frames. The design of the wall foundations depends on the magnitude of the 

forces and the soil properties. If overturning forces are greater than the available counteracting 

building mass or the soil is poor, some type of pile foundation will be necessary; otherwise, strip 

footings may suffice. 
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Wall locations: Preference of shear wall locations should be given to existing bays of moment 

frames to utilize the layout of the existing collectors and the strengths of the existing members 

and connections. If this is not possible and shear walls are added at other locations, the existing 

moment frames should be considered when forces are distributed to the lateral force-resisting 

system. Deficiencies in the moment frames may still require mitigation whether the frames are 

included in the lateral system or not. 

Continuous walls vs. frames with infill walls: A continuous wall that encases the existing frame 

beams and columns, shown in Figures 8.4.2-1 and 8.4.2-2, respectively, provides the most robust 

and simple design. It may be desired to use a system that utilizes independent wall panels that 

appear similar to older style steel frame buildings with infill walls. The detailing and behavior 

would be much different from the older buildings though. The wall panels in each bay do not 

encase the beams and columns, as shown in Figures 8.4.2-3 and 8.4.2-4, respectively. One 

advantage of this system is that it may allow the elimination of shoring by not having to remove 

the metal deck though some slab concrete removal may still be required. Another advantage is 

that beams that have existing penetrations and the elements penetrating the beams would not be 

affected. The main drawback is that this system may be limited to use in buildings with lower 

seismic forces. The wall strengths would be limited by the number of studs that can be installed 

and the capacity of the beam webs. Another option is to encase the beams but not the columns. 

Coupled walls: Beams in existing bays of moment frames that are located between new walls 

will behave like coupling beams, particularly if the bay is relatively short. These beams, subject 

to high shears and moments, require ductile detailing. The level of coupling between the walls is 

a function of the stiffness of the beams, which can be determined by a computer model of the 

system. 

Nonstructural issues: The addition of walls to an existing structure changes the architectural 

character of the building. Walls at the exterior will be visible in buildings with clear glazing. At 

interior bays, walls have to be configured to avoid obstruction of existing corridors, doorways, 

and other building systems. Walls can be exposed and incorporated into the interior architecture 

or hidden in partition walls. The architect should be aware that these walls will be thicker than 

typical partition walls. Encased beams and new collectors affect nonstructural components by 

reducing clear floor heights. These components typically include suspended ceilings, pipes, 

conduits, and ducts. Encased columns reduce usable floor space. Coordination with the 

architect and other trades should not be overlooked or underestimated. 

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 

available, and conducting comprehensive field surveys, the following issues should be noted: 

Connection to existing frame: Positive connection should be provided to the existing frame to 

transfer seismic loads to the walls and to provide overturning resistance. Welded shear studs or 

reinforcing bars are typically used. In much older buildings where cast iron members still exist, 

it is difficult to weld to cast iron due to its high carbon content. Instead, holes for the reinforcing 

steel bars may have to be drilled in the members to rely on direct bearing for the load paths. 
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Figure 8.4.2-1: Cast-in-Place Concrete Wall at Existing Beam 

Wall reinforcing: Reinforcing steel in the walls should meet all ACI 318 (ACI, 2005) 

requirements, including the seismic provisions. Columns with positive connections to the walls 

could be counted as boundary reinforcing steel limited by the strength of the shear connections. 

Congestion issues may be encountered at the ends of the walls due to the presence of the 

columns, confinement steel, and anchorage for horizontal wall reinforcing steel. It may be 

necessary to drill holes through the columns for the horizontal steel. 
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Figure 8.4.2-2: Cast-in-Place Concrete Wall Encasing Existing Column 

Mechanical couplers: Consider using reinforcing steel couplers at areas of congestion. Note that 

diameters of the couplers could be as much as twice that of the bars and must meet ACI cover 

requirements for reinforcing steel. 

Effective wall thickness: At beam flanges, the wall thickness could be reduced considerably, 

which compromises the shear strength of the wall. There are several ways to add the shear 

strength of the wall at these locations. Shear studs can be added to the beams to transfer some 

forces through the beam web. Thickness of the wall can be increased at the beam. Extra shear 

reinforcing steel can be placed in the wall provided that ACI limits for reinforcing steel are not 

exceeded. 
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Figure 8.4.2-3: Discontinuous Wall at Existing Beam 

Wall connections at slabs: To avoid damaging the slab reinforcing steel critical for transferring 

diaphragm forces to the shear walls, the concrete slab could be chipped away without damaging 

the slab steel. This still allows for a monolithic wall construction with a construction joint most 

likely located at the top of slab. If the shear forces in the walls are sufficiently low, the entire 

slab may be preserved and merely roughened at the construction joint. Holes can be drilled in 

the slab to allow vertical reinforcing steel to pass through. 
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Figure 8.4.2-4: Wall at Existing Column 

Frame with infill walls: Since the purpose of this technique is to create a shear wall system, the 

detailing should meet this goal by providing continuous load paths for both the laterally induced 

shear and overturning forces. Welded shear studs or reinforcing bars should be provided along 

all wall panel edges to achieve this. The horizontal shear force transfers from one wall to 

another through the studs on the top and bottom flanges and the webs of the beams at each floor. 

Additional plates could be welded between flanges to reduce stresses in beam webs. The 

columns become the boundary elements of the shear walls. Their effectiveness in providing 

overturning resistance may be limited by the number of shear studs or reinforcing bars that can 

be installed. Consider welding one stud on top of another to increase the embedment depth to 

avoid pullout failure. 

Offset walls: Walls do not necessarily have to be centered on the existing beams and columns. 

By offsetting a wall and using the beam and column webs as the edge of the wall, shoring and 

metal deck removal only has to be performed on one side of the frame. An offset wall also lends 

itself to shotcrete construction, while walls with centered steel members have to be cast-in-place 

since it is not possible to place shotcrete on both sides of a steel member. 
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Cost/Disruption 

Adding concrete walls to an existing building is costly, though some savings can be found in the 

wall construction. Shotcrete walls are typically cheaper and faster to construct than conventional 

concrete walls due to the savings in materials and labor associated with formwork. Cost savings 

can be even greater if shotcrete is applied against an existing wall at a stair or elevator and 

mechanical shafts. CMU walls are generally less expensive than either shotcrete or cast-in-place 

concrete. New foundations are almost always required for new walls and could be extremely 

costly if deep foundations, such as drilled piers, are added. 

Installing new walls is disruptive to the occupants because of the noise and vibrations associated 

with construction. Even if tenants are relocated to parts of the building where the work is not 

being performed, vibrations associated with cutting, chipping, and drilling of concrete can 

transmit through the structure. The disruption can be reduced somewhat if the walls are installed 

at the perimeter. 

Construction Considerations 

The engineer’s involvement during the construction phase is critical during a seismic 

rehabilitation. The design of the retrofit scheme must not neglect the construction phase and 

should consider the issues below at a minimum. 

Shotcrete walls: The quality of a shotcrete wall is highly dependent on the skill of the nozzle 

operator. Building codes typically require preconstruction test panels constructed and reinforced 

similarly to the actual walls. This allows for inspection of the finished product, sawcutting to 

verify the quality of the shotcrete, and coring to determine strength. Additional test panels for 

overhead joints should also be requested to allow for cores to be taken to inspect the surface 

preparation and the joint bond. A series of ACI 506 publications provide general information, 

specifications, certification of nozzle operators, and evaluation of shotcrete (ACI, 1991, 1994, 

1995a, 1995b, 1998). 

Congestion: Congestion issues may be encountered at the ends of the walls due to the presence 

of the columns, confinement steel, and anchorage for horizontal wall reinforcing steel. Consider 

using mechanical couplers in heavily reinforced boundary elements and walls where lap splices 

are impractical. 

Concrete/shotcrete placement: The type of construction should be determined during the design 

phase because details for cast-in-place concrete and shotcrete construction are different. It may 

not always be possible to confidently use either type of construction; and, thus, the details 

creating such situations should be avoided. During construction, some locations require 

particular attention when shotcrete and concrete are being placed to ensure that all gaps are filled 

– tops of wall panels, k-region of steel sections, and construction joints. Figure 8.4.2-5 shows a 

pilaster at a steel column where shotcrete shadowing restrictions require the use of cast-in-place 

construction for the pilaster but still allow shotcrete construction away from the pilaster. See 

Section 21.4.5 for additional discussion concerning concrete and shotcrete construction. 
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Removal of existing nonstructural elements: Exterior cladding and interior partitions may have to 

be removed to deliver concrete formwork and other equipment. Walls that encase beams at the 

roof require temporary removal of the roofing and waterproofing. Installation of the walls will 

affect ceilings, lights, and other mechanical/electrical/pluming components. Nonstructural 

elements located within the frame bays being strengthened have to be moved for construction of 

the walls. Permanent relocation may be desired for some of these elements to minimize 

openings in the new walls. Older buildings with items of historical significance require 

additional coordination and effort so that these items are not damaged when temporarily 

removed and are restored properly. 

Removal of existing structural elements: See general discussion in Section 8.4.1. 

Construction loads: See general discussion in Section 8.4.1.

Proprietary Concerns 

Mechanical couplers for reinforcing steel are proprietary. 

Figure 8.4.2-5: Combined Shotcrete and Cast-In-Place Construction 
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8.4.3 Add Steel Cover Plates or Box Existing Steel Member 

Deficiency Addressed by Rehabilitation Technique 

Frame members that are inadequate to resist the seismic demands are strengthened with cover 

plates or by adding side plates to W-shapes to create box sections. This reduces axial and 

flexural stresses in beams and columns and could also be used to increase the shear strengths of 

these members. 

Description of the Rehabilitation Technique 

Cover plates are welded to the outside of existing flanges, which in effect, increase the flange 

areas. When strengthening a beam, a cover plate is more commonly attached only to the 

underside of the beam since the presence of an existing diaphragm makes it difficult and costly 

to weld to the top of the beam or attach side plates to box the beam. However, for strengthening 

non-composite beams, top cover plates are virtually indispensable. Conversely for columns, 

plates are typically welded to the sides of W-shapes because it is much more effective to increase 

the axial and flexural capacities by converting a column to a box section and the shear capacity 

by essentially providing two additional webs. 

Design Considerations 

Thorough knowledge of the existing material behaviors and strengths are necessary for the new 

and existing elements to interact in the desired manner. If welding to the existing components, 

the carbon equivalent of these components need to be verified through as-built records or new 

testing to determine their weldability; see FEMA 356 (FEMA, 2000e) for further discussion of 

this issue. Other design issues include the following: 

Research basis: No references directly addressing the addition of steel cover plates to existing 

frame members have been identified.

Beams: The flexural strength of a beam can be improved by welding cover plates to the bottom if 

there is a composite slab present. If there is only a bare metal deck, a cover plate on only one 

side of the beam may not be very effective. However, it could be useful for strengthening beams 

with large axial forces, primarily collector members. 

Columns: The overall capacity of a column is determined by axial-flexural interaction. Boxing a 

column decreases its slenderness and, therefore, increases its axial and flexural capacities. 

Whether the areas of the new plates can be directly included in computing these capacities 

depends on their continuity and detailing at a beam-column joint. Except for one side of the 

exterior columns, beams framing into the columns at each floor will disrupt the continuity of the 

new plates. 

Foundations: Where cover plates are added to the columns at their base, a reevaluation of the 

foundation system is warranted. It is not uncommon for frame columns to develop plastic hinges 

at their bases and thus, the increase in demand on the foundation may be greater than intended. 

The strength and stiffness of the base plate and the anchor rods should be evaluated and 

upgraded accordingly. A more in-depth discussion of column connections to the foundation is 
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provided in Section 8.4.5. Foundation upgrades are not common with this technique, but if 

required, refer to the chapter on foundations. 

Nonstructural issues: The beam and column upgrades hardly take up any additional space but 

minor modifications and temporary relocations may be required for some architectural and 

M/E/P elements.

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 

available, and conducting comprehensive field surveys, the following issues should be noted: 

Beams: When a section of beam is strengthened, the cover plate typically attaches to the bottom 

flange with fillet welds, as shown in Figure 8.4.3-1. The length and size of the welds determine 

the shear flow between the plate and the existing beam, similar to interactions between other 

composite elements. 

Columns: Side plates used to create a box section can be attached to the column flanges with 

CJP, PP, or fillet welds, as shown in Figure 8.4.3-2. The fillet weld is the simplest to implement 

since it does not require additional preparations like the other welds, particularly the use of 

backing plates and beveling of the welded edges. The type of connection used may limit the 

shear flow between the plates and the existing column. 

Existing floors: Whether the columns are upgraded with cover plates or box sections, the new 

plates have to terminate at the existing beams and slabs. Though it is possible to achieve 

continuity by using additional plates or sections if necessary, this would create detailing and 

construction complexities and should be avoided. Furthermore, excessive welding around a 

beam-column joint could create undesirable residual stresses in the joint. 

Cost/Disruption 

Schemes that involve slab removal, work around a connection, and foundation work are costly. 

As typical with seismic upgrades, cost and disruption is minimized when schemes are kept 

simple. 

Construction Considerations 

See Section 8.4.1 for general discussion of welding issues, removal of existing nonstructural and 

structural elements, and construction loads. 

Proprietary Concerns 

There are no known proprietary concerns with this technique. 
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Figure 8.4.3-1: Cover Plate at Existing Beam 

Figure 8.4.3-2: Box Section at Existing Column 
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8.4.4 Provide Collector in a Concrete Fill on Metal Deck Diaphragm 

Deficiency Addressed by Rehabilitation Technique 

Beams that are inadequate to transfer collector loads are strengthened. 

Description of the Rehabilitation Technique 

Plates can be welded to various portions of the existing beam. Cover plates, discussed in Section 

8.4.3, are commonly attached to the underside of beams. To attach plates to top flanges, it may 

be simplest to orient the plates vertically and weld to the underside of the flanges, as shown in 

Figure 8.4.4-1, or similarly on top of the bottom flanges. Given all the possible options, the plate 

locations may be dictated by the eccentricities, continuity options at a beam-column joint, or the 

presence of nonstructural elements. 

Design Considerations 

Though the purpose of this technique is to strengthen a beam axially, it also changes the flexural 

properties of the beam by increasing its stiffness. It should be verified that this does not have 

unintended consequences, such as converting the beam into a frame member. Eccentricities in 

collectors are typically neglected. However, deep collectors and/or collectors with large forces 

could be subject to significant moments from these eccentricities. The eccentricities could be 

reduced if plates are attached to the top flanges but requires other detailing considerations at a 

beam-column joint.

Detailing Considerations 

The new plates should attach to the columns with complete joint penetration (CJP) welds. 

Consider replacing the welds in the existing collector with high notch toughness welds in a high 

seismic region. Continuity plates directly aligned with the collector elements at a beam-column 

joint should be used as much as possible but may be offset if all eccentricities and their effects on 

the joint are considered. 

Cost/Disruption 

This technique is relatively inexpensive compared to other types of steel frame upgrades. The 

most expensive part of the structural upgrade is associated with the work at the beam-column 

connections. The nonstructural disruptions to the architectural and M/E/P elements are typical. 

Construction Considerations 

See Section 8.4.1 for general discussions of welding issues, removal of existing nonstructural 

and structural elements, and construction loads. 

Proprietary Concerns 

There are no known proprietary concerns with this technique. 
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Figure 8.4.4-1: Plate Collectors at Existing Beam 

8.4.5 Enhance Connection of Steel Column to Foundation 

Deficiency Addressed by Rehabilitation Technique 

Frame columns are subject to axial (including possible tension), flexural, and shear forces. To 

this end, columns with inadequate anchorage to the foundation limit the capacity of a frame. The 

columns could be part of an existing lateral force-resisting system that do not meet current 

standards or part of an upgraded system with larger forces resulting from increased stiffness. 
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Description of the Rehabilitation Technique 

Two methods are common for enhancing the column connection to the foundation. First, 

modifications at the base plate could include the addition of anchor rods, welding shear lugs to 

the base plate, and/or enlargement of the base plate. The other method is to encase the column in 

a concrete pedestal. It is also possible to use both of these methods together. The foundation 

system itself is not addressed here but is discussed in Chapter 23. 

Anchor rods can be used to resist tensile forces due to uplift and flexure. Holes drilled in the 

existing footing for these rods can be filled with a nonshrink high strength grout or chemical 

adhesive. Shear lugs may be used to transfer shear forces into the foundation. If the existing 

base plate is not large enough to accommodate the new rods or lugs, new plates can be welded to 

the existing plate to enlarge its area. This is also necessary if the allowable bearing stress is 

exceeded due to increased column compression. The increased base plate size leads to greater 

shear and flexural forces in the plate. It is likely that the plate will not be thick enough to resist 

these forces. Thus, stiffeners that act as supports can be welded to the base plate to reduce the 

plate forces. The stiffeners also provide additional load paths from the column to the base plate. 

Figure 8.4.5-1 shows some of these modifications. 

Base plate modifications may not always be practical or possible. Instead, the column can be 

encased in a concrete pedestal above the footing, as shown in Figure 8.4.5-2. Shear forces would 

transfer through direct bearing of the column against the pedestal. The pedestal could also be 

used to transfer uplift and flexural forces by relying on the existing base plate and other 

mechanisms, such as welded shear studs along the column. The pedestal itself should be detailed 

as a reinforced concrete column that meets all ACI requirements and seismic provisions. 

Design Considerations 

Thorough knowledge of the existing material behaviors and strengths are necessary for the new 

and existing elements to interact in the desired manner. A complete presentation of column base 

plate design can be found in AISC Design Guide 1 (Fisher and Kloiber, 2005). Key design 

issues include the following: 

Research basis: No references directly addressing the upgrade of connections of steel columns to 

foundations have been identified. 

Shear lugs: Columns that are in compression offer shear resistance through friction below the 

base plate. AISC recommends different friction coefficients depending on the location of the 

base plate with respect to the top of concrete. Additional shear resistance is required when the 

compressive force is not great enough or there is tension in the column. Though not common in 

retrofit applications, shear lugs, or simply plates welded to the bottom of a base plate, have the 

advantage of shallow embedment compared to anchor rods. The lugs also receive confinement 

from the base plate above and simplify the design by allowing anchor rods to only resist tension. 

The design of a shear lug is a function of the allowable bearing stress in the surrounding grout 

and flexural forces in the lug. 

Anchor rods: Tension in an anchor rod is developed through bond along its length and/or direct 

bearing through a hook, bolt head, or nut at the end of the rod. Though it is possible to drill a 
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hole large enough in an existing footing to accommodate a bolt head or nut, manufacturers of 

grouts and adhesives sometimes limit the size of the a hole for a given rod diameter in which its 

product can be used as an infill. In this case, the tensile force has to be developed entirely 

through bond. For bond development, the rod has to be threaded or some other means of 

mechanical anchorage has to be provided along the rod. Anchor rods are not typically used to 

transfer shear to the foundation because the mechanism for shear transfer is difficult to define 

and still subject to debate. If unavoidable, several factors should be considered, including 

bending of the rod through the oversized hole in the base plate and shear and tension interaction 

on the rod. If the column shear force is being distributed between both new and existing rods, 

note that the holes for the existing rods are likely to be oversized and welded plate washers 

should be verified or provided. 

Figure 8.4.5-1: Modified Base Plate to Increase Uplift Capacity 
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Figure 8.4.5-2: Concrete Pedestal at Existing Column 
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Stiffeners: The simplest way to reduce flexural stresses in a base plate is to add stiffeners that 

essentially provide multiple supports along the base plate. These stiffeners also assist in 

transferring uplift and flexural forces to the base plate and reduce the weld stresses at the base of 

a column.

Nonstructural issues: Base plates are typically hidden in the slab and thus, the upgrade could 

remain hidden if it only involves adding anchor rods. Pedestals, on the other hand, require 

additional space and could present aesthetic issues. However, basements are often used for 

parking, storage, or as equipment locations, where aesthetic considerations may be negligible.

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 

available, and conducting comprehensive field surveys, the following issues should be noted: 

Base plates: If the existing base plate can accommodate new anchor rods or lugs, burning holes 

in the plate is usually acceptable. Otherwise, enlargement of an existing base plate is 

accomplished by welding new plates to the existing plate. Only partial penetration (PP) welds 

can be used since the lack of backing precludes using complete joint penetration (CJP) welds 

unless concrete is removed to allow placement of backing plates. The depth of the PP weld 

limits the effective thickness of a plate. Also of note is that AISC relaxes its typical edge 

distance provisions if there is no lateral load on a base plate. The only recommendation is that 

enough edge distance remains such that the drill or punch does not drift when a hole is made. 

AISC Design Guide 1 suggests that one quarter of an inch is enough to meet this condition. 

Anchor rods: Detailing differs depending on if an anchor rod is designed to resist small shear 

forces and tension or tension only. For the latter, an oversized hole in the base plate and a thick 

plate washer over the hole is considered adequate. Since oversized holes are intended to allow 

for inaccuracies in anchor rod placement when concrete footings are placed, it should be possible 

to use smaller holes if the holes are being drilled. In addition to the design considerations 

mentioned above, anchor rods with shear forces require welding heavy plate washers to the base 

plate. The washers should have close-fit holes to minimize the movement required to engage all 

of the rods together. Nuts are sometimes welded to the washers but the mechanical properties of 

the nuts could lead to welding complications. 

Pedestals: Reinforcement of the pedestal is similar to a reinforced concrete column. Vertical 

bars provide tensile and flexural strength while hoops or ties provide confinement. At the top of 

footing, the shear is transferred through shear friction in the anchored reinforcing steel dowels. 

The concrete surface should be roughened, which also increases the shear friction values. If the 

pedestal is also used to transfer uplift and flexural forces, additional dowels are required for 

tension. A single dowel should not be considered to provide both shear and tensile resistance. 

One mechanism to transfer uplift and flexural forces from the column to the pedestal is the use of 

welded shear studs along the column. The size and spacing of the shear studs may determine the 

minimum pedestal diameter and height. 

Concrete anchorage: Installation of the anchor rods and dowels requires drilling into the existing 

footing. Consider using a scanning device to avoid damaging the reinforcing steel in the top 
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layer of the footing. Cores may have to be taken to confirm the concrete strength. Many 

commercial grouts and adhesives are available for bonding the anchor to the concrete. Verify 

that the selected product is appropriate for the specific anchor in a seismic application. On site 

inspection and testing of the anchors are mandatory since their performance rely heavily on the 

installation process. The grout and adhesive product vendors’ ICC Evaluation Service reports 

provide standardized installation procedures and anchorage capacities. 

Cost/Disruption 

The cost of a column connection to foundation upgrade is not very expensive. However, this 

technique is not usually performed only by itself. Costs would be small relative to an overall 

lateral force-resisting system upgrade and a foundation upgrade. Disruption could be minimal 

since typically, there are no tenants in the basement. Even if the basement were used for tenant 

access, such as parking, only a few columns would be affected at a time. 

Construction Considerations 

The engineer’s involvement during the construction phase is critical during a seismic 

rehabilitation. The design of the retrofit scheme must not neglect the construction phase and 

should consider these issues at a minimum: 

Welding issues: See general discussion in Section 8.4.1. 

Removal of existing nonstructural elements: This technique tends to be less disruptive of 

nonstructural functions compared to other techniques. However, there may still be some 

architectural and M/E/P elements, such as partitions and pipes adjacent to the columns, that 

require temporary removal. See Section 8.4.1 for discussions of fireproofing, asbestos, and 

concrete encasement. 

Removal of existing structural elements: To access the base plate, an existing slab and slab 

reinforcement will probably have to be removed and replaced. Care should be taken to not 

damage any of the existing structural elements. If the existing base plate is being replaced 

entirely, a column shoring scheme has to be devised. The existing anchor rods should be cut at 

the top of footing and the new rods have to be relocated. 

Construction loads: See general discussion in Section 8.4.1. Construction loads at the basement 

level are not typically a concern if a slab-on-grade is present. 

Proprietary Concerns 

Many grout and adhesive products are available. 

8.4.6 Enhance Beam-Column Moment Connection 

Deficiency Addressed by Rehabilitation Technique 

Riveted, bolted, and WSMF connections are upgraded to improve their ability to withstand 

inelastic rotational demands and develop the plastic moment capacity of the beams. 
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Description of the Rehabilitation Technique 

The techniques discussed in this section were developed specifically to address pre-Northridge 

WSMF connections. These techniques can be adapted with prudence to existing riveted and 

bolted connections that are found to be inadequate or to upgrade partially restrained connections. 

These techniques are covered thoroughly in AISC Design Guide 12 and FEMA 351 and only 

briefly presented here. The reduced beam section (RBS) is the only technique that weakens the 

beam in flexure, which in turn, moves the plastic hinge away from the column and reduces the 

demand on the complete joint penetration (CJP) welds. Two other methods - welded haunch and 

bolted bracket – also move the hinge away from the column but strengthen the existing 

connection and seek to maintain the original flexural capacity of the beam. A similar method 

employs cover plates over the beam flanges, requiring little additional space. This method is not 

discussed in detail here but more information can be found in FEMA 351. Additional 

modifications that should be performed for each technique include adding or verifying the 

capacity of beam flange continuity plates across the column web and strengthening the panel 

zone; see the references mentioned above and AISC Design Guide 13 (Carter, 1999) for 

reference. 

The selection of a particular connection modification depends on specific project factors. There 

are advantages and disadvantages to each of the methods. While each of the three connections 

discussed below consistently developed a minimum plastic rotation of 0.02 radian in cyclic 

loading experiments, the welded haunch and bolted bracket demonstrated higher levels of 

performance and reliability (Gross et al., 1999). On the other hand, the RBS modification 

requires no additional space and reduces the beam capacity to enforce a strong-column weakbeam condition. The welded haunch is the only modification that exhibited desirable behavior in 

tests where the beam top flanges were left in a pre-Northridge condition (Gross et al., 1999). 

The bolted bracket scheme requires top flange reinforcement but eliminates field welding. 

Design Considerations 

Connection modifications merely ensure that the connections behave as originally intended in a 

moment frame. These modifications by themselves do not reduce the frame forces in an 

earthquake. Other structural upgrades are necessary if the frame does not have sufficient global 

strength or stiffness to resist the demands. 

Research basis: Numerous experimental programs have been performed under the auspices of 

the SAC Joint Venture, the National Science Foundation, the National Institutes of Standards and 

Technology, and AISC. Most of the results of these tests have been summarized in AISC Design 

Guide 12 and the FEMA documents listed in the references. The user is cautioned to extrapolate 

beyond the conditions that have been tested only with proper considerations of the differences 

between the tested connections and the building conditions.

Design forces: AISC enforces the strong-column weak-beam concept by specifying a minimum 

column-beam moment ratio in the AISC Seismic Provisions. The general intent is to prevent 

plastic hinges from developing in the columns, which could lead to a soft story. Some yielding 

may still occur in the columns without causing loss in frame strengths. The required flexural 

strength of a column is determined from several variables associated with the beam. The beam 

plastic moment at the critical plastic section – centerline of RBS or tip of haunch and bracket – 
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includes a strain hardening factor and the expected yield stress of the flanges. An additional 

moment is a function of the shear in the beam and the distance from the critical plastic section to 

the column centerline. This shear is a combination of the shear associated with the plastic 

moment and the gravity loads. 

RBS: The design of moment frames is often governed by global and interstory drift limits instead 

of strength requirements. Thus, the reduction in beam strength and stiffness from an RBS could 

be acceptable within practical limits. If the presence of an existing slab poses construction and 

design issues, then the top flange may remain intact and only the bottom flange has to be cut; 

though minimal removal of the existing slab is still required to replace the existing top flange 

weld. Note that a bottom flange modification only achieves a minimal stress reduction and that 

both the top and bottom flange CJP welds still require replacement with high toughness weld 

metal. Various RBS cut shapes have been tested successfully, though the radius cut RBS 

minimizes stress concentrations. The critical dimension is the depth of the cut, which can also be 

expressed in terms of flange width reduction as a percentage. AISC recommends a maximum 

flange reduction of 50% as a practical limit for field modifications. To determine if a flange 

reduction is adequate, the maximum moment at the face of the column can be computed for the 

modified beam. If the ratio of this moment to the plastic moment of the beam exceeds 1.05, a 

bottom flange RBS modification is not recommended. Either a top flange RBS should also be 

provided or another method should be used. 

Welded haunch: A tapered haunch welded to a beam bottom flange changes the force transfer 

mechanism at the connection. Analytical and experimental studies have found that the beam 

shear transfers through the haunch flange (FEMA, 2000b). This, in turn, reduces the shear 

stresses in the flange welds and increases the depth of the column panel zone. The welded 

haunch is the only connection out of the three presented for this technique to reach plastic 

rotations on the order of 0.03 radian while allowing the top flange weld to remain in a preNorthridge condition in tests. Additional rotational capacity can be obtained by upgrading the 

top flange groove weld to a higher toughness, adding a cover plate, or adding another haunch, in 

order of increasing performance. The two latter modifications can be performed without 

replacing the top flange weld. 

Bolted bracket: Considered an alternative to the welded haunch, this option does not require any 

field welding. A portion of the flange force is transferred to the bracket and reduces the weld 

stresses. Several bolted bracket configurations are possible and can be applied to either beam 

flange or both flanges for large beams. Common types of brackets include the haunch bracket, 

pipe bracket, angle bracket, and double angle bracket. Pipe and angle brackets are relatively 

compact and can be covered up by the slab for a top flange application. For bottom flange 

haunch only schemes, a stiff angle is still recommended at the top flange (Gross et al., 1999). 

The pre-Northridge weld in the top flange can remain in this case even if it is found to have 

defects. 

Material strength: Yield strengths of existing materials are best determined from tests of samples 

taken from the actual members. Samples taken from flanges are preferred over the web. When 

samples are not available, AISC Design Guide 12 prescribes some overstrength values for 

different grades of steel. 
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Minor axis connections: Moment connections to the minor axis of a column are not as common, 

but sometimes present in older moment frame construction. Their performance in the Northridge 

earthquake did not result in significant damage. Thus, little work has been done to study the 

rehabilitation of these connections. Similar concepts presented in this section may be used to 

rehabilitate these connections with the recognition that the columns may be weaker than the 

beams, which would also limit the inelastic demands on the connections. Note the governing 

jurisdiction may require qualification through the testing program described in FEMA 351. 

Nonstructural issues: Welded haunches and bolted brackets are considerable in size. The 

recommended welded haunch and bolted bracket depths are approximately one third and one half 

times the beam depths, respectively. For a W36 beam used in a moment frame, this adds up to 

18 inches to the connection depth. This makes the option of adding components to only the 

bottom flange more attractive, though it could still interfere with the ceiling. At the top flange, 

the modification would certainly have to be hidden or incorporated into an architectural feature.

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 

available, and conducting comprehensive field surveys, the following issues should be noted: 

Weld filler metal matching and overmatching: Weld filler metals with greater tensile strength 

than the connected steel should be used. Flux cored arc welding and shielded metal arc welding 

electrodes that conform to E70 specifications exhibit overmatching properties compared to 

common steel specifications, including ASTM A36, A572 (Grade 42 and 50), A913 (Grade 50), 

and A992 (FEMA, 2000b). 

Weld metal toughness: Tests on existing connections modified with RBS but no modifications to 

the existing beam flange welds showed poor performance (Gross et al., 1999). Improved 

connection performance was achieved for connections where the top and bottom flange welds 

were replaced with higher toughness weld metal. AISC Seismic Provisions now require weld 

metals with minimum Charpy V-Notch (CVN) toughness of 20 ft-lbs at -20oF and 40 ft-lbs at 

70oF for demand critical welds. 

Weld backing and access holes: Backing plates provided for flange welds create a notch effect 

and also hinder detection of weld flaws at the weld root. Backing should be removed at flanges 

receiving new CJP groove welds and reinforced with fillet welds. The removal of existing 

backing without weld replacement is an ineffective upgrade since the existing weld is still likely 

to have low toughness. The size and shape of the weld access hole should be configured to 

provide welder access and minimize stress concentrations. FEMA 351 provides a specific weld 

access hole detail that meets these criteria. 

RBS (Figure 8.4.6-1): The distance from the face of column to the start of the RBS cut and the 

length of the cut, more concisely known as the offset and the chord length, respectively, should 

be kept small as to not allow the moment to increase significantly from the plastic hinge to the 

column. Yet, the distance to the start of the cut should also be large enough to allow for the 

flange force from the RBS to be uniformly distributed across the flange at the CJP weld. The 
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length of the cut should also be long enough to control the inelastic strains along the RBS. The 

radius of the cut is a function of its depth and the chord length. See AISC Design Guide 12 for 

specific recommendations and guidelines for selecting these dimensions. 

Figure 8.4.6-1: Reduced Beam Section at Bottom Flange of Existing Beam 

(adapted from AISC Design Guide 12) 

Welded haunch (Figure 8.4.6-2): Haunches can be cut from WT- or W-shapes. Typically, the 

haunch flange is groove welded to the beam and column flanges while the haunch web is fillet 

welded to these elements. Experimental programs have consistently used the same haunch 

geometry—length and depth approximately one half and one third that of the beam depth, 
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Figure 8.4.6-2: Welded Haunch at Bottom Flange of Existing Beam 

(adapted from AISC Design Guide 12) 

respectively. The haunch flange is sized to resist most of the shear force from the beam. The 

haunch web size can be established to achieve equilibrium. At the intersection of the beam and 

haunch flanges, web stiffeners should be provided for the beam. Similarly at the column, add 

continuity plates to align with the haunch flange. 

Bolted bracket: The preferred configuration for this modification is a haunch bracket at the 

bottom flange and an angle bracket at the top flange that can be hidden within the slab. If 

additional reinforcement is necessary at the top flange, angles can be added below the top flange 
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on each side of the web, as shown in Figure 8.4.6-3. Close-fit holes minimize slip in this 

connection. Tests have demonstrated that these connections are essentially fully rigid and can 

reach large plastic rotations while allowing the pre-Northridge welds at the flanges to remain 

(Gross et al., 1999). AISC recommends neglecting the existing welds and designing each 

bracket for the entire flange tension force. The bottom flange haunch bracket will be slightly 

larger than the welded haunch for the same beam size. The top flange angle bracket may require 

several rows of bolts along the beam flange and has to be designed for the prying force on the 

column flange. Due to limitations in the available L-shapes, the angle bracket may have to be 

cut from a W-shape. 

Cost/Disruption 

Connection modifications are locally very disruptive. Modification of a connection typically 

requires access from two floors to perform the work on each flange. Noise associated with this 

type of work will spread and disrupt tenants on other floors unless the work is done during offhours. The RBS is probably cheaper than the other two modifications since it requires the least 

amount of material and labor. With older buildings, there may be asbestos present in the 

fireproofing around the steel members, which could require costly abatements to expose the 

connections. 

Construction Considerations 

The engineer’s involvement during the construction phase is critical during a seismic 

rehabilitation. The design of the retrofit scheme must not neglect the construction phase and 

should consider these issues at a minimum: 

Welding/bolting issues: See general discussion in Section 8.4.1. The bolted bracket modification 

can be performed without any field welding. Primary issues associated with the bolted bracket 

consist of typical field bolting issues such as set up, fit-up, and alignment. 

Removal of existing nonstructural elements: The connection work will affect ceilings, lights, and 

other mechanical/electrical/pluming components. Connection modifications at the roof level 

may warrant removal of the roofing and waterproofing. See Section 8.4.1 for discussions of 

fireproofing, asbestos, and concrete encasement. 

Removal of existing structural elements: See general discussion in Section 8.4.1. 

Construction loads: See general discussion in Section 8.4.1. 

Proprietary Concerns 

There are several proprietary connections that have been developed to upgrade pre-Northridge 

connections, generally sharing similar design intents. Some of these connections briefly 

presented in FEMA 351 include the Side Plate connection system, the Slotted Web connection, 

and one particular type of the bolted bracket connection. The reader should contact the licensors 

of these technologies for more information. 
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8.4.7 Enhance Column Splice 

Deficiency Addressed by Rehabilitation Technique 

Strengthen welded or bolted column splices that do not meet the detailing and minimum design 

strength requirements in AISC Seismic Provisions. 

Description of the Rehabilitation Technique 

Numerous options exist for upgrading a column splice. The approach and the level of 

strengthening depend on the type of lateral system since special moment frames have different 

requirements from other systems in the AISC Seismic Provisions. It also depends on if the 

existing splice is welded or bolted and if field welding is permitted. Field welding may be a 

necessity in some cases since the AISC Seismic Provisions do not allow bolts and welds to share 

loads on the same faying surface. 

Most existing welded splices are likely to be complete joint penetration (CJP) welded or partial 

joint penetration (PP) welded. CJP welds require beveled transitions to avoid stress 

concentrations. A beveled transition can be constructed in the field by building up the weld over 

the thicker column flange or grinding away a portion of the flange. PP welds inherently possess 

stress concentrations at the unwelded portion of the joint and thus, have higher strength 

requirements, but do not require beveled transitions. PP welds can be strengthened by welding 

plates or stiffeners across the splice. When field welding is impractical or undesirable, it may be 

possible to use bolted plates at an existing welded splice if the bolts are designed to resist the 

entire load. 

The rehabilitation method for a bolted splice depends on the controlling design mode for the 

splice – whether the splice is governed by net section fracture of the column, yielding of the 

bolts, or gross yield or net section fracture of the splice plate. Bolts that govern the existing 

splice capacity could be replaced with stronger bolts, such as replacing A325 with A490. The 

upgrade may also be as simple as replacing bolts that have threads included in the shear plane 

with longer bolts. Alternatively, more bolts or larger bolts could be provided, but both would 

require more extensive field work. Splice plates with insufficient strength can be replaced or 

new plates could be provided on the opposite side of the existing plate, which would require 

longer bolts. The existing plates may also be extended by welding new plates to the ends. 

Design Considerations 

Considerations for a column splice include the justification of a load path, preservation of 

symmetry to avoid eccentric loads, and deformation compatibility if welds and bolts are both 

used. In addition to the requirements in the AISC Seismic Provisions, the AISC Steel 

Construction Manual (AISC, 2005c) has a general discussion of column splices and contains 

typical details of W-shape splices. Information can be found regarding filler plate sizes for 

different column depths and erection tolerances. Important design issues include the following: 

Research basis: No references directly addressing the upgrade of existing column splices have 

been identified.
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Design strength: The AISC Seismic Provisions specify a minimum strength requirement for all 

splices as a function of the expected yield strength and the flange area of the column. In 

addition, PP welded splices have higher strength requirements due to the stress concentration 

manifested in the crack-like notch at the unwelded side of a flange. Lastly, column splices in 

special moment frames have to be designed for the full flexural strength of the smaller column 

and their web splices require a shear strength of twice the plastic flexural strength of the column 

divided by the story height. 

PP welded splices: For flange thickness differences between the columns up to 1/8”, steel shims 

can be fillet welded to the inside of the flanges. Greater than 1/8”, it may be more practical to 

attach a plate to the flanges using CJP welds. 

Plate locations: Plates can be extremely versatile whether they are welded or bolted to the 

columns. Bolted plates are commonly placed at the outside of flanges, where extra filler plates 

can be used to make up differences in flange thicknesses. Filler plates are not necessary at the 

inside of flanges, where a flush surface is already provided if the same nominal depth columns 

are present. Plates can also be welded from one flange tip to another, which provides increased 

shear strength and stability across the joint. See Figure 8.4.7-1. 

Nonstructural issues: The column splices hardly take up any additional space but minor 

modifications and temporary relocations may be required for some architectural and M/E/P 

elements.

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 

available, and conducting comprehensive field surveys, the following issues should be noted:

Welded splices: Large shrinkage strains could develop when welding heavy sections. Bolted or 

fillet welded plate slices should be considered as an alternative if there is a possibility of a brittle 

weld fracture. 

Bolted splices: If new holes are drilled in the existing column or splice plate, tolerances of the 

existing holes should be verified to ensure that bolts will be loaded evenly. The capacity of the 

net section should also be checked. See Figure 8.4.7-2 for a sample detail. 

Web splices: Column webs in moment frames that use bolted web splices require plates or 

channels on both sides to reduce eccentricities. This should be considered for column webs in 

other lateral force-resisting systems though it is not a specific requirement in the AISC Seismic 

Provisions. 

Cost/Disruption 

The cost of implementing this technique is highly dependent on the level of modification 

performed to the existing splice. The level of disruption is typical for that of steel frame 

upgrades, involving vacancy of space, removal of all nonstructural elements around the column, 

and significant noise. 
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Figure 8.4.7-1: Welded Splice Upgrade at Existing Column 

Construction Considerations 

The engineer’s involvement during the construction phase is critical during a seismic 

rehabilitation. The design of the retrofit scheme must not neglect the construction phase and 

should consider these issues at a minimum: 

Welding issues: See general discussion in Section 8.4.1. Though weld fractures were not found 

at column splices after the Northridge earthquake, an environment in which the welder can 

perform quality welds is critical. 

Removal of existing nonstructural elements: See general discussion in Section 8.4.1. Depending 

on the location of the splice, the upgrade could affect ceilings, lights, and other 

mechanical/electrical/pluming components. 
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Figure 8.4.7-2: Bolted Splice Upgrade at Existing Column 

Removal of existing structural elements: Due to the critical nature of columns, the removal of 

existing welds or bolts at a column should be minimized. Column alignment and stability should 

be maintained at all times. 

Construction loads: See general discussion in Section 8.4.1. Typically, welding on a loaded 

column should not create a safety issue, although stability during construction should always be 

considered. At a minimum, see section in AISC Steel Construction Manual (2005c) on column 

splices and the AISC Code of Standard Practice for Steel Buildings and Bridges (AISC, 2005a) 

for other construction considerations. 

Proprietary Concerns 

There are no known proprietary concerns with this technique. 
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8.4.8 Add Steel Plate Shear Wall (Connected to An Existing Steel Frame) 

Deficiency Addressed by Rehabilitation Technique 

Moment frames buildings that are insufficient to resist lateral forces or too flexible to control 

building drifts can be strengthened and stiffened by adding steel plate shear walls (SPSW), as 

shown in Figure 8.4.8-1A. The shear walls may be used alone as the new lateral force-resisting 

system or in conjunction with the moment frames. 

Description of the Rehabilitation Technique 

Shear walls can add considerable strength and stiffness to a structure. It should be considered as 

an alternative to adding braced frames if additional stiffness is required. Compared to concrete 

shear walls, steel plate shear walls are lighter and add less seismic mass to a structure. They also 

take up less space and may be more economical to construct, particularly in taller buildings 

where the costs of delivering formwork and pumping concrete are significant. The behavior of 

this system is analogous to braced frames that rely on tension-only braces, as well as plate 

girders whereby tension fields develop along the diagonals. However, neither of these examples 

completely characterizes the behavior of an SPSW. The beams and columns in an SPSW behave 

as boundary elements that are subject to a complex array of forces and require a considerable 

amount of stiffness to develop the capacity of the steel panels. Large strut forces are imposed on 

the beams while columns are subject to axial, flexural, and shear forces. Thus, preference of 

shear wall locations should be given to existing bays of moment frames, in which the members 

and connections are less likely to require modifications for use in the SPSW system. The energy 

dissipating mechanisms in an earthquake include tension yielding and eventual tearing of the 

diagonal tension fields, shear yielding of the plates, compressive buckling of the plates along the 

diagonal compression fields, and slipping of the bolted connections to the fin plates when used. 

In an existing building, fin plates at the wall boundaries would be field welded to the beams and 

columns first. Large steel panels would then be welded or bolted to these fin plates, as illustrated 

in Figure 8.4.8-1B. Panel splices could also be welded or bolted in the field. If welded splices 

are used, it is recommended to provide full penetration welds with the backing bars removed 

after welding. Openings are acceptable if stiffeners are provided around the edges. The panels 

themselves may be unstiffened or stiffened. Provisions for the design of this system have been 

included in the 2005 AISC Seismic Provisions. AISC is also in the process of publishing a 

design guide for unstiffened SPSW. 

Design Considerations 

Adding shear walls to a moment frame building increases its stiffness considerably. The 

upgraded structure should be evaluated for higher lateral and overturning forces accordingly. 

This system is almost always controlled by shear due to the substantial flexural strength provided 

by the steel column boundary elements. 

Research basis: No references directly addressing the addition of SPSW to moment frame 

buildings have been identified. However, if the inelastic deformations can be limited to the steel 

plates, the seismic performance of the strengthened structure is not expected to differ from new 
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Figure 8.4.8-1A: Unstiffened Steel Plate Shear Wall 

buildings utilizing SPSW. A summary of the major research on this system as well as its seismic 

performance in past earthquakes can be found in Astaneh-Asl (2001).

Unstiffened vs. stiffened walls: Stiffened walls tend to exhibit higher shear strength though both 

types of walls can be expected to exhibit ductile behavior. Buckling of the steel plates in 

unstiffened walls allow tension fields to develop and resist the lateral forces. Stiffeners, such as 

plates or channels, can be welded to the steel panels to prevent buckling of steel plates. These 

walls are more likely to yield in shear instead of developing tension fields. The use of stiffeners 

also permits the panels themselves to be thinner than panels in unstiffened walls. The panels in 

stiffened wall participate in resisting the overturning forces because buckling does not occur in 

the panels. Consequently, overturning forces in unstiffened walls are primarily resisted by the 

columns. 
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Figure 8.4.8-1B: Fin Plate Connection Options 

Nonstructural issues: The addition of walls to an existing structure changes the architectural 

character of the building. Walls at the exterior will be visible in buildings with clear glazing. At 

interior bays, walls have to be configured to avoid obstruction of existing corridors, doorways, 

and other building systems. Walls can be exposed and incorporated into the interior architecture 

or hidden in partition walls. 
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Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 

available, and conducting comprehensive field surveys, the following issues should be noted: 

Steel connections: Welds from the fin plates to beams and columns should develop the capacity 

of the steel panels through full penetration welds or fillet welds on both sides of the plates. To 

allow for construction and field tolerances, the steel panels can be lapped with the fin plates and 

connected using fillet welds along both edges of the lap. 

Connections at slabs: To avoid damaging the slab reinforcing steel critical for transferring 

diaphragm forces to the shear walls, the concrete slab could be chipped away without damaging 

the slab steel and individual fin plates could be placed between the reinforcing steel. The 

diaphragm forces are transferred to the wall through shear studs on the beams. Alternatively, if 

continuous fin plates are required and/or the shear studs are inadequate, the reinforcing steel in 

could be cut and welded directly to the fin plates. In this case, forces transfer from the slab into 

the wall through shear friction. 

Cost/Disruption 

As always, the cost of a strengthening scheme depends on the project and its unique 

requirements. There are no issues with the SPSW system that is known to cost significantly 

more than adding braced frames or concrete shear walls to a moment frame building. 

Unstiffened walls are cheaper and less labor intensive than stiffened walls. New foundations are 

almost always required for new walls and could be extremely costly if deep foundations, such as 

drilled piers, are added. 

Installing new walls is disruptive to the occupants because of the noise and vibrations associated 

with construction. Even if tenants are relocated to parts of the building where the work is not 

being performed, vibrations associated with cutting, chipping, and drilling of concrete as well as 

the installation of steel panels can transmit through the structure. The disruption can be reduced 

somewhat if the walls are installed at the perimeter. 

Construction Considerations 

See Section 8.4.1 for general discussions of welding issues, removal of existing nonstructural 

and structural elements, and construction loads. 

Proprietary Concerns 

There are no known proprietary concerns with this technique. 

8.4.9 Convert an Existing Steel Gravity Frame to a Moment Frame 

Deficiency Addressed by Rehabilitation Technique 

This technique addresses moment frames buildings that only require slight gains in strength 

and/or stiffness. The extent of connection modification depends on the seismic hazard and the 

type of moment frame. 
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Description of the Rehabilitation Technique 

Converting existing gravity frame connections to moment frame connections does not increase 

the strength or stiffness of a structure significantly unless a large majority of the gravity frame 

column connections are made moment resisting. The strength and stiffness gain are also limited 

because the existing beams and columns used in the gravity frame are typically much lighter than 

the moment frame members. However, these members could also be strengthened as part of the 

rehabilitation scheme. This technique has less impact on the architectural character of the 

building than adding braced frames or shear walls. 

The simplest method for implementing this technique is through the addition of welded flange 

cover plates from the beam to the column without any modification of the bolted shear tab. The 

beam flanges remain unattached from the column since the gap typically exceeds the maximum 

permitted root opening size for full penetration welds. This method should only be used with 

Ordinary Moment Frame applications. 

In high seismic regions where moment frames have to meet Special Moment Frame 

requirements, a more sophisticated connection upgrade that forces the beam yielding to occur 

away from the connection should be provided. This could range from adding welds to the bolted 

shear tab for the method described above to welding top and bottom haunches from the beam to 

the column. Several methods are presented in Section 8.4.6 as well as FEMA 350 (FEMA, 

2000a) and FEMA 351 (FEMA, 2000b). The level of upgrade ultimately depends on the 

expected ductility demand and the performance objective. 

Design and Detailing Considerations 

The strong-column weak-beam concept should still be a primary design consideration. All 

detailing issues related to the design of moment frame connections need to be considered, 

including weld filler metal matching, weld metal toughness, removal of weld backing, and 

column flange and web reinforcing. For Ordinary Moment Frames, joint reinforcing should be 

limited in order to permit some yielding, as to not place excessive demands on the flange to 

column connections. For Special Moment Resisting Frames, other requirements such as widththickness limitations, lateral bracing requirements, etc., should be checked to be in accordance 

with the AISC Seismic Provisions. In most cases, it is expected that the existing gravity beam 

and column configurations will be such that it will be difficult to meet the requirements for 

Special Moment Frames without other major modifications. As a result, it is expected that in 

most cases, the Ordinary Moment Frame requirements would apply. 

Cost/Disruption 

These issues are discussed in Section 8.4.6 for moment connections. Connection upgrades are 

typically less disruptive than adding braced frames or shear walls. The costs could vary 

depending on if existing moment frame connections are also being upgraded and the total 

number of connections being modified. 

Construction Considerations 

See Section 8.4.6 for general discussions of issues related to moment connections. 
Techniques for the Seismic Rehabilitation of

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8.5 References 

ACI, 1991, Guide to Certification of Shotcrete Nozzlemen, ACI-506.3R, American Concrete 

Institute, Farmington Hills, MI. 

ACI, 1994, Guide for the Evaluation of Shotcrete, ACI-506.4R, American Concrete Institute, 

Farmington Hills, MI. 

ACI, 1995a, Guide to Shotcrete, ACI-506R, American Concrete Institute, Farmington Hills, 

Michigan. 

ACI, 1995b, Specification for Shotcrete, ACI-506.2, American Concrete Institute, Farmington 

Hills, MI. 

ACI, 1998, State of the Art Report on Fiber Reinforced Shotcrete, ACI-506.1R, American 

Concrete Institute, Farmington Hills, MI. 

ACI, 2005, Building Code Requirements for Reinforced Concrete and Commentary, ACI 318, 

American Concrete Institute, Farmington Hills, MI. 

AISC, 2005a, Code of Standard Practice for Steel Buildings and Bridges, American Institute of 

Steel Construction, Chicago, IL. 

AISC, 2005b, Seismic Provisions for Structural Steel Buildings, American Institute of Steel 

Construction, Chicago, IL. 

AISC, 2005c, Steel Construction Manual, 13th Edition, American Institute of Steel Construction, 

Chicago, IL. 

Astaneh-Asl, A., 2001, Seismic Behavior and Design of Steel Shear Walls, Structural Steel 

Educational Council, Moraga, CA. 

Carter, C.J., 1999, AISC Design Guide 13, Stiffening of Wide-Flange Columns at Moment 

Connections: Wind and Seismic Applications, American Institute of Steel Construction, Chicago, 

IL. 

FEMA, 2000a, Recommended Seismic Design Criteria for New Steel Moment-Frame Buildings, 

FEMA 350, Federal Emergency Management Agency, Washington, D.C. 

FEMA, 2000b, Recommended Seismic Evaluation and Upgrade Criteria for Existing Steel 

Moment-Frame Buildings, FEMA 351, Federal Emergency Management Agency, Washington, 

D.C. 

FEMA, 2000c, State of the Art Report on Connection Performance, FEMA 355D, Federal 

Emergency Management Agency, Washington, D.C. 

FEMA, 2000d, State of the Art Report on Past Performance of Steel Moment-Frame 
Techniques for the Seismic Rehabilitation of

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Buildings in Earthquakes, FEMA 355E, Federal Emergency Management Agency, Washington, 

D.C. 

FEMA, 2000e, Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 

356, Federal Emergency Management Agency, Washington, D.C. 

FEMA, 2003, NEHRP Recommended Provisions for Seismic Regulations for New Buildings and 

Other Structures, FEMA 450, Federal Emergency Management Agency, Washington, D.C. 

Fisher, J.M. and L.A. Kloiber, 2006, AISC Design Guide 1, Base Plate and Anchor Rod Design, 

American Institute of Steel Construction, Chicago, IL. 

Griffis, L.G., 1992, AISC Design Guide 6, Load and Resistance Factor Design of W-Shapes 

Encased in Concrete, American Institute of Steel Construction, Chicago, IL. 

Gross, J.L., Engelhardt, M.D., Uang, C.M., Kasai, K., and N.R. Iwankiw, 1999, AISC Design 

Guide 12, Modification of Existing Welded Steel Moment Frame Connections for Seismic 

Resistance, American Institute of Steel Construction, Chicago, IL. 

Lee, H. and S.C. Goel, 1990, “Seismic Behavior of Steel Built-up Box-Shaped Bracing 

Members, and Their Use in Strengthening Reinforced Concrete Frames,” Report No. UMCE 90-

7, University of Michigan, Ann Arbor, MI. 

Uriz, P. and S. Mahin, 2004, Summary of Test Results for UC Berkeley Special Concentric 

Braced Frame No. 1 (SCBF-1) Draft, Version 1.1, Department of Civil and Environmental 

Engineering, University of California, Berkeley, CA. 

Yang, F. and S. Mahin, 2005, Limiting Net Section Fractures in Slotted Tube Braces, Structural 

Steel Educational Council, Moraga, CA. 
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Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames

9-1 

Chapter 9 - Building Types S2/S2A: Steel Braced Frames 

9.1 Description of the Model Building Type 

Building Type S2 consists of a frame assembly of steel beams and columns. Lateral forces are 

resisted by diagonal steel members placed in selected bays. Floors are cast-in-place concrete 

slabs or concrete fill over metal deck. These buildings are typically used for buildings similar to 

steel moment frames, although more often for low-rise applications. Figure 9.1-1 shows an 

example of this building type. 

Building Type S2A is similar but has floors and roof that act as flexible diaphragms such as 

wood or untopped metal deck. This is a relatively uncommon building type and is used 

primarily for small office or commercial buildings in which the fire rating of concrete floor is not 

needed. 

Figure 9.1-1: Building Type S2: Steel Braced Frames 
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Variations Within the Building Type 

The two principal types of braced frame configurations are concentrically braced frames (CBFs) 

and eccentrically braced frames (EBFs). In a CBF, the centerlines of members that meet at a 

joint all intersect at a single point. These frames behave as vertical truss systems by transferring 

lateral loads primarily through axial loading of beams, columns, and braces. During 

earthquakes, inelastic behavior is typically limited to the braces and connections. Common CBF 

configurations include diagonal bracing, X-bracing (or 2-story X-bracing), V-bracing (or 

inverted-V-bracing), and K-bracing. The diagonal braces in an EBF are offset at joints such that 

link beams separate the ends of braces from columns or other braces. Inelastic behavior is 

concentrated in the links while all members outside of the links remain elastic or near elastic. 

Link beams can be located adjacent to a column or at the center of a beam. Common types of 

braces include W-shapes, hollow structural sections (HSS), steel pipes, double angles, and 

double channels. 

Braces are either welded directly to the beams and columns or welded to gusset plates. It is 

standard practice on the West Coast to weld gusset plates to the beams and columns. Away from 

the West Coast, the plate is more commonly welded to the beam and bolted to the column with a 

pair of angles or a WT-shape. Beam-column connections vary depending on whether there is a 

brace at the joint or not. Connections range from simple shear tabs to fully welded moment 

connections. 

Floor and Roof Diaphragms 

Diaphragms associated with this building type may be either rigid or flexible. The typical rigid 

diaphragm found in modern buildings consists of structural concrete on metal deck. Diaphragm 

forces transfer to the frames through shear studs welded to the beams. Older steel buildings that 

were constructed before metal decks were commonly used may have concrete slabs or masonry 

arches that span between the beams. Flexible diaphragms include bare metal deck or metal deck 

with nonstructural fill. These are frequently used on roofs that support light gravity loads. 

Decks could be connected to the steel members with shear studs, puddle welds, screws, or shot 

pins. The steel members also act as chords and collectors for the diaphragm. 

Foundations 

There is no typical foundation for this building type. Foundations can be of any type, including 

spread footings, mat footings, and piles, depending on the characteristics of the building, the 

lateral forces, and the site soil. Spread footings are used when lateral forces are not very high 

and a firm soil exists. For larger forces and/or poor soil conditions, a mat footing below the 

entire structure is commonly used. Pile foundations are used when lateral forces are extremely 

large or poor soil is encountered. The piles can be either driven or cast-in-place. Vertical forces 

are distributed to the underlying soil through a combination of skin friction between the pile and 

soil and/or direct bearing at the end of the pile; lateral forces are resisted primarily through 

passive pressure on the vertical surfaces of the pile cap and piles. 

9.2 Seismic Performance Characteristics 

Braced frames are generally considered to be stiff systems in the elastic range. Their nonlinear 

response depends on their ability to redistribute forces between bays and drifts between stories. 
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9-3 

Braces and connections in CBF undergo large inelastic deformations in tension and compression 

into the post-buckling range. Ductility of CBF systems in past earthquakes have been limited by 

local failures of braces and connections. It is thought that CBF systems that are properly 

designed and detailed can possess ductility in excess of that previously assigned to these systems 

(AISC, 2005). Yet, recent experimental testing at UC Berkeley found that special concentrically 

braced frames (SCBF), designed with an inverted-V configuration using the 1997 AISC Seismic 

Provisions, fractured the HSS braces after only a few cycles of loading in the inelastic range 

(Uriz and Mahin, 2004). Also, the damage concentrated at the level that first experienced brace 

buckling, resulting in weak story response. As of this writing, these results are still under 

investigation. 

EBF systems approach the higher performance levels of structural systems such as bucklingrestrained braced frames and fluid or friction damped frames (Horne et al., 2001). By limiting 

nonlinear action to the link beams, the post-yield behavior of the system and the maximum 

demand on the frame can be better predicted. The ductility of an EBF system is dependent on 

the length and detailing of the link beams and whether shear or flexural yielding governs their 

inelastic response. 

9.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

Undesirable behaviors of CBF systems that have been observed in past earthquakes include 

fracture of connection elements, fracture of braces, and local buckling of braces. These failures, 

in turn, cause excessive demands on other elements in the system and lead to overall frame 

failure. See Table 9.3-1 for deficiencies and potential rehabilitation techniques particular to this 

system. Selected deficiencies are further discussed below by category. 

Global Strength 

The lack of global strength to resist the seismic demands is a direct result of weak frames. In 

some cases, the existing beams and columns in a frame may possess enough capacity to 

accommodate upgrades to the braces and connections. If the beams and columns cannot make 

this accommodation, braces can be added to other bays to create new frames and thereby, reduce 

the demand on the existing frames. This would be more easily achieved for CBF systems than 

for EBF systems, which have special link beam requirements (AISC, 2005). 

Global Stiffness 

Braced frames are extremely stiff in the elastic range. Concerns with global stiffness occur in 

the inelastic range when braces are prone to buckling and cause a loss of stiffness. Limited postelastic stiffness is potentially provided by the frame and non-frame columns in the system (Tang 

and Goel, 1987; Hassan and Goel, 1991). 

Configuration 

Concentrically braced frames: K-bracing and inverted-V-bracing configurations exhibit 

undesirable behavior when a brace buckles. The remaining tension brace at a joint imparts an 

unbalanced force onto the column or beam. This is particularly hazardous for frames with Kbracing since it may cause a column to fail entirely. In frames that use inverted-V-bracing, the 

unbalanced force is additive to gravity loads on beams. 
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9-4 

Table 9.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S2/S2A Bui

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce De

Global Strength 

Insufficient frame strength 

Braced frame [8.4.1] 

Concrete/masonry shear wall [8.4.2] 

Steel plate shear wall [8.4.8] 

Strengthen braces [9.4.2], beams [8.4.3], columns [8.4.3], and/or connections [9.4.1] 

Seismic is[24.3] 

Suppleme

damping [

Global Stiffness 

Excessive drift Braced frame [8.4.1] 

Concrete/masonry shear wall [8.4.2] 

Steel plate shear wall [8.4.8] 

Strengthen braces [9.4.2], beams [8.4.3], columns [8.4.3], and/or connections [9.4.1] 

Supplemedamping [

Configuration Soft story Braced frame [8.4.1] 

Concrete/masonry shear wall [8.4.2] 

Steel plate shear wall [8.4.8] 

 Re-entrant 

corner 

Braced frame [8.4.1] 

Concrete/masonry shear wall [8.4.2] 

Collector [8.4.4] 

Enhance detailing 

[8.4.3], [8.4.4] 

Load Path Missing 

collector 

Collector [8.4.4] 

 Inadequate 

shear, flexural, and uplift anchorage to foundation 

Embed column into a pedestal bonded to other existing foundation elements [8.4.5] 

Provide steel shear lugs or anchor bolts from base plate to foundation [8.4.5] 

 Inadequate outof-plane 

anchorage at walls connected to diaphragm 

Tension anchors 

[16.4.1] 
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9-5 

Table 9.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S2/S2A Bui

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce De

Component Detailing 

Inadequate capacity of braces and/or connection 

Replace braces [9.4.1] Increase area of braces 

[9.4.2] 

Make braces composite elements [9.4.2] 

Improve b/t ratios [9.4.2] 

Add bolts and welds [9.4.1] 

Increase size of gusset plates [9.4.1] 

 Inadequate 

capacity of beams, columns, and/or connections 

Add cover plates or box members [8.4.3] 

Provide gusset plates or knee braces [9.4.1] 

Provide gusset plates [9.4.1] 

 EBFs not 

conforming to current standards 

Check current EBF design standards [9.4.2] 

 Inadequate 

capacity of horizontal steel bracing 

Provide additional secondary bracing [9.4.2] 

Strengthen bracing elements [9.4.2] 

Reduce unbraced lengths [9.4.2] 

Strengthen connections [9.4.1] 

Diaphragms Inadequate 

in-plane strength and/or stiffness 

Collectors to distribute forces [8.4.4] 

Moment frame 

Braced frame [8.4.1] 

Concrete/masonry shear wall [8.4.2] 

Concrete topping slab overlay 

Wood structural panel overlay at flexible diaphragms [22.2.1] 

Strengthen chords [8.4.3], [8.4.4], and [22.2.2] 

Add nails at flexible diaphragms [22.2.1] 

Inadequate shear transfer to frames 

Provide additional shear studs, anchors, or welds [22.2.7] 

Inadequate chord capacity 

Add steel members or reinforcement [8.4.3], [8.4.4] 
Techniques for the Seismic Rehabilitation of

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Table 9.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S2/S2A Bui

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce De

Diaphragms (continued) 

Excessive stresses at openings and irregularities 

Add reinforcement [8.4.3] 

Provide drags into surrounding diaphragm [8.4.4] 

Foundations See Chapter 23 [ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 
Techniques for the Seismic Rehabilitation of

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9-7 

Eccentrically braced frames: The inelastic response of a link beam is influenced by its length 

relative to the ratio MP/Vp of the link section (AISC, 2005). Link beams that exhibit shear 

yielding have greater inelastic deformation capacity than ones that exhibit flexural yielding. 

AISC Seismic Provisions (AISC, 2005) permit shear yielding links to have four times the plastic 

rotation angles of flexural yielding links. Configurations with link beams adjacent to columns 

are susceptible to weld fractures found in pre-Northridge connections unless special detailing 

measures are taken to reduce the demands on these welds. 

Load Path 

Load path deficiencies in steel braced frame buildings include inadequate connections, 

collectors, and frame anchorage to foundations. Brace connections may have less strength than 

the braces. In Type S2 buildings, seismic forces transfer from the diaphragm to the frame 

through shear studs welded to collectors or directly to the frame beams. The collectors or the 

connections to the frame may be too weak to transfer these forces. Connections from columns to 

base plates or pile caps must resist shear, flexural, and potential uplift forces. Connections that 

cannot develop these frame forces prevent the frame from developing its full capacity. 

Component Detailing 

CBF: HSS and pipe braces with high b/t ratios and other steel shapes that lack compactness are 

subject to local buckling or fracture after a limited number of inelastic cycles. Ductility of these 

braces can be improved by infilling with concrete or adding longitudinal stiffeners. 

Alternatively, the braces can be replaced with double HSS sections, which can be used in 

configurations similar to double angles or channels (Lee and Goel, 1990). HSS and pipe braces 

are also subject to net section fracture at the gusset plate slots (Uriz and Mahin, 2004). This 

deficiency can be mitigated by adding reinforcing plates to the sides of the HSS without the 

slots, as shown in Figure 8.4.1-1 and Figure 8.4.1-2. For pipes, the reinforcing plates can be 

oriented at right angles to the pipe and appear like stiffeners. Brace connections that are only 

designed for the axial capacity of the braces may not be adequate to generate the full strength of 

the braces. To achieve good post-inelastic response, all eccentricities in the connection must be 

considered. A brace that buckles in the plane of the gusset plates should have its end 

connections designed for the full axial load and flexural strength of the brace (AISC, 2005). A 

brace that buckles out-of-plane should ensure that the gusset plates can develop restraint-free 

plastic rotations without buckling. 

EBF: Frames that rely on link-to-column connections have traditionally utilized similar detailing 

as pre-Northridge connections at beam-column joints. These connections should be reevaluated 

in the wake of the findings following the Northridge earthquake. Also, experimental research 

has found that link beams at the first floor undergo the largest inelastic deformation and have the 

potential to create a soft story (AISC, 2005). 

Diaphragm Deficiencies 

Common diaphragm deficiencies include insufficient in-plane shear strengths, inadequate 

chords, and excessive stresses at openings. Causes for these deficiencies could be due to lack of 

slab or fill thickness, lack of reinforcing steel in the slabs, insufficient connections to chord 

elements, and poor detailing at openings. See Chapter 22 for common rehabilitation techniques. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames

9-8 

Foundation Deficiencies 

Foundations that are inadequate do not develop the full capacity of the lateral force-resisting 

system. Their deficiencies result from insufficient strengths and sizes of footings, grade beams, 

pile caps, and piles. See Chapter 23 for common rehabilitation techniques. 

9.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

9.4.1 Enhance Braced Frame Connection 

Deficiency Addressed by Rehabilitation Technique 

Adequate capacities of connections are essential to the proper performance of a braced frame. 

Connections with insufficient strength and/or ductility to develop stable inelastic frame behavior 

are strengthened or replaced. 

Description of the Rehabilitation Technique 

Brace end connections commonly rely on additional connection elements (e.g., gusset plates), 

but the braces may also be directly welded to the beam and column through a moment 

connection. Moment connections are also used for link-to-column connections in an EBF and 

sometimes used for beam-to-column connections in both EBF and CBF systems. The mitigation 

approach is different depending on whether the existing connection relies on a connection 

member or a moment connection. 

If the existing connection members have sufficient capacity, the most economical alternative 

may be to increase the connection capacity by providing additional welds or bolts. This typically 

only allows for a limited increase in capacity since existing brace connection configurations can 

rarely accommodate significant modifications. If the existing connection members have 

inadequate capacity, the existing configuration and accessibility need to be assessed to determine 

whether adding supplemental connection members or replacing the existing connection members 

with members of greater capacity is more economical. Supplementing the existing connection 

eliminates the challenges associated with removal of existing connection welds and temporary 

support of the braces. 

The primary concern with moment connections used in braced frames is the use of low notch 

toughness weld metals. For braces that are expected to develop plastic hinges at their ends, 

consider replacing the existing welds with a high notch toughness weld metal. Beam-to-column 

connections in CBF do not typically experience large flexural forces and likely do not need to be 

upgraded. An exception occurs when the braces or their end connections fail and frame action 

becomes the primary mechanism for lateral force resistance; however, this is not a recommended 

design approach, as it does not ensure stable post-elastic behavior. In EBF configurations where 

the link beam is adjacent to the column, an upgrade should be considered given the large 

demands on the link beam and its critical nature. 
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9-9 

Design Considerations 

AISC Seismic Provisions does not permit the sharing of loads by both welds and bolts on the 

same faying surface. Thus, a bolted brace to gusset plate connection should only be enhanced 

with bolts or replaced entirely with welds. Note it is not uncommon in some regions for a gusset 

plate to be bolted to the column through a shear plate and welded to the beam since these are 

separate faying surfaces. In addition to having thorough knowledge of the existing material 

behaviors to ensure that the new and existing elements to interact in the desired manner, other 

design issues include the following: 

Research basis: No references directly addressing upgrades of braced frame connections have 

been identified. 

Design forces: Brace connections that are only designed for the axial capacity of the braces may 

not be adequate to generate the full strength of the braces. A brace that buckles in the plane of 

the gusset plates should have its end connections designed for the full axial load and flexural 

strength of the brace (Astaneh-Asl et al., 1986). This recommendation may be more appropriate 

for high seismic applications. 

Bolted connections: Bolts that govern the existing connection capacity could be replaced with 

stronger bolts, such as replacing A325 with A490. The upgrade may also be as simple as 

replacing bolts that have threads included in the shear plane with longer bolts. More bolts could 

be added if the existing configuration allows for it. Larger bolts could be provided but this 

would reduce the net section capacities due to enlargement of the holes. 

Welded connections: Existing fillet welds can be thickened provided the welds are not of low 

notch toughness weld metals or found to be inadequate through material testing. Otherwise, the 

existing welds should be removed and replaced with high notch toughness weld metals. A 

typical fully welded connection appropriate for use in SCBF is shown in Figure 8.4.1-1. For low 

to moderate seismic applications, Figure 8.4.1-2 shows a more compact connection. Welded 

brace connections to the weak axis of columns are complex and expensive; an example of this 

type of connection is shown in Figure 8.4.1-3. 

Moment connections: As mentioned above, moment connections that are subject to large flexural 

forces should be upgraded with high notch toughness weld metals. This would primarily apply 

to braces that directly connect to beams and columns in SCBF and some OCBF in high seismic 

applications. EBF link beams that are adjacent to columns also fall into this category. Link 

beams develop large flexural forces whether shear or flexural yielding governs. 

Nonstructural issues: Gusset plates designed in accordance with the AISC Seismic Provisions for 

SCBF can be fairly large and should be discussed with the architect and tenants. 

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 

available, and conducting comprehensive field surveys, the following issues should be noted: 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames

9-10 

Gusset plates: A brace in a SCBF that buckles out-of-plane could form plastic hinges at midspan 

and in the gusset plates at each end. The gusset plates should provide restraint-free plastic 

rotations without buckling. AISC Seismic Provisions suggests a minimum distance of two times 

the plate thickness between the end of the brace and the assumed line of restraint to achieve this. 

Consider increasing this distance to three times the plate thickness to accommodate over cutting 

of the slots in HSS and other erection tolerances. Connections that do not allow for restraint-free 

plastic rotations out-of-plane should be used primarily in situations where in-plane buckling of 

the braces govern or ductility demands are low. 

Bolted connections: If new holes are drilled in the existing brace and connection member, 

tolerances of the existing holes should be verified to ensure that bolts will be loaded evenly. 

Weld filler metal matching and overmatching: Weld filler metals with slightly greater tensile 

strength than the connected steel should be used. Flux cored arc welding and shielded metal arc 

welding electrodes that conform to E70 specifications exhibit overmatching properties compared 

to common steel specifications, including ASTM A36, A572 (Grade 42 and 50), A913 (Grade 

50), and A992 (FEMA, 2000b). 

Weld metal toughness: AISC Seismic Provisions now require weld metals with minimum Charpy 

V-Notch (CVN) toughness of 20 ft-lbf at -20oF for all welds in the lateral force-resisting system. 

There is also an additional requirement of 40 ft-lbs at 70oF for demand critical welds. 

Cost/Disruption 

Connection modifications are locally very disruptive. Noise associated with this type of work 

will spread and disrupt tenants on other floors unless the work is done during off-hours. These 

modifications could be particularly costly if existing gusset plates have to be replaced. 

Construction Considerations 

See Section 8.4.1 for general discussions of welding issues, removal of existing nonstructural 

and structural elements, and construction loads. 

Proprietary Concerns 

There are no known proprietary concerns with this technique. 

9.4.2 Enhance Strength and Ductility of Braced Frame Member 

Deficiency Addressed by Rehabilitation Technique 

Inadequate beams, columns, and braces are strengthened or replaced to achieve ductile frame 

behavior. 

Description of the Rehabilitation Technique 

The ductility of an existing brace can be enhanced by reducing its slenderness, which can be 

accomplished by decreasing its unbraced length, infilling hollow sections with concrete, or 

adding longitudinal stiffeners. The unbraced length of a brace can be reduced by adding 

secondary bracing members that are not part of the primary lateral force-resisting system. 

Infilling existing hollow sections with concrete can reduce the severity of local buckling (Liu and 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames

9-11 

Goel, 1988; Lee and Goel 1987). An effective width-thickness ratio for the infilled member is 

determined by multiplying the width-thickness ratio of the section by the factor (0.0082 x KL/r + 

0.264), applicable to braces with KL/r values between 35 and 90 (Goel and Lee, 1992). Adding 

longitudinal stiffeners presents the least field complications; the stiffeners could consist of plates 

or small angle sections. 

If both strength and ductility are required, new braces have to be added. Some configurations 

may lend themselves to schemes that allow the existing braces to remain. These include single 

angle, double angle, and channel braces that can be doubled; rolled sections can also be cover 

plated. In other cases, it is more practical to replace the existing brace with a new brace, of 

which numerous options exist. The increase in brace strength may require upgrades to other 

components of the braced frame, such as the brace connections, beams, and columns. 

Connection upgrades are discussed in the Section 9.4.1. In many cases, the most cost-effective 

alternative for increasing the capacity of the existing beams and columns in is to add cover plates 

or side plates to create box sections. This technique is discussed in Section 8.4.3. 

Design Considerations 

It would be preferable to limit the strengthening of the existing braces to the capacity of the other 

members of the lateral force-resisting system, including the foundations, to avoid triggering too 

many upgrades. Thorough knowledge of the existing material behaviors and strengths are 

necessary for the new and existing elements to interact in the desired manner. Other design 

issues include the following: 

Research basis: No references directly addressing upgrades of braced frame members have been 

identified. 

Existing brace strengthening: Significant modifications to an existing brace could trigger 

strengthening or redesign of its end connections. Strengthening of existing K- or inverted-Vbracing should be undertaken only after careful evaluation of the additional bending forces 

following the buckling of the compression bracing. Where the existing bracing in these systems 

is found to have inadequate capacity, the preferred solution is to replace it with a diagonal or Xbracing configuration. 

Secondary bracing: A brace member is designed to resist both tension and compression forces, 

but its capacity for compression stresses is limited by potential buckling and is therefore less 

than the capacity for tensile stresses. Since the design of the system generally is based on the 

compression capacity of the brace, some additional capacity may be obtained by simply reducing 

the unsupported length of the brace by means of secondary bracing provided the connections 

have adequate reserve capacity or can be strengthened for the additional loads. 

New brace selection: If existing braces are replaced, use compact and non-slender sections 

whenever possible to avoid premature fracturing or buckling of the braces during post-yield 

behavior. Two particular brace types not common in older braced frame buildings are double 

HSS sections and buckling-restrained braces. Double HSS sections can be used in 

configurations similar to double angles or channels (Lee and Goel, 1990). They provide reduced 

fit-up issues and smaller width-to-thickness ratios compared to a single HSS, resulting in 
Techniques for the Seismic Rehabilitation of

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9-12 

increased energy dissipation capacity. The other type of brace, used in a buckling-restrained 

braced frame (BRBF), is typically used in new buildings but has also been used successfully in 

new BRBF systems in existing buildings. One example of a brace used in a BRBF consists of a 

steel core inside a casing, which consists of a hollow structural section (HSS) infilled with 

concrete grout. Proprietary materials separate the steel core and concrete to prohibit bonding 

between the two materials. There are other buckling-restrained braces that do not use grout or 

additional separating agents between the steel and grout. The main advantage of these braces is 

the ability of the casing to restrain the buckling of the steel core without providing any additional 

axial force resistance beyond the capacity of the steel core. Provisions for new building BRBF 

design are included in the NEHRP Recommended Provisions (FEMA, 2003) and the AISC 

Seismic Provisions for Structural Steel Buildings. Note that significant connection modifications 

may be required when braces are replaced. 

Nonstructural issues: Brace modifications, when exposed, will affect the interior architecture or 

if hidden in partition walls, these walls may be thicker than typical walls. Beams that are 

increased in size affect nonstructural components by reducing clear floor heights. These 

components typically include suspended ceilings, pipes, conduits, and ducts. Coordination with 

the architect and other trades should not be overlooked or underestimated.

Detailing Considerations 

In addition to obtaining the latest drawings for the building including as-built drawings, if 

available, and conducting comprehensive field surveys, the following issues should be noted: 

Built-up brace members: While double angles, double channels, and double HSS offer 

advantages for installation, special criteria apply to these members when used in a SCBF. 

Buckling of these types of braces imposes large shear forces on the stitches. Therefore, closer 

stitch spacing and higher stitch strengths are required. More stringent member compactness is 

also necessary for ductility and energy dissipation. 

Reinforcing cover plates: HSS and pipe braces are subject to net section fracture at the gusset 

plate slots (Uriz and Mahin, 2004). This brittle failure mode can be eliminated by adding 

reinforcing cover plates to the sides of the HSS without the slots, such as the ones shown in 

Figures 8.4.1-1 and 8.4.1-2. For pipes, the reinforcing plates can be oriented at right angles to 

the pipe and appear like stiffeners. Additional information regarding the design of these plates 

can be found in Limiting Net Section Fracture in Slotted Tube Braces (Yang and Mahin, 2005). 

Cost/Disruption 

Designs that are simple and details that are not overly complicated will minimize costs 

associated with this technique. This could include maximizing the use of existing members, 

minimizing connection upgrades, and reducing the amount of field welding. 

Costs can also be reduced if disruption is minimal during construction. Installing braces at the 

perimeter frames reduces logistical issues associated with working in confined spaces and 

temporary removal of the nonstructural elements. Noise associated with this type of work is loud 

and disturbing to the tenants if the building is occupied while the work is being performed. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames

9-13 

Construction Considerations 

See Section 8.4.1 for general discussions of welding issues, removal of existing nonstructural 

and structural elements, and construction loads. Connecting members, such as gusset plates, that 

are not being replaced should be protected when braces are removed. 

Proprietary Concerns 

Braces used in buckling-restrained braced frames (BRBF) are proprietary. There are a limited 

number of manufacturers of the braces used in BRBF. 

9.5 References 

AISC, 2005, Seismic Provisions for Structural Steel Buildings, American Institute of Steel 

Construction, Chicago, IL. 

Astaneh-Asl, A., Goel, S.C., and R.D. Hanson, 1986, “Earthquake-Resistant Design of Double 

Angle Bracing,” Engineering Journal, Vol. 23, No. 4, American Institute of Steel Construction, 

Chicago, IL. 

FEMA, 2000a, Recommended Seismic Evaluation and Upgrade Criteria for Existing Steel 

Moment-Frame Buildings, FEMA 351, Federal Emergency Management Agency, Washington, 

D.C. 

FEMA, 2003, NEHRP Recommended Provisions for Seismic Regulations for New Buildings and 

Other Structures, FEMA 450, Federal Emergency Management Agency, Washington, D.C. 

Goel, S.C. and S. Lee, 1992, “A Fracture Criterion for Concrete-Filled Tubular Braces,” 

Proceedings of the 1992 ASCE Structures Congress, pp. 922-925, ASCE, Reston, VA. 

Hassan, O. and S.C. Goel, 1991, “Seismic Behavior and Design of Concentrically Braced Steel 

Structures,” Report No. UMCE 91-1, University of Michigan, Ann Arbor, MI. 

Horne, J., Rubbo, A., and J. Malley, 2001, “AISC-LRFD Design and Optimization of Steel 

Eccentrically Braced Frames,” Proceedings of 2001 Convention of the Structural Engineers 

Association of California, Sacramento, CA. 

Lee, H. and S.C. Goel, 1990, Seismic Behavior of Steel Built-up Box-Shaped Bracing Members, 

and Their Use in Strengthening Reinforced Concrete Frames, Report No. UMCE 90-7, 

University of Michigan, Ann Arbor, MI. 

Lee, S. and S.C. Goel, 1987, Seismic Behavior of Hollow and Concrete-Filled Square Tubular 

Bracing Members, Report No. UMCE 87-11, University of Michigan, Ann Arbor, MI. 

Liu, Z.,and S.C. Goel, 1988, “Cyclic Load Behavior of Concrete-Filled Tubular Braces,” Journal 

of Structural Division, Vol. 114, No. 7, ASCE, Reston, VA. 

Tang, X. and S.C. Goel, 1987, Seismic Analysis and Design Considerations of Braced Steel 

Structures, Report UMCE 87-4, University of Michigan, Ann Arbor, MI. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 9 - Types S2/S2A: Steel Braced Frames

9-14 

Uriz, P. and S. Mahin, 2004, Summary of Test Results for UC Berkeley Special Concentric 

Braced Frame No. 1 (SCBF-1) Draft, Version 1.1, Department of Civil and Environmental 

Engineering, University of California, Berkeley, CA. 

Yang, F. and S. Mahin, 2005, Limiting Net Section Fractures in Slotted Tube Braces, Structural 

Steel Educational Council, Moraga, CA. 
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10-1 

Chapter 10 - Building Type S4: Steel Frames with Concrete 

Shear Walls 

10.1 Description of the Model Building Type 

Building Type S4 consists of an essentially complete frame assembly of steel beams and 

columns. The floors are concrete slabs or concrete fill over metal deck. These buildings feature 

a significant number of concrete walls effectively acting as shear walls, either as vertical 

transportation cores, isolated in selected bays, and/or as a perimeter wall system. The steel 

column and beam system may act only to carry gravity loads or may have rigid connections to 

act as a moment frame to form a dual system. This building type is generally used as an alternate 

for steel moment or braced frames in similar circumstances. These buildings will usually be 

mid- or low-rise. Figure 10.1-1 shows an example of this building type. 

Figure 10.1-1: Building Type S4: Steel Frames with Concrete Shear Walls 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 10 - Type S4: Steel Frames with Concrete Shear Walls

10-2 

10.2 Seismic Performance Characteristics 

In older buildings, the steel frame carries only gravity loads while all lateral loads are resisted by 

the concrete shear walls. In modern buildings, both lateral systems work together in proportion 

to relative rigidity. Generally, except in tall buildings, these systems tend to behave more like 

shear wall structures due to the much greater stiffness of the walls. The contribution of the steel 

moment frame to the lateral capacity of the building is a function of the number of frames and 

the detailing of the beam-column joints. See performance characteristics described in Section 

14.2 for concrete shear wall buildings and Section 8.2 for steel moment frame buildings. 

10.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

See deficiencies and techniques described in Section 13.3 and 14.3 for concrete shear wall 

buildings and Section 8.3 for steel moment frame buildings. 

10.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

See recommended techniques in Section 13.4 for concrete shear wall buildings and Section 8.4 

for steel moment frame buildings. 

10.5 References 

See references in Section 13.5 and 14.5 for concrete shear wall buildings and Section 8.5 for 

steel moment frame buildings. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 10 - Type S4: Steel Frames 

10-3 

Table 10.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S4 Build

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce Dem

Global Strength 

Insufficient inplane wall shear strength 

Concrete/masonry shear wall [8.4.2] 

Braced frame [8.4.1] 

Concrete wall overlay [21.4.8] 

Fiber composite wall overlay [13.4.1] 

Steel overlay 

Seismic isolation [2

Reduce flecapacity [13.4.4] 

 Insufficient 

flexural capacity 

Concrete/masonry shear wall [8.4.2] 

Braced frame [8.4.1] 

Add or enhance chords 

 Insufficient 

frame strength 

Moment frame 

Braced frame [8.4.1] 

Concrete/masonry shear wall [8.4.2] 

Strengthen beams [8.4.3], columns [8.4.3], and/or connections [8.4.6] 

Seismic iso

[24.3] 

Global Stiffness 

Excessive drift Concrete/masonry shear 

wall [8.4.2] 

Braced frame [8.4.1] 

Moment frame 

Strengthen beams [8.4.3], columns [8.4.3], and/or connections [8.4.6] Concrete wall overlay [21.4.5] 

 Inadequate 

capacity of coupling beams 

Concrete/masonry shear wall [8.4.2] 

Braced frame [8.4.1] 

Strengthen beams [13.4.2] 

Improve ductility of beams [13.4.2] 

Configuration Discontinuous 

walls 

Concrete/masonry shear wall [8.4.2] 

Enhance existing column for overturning loads 

Improve connection to diaphragm [13.4.3] 

 Soft story Concrete/masonry shear 

wall [8.4.2] 

Braced frame [8.4.1] 

 Re-entrant 

corner 

Moment frame 

Braced frame [8.4.1] 

Concrete/masonry shear wall [8.4.2] 

Collector [8.4.4] 

Enhance detailing [8.4.3], [8.4.4] 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 10 - Type S4: Steel Frames 

10-4 

Table 10.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S4 Build

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce Dem

Configuration (continued) 

Torsional layout 

Add balancing walls [8.4.2], braced frames [8.4.1], or moment frames 

Load Path Missing 

collector 

Add collector [8.4.4] Strengthen existing 

beam [8.4.3] or slab 

Enhance splices or connections of existing beams [8.4.4] 

 Discontinuous 

Walls 

Provide new wall support components to resist the maximum expected overturning moment 

Strengthen the existing support columns for the maximum expected overturning moment [8.4.3] 

Provide elements to distribute the shear into the diaphragm at the level of discontinuity [13.4.3] 

 Inadequate 

shear, flexural, and uplift anchorage to foundation 

Embed column into a pedestal bonded to other existing foundation elements [8.4.5] 

Provide steel shear lugs or anchor bolts from base plate to foundation [8.4.5] 

 Inadequate outof-plane 

anchorage at walls connected to diaphragm 

Tension anchors 

[16.4.4] 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 10 - Type S4: Steel Frames 

10-5 

Table 10.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S4 Build

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce Dem

Component Detailing 

Wall inadequate for out-of-plane bending 

Add strongbacks [21.4.3] Concrete wall 

overlay [21.4.5] 

 Wall shear 

critical 

Concrete wall overlay [21.4.5] 

Fiber composite wall overlay [13.4.1] 

Reduce flecapacity of [13.4.4] 

 Inadequate 

capacity of beams, columns, and/or connections 

Enhance beamcolumn connections [8.4.6] Add cover plates or box members [8.4.3] Provide gusset plates or knee braces [9.4.1] Encase columns in concrete [8.4.2] 

 Inadequate 

capacity of panel zone 

Provide welded continuity plates [8.4.6] Provide welded stiffener or doubler plates [8.4.6] 

 Inadequate 

capacity of horizontal steel bracing 

Provide additional secondary bracing [9.4.2] 

Strengthen bracing elements [9.4.2] Reduce unbraced lengths [9.4.2] 

Strengthen connections [9.4.1] 

Diaphragms Inadequate 

in-plane strength and/or stiffness 

Collectors to distribute forces [8.4.4] 

Moment frame 

Braced frame [8.4.1] 

Concrete/masonry shear wall [8.4.2] 

Concrete topping slab overlay Strengthen chords [8.4.3], [8.4.4] 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 10 - Type S4: Steel Frames 

10-6 

Table 10.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S4 Build

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce Dem

 Inadequate 

shear transfer to frames 

Provide additional shear studs, anchors, or welds [22.2.7] 

Diaphragms (continued) 

Inadequate chord capacity 

Add steel members or reinforcement [8.4.3], [8.4.4] 

 Excessive 

stresses at openings and irregularities 

Add reinforcement [8.4.3] 

Provide drags into surrounding diaphragm [8.4.4] 

Foundations See Chapter 23 [ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 
Techniques for the Seismic Rehabilitation of Chapter 11 – Types S5/S5A: Steel Frames 

Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

11-1 

Chapter 11 - Building Types S5/S5A: Steel Frames with Infill 

Masonry Shear Walls 

11.1 Description of the Model Building Type 

Building Type S5 is normally an older building that consists of an essentially complete gravity 

frame assembly of steel floor beams or trusses and steel columns. The floor consists of masonry 

flat arches, concrete slabs or metal deck and concrete fill. Exterior walls, and possibly some 

interior walls, are constructed of unreinforced masonry, tightly infilling the space between 

columns and between beams and the floor such that the infill interacts with the frame to resist 

lateral movement. Windows and doors may be present in the infill walls, but to effectively act as 

a shear resisting element, the infill masonry must be constructed tightly against the columns and 

beams. The steel gravity framing in these buildings may include truss spandrels or knee braces 

on the exterior walls, or partially restrained beam-column connections in a more extensive 

pattern. The steel frame also is often cast in concrete for fireproofing purposes. The buildings 

intended to fall into this category normally feature exposed clay brick masonry on the exterior 

and are common in commercial areas of cities with occupancies of retail stores, small offices, 

and hotels. Figure 11.1-1 shows an example of this building type. 

Figure 11.1-1: Building Type S5: Steel Frames with Infill Masonry Shear Walls 
Techniques for the Seismic Rehabilitation of Chapter 11 – Types S5/S5A: Steel Frames 

Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

11-2 

The S5A building type is similar but has floors and roof that act as flexible diaphragms such as 

wood, or untopped metal deck. This type of building will almost always date to the 1930 or 

earlier. 

Variations Within the Building Type 

The building type was identified primarily to capture the issues of interaction between 

unreinforced masonry and steel gravity framing. The archetypical building has solid clay brick 

at the exterior with one wythe of brick running continuously past the plane of the column and 

beam and two or more wythes infilled within the plane of the column and beam. The exterior 

wythe of clay brick forms the finish of the building although patterns of terra cotta, stone, or 

precast concrete may be attached to the brick or laid up within the brick. However, there can be 

many variations to this pattern depending on the number and arrangement of finished planes on 

the exterior of the building. For example, the full width of the infill wall may be located with the 

plane of the column and beam with a pilaster built out and around the column and a horizontal 

band of brick or other material covering the beam. The beam is often placed off center of the 

column, usually on the out-board side. In extreme conditions, the primary plane of the masonry 

wall may not directly engage the column at all. In these cases, strut compression must be 

transferred eccentrically through the masonry surrounding the column, reducing effectiveness. 

In some buildings the steel frame is encased in concrete, primarily for fireproofing. This 

encasement is normally reinforced with mesh and may contribute to overall frame stiffness and 

to connection strength and stiffness of partially restrained steel connections. Importantly, at the 

perimeter frames, the concrete encasement forms a smooth surface at the masonry interface and 

probably encouraged a neater fit during construction. Concrete encasement of columns also will 

assist in transferring eccentric strut loading into the column-frame system 

Hollow clay tile masonry may also be used as an exterior infill material. Although this material 

often has a very high compression strength, the net section of material available to form the 

compression strut within the frame will normally contribute a lateral strength of only a small 

percentage of the building weight. The material being brittle and the wall being highly voided, 

these walls may also lose complete compressive strength quite suddenly. Therefore, walls of 

hollow clay tile infill will probably not contribute a significant portion of required lateral 

resistance except in areas of low seismicity and/or when walls are arranged as infill on both the 

exterior and interior of the building. 

More recent buildings may have unreinforced concrete block masonry configured as an exterior 

infill wall, with a variety of finish materials attached to the outside face of the concrete block. 

Similar to hollow clay tile walls, these walls may exhibit moderate to low compressive strength 

and brittle behavior that marginalizes their usefulness as lateral elements. In addition, hollow 

concrete block exterior walls often will not be installed tight to the surrounding framing, 

eliminating infill compression strut behavior. 
Techniques for the Seismic Rehabilitation of Chapter 11 – Types S5/S5A: Steel Frames 

Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

11-3 

Floor and Roof Diaphragms 

The earliest version of this building type may include floors constructed of very shallow masonry 

arches spanning between steel beams. A relatively flat top surface is created with masonry 

rubble or light-weight cementitious fill and the floor is finished with wood sheathing. In some 

cases, the thrust from the arches is resisted by tie rods running perpendicular and through the 

steel beams. The only diaphragm action provided by such floors is the finish wood sheathing 

and the lateral flexibility of this system is incompatible with the stiff but brittle masonry arches. 

Building Type S5A will have heavy timber floors with one or more layers of sheathing forming a 

diaphragm. The flexibility of such diaphragms will often form a seismic deficiency because, 

assuming no interior shear elements, the large drift at the diaphragm mid-span will damage 

perpendicular walls and gravity framing. Specific strengthening techniques for this building type 

are not covered here. For generalized strengthening of diaphragms, see Chapter 22. 

Most typically, the floor and roof are cast-in-place concrete slabs spanning between beams. The 

concrete slab is often integral with lightly reinforced concrete surrounding each beam. This 

building type can also be found with metal deck and concrete floor slabs. 

Foundations 

There is no typical foundation for this building type. Foundations can be found of every type 

depending on the height of the building, the span of the gravity system and the site soil. The 

exterior walls are exceptionally heavy and typically will be supported by a continuous concrete 

footing or often a continuous concrete wall forming a basement space below. 

11.2 Seismic Response Characteristics 

Most steel frame infill buildings will incorporate some beam column connections with moment 

resistance, either from top and bottom chord truss connections, knee bracing, or partially 

restrained tee or angle connections. The restraint is often enhanced by cast-in-place concrete 

cover. The lateral strength and stiffness of these systems is difficult to assess, although some 

testing has been done (Roeder et al., 1996). See also Abrams (1994). Unless the perimeter infill 

is penetrated with large openings, the frame will be far more flexible than the infill. Therefore, 

both in terms of stiffness and strength, the exterior infill walls typically will form the effective 

lateral system for this building type. The effectiveness of the system depends on the size and 

extent of openings and articulation of the plane of the wall. With solid or nearly solid infill 

panels, strut action will be stiff and strong. As openings in panels increase in size, struts or 

combinations of struts cannot effectively form around the opening and the steel columns and 

beams will begin to work as a moment frame, with “fixity” at the beam-column joint provided by 

the masonry. For low and moderate intensity shaking, the exterior walls may provide adequate 

strength to satisfy the specified performance objective. As the shaking demand increases, the 

masonry will tend to crack and spall, losing stiffness and potentially creating a falling hazard. 

The complete steel gravity system, characteristic of this building type, is generally expected to 

provide sufficient stability to prevent collapse, particularly if designed for lateral resistance. 

However, in configurations with large height-to-width ratios, end or corner columns could fail in 

compression or at tension splices, potentially leading to partial collapse. 
Techniques for the Seismic Rehabilitation of Chapter 11 – Types S5/S5A: Steel Frames 

Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

11-4 

This building type is often characterized by a commercial store-front first floor with little or no 

infill at that level on one or more faces of the building. This condition can cause a soft story 

condition or a severe torsional response if open on one or two sides only. Such conditions can 

lead to concentration of seismic deformation at the open level, potentially leading to local P-delta 

failure. This open commercial story was a common feature in many buildings of this type that 

were shaken in the 1906 San Francisco earthquake, but there were no story-mechanisms 

reported. It is speculated that the soft story provided isolation for the upper stories and that the 

displacement demand, for reasons unknown, did not exceed the story capacity. In fact, there 

have been no reports of collapses or damage that suggested imminent collapse in typical U.S. 

multistory office-like steel infill buildings in strong ground motion. Earthquakes providing such 

tests include the 1906 San Francisco, the 1933 Long Beach, and to a lesser extent, the 1994 

Northridge events. In general, current seismic evaluation technology does not reach the same 

conclusion. 

11.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

See Table 11.3-1 for deficiencies and potential rehabilitation techniques particular to this system. 

Deficiencies related to steel moment frames and masonry shear walls are shown in Table 5.3-1 

and Table 18.3-1, respectively. Selected deficiencies are further discussed below by category. 

Global Strength 

The overall strength provided by the exterior walls may be insufficient to prevent serious 

degradation and resulting amplified displacements in the building that can lead to irreparable 

damage or even instability. The strength may be limited by inadequate number of panels of 

infill, excessive openings, or masonry weak in compressive strength. The standard approach to 

such deficiencies will be to add new, relatively stiff lateral force-resisting elements such as 

concrete shear walls or steel braced frames often located on the interior between existing 

columns. Concrete walls can also be added at the perimeter on the inside face of the masonry. 

This procedure is usually conceptualized and analyzed as a concrete shear wall rather than an 

infill to the frame. 

Fiber composite layers also can be added to the face of masonry to enhance infill strut action. 

Although this technique has been tested for increasing shear strength of URM walls, little 

research is available directly on the effects of adding these layers to infill panels. 

Unless the masonry is completely doweled or connected to supporting backing, the damage state 

of the masonry wall must be estimated for the expected drifts of the combined system to 

determine if the desired performance has been achieved. 

Global Stiffness 

For this building type, the methods for adding stiffness are similar to those adding strength. 

Configuration 

Two global configurational deficiencies are common in this building type. The first is a soft and 

weak story at the street level created by commercial occupancies with exterior bays with little or 
Techniques for the Seismic Rehabilitation of Chapter 11 – Ty

Existing Buildings: FEMA 547 with I

11-5 

Table 11.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S5/S5A Buil

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Dem

Global Strength 

Inadequate length of exterior wall 

Interior concrete walls [8.4.2] Interior steel braced frames [8.4.1] 

Concrete wall overlay [21.4.5] Fiber composite wall overlay [21.4.6] 

 Excessive sized 

openings in infill panels 

Interior concrete walls [8.4.2] Interior steel braced frames [8.4.1] 

Infill selected openings [21.4.7] Concrete wall overlay [21.4.5] Fiber composite wall overlay [21.4.6] 

 Inadequate 

columns for overturning forces 

Add cover plates or box members [8.4.3] Encase columns in concrete 

 Weak or 

deteriorated masonry 

Interior concrete walls [8.4.2] Interior steel braced frames [8.4.1] 

Point outside and/or inside wythes of masonry Inject wall with cementitious grout Add concrete or fiber composite overlay on exterior walls pier and/or spandrel [21.4.5], [21.4.6] 

Global Stiffness 

See inadequate strength 

Configuration Soft or weak story Interior concrete walls 

[8.4.2] Interior steel braced frames [8.4.1] 
Techniques for the Seismic Rehabilitation of Chapter 11 – Ty

Existing Buildings: FEMA 547 with I

11-6 

Table 11.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S5/S5A Buil

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Dem

Configuration (continued) 

Torsion from one or more solid walls 

Balance with Interior concrete walls Balance Interior steel braced frames 

 Irregular Plan 

Shape 

Balance with interior concrete walls Balance with interior steel braced frames 

Load Path Out-of-plane 

failure of infill due to loss of anchorage or slenderness of infill 

Provide vertical strongback wall supports [21.4.3] 

Concrete wall overlay [21.4.5] Fiber composite wall overlay [21.4.6] 

 Inadequate 

connection of finish wythe to backing 

Add interwythe tie [21.4.12] 

 Inadequate 

collectors 

Add steel collector on surface of concrete [12.4.3] Embed or add collector in concrete floor slab [12.4.3] 

Strengthen beam to column or beam to beam splices 

Component Detailing 

Inadequate columns splice for tension due to uplift force induced by infill 

Add splice plates Provide splice through added reinforced concrete encasement 

 Inadequate beam 

column connection to resist compression thrust 

Strengthen connection in shear 
Techniques for the Seismic Rehabilitation of Chapter 11 – Ty

Existing Buildings: FEMA 547 with I

11-7 

Table 11.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for S5/S5A Buil

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Dem

Component Detailing (continued) 

Weak or incompletely filled joint between masonry and surrounding steel components 

Repair or fill voids to provide essentially continuous bearing. 

Diaphragms Flat masonry arch 

diaphragm 

Add diagonal steel braced diaphragm under floor [22.2.8] Remove top layers of floor construction and add concrete slab diaphragm 

Add tension ties to prevent loss of arch action [22.2.8] 

Foundation See Chapter 23 [ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 
Techniques for the Seismic Rehabilitation of Chapter 11 – Types S5/S5A: Steel Frames 

Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

11-8 

no infill. This deficiency can be corrected by adding selected bays of infill or by adding shear 

walls or braced frames at this level. The second common issue is a plan torsional irregularity 

created by solid masonry walls on property lines coupled with walls with many openings on 

street fronts. If shown by analysis to be necessary, torsional response can be minimized by 

stiffening the more flexible side of the building with more infill or by the addition of lateral 

elements. In rare cases, the solid walls can be balanced with the open side by selected removal 

of panels or disengagement of the infill strut action. 

Load Path 

The primary load path issue with this building type is to assure that the mass of the exterior walls 

will not become disengaged from the frame which will both prevent infill strut action as well as 

to create a significant falling hazard on the street below. 

In-plane, the articulation of the exterior walls may result in offsets of the wall plane between 

floors. The presence of a complete load path and maintenance of confinement for strut formation 

must be reviewed in such instances. 

If new lateral load-resisting elements are added, existing slab and steel beam construction may 

need to be strengthened to provide adequate collectors. 

Component Detailing 

In order to qualify as an infill lateral force-resisting element, the infill must be installed tight to 

the surrounding steel frame. Loose or incomplete infill can be mitigated with local patching of 

the masonry or by injection of cemetitious or epoxy grout. However, unless the building is 

gutted for remodeling purposes, this procedure will be extremely disruptive. 

The detailing of the steel frame forming the confinement for the masonry is important to achieve 

infill strut behavior. The connection of beam to column must be capable of resisting the strut 

compression forces from the masonry. Many different configurations are possible, each with a 

different potential weakness, but the shear capacity of the beam-to-column connection is often 

critical. In addition, column splices may be inadequate to transfer the overturning forces created 

by strut action. Critical connections normally can be strengthened with steel plates. 

Diaphragm Deficiencies 

A wide variety of concrete diaphragms can be found in this building type. Solid slab-type floors 

will often provide an adequate diaphragm while joisted floors may include only a thin, poorly 

reinforced continuous slab with low shear capacity. The connection of slabs to exterior wall 

should be reviewed because dowels or other positive connections may not have been provided. 

See Chapter 21 on URM construction for discussion of wood diaphragms in this type of building. 

Flat masonry arch floors are problematic. The diaphragm capacity of such built up construction 

has not been established. Damage causing loss of arch action can create falling hazards or 

vertical load failures. Removal and replacement may not be feasible, either from a pure 

economic standpoint or due to historical preservation issues. The added weight of a new 
Techniques for the Seismic Rehabilitation of Chapter 11 – Types S5/S5A: Steel Frames 

Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

11-9 

concrete slab is often difficult to accommodate, even if top layers of the existing floor are 

removed. 

If space is available, a new steel diagonal frame diaphragm can be added underneath such floors. 

FRP can be layered on the masonry arches to better secure them in place. New lateral forceresisting elements can be added to minimize the need for diaphragm action. 

Foundation Deficiencies 

No systematic deficiency in foundations should be expected solely due to the characteristics of 

this building type. 

Other Deficiencies 

Although deterioration of material, in general, is not covered in this document, it is known that 

most buildings of this type have no reliable waterproofing system for the exterior steel framing, 

particularly the columns. Significant damage to columns from water infiltration has been noted 

in several cases, and this condition should be investigated before assuming that the perimeter 

frame is a significant lateral force-resisting element. 

11.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

Most significant recommendations listed in Table 11.3-1 are similar to techniques more 

commonly associated with other building types such as steel framed buildings (S1, S2, or S4), 

unreinforced masonry bearing wall buildings (URM), or general techniques applied to concrete 

diaphragms. Details concerning these techniques can be found in other chapters. 

11.5 References 

Abrams, D.P. (Editor), 1994, Proceedings from the NCEER Workshop on Seismic Response of 

Masonry Infills, Technical Report NCEER-94-0004, National Center for Earthquake Engineering 

Research, Buffalo, NY. 

Roeder, C.W., Leon, R.T., and Preece, F.R., 1996, “Expected Seismic Behavior of Older Steel 

Structures,” Earthquake Spectra, EERI, Vol. 12, No. 4, Oakland, CA, pgs 805-824. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 12 – Type C1: Concrete Moment Frames

12-1 

Chapter 12 - Building Type C1: Concrete Moment Frames 

12.1 Description of the Model Building Type 

These buildings consist of concrete framing, either a complete system of beams and columns or 

columns supporting slabs without gravity beams. Lateral forces are resisted by cast-in-place 

moment frames that develop stiffness through rigid connections of the column and beams. The 

lateral force-resisting frames could consist of the entire column and beam system in both 

directions, or the frames could be placed in selected bays in one or both directions. An important 

characteristic is that no significant concrete or masonry walls are present, or that they are 

adequately separated from the main structure to prevent interaction. Some buildings of this type 

have frames specifically designed for lateral loads, but also have interacting walls apparently 

unaccounted for in the design. These buildings could be classified as moment frames and the 

wall interaction would immediately be considered a seismic deficiency. Alternately, these 

buildings could be classified as Building Type C2f (Shear Wall with Gravity Frames). Older 

concrete buildings may include frame configurations that were not designed for lateral load, but 

if no walls or braces are present, the frames become the effective lateral force system and should 

be included in this building category. Buildings of this type that include integral concrete or 

masonry walls on the perimeter should be considered as Building Type C2f or C3. Floors may 

be a variety of cast-in-place or precast concrete. Buildings with concrete moment frames are 

generally used for most occupancies listed for steel moment frames, but are also used for 

multistory residential buildings. 

Figure 12.1-1: Building Type C1: Concrete Moment Frames 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 12 – Type C1: Concrete Moment Frames

12-2 

Variations Within the Building Type 

The primary variation within this type is the type of frame and the number of frames included. 

Frames can range from column-girder systems of one bay on each face of the building to systems 

that employ every column coupled with two-way slabs. Frames classified by code as ductile or 

semi-ductile by code beginning in the late 1960s and early 1970s are far more constrained in 

configuration due to prescriptive rules governing girder configuration, strong column-weak 

beam, and limitations on joint shear. 

Floor and Roof Diaphragms 

The floor and roof diaphragms in this building type are essentially the same as the bearing wall 

system, and are almost always cast-in-place concrete. The diaphragms are stiff and strong in 

shear because the horizontal slab portion of the gravity system is either thick or frequently 

braced with joists. However, one way joist systems could be inadequate in shear in the direction 

parallel to the joists. Collectors are seldom in place and transfer of load from diaphragm to shear 

wall must be carefully considered. 

Foundations 

There is no typical foundation for this building category. Foundations could be found of every 

type depending on the height of the building, the span of the gravity system and the site soil. 

12.2 Seismic Response Characteristics 

This building type must be separated into older frame systems, often not even designed for 

lateral loads and including few, if any, features that would assure ductile behavior, and frames 

specifically designed to exhibit ductility under seismic loading. Rules for design of ductile 

concrete frames were developed during the 1960s. 

Older, non-ductile frame buildings, assuming an insignificant amount of concrete or masonry 

walls are present, will be far more flexible than other concrete buildings, and will probably be 

relatively weak. Most importantly, columns are often not stronger than beam or slab system, 

forcing initial yielding in these key elements. In addition, unless spiral ties were used, the 

column will typically fail in shear before a flexural hinge can form. Buildings with these 

characteristics are among the most hazardous in the U.S. inventory and are in danger of collapse 

in ground motion strong enough to initiate shear failures in the columns. Buildings of this type 

that are configured such that initial hinging occurs in the floor system will exhibit stiffness and 

strength degradation and large drifts, but unless exceptionally weak, are far less likely to 

collapse. The ratio of the inherent strength of the frame—designed for lateral loads or not—

compared to the seismic demand has a large influence on the performance, and frames in low 

and moderate seismic zones may be at less risk for this reason. 

Semi-ductile frames, with some but not all of current design features for concrete frames, likely 

will perform better, particularly if the columns are protected by basic strength and are designed 

to be flexurally controlled. However, many of these early concrete frames may be excessively 

weak and suffer from high ductility demands which could have serious consequences if a soft or 

weak story is present due to architectural configuration or column layout. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 12 – Type C1: Concrete Moment Frames

12-3 

Buildings with “fully ductile” frames are expected to perform well, unless vertical or horizontal 

configuration irregularities concentrate inelastic deformation on certain structural components. 

12.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

See Table 12.3-1 for deficiencies and potential rehabilitation techniques particular to this system. 

Selected deficiencies are further discussed below by category. 

Global Strength 

Although lack of ductility is the overwhelming deficiency for this building category, low 

strength may contribute to poor performance. It is difficult to add significant strength within the 

confines of the existing frames and most often new elements of braced frames or shear walls are 

added in these buildings. 

Global Stiffness 

See Global Strength. 

Configuration 

The most common configuration issue in this building type is a soft or weak story created by a 

non-typical story height. If the building is not to receive new walls or frames as part of a global 

retrofit, such configuration deficiencies can be minimized or eliminated with local strengthening 

of columns. 

Load Path 

There are no load path issues particular to this building type. 

Component Detailing 

The major deficiencies of this building type are due to inadequate component detailing, namely 

the structural components of the frame. Current requirements for “ductile frames” include 

capacity design techniques to assure flexural yielding in both girders and columns, as well as, for 

the most part, to limit yielding to the floor system. Retrofit procedures to obtain this ductile 

behavior of the frames are difficult, disruptive, and expensive, and are therefore seldom done. In 

high seismic zones, retrofit of these buildings is normally accomplished by adding new, stiffer 

lateral force-resisting elements that prevent significant ductility demand on the frames. 

Some research has been completed to investigate methods of retrofit for concrete moment frames 

(see Section 12.4.6), and in lower seismic zones where demands over and above gravity designs 

are not great, local strengthening and confinement of frame elements may be practical. 

Diaphragm Deficiencies 

The most common diaphragm deficiency in this building type is a lack of adequate collectors. 

The addition of effective collectors in an existing diaphragm is difficult and disruptive. Existing 

strength to deliver loads to the shear walls should be studied carefully before adding new 

collectors. 
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Table 12.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C1 Buildi

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Deman

Global Strength 

Insufficient number of frames or weak frames 

Concrete/masonry shear wall [12.4.2] 

Steel braced frame [12.4.1] 

Concrete or steel moment frame 

Steel moment frame 

Increase size of columns and/or beams [12.4.5] 

Remove upper story or storie[24.2] 

Seismically isolate [24.3] 

Supplemental damping [24.4

Global Stiffness 

Insufficient number of frames or frames with inadequate stiffness 

Concrete/masonry shear wall [12.4.2] Steel braced frame [12.4.1] Concrete or steel 

moment frame 

Increase size of columns and/or beams [12.4.5] 

Fiber composite wrap of gravity columns [12.4.4] 

Concrete/steel jacket of gravity columns [12.4.5] 

Provide detailing of all other elements to accept drifts 

Supplemental damping [24.4

Configuration Soft story or 

weak story 

Add strength or stiffness in story to match balance of floors 

 Re-entrant 

corner 

Add floor area to minimize effect of corner 

Provide chords in diaphragm 

 Torsional layout Add balancing 

walls, braced frames, or moment frames 
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Table 12.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C1 Buildi

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Deman

Configuration (continued) 

Incidental walls failing or causing torsion 

Add balancing walls, braced frames, or moment frames 

Uncouple incidental walls 

Convert incidental walls to lateral elements walls 

Load Path Inadequate 

collector 

Add or strengthen collector [12.4.3] 

Component Detailing 

Lack of Ductile detailing--general 

Perform selected improvements to joints [12.4.6]

Seismic isolat[24.3] 

 Lack of ductile 

detailing: Strong columnweak beam 

Jacket columns [12.4.4] 

 Lack of ductile 

detailing: Inadequate shear strength in column or beam 

Fiber composite wrap [12.4.4] 

Concrete/steel jacket [12.4.5] 

 Lack of ductile 

detailing: Confinement for ductility or splices 

Fiber composite wrap [12.4.4] 

Concrete/steel jacket [12.4.5] 

Diaphragms Inadequate inplane shear 

capacity 

Concrete or masonry shear wall [12.4.2] 

Braced frame [12.4.1] 

Moment frame 

R/C topping slab overlay 

FRP overlays [22.2.5] 
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Table 12.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C1 Buildi

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Deman

Diaphragms (continued) 

Inadequate chord capacity 

New concrete or steel chord member [12.4.3] 

 Excessive 

stresses at openings and irregularities 

Add chords [12.4.3] 

Foundations See Chapter 23 

[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 
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12.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

12.4.1 Add Steel Braced Frame (Connected to a Concrete Diaphragm) 

Deficiencies Addressed by the Rehabilitation Technique 

Inadequate global shear capacity 

Inadequate lateral displacement (global stiffness) capacity 

Description of the Rehabilitation Technique 

Addition of steel diagonal braced frames to an existing concrete moment frame building is a 

method of adding strength and/or stiffness to the structural system. The steel braces can be added 

without a significant increase in the building weight. The new braces will commonly be some 

configuration of concentric braced frame (CBF); it is very uncommon to use an eccentrically 

braced frame (EBF) due to costs and difficult detailing issues associated with the link 

mechanism. Any of a variety of diagonal brace configurations may be used, as well as a variety 

of brace member section types. Figure 12.4.1-1 shows several common configurations. 

Common connections of the new brace to the existing concrete structure are shown in Figures 

12.4.1-2A, 12.4.1-2B, and 12.4.1-2C. 

Design Considerations 

Research basis: Design of the lateral force-resisting system for the building should account for 

the stiffness of both the braced frame system and the existing concrete moment frames. While 

basic research regarding adding braced frames at the interior of a concrete moment frame 

building has not been identified, research in the 1980s at the University of Texas at Austin on 

frames at the exterior façade demonstrated the ability of the new steel braced frames to increase 

the deformation capacity of the non-ductile concrete frames (Jones and Jirsa, 1986). A 

schematic detail of the connection used in this testing is shown in Figure 12.4.1-3. 

Braced frame – concrete frame interaction: Most designs of braced frame retrofits will be 

governed by maintaining drifts within the range of acceptability for the existing concrete 

elements. This can be accomplished by setting up a model that includes both the stiffness of the 

braced frame and of the concrete frame and meeting acceptability requirements for the 

displacements (or psuedo forces) in the concrete elements. Some engineers prefer to consider 

only the braced frames as a new lateral system, determine real drift demand for that system, and 

then check that drift for acceptability superimposed on the existing frame 

In taller buildings, the possible incompatibility between vertical cantilever behavior of discrete 

braced frames and the existing moment frames must be assessed. Existing beams or slabs, if 

unusually thick, that frame directly into the ends of new braces may restrain the global flexural 

deformation of the brace and require special consideration. Finally, due to the wide variety of 

nonlinear behavior of braced frames that is dependent on configuration and detailing, it may be 

difficult to obtain an adequate understanding of overall deformation compatibility using linear 

methods. 
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Figure 12.4.1-1: Typical Braced Frame Configurations 

Braced frame location: The new braces may be located on the exterior or interior of the building. 

An exterior location generally allows for easier construction access and perhaps less cost, but is 

visible, exposed to the environment and probably will impact exterior building finishes. Braces 

placed parallel to the façade can be connected to the exterior faces of perimeter spandrel beams, 

perimeter moment frames or edges of floor and roof diaphragms relatively easily, but will most 

likely cross in front of some windows. Alternatively, exterior bracing may be placed as 

buttresses, perpendicular to the existing façade. This configuration will probably require more 

extensive new collectors to deliver lateral forces from the diaphragms but may allow creation of 

new stair or elevator shafts, or perhaps additional floor area. For projects that include expansion 

of or additions to the existing building, the new braces could be located in the adjacent new 

construction, tied to the existing building. 
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Figure 12.4.1-2A: Typical Connection to Concrete Diaphragm 

Interior braces will most commonly be located along existing frame lines, particularly at moment 

frame bays. This will allow for best use of any existing diaphragm chords and collectors and for 

best moment frame – braced frame interaction. In some cases however, interior braces will be 

located offset from existing column-frame lines to minimize direct impact on existing structural 

or architectural components or to simplify the frame-diaphragm connections. 

The addition of new braced frames to a building will always impact the architectural character 

and functional uses of the building to some degree. Selection of preferred brace locations must 

be made considering these issues, such as space layout, corridor locations, doorways, windows, 

main M/E/P distribution runs, as well as the structural or construction considerations. 

Braced frame configuration and member section type: In most cases where diagonal steel braces 

are used to strengthen or stiffen a concrete frame building, a complete braced frame including 

horizontal beam and column members, as well as the diagonal braces themselves, is employed. 

Installation of diagonal bracing members between existing concrete columns is difficult because 

transfer of a large concentrated axial force from the concrete members through a localized 

connection with a limited number of anchors is rarely feasible. The steel columns are often 

continuous, passing through the floors, from foundations up to the roof or highest level required 

to avoid transfer of load from the steel system in and out of the concrete at each floor. In some 

cases, columns can be connected to adjacent concrete columns, but if the concrete column 

becomes part of the primary chord, reinforcing splice locations must be carefully considered. 
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Figure 12.4.1-2B: Typical Connection to Existing Concrete Beam 

New steel horizontal elements are similarly needed to facilitate the connections of the diagonal 

and to transfer forces from each floor into the frame. These steel elements are generally placed 

below the floor and roof diaphragms or adjacent to beams or spandrels. The diagonal steel 

braces may be placed in any of the commonly used configurations indicated in Figure 12.4.1-1; 
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Figure 12.4.1-2C: Typical Connection to Existing Concrete Column 
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Figure 12.4.1-3: Test Specimen Connection Detail for Braced Frame 

single diagonal or X-shaped, V-shaped, chevron (inverted-V) shaped, or “super-X” shaped (a 

combination of chevron and V braces in alternate stories forming a two-story X shape). Twostory X-bracing has the advantage over V- or inverted-V-bracing should a compression brace 

buckle. In the latter configurations, the remaining tension brace has an unbalanced vertical 

component that has to be resisted by the beam. For an X-bracing configuration, even if a 

compression brace buckles, the force in the remaining tension brace is transmitted directly to the 

tension brace on the opposite side of the beam. 

Configuration will be selected based on consideration of structural issues, relative strength, 

stiffness and performance, as well as of several other issues including aesthetics, conflicts with 

doorways, corridors or windows, M/E/P systems, or the number of connections and penetrations. 

Column and beam members are often W-shapes, but may be other shapes such as channels or 

hollow structural section (HSS) tubes to improve aesthetics or to ease detailing. Diagonal 

members may be of any typically used sections including W-shapes, hollow (HSS) pipes or 

tubes, or double channels, angles or HSS tubes. 
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Buckling-Restrained Braced Frames (BRBFs), in which steel plates or cruciform shaped braces 

are surrounded by unbonded concrete in such a way as to prevent bucking of the brace, act 

essentially the same in tension and compression. The yield strength of a bay braced with one or 

more of these braces can be relatively accurately predicted. In situations where many, lightly 

loaded braces will be employed, sufficient global strength can be obtained by designing the 

braces to yield prior to yielding or other failure of the existing columns, preventing the need to 

retrofit the columns. 

Detailing Considerations 

Connection to existing concrete floor and roof diaphragms: A significant concern associated 

with installing a new steel braced frame in a concrete building is the connection of the beam at 

the top of the frame in each story to the underside of the existing concrete diaphragm overhead. 

The primary concern is that a relatively large shear force must be transferred from the overhead 

diaphragm into the new steel bracing below through a relatively localized connection using 

discreet anchors/bolts. The connection is generally made by one or more rows of concrete 

anchors as shown in Figure 12.4.1-2A. Typically, the anchors are threaded rods set in epoxy, but 

drilled expansion anchors may be used if they provide sufficient force transfer capacity and 

adequate testing to show they can resist cyclic loading. An alternate connection method, 

installed from the top down, consists of providing large holes in the concrete slab to expose the 

steel beam sufficiently to installed welded dowels to the top flange. The hole is then backfilled 

with cementitious or epoxy grout. In many cases however, the shear capacity of the existing 

concrete diaphragm is inadequate to deliver the relatively large shear force within the length of 

the braced frame. In those cases, a collector will be required (refer to Section 12.4.3). 

Connection to existing moment frames: New braced frames are often located on or alongside of 

the existing moment frame lines. This generally allows for better use of the existing collectors 

(beams) to deliver diaphragm forces to the bracing and, perhaps, use of the existing frame 

columns and footings to help resist overturning and uplift forces. It is generally preferable to 

locate the new braces alongside of the existing moment frames instead of as an “infill” within the 

width of the existing concrete frame beams and columns. 

For diagonal braces installed in an “infill” configuration, it is often extremely difficult to transfer 

large seismic forces from the surrounding concrete members through very localized connections 

with a limited number of discreet anchors. Also, if steel columns or vertical members are used in 

the “infill” frame, it is virtually impossible to provide vertical continuity from floor to floor 

through the existing concrete beams. Furthermore, if connections of sufficient strength can be 

made, the anchors must be threaded into or through the relatively densely reinforced beams and 

columns and, where collector strengthening is required, the added collector components will not 

connect directly to the new braces. In addition, physical installation and fit-up of the new braces 

and their connecting gusset plates often becomes significantly more difficult. 

These detailing difficulties can be reduced or avoided by placing the new braced frame alongside 

of the concrete moment frame. In most cases, placing the new bracing alongside an exterior 

frame will allow the greatest ease in detailing. Bracing in this location will almost always require 

installation of a complete new steel braced frame instead of only the new diagonal braces 

themselves. In this configuration, the connection of the concrete diaphragm to steel braced frame 
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can be made as discussed above or by installing anchors into the side of the adjacent concrete 

frame beam as shown in Figure 12.4.1-2B. Braced frame overturning forces are carried directly 

by the new steel column members. However, if concrete beam framing occurs in two directions, 

the new steel columns will generally need to be offset from the existing concrete column on a 

45-degree diagonal to provide continuity of the new column through the floors without 

interference with the existing concrete beams. Overturning resistance can be obtained by 

connecting the new steel column to the existing concrete column (see Figure 12.4.1-2C1) and 

footing. For cases where the new braced frame can be placed on the exterior of the building, the 

new steel columns can be continuous and the connections to the adjacent concrete columns or 

pilasters can be made with relative ease. If diaphragm collector strengthening is required, the 

additional collector can be installed alongside of the existing beam line and can be connected 

directly to the new braced frame. 

Exterior bracing at offset columns: There are some buildings where the exterior columns 

protrude farther out than the exterior beams. The Jones and Jirsa (1986) research can be applied 

in these situations, where the new steel framing is placed adjacent to the beams and in the plane 

of the outer portion of the protruding concrete column as shown in Figures 12.4.1-3 through 

12.4.1-5. The primary challenges lie in connecting the two types of frames and delivering loads 

into the braced frames. As an alternative to drilling numerous holes for bolts or dowels into the 

concrete columns, steel lugs can be provided at each floor. In this approach, steel pipes are 

inserted through cores drilled through the concrete columns and filled with grout. Next, the 

pipes are welded to steel plates on the sides of the concrete columns, which then provide surfaces 

for welding to the columns of the braced frames. An example of this connection is shown in 

Figure 12.4.1-4. If required, horizontal forces can also be transferred directly to the braced 

frames through the braced frame beams. The beams are welded to steel plates, which are 

connected to the concrete slab or beams at the building perimeter with dowels, bolts, or lugs, as 

shown in Figure 12.4.1-5. 

Exposed exterior braced frames require simple and clean connections that fit the architectural 

character of the building. Use of W-shapes for the braced frame members can eliminate gusset 

plates and allow direct connection of the members through complete joint penetration welds. 

Shop welding of the connections and on-site prefabrication of the braced frames will minimize 

field welding on the structure. W-shapes also simplify other architectural issues by not allowing 

rainwater or debris to accumulate. 

Installation of additional collectors: Installation of new braced frames in a concrete frame 

structure, especially in one with a distributed frame system, will often result in increased 

diaphragm demands at the individual braces. An advantage of locating the new brace at an 

existing frame line is that the existing beams can then be used as a collector. However, 

insufficient continuity and/or laps of reinforcing steel combined with highly concentrated 

diaphragm demand may still require strengthening of the existing collector (refer to 

Section 12.4.3). 

Footings: Addition of steel braced frames to an existing building will almost always require 

construction of new footings, or augmentation of existing ones, to resist the concentrated 

overturning demand. In many cases, the overturning uplift demand will require installation of tie 
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Figure 12.4.1-4: Braced Frame to Concrete Column Connection 

downs. Alternatively, the new frame can be located between two existing column frame lines, 

instead of directly on or along one frame line, and new foundations or grade beams can be used 

to engage more than one existing column to resist the uplift demand. 

Cost/Disruption Considerations 

The cost and level of disruption associated with installation of steel bracing is generally less than 

that of shear walls. The number of penetrations that need to be cut through the existing concrete 

structure and of drilled dowels and anchors may be less than for the shear wall alternative, and 

the work is generally not as wet or messy. Also, it will not be necessary to prepare any existing 

concrete surfaces that will be in contact with new steel members. The new members are discrete 

and welded or bolted connections are localized. However, there will be noise and vibrations 

resulting from the required cutting and drilling that will make continued occupancy difficult. 
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Figure 12.4.1-5: Braced Frame to Slab Connection 

Invariably, some architectural and M/E/P system components will require relocation or 

replacement. 

Construction Considerations 

The primary construction consideration will be fit up and installation of the steel braces and their 

connections. The desire to limit the number of splice connections must be balanced against the 

difficulties of installing longer members such as multistory columns. Installation of the diagonal 

braces will require careful planning and will often require member splices in the field. 

Installation of drilled threaded rod or expansion anchors will require some precision and 

extensive use of templates and oversized holes, to assure proper fit with the steel members. In 

some cases, the steel members themselves could be used as the template for the anchors. 

Proprietary Concerns 

In general, there are no proprietary concerns related to installation of steel diagonal braced 

frames in a building. The one exception, however, occurs if buckling-restrained braces are used. 
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12.4.2 Add Concrete or Masonry Shear Wall (Connected to a Concrete 

Diaphragm) 

Deficiencies Addressed by the Rehabilitation Technique 

Inadequate global shear capacity 

Inadequate lateral displacement (global stiffness) capacity 

Description of the Rehabilitation Technique 

Addition of shear walls to an existing concrete frame building is a common method of adding 

significant strength and/or stiffness to the structure. The new walls may be of cast-in-place 

concrete, shotcrete or fully grouted concrete masonry unit (CMU) construction. 

Design Considerations 

Research basis: No research focused on the overall effects of adding shear walls to existing 

concrete frames has been identified. The effects of surface preparation, concrete strength, and 

interface reinforcement on interface shear capacity between new and existing concrete were 

examined by Bass, Carrasquillo, and Jirsa (1985). These tests indicate that surfaces prepared 

with heavy sandblasting exhibit shear capacities greater than or equal to those exhibited by 

chipped surfaces or surfaces prepared with shear keys. Increased concrete strength resulted in 

increased interface shear capacity for chipped surfaces and those prepared with shear keys, but it 

had little effect on the shear capacity of interface surfaces prepared by sandblasting. Specimens 

in which drypack mortar was used exhibited a significantly smaller shear capacity than those 

where new concrete was cast directly against the interface. Increasing the amount or embedment 

depth of reinforcement across the interface resulted in greater interface shear capacity. 

Frame-wall interaction: Most designs of shear wall retrofits will be governed by maintaining 

drifts within the range of acceptability for the existing concrete frame elements. This can be 

accomplished by setting up a model that includes both the stiffness of the shear walls and of the 

concrete frame and meeting acceptability requirements for the displacements (or psuedo forces) 

in the concrete frame elements. Some engineers prefer to consider only the shear walls as a new 

lateral system, determine the expected drift demand for that system and then check that drift 

superimposed on the existing frame for acceptability. 

In taller buildings, the possible incompatibility between vertical cantilever behavior of discrete 

shear walls and the existing concrete frames must be assessed. Existing beams or slabs, if 

unusually thick, that frame directly into the ends of new walls may restrain the global flexural 

deformation of the wall and require special consideration. Two shear walls are often purposely 

placed in line and connected by a short beam to form a coupled shear wall system. In this 

system, the coupling beams are specially designed to accept significant inelastic deformations. 

Seldom can two such walls utilize existing beams as coupling beams due to inadequate detailing. 

Thus coupling beams, when employed, are installed new, as part of the system.

Frame-wall configuration: A primary design consideration is determination of whether or not the 

existing concrete frames may be used as an effective part of a combined system. Are the existing 

frame columns strong enough and/or well detailed enough to serve as the chord/boundary 

member of a shear wall without improvement? Are the frame beams detailed well enough to 
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serve as coupling beams or to be incorporated into the wall itself? These considerations may 

limit the choices of wall-frame physical relationship: that is, should the walls be placed 1) 

within the plane of the existing concrete frames, 2) as vertically continuous walls alongside of, 

and joined to, the existing frames, or 3) as separate vertical elements independent of the frames? 

The first alternative is often best avoided as noted in the Detailing Considerations discussion 

below. Considering alternates 2 and 3, it must be determined if the frames are capable of 

becoming part of the shear wall (primarily as chord elements) or if it is beneficial to prevent 

direct interaction by placing the shear walls free of the existing concrete frame elements. In 

some cases, it is not feasible to stiffen the building into the range of acceptable deformation of 

the existing frames, and improvement in deformation capacity may be required in addition to the 

addition of new walls. 

Wall location: The new walls may be placed on the exterior or interior of the building. An 

exterior location generally allows for easier construction access and perhaps less cost, but is 

visible, exposed to the environment and may impact exterior building finishes. Walls placed 

parallel to the façade can be connected to the exterior edges of floor and roof diaphragms or 

perimeter concrete frames relatively easily, but will most likely require closure or reduction in 

size of some windows. Alternatively, exterior walls may be placed as buttresses perpendicular to 

the existing façade. This configuration will probably require more extensive new collectors to 

deliver lateral forces from the diaphragms but may allow creation of new stair or elevator shafts, 

or even of additional floor area. For projects that include expansion of or additions to the existing 

building, the new walls could be located in the adjacent new construction. 

Interior walls located along frame lines, particularly at moment frame bays will often allow for 

best use of any existing diaphragm chords and collectors. Beams that frame directly into the ends 

of new walls may behave like coupling beams as described above. In some cases, interior walls 

are better located offset from existing column-frame lines to minimize direct impact on existing 

structural or architectural components or to simplify the wall-diaphragm connections. 

The addition of shear walls to a building will always impact the architectural character and 

functional uses of the building. Selection of preferred wall locations must be made considering 

these issues, such as space layout, corridor locations, doorways, windows, main M/E/P 

distribution runs, as well as the structural or construction considerations. 

Detailing Considerations 

Connection to existing concrete floor and roof diaphragms: Arguably, the most significant detail 

associated with installing a new shear wall in a concrete building is the connection at the top of 

the new wall to the underside of the existing concrete diaphragm overhead. The construction 

joint must be made tight, without any gapping, to facilitate transfer of shear forces from the 

overhead diaphragm into the new wall below and to minimize the possibility of joint slip. See 

the discussion under the Research basis section. 

Typical details of this connection for a new cast-in-place concrete wall below an existing 

concrete flat slab are shown in Figures 12.4.2-1A and 12.4.2-1B. The vertical dowels must be 

sufficient to transfer forces from the existing diaphragm and from the new wall above (if it 

exists), to the lower wall. Shears can also be transferred across this joint with large diameter 
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Figure 12.4.2-1A: Concrete Wall Connection to Concrete Slab 

pipes or structural shapes. The holes made through the existing slab must serve not only to 

install the dowels, but also to allow for placement and consolidation of the wall concrete. The 

concrete head created by placement up to the top of slab coupled with cleaning and roughening 

the existing concrete contact surface by either sandblasting or chipping will provide the best joint 

available. The larger holes through the slab will also be more like intermittent shear keys. The 

holes should be drilled or made with impact tools instead of saws or core drills to avoid cutting 

or damaging existing slab reinforcement. Prior to cutting the holes, temporary shores may be 

required below the slab along each side of the row of holes. The concrete should be placed 

through the slab openings into the forms below, up to top of slab, to provide some head on the 

joint at the underside of the diaphragm. 
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Figure 12.4.2-1B: Concrete Wall Connection to Concrete Slab – Partial Elevation View 
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If the new wall is shotcrete, special care is required by the nozzle operator when placing the 

shotcrete directly at the underside of the slab to provide a tight, well-bonded joint free of 

rebound or gaps. To minimize the possibility of creating a gap due to sagging, the last lift of 

shotcrete should be a short one. In the end, however, such a well-bonded joint is often not 

achieved at the slab soffit and remedial work, similar to crack-injection repairs, is likely to be 

needed. As an alternative, the holes through the slab needed for placement of the vertical rebar 

dowels could be made oversized, sufficient to allow placement of pourable, cementitious grout at 

the top of the shotcrete wall below, similar to the cast-in-place concrete alternative. 

For CMU wall construction, the masonry units will typically be constructed up to within one or 

two courses of the overhead slab soffit, leaving enough of a gap to allow placement of the upper 

lift of grout. Preferably, the gap should be formed and grouted from above through holes in the 

slab similar to cast-in-place concrete alternative described above. Consolidation of the upper lift 

of grout should be performed through holes in the slab above. Although the gap can be dry 

packed from below, this is a considerably less effective alternative as confirmed by the research 

results noted above. 

Regardless of whether the new wall is cast-in-place concrete, shotcrete or CMU, some shrinkage 

or sagging will probably occur creating a crack at the joint. To account for the resulting 

reduction in effective aggregate interlock along this joint, it may be prudent to use a lower 

coefficient of friction, and increase the size the vertical dowels. 

Figure 12.4.2-1C shows the conditions where the existing concrete diaphragm is in a pan joist or 

waffle slab system instead of a flat slab. For these types of floor or roof systems, the joists or 

waffle ribs must be preserved to avoid shoring. However, there is likely to be more flexibility in 

the extent of the openings that can be made through the slab between the joists/ribs, and 

temporary shoring will generally not be required. Where the new wall is parallel to the joists, it is 

preferable to locate the wall offset from the joist as shown in the detail. The holes in the slab may 

be made as intermittent keys, similar to the flat slab condition discussed above, or they can be 

made as relatively long slots or as a continuous opening the length of the wall. Additional 

diaphragm to wall shear transfer capacity can be obtained by doweling into the side of the 

adjacent rib. 

Where the new wall is perpendicular to the joists, or at a waffle slab condition, the slab can be 

removed between the ribs as indicated in Figure 12.4.2-1D. Since installation of continuous 

horizontal wall bars through the perpendicular ribs is generally not possible, installation of one 

or two horizontal hoop ties may be required at the upper portions of the wall between the ribs. 

For CMU wall construction, the masonry will stop below the joists or ribs, and the large vertical 

gap up to the slab, between the ribs, will be completed with poured concrete. 

Connection to existing frames: New walls are often located on or alongside existing frame lines. 

This generally allows for better use of the existing diaphragm collectors (beams) and of the 

existing frame columns as wall chords or boundary elements. It is almost always preferable to 

locate the wall alongside the frame beams instead of as an “infill,” within the width of the frame 
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Figure 12.4.2-1C: Connection of Concrete Wall to Concrete Joists or Waffle Slab 

beams and columns. When placed alongside the frame, the wall-diaphragm connections are as 

discussed above, and additional shear transfer and wall chord capacity can be obtained by 

doweling into the side of the beam and the column, respectively. Also, if diaphragm collector 

strengthening is required, the additional collector can be installed alongside the existing beam 

line and will be lead directly to the new wall. In the “infill” configuration, the vertical wall 

dowels must be threaded through the relatively densely reinforced beams, concrete placement 

and consolidation becomes significantly more difficult, and additional collectors do not connect 

directly to the new wall. 
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Figure 12.4.2-1D: Concrete Wall Connection to Waffle Slab – Partial Elevation View 

If the existing columns have sufficient strength and appropriate reinforcement detailing, they 

may be used as the wall chord or boundary element, by doweling into the column. The 

effectiveness of this is limited by the amount of doweling that can be installed. In many cases, 

however, the existing column will require strengthening or jacketing, or new wall chords will be 

needed. 

Installation of additional collectors: Installation of shear walls in a frame structure, especially in 

one with a complete frame system, will result in increased diaphragm demands at the individual 

walls. An advantage of locating the new wall at an existing frame line is that the existing beams 

can then be used as a collector. However, insufficient rebar continuity and/or laps combined with 

highly concentrated diaphragm demand may still require strengthening. Refer to Section 12.4.3. 

Footings: Addition of concrete or masonry shear walls will almost always require construction 

of new footings, or augmentation of existing ones, to support the added weight as well as to 
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resist the increased and/or concentrated overturning demand. In many cases, the overturning 

uplift demand will require installation of tie downs. Where the new wall is located between 

column frame lines, instead of directly on or along one frame line, new foundations can be used 

to engage more than one column to resist the uplift demand. 

Cost/Disruption Considerations 

In general, shotcrete walls are less expensive than cast-in-place concrete because at least one side 

of the wall forming is eliminated. If shotcrete can be applied against an existing wall at stair, 

elevator or mechanical shafts, the cost savings of shotcrete is even greater. CMU walls are 

generally less costly, per square foot, than either shotcrete or cast-in-place concrete walls. 

However, CMU walls may not provide comparable strength or stiffness, requiring the addition of 

more linear feet of CMU walls than either cast-in-place concrete or shotcrete walls. 

Construction of new shear walls in an existing building can be very disruptive to any building 

occupants. Noise, vibration, and dust associated with many operations, especially cutting holes 

through and drilling dowels into concrete, can be transmitted throughout a concrete structure. 

Placing cast-in-place concrete, shotcrete or even grouted masonry is a wet process and very 

messy. Shotcreting in an enclosed area creates differential pressures that can spread debris 

beyond nominal construction barriers. Also, excavation and drilling operations and the use of 

mechanized and/or truck mounted equipment associated with installation of new foundations can 

be very disruptive. 

Construction Considerations 

The existing concrete surfaces to be in contact with the new concrete walls should be cleaned of 

all finishes, paint, dirt, or other substances and then be roughened to at least attempt to provide 

1/4” minimum amplitude aggregate interlock at joints and bonded surfaces. At overhead joints 

where such preparations may be less effective, as discussed in the Detailing Considerations

section above, additional dowels can be used with less roughening. 

For shotcrete applications, separate trial test “panels” at the overhead joints should be included 

with the normal preconstruction test panels. These test joints should be cored to inspect the 

adequacy of the surface preparation and the joint bond. Nozzle operators should have several 

years experience with similar structural seismic improvement applications. 

In addition to the usual concrete/shotcrete core sampling and testing, the overhead joints should 

be cored to allow inspection of the joint quality and determine whether or not repairs are needed. 

For CMU shear walls, practical limitations on placement of wall reinforcing steel must be 

considered. In particular, use of “seismic comb” type of joint reinforcement (a prefabricated 

mesh of welded wire reinforcement used as transverse reinforcement at boundary elements of 

CMU walls) has often proven to be very difficult to install and the resulting rebar congestion 

interferes with grouting operations. 
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12.4.3 Provide Collector in a Concrete Diaphragm 

Deficiencies Addressed by the Rehabilitation Technique 

Inadequate or missing collector 

Inadequate diaphragm chord capacity 

Description of the Rehabilitation Technique 

Addition of a new collector or strengthening of an existing collector is often needed when new 

steel braced frames or concrete shear walls are added to an existing building. The new collector 

must extend as far as necessary, often one or more bays from one or both ends of the new brace 

or wall, to draw the required shear demand from the existing diaphragm. The new collector will 

be constructed of reinforced concrete or steel, generally depending on whether the general 

building upgrade involves installation of new concrete shear walls or steel braced frames. The 

new collector will most often be installed at the underside of floor. At roofs, the collector may be 

placed either from below or above the roof. 

In cases where the existing diaphragm chord is absent or inadequate, the mitigation approach 

will be similar to that used for collectors. 

Design Considerations 

Research basis: For new reinforced concrete collectors, see the discussion of tests by Bass, 

Carrasquillo, and Jirsa (1985) in Section 12.4.2. 

For new steel collectors, Jiménez-Pacheco and Kreger (1993) tested single anchor connections 

between existing concrete and new steel members in order to examine shear transfer along the 

interface between these two elements. Results indicate that sandblasting the steel surface and 

applying a layer of epoxy at the interface between steel and concrete can substantially increase the 

force level at which the interface begins to slip. Also, the use of spring washers may reduce longterm anchor bolt relaxation, maintaining the first-slip force capacity over time. For applications 

where significant inelastic deformations are expected, a thick layer of nonshrink grout between the 

steel and concrete was found to increase deformation capacity, though it decreased ultimate strength 

slightly. Filling the annulus between the bolt and washer with epoxy resulted in greater connection 

stiffness than that exhibited by specimens with unfilled annuli or those filled with nonshrink grout. 

Material selection - reinforced concrete or steel: In reinforced concrete buildings with some sort 

of concrete slab floor system, especially one with joists, waffle ribs or beams crossing the path of 

the collector, the most common material choice for the new collector is reinforced concrete. 

Often, this choice is made because concrete is aesthetically compatible with the surrounding 

structure, especially in a condition exposed to view. However, concrete is selected principally 

because it is compatible with the deformation characteristics of the diaphragm it is connected to. 

A concrete collector is bonded to, and is integral with, the concrete slab diaphragm and the strain 

deformations of the collector are the same as the deformations of the diaphragm system. At a 

steel plate collector, the elongation of the plate is not compatible with the diaphragm slab. As the 

collector load accumulates towards the connection to the new wall or brace, the elongation of the 

plate accumulates as well. The threaded rod anchors connecting the plate collector to the 
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diaphragm in the zone of greatest elongation can become overloaded to failure by the plate 

bearing on the bolts. This can lead to a “zipper-like” failure mode as the adjacent anchors assume 

the load of the failed anchors and become overloaded in turn. This behavior can occur even at 

relatively short collectors if the elongation exceeds the available annulus gap around the anchor. 

To avoid this, special detailing is required as discussed in the Detailing Considerations section 

below. 

Impact on architectural and M/E/P systems and components: A new collector often must extend 

one or more entire bays away from the new wall or brace in order to draw the necessary load 

from the existing concrete diaphragm. Installation of the new collector at the underside of the 

existing floor slab impacts any existing ceilings, partitions, ductwork, plumbing, lighting, etc., 

located along its entire length. As a result, the new collectors will often have a greater impact on 

the building’s other systems than the new walls or braces themselves. Furthermore, consideration 

of these impacts will often affect placement of the new walls or braces. In many cases, the new 

walls and their associated collectors are located along the exterior edge of the building 

specifically to avoid or minimize these impacts on other building systems, especially in a case 

where building occupancy is maintained during the construction. 

In some cases, it may be possible to locate the new collector at the top surface of the existing 

diaphragm. At roofs, new collectors can be placed on top of the roof diaphragm, provided that 

any conflicts with roof mounted equipment, pads or penthouses can be accommodated or 

avoided. More importantly, placement of collectors on top of the roof slab requires careful 

consideration of the impact on roof drainage and waterproofing systems. At floors, the 

opportunity exists if a new concrete topping or structural overlay is proposed. In this case, the 

reinforcement for the collector can be embedded in the topping. Also, if a new raised floor 

system is being installed, it may be possible to locate a new collector in the space beneath the 

new floor. 

Weight of new collector: The gravity load capacity of the existing slab, waffle ribs, joists or 

adjacent beams must be adequate to support the additional weight of the new collector, 

especially for a new concrete collector that may represent a considerable load. In some cases, 

the new collector may need to be designed to support itself as it spans between existing girders 

or columns. In others, it may be required to adjust the location of the collector, and the new shear 

walls, if the existing floor or roof slab cannot support the new loads. 

Detailing Considerations 

Connection of a reinforced concrete collector to existing concrete diaphragms or collectors: A 

typical detail of the installation of a new reinforced concrete collector to the underside of an 

existing concrete slab diaphragm is shown in Figure 12.4.3-1. The primary considerations are to 

provide a good bond between the new and existing concrete and to provide adequate access ports 

for concrete placement and consolidation. The contact surface must be thoroughly cleaned and 

roughened for good shear transfer performance. It is best to place concrete from above through 

pour ports made in the diaphragm. The ports will need to be at least 4 inches in diameter. Care 

must be taken to locate existing diaphragm reinforcement before cutting the ports to avoid 

cutting any bars in what is likely to be a lightly reinforced slab. 
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Figure 12.4.3-1: Concrete Collector at Concrete Slab 

The required length of collector will be determined primarily by the existing diaphragm shear 

capacity. Lightly reinforced diaphragms can deliver only a limited load per foot, requiring long 

collectors. Also, for thin diaphragm slabs, the shear capacity of each drilled dowel will be 

limited, requiring more dowels. If the collector crosses any existing beam or girder, a splice must 

be made through the existing member. Horizontal holes can be drilled through the member and 

dowels installed to lap with the main collector reinforcing bars on each side. Care must be taken 

to avoid cutting any reinforcement, either main longitudinal bars or stirrups, in the existing 

beam. 

If the existing floor or roof diaphragm is a waffle or pan joist system, the continuous collector 

will almost always be placed below the ribs, as shown in Figure 12.4.3-2, to avoid excessive 

drilling and rebar splicing. In this condition, the voids between the ribs, above the dropped 

collector, will be filled with reinforced concrete. Advantages of this condition are that the drilled 

dowels can be installed into the sides of the ribs instead of the relatively thin cover slab, and 

making pour ports through the slab is likely to be less problematic. Also, although the new 

collector may weigh more in this condition, the waffle or joist ribs are much more likely to have 

adequate strength to support the added weight. 
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Figure 12.4.3-2: Concrete Collector at Waffle Slab 

In many cases, the new collector will occur along an existing beam or girder line. Often, this will 

occur if the task is to strengthen a diaphragm edge chord or if the new walls occur at the 

building’s exterior. Figure 12.4.3-3 shows two generic conditions that can be used in this case. In 

this condition, the dowels will always be placed into the beam, and the combined beam-collector 

member will easily be designed support the added weight. However, special care should be taken 

to avoid cutting any beam stirrups or slab diaphragm reinforcement, especially at an exterior 

edge condition, with the pour holes. 

In any of these collector configurations, a significant portion of the main reinforcement can be 

provided by a steel plate instead of by bar reinforcement (refer to Figure 12.4.3-2). This option 

may be best for conditions of very high loads, where installation of a high strength steel plate 

may be preferred over placement of many large bars. 

Connection of a steel collector to existing concrete diaphragms: Steel plate also may be used as 

the collector in lieu of a reinforced concrete member. A steel collector will have to be installed in 
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Figure 12.4.3-3: Concrete Collector at Existing Beam 

manageable sections, generally about 10 to 20 feet in length, and will be connected to the 

concrete diaphragm with drilled threaded rod anchors set in adhesive or epoxy. In almost all 

cases, the steel plates will be installed at the top of the diaphragm as shown in Figure 12.4.3-4. 

Although possible, it is extremely difficult to install heavy plate sections, connect the bolts and 

make the necessary welded splices from below. 

As discussed in the Design Considerations section above, the primary concern with a steel plate 

collector is its lack of strain compatibility with the concrete diaphragm, unless the collector is 
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Figure 12.4.3-4: Steel Plate Collector 

very short. The strain deformation of a steel collector will vary from zero at its free end to a 

maximum at the connection to the wall or brace while the concrete diaphragm will not 

experience similar deformations. In effect, the steel collector will stretch like a very stiff rubber 

band relative to the concrete diaphragm. This relative deformation is difficult to accommodate, 

especially in relatively long collectors. To do that, several conditions must be considered. First, 

the various plate sections of the collector must be stepped in size so the strain is distributed 

relatively equally along the length of the collector. Second the plates must be sized to limit the 

maximum elongation to a reasonable amount of about one or two inches. Third, the threaded rod 

anchors must be installed in slotted holes to allow the design elongation to occur without bearing 

on and overloading the anchors. Fourth, to allow the slip to occur between the collector and 

diaphragm, load transfer must be accomplished by friction using specially calibrated spring 

washers to generate the appropriate clamping force in the anchors. 
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Cost/Disruption Considerations 

Collectors have significant cost/disruption impact in a retrofit project primarily due to their 

length. They impact many building systems over a relatively large area compared to the impact 

associated with the walls themselves. This is especially true if general renovation of architectural 

and M/E/P systems is not included in the project. Thus, any available means of reducing 

collector length will probably be cost effective. A collector installed at the exterior edge of a 

diaphragm will generally be less costly than one installed in the interior and one installed above 

the diaphragm will be easier to install and, generally, less costly than one installed from below. 

However, installation of any collector can be very disruptive to any building occupants, due to 

the noise and vibration caused by drilling and coring through concrete, as well as the likely need 

to relocate various utilities and service distribution systems. 

A comparison of the cost between reinforced concrete and steel plate collectors is very difficult 

to make. In general, the cost of either type of collector installed from below the diaphragm will 

likely be similar, because so much of the cost will be related to the impact on other systems. The 

cost of a steel plate collector may be less than one of reinforced concrete, but only for collectors 

installed from above the diaphragm, and particularly on a roof. 

Construction Considerations 

Existing concrete surfaces to be in contact with new concrete or steel plate collector should be 

cleaned of all finishes, paint, or other substances that could impair bond and shear transfer 

capability. Surfaces to receive new concrete should be roughened to provide 1⁄4” amplitude 

aggregate interlock to prevent slip. However, since slip is expected to occur as a steel collector 

elongates, only light sandblasting may be required to assure development of the appropriate 

friction. 

Installation of grouted anchors and/or dowels for steel plate collectors will require relative 

precision. They must be installed at the middle of the long slotted holes, with only a small 

tolerance, to allow the plate to elongate without bearing on the anchor. If existing rebar is 

encountered at a location, that location should be abandoned and the anchor installed in an 

available adjacent or nearby slot. Since it is reasonable to expect that this will occur with some 

frequency, a substantial amount of extra slotted holes must be available. For instance, if anchors 

are required at 12” on center, slotted holes should be provided at 6” on center. 

The complete penetration welded splices of the relatively thick steel plate collectors are likely to 

be problematic. Making one-sided complete penetration welds in relatively thick plates will 

cause the plates to curl. To control this, the plate sections may need to be anchored down, with 

anchors placed in addition to the required shear anchors, and welded in place. Also, removal of 

backup bars will be difficult or impossible. Notches may be made into the concrete slab, and any 

remaining gap between the bottom of the steel plate collectors and the concrete diaphragm slab 

must be filled with grout to assure adequate friction at the concrete-steel interface. 

Overhead construction of reinforced concrete collectors will require careful consideration and 

planning of how the reinforcing steel is placed and secured, prior to closing up the forms from 

below. Making the pour access ports and any sleeve holes for continuous rebar will require 
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careful scanning of the slabs, waffle ribs, and beams to locate existing reinforcement to avoid 

cutting any existing rebar. 

While the inspection, sampling, and testing required for reinforced concrete collectors is not 

particularly different from what is required for other seismic force-resisting reinforced concrete 

work, some special considerations do occur for steel plate collectors. The welded splices will 

require careful, non-destructive testing and thorough inspection. The shear anchors must be 

located at the middle of the slotted holes with some precision, and they must be extensively 

proof tested in tension. The holes with anchors must be free of any grout that could reduce the 

range of slip. Any gap between the bottom of the plate and the concrete slab must be grouted. 

The installation of the spring washers must be carefully inspected and tested to assure 

development of the design clamping force. 

Proprietary Concerns 

The basic materials are generic. 

12.4.4 Enhance Column with Fiber-Reinforced Polymer Composite Overlay 

Deficiencies Addressed by the Rehabilitation Technique 

Inadequate shear capacity 

Inadequate concrete compression strain and stress capacity due to lack of concrete 

confinement 

Inadequate lap splice 

Description of the Rehabilitation Technique

The use of a fiber-reinforced polymer (FRP) overlay with columns has proven to be an efficient 

rehabilitation technique in both the building and bridge construction industries. Columns are 

overlaid with unidirectional fibers in a horizontal orientation, thus providing shear strengthening 

and confinement similar to that provided by hoops and spirals used with circular columns, and 

stirrups and ties used with rectangular columns. The confinement enhances the concrete 

compression characteristics, provides a clamping action to improve lap splice connections, and 

provides lateral support for column longitudinal bars. 

The preferred strength hierarchy for a building type structure is strong-column, weak-beam. 

Where the strength hierarchy results in weak-column, strong-beam (and is not considered 

acceptable by the designer due to, perhaps, concern for a soft story mechanism), the use of FRP 

overlay as flexural strengthening should not be used, unless there are extenuating circumstances 

and a very detailed analysis and design are performed. The uncertainty of strain compatibility 

between the FRP and column longitudinal bars and between the FRP and substrate, the lack of 

vertical strain capacity as a result of using FRP as longitudinal reinforcement, and the anchorage 

of the FRP at column ends and at points of contra-flexure deem this approach as undesirable. 

Other techniques presented in this document should be used in this situation. 

See Section 13.4.1, “Enhance Shear Wall with Fiber-Reinforced Polymer Composite Overlay, 

Fiber-Reinforced Polymer Composite Overview,” for background information. 
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Design Considerations 

Research basis: Seible and Innamorato (1995) is one of the original ground-breaking research 

efforts of this topic and serves as an excellent source for understanding and design equations. 

The primary deficiencies of a column are typically the lack of shear strength capacity and postyield shear deformation capacity, as observed during many earthquake events and in laboratory 

testing. For shear assessment, two column locations need to be evaluated for shear: the end 

region within the plastic hinge zone (where the concrete shear strength degrades), and the region 

away from the flexural hinges, where there is no concrete shear strength degradation. 

The FRP overlay provides confinement to enhance the concrete strain and stress capacity. 

Confinement is more effective for circular sections than rectangular sections. For circular 

sections, the passive radial pressure exerted by the FRP overlay on the gross concrete section, 

which is a result of the concrete lateral dilation, provides confinement. Dilation (similar to the 

concrete splitting action when performing a pure axial compression test on a concrete cylinder) 

occurs as result of the compression force, which is influenced by the level of axial load and 

flexural forces. For a rectangular section, dilation is only arrested at the corners of the section, 

thus relying on the concrete to arch between the corners, resulting in a reduced concrete core 

size. Due to the lack of effective confinement by the FRP, it is recommended to limit the 

rectangular section to a 1.5 depth-to-width aspect ratio and a width or depth dimension less than 

36 inches, unless a special study is performed. 

The confinement afforded by this technique does marginally increase the flexural strength and 

stiffness of the column, but not to the degree of concrete jacketing. The marginal increase is due 

to the higher concrete stress capacity of the cover and core concrete, hence reduced neutral axis 

depth, and is located within the what is called the primary hinge zone. This increase is over 

about half the depth of the column at each column end (where double-curvature occurs). 

Consequently, there is a greater moment demand just beyond this region, within the secondary

hinge zone. Confinement enhancement, therefore, extends from the end of the column through 

the primary and secondary hinge zones, as shown in Figure 12.4.4-1A. Note that the 

categorization of primary and secondary hinge zones comes from Seible and Innamorato(1995). 

The confinement pressure also serves to clamp the lap splice connection of the column 

longitudinal bars. The thickness (effective clamping pressure) needed for lap splices is derived 

differently from the confinement requirements, however, as test results indicate that, at a dilation 

strain of about 0.001, lap splice slippage is initiated. These results, combined with the 

empirically derived radial pressure requirement to prevent slippage, determine the FRP overlay 

thickness. 

The FRP overlay thickness is determined for each of the three deficiencies; zones for these are 

shown in Figure 12.4.4-1A. As noted by Priestley, Seible, and Calvi (1996), the maximum of the 

three at any section should be provided; it should not be the sum, as reported in some other 

documents. This is because the lap-splice clamping action and compression concrete 

confinement occurs on opposite sides of the column, so the maximum requirement of the two 

will serve both well. Shear resistance of the FRP occurs along the column face parallel to shear 

load direction. The FRP anchors the concrete compression strut and is designed to maintain 
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Figure 12.4.4-1A: Seismic Retrofit of Columns Using FRP Composites 
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Figure 12.4.4-1B: Seismic Retrofit of Columns Using FRP Composites 

aggregate interlock over the crack length. This contribution is a by-product of the hoop tension 

required for the confinement, so FRP composite thickness for shear need not be added to that 

required for confinement. 

With successful mitigation of the three deficiencies, column flexural hinges can be developed 

and deformation capacity will be increased. 
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Detailing Considerations 

The lap splices should be staggered, like that done with steel reinforcement, to mitigate a weak 

plane. As shown in Figure 12.4.4-1A, a nominal gap of about 1⁄2” between the FRP composite 

and the boundary elements (slab, beam or footing) is provided to prevent the overlay from 

bearing and, consequently, increasing the flexural capacity or stiffness of the column. 

Cost/Disruption 

To appropriately evaluate the cost of a retrofit scheme using and FRP overlay in comparison to 

traditional retrofit concepts (such as concrete or steel jacketing), one needs to consider the cost of 

the raw material, the level of specialization required by the contractor to install the system, the 

cost of labor and equipment, the cost of quality control and quality assurance, the temporary 

impact of disruption during construction, and the permanent impact to the building functions. 

Although FRP overlays are relatively expensive compared to steel and concrete, they can offer 

advantages when only limited access is available or minimal disruption of existing conditions is 

desired. 

Construction Considerations 

Access all around the column is often limited due to partition walls, ceilings and other 

architectural components, as well as structural elements of a building. Conditions at the base of 

the column in the lowest story must be carefully considered. The extent of FRP application 

normally will extend to the top of footing which may require local slab removal. The floor slab 

in the area may also interact with the column and affect strengthening requirements. This effect 

should be considered or a gap placed between the slab and column to prevent interaction. 

Proprietary Concerns 

See Section 13.4.1 for brief discussion of proprietary concerns. 

12.4.5 Enhance Concrete Column with Concrete or Steel Overlay 

Deficiencies Addressed by the Rehabilitation Technique 

Inadequate shear capacity 

Inadequate axial compression capacity 

Inadequate flexural plastic hinge confinement 

Inadequate lap splice 

Description of the Rehabilitation Technique

Adding a fiber reinforced polymer (FRP) composite overlay to a concrete column is a recent 

approach to addressing seismic deficiencies, and is discussed in Section 12.4.4. Adding a 

concrete or steel jacket is a more traditional method of enhancing a deficient concrete column, 

Figure 12.4.5-1 provides examples of concrete and steel jacketing for a rectangular column. 

Because FRP overlays have become more common, this section will focus on FRP overlays. 
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Figure 12.4.5-1: Concrete and Steel Overlays for Concrete Columns 
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Design Considerations 

Research basis: Some research on concrete overlays is contained in FIB (2003); other related 

work is discussed in Section 12.4.5. Design of the overlay or jacket uses typical ACI 318 

concrete design principles. Sufficient drilled dowels between the overlay and existing concrete 

should be provided to achieve composite action. Research on steel overlays includes Engelhardt, 

et al., (1994). 

Detailing Considerations 

Concrete overlays: Concrete jacketing will take up a larger cross section than either FRP 

overlays or steel overlays. The surface of the existing concrete must be roughened appropriately. 

Reinforcing steel will need to be in at least two pieces to get it around the existing column. 

135 degree hooks are required for confining ties and may dictate the size of the overlay in order 

to provide enough room for the hook extension. 

Steel jackets: Steel jackets require at least two pieces to get around the existing column and 

involve field welding. Like FRP overlays, when the aspect ratio of a rectangular column gets too 

large, the jacket becomes less effective. The California Department of Transportation (Caltrans) 

uses elliptical jackets in these situations. The corners of the existing column will need to be 

trimmed so the steel can pass by. There is a gap between the steel and the concrete of at least 1⁄4” 

that is filled with grout. A gap is also provided at the ends of the column to permit rotation 

without engagement in bearing of the steel jacket. 

Cost/Disruption 

To appropriately evaluate the cost of a retrofit scheme using and FRP overlay in comparison to 

traditional retrofit concepts (such as concrete or steel jacketing), one needs to consider the cost of 

the raw material, the level of specialization required by the contractor to install the system, the 

cost of labor and equipment, the cost of quality control and quality assurance, the temporary 

impact of disruption during construction, and the permanent impact to the building functions. 

Although FRP overlays are relatively expensive compared to steel and concrete, they can offer 

advantages when only limited access is available or minimal disruption of existing conditions is 

desired. 

Construction Considerations 

Concrete overlays: Because of the difficulty of placing an overlay on all sides of an existing 

column, concrete overlays are typically done with cast-in-place concrete, rather than shotcrete. 

The need for formwork is a significant disadvantage for concrete overlays, compared to FRP and 

steel overlays. Placing the concrete and vibrating are also challenging due to access limitations 

at the top of the column where it runs into beams or slabs. Pour ports or holes in the diaphragm 

are needed. 

Steel overlays: Steel overlays are typically 3/16” or 1⁄4” thick and become quite heavy. Access 

and lifting issues in existing buildings can force the overlay to be broken down into pieces, 

increasing the amount of field welding necessary to join the pieces back together. 
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Proprietary Concerns 

Unlike FRP overlays, no proprietary issues have been identified with using concrete of steel 

jackets. 

12.4.6 Enhance Concrete Moment Frame 

Deficiencies Addressed by the Rehabilitation Technique 

Inadequate global shear capacity 

Inadequate lateral displacement (global stiffness) capacity 

Inadequate ductile detailing for shear strength, confinement, or strong column/weak beam 

Description of the Rehabilitation Technique 

An alternative method to either add strength and stiffness to an existing concrete moment frame 

or correct non-ductile detailing deficiencies of the frame members is by direct enhancement: 

increasing the size of the columns and beams of the frame with new reinforced concrete. This 

method entails adding a jacket of reinforced concrete around the existing columns and beams, an 

approach similar to jacketing by steel or fiber wrap. The new concrete may be either cast-inplace or shotcrete. This approach is relatively rarely employed in the U.S. because it is labor 

intensive, but when existing openings must be maintained and walls or braced frames cannot be 

installed, it may be rehabilitation technique worth considering. 

Design Considerations 

Research basis: Alcocer and Jirsa (1991) tested reinforced concrete beam-column connections 

subjected to unidirectional and bidirectional cyclic loading up to displacements equivalent to 4% 

interstory drift. Jacketing of columns resulted in strong column – weak beam behavior, with 

increases in peak strength of over four times the existing assembly. Jacketing of beams as well 

as columns improved joint confinement, decreased stresses on existing beam reinforcement, and 

provided some additional strength increases. The use of jacketing to repair damaged frames was 

also tested. When comparing the performance of a specimen in which a damaged column and 

joint were repaired by jacketing to that of a similar specimen in which the original column and 

joint were undamaged, the damaged and repaired assembly exhibited 65% of the strength at 2% 

drift and 50% of the stiffness at 0.5% drift relative to the undamaged retrofitted specimen. 

Reinhorn, Bracci, and Mander (1993) performed shake table tests on a one-third scale model of a 

three-story, one-bay by three-bays, concrete moment frame. The tests compared the 

performance of an unretrofitted building, subjected to peak ground acceleration 0.30 g, to that of 

a similar structure retrofitted by jacketing interior columns, post-tensioning added longitudinal 

column reinforcement, and providing a reinforced concrete “fillet” infill around beam-column 

joints. These retrofit measures were intended to ensure strong column - weak beam behavior, 

enhance joint shear capacity, and improve anchorage for discontinuous beam reinforcement. 

Tests demonstrated that the retrofitted structure exhibited significantly reduced column damage, 

especially in the first story, and improved ductility in yielding beams; however, lateral 

displacements remained quite large, reaching a maximum inter-story drift of 2.1% at the 

first floor. 
Techniques for the Seismic Rehabilitation of

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12-40 

Impact on architectural and M/E/P systems and components: Enlarging columns, beams, or 

joints may impact existing ceilings, partitions, ductwork, plumbing, lighting, or usable floor 

space. The designer should keep in mind that the space required for access, formwork, and 

finishes will be greater than the final dimensions of the jacketed member. 

System design: New transverse and longitudinal reinforcement should be designed such that 

members are flexurally-governed and beams yield before columns do. The jacketing of the 

frame will result in increases in system stiffness that may result in increased attracted load. Both 

new and existing concrete should be considered in developing composite properties for modeling 

and design. Joint shear can be the weak link in existing moment frames. A significant focus of 

the Alcocer and Jirsa (1991) research was to provide special angles at the joint region to improve 

joint confinement. Stiffness at 0.5% drift was well predicted when cracked properties of 0.5EI 

were used for the beams and 0.5EI were used for the columns. Stiffness decreased about 40% to 

50% as drifts increased from 0.5% to 2.0%. 

Detailing Considerations 

Surface preparation: In order for a new reinforced concrete jacket to act compositely with an 

existing member, sufficient bond must exist between the new and existing concrete. The Alcocer 

and Jirsa (1991) tests made use of findings from Bass, Carrasquillo, and Jirsa (1985). In the 

Alcocer and Jirsa (1991) tests, they used a handheld electric chipping hammer to reveal some 

aggregate. The amplitude is not documented in the report, but the implication is that it was less 

than the traditional 1⁄4” amplitude value. Dust was cleaned with a thick brush and vacuum 

cleaner. A bonding agent was not used, and the existing concrete surfaces were not saturated in 

all cases. Bonding agents in current practice are not common, but some engineers do 

recommend prewetting the existing concrete. 

Column jacketing: Beams are typically the same plan dimension or narrower that the supporting 

column. New column bars can then pass by the existing beam bars through holes cut in the floor 

slab. Alcocer and Jirsa (1991) explored distributed longitudinal bars vs. bundled bars in the 

corners of the new jacket, but did not find significant differences. It is not possible to put a onepiece closed hoop around an existing column. One approach is to place a U-bar on three sides 

with a single leg closure piece, such as shown in Figure 12.4.4-2A. Alcocer and Jirsa (1991) 

used two overlapping L-shaped bars, each with 135-degree hooks at the ends. A minimum 

column jacket thickness of 4” was recommended. 

Beam jacketing: When beams are jacketed, the bottom and sides can be increased in dimension, 

but typically the top of beam must remain at the existing top of slab level. In the Alcocer and 

Jirsa (1991) tests, slots were cut in the top of the slab just above the beam with holes at the ends 

of the slot through slab. Inverted U-bars were placed in the slot and through the holes. The 

bottom of the inverted U-bars overlapped around beam longitudinal steel with U bars 

surrounding the beam. A minimum beam jacket thickness of 3” is recommended. Care must be 

taken to investigate existing beam reinforcing, so that appropriate locations for new beam 

longitudinal bars can be located to miss existing reinforcing in orthogonal beams and in columns. 

Joint enhancement: In the Alcocer and Jirsa tests (1991), vertical steel angles were placed in 

holes cut through the slab at the four corners of the column and then welded to horizontal steel 
Techniques for the Seismic Rehabilitation of

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12-41 

bars above the slab and below the beam, to form a cubical cage. In the Reinhorn, Bracci, and 

Mander (1993) tests, joint strength was enhanced by infilling between orthogonal beams with 

concrete to create a diamond-shaped “fillet” in plan. Reinforcing at the outer ends of the fillet 

was placed through holes cut into the beam just under the slab and at the base of the beam. 

Cost/Disruption 

The cost and disruption associated with jacketing a concrete frame is significant because it 

potentially involves complicated formwork, reinforcement, and concrete pouring over much of 

the building, as opposed to more common shear wall approaches where the walls are only placed 

in localized areas. Beam jacketing is much more invasive and time consuming and is typically 

less important than the column jacketing. Alcocer and Jirsa (1991) reported that the construction 

time required for jacketing beams and columns was nine times that required for jacketing 

columns alone. 

Construction Considerations 

Noise and disruption: Removing existing column, beam, ceiling and floor finishes is disruptive. 

Chipping the cover concrete off of existing frame members is noisy and disruptive to occupants, 

as is cutting holes in the floor slab. As such this particular technique can be less desirable than 

many others. 

Mix design: Due to the narrow thickness of the jacket and difficult working conditions, the mix 

design should emphasize ease of placement, by using small size aggregate and water-reducing 

admixtures. In the Alcocer and Jirsa (1991) tests, they used 3/8” maximum size aggregate and 

superplasticizer. 

Concrete placement: Concrete should be placed from above through holes in the slab for both 

columns and slabs. Holes have to be big enough for both the concrete hose and the vibrator. 

Consideration can be given to leaving gaps at the top of the forms on the sides of the beams 

where they meet the slabs for air relief vents. 

Proprietary Concerns 

The basic materials are generic. 

12.5 References 

ACI 440.2R-02, 2002, “Guide for the Design and Construction of Externally Bonded FRP 

Systems for Strengthening Concrete Structures,” American Concrete Institute, Chicago, IL. 

Alcocer, S.M. and J.O. Jirsa, 1991, Reinforced Concrete Frame Connections Rehabilitated by 

Jacketing, Phil M. Ferguson Structural Engineering Laboratory Report 91-1, University of Texas 

at Austin, Austin, TX, July. 

Bass, R.A., R.L. Carrasquillo, and J.O. Jirsa, 1985, Interface Shear Capacity of Concrete 

Surfaces Used in Strengthening Structures, Phil M. Ferguson Structural Engineering Laboratory 

Report 85-4, University of Texas at Austin, Austin, TX, December. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 12 – Type C1: Concrete Moment Frames

12-42 

Engelhardt, M.D., et al., 1994, "Strengthening and Repair of Nonductile Reinforced Concrete 

Frames Using External Steel Jackets and Plates," Repair and Rehabilitation Research for Seismic 

Resistance of Structures, ed. James Jirsa, Report R/R 1994-1, June. 

FIB (Federation Internationale de Beton), 2003, Seismic Assessment and Retrofit of Reinforced 

Concrete Buildings, State-of-Art Report, Bulletin 24. 

Jiménez-Pacheco, J. and M.E. Kreger, 1993, Behavior of Steel-to-Concrete Connections for Use 

in Repair and Rehabilitation of Reinforced Concrete Structures, Phil M. Ferguson Structural 

Engineering Laboratory Report 93-2, University of Texas at Austin, Austin, TX, March. 

Jones, E.A. and J.O. Jirsa, 1986, Seismic Strengthening of a Reinforced Concrete Frame Using 

Structural Steel Bracing, PMFSEL Report No. 86-5, Department of Civil Engineering/Bureau of 

Engineering Research, University of Texas, Austin, TX. 

Priestley, N., Seible, F. and G.M. Calvi, 1996, “Seismic Design and Retrofit of Bridges,” John 

Wiley & Sons, New York, NY. 

Reinhorn, A.M., J.M. Bracci, and J.B. Mander, 1993, “Seismic Retrofit of Gravity Load 

Designed Reinforced Concrete Buildings,” Proceedings of the 1993 National Earthquake 

Conference: Earthquake Hazard Reduction in the Central and Eastern United States: A Time for 

Examination and Action, Memphis, TN, May, Vol. 2, pp. 245-254. 

Seible, F. and D. Innamorato, 1995, “Earthquake Retrofit of Bridge Columns with Continuous 

Carbon Fiber Jackets, Volume II, Design Guidelines,” Report to Caltrans, Division of 

Structures, University of California, San Diego, La Jolla, CA. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 13 – Type C2b: Concrete Shear Walls

13-1 

Chapter 13 - Building Type C2b: Concrete Shear Walls 

(Bearing Wall Systems) 

13.1 Description of the Model Building Type 

Reinforced concrete walls in a building will act as shear walls whether designed for that purpose 

or not. Therefore, cast-in-place concrete buildings that contain any significant amount of 

concrete wall will fall into this category. However, there are two distinctly different types of 

concrete wall buildings, those that contain an essentially complete beam/slab and column gravity 

system, and those that use bearing walls to support gravity load and have only incidental beam 

and column framing. In this document, these building types have been separated and are 

designated C2f for the gravity frame system and C2b for the bearing wall. This section covers 

the bearing wall type. In this type of building, all walls usually act as both bearing and shear 

walls. The building type is similar and often used in the same occupancies as Building Type 

RM2, namely in mid- and low-rise hotels and motels. This system is also used in residential 

apartment/condominium type buildings. 

Figure 13.1-1: Building Type C2b: Concrete Shear Walls (Bearing Wall Systems) 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 13 – Type C2b: Concrete Shear Walls

13-2 

Variations Within the Building Type 

In order for this framing system to be efficient, a regular and repeating pattern of concrete walls 

are required to provide support points for the floor framing. In addition, since it is difficult and 

expensive to make significant changes in the plan during the life of the building, planning 

flexibility is not normally an important characteristic when this structural system is employed. 

The occupancy type that most often fit these characteristics are residential buildings, including 

dormitories, apartments, motels, and hotels. These buildings will often be configured with 

reinforced concrete bearing walls between rooms—also acting as shear walls in the transverse 

direction, and reinforced concrete walls on the interior corridor acting primarily as shear walls in 

the longitudinal direction. Sometimes the longitudinal lateral system includes the exterior wall 

system, although this wall is normally made as open as possible. In any case, the wide variation 

in structural layouts and occupancies that is included in Building Type C2f is not seen in 

Type C2b. 

It is seldom possible to plan a building layout that provides complete gravity support with walls. 

Often, local areas are supported with isolated columns, and sometimes beams and girders are 

also necessary, but story heights in these buildings are usually small and added depth in the floor 

framing system is difficult to obtain. The extent of such beam and column framing often causes 

confusion between Building Types C2b and C2f, but buildings should have an essentially 

complete gravity frame system to be placed in C2f. If significant plan area is supported solely 

by walls, the structures are normally classified as C2b. 

There are important variations in floor framing systems employed in this building type, and their 

adequacy to act as a diaphragm is an important characteristic of this building type as discussed 

below. 

Floor and Roof Diaphragms 

The parallel layouts of supporting walls and the need to minimize story heights normally leads to 

the use of one-way uniform-depth concrete floor systems. Cast-in-place and precast systems, 

both conventionally reinforced and prestressed, have been employed. The precast systems are 

often built up of narrow planks, which may not provide an adequate diaphragm unless a cast-inplace topping is provided. In addition, the precast systems may be placed with only a very 

narrow bearing area on the supporting walls, which may be inadequate to provide vertical 

support during seismic movements. The adequacy of the shear connection between slab and 

walls is also often an issue for both cast-in-place and precast systems. 

Foundations 

The bearing walls obviously require some kind of starter beam at grade for construction purposes 

and this often leads to a simple continuous grade beam system. In poor soils, piles or drilled 

piers may be added below the grade beam. A continuous mat foundation may also be employed 

due to the short spans and total length of bearing points in this building type. 
Techniques for the Seismic Rehabilitation of

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13-3 

13.2 Seismic Response Characteristics 

Due to the extent of wall, bearing wall buildings will be quite stiff. Elastic and early post-elastic 

response will therefore be characterized with lower-than average drifts and higher-than-average 

floor accelerations. Damage in this range of response should be minimal. 

Overall post-elastic response may often include rocking at the foundation level. If rocking does 

not occur, the height-to-length ratio of shear walls in these buildings may force shear yielding 

near the base, which may lead to strength and stiffness degradation. 

Global stability may also be compromised by poor connections between floor slab construction 

and bearing walls. 

Shear Wall Behavior 

When subjected to ever increasing lateral load, individual shear walls or piers will first often 

force yielding in spandrels, slabs, or other horizontal components restricting their drift, and 

eventually either rock on their foundations, suffer shear cracking and yielding, or form a flexural 

hinge near the base. Shear and flexural behavior is quite different, and estimates of the 

controlling action are affected by the distribution of lateral loads over the height of the structure. 

Yielding of spandrels, slabs, or other coupling beams can cause a significant loss of stiffness in 

the structure. Flexural yielding will tend to maintain the strength of the system, but shear 

yielding, unless well detailed, will degrade the strength of the coupling component and the 

individual shear wall or pier will begin to act as a cantilever from its base. In this building type, 

the coupling elements are often slabs, and their lack of bending stiffness may reduce or eliminate 

significant coupling action. 

Rocking is often beneficial, limiting the response of the superstructure. However, the amplified 

drift in the superstructure from rocking must be considered. In addition, if varying wall lengths 

or different foundation conditions lead to isolated or sequencing rocking, the transfer of load 

from rocking walls must be investigated. In buildings with basements, the couple created from 

horizontal restraint at the ground floor diaphragm and the basement floor/foundation (sometimes 

called the “backstay” effect) may be stiffer and stronger than the rocking restraint at the 

foundation and should be considered in those configurations. 

Shear cracking and yielding of the wall itself is generally considered undesirable, because the 

strength and stiffness will quickly degrade, increasing drifts in general, as well as potentially 

creating a soft story or torsional response. However, in accordance with FEMA 356, shear 

yielding walls or systems can be shown to be adequate for small target displacements. Type C2b

buildings will often fall into this category. 

Flexural hinging is considered ductile in FEMA 356 and will degrade the strength of the wall 

only for larger drifts. Similar to rocking, the global effect of the loss of stiffness of a hinging 

wall must be investigated. 
Techniques for the Seismic Rehabilitation of

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13-4 

13.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

See Table 13.3-1 for deficiencies and potential rehabilitation techniques particular to this system. 

Selected deficiencies are further discussed below by category. 

Global Strength 

Due to the extensive use of walls, buildings of this type seldom have deficiencies in this 

category, unless significant degradation of strength occurs due to shear failures. 

Global Stiffness 

Similar to strength, global stiffness is seldom a problem in this building type. However, the 

effect of coupling slabs on initial stiffness and the potential change in stiffness due to yielding of 

these coupling slabs or wall-beams over doors should be investigated. 

Configuration 

The most common configuration deficiencies in this building type are weak or soft stories 

created by walls that change configuration or are eliminated at the lower floors. It is difficult to 

provide the needed ductility at the weak story, and often strength must be added. Completely 

discontinuous walls also create a load transfer deficiency for both overturning and shear. In such 

cases, collectors are often needed in the floor diaphragm, and supporting columns need axial 

strengthening. 

Load Path 

A common deficiency in this building is weakness in the load path from floor to walls, either 

collector weaknesses or shear transfer weakness immediately at the floor wall interface. Local 

transfer can be strengthened by adding concrete or steel corbel elements, dowels, or 

combinations of these components. As indicated above discontinuous walls also often create 

load path deficiencies. 

Component Detailing 

The most common detailing problem in this building type is an imbalance of shear and flexural 

strength in the walls, leading to pre-emptive shear failure. This deficiency may be shown to be 

not critical with small displacement demands, walls can be strengthened in shear with overlays 

of concrete, steel, or FRP. 

The layout of walls often forces coupling between walls through the slab system or across 

headers of vertically aligned doors. These coupling components are seldom designed for the 

coupling distortions that they will undergo, particularly in older buildings. Short lengths of slabs 

between adjacent walls receive damage by coupling action that could compromise the gravity 

capacity. It is difficult to add strength or ductility to these slab areas, but vertical support at 

support points can be supplemented by corbels of steel or concrete. Damage to headers over 

doors often does not contribute to deterioration of overall response and can sometimes be 

acceptable. Local areas of wall can also be strengthened by overlays of concrete, steel, or FRP. 
Techniques for the Seismic Rehabilitation of

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13-5 

Table 13.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C2b Buil

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Dem

Global Strength 

Insufficient inplane wall shear strength 

Concrete/masonry shear wall [12.4.2] 

Concrete wall overlay [21.4.5] Fiber composite wall overlay [13.4.1] Steel overlay 

Seismic isol[24.3] 

Reduce flexcapacity [13

 Insufficient 

flexural capacity 

Concrete/masonry shear wall [12.4.2] 

Add chords [12.4.3] 

 Inadequate 

capacity of coupling beams 

Concrete/masonry shear wall [12.4.2] 

Strengthen beams [12.4.2] Improve ductility of beams [12.4.2] 

Global Stiffness 

Excess drift (normally near the top of the building) 

Concrete/masonry shear wall [12.4.2] 

Fiber composite wrap of columns to improve lateral displacement capability [12.4.4] Provide detailing of all other elements to accept drifts Concrete wall overlay [21.4.5] 

Supplementdamping [24

Configuration Discontinuous 

walls 

Add wall or adequate columns beneath [12.4.2] 

Fiber composite wrap of supporting columns [12.4.4] Concrete/steel jacket of supporting columns [12.4.5] 

Improve connection to diaphragm [13.4.3] 

 Soft story or weak 

story 

Add strength or stiffness in story to match balance of floors 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 13 – Type C

13-6 

Table 13.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C2b Buil

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Dem

Configuration (continued) 

Re-entrant corner Add floor area to 

minimize effect of corner 

Provide chords in diaphragm [12.4.3] 

 Torsional layout Add balancing walls 

[12.4.2] 

Load Path Inadequate 

collector 

Add steel or concrete collector [12.4.3] 

 Inadequate slab 

bearing on walls 

Add diagonal dowels [13.4.3] Add steel ledger [13.4.3] 

Component Detailing 

Wall inadequate for out-of-plane bending 

Add strongbacks [21.4.3] 

Concrete wall overlay [21.4.5] 

 Wall shear critical Concrete wall overlay 

[21.4.5] Fiber composite wall overlay [13.4.1] 

Reduce flexucapacity [13.

Diaphragms Precast 

components without topping 

Improve interconnection [22.2.11] Add topping 

 Inadequate inplane shear 

capacity 

R/C topping slab overlay Fiber composite overlay [22.2.5] 

 Inadequate shear 

transfer to walls 

Add diagonal drilled dowels [13.4.3] Add steel angle ledger [13.4.3] 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 13 – Type C

13-7 

Table 13.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C2b Buil

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Dem

Diaphragms (continued) 

Inadequate chord capacity 

New concrete or steel chord member [12.4.3] 

 Excessive stresses 

at openings and irregularities 

Add chords [12.4.3] 

Foundations See Chapter 23 

[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 
Techniques for the Seismic Rehabilitation 

of Existing Buildings: FEMA 547 Chapter 13 – Type C2b: Concrete Shear Walls 

13-8 

Diaphragm Deficiencies 

Precast floor systems used in this building type often provide inadequate diaphragm behavior 

that could lead to bearing failures at the floor wall interface, particularly when no topping slab is 

present. Some topping slabs used primarily for leveling and smoothing the floor are 

inadequately tied to the precast elements or the walls, and are too thin or poorly reinforced to act 

as diaphragms on their own. See Chapter 22. 

Foundation Deficiencies 

This building type often places large demands on the foundation system. If rocking is shown to 

be a controlling displacement fuse for the building, the foundations must be investigated to 

assure that these displacements can safely occur. See Chapter 23. 

13.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

13.4.1 Enhance Shear Wall with Fiber-Reinforced Polymer Composite Overlay 

Fiber-Reinforced Polymer Composite Overview 

In addition to this section, seismic rehabilitation techniques using fiber-reinforced polymers are 

described in several other sections in this document, including Section 12.4.4, “Enhance Column 

with Fiber-reinforced Polymer Composite Overlay,” and Section 22.2.5, “Enhance Slab with 

Fiber Reinforced Polymer Composite Overlay.” This section provides a general overview of FRP 

characteristics. 

Composite makeup and application: The construction industry’s term Fiber-Reinforced Polymer

(FRP) refers to a composite material made up of carbon, fiberglass or aramid (Kevlar) fibers that 

are bound together by either a resin or ester polymer. These are commonly referred to as glass 

fiber-reinforced polymer (GFRP) composite, carbon fiber-reinforced polymer (CFRP) 

composite, or aramid fiber-reinforced polymer (AFRP) composite, respectively. The raw fibers 

(synonymous to filaments) can be woven to form mesh (uni- or bi-directional with the 

orientation of the warp to the weft at 45 or 90 degrees), collected together to form a carbon tow 

winding (an untwisted bundle of continuous filaments) or sheet, or pultruded to form a 

prefabricated plate or other shape. 

Usually, at least two layers of FRP are applied to the exterior concrete (substrate) surface. For 

beam and column applications, one layer should be considered sacrificial, due to the possibility 

of abrasion and the fact that lap splices are used and potentially compromise a layer’s 

effectiveness. 

Mechanical properties: The mechanical property in the direction of the fibers of GFRP, CFRP 

and AFRP is an essentially linear-elastic response followed by sudden rupture. The three 

material systems have different rupture strains and moduli. This variation becomes an important 

consideration when selecting the fiber type. While the rupture strain of each material is 

different, they are all significantly less (by approximately an order of magnitude) than that of 

conventional concrete reinforcement, which leads to compatibility issues. 
Techniques for the Seismic Rehabilitation 

of Existing Buildings: FEMA 547 Chapter 13 – Type C2b: Concrete Shear Walls 

13-9 

Creep (increase in strain), stress rupture (reduced tensile capacity) and stress corrosion 

(corrosion that is dependent on presence of stress to occur) are phenomena that occur when FRP 

is subjected to sustained loads, such as flexural strengthening of slabs for long-term gravity 

loads. Each type of composite responds differently to loading regime, environmental conditions, 

and matrix and fiber make-up: AFRP is prone to creep under sustained load and moderately so to 

stress corrosion, GFRP is prone to stress rupture (as low as 20% of the ultimate), but CFRP is 

robust to sustained loads. Page 115 of FIB (2001) reference provides more information. This 

document addresses seismic loading, which is of short duration; hence, FRP is not affected by 

these characteristics. Accidental sustained stress loading for seismic applications may occur and 

should be given special study. An example is the FRP diaphragm overlays where the fiber 

composite transfer on top of a floor to a wall can act as negative slab reinforcement and resist 

subsequent live loads. Gravity load enhancement is not addressed in this document, but it should 

be noted that it is affected by this characteristic. 

For the same FRP composite, mechanical properties can vary between manufacturers and 

sometimes within the same manufacturer over the course of material production for a project. It 

is recommended that the contract documents clearly specify the performance requirements (such 

as force per unit width for each application) and the minimum ultimate rupture strain. 

Requirements at the FRP-to-substrate interface: There are “contact-critical” and “bond-critical” 

applications between the composite material and the substrate surface. 

Contact-critical applications are mostly limited to beam and column shear and confinement 

enhancement techniques, and it is preferred that the composite be wrapped around all sides of the 

element (i.e., made continuous). These applications do not require shear flow capacity between 

the composite to the substrate, so paint and other smooth finish materials may be left in place. 

Bond-critical applications, such as discontinuous applications and wall applications, require 

shear flow capacity between the FRP overlay and the substrate. (Note that flexural strengthening 

for gravity load enhancement of slabs and beams is a bond-critical application, but it is not 

included in this document.) The surface preparation is important and a concrete surface profile of 

3 (CSP 3) is required (as described in ICRI, 2003), which calls for all loose laitance, the weaker 

outer cement paste layer, and any unsound concrete to be removed. Light sand or water blasting 

readily achieves the desired result. The surface should be dust free at the time of applying the 

first resin layer. Resin putty is used to fill the voids, provide a smooth surface, and create a 

chamfered corner. Paint and other surface finishes must be checked for hazardous materials 

before removal. 

All of the fiber composites have the inherent tendency to rupture prematurely at stress 

concentrations. Such concentrations are formed by burrs, sharp edges, protrusions, etc., in the 

substrate. Any sharp edge, or protrusion of any kind, must be removed during the preparation of 

the substrate to ensure a smooth surface free of dirt, grease, oil and finishes. Existing elements 

damaged by cracks or corrosion should be repaired prior to applying this rehabilitation 

technique. 
Techniques for the Seismic Rehabilitation 

of Existing Buildings: FEMA 547 Chapter 13 – Type C2b: Concrete Shear Walls 

13-10 

The substrate surface should be essentially flat, so that the fibers are straight when positioned. 

An uneven surface prevents the fibers from lying straight, and upon loading, will tend to 

straighten, which will compromise the bond capacity. An out-of-plane angle of 1-2% will 

compromise the bond strength by initiating peeling of the FRP overlay from the substrate. 

Typically, the weak link in bond-critical applications is the tensile rupture of the substrate 

(although the resin should also be checked). This weak link obtains peak strength at about 1/32-

inch displacement, followed by complete loss of bond strength at about 1/3-inch. Further, test 

results and observed behavior show a finite development length of about a few inches and 

beyond that no additional load is developed in the composite material, meaning that the fiber 

material does not necessarily develop its full strength (see Teng et al. (2002) Figures 2.3 and 

2.7). This behavior is unlike rebar reinforcement, where the bar development length is sized, and 

the bond area is provided to attain full rebar strength. The design, therefore, for FRP composite 

requires focus on bond strength and an awareness of crack patterns and the locations of inelastic 

deformations. 

Fiber and mechanical anchors: Significant research on new means of anchoring FRP composite 

to a substrate is occurring at this time. The most common anchor consists of inserting carbon 

fiber or a mixture of carbon and glass fibers into a drilled hole in the substrate. The hole is then 

filled with epoxy, and the protruding strands are splayed to a cone or fan shape that is used to lap 

with the composite overlay. The splay is located between the overlay layers to enhance the 

splice connection. There are many variations of this approach being studied. Anchor spacing, 

edge distance and other issues that influence the performance are also being investigated. The 

design of the anchor and FRP composite overlay system, therefore, should be case specific and 

based on the most recent research. 

Durability: Composite materials, if manufactured correctly and with the appropriate finish 

applied, can provide corrosion resistance, ultraviolet light resistance, fire resistance, and 

tolerance to variations in temperature – making them generally suitable for most environments. 

Specific consideration should be given to galvanic corrosion of CFRP, particularly where there is 

an electrolyte present and potential for mixing of metals. Although CFRP is otherwise corrosion 

resistant, it is a conductor and will change properties when heated. Locations where lightning 

strikes may occur, such as garage rooftops or exterior of buildings, will warrant the use of other 

FRP materials or grounding of metal grid to protect the CFRP. See FIB (2001), Section 9.11. 

For each environment application, the experience of manufacturers and researchers should be 

considered, and the manufacturer’s warranties should be carefully considered. The designer 

must consider the myriad of environmental factors and develop an FRP solution that is 

appropriate for those conditions, and, more importantly ensure that the appropriate fiber and 

polymer material, surface protection, and finish is specified. As an example, where high 

humidity and/or high alkalinity are present, carbon fiber is the preferred choice. Where 

ultraviolet light is present, surface protection of the composite can be achieved with an acrylic or 

polyurethane based paint applied when the polymer is still tacky. For more durability 

information, refer to FIB (2001), Chapter 9. 

Constructibility: In addition to the costs of material and installation of FRP composite, concrete 

surface preparation and final appearance requirements must also be considered. The surface 
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finish requirements, as described above in the Interface Requirements section, need to be 

included and made clear in the contract documents. The woven texture of the mat applications 

does read through to the finish surface. If additional surface finishing is required, such as paint 

or a cementitous appearance, this can add significant cost, and should be coordinated with the 

architect and owner. For seismic rehabilitation applications, the FRP overlay is similar to 

structural steel braces, in that they are not typically required to support gravity loads and thus do 

not require fireproofing. However, some local jurisdictions do require that a fire-protecting 

surface be provided, such as intumescent paint. 

The resin’s shelf-life, pot-life, ambient temperature, ventilation, substrate moisture state and 

other environmental factors need to be carefully considered by the design team. For example, 

due to issues with offgassing during and following installation for some time, not only will 

ventilation be needed, but occupants may need to be temporarily displaced. Quality control 

procedures, which should be required by the contract documents, need to address these issues 

and be verified as acceptable by the engineer of the project and the inspector of record. 

Research basis: Teng, et al. (2002) provides a summary of recent research projects and design 

equations suggested by researchers. Design equations and construction quality assurance 

requirements are also present in model codes, such as ACI 440.2R-02 (2002) and FIB 14 (2001). 

For a summary of the mechanical properties and design procedures, see Priestley, et al. (1996). 

For detailed material mechanical properties and chemical background information, see Kaw 

(1997). 

Deficiency Addressed by the Rehabilitation Technique 

Inadequate shear capacity in a concrete shear wall 

Description of the Rehabilitation Technique 

An FRP overlay is a technique that is used to enhance the in-plane shear capacity of a reinforced 

concrete shear wall. The overlay can be applied to one or both sides of a wall, and where 

possible, should wrap around the ends of the wall to aid in anchoring the overlay. The 

rehabilitation technique is bond-critical, regardless of whether or not the material wraps around 

the end of the walls. Uni-directional (horizontal-oriented) fibers are used to enhance the shear 

capacity, creating a predominately flexural post-yield response. Vertically-oriented fibers in bidirectional layouts will limit the vertical strains to that of the FRP composite, inhibiting the 

ductile behavior. Therefore, the rehabilitation technique for walls is limited to horizontally 

oriented fibers, unless there are extenuating circumstances. The shear enhancement may change 

the wall’s response from a shear-dominated behavior to a flexural, sliding shear, or rocking 

behavior. At coupling beams, vertically-oriented fibers are typically used. See Figure 13.4.1.1A 

for examples of wall and coupling beam layouts. 

Design Considerations 

Research basis: For wall-specific criteria, there has been an increasing amount of research on 

strengthening of unreinforced masonry walls, but less information on reinforced concrete walls. 

Ghobarah (2004) is one source; and Laursen, Seible, and Hegemier (1995), though based on 

CMU specimens, can be extended to some degree to reinforced concrete walls. 
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Figure 13.4.1-1A: Shear Strengthening of Concrete Shear Walls Using FRP Composite 

The shear resistance contribution from the FRP is obtained in a similar manner to that used for 

wall reinforcement; the horizontal bars and FRP resist the horizontal shear (see Section 5.1 in 

Laursen, Seible, and Hegemier, 1995). The effective fiber area per unit width, and its 

contribution to shear resistance, is limited to the bond and anchorage strength capacity; providing 

additional fiber area will not provide additional shear capacity. Testing has been limited to 

mostly single and double layers of FRP per side of wall. Wrapping the material around the ends 

of the wall and/or providing fiber anchors will enhance the effectiveness of the overlay, 

particularly where cracks form, by increasing the anchorage and bond of the FRP to the substrate 
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Figure 13.4.1-1B: Fiber Anchor Details 

In terms of wall shear strength, the FRP overlay can be additive to the wall’s concrete and 

reinforcement contribution. The ductility, however, is dependent on the type of governing 

mechanism. For shear-dominated walls (where the post-yield deformations require slippage at 

crack locations), the FRP has limited ability to accommodate such deformation; hence, there is 

likely very limited ductility. Conversely, a wall dominated by flexural yielding is able to 

accommodate the plastic deformations by the yielding of the wall vertical reinforcement, 

provided that the wall vertical reinforcement is developed. As a result, the typical goal in adding 

an FRP overlay is to make a wall flexurally-critical. For the case where flexural yielding occurs 

first, and is then followed by shear yielding as a result of the reduced concrete shear contribution 

(due to the reduced aggregate interlock effectiveness), ductility is significantly less than that with 

a flexural yielding response. Moreover, there is essentially no confining pressure afforded by the 

use of FRP overlay, except locally, where a fiber anchor is used. Therefore, lap splice 
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performance is not enhanced with this technique and, depending on the size of bar, cover, and 

lap length, limited ductility may result. 

Although testing with bi-directional (at plus and minus 45 degrees) fibers in wall overlays shows 

enhanced shear and flexural strengths and moderately enhanced displacement ductility capacity, 

the vertical force component of the fibers contribution to the flexural strength may have adverse 

affects. This may reduce the distribution of inelastic straining and/or change the strength 

hierarchy to that of shear controlled. Until further research is performed, horizontally-oriented 

FRP strengthening is recommended. 

Detailing Considerations 

Given the high dependence on the bond strength of the FRP overlay to the substrate, in-situ bond 

testing should be included as a requirement in the contract documents. A testing program will 

verify the design assumptions and assist in providing quality assurance. If pilasters are present, 

either within the wall length or at wall ends, installing fiber anchors or removing portions of the 

pilaster should be considered to enhance the anchoring of the FRP overlay. 

Construction Considerations 

As discussed in the FRP Composite Overview section earlier, the engineer should inspect the 

surface of the wall elements to be rehabilitated and note in the contract documents the surface 

condition and wall configuration (e.g., wall corner profile and wall-to-slab configuration). To 

aid in developing a sound bid price, the contract documents need to record the as-built condition, 

including surface anomalies and configuration, and the surface preparation requirements. If the 

surface has been board formed with wood planks, for example, calling this out in the 

construction documents will enable a more accurate bid. Bid documents should also require the 

contractor to view existing conditions before bidding. 

Proprietary Concerns 

Although the basic materials are generic, the fabric that will be supplied is proprietary. Strand 

orientation and density, epoxy overlays, preparation requirements, and installation procedures 

may be different between suppliers. Some suppliers may not have experienced applicators in the 

area of the project. These variations must be considered to achieve an adequate specification, 

particularly if competition between suppliers is desirable. 

13.4.2 Enhance Deficient Coupling Beam or Slab 

Deficiencies Addressed by the Rehabilitation Technique 

Inadequate shear and bending capacity of coupling beams or slabs 

Inadequate ductile detailing for shear strength 

Description of the Rehabilitation Technique 

Coupling beam deficiencies can be encountered in any type of building structure that has shear 

walls for their primary lateral force-resisting system. In a bearing wall structure, however, the 

coupling beam is often not a beam at all but only the relatively thin concrete slab linking the 

adjacent walls across a corridor. In some cases, particularly near the top of taller buildings, the 
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restraint of a slab or beam in-line with the flexural deformed shape of a narrow shear wall can 

create coupling beam type issues at the tip of a single wall. In many cases, the slab consists of 

precast (and prestressed) hollow core planks, with or without a topping, with limited reinforcing 

steel at the link beam-wall joint instead of cast-in-place concrete. Furthermore, because bearing 

wall buildings are often residential buildings with short story heights, there is often no ability to 

create a dropped beam, and it is very difficult or impossible to increase the strength or ductility 

of the linking slab-beam itself. In these cases, the mitigation approach is to install a steel or 

concrete corbel to provide supplemental vertical support. 

For those cases where there actually is a beam or deeper header linking the walls, the most 

common mitigation method to strengthen it or to correct its non-ductile detailing deficiencies is 

by direct enhancement with new added reinforced concrete. The new concrete is generally added 

to one face of the coupling beam, in concert with concrete strengthening of the adjacent shear 

walls as well. In some cases, new reinforced concrete can be added to both faces and extended 

along the walls for development. In addition, or as an alternative, it may be possible to install a 

limited number of diagonal dowels to enhance the shear capacity of the coupling beam to wall 

joint. In a case where there is enough beam or header depth, two groups of tied bars could be 

placed in an X-configuration, embedded within the new beam reinforcement cage. The new 

concrete may be either cast-in-place or shotcrete. 

Design Considerations 

If the restraint of the slabs and beams described above is not needed for overall structural 

stiffness, the element should be evaluated for gravity support in a damaged state. No mitigation 

may be required, particularly in cases where cast-in-place concrete slabs exist. For cases where 

gravity support is inadequate, the approach may be the installation of a supplemental support in 

the form of new corbels placed under locations of damage. If there is no ceiling, care should be 

taken to minimize the visual impact of the corbel and connectors. If existing M/E/P system 

components conflict with installation of these supplemental supports, they should be relocated as 

required to make this minimum mitigation effort. 

For cases where an actual coupling beam or header is being augmented, the contribution of both 

the existing and new portions of the now composite member should be considered. Also, since it 

will be very difficult to provide sufficient strength in the augmented beam, the primary goal 

should be to provide adequate ductility to survive the expected rotations and continue to provide 

gravity load support and shear transfer capability. If the new concrete work is exposed to view, 

consideration should be given to the nature of the forming and surface finishes desired in relation 

to the surrounding existing elements. For cases where a deeper beam or header exists, the new 

ductile member should be designed to provide all required strength and ductility. 

Detailing Considerations 

Connection of new corbel to existing concrete elements: A typical detail showing the installation 

of a new steel corbel is shown in Figure 13.4.2-1. The steel angle provides supplemental support 

for the coupling slab at the wall joint where the most damage is expected to occur. For a precast 

plank slab condition (as shown in the detail), especially without a cast-in-place topping, 

consideration should be given to extending the supplemental steel supports along the full length 

of the link between the coupled walls. Installation of drilled anchors into the bottom of the 
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Figure 13.4.2-1: Typical Corbel at Linking Slab 

planks may be considered. However, it may be difficult to avoid both prestressing tendons and 

hollow core voids. Consider use of screen-tube anchors (see Chapter 21) if voids cannot be 

avoided. 

Connection of new concrete to the existing concrete beam and adjacent walls: Figure 13.4.2-2 

shows an elevation and section view of a typical augmentation of a relatively shallow coupling 

beam or header. The surface of the existing concrete should be thoroughly cleaned and 

roughened to provide a good bond and interaction between the existing and new portions of the 
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Figure 13.4.2-2: Typical Strengthening of Shallow Coupling Beam 
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composite member. The new concrete should be extended along the walls far enough to develop 

the longitudinal bars. If cast-in-place concrete is used, access holes will be required through the 

slab for placement and consolidation. Care must be taken to avoid damage to any prestressing 

tendons in the precast planks. Figure 13.4.2-3 shows typical details for installation of X-shaped 

rebar cages at deeper coupling beams or headers. The new concrete should be thick enough to 

allow placement of the crossing longitudinal reinforcement cages inside of the confining beam 

ties. 

Cost/Disruption Considerations 

Installation of steel corbel supplemental supports is an inexpensive, minimal mitigation 

approach. However, costs could increase substantially if relocations of M/E/P systems are 

required. The disruption associated with a single installation of this work is relatively local, but 

similar work will likely be required at all or most of the walls throughout the building. At all 

cases where new concrete or shotcrete is added to existing coupling beams, the cost and level of 

disruption will increase substantially. Refer to similar discussion in Sections 12.4.2 and 12.4.3. 

Construction Considerations 

For installation of a steel corbel, care should be taken to locate existing reinforcing steel in the 

wall prior to drilling or coring. Also, the pattern of prestressing tendons should be located before 

any drilling into precast planks is begun. For augmentation of coupling beams, refer to similar 

discussions at Sections 12.4.2 and 12.4.3. 

Proprietary Concerns 

The basic materials are generic. 

13.4.3 Enhance Connection Between Slab and Walls 

Deficiencies Addressed by the Rehabilitation Technique 

Inadequate bearing for precast slab planks at wall 

Inadequate shear transfer capacity from precast plank diaphragms to shear wall 

Description of the Rehabilitation Technique 

For most cases involving inadequate bearing support for precast floor planks at the walls, the 

typical mitigation approach will be to install steel (or concrete) ledgers secured to the wall under 

the slab-wall joint. This approach can also be used to increase joint shear capacity by including 

concrete anchors drilled up into the slab. An alternative approach to increasing joint shear 

capacity is to install diagonal dowels, or shear pins, through the joint. Occasionally, if adequate 

story height is available and gravity load issues can be addressed adequately, a new reinforced 

concrete topping slab could be placed over the existing diaphragm. 

Design Considerations 

For cases where the concern is focused only on providing adequate bearing, installation of a steel 

ledger angle anchored to the wall is a very simple and direct technique. Provision of about six 

inches of additional bearing will be enough to assure against loss of vertical support, so anchors 

drilled into the slab or precast concrete planks may be omitted. Use of a concrete ledger is 
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Figure 13.4.2-3: Strengthening of Deep Coupling Beam 
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certainly possible, but would seem appropriate only if concrete collectors were also being 

installed. However, a steel angle or channel section ledger could still be used, linking the end of 

a concrete collector to the wall. 

For cases where diaphragm-wall shear transfer is to be augmented, either vertical slab anchors or 

diagonal dowels, drilled and placed from above, may be added. However, the added shear 

capacity that can be provided by these techniques is limited by the shear capacity of the 

diaphragm slab immediately adjacent to the wall. If the local diaphragm capacity is inadequate, 

then a new collector will be required to engage a greater extent of the diaphragm (refer to 

Section 12.4.3). 

If a new concrete topping slab is to be placed, sufficient reinforcement probably can be included 

to serve as the collector. However, new diagonal dowels or perhaps a new ledger will likely be 

needed to insure transfer of the collected diaphragm shear demand down through the lightly 

reinforced slab-wall construction joints into the wall below. 

Consideration should be given to the treatment of any architectural finishes and M/E/P system 

components mounted on the wall in the affected areas. 

Detailing Considerations 

Typical details for installation of new steel ledgers and drilled diagonal dowels are shown in 

Figures 13.4.3-1 and 13.4.3-2, respectively. For these techniques, little or no prior cleaning or 

preparation of the existing concrete surfaces will be required. Alternatively, refer to Section 

12.4.3 for discussion related to installation of a concrete ledger. Where drilled concrete anchors 

or drilled diagonal dowels (acting as shear pins through the slab-wall joint region) are to be 

installed through precast concrete planks, the prestressed tendons must be avoided. Also, 

consideration should be given to means of dealing with any hollow core voids in the planks that 

may be encountered. Screen tubes similar to those used in brick masonry anchorage details could 

be used, or the voids could be filled, at least locally, with grout. 

Cost/Disruption 

Installation of steel section ledgers and drilled concrete anchors and dowels are simple and well 

known basic techniques that should be relatively inexpensive. There will be noise and vibration 

associated with the drilling, but the work is essentially “dry” and not particularly messy. 

However, given the nature of this type of structure, there are likely to be many walls distributed 

throughout the building, thus the work will be pervasive. If concrete ledgers are installed, the 

level of cost and disruption will be considerably higher (refer to the discussions regarding 

installation of concrete collectors at Section 12.4.3). 

Construction Considerations 

Installation of the drilled concrete anchors or diagonal dowels may require some precision to 

avoid prestressing tendons or hollow core voids. Once the pattern of these items is defined, a 

steel template, perhaps the steel ledger itself, can be used. If some flexibility in the exact location 

of the anchors is required, consider using oversized holes and welded plate washers, and provide 

extra holes. 
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Figure 13.4.3-1: Added Support and Shear Strength at Slab-Wall Joint 

The drilled anchors and dowels will require testing, which should be performed by persons 

experienced in torque and/or tension testing of diagonal dowels and overhead installations. For 

the drilled shear pin dowels that do not project out of the slab, additional similar dowels can be 

installed for testing purposes only. 

Proprietary Concerns 

The basic materials are generic. 

13.4.4 Reduce Flexural Capacity of Shear Walls to Reduce Shear Demand 

Deficiency Addressed by the Rehabilitation Technique 

Degradation of shear-critical walls 

Description of the Rehabilitation Technique 

In buildings with many walls and high strength, many of the walls may be shear-critical and 

prone to strength degradation and/or possible reduction of gravity support capacity. A sudden 
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Figure 13.4.3-2: Added Shear Capacity at Slab-Wall Joint 

loss of lateral strength and stiffness, particularly all at one floor, should be avoided. To avoid 

this condition, an alternative to shear enhancement is to reduce the level of possible shear 

demand that the wall can experience by reducing flexural capacity. This approach may be 

applied to one or more walls, in a system with adequate global strength, to change the expected 

post-yield behavior from a brittle shear failure to a more ductile flexural yielding. This approach 

also may be applied in other circumstances to an individual wall that cannot be strengthened for 

some reason, but must be protected from serious damage, even as other walls are strengthened or 

new walls are added in other locations within the system. This technique will generally involve 

cutting a number of the existing wall chord bars. However, an alternative approach that may be 

applied to a long wall would be to make one or more vertical cuts in the wall, creating a series of 

shorter panels with reduced flexural capacity. 

Design Considerations 

Application of this technique to a particular shear wall is intended to replace a brittle shear 

failure mechanism with a more ductile flexural hinging mechanism. It is not recommended that 

this technique be applied in conditions that may result in creating a brittle tension failure 

mechanism. Therefore, the number and location of cut bars must be determined with careful 

analysis of the detail of the expected mechanism (strain compatibility, flexural hinge length, 

etc.). Bar cuts should be sufficiently staggered and a major reduction in flexural capacity should 

not be attempted with this method. The impact of this technique on architectural and M/E/P 
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systems and components is likely to be relatively limited. Only fixtures in the immediate vicinity 

of the chord bar cuts or vertical wall slices are likely to be affected. However, client perception 

of this reductive technique may be decidedly negative. 

Detailing Considerations 

Cutting the selected existing wall chord bars must be done with great care and precision to avoid 

cutting adjacent chord bars or the confining transverse reinforcement. If the wall reinforcement 

can be adequately mapped by using metal detector, x-ray or ground penetrating radar techniques, 

then the selected bars could be cut by coring through the concrete cover. However, congested 

chord reinforcement and/or closely spaced transverse ties may require that the concrete cover be 

chipped away to expose the bars before cutting with either a core drill or a torch. 

In the case of reducing the overall length of a long wall with vertical cuts, the cuts can be made 

with a circular concrete saw. In order to extend the cuts as close to the floor slabs as possible the 

cuts may be made from each side of the wall, to keep the depth of the cut and the radius of the 

saw blade to about half the wall thickness. 

Cost/Disruption 

This technique will be less costly than alternative methods to increase the shear capacity of a 

wall. The extent of the work is very localized and the impact on nonstructural components will 

be limited. However, there will always be noise and vibration associated with any concrete 

chipping, coring or sawcutting, and the latter two operations can be very wet and messy. 

Construction Considerations 

Access must be available to the locations where chord bars are to be cut or the wall is to be 

sawcut. However, this will certainly be less of an imposition than any other alternative to 

increase the shear capacity of the wall. 

Employment of this technique will most likely require special scheduling and/or sequencing 

considerations relative to the other strengthening work, to avoid creating a weakened structure 

during the course of the overall retrofit project. 

Proprietary Concerns 

The basic materials are generic. 

13.5 References 

ACI 440.2R-02, 2002, “Guide for the Design and Construction of Externally Bonded FRP 

Systems for Strengthening Concrete Structures,” American Concrete Institute, ACI Committee 

440, Chicago, IL 

ICRI (International Concrete Repair Institute), 2003, “Guide for Surface Preparation for Repair 

of Deteriorated Concrete Resulting from Reinforcing Steel Corrosion,” Publication 03730, 

International Concrete Repair Institute. 
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FIB (Federation Internationale de Beton), 2001, Externally Bonded FRP Reinforcement for RC 

Structures, Technical Report, Task Group 9.3 FRP Reinforcement for Concrete Structures 

Bulletin 14, July. 

Ghobarah, A., 2004, “Seismic Retrofit of RC Walls Using Fibre Composites,” McMaster 

University, Hamilton, Onterio, Canada. 

Kaw, A.K., 1997, “Mechanics of Composite Materials,” CRC Press LLC, FL. 

Laursen, P.T., Seible, F., and G.A. Hegemier, 1995, Seismic Retrofit and Repair of Reinforced 

Concrete with Ovelays, University of California, San Diego, CA. 

Priestley, N., Seible, F. and G.M. Calvi, 1996, Seismic Design and Retrofit of Bridges, John 

Wiley & Sons, New York, NY. 

Teng, J.G., J.F. Chen, S.T. Smith, and L. Lam, 2002, FRP Strengthened RC Structures, John 

Wiley & Sons, New York, NY. 
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Chapter 14 - Building Type C2f: Concrete Shear Walls 

(Gravity Frame Systems) 

14.1 Description of the Model Building Type 

Reinforced concrete walls in a building will act as shear walls whether designed for that purpose 

or not. Therefore, concrete buildings that contain any significant amount of concrete wall will 

fall into this category. However, there are two distinctly different types of concrete wall 

buildings: those that contain an essentially complete beam/slab and column gravity system, and 

those that use bearing walls to support gravity load and have only incidental beam and column 

framing. In this document, these building types have been separated and are designated C2f for 

the gravity frame system and C2b for the bearing wall. This section covers the building with 

gravity framing system. Although it is typically assumed that the gravity framing is not part of 

the lateral force-resisting system, the framing could add stiffness to the building, particularly 

near the top of taller buildings. This building type is very common and has been used in a wide 

variety of occupancies and in all sizes. 

Figure 14.1-1: Building Type C2f: Concrete Shear Walls 

(Gravity Frame Systems) 
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Variations in Framing Systems 

There are wide overall variations within this building type due to the possible configuration and 

extent of the concrete walls, the many types of vertical framing systems used, and the lateral 

stiffness interaction between the two. In buildings with incidental concrete walls and a 

substantial beam-column gravity frame system, this building type merges with Building 

Type C1. If the building type is unclear, reference should be made to both Chapter 12 and this 

chapter. 

Gravity frame systems in this building type include cast-in place concrete beam and slab, oneway joists, two-way waffles, and two-way or flat slabs. 

In older buildings that are seismically deficient, the walls were often intended for fire protection 

of vertical shafts, as exterior closure walls, or as bearing walls. However, buildings built in 

regions of high seismicity in the 1950s, 1960s or early 1970s often were designed with a shear 

wall lateral force-resisting system, but they are now found deficient due to low global strength, a 

highly torsional plan layout or detailing that leads to premature shear failure 

In buildings designed with shear walls, the walls are either strategically placed around the plan, 

or at the perimeter. Shear walls systems placed around the entire perimeter almost always 

contain windows and other perimeter openings and are often called punched shear walls. Older 

buildings will have concrete walls somewhat arbitrarily placed in plan. 

Floor and Roof Diaphragms 

The floor and roof diaphragms in this building type are essentially the same as the bearing wall 

system and are almost always cast-in-place concrete. The diaphragms are stiff and strong in 

shear because the horizontal slab portion of the gravity system is either thick or frequently 

braced with joists. However, one way joist systems could be inadequate in shear in the direction 

parallel to the joists. Collectors are seldom in place and transfer of load from diaphragm to shear 

wall must be carefully considered. 

Foundations 

There is no typical foundation for this building category. Foundations could be found of every 

type depending on the height of the building, the span of the gravity system and the site soil. 

14.2 Seismic Response Characteristics 

Shear wall buildings, unless configured with only incidental or minimal walls, will typically be 

quite stiff. Elastic and early post-elastic response will therefore be characterized with lower-than 

average drifts and higher-than-average floor accelerations. Damage in this range of response 

should be minimal. 

Overall post-elastic response is highly dependent on the specific characteristics of the shear walls 

and the gravity frame components. 
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Shear Wall Behavior 

The walls must first be evaluated to determine if they contribute sufficient strength or stiffness to 

be considered significant. Often walls around vertical shafts are thin and lightly reinforced and 

will have little effect on the overall building response. Although retrofit techniques are similar 

such a building may be classified as Type C1. 

When subjected to ever increasing lateral load, individual shear walls or piers will first often 

force yielding in spandrels or other horizontal components restricting their drift, and eventually 

either rock on their foundations, suffer shear cracking and yielding, or form a flexural hinge near 

the base. Shear and flexural behavior is quite different, and estimates of the controlling action 

are affected by the distribution of lateral loads over the height of the structure. 

Yielding of spandrels or other coupling beams can cause a significant loss of stiffness in the 

structure. Flexural yielding will tend to maintain the strength of the system, but shear yielding, 

unless well detailed, will degrade the strength of the coupling component and the individual 

shear wall or pier will begin to act as a cantilever from its base. 

Rocking is often beneficial, limiting the response of the superstructure. However, the amplified 

drift in the superstructure from rocking must be considered. In addition, if varying wall lengths 

or different foundation conditions lead to isolated or sequencing rocking, the transfer of load 

from rocking walls must be investigated. In buildings with basements, the couple created from 

horizontal restraint at the ground floor diaphragm and the basement floor/foundation (often 

termed the “backstay” effect) may be stiffer and stronger than the rocking restraint at the 

foundation and should be considered in those configurations. 

Shear cracking and yielding of the wall itself is generally considered undesirable, because the 

strength and stiffness will quickly degrade, increasing drifts in general, as well as potentially 

creating a soft story or torsional response. However, in accordance with FEMA 356 (FEMA 

2000), shear yielding walls or systems can be shown to be adequate for small target 

displacements. 

Flexural hinging is considered ductile in FEMA 356 and will degrade the strength of the wall 

only for larger drifts. Similar to rocking, the global effect of the loss of stiffness of a hinging 

wall must be investigated. 

Gravity Frame Behavior 

The lateral strength and stiffness of gravity frames will vary considerably among buildings in 

this type. In some configurations of this building type, the gravity frame will not significantly 

participate in the response. However, it is not uncommon in these buildings for a stiff and brittle 

gravity system to dominate both response and the extent of damage. For example, if concrete 

spandrels or sills on the perimeter of the building restrain the gravity columns (the “short 

column”), the column must take the full story drift over a short height, potentially causing shear 

failure and loss of gravity load capacity. Other gravity systems, such as flat slab or heavy beam 

and column systems, can also be sensitive to drifts, particularly to the increased drifts near the 

top of buildings with walls of a height-to-width ratio over 3. The frame action of the gravity 
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system of these buildings may be beneficial or could form a deficiency, but in any case the 

interaction with the shear walls should be considered. 

14.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

See Table 14.3-1 for deficiencies and potential rehabilitation techniques particular to this system. 

Selected deficiencies are further discussed below by category. 

Global Strength 

Older buildings placed in this category may have minimal shear walls and seismic displacements 

will likely put excessive demand on the walls regardless of the shear or flexural behavior. The 

most common method of mitigating this deficiency is to add more shear walls, although for 

smaller buildings steel braced frames have been used. The nonlinear behavior of steel braced 

frames must be carefully studied for compatibility with the other shear walls and the gravity 

frame. Additional bending resistance of existing walls can be obtained by enhancing existing 

chords, although such walls should not be strengthened to become shear critical. If the majority 

of walls are shear-critical and strength degradation is the primary concern, shear strength can be 

added to the existing walls with concrete or FRP overlays. 

Global Stiffness 

The most important issue in a building of this type that exhibits large interstory drifts is the 

ability of the gravity system to accept the drifts while sustaining their loads. Excess drifts could 

be caused by inadequate length of wall, by rocking at the foundations, or, at the upper stories, by 

a deformed shape characterized by bending deformations. In most buildings, strength and 

stiffness are closely related, and inadequate stiffness is mitigated by adding new elements or 

stiffening existing ones, which generally will also increase strength. Damping can also be added 

to reduce drifts but care must be taken to achieve the desired damping with small displacement 

expected in a shear wall building. 

Configuration 

The two most common configuration deficiencies in this building type are 1) severe torsion 

caused by eccentrically placed shafts or towers, and 2) shear walls or full stories that are weak or 

soft from openings in the walls or from discontinuous walls that may not run through the ground 

floor or basement floor. Completely discontinuous walls also create a load transfer deficiency 

for both overturning and shear. 

Load Path 

Common load path deficiencies include discontinuous shear walls, as discussed above, and 

collectors for the shear walls. Collectors can be added with steel members or new reinforcing 

and concrete. 

Component Detailing 

Shear walls in most older buildings meet none of the current detailing requirements covering 

minimum shear reinforcement, for confinement of chords, and the walls are commonly shear 

critical. FEMA 356 allows these deficiencies at controlled displacement levels. A common 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 14 – Type 

14-5 

Table 14.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C2f Build

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Dem

Global Strength 

Insufficient inplane wall shear strength 

Concrete/masonry shear wall [12.4.2] 

Steel braced frame [12.4.1] 

Steel plate shear wall 

Concrete wall overlay [21.4.5] 

Fiber composite wall overlay [13.4.1] 

Steel wall overlay 

Seismic isola[24.3] 

Reduce flexucapacity [13.

 Insufficient 

flexural capacity 

Concrete/masonry shear wall [12.4.2] 

Steel braced frame [12.4.1] 

Add or enhance chords [12.4.3] 

 Inadequate 

capacity of coupling beams 

Concrete/masonry shear wall [12.4.2] 

Steel Braced frame [12.4.1] 

Strengthen beams 

Improve ductility of beams [13.4.2] 

Global Stiffness 

Excess drift (normally near the top of the building) 

Concrete/masonry shear wall [12.4.2] 

Steel braced frame [12.4.1] 

Fiber composite column wrap [12.4.4] 

Concrete/steel column jacket [12.4.5] 

Provide detailing of all other elements to accept drifts 

Thicken walls 

Supplementadamping [24

Configuration Discontinuous 

walls 

Concrete/masonry shear wall [12.4.2] 

 Enhance existing 

column for overturning loads 

Improve connection to diaphragm [13.4.3] 

 Soft story or weak 

story 

Add strength or stiffness in story to match balance of floors 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 14 – Type 

14-6 

Table 14.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C2f Build

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Dem

Configuration (continued) 

Re-entrant corner Add floor area to 

minimize effect of corner 

Provide chords in diaphragm [12.4.3] 

 Torsional layout Add balancing 

walls, braced frames, or moment frames 

Load Path Inadequate 

collector 

Add steel collector [12.4.3] 

Add concrete collector [12.4.3] 

Strengthen existing beam or slab 

Enhance splices or connections of existing beams 

 Discontinuous 

Walls 

Provide new wall support components to resist the maximum expected overturning moment 

Strengthen the existing support columns for the maximum expected overturning moment 

Provide elements to distribute the shear into the diaphragm at the level of discontinuity 

Component Detailing 

Wall inadequate for out-of-plane bending 

Add strongbacks Concrete wall 

overlay [21.4.8] 

 Wall shear critical Concrete wall 

overlay [21.4.8] 

Fiber composite wall overlay [13.4.1] 

Reduce flexucapacity of w[13.4.4] 

 Inadequate 

displacement capacity of gravity columns 

Enhance ductility (see also global stiffness) 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 14 – Type 

14-7 

Table 14.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C2f Build

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Dem

Diaphragms Inadequate inplane shear 

capacity 

Concrete topping slab overlay 

Fiber composite overlays [22.2.5] 

 Inadequate chord 

capacity 

New concrete or steel chord member [12.4.3] 

 Excessive stresses 

at openings and irregularities 

Add chords [12.4.3] 

Foundations See Chapter 23 

[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 14 – Type C2f: Concrete Shear Walls

14-8 

improvement to these walls is to enhance shear strength to be equal or greater than the maximum 

that can be developed in the wall, based on bending strength. 

Diaphragm Deficiencies 

The most common diaphragm deficiency in this building type is a lack of adequate collectors. 

The addition of effective collectors in an existing diaphragm is difficult and disruptive. Existing 

strength to deliver loads to the shear walls should be studied carefully before adding new 

collectors. 

Foundation Deficiencies 

See Chapter 23. 

14.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

The most relevant recommendations listed in Table 14.3-1 are similar to techniques also 

associated with other concrete building types such as C1 and C2b or general techniques applied 

to concrete diaphragms. Details concerning these techniques can be found in other chapters. 

14.5 References 

FEMA, 2000, Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 

356, Federal Emergency Management Agency, Washington, D.C.
Techniques for the Seismic Rehabilitation of Chapter 15—Types C3/C3A Concrete Frames 

Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

15-1 

Chapter 15 - Building Types C3/C3A: Concrete Frames with 

Infill Masonry Shear Walls 

15.1 Description of the Model Building Type 

Building Type C3 is normally an older building that consists of an essentially complete gravity 

frame assembly of concrete columns and floor systems. The floors consist of a variety of 

concrete systems including flat plates, two-way slabs, and beam and slab. Exterior walls, and 

possibly some interior walls, are constructed of unreinforced masonry, tightly infilling the space 

between columns horizontally and between floor structural elements vertically, such that the 

infill interacts with the frame to form a lateral force-resisting element. Windows and doors may 

be present in the infill walls. The buildings intended to fall into this category often feature 

exposed clay brick masonry on the exterior. Figure 15.1-1 shows an example of this building 

type. 

It is important to note that similar buildings with exterior masonry infill sills below windows that 

extend column to column do not behave with strut action and should be classified as Building 

Type C1 or C2f. In fact, such infill sills often create “short columns” that must absorb the entire 

story drift over the unrestrained height of the window, which can often be an extreme deficiency 

in poorly reinforced columns. The C3A building type is similar but has floors and roof that act 

as flexible diaphragms such as wood, or untopped metal deck. 

Figure 15.1-1: Building Type C3: Concrete Frames with Infill Masonry Shear Walls 
Techniques for the Seismic Rehabilitation of Chapter 15—Types C3/C3A Concrete Frames 

Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

15-2 

Variations Within the Building Type 

The building type was identified primarily to capture the issues of interaction between 

unreinforced masonry and concrete gravity framing. The archetypical building has solid clay 

brick at the exterior with one wythe of brick running continuously past the plane of the column 

and beam and two or more wythes infilled within the plane of the column and beam. The 

exterior wythe of clay brick forms the finish of the building although patterns of terra cotta, 

stone, or precast concrete may be embedded into the brick. However, there can be many 

variations to this pattern depending on the number and arrangement of finished planes on the 

exterior of the building. For example, the full width of the infill wall may be located with the 

plane of the column and beam with a pilaster built out and around the column and a horizontal 

band of brick or other material covering the beam; the beam may also be slightly offset from the 

centerline of the column to accommodate the pattern of exterior finishes. 

Hollow clay tile masonry may also be used as an exterior infill material. Although this material 

often has a high compression strength, the net section of material available to form the 

compression strut within the frame will normally contribute a lateral strength of only a small 

percentage of the building weight. The material being brittle and the wall being highly voided, 

these walls may also lose complete compressive strength quite suddenly. Therefore, walls of 

hollow clay tile infill will probably not contribute a significant portion of required lateral 

resistance except in areas of low seismicity and/or when walls are arranged as infill on both the 

exterior and interior of the building. 

More recent buildings may have unreinforced concrete block masonry configured as an exterior 

infill wall, with a variety of finish materials attached to the outside face of the concrete block. 

Similar to hollow clay tile walls, these walls may exhibit moderate to low compressive strength 

and brittle behavior that marginalizes their usefulness as lateral elements. In addition, hollow 

concrete block exterior walls often will not be installed tight to the surrounding framing, 

eliminating infill compression strut behavior. 

Floor and Roof Diaphragms 

Floors are often flat plates or two-way slabs. Beam and slab or beam and joist systems will also 

be found in this building type. Typically, these slabs provide adequate diaphragms. 

Building Type C3A will have heavy timber floors with one or more layers of sheathing forming 

a diaphragm. The flexibility of such diaphragms will often form a seismic deficiency because, 

assuming no interior shear elements, the large drift at the diaphragm mid-span will damage 

perpendicular walls and gravity framing. Specific strengthening techniques for this building type 

are not covered here. For generalized strengthening of diaphragms, see Chapter 22. 

Foundations 

There is no typical foundation for this building type. Foundations can be found of every type 

depending on the height of the building, the span of the gravity system and the site soil. The 

exterior walls are exceptionally heavy and typically will be supported by a continuous concrete 

footing or often a continuous concrete wall forming a basement space below. 
Techniques for the Seismic Rehabilitation of Chapter 15—Types C3/C3A Concrete Frames 

Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

15-3 

15.2 Seismic Response Characteristics 

Both in terms of stiffness and strength, the exterior infill walls will form the effective lateral 

system for this building type. The effectiveness of the system depends on the size and extent of 

openings and articulation of the plane of the wall. With solid or nearly solid infill panels, strut 

action will be stiff and strong. As openings in panels increase in size, struts or combinations of 

struts cannot effectively form around the opening, and the concrete columns and beams may 

begin to work as a moment frame, with “fixity” at the beam-column joint provided by the 

masonry. For low and moderate intensity shaking, the exterior walls may provide adequate 

strength to satisfy the specified performance objective. As the shaking demand increases, the 

masonry will tend to crack and spall, losing stiffness and potentially creating a falling hazard. 

The complete concrete gravity system, characteristic of this building type, will provide additional 

stability, but may quickly degrade in strength and stiffness due to inadequate column reinforcing. 

However, in wall configurations with large height-to-width ratios, end or corner columns could 

experience large compression or tension loads from overturning, leading to rapid degradation of 

column lateral and gravity capacity. 

This building type is often characterized by a commercial store-front first floor with little or no 

infill at that level on one or more faces of the building. This condition can cause a soft story 

condition or a severe torsional response if open on one or two sides only. Such conditions can 

lead to concentration of seismic deformation at the open level, degradation of the columns, and 

possible P-delta failure. 

15.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

See Table 15.3-1 for deficiencies and potential rehabilitation techniques particular to this system. 

Deficiencies related to specifically to concrete moment frames and to masonry shear walls are 

shown in Table 12.3-1 and Table 18.3-1, respectively. Selected deficiencies are further 

discussed below by category. 

Global Strength 

The overall strength provided by the exterior walls may be insufficient to prevent serious 

degradation and resulting amplified displacements that can lead to irreparable damage or even 

instability. The strength may be limited by inadequate number of panels of infill, excessive 

openings, or masonry weak in compressive strength. The standard approach to such deficiencies 

will be to add new, relatively stiff lateral force-resisting elements such as concrete shear walls or 

steel braced frames often located on the interior between existing columns. The infill itself could 

also be strengthened by adding a layer of reinforced concrete on the interior surface, although 

whether such a system is acting as infill or as a shear wall must be checked by analysis. Unless 

all infill has been backed by a support system, the damage state of the infill wall must be 

estimated for the expected drifts of the combined system to determine if the desired performance 

has been achieved. 

Global Stiffness 

The most important issue in a building of this type that exhibits large interstory drifts is the 

ability of the gravity system to accept the drifts while sustaining their loads. In most buildings, 
Techniques for the Seismic Rehabilitation of Chapter 15—Types 

Existing Buildings: FEMA 547 with I

15-4 

Table 15.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C3/C3A Bui

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Deman

Global Strength 

Inadequate length of exterior wall 

Interior concrete walls [12.4.2] Interior steel braced frames [12.4.1] 

Concrete wall overlay [21.4.5] Fiber composite wall overlay [21.4.6] 

 Excessive sized 

openings in infill panels 

Interior concrete walls [12.4.2] Interior steel braced frames [12.4.1] 

Infill selected openings [21.4.7] Fiber composite wall overlay [21.4.6] 

 Inadequate 

columns for overturning forces 

Add confinement Add tensile capacity on outside surface of column. 

 Weak or 

deteriorated masonry 

Interior concrete walls [12.4.2] Interior steel braced frames [12.4.1] 

Point outside and/or inside wythes of masonry Inject wall with cementitious grout Fiber composite overlay [21.4.6] 

Global Stiffness 

See Global Strength 

Configuration Soft or weak story Interior concrete 

walls [12.4.2] Interior steel braced frames [12.4.1] 

 Torsion from one 

or more solid walls 

Balance with Interior concrete walls Balance interior steel braced frames 
Techniques for the Seismic Rehabilitation of Chapter 15—Types 

Existing Buildings: FEMA 547 with I

15-5 

Table 15.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C3/C3A Bui

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Deman

Configuration (continued) 

Irregular Plan Shape 

Balance with Interior concrete walls Balance with Interior steel braced frames 

Load Path Out-of-plane 

failure of infill due to loss of anchorage or slenderness of infill 

Provide surface wall supports [13.4.3] Shotcrete overlays [21.4.5] Fiber composite overlay [21.4.6] 

 Inadequate 

connection of finish wythe to backing 

Add connections 

 Inadequate 

collectors 

Add steel collector or concrete collector [12.4.3] 

Component Detailing 

Inadequate columns splice for tension due to uplift force induced by infill 

Add splice plates Provide splice through added reinforced concrete encasement 

 Inadequate beam 

column connection to resist compression thrust 

Strengthen connection in shear with steel or fiber composite 
Techniques for the Seismic Rehabilitation of Chapter 15—Types 

Existing Buildings: FEMA 547 with I

15-6 

Table 15.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for C3/C3A Bui

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Deman

Component Detailing (continued) 

Weak or incompletely filled joint between masonry and surrounding steel components 

Create appropriate clean void and repack with masonry and/or mortar. Inject voids with cementitious grout or epoxy 

Diaphragms See Chapter 22 Foundations See Chapter 23 

[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 
Techniques for the Seismic Rehabilitation of Chapter 15—Types C3/C3A Concrete Frames 

Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

15-7 

strength and stiffness are closely related, and inadequate stiffness is mitigated by adding new 

elements or stiffening existing ones, which generally will also increase strength. Damping can 

also be added to reduce drifts, but care must be taken to achieve the desired damping with the 

small displacement expected in a shear wall building. 

Configuration 

Two configurational deficiencies are common in this building type. The first is a soft and weak 

story at the street level created by commercial occupancies with exterior bays with little or no 

infill. This deficiency can be corrected by adding selected bays of infill or by adding shear walls 

or braced frames at this level. The second common issue is a plan torsional irregularity created 

by solid masonry walls on property lines coupled with walls with many openings on street fronts. 

If shown by analysis to be necessary, torsional response can be minimized by stiffening the more 

flexible side of the building with more infill or by the addition of lateral elements. In rare cases, 

the solid walls can be balanced with the open side by selected removal of panels or 

disengagement of the infill strut action. 

Load Path 

The primary load path issue with this building type is to assure that the mass of the exterior walls 

will not become disengaged from the frame which will both prevent infill strut action as well as 

to create a significant falling hazard on the street below. 

In-plane, the articulation of the exterior walls may result in offsets of the wall plane between 

floors. The presence of a complete load path and maintenance of confinement for strut formation 

must be reviewed in such instances. 

If new lateral load-resisting elements are added, existing slab and/or beam construction may need 

to be strengthened to provide adequate collectors. 

Component Detailing 

In order to qualify as an infill lateral force-resisting element, the infill must be installed tight to 

the surrounding concrete elements. Loose or incomplete infill can be mitigated with local 

patching of the masonry or by injection of cemetitious or epoxy grout. 

As previously noted, the infill adjacent to columns must be of sufficient stiffness to provide a 

floor to floor diagonal strut. With narrow piers surrounding columns, the jamb and header 

masonry may restrain the column such that the entire story drift must be absorbed in the central 

length of the column, often creating dangerous shear failures. 

The detailing of the concrete frame forming the confinement for the masonry is important to 

achieve infill strut behavior. The connection of beam to column must be capable of resisting the 

strut compression forces from the masonry. The shear capacity of the beam-to-column 

connection is often critical. Strengthening of these connections may require removal of 

considerable masonry to obtain adequate access. Some of the techniques developed to 

strengthen concrete moment frames may be applicable. In addition, splices in vertical 

reinforcing of column splices may be inadequate to transfer the tensile overturning forces created 
Techniques for the Seismic Rehabilitation of Chapter 15—Types C3/C3A Concrete Frames 

Existing Buildings: FEMA 547 with Infill Masonry Shear Walls

15-8 

by strut action. These areas can be confined to reduce the required splice length or augmented 

with additional tensile strength from the surface. 

Diaphragm Deficiencies 

Concrete slab diaphragms often will be adequate. The connection of slab to exterior wall should 

be reviewed. 

See Chapter 21 on URM construction for discussion of wood diaphragms in this type of building. 

Foundation Deficiencies 

No systematic deficiency in foundations should be expected solely due to the characteristics of 

this building type. 

15.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

Most significant recommendations listed in Table 15.3-1 are similar to techniques more 

commonly associated with other building types such as various concrete buildings (C1, C2b, or 

C2f), unreinforced masonry bearing wall buildings (URM), or general techniques applied to 

concrete diaphragms. Details concerning these techniques can be found in other chapters. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 16 – Type PC1: Tilt-up Concrete Shear Walls

16-1 

Chapter 16 - Building Type PC1: Tilt-up Concrete Shear Walls 

16.1 Description of the Model Building Type 

Building Type PC1 is constructed with concrete walls, cast on site and tilted up to form the 

exterior of the building. PC1 buildings are used for many occupancy types including warehouse, 

light industrial, wholesale and retail stores, and office. The majority of these buildings are one 

story; however, tilt-up buildings of up to three and four stories are common, and a limited 

number with more stories exist. For many years, tilt-up buildings have been primarily large boxtype buildings with the tilt-up walls at the building perimeter; this is by far the largest group of 

PC1 buildings in the U.S. inventory. In recent years, tilt-up construction has expanded to more 

varied uses and building configurations. Figure 16.1-1 illustrates one example of a PC1 building. 

Figure 16.1-1: Building Type PC1: Tilt-Up Concrete Shear Walls 

Key to the PC1 building type as addressed by this chapter is the combination of a flexible roof 

diaphragm and rigid walls. Lateral forces in PC1 buildings are resisted by flexible wood 

sheathed or steel deck roof diaphragms, wood, composite steel deck, or precast floor diaphragms, 

and tilt-up concrete shear walls. In some local areas, walls are concrete panels or T-beams cast 

off-site, rather than on-site; PC1 also applies to this variation. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 16 – Type PC1: Tilt-up Concrete Shear Walls

16-2 

One-story PC1 buildings with flexible roof diaphragms are the primary focus of this chapter; 

however, discussion of wall-to-flexible diaphragm anchorage is equally applicable to the roof of 

multistory PC1 buildings. For other rehabilitation issues in multistory buildings, refer to the C1

Building Type. For variations with rigid diaphragms at floors and roof, see Building Type PC2. 

Guidelines for Seismic Evaluation and Rehabilitation of Tilt-Up Buildings and Other Rigid 

Wall/Flexible Diaphragm Structures or SEAONC Guidelines (SEAONC, 2001) provides a 

substantial collection of information on West Coast PC1 building configurations, experience 

with earthquake performance, rehabilitation priorities, and techniques for rehabilitation. This 

chapter highlights the major rehabilitation considerations from the SEAONC Guidelines

document, and provides suggested adaptations for construction variations. Provisions addressing 

rehabilitation can also be found in International Existing Building Code (IEBC) Appendix 

Chapter A2 (ICC, 2003b), City of Los Angeles Building Code Chapter 91 (City of Los Angeles, 

2002) and Guidelines for Seismic Retrofit of Existing Buildings (GSREB) (ICBO, 2001). These 

provisions focus on wall anchors, diaphragm cross-ties, and collectors, with the goal of hazard 

reduction. This chapter and the SEAONC Guidelines address a broader range of rehabilitation 

issues and techniques. 

The Tilt-up Construction and Engineering Manual (Tilt-Up Concrete Association, 2004), now in 

its sixth edition, serves as a primary resource for design and detailing practice for new tilt-up 

construction. 

Walls 

Tilt-up exterior walls are the primary vertical elements in the lateral force-resisting system. Tiltup buildings with large plan areas may have interior tilt-up walls or braced steel frames 

providing additional lateral resistance. Large-scale construction of tilt-up buildings began in the 

1950s and 1960s with primarily solid wall panels for warehouse and light-industrial use. As tiltup construction expanded to commercial use in the 1970s and 1980s, the wall panels changed to 

include large window and door openings and multistory construction. The publications 

Recommended Tilt-up Wall Design (SEAOSC, 1979) and Test Report on Slender Walls (ACISEASC, 1982) document the change from use of code-prescribed height to thickness (h/t) limits 

of 25 for bearing walls and 36 for other walls to use of much higher h/t ratios, in combination 

with rational analysis of slenderness effects. Other code changes of interest have occurred for 

wall pier reinforcing requirements and wall panel connection to the foundation or slab-on-grade. 

Most tilt-up panels are a single piece from the foundation to the top of the building; however, 

some tilt-up systems use separate wall panels at upper stories, or lintel (spandrel) panels that are 

supported on other tilt-up wall panels. Welded connections of these panels can be damaged when 

they restrict panel movement under earthquake load. Prior to start of rehabilitation design, it is 

important to identify the tilt-up panel joint and support locations and connection condition. 

Surface treatments (such as exposed aggregate) and reveal joints are commonly used to visually 

enhance tilt-up wall panels. These treatments effect the location and dimensions of critical 

sections for design. 

Connections between adjacent tilt-up panels are often relied on to provide continuity for 

diaphragm chords and collectors. Connection types have varied over the years. Early tilt-up walls 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 16 – Type PC1: Tilt-up Concrete Shear Walls

16-3 

were often joined by cast-in-place pilasters. Later, welded connections between cast-in 

embedments or between horizontal reinforcing steel were commonly used. Both of these 

connection systems experience some problems with fractures at welds due to panel shrinkage 

and temperature movement. As a result, the codes placed carbon equivalent requirements on 

rebar to be welded, and detailing to minimize restraint of panel movement was pursued. In 

Recommended Tilt-up Wall Design (SEAOSC, 1979), minimal interconnection between the 

panels and special detailing of chord and collector connections are suggested to allow movement. 

Detailing suggestions included 1) breaking the bond between the chord reinforcing and concrete 

for one fourth the panel width in from each panel end and 2) use of steel angle chord/collector 

members connected in the center portion of the panel, but slotted near the panel ends. The panel 

connection detailing should be considered in evaluating the behavior of the chords and collectors 

and in the distribution of shear to the tilt-up panels. 

Gravity Load Support at the Building Perimeter 

It is most common for roof and floor framing members to be supported on the tilt-up panels at 

the building perimeter; however, some contractors have found it advantageous to provide steel 

gravity columns at the inside face of the tilt-up walls. This separation of the gravity and lateral 

systems makes construction tolerances for items embedded in the tilt-up panels and construction 

sequence less critical. The use of columns requires some modification of wall anchorage details 

at girders, but has little or no effect on the typical wall to diaphragm attachments, or fundamental 

building behavior. 

Floor and Roof Diaphragms 

The PC1 roof system will generally be of either wood or steel construction. In the western states 

(primarily California, Oregon, Washington), roof systems are almost exclusively wood structural 

panel sheathed. Subpurlins (joists), purlins, and girders are most often wood; however, open web 

trusses with wood chords or nailers are also used. Wood girders are most often supported on 

steel columns at the building interior. Sheathing fastening and therefore unit shear capacity in 

wood structural panel diaphragms generally varies by nailing zone or area. In older West Coast 

PC1 buildings, roofs can be lumber sheathed, and roof framing can include bow-string trusses. 

Outside of the western states, roof systems are almost exclusively sheathed with steel decking 

with rigid foam or nonstructural concrete insulation. The roof framing system is most commonly 

of steel open web trusses (bar joists) used in combination with truss girders or hot-rolled steel 

beams. 

Interior Additions 

Mezzanines and interior second stories constructed within tilt-up buildings are common. The 

interior addition may be seismically separated and braced independently of the building shell 

(exterior walls and roof), or may be attached to and supported by the shell. Attachment to the 

building shell raises two potential issues for the tilt-up: 1) whether the seismic load to the tilt-up 

system is significantly increased beyond that considered in initial design, and 2) whether the 

interior addition restrains seismic deflections of the building and creates an unintended load path. 

Both of these issues should be addressed in evaluation and rehabilitation. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 16 – Type PC1: Tilt-up Concrete Shear Walls

16-4 

Foundations 

Tilt-up buildings are often constructed on continuous or isolated spread footings; however, 

drilled pier and grade beam foundations are common in some regions. Often tilt-up wall panels 

are set on top of shims and grout pads on top of the foundation. Where continuous foundations 

are provided, grout may then be placed under the full length of the panel to provide continuous 

bearing. Often, no direct connection is provided between the wall panel and foundation. While 

not common practice in older tilt-up construction, in many buildings constructed in the 1980s 

and later, wall panel connection to the slab-on-grade were provided, allowing transfer of 

horizontal seismic forces to the slab. The most common connection uses rebar dowels, cast into 

the wall panel and a slab closure strip. Other approaches include welded or bolted connections 

between cast-in connection plates or threaded inserts. See Section 16.4.5 for further discussion. 

Due to the stiffness of the wall system, tilt-up buildings are best located on sites with very stable 

soils; however, they are often relegated to poor soil sites. Where tilt-up buildings are located on 

sites with soils subject to expansion, consolidation, or liquefaction, the effects of any damage to 

the foundation and wall connections due to soil movement should be considered in building 

evaluation. 

16.2 Seismic Response Characteristics 

The seismic response of the classic large box-type one-story PC1 building is characterized by 

rigid wall and flexible diaphragm behavior. In this type of building, the concrete walls will have 

a very short period, while the diaphragm has long-period behavior. Amplification of seismic 

forces near the center of wood sheathed diaphragms loaded in the transverse direction can be 

significant. This creates high demand on the diaphragm and out-of-plane wall anchorage near the 

center of the diaphragm, and it can also generate very high shear demand at the diaphragm 

connection to the end walls. This behavior has been replicated in instrumented buildings and in 

laboratory testing (Fonseca, Wood and Hawkins, 1996). Roof diaphragm amplification in the 

longitudinal direction can be lower, depending on the diaphragm aspect ratio; however, 

significant overstrength can result in large anchorage forces in this direction also. 

Observation of the earthquake performance of these buildings has identified as important 1) 

understanding the magnitude of the wall anchorage force and 2) detailing to eliminate weak links 

in the wall anchorage connection. With each subsequent earthquake since 1971, building code 

requirements have been revised to reflect changing understanding of the magnitude of forces 

generated in wall anchors and requirements for proper detailing; simply providing positive 

connections from the walls to the diaphragm has not resulted in adequate performance. A variety 

of weaknesses in the connections have kept anchorages from developing adequate capacity. 

Included are cross-grain tension of wood ledgers, rotation of non-symmetric connectors, lowcycle fatigue of straps that buckle under compression loads, net section fracture of straps with 

punches bolt holes, etc. The current code wall anchorage force and detailing requirements reflect 

a history of knowledge gained for wood structural panel diaphragms in large box-type buildings. 

Recent construction of PC1 buildings has moved away from this classic large box-type building 

with wood structural diaphragms. In most regions of the United States, PC1 buildings are now 

constructed with steel deck diaphragms. Many PC1 box-type buildings are being constructed 

with very tall walls, in which out-of-plane behavior could potentially modify building seismic 
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response. Tilt-up construction has expanded to include a wide variety of building uses with 

different building characteristics, including smaller and less regular plans, less regular diaphragm 

configurations, and some with interior as well as exterior tilt-up walls. The seismic response 

characteristics of these building configurations are not known. Their behavior is of much 

interest, however, because current building code provisions may not adequately address these 

variations and materials. The Tilt-Up Concrete Association has just initiated the TCA Seismic 

Performance Initiative, with the intent of identifying and developing strategic plans to resolve 

issues of building performance and code requirements for design and detailing. Initial work will 

focus on design models developed from instrumented building behavior (Freeman, Searer and 

Gilmartin, 2002). Research on tilt-up building performance is also in progress at Canterbury 

University, New Zealand. 

16.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

PC1 buildings with wood sheathed roof diaphragms have experienced structural damage and 

partial building collapse in a number of California earthquakes, as well as the 1964 Alaska 

earthquake. Partial collapse is almost exclusively associated with inadequate connection of the 

walls to the flexible roof diaphragm for out-of-plane loading. A variety of other types of damage 

to the wall panels, connections and roof diaphragms have been observed, including some interior 

diaphragm failures observed in the Northridge earthquake. These observations have been almost 

exclusively of buildings with wood diaphragms. 

Currently used design provisions for new buildings [IBC (ICC, 2003a), ASCE 7 (ASCE, 2005), 

and NEHRP (FEMA, 2004)] contain one set of requirements addressing wall anchors and crossties for concrete and masonry walls used with flexible diaphragms. These requirements do not do 

not differentiate between wood and steel deck diaphragms in application these requirements, nor 

do they differentiate between different possible tilt-up building configurations and behaviors. 

Opportunities to observe earthquake performance of steel deck diaphragms and newer building 

configurations have been limited to date. Until performance information is available from 

research or earthquake experience, the vulnerable behavior seen in wood diaphragm PC1

buildings needs to be considered a possibility for all PC1 buildings. The fundamental 

rehabilitation concept is positive anchorage of tilt-up walls to supporting diaphragms, with 

anchorage load paths adequate for forces both away from and towards the diaphragm. 

The SEAONC Guidelines document provides a detailed list of rehabilitation measures and 

relative priorities based on both potential hazard and the level of design of the existing 

construction. The major categories of deficiencies and rehabilitation (not prioritized) are: 

Out-of-plane wall anchors to walls and pilasters 

Diaphragm cross-ties 

Collectors 

Diaphragm strength, stiffness and openings 

Wall in-plane shear connections 

Wall in-plane capacity 

Wall in-plane base anchorage 

Wall out-of-plane bending 
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These deficiencies and rehabilitation measures are included in Table 16.3-1 and the general 

discussion below; however, the compilation of information in the SEAONC Guidelines is 

recommended as a useful resource. 

Global Strength and Stiffness 

Global strength and stiffness of the tilt-up walls have not been seen as a significant source of 

damage in PC1 buildings to date. This is likely due to the length of solid wall provided in older 

buildings, minimum reinforcing ratios, and the tendency to design at low stress levels so that a 

single layer of reinforcing without special detailing can be used. Strength and stiffness 

deficiencies are most likely to occur at wall lines with significant number of large penetrations 

and other locations where loads are carried by a limited number of panels. Rehabilitation 

measures include addition of new vertical elements, enhancing existing walls, and infilling 

openings in existing walls. 

Configuration 

Poor distribution of shear walls can result in torsionally irregular behavior of PC1 buildings. 

Common occurrences include an entire line of highly perforated tilt-up panels such as at loading 

dock walls in distribution and storage facilities and street front walls in commercial buildings. 

Concrete cracking and spalling have been seen in perforated wall panels that act as frames. The 

most direct approach to rehabilitation of this condition is the addition of strength and stiffness in 

line with the perforated wall. This can be accomplished through addition of new shear walls, 

enhancing of existing shear walls, or addition of steel braced frames. 

Re-entrant corners are reasonably common in large box-type PC1 buildings, either due to in-set 

panels (Figure 16.3-1), or an L-shaped building plan. The in-set walls create a hard spot in the 

diaphragm, restraining it from the deflection required to transmit load to the end wall. In most 

cases, the in-set walls will have to be considered shear walls supporting the diaphragm. Where 

chords and collectors have not been provided at the re-entrant corner, the diaphragm has been 

seen to pull away from the wall, damaging gravity and lateral load connections. Rehabilitation at 

re-entrant corners requires the provision of adequate chords and collectors, shear transfer to the 

in-set wall panels, and possibly the strengthening of the diaphragm, wall panels, and connections 

to the foundation. In some cases, high earthquake loads in existing elements may make it 

necessary to add new vertical elements in line with the re-entrant corner. The SEAONC 

Guidelines suggest that there may be diaphragm continuity over this interior diaphragm support, 

increasing the diaphragm reaction to the in-set wall line. Alternately, it may be possible to allow 

the wall at the re-entrant corner to rock, or to separate the wall from the diaphragm allowing the 

diaphragm to span to the exterior wall. These approaches require complex detailing, however. 

Some very large tilt-up buildings are constructed in configurations that resemble multiple 

buildings alongside each other, as seen in Figure 16.3-2. This figure illustrates a single large tiltup building constructed in three sections, separated by roof expansion joints. Thermal 

movements in the steel deck diaphragms make the expansion joints necessary. If each of the 

three building sections were provided a complete lateral force-resisting system and separated by 

adequate seismic joints, this building configuration would be of little concern. As shown in 

Figure 16.3-2, however, building sections are often laterally braced off of adjacent building 

sections, using shear transfer through the expansion joint. This configuration raises concerns of 
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Table 16.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for PC1 Buil

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Red

Dem

Global Strength 

Insufficient in-plane strength of shear walls or frames 

Steel braced frame [7.4.1] Concrete/masonry shear wall [21.4.8] 

Fiber composite wall 

overlay [13.4.1] 

Concrete wall overlay 

[21.4.5] 

Infill openings 

Global Stiffness 

Insufficient in-plane stiffness of shear walls or frames 

Steel braced frame [7.4.1] 

Concrete/masonry shear wall [21.4.8] 

Fiber composite wall 

overlay [13.4.1] 

Concrete wall overlay 

[21.4.5] 

Infill openings 

Torsionally irregular plans (highly perforated wall line) 

Steel braced frame [7.4.1] 

Concrete/masonry shear wall [21.4.8] 

Enhance existing collector [7.4.2] Fiber composite wall overlay [13.4.1] Concrete wall overlay [21.4.5] Infill openings 

Re-entrant corners Steel braced frame 

[7.4.1] Collector [7.4.2] Concrete/masonry shear wall [21.4.8] 

Enhance existing collector [7.4.2] Fiber composite wall overlay [13.4.1] Concrete wall overlay [21.4.5] Infill openings 

Configuration 

Incidental bracing 

Load Path Inadequate or missing 

wall-to-diaphragm tie for out-of-plane load – exterior and interior walls 

Wall-to-diaphragm anchorage [16.4.1] plus diaphragm cross-ties [22.2.3] 
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Table 16.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for PC1 Buil

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Red

Dem

Inadequate anchorage to diaphragms for inplane forces – exterior and interior walls 

Wall-to-diaphragm shear anchors [21.4.8] 

Beam or girder connection to tilt-up wall inadequate for wall out-of-plane loads 

Enhance beam or girder connection [16.4.2] 

Inadequate connection at base of tilt-up panel 

Wall-to-foundation connections [16.4.3] 

Load Path (continued) 

Inadequate collectors Add collector 

[7.4.2] 

Enhance existing collector [7.4.2] 

Wall inadequate for out-of-plane bending 

Wall strongback or pilaster [21.4.3] 

Component Detailing 

Inadequate detailing of narrow wall piers 

Steel braced frame [7.4.1] Concrete/masonry shear wall [21.4.7] 

Supplement component to provide adequate load path [13.4.1], [21.4.5] Add backup vertical supports where bearing might be lost [21.4.11] 

Inadequate in-plane strength and/or stiffness 

Steel braced frame [7.4.1] Concrete or masonry wall [21.4.8] 

Enhance existing diaphragm [22.2.1] Horizontal braced frame 

Diaphragms 

Inadequate chord capacity 

Add chord [22.2.2] Enhance existing chord 

 Excessive stresses at 

openings and irregularities 

Enhance diaphragm detailing 

 Excessive stresses at 

openings and irregularities 

Enhance diaphragm detailing 

Foundations See Chapter 23 [ ] Numbers in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 
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Figure 16.3-1: Plan of PC1 Building with Re-Entrant Corner at In-Set Panels 

deformation compatibility between the building sections under earthquake loading. Incompatible 

deformations are likely to significantly compromise shear transfer through the expansion joint. In 

addition, common details used for the expansion joint may not perform adequately under 

expected seismic forces. Details often involve significant eccentricities. The eccentricities may 

create forces in roof framing members that were not envisioned in the framing member design. 

The most direct approach to rehabilitation of this building type is the addition of vertical 

elements such that each building section is independently braced. Alternatively, the expansion 

joint connection can be improved such that it can reliably transfer anticipated earthquake forces 

while accommodating anticipated building movement. 

Another configuration concern is the occurrence of components (such as mezzanines) that act as 

incidental bracing, creating an unintended load path. For the PC1 building seismic resisting 

system to work as intended, the diaphragm must be able to deflect, and the walls must be able to 

deflect out-of-plane to follow the diaphragm. Ideally rehabilitation of incidental bracing would 

involve isolation from the building shell. Alternately, the incidental bracing could be analyzed as 

part of the structural bracing system. 
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Figure 16.3-2: Plan and Detail of Large PC1 Building Constructed in Three Sections 
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Load Path 

As previously mentioned, out-of-plane anchorages between the tilt-up walls and the diaphragm 

have been the primary source of damage and focus of rehabilitation in wood diaphragm PC1

buildings. Conceptually, the approach for both new construction and rehabilitation has been to 

create continuous diaphragm cross-ties between exterior walls on opposite sides of the building. 

Rehabilitation of wall out-of-plane anchorage is discussed in this chapter. Diaphragm cross-ties 

are discussed in Chapter 22. Wall anchorage for in-plane shear commonly uses different 

connectors than for out-of-plane loads. In-plane shear connection is discussed as part of the 

diaphragm chord discussion in Chapter 22. 

Girder gravity load connections to tilt-up walls provide diaphragm-to-wall anchorage and will 

therefore have to resist wall out-of-plane anchorage forces. These anchorage forces may or may 

not have been considered in initial connection design. Even if considered, the connection may 

not be adequate for currently required loads. Rehabilitation of these connections is often 

required. 

The addition of or enhancement of existing collectors may be required in order to transmit 

diaphragm forces to the resisting shear walls. This is particularly of concern when a limited 

number of solid panels are intended to carry a significant portion of the building shear. Although 

not as common, there is also significant concern when vertical offsets in the roof diaphragm 

result in incomplete chords or collectors. Any breaks or offsets in chords or collectors need to be 

carefully evaluated. In the 1993 Guam earthquake (EERI, 1995), a high bay portion of a forklift 

repair shop was braced off of lower bays on each side. Incomplete collectors from the high-bay 

diaphragm to low bay shear walls resulted in damage. 

The anchorage at the tilt-up panel at the wall base may also be deficient. In some older tilt-ups, 

no positive connections were made from the wall to the foundation or slab; friction was relied on 

for force transfer both in-plane and out-of plane. Most tilt-up panels in recently constructed 

buildings will have a base connection, either to the foundation or more likely to the adjacent 

slab-on-grade; however, it is possible for the connection to be inadequate. Rehabilitation most 

often involves the addition of new wall to slab connections. 

Component Detailing 

Component detailing deficiencies include inadequate out-of-plane wall capacity. This deficiency 

may occur due to increases in design seismic forces or inadequate consideration of panels with 

openings. It is seldom practical to address wall capacity by adding reinforcing and concrete 

thickness to individual wall sections, so addition of wall pilasters or strongbacks is common. 

Where pilasters are added to tilt-up walls, the pilasters stiffen the wall for out-of-plane forces, 

allowing two-way spanning of the wall and attracting high out-of-plane forces to the pilaster and 

pilaster-to-diaphragm connection. A pilaster-to-roof diaphragm anchorage must be provided to 

accommodate the concentrated wall out-of-plane force. 

Tilt-up panels with large openings generally have narrow wall panels on each side. These panels 

and the spandrel wall over the opening act as a concrete frame. In loading dock and storage 

facilities, entire walls can be made up of frames. Because the narrow wall piers do not meet the 

ACI 318 (ACI, 2005) definition of a column, many have been constructed with standard wall 
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detailing. The 1994 and later editions on the UBC (ICBO, 1994) defined and created reinforcing 

requirements for wall piers, which have been brought into the IBC (ICC, 2003a), but are not in 

ACI 318. Where the wall piers are required for resistance to gravity and lateral loads, 

rehabilitation may be required. Rehabilitation for lateral loads may involve the additional of new 

vertical elements, enhancing of existing elements, or filling in some of the openings. In addition, 

loss of vertical support for the roof adjacent to the panel may be of concern. Rehabilitation for 

vertical loads may be approached by either enhancing the wall element, or providing back-up 

vertical supports. 

Diaphragm Deficiencies 

Due to changes in building code requirements, it is very common for diaphragms in areas of high 

seismic hazard to have inadequate in-plane shear capacity. Regardless, the SEAONC Guidelines

indicates that diaphragm overstresses have rarely been associated with significant earthquake 

damage. In areas of moderate seismic hazard, this may or may not be a significant deficiency. 

Diaphragm strength and stiffness deficiencies are most often rehabilitated by enhancing the 

existing diaphragm. The addition of new vertical elements to reduce diaphragm span and 

therefore diaphragm shears is also an effective approach, but such an approach is not as 

commonly used. Diaphragm enhancement is addressed in Chapter 22. 

Many California tilt-up buildings have been constructed with solid-sawn roof purlins with sizes 

ranging from 4x12 to 4x16. These framing members may have calculated overstresses under 

existing dead and live loads due to reductions in allowable stresses from in-grade values 

introduced in the 1991 NDS (AF&PA, 1991). When this is the case, removal and replacement of 

roof sheathing is a preferred alternative to overlays in order to keep additional dead load a 

minimum. Diaphragm enhancement using staples may also provide required strength. 

Sometimes, lighter roofing finish materials can be installed where existing roofing is removed, to 

further reduce overstresses. 

Other diaphragm deficiencies include inadequate chord capacity and stress concentrations at 

large diaphragm openings and re-entrant corners. Rehabilitation at re-entrant corners primarily 

involves the provision of adequate chords and collectors. The same is true at large diaphragm 

openings. 

16.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

16.4.1 Enhance Wall-to-Diaphragm Anchorage 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses enhanced anchorage of walls into wood sheathed or steel 

deck diaphragms for out-of-plane loads. Cross-ties are a required continuation of the wall 

anchorage system. See Section 22.2.3. 

Wall anchors and cross-ties should be the highest priority for rehabilitation of wood diaphragm 

PC1 buildings. This section illustrates the basic rehabilitation requirements for wall anchorage of 
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wood and steel diaphragm buildings. Refer to the SEAONC Guidelines for more exhaustive 

treatment of detailing. 

The SEAONC Guidelines provides details of code requirements and observed damage over many 

years. Earthquake damage to wall anchorage has been observed not only in older buildings, but 

also in recently constructed buildings, where the type and installation of wall anchors was critical 

to their ability to perform. Just providing anchors is not enough; attention to detailing and field 

installation is critical. 

Description of the Rehabilitation Technique 

Figure 16.4.1-1A illustrates a roof plan for a PC1 building with a wood diaphragm. Figures 

16.4.4-1B through 16.4.4-1D, located as shown on the roof plan, illustrate common rehabilitation 

measures for wall-to-diaphragm anchorage. Also illustrated in Figure 16.4.1-1A are 

subdiaphragms used as part of the diaphragm cross-tie system, described in detail in Chapter 22. 

While subdiaphragms are always used in wood structural panel diaphragms, they are only 

occasionally used in steel deck diaphragms. 

Figures 16.4.1-1B and 16.4.1-1C illustrate anchorage from the west wall to a subdiaphragm 

extending between Lines A and B. Similarly, the east wall is anchored to a subdiaphragm 

between Lines G and H. The primary objective is to create an adequate load path for out-of-plane 

loads acting both into the diaphragm (compression) and away from the diaphragm (tension). This 

type of anchorage is generally provided every four to eight feet on center. The load path into the 

diaphragm includes both wall to framing member anchorage and sheathing nailing to transfer 

loads from the framing member to the sheathing. Figure 16.4.1-1B relies on existing sheathing 

nailing, while Figure 16.4.1-1C adds sheathing nailing. 

Where framing member ends are tight against the wood ledger and the wood ledger is tight 

against the exterior wall, the compression load path can be carried by the framing members. 

Experience has shown that gaps often occur between member ends and the ledger. The 

movement required to close these gaps has been enough to damage the roof sheathing and 

damage devices used for tension anchorage. 

Tie-down devices can provide both the tension and compression load path. In order to do this, 

the device must be rated for both tension and compression loading by the manufacturer, and it 

must be installed in accordance with the manufacturer’s instructions for compression load. This 

generally involves limiting the unsupported length of the tie-down rod and providing additional 

nuts and washers at the tie-down seat. Very few tie-down devices are currently available that are 

rated for both tension and compression loads. Building codes such as the now require that forces 

and stresses induced by eccentricities in the connection be addressed, and the SEAONC 

Guidelines encourages use of tie-downs placed symmetrically on the purlin or sub-purlin in order 

to minimize eccentric beam loading and encourages use of stiff tie-downs in order to minimize 

deformation demand and possible diaphragm damage resulting from deformation. For girders, 

symmetrical tie-downs are recommended but not as critical as with smaller framing members. 
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Figure 16.4.1-1A: Roof Plan with Wall Out-of-Plane Anchorage 

for Flexible Wood Diaphragm 

Figures 16.4.1-1A and 16.4.1-1C do not show wall anchors attached to single existing 2x 

subpurlins. It is recommended that anchorage be to a 3x or wider member, or multiple 2x’s as 

shown in Figure 16.4.1-1B. While it may be possible for a single 2x4 or 2x6 to be shown 

adequate in calculation, adequate installation and performance are extremely difficult to achieve. 

The wall anchorage system needs to extend across the width of the subdiaphragm. In Figure 

16.4.1-1C, this involves providing extra pairs of tie-down devices between subpurlins across the 

first purlin. Again, both tension and compression load paths are needed. The required 

subdiaphragm depth is determined from the number of nails required to transfer subdiaphragm 

forces into the main diaphragm. Depending on subdiaphragm requirements, additional tie-down 

pairs could be required across more purlins. Subdiaphragm requirements are discussed in 

Chapter 22. 

Details 16.4.1-1B and 16.4.1-1C show work from top of the diaphragm. See Chapter 22 for 

alternates for working from below. Location of access needs to be decided early on in the design 

process and will drive both calculations and detailing of the rehabilitation work. 
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Figure 16.4.1-1B: Wall Out-of-Plane Anchorage for Flexible 

Wood Diaphragm at Subpurlins – Roofing Not Removed 
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Figure 16.4.1-1C: Wall Out-of-Plane Anchorage for Flexible 

Wood Diaphragm at Subpurlins – Roofing Removed 
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Figure 16.4.1-1D illustrates anchorage of the north and south walls to a purlin. In this case, the 

purlin is long enough to extend across the subdiaphragm width (extending between Lines 1-2 and 

3-4), so additional pairs of tie-downs are not needed. As is previous details, both tension and 

compression load paths must be maintained. 

Figure 16.4.1-1D: Wall Out-of-Plane Anchorage 

for Flexible Wood Diaphragm at Purlins 

Figure 16.4.1-2A illustrates a similar roof plan for a PC1 building with a steel deck diaphragm. 

It is important to note in this figure that subdiaphragms (as shown in Figure 16.4.1-1A) are not 

used. Instead, the steel deck provides a continuous cross-tie in the east-west direction, while in 

the north-south direction open web joists provide direct cross-ties across the entire diaphragm 

width at each wall anchor location. This is the primary approach used in new steel deck 

diaphragm construction. Subdiaphragm concepts can be applied to steel deck construction, but 

are not common. 

Figures 16.4.1-2B and 16.4.1-2C provide wall to diaphragm anchorage details. In Figure 16.4.1-

2B, wall anchorage forces are transmitted to the steel deck. The deck section, deck edge 

fastening, and deck end splices need to be checked for wall anchorage tension and compression 

forces. Justification of the capacity may be by calculation or testing. The balance of the load path 

also needs to be checked and enhanced as required. In the detail shown, supplemental adhesive 
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Figure 16.4.1-2A: Roof Plan with Wall Out-of-Plane Anchorage 

for Flexible Steel Diaphragm 

anchors to the wall and a second ledger angle are provided. Prying action in the steel ledger 

angle needs to be considered in determining wall anchor forces. 

Figure 16.4.1-2C shows wall anchorage to a steel open web joist. The wall anchorage connection 

may be through the joist seat in new construction; however, a supplemental anchor to the joist is 

likely needed in rehabilitation. The joist must be checked for wall anchorage forces and any 

applicable eccentricities. Details in Chapter 22 address joist to joist connections to complete the 

cross-tie. 

Compression forces can be carried in tie-down devices, if rated for compression by the 

manufacturer. Unsupported tie-down rod lengths must be kept short, and additional nuts and 

washers are needed to transfer compression. Again, wall anchorage loads need to be transferred 

into the decking. In this case existing welding or screwing of the decking is relied upon. If this 

fastening is not adequate, the roofing will need to be removed to allow additional fastening. 
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Figure 16.4.1-2B: Wall Out-of-Plane Anchorage for Flexible Steel Diaphragm 

– to Decking for Load Parallel to Flutes 

Design and Detailing Considerations 

Research basis: No research applicable to the performance or adequacy of enhanced anchorage 

methods has been identified; however, the demands created in flexible diaphragms have been 

studied by Fonseca, Wood and Hawkins (1996); Hamburger and McCormick (1994); Ghosh and 

Dowty (2000); and Freeman, Searer, and Gilmartin (2002). 

As discussed in Section 16.1, even wall anchorages constructed or rehabilitated in the 1980s and 

early 1990s were observed to have been damaged in the 1994 Northridge earthquake. The reader 

is referred to the extensive discussion in the SEAONC Guidelines for design and detailing 

considerations and lessons learned. 

Anchor type and installation: A variety of proprietary anchors are available for anchorage to 

existing concrete walls. Both manufacturer literature and ICC Evaluation Service reports should 

be consulted for information on conditions of use, allowable loads, and installation and 

inspection requirements. It is important to make sure that the anchor type is appropriate for the 

material to which it will be connected and is approved for seismic loads. The diameter of drilled 

holes is specified in installation requirements for each anchor type; variation from this size often 
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Figure 16.4.1-2C: Steel Open Web Joist Anchorage to Exterior Wall 

leads to inadequate anchor capacity. Most manufacturers have caulking gun-like devices that 

make field placement of epoxy fairly simple and automatically mix two-part adhesives. 

Generally, these types of adhesives provide more than adequate strength, and there is no need to 

use more complicated high-strength adhesive types. The cleaning of holes prior to placing 

adhesive anchors is paramount for anchor capacity. When not well cleaned, the anchors can pull 

out at a small fraction of the design load. It is common to pull-test a portion of the adhesive 

anchors to verify adequate installation. The pull test load is usually in the range of one to two 

times the tabulated allowable stress design tension load. The bridge used for testing generally 

makes a concrete cone pull-out failure unlikely. The test load should not be near yield load for 

bolts or adhesive pull-out failure loads. 

Anchors added in rehabilitation will often have to work in combination with existing cast-in 

anchors. In order to allow load sharing, anchorage of similar stiffness is desirable. This is often 

best achieved with adhesive anchors. In addition, anchorage provisions for new buildings have 

moved towards having the attachment to concrete be capable of developing the yield capacity of 
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the steel anchor in order to promote ductile connection behavior. Again, this is best achieved 

with adhesive anchors. Anchor types other than adhesives should be carefully evaluated. 

Cost, Disruption and Construction Considerations 

When rehabilitation work is undertaken on the roof diaphragm of a PC1 building, it is important 

that the cost and the preferred location for work (from the underside or top of roof) take into 

account the combination of work, rather than considering one piece at a time. If several 

diaphragm measures will be undertaken, it will quickly become cost-effective to remove the roof 

and allow work from the top. This is particularly true if a steel deck requires several 

rehabilitation measures. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique, other than the use of 

proprietary connectors and adhesive anchors as part of the assemblage. 

16.4.2 Enhance Beam and Girder Connection to Supporting Elements 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses enhancement of girder gravity load connections to tilt-up 

walls. While primarily intended to carry gravity loads, these connectors should also be adequate 

to resist wall out-of-plane loads. 

Description of the Rehabilitation Technique 

Tilt-up walls may have pilasters supporting girder loads. In older tilt-up buildings, these are often 

cast-in-place pilasters, while in newer buildings pilasters are cast as part of the tilt-up panel. The 

pilaster acts as a wall stiffener, allowing the wall to span both horizontally and vertically under 

out-of-plane earthquake loading. This attracts more load to the pilaster, and the top of pilaster 

reaction for out-of-plane loading will be higher than at a typical wall anchor. The gravity load 

connection generally also serves as the girder-to-pilaster connection for out-of-plane loads. Two 

deficiencies are common with this connection: 1) inadequate confinement around anchor bolts 

embedded in the pilaster top and 2) inadequate connection to the girder. Tension loads on the 

connector have led to splitting of the column top, pulling away the wedge of concrete in front of 

the anchor bolts. In recent codes, the placement of three closely spaced ties at anchor bolts has 

been required to reinforce across the anticipated concrete crack. Where added ties have not been 

provided, a collar around the pilaster top can provide external reinforcement (Figure 16.4.2-1). 

The collar should be relatively stiff to minimize splitting of the concrete before load is taken up. 

The second issue is inadequate connection to the wood girder, including bolt capacity and 

placement. Where the girder seat connection to concrete is adequate or can be enhanced, 

inadequate bolt capacity can be mitigated with addition of bolt tabs (Figure 16.4.2-2). 

Alternately, it is possible to use new wall anchors from the girder to the wall, bypassing the 

girder seat (Figure 16.4.2-3). Again, this connection must be stiff. 
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Figure 16.4.2-1: Enhanced Girder Connection – Collar at Pilaster 

Adapted From SEAONC (2001) 

Girders that are supported directly on a flat wall panel using a steel U-bracket bolted or welded 

to the panel (Figure 16.4.2-3) will also attract wall out-of-plane forces. As is true with wall 

pilasters, a girder and U-bracket are likely to provide a stiffer load path for wall out-of-plane 

loads than adjacent anchors. For this reason, use of a wall anchorage force greater than used for 

adjacent anchors is encouraged. The girder connection should have the ability to resist wall 

anchorage loads in combination with gravity loads. Anchorage of the bracket to the panel will 

often be adequate for both gravity and lateral loads; however, the bracket attachment to a wood 

girder will often not have the quantity or placement of bolts required for tension loads. Addition 

of steel tabs and bolts will add capacity and place bolts where end distances are adequate for 

tension loads. Where the steel connection to the concrete is not adequate, the out-of-plane 

anchor might bypass the existing connection and connect the girder directly to the wall. Figure 

16.4.2-2 shows two approaches, one with a tie-down on each side of the girder and a second with 

a tie-down on the girder bottom. The out-of-plane wall anchor should be as stiff as possible to 

minimize damage to the gravity connection. 

Design and Detailing Considerations 

Research basis: No research applicable to this rehabilitation measure has been identified. 

See Section 16.4.1 wall anchorage and the SEAONC Guidelines for additional detailed 

discussion. 
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Figure 16.4.2-2: Enhanced Girder Connection at U-hanger 

Adapted From SEAONC (2001) 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique, other than the use of 

proprietary connectors as part of the assemblage. 

16.4.3 Enhance Anchorage at Base of Tilt-Up Panels 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses the addition or enhancement of connections between the 

tilt-up wall panel and foundation or adjacent slab-on-grade to resist in-plane shear and 

overturning forces and out-of-plane wall anchorage forces. 

Description of the Rehabilitation Technique 

Rehabilitation of base connections in PC1 buildings for in-plane and out-of-plane shear loads is 

most commonly accomplished by addition of steel angles and adhesive anchors between the wall 

panel and adjacent slab-on-grade. This is illustrated in Figure 16.4.3-1. In some instances, the 

slab-on-grade may not have been thickened adjacent to the tilt-up panel. When this is the case, it 

may be necessary to remove and recast a thicker pour strip in order to get adequate anchorage. 

The connection shown would flex if the wall were to uplift. Where uplift connection capacity is 

required, a direct tension connection of the wall to the footing below is recommended. 
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Figure 16.4.2-3: Enhanced Girder Connection at Pilaster 

Adapted From SEAONC (2001) 

Variations in base conditions include 1) older PC1 buildings that may not have any doweling 

because friction was relied on to resist forces at the base of the panel and 2) welded connections 

between cast-in embeds in the wall panel and slab, similar to PC2 wall panel connections. 

Design and Detailing Considerations 

Research basis: No research applicable to this rehabilitation technique has been identified. 

The SEAONC Guidelines provide discussion of a variety of possible existing conditions, changes 

in code requirements, and implications for retrofit. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique, other than the use of 

proprietary adhesive anchors as part of the assemblage. 
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Figure 16.4.3-1: Enhancement of Tilt-up Panel Base Connection 

16.5 References 

Extensive reference and bibliography listings are included in SEAONC (2001). The following 

only includes references specifically cited in this chapter. 

ACI, 2005, Building Code Requirements for Structural Concrete, American Concrete Institute, 

Farmington Hills, MI. 

ACI-SEASC Task Group on Slender Walls, 1982, Report of the Task Group on Slender Walls

(Green Book), American Concrete Institute, Southern California Chapter and the Structural 

Engineers Association of Southern California, Los Angeles, CA. 

AF&PA, 1991, ANSI/NFoPA NDS-1991 National Design Specification for Wood Construction, 

American Forest and Paper Association, Washington, D.C. 

City of Los Angeles, 2002, “Chapter 91,” City of Los Angeles Building Code, Los Angeles, CA. 

EERI, 1996, Supplement C to Volume 11, Northridge Earthquake of January 17, 1994, 

Reconnaissance Report, Earthquake Spectra, Volume 2, EERI, Oakland, CA. 
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EERI, April 1995, Guam Earthquake of August 8, 1993 Reconnaissance Report, Earthquake 

Spectra, Supplement B to Volume 11, Earthquake Engineering Research Institute, Oakland, Ca. 

FEMA, 2004, NEHRP Recommended Provisions for Seismic Provisions for New Buildings and 

Other Structures, FEMA 450, Federal Emergency Management Agency, Washington, D.C. 

Freeman, S., S. Searer and U. Gilmartin, 2002, “Proposed Seismic Design Provisions For Rigid 

Shear Wall / Flexible Diaphragm Structures,” Proceedings of the Seventh U.S. National 

Conference on Earthquake Engineering, EERI, Oakland, CA. 

Fonseca, F., S. Wood and N. Hawkins, 1996, “Measured Response of Roof Diaphragms and 

Wall Panels in Tilt-Up Systems Subject to Cyclic Loading,” Earthquake Spectra, Volume 12, 

Number 4, Earthquake engineering Research Institute, Oakland, CA. 

Ghosh, S.K. and S. Dowty, 2000, Anchorage of Concrete or Masonry Walls to Diaphragms 

Providing Lateral Support, Draft 2000, Not Published. 

Hamburger, R.O. and D. McCormick, 1994, “Implications of the January 17, 1994 Northridge 

Earthquake on Tilt-up Wall and Masonry Wall Buildings with Wood Roofs,” Proceedings of the 

1994 Convention of the Structural Engineers Association of California, Sacramento, CA. 

ICBO, 1994, Uniform Building Code, International Conference of Building Officials, Whittier, 

California. 

ICBO, 2001, Guidelines for Seismic Retrofit of Existing Buildings (GSREB), International 

Conference of Building Officials, Whittier, CA. 

ICC, 2003a, International Building Code, International Code Council, Country Club Hills, IL. 

ICC, 2003b, International Existing Building Code, International Code Council, Country Club 

Hills, IL. 

SEAONC (Structural Engineers Association of Northern California), 2001, Guidelines for 

Seismic Evaluation and Rehabilitation of Tilt-Up Buildings and Other Rigid Wall/Flexible 

Diaphragm Structures, SEAONC, San Francisco, California. 

SEAOSC (Structural Engineers Association of Southern California), 1979, Recommended Tilt-up 

Wall Design (Yellow Book), SEAOSC, Los Angeles, CA. 

 

Tilt-up Concrete Association, 2004, Tilt-up Construction and Engineering Manual, Sixth 

Edition, Tilt-up Concrete Association, Mount Vernon, IA. 
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Chapter 17 - Building Type PC2: Precast Concrete Frames 

with Shear Walls 

17.1 Description of the Model Building Type 

Buildings designated as PC2 include wide ranging combinations of precast and cast-in-place 

concrete elements. Precast members may be limited to a floor system of hollow core or T-beam 

construction, or may include all elements of the gravity and lateral load systems. For this chapter, 

Building Type PC2 includes concrete wall or frame buildings in which any of the horizontal or 

vertical elements of the lateral load system are of precast concrete, except for flexible diaphragm 

buildings which are addressed as Building Type PC1 in Chapter 16. 

Figure 17.1-1: Building Type PC2: Precast Concrete Gravity Frames with Shear Walls 

Extensive use of hollow core floor systems in buildings with concrete and masonry walls in 

southern regions of the United States makes this the single largest group of PC2 buildings. 

Parking garages (used exclusively for parking rather than mixed use) represent the next largest 

group of PC2 buildings and a substantial portion of the current PC2 building inventory in the 

U.S. The PC2 building type has also been used for a variety of other occupancy types in the U.S. 

and internationally, including mid-rise office, hotel, and residential buildings, low-rise 

residential, commercial, and prison buildings. 
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Over the past decade, significant effort has been devoted to development and testing of precast 

ductile moment frame systems through the PRESSS (Priestley et al., 1999). These systems are 

not addressed in this chapter due to their very recent development and state-of-the-art detailing. 

Gravity-Carrying Load Systems 

Special attention is needed to PC2 buildings in which concrete frames (beams, girders, and 

columns or moment frames) resist gravity load, or a combination of gravity and seismic load. 

Very important to the performance of all concrete buildings with frames, including PC2

buildings, is the lack of ductile detailing in concrete columns not designated as part of the 

seismic force-resisting system. These columns in many instances do not have confining steel 

adequate to accommodate the drift imposed by the seismic force-resisting system and as a result 

fail through longitudinal bar buckling and concrete crushing. Requirements for estimation of 

building drift have changed over time, and understanding of potential building deflection has 

improved with each observation of earthquake performance. As a result, it is important to revisit 

the ability of non-ductile columns to accommodate estimated drifts, even if they were checked 

when initially designed. In some precast buildings, the division of initial design responsibility 

between one engineering firm for the gravity load system and a second firm for the seismic 

force-resisting system may have contributed to inability to accommodate estimated building 

deflections. In earthquake performance to date, diaphragm deflections have been a large 

contributor to deflection of non-ductile gravity systems. Vertical elements, and most particularly 

moment frames, could also contribute significantly to gravity system deflection. 

Following the 1994 Northridge earthquake, column detailing requirements and methods of 

estimating building deflection for purposes of gravity frame design were modified in codes and 

standards. Concrete buildings constructed recently in areas of high seismic hazard should 

perform significantly better than those designed under older provisions. Research continues to 

develop a better understanding of sources of diaphragm deflection. 

When considering the ability of gravity load-carrying systems to accommodate building 

deflection, a related issue of importance is proper accounting for column stiffness and restraint in 

analysis. This, again, is a concern for all concrete buildings including PC2 buildings. Short 

columns will attract higher forces due to increased stiffness and have been seen to fail as a result. 

Short columns can be created accidentally due to inadequate separation of the column from 

nonstructural components such as guardrails. In addition, systematic problems with short 

columns can occur at parking garage ramps. Analysis models need to pay special attention to 

these and other sources of shortened columns or columns with increased end fixity. 

PC2 buildings with gravity and lateral loads supported exclusively by structural walls do not 

have the same issues of deflection of non-ductile columns. Connections tend to be the primary 

issue of importance to these systems for both gravity and seismic load systems. 

Shear Walls and Frames 

Building Type PC2 may have a lateral force-resisting system of concrete shear walls or moment 

frames, cast-in-place or precast. In PC2 buildings, critical behavior of shear walls is generally 

governed by connections including: diaphragm to shear wall, shear wall above to shear wall or 

foundation below, and interconnection of shear walls within a story. In PC2 buildings with 
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precast frames, field connections within the frame are critical to performance, as is ductile 

detailing. Connection practice has varied widely over time and by geographic region. 

Floor and Roof Diaphragms 

In California, precast floor T-beams or hollow core planks are covered by a cast-in-place topping 

slab, reinforced to provide diaphragm action. These toppings need to be clearly differentiated 

from topping slabs plant-applied to individual precast members, which do not serve the same 

function of structurally interconnecting adjacent members. Welded connections between 

embedded inserts or plates may also be used to aid in alignment of members during erection, but 

are generally not relied on for diaphragm action. Reinforcing bars are often added in the cast-inplace topping slab to act as diaphragm chords and collectors, and welded wire fabric is used to 

provide shear reinforcement. 

In other areas of the United States, common methods of joining floor sections include use of 

grouted hollow core joints (grout placed in the joint between two adjacent panels, relying on 

adhesion and/or friction for shear transfer) or welded insert plates. Cast-in-place topping slabs 

are not commonly used. In some areas outside the U.S., hollow core planks are installed with no 

connection or grouting between adjacent planks. 

As per the discussion of gravity load systems, deflection of the diaphragm system has been seen 

as a significant contributor to building deflection is past earthquakes. Discussion of diaphragm 

behavior and rehabilitation can be found in this chapter, Chapter 20 for masonry wall buildings, 

and Chapter 22 for detailed discussion of diaphragm rehabilitation. 

Parking Structure Issues 

Parking structure PC2 buildings have unique characteristics that deserve specific discussion, 

some applicable to parking structures regardless of structural system and others specific to 

precast construction. These issues include the following. 

Many parking structures have large plan areas, and considerations of security and 

restraint against temperature, creep, and shrinkage movement lead to concentration of the 

shear walls at the building perimeter near the center of each side (Figure 17.2-1). This 

configuration leads to long diaphragm spans with significant shear, moment, and 

collector demands. With these high demands, it is possible for the diaphragm, rather than 

the vertical elements to control building dynamic behavior. This is of concern in all 

systems, but particular in precast systems due to the lower level of inherent diaphragm 

continuity. 

Compared to other building uses, parking garages have greatly minimized finish and 

cladding systems, resulting in low levels of nonstructural damping and energy 

dissipation. 

Ramps in parking structures may act as tension and compression struts between floors, 

resulting in demands not anticipated during design. This behavior can be avoided by 

inclusion of seismic joints at one end of each ramp; however, seismic joint detailing is 

difficult to accommodate in precast concrete construction, making use of a fixed 

connection more likely. Unless the effect of the ramp is specifically considered in 

analysis, force transfer through the ramp can result in seismic forces bypassing the 
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intended resisting system and significant redistribution of forces in the diaphragms and 

vertical elements. An analytical study of ramp effects discussed in Lyons, Bligh, 

Purlinton, and Beaudoin (2003) suggests that while the effect of ramps is significant in 

moment frame buildings, it is less significant and can often be managed in design of 

shear wall buildings. The effect of ramps should be considered in evaluation and 

rehabilitation of parking structure buildings. 

Figure 17.2-1: Plan of Common Parking Structure Configuration 

17.2 Seismic Response Characteristics 

PC2 buildings occur with a wide range of vertical element types. In most cases, the vertical 

element type will dictate the building seismic response: shear wall buildings will have short 

period response, while frame buildings will have a longer period. In PC2 buildings, stiff 

diaphragm behavior will generally be intended. Parking structure PC2 buildings with long 

diaphragm spans, however, have been observed to have inelastic behavior concentrated in the 

diaphragms rather than vertical shear wall or frame elements. To date, this has been brittle 

behavior resulting in premature diaphragm failure; however, with development of proper 

detailing it may be possible to achieve stable long-period diaphragm behavior. 
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17.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

Construction of PC2 buildings in areas of high seismic hazard in the U.S. has been of limited 

quantity and recent compared to most other building types, resulting in limited opportunities to 

observe earthquake performance; the City of Los Angeles and SEAOSC (1994) and EERI (1996) 

reported on performance of one group of PC2 buildings following the 1994 Northridge 

earthquake. Out of an estimated 100 parking garages (precast and cast-in-place) in heavily 

shaken areas in the Northridge earthquake, eight had partial collapses, and an additional 20 had 

at least 25% damage (City of Los Angeles and SEAOSC, 1994). The task force looked at 

approximately 30 structures; of 26 structures with damage, approximately half contained some 

precast elements (Mooradian, 2005). 

Within limited experience to date, life-safety performance of other PC2 buildings in the U.S. has 

been good; however, performance in other countries has identified concerns that could be 

applicable to U.S. construction. See below for general discussion and Table 17.3-1 for a detailed 

compilation of common seismic deficiencies and rehabilitation techniques for the PC2 building 

type. 

Global Strength and Stiffness 

Insufficient in-plane shear wall strength and stiffness are possible seismic deficiencies in PC2

buildings and particularly in parking garages where shear wall length is generally limited. 

Rehabilitation to address inadequate shear wall strength and stiffness can include addition of new 

vertical elements or strengthening of existing elements, as summarized in Table 17.3-1. Addition 

of and enhancement to elements in PC2 buildings is very similar to other concrete building 

types; however, several additional cautions are in order. First, the configuration of existing 

precast members, including cast-in voids and prestressing tendons must be carefully studied to 

allow connection of new elements to existing construction. Second, precast and post-tensioned 

systems are configured to minimize damaging effects of movement due to temperature variation, 

shrinkage and creep; these effects should be considered in the addition of new vertical elements. 

Insufficient in-plane moment frame strength is a possible seismic deficiency in PC2 buildings 

and particularly of concern where the frame might not have been initially designed for seismic 

loads. Where strength is a concern, it is likely that stiffness, connections, and ductile detailing 

will also be inadequate and that major addition or enhancement of vertical elements is required. 

Configuration 

Torsional irregularities can lead to possible seismic deficiencies in PC2 buildings, as in any 

other building type, increasing deformation demand in local portions of the structure. One of the 

parking garages investigated by the City of Los Angeles and SEAOSC Task Force (City of Los 

Angeles and SEAOSC, 1994) had shear walls on three sides and an open front of the fourth side. 

The report speculates that excessive deflection at the open-front allowed girders to move 

sideways off of supporting columns, resulting in total collapse. The torsionally irregular building 

configuration appears to have contributed to collapse, along with inadequate diaphragm stiffness 

and component connections. Torsional deficiencies are most directly addressed through the 

addition of new vertical elements, as indicated in Table 17.3-1. Design to accommodate the 

concentrations of force and deformation demand may be an alternative. 
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Table 17.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for PC2 Buil

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce 

Demand 

Global Strength 

Insufficient in-plane strength of shear walls or frames 

Steel braced frame [7.4.1] Concrete/masonry shear wall [12.4.2] 

Fiber composite wall overlay [13.4.1] Concrete wall overlay [21.4.5] Infill openings 

Global Stiffness 

Inadequate stiffness of shear walls or frames 

Steel braced frame [7.4.1] Concrete/masonry shear wall [12.4.2] 

Fiber composite wall overlay [13.4.1] Concrete wall overlay [21.4.5] Infill wall openings 

Torsional irregularity 

Steel braced frame [7.4.1] Concrete/masonry shear wall [12.4.2] 

Incompatible deformation of building sections 

Configuration 

Distance between shear walls too large 

Steel braced frame [7.4.1] Concrete/masonry shear wall [12.4.2] 

Load Path Inadequate force 

transfer, diaphragm to shear wall, shear wall above to shear wall below, shear wall to foundation 

Enhance anchorage 

[17.4.2] 
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Table 17.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for PC2 Buil

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements 

Reduce 

Demand 

Load Path (continued) 

Inadequate connection of beam or girders to supporting elements 

Supplemental vertical supports [21.4.11] 

Enhance anchorage 

 Inadequate 

collectors 

Add collector [17.4.1] 

Enhance existing collector [17.4.1] 

Gravity columns inadequate to accommodate drift 

Reduce building drift to level acceptable for gravity elements – see global stiffness 

Enhance column ductility with jacketing [12.4.4] 

Inadequate wall strength 

See Global Strength See Global Strength 

Component Detailing 

Inadequate frame connection detailing 

Steel braced frame [7.4.1] Concrete/masonry shear wall [12.4.2] 

Enhance existing frame connections 

Inadequate strength and/or stiffness 

Steel braced frame [7.4.1] Concrete/masonry shear wall [12.4.2] 

Enhance shear transfer within diaphragm [22.2.11] 

Inadequate shear transfer to walls 

See load path 

Inadequate chord capacity 

Add chord [17.4.1] Enhance existing 

chord [17.4.1] 

Excessive stresses at openings and irregularities 

Enhance diaphragm detailing 

Diaphragms 

Re-entrant corners Enhance diaphragm 

detailing 

Foundations See Chapter 23 [ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation measures. 
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One of the observations from the Northridge earthquake was that performance of parking 

garages decreased as the distance between supporting shear walls increased. Long diaphragm 

spans resulted in high shear and flexure demands on the diaphragms, causing yielding and 

fracture of diaphragm reinforcing (City of Los Angeles and SEAOSC, 1994; and Wood, Stanton 

& Hawkins, 2000). This can be considered either a configuration deficiency or a diaphragm 

deficiency. The addition of vertical elements will contribute significantly to reduction of demand 

and deflection. See global strength and stiffness discussion. Diaphragm shear capacity, chords, 

and collectors can also be enhanced, as discussed in the diaphragm deficiency section. 

Where PC2 buildings are comprised of several sections separated by seismic joints, movement 

can be incompatible and separation inadequate. 

Load Path 

Deficiencies in the load path connections between diaphragms and vertical elements, story to 

story and wall to foundation are of significant concern in every type of PC2 building. 

Connection and load path detailing in the existing PC2 building stock is thought to range from 

systems with no positive connections at all, to potentially brittle welded connections, to recently 

developed connections and systems that may allow better performance than cast-in-place 

buildings. Good performance was reported for limited examples of low-rise PC2 shear wall 

buildings in the Northridge earthquake (Iverson and Hawkins, 1994) and the Kobe earthquake 

(Ghosh, 1995). In the Guam earthquake (EERI, 1995) damage to the shear wall to foundation 

connections in a mid-rise hotel building caused extensive local spalling, apparently related to 

eccentricities within the connections. In the Armenia earthquake (EERI, 1989) low to mid-rise 

large panel precast buildings performed well. The good performance was attributed to floor 

panels that spanned to bearing walls on all four sides, and to the redundancy of these systems. In 

contrast, the connections between precast column members in frame buildings were very 

vulnerable, likely due to eccentricities introduced in site modifications to longitudinal bar splices 

and poor column confinement. Overall load path connections are important to the performance of 

PC2 buildings, and attention to detail and eccentricities is important. Rehabilitation of load path 

connections in shear wall type buildings will generally involve external mechanical connections. 

While not as obvious an issue in PC2 buildings as in PC1, as part of load path considerations 

concrete wall panels must be adequately connected to resist out-of-plane forces. The connections 

that are transferring diaphragm forces to the shear walls are generally also used to resist out-ofplane forces, so rehabilitation of these connections should consider both demands. 

The connection between girders and supporting columns and other similar connections may 

require rehabilitation in order to provide a continuous load path. The movement of a girder off of 

the supporting column due to building deflection is the most obvious concern. It has additionally 

been postulated that, in the Northridge earthquake, some gravity members may have been pulled 

free of supports due to high vertical accelerations or vibrations. At one time, it was common to 

have fairly heavy connections between girders and columns at the point of bearing. Connection 

design would likely have controlled by calculated forces, and welded connection behavior could 

be brittle. It is now more common to minimize or eliminate connection between the column and 

girder at the bearing point and resulting restraint of support movement and to rely instead on the 

column and girder each being doweled into the diaphragm system. This approach minimizes 
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unintended restraint and resulting damage. As in new construction, rehabilitation of this 

connection would best be accomplished by connection of each member into the diaphragm 

system. This approach is best suited to the intended behavior of the building system. Alternate 

approaches could include use of restraint cables, as is common in bridge rehabilitation or use of 

secondary vertical supports, as is common in rehabilitation of unreinforced masonry buildings 

(Section 21.4.10). There is no broad consensus on the contribution of vertical accelerations or 

vibrations to damage of PC2 buildings in the Northridge earthquake. In locations where vertical 

thrust is a concern, rehabilitation measures could specifically take vertical demands into 

consideration. 

In some precast systems, design of connections between elements has been based on the concept 

that the precast system emulates performance of an equivalent cast-in-place system. In frame 

members, the splicing of reinforcing steel is key to this performance. Over the years, the 

understanding of demands on splices and adequacy of splice technologies under earthquake 

loading has changed, as well as understanding of desirable locations of splices and controlling 

behavior modes within concrete frame systems, both precast and cast-in-place. Rehabilitation of 

connections between members in precast frames is difficult, and addition of vertical elements to 

limit frame drift may be a more practical solution. 

Component Detailing 

The inability of columns in gravity load systems to accommodate building drift has been pointed 

out as a significant deficiency in earlier discussion. Rehabilitation approaches include either 

adding additional vertical elements to reduce drift or enhancing columns with fiber-reinforced 

polymer wraps or similar systems to allow ductile behavior. These rehabilitation measures are 

discussed further in Chapter 12. The City of Los Angeles/SEAOSC report (1994) identified 

column wrapping as the only rehabilitation measure for parking garages that could be 

recommended as both practical and economical. Rehabilitation for the related issue of short 

columns can sometimes be as simple as creating an adequate joint between the column and the 

incidental restraint. Where this is not possible, reducing drift or enhancing column ductility are 

recommended rehabilitation approaches. 

Diaphragm Deficiencies 

Insufficient in-plane strength and stiffness of diaphragms are of significant concern over a range 

of precast systems. 

In parking garages with long diaphragm spans, insufficient shear strength has been identified as a 

likely contributor to poor performance of PC2 parking garages in the 1994 Northridge 

earthquake (Wood, Stanton & Hawkins, 2000), where pre-earthquake cracking of cast-in-place 

topping slabs occurred along the joists between T-beams. Analytical studies identified two 

strength-related issues that had not been considered previously. First, the pre-earthquake 

cracking of the topping slab along T-beam joints meant that the topping slab concrete was not 

contributing shear strength, leaving the reinforcing behavior acting in a shear-friction mode. 

Second, the limited strain capacity of the welded wire fabric reinforcing was being exceeded in 

commonly observed pre-earthquake crack widths, leaving the reinforcing vulnerable to brittle 

fracture during earthquake loading. Further, inadequate performance of chord and collector 

reinforcing in topping slabs has also been identified as a deficiency contributing to damage. In 
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the Northridge earthquake, reinforcing bars serving as chords and collectors in cast-in-place 

topping slabs appear to have yielded and subsequently buckled (City of Los Angeles/SEAOSC, 

1994). This behavior is a combination of strength and stiffness issues. In order to limit 

diaphragm deflection, maintain the integrity of the chord and collector members, and have the 

vertical elements control building dynamic behavior, it would be desirable to not have the chord 

and collector reinforcing yield. Since the Northridge earthquake, ACI 318 (ACI, 2005) and the 

building codes have taken initial steps towards addressing observed problems. Nakaki (1998) 

observed that the prescriptive steps taken by ACI 318 did not always result in improved 

behavior. Nakaki proposed simplified approaches to estimation of force and deformation demand 

for the purpose of ensuring elastic diaphragm design. This work has been incorporated into an 

appendix chapter of the NEHRP Provisions (FEMA, 2004) for untopped diaphragms.

In diaphragms without cast-in-place topping slabs, connections between adjacent planks or Tbeams often use embedded steel plates with field-welded connections, or grout connections. 

Questions arise as to the pre-earthquake adequacy of these connections. Welded connections are 

often used to correct differences in camber between adjacent members during initial erection and 

are often stressed by moving vehicle point loads and shrinkage and creep movement of the 

building. Observations of connections suggest that reduced capacity prior to earthquake loading 

may be common. This combines with a changing understanding of earthquake demands on the 

connection and the interaction of shear demands and deformation due to flexural or tension 

loading. The complete lack of connection between hollow core floor planks within diaphragms 

appears to have been a primary contributor to collapse of nine-story residential precast concrete 

frame buildings in the 1988 Armenia earthquake (EERI, 1989). 

A significant integrated analytical and experimental research program is currently underway to 

develop a comprehensive design methodology for precast concrete diaphragm systems. The 

project intends to address the discrepancy between current design practice, based on inelastic 

behavior concentrating in vertical elements and observed performance in which substantial 

inelastic behavior has occurred in diaphragms (Wan et al., 2004; and Naito and Cao, 2004). The 

project proposes to determine force and deformation demands required for design, connection 

details to support the performance, and address deformation relative to the gravity load-carrying 

system. This information will be invaluable for both new design and rehabilitation. Testing will 

include individual connections, joints, and half-size components. Analytical modeling of full 

buildings is being used to identify critical demands. Of particular interest is the simultaneous 

occurrence of shear and tension or compression on connections normally considered to carry 

only shear. Published information to date (Naito and Cao, 2004) provides a database of 

connector properties from existing literature and suggests a simplified analysis model based on 

initial finite element testing. Additional information should be available over the next several 

years. 

Rehabilitation of diaphragm chord and collector members is reasonably practical due to the 

focused locations of work. Where possible, it is easiest to add reinforcing steel collectors on the 

top of the floor system in new cast concrete curbs. Where chords and collectors need to occur at 

building interior locations where traffic must cross, more complex solutions are required. 

Unstressed post-tensioning tendons may be a desirable alternative to rebar in some locations; 

however, it must be kept in mind that stresses must be kept low in order to minimize 
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deformation, so the high strength does not give particular advantage. In undertaking 

rehabilitation of the chord and collector members, it must be acknowledged that there is lack of 

consensus on the interaction of shear and flexure demand in these buildings. Care must be taken 

not to induce brittle shear failure of the floor diaphragm as a result of flexural strengthening. 

Rehabilitation of inadequate shear capacity is significantly more difficult. Most precast floor 

systems will have little capacity to support vertical load from additional topping slab thickness, 

and removal of existing toppings over large areas is not practical. Research conducted on 

connections between precast wall panels can be applied to connections between precast 

diaphragm members. Research by Pantelides, Volnyy, Gergeley, and Reaveley (2003), discussed 

in relation to load path connection in Section 17.4.2, could be applied to precast diaphragms. In 

addition a research program is currently being conducted by the Precast/Prestressed Concrete 

Institute investigating development of ductile panel-to-panel connections in precast diaphragms. 

See Section 22.2.11 for further discussion. 

Rehabilitation for large openings and re-entrant corners in PC2 buildings involves providing 

adequate chord and collector members in the vicinity of the opening or corner. Rehabilitation 

methods discussed in Section 17.4.1 are applicable. Framing bays at ramps in parking garages 

may need to be treated as openings for purposes of diaphragm design. 

17.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

To date, very little rehabilitation of PC2 buildings has occurred in the U.S. As a result, the 

following discussion of rehabilitation measures draws from limited available research, suggested 

details for new PC2 construction, and application of rehabilitation techniques for concrete 

buildings to the specific configurations of precast elements. 

17.4.1 Add or Enhance Collector or Chord in Existing Precast Diaphragm 

Deficiencies Addressed by Rehabilitation Technique 

This rehabilitation technique addresses deficient diaphragm boundary members – chords and 

collectors at diaphragm boundaries and at interior openings and re-entrant corners. 

Description of the Rehabilitation Technique 

In diaphragms with cast-in-place topping slabs, existing chords and collectors are likely to be 

reinforcing bars at the edge of the cast-in-place slab and in line with shear walls. The most 

practical rehabilitation technique is addition of structural steel sections or reinforcing bars at the 

diaphragm boundary locations. Where boundary members occur at the perimeter of parking 

structures, it may be possible to encase the steel sections or reinforcing in new concrete curbs on 

top of the existing deck, as shown conceptually in Figure 17.4.1-1. Where boundary members 

extend across the floor plate where foot or vehicle traffic will occur, alternate chord and collector 

locations are required, as shown conceptually in Figures 17.4.1-2 and 17.4.1-3. For both chords 

and collectors, shear transfer capability between the boundary member and the structural 

diaphragm needs to be provided. Adhesive anchors or reinforcing dowels are the most likely 
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Figure 17.4.1-1: Added or Enhanced Chord or Collector at Floor Perimeter 

with Cast-In-Place Topping Slab 

Figure 17.4.1-2: Added or Enhanced Chord or Collector at Floor Interior 

with Cast-In-Place Topping Slab 
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Figure 17.4.1-3: Added or Enhanced Chord or Collector at Floor Interior with 

Cast-In-Place Topping Slab 

methods of attachment. Chord and collector splices must also be detailed. For collectors, transfer 

of the collector force to the vertical elements of the lateral force-resisting systems is required. 

Figure 17.4.1-4 shows a conceptual approach where the chord/collector curb runs by the face of 

the shear wall and dowels in over the full shear wall length. Figure 17.4.1-5 illustrates 

considerations when collector steel is to be doweled into the end of an existing shear wall, which 

may not be advisable. 

In diaphragms without cast-in-place topping slabs, reinforcing in wall panels, floor panels, or 

beams will likely serve as chord and collector reinforcing. Where these floor panels or beams are 

precast, the connection between members is likely to be the weak link in chord and collector 

capacity. Figures 17.4.1-6 and 17.4.1-7 show the concept of added steel angles used as chords 

and collectors. The steel angle also serves as the connection between the wall and diaphragm for 

in-plane and out-of-plane forces. 

Design Considerations 

Research basis: No research applicable to this rehabilitation measure has been identified. 

The objective of this rehabilitation method is to enhance the ability of the diaphragm to perform 

adequately and to deliver forces to the vertical elements of the lateral force-resisting system. In 
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Figure 17.4.1-4: Added Collector Anchorage to Shear Wall with 

Cast-In-Place Topping Slab 

order to achieve this, careful consideration of the existing diaphragm configuration and shear 

strength is needed in order to determine what behavior can be achieved and what detailing 

approach is best. In long-span diaphragms, yielding of a chord member may be a preferred 

behavior in order to protect against in-plane shear failure. Where this is true, it should be 

anticipated that the chord member is likely to elongate in one or more locations under tension 

loading, opening up gaps between adjacent precast members. When loading reverses, 

compression will be carried by the chord member until gaps close. In order to prevent local 

buckling failure of the chord, it is advisable to either use a structural steel section that can be 

adequately braced against buckling with a reasonable adhesive anchor spacing, or provide 

confinement of reinforcing bars, as would be provided for a concrete column. These alternatives 

are shown at the right hand side of Figure 17.4.1-1. Both of these alternatives will be more costly 

and difficult to install. In diaphragm configurations where yielding of the chord and collector 

members can be avoided, providing additional reinforcing may be less costly than detailing for 

buckling restraint. Where this approach is taken, a careful evaluation of anticipated forces is 

needed. In general, it is the philosophy of ASCE 7 seismic provisions for inelastic behavior to be 

focused in vertical elements rather than collectors, allowing the vertical elements of the seismic 

force-resisting system to control building dynamic behavior. This may not be achievable in 

diaphragms where the collector serves as a chord for loading in the perpendicular direction, and 

avoiding shear failure is paramount. In this case, use of a detail that restrains buckling is 

recommended. 
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Figure 17.4.1-5: Rebar Embedment Considerations in Collector Anchorage 

to Existing Shear Wall 

For the chord configurations shown in Figure 17.4.1-6, the existing original connection (if any) 

likely consists of intermittent cast-in embed plates and a welded plate connection. The capacity 

of anchors embedded in the concrete and welds would have been sized to meet load requirements 

(wind or earthquake) applicable at time of construction. Even if earthquake loading were 

considered, the need to allow for forces in excess of design levels, ductility, and energy 

dissipation likely would not have been considered. Because inelastic behavior in the precast 

walls and diaphragms is very unlikely, it must be anticipated that inelastic behavior will 

concentrate in the connections between members. With many existing connections, evaluation 

would likely identify failure of the anchors embedded in concrete as the weak link. This is an 

undesirable weak link due to lack of ductility. Unless extreme overstrength has been provided, to 

allow the connection to remain elastic, rehabilitation of the connections is needed in order to 

avoid this weak link. ACI 318 Appendix D (ACI, 2005) provisions for anchorage to concrete 

require that that design be governed by tensile or shear strength of a ductile steel element rather 
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17-16 

the concrete capacity. This requirement is particularly appropriate for rehabilitation of 

connections in precast wall buildings. 

Figure 17.4.1-6: Steel Chord or Collector at Floor Perimeter 

without Cast-In-Place Topping 

Figure 17.4.1-7: Steel Chord or Collector at Floor Perimeter 

without Cast-In-Place Topping 
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The deformation compatibility of the existing and added connections needs to be carefully 

considered. It may be desirable to design new connections to carry all forces, neglecting the 

contribution of the existing connections. In some cases it might even be desirable to cut the weld 

on the original connection to ensure that it will not carry load. 

Detailing Considerations 

Use of reinforcing dowels to anchor new chord or collector reinforcing to existing concrete, as 

shown in Figures 17.4.1-1 and 17.4.1-2, relies on shear friction. Roughening of the existing 

concrete to the amplitude required by ACI 318 requires a very heavy sandblasting or chipping 

with a jackhammer, in order to expose the aggregate. This is expensive and messy. It is often 

preferable to use a reduced μ coefficient of less than 1.0 and add more dowels, rather than 

roughening the surface. Removal of finishes and cleaning of the concrete surface is still required. 

Use of shear friction also requires that the yield capacity of the dowel be developed on both sides 

of the joint. This forces use of smaller dowels and a curb dimension adequate to develop 

standard hooks per ACI 318 requirements. Adhesive anchor embedment requirements for 

development of the bar yield are generally available from the manufacturer. 

Figure 17.4.1-2 places new chord or collector reinforcing at the underside of the floor so that 

vehicle or foot traffic above is not disrupted. Where chord or collector reinforcing runs parallel 

to the T-beams, use of reinforcing bars is likely feasible. Where the chord or collector runs 

perpendicular, reinforcing bars would have to be installed in short lengths threaded thru the Tbeam webs. An unstressed tendon that can be more easily placed might be a preferable 

alternative. 

Figure 17.4.1-3 uses the existing girder as the chord or collector member and adds angles for 

shear transfer between the slab and collector. This must be used in combination with enhanced 

connections at the girder ends. The figure illustrates the possible pre-earthquake cracks in the 

topping slab described by Wood, Stanton and Hawkins (2000). The shear transfer angles bypass 

this potentially weak location. Where steel splice angles are used, splice details will be required. 

A field welded splice would be common in concrete rehabilitation. Where yielding of the steel 

chord/collector member can be anticipated, it may be desirable to proportion the splice plates and 

welds to develop the anticipated strength of the chord/collector member. 

Figure 17.4.1-5 illustrates considerations when collector steel is to be doweled into the end of an 

existing shear wall. This is only possible when there is adequate reinforcing within the wall in 

the vicinity of the collector anchorage to distribute the collector force over the wall length. 

Anchoring two No. 9 bars to the end of walls with only two No. 4 bars should not occur. Further, 

it is important that the collector lap adequately with reinforcing in the shear wall that can 

distribute the collector force over the wall length, as shown in the upper two figures. If 

embedment were limited to the requirements of the anchor manufacturer, as shown in the bottom 

figure, failure of the collector may occur due to inability of the wall reinforcement to develop. 

Cost/Disruption and Construction Considerations 

The work required to add or enhancement chords and collectors is generally spread out over 

considerably over the building area. This distribution of work is reasonably easy in parking 

garages because there are generally no finishes to remove and replace, all areas are reasonably 
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accessible for materials and equipment, and user relocation does not involve a big effort. In 

contrast, the distribution of work will create significantly greater cost and coordination in other 

building types such as commercial or residential. In these building types, it may well be easier 

and more cost effective to add vertical elements, rather than enhancing chords and collectors, 

because this work may be located in one or more local areas. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique other than the use of 

proprietary anchors as part of the assemblage. 

17.4.2 Enhance Connections Between Existing Precast Diaphragm, Shear Wall 

and Foundation Elements 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses rehabilitation of inadequate connections between precast 

concrete elements including: diaphragm to shear wall, shear wall above to shear wall below, 

shear wall to foundation, and shear wall panel to panel connections. 

Description of the Rehabilitation Technique 

Existing connections of precast wall and diaphragm elements are generally either welded 

connections between cast-in embed plates, or connections involving mechanical, grouted or 

welded rebar splices. Where precast elements are combined with cast-in-place elements, Figures 

17.4.2-1 and 17.4.2-2 show connections between precast wall panels and foundation and 

diaphragms. The connections depict one possible configuration for existing welded connections 

and rehabilitation approaches for enhancing connection shear capacity. The rehabilitation 

measures involve adding new steel angles and adhesive anchors as required to carry seismic 

forces. Figures 17.4.2-1B shows added anchorage to a hollow core wall section using grout, 

which provides a high connection capacity. Alternately an anchor specifically designed for 

attachment to hollow masonry could be used, resulting in a lower connection capacity. 

Design Considerations 

Research basis: FRP composite connections between wall panels have been tested at the 

University of Utah (Pantelides, Volnyy, Gergeley, and Reaveley, 2003). 

Existing precast shear wall connections most often use cast-in embed plates and welded plate 

connections. The capacity of anchors embedded in the concrete and welds would have been sized 

to meet load requirements (wind or earthquake) applicable at time of construction. Even if 

earthquake loading were considered, the need to allow for forces in excess of design levels, 

ductility and energy dissipation likely would not likely have been considered. Because inelastic 

behavior in the precast walls and diaphragms is very unlikely, it must be anticipated that inelastic 

behavior will concentrate in the connections between members. With many existing connections, 

evaluation would likely identify failure of the anchors embedded in concrete as the weak link. 

This is an undesirable weak link due to lack of ductility. Unless extreme overstrength has been 

provided, allowing the connection to remain elastic, rehabilitation of the connections is needed in 

order to avoid this weak link. ACI 318 Appendix D (ACI, 2005) provisions for anchorage to 

concrete require that the design be governed by tensile or shear strength of a ductile steel element 
Techniques for the Seismic Rehabilitation of

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17-19 

rather the concrete capacity. This requirement is particularly appropriate for rehabilitation of 

connections in precast wall buildings. The deformation compatibility of the existing and added 

connections needs to be carefully considered. It may be desirable to design new connections to 

carry all forces, neglecting the contribution of the existing connections. In some cases, it might 

even be desirable to cut the weld on the original connection to ensure that it will not carry load. 

Figure 17.4.2-1: Added or Enhanced Precast Wall-to-Foundation Connection 

Each precast wall panel tends to have two existing welded connections at the wall top and two at 

the wall bottom, and connections are generally not provided panel to panel. As a result, the two 

connections at the wall bottom must resist both shear and overturning forces from in-plane loads 

as well as out-of-plane loads. This is illustrated in Figure 17.4.2-3A. Where multiple wall panels 

are in line, as shown in Figure 17.4.2-3B, it may be possible to add panel to panel connections to 

resist overturning, reducing the demand on the bottom of panel connections. Overturning 

capacity will still be needed at each end of the panel group. The perpendicular wall panel at the 

right side of Figure 17.4.2-3 should not be counted on to resist the uplift, since it may also be 

overturning. 

Figure 17.4.2-4 illustrates an FRP composite connection between wall panels developed and 

tested at the University of Utah (Pantelides, Volnyy, Gergeley, and Reaveley, 2003) that would 

be applicable to this use. While various surface preparations and FRP applications were 

investigated, the researchers settled on use of a high-pressure water jet preparation of the 

concrete surface to expose aggregate, a bonding agent and dry lay-up of carbon fibers, and 

saturation with an epoxy resin. A lay up of six layers of 12k (12,000 threads per tow) carbon 

fiber reinforcing in a 16 inch by 48 inch rectangle on one face of the wall panels provided failure 
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loads on the order of 40 kips. The failures were sudden and brittle, indicating that design of this 

connection type should be considered force-controlled, and ductility should be provided in other 

connections or members. See Section 13.4.1 for general information on FRP composite 

overlays. 

Figure 17.4.2-2: Precast Wall Connection 

The new connections shown have both vertical and horizontal eccentricities that must be 

considered in design of the angle section and the anchors. 

Detailing Considerations 

Anchorage to hollow core wall panels is more difficult than solid precast panels because the 

voids do not leave solid concrete sections large enough to meet embedment and edge distance 

requirements for adhesive anchors. For most anchorage, drilling grout access and inspection 

holes and filling the lower portion of the void with grout can provide a solid cell for anchor 

placement. Similar grouting of floor planks is sometimes used in new construction. For very light 

anchorage loads, it may be possible to use screen-tube adhesive anchors specifically designed for 

attachment to hollow masonry units. Along with low capacity, failure of this anchorage should 

be expected to have brittle behavior. 
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Figure 17.4.2-3: Modification of Demand on Anchors 

Through Use of Panel-to-Panel Connections 

Cost, Disruption and Construction Considerations 

The cost of retrofitting connections in areas covered by finishes can be very expensive due to 

finish removal and replacement. The use of high-pressure water jets for surface preparation is not 

practical for an occupied building with finishes. It may be possible in an open building such as a 

parking garage. 

Proprietary Concerns 

FRP systems are proprietary. Manufacturers should be contacted for appropriate uses and 

limitations. Proprietary adhesives are used as part of connection details. 
Techniques for the Seismic Rehabilitation of

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Figure 17.4.2-4: Panel Setup for Testing of FRP Panel-to-Panel Connections 

17.4.3 Mitigate Configurations Creating Short Columns 

Deficiency Addressed by Rehabilitation Technique 

This technique addresses mitigation of unintended shortening and fixity of concrete columns. 

Description of the Rehabilitation Technique 

Figures 17.4.3-1, 17.4.3-2 and 17.4.3-3 illustrate unintended shortening and stiffening of 

columns, common in parking garages. The condition in Figure 17.4.3-1 is very easily mitigated 

by sawcutting the required gap between the column and the guardrail. The connection at the base 

of the guardrail may need to be improved as a result. The condition in Figure 17.4.3-2 can be 

improved by creating a hinge or joint in the column at the top of slab level. This will require 

shoring to take the load off of the column during modification, and the development of a shear 

and tension connection to the pedestal and footing. The condition in Figure 17.4.3-3 is difficult 

to address. Columns with this configuration are not easily confined. A solution that can minimize 

column shortening for new construction is to provide separate columns for the ramp and level 

floors. This approach could also be attempted for a retrofit, but would be costly. 

Design Considerations 

Research basis: No research applicable to this rehabilitation measure has been identified. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique. 
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Figure 17.4.3-1: Accidental Reduction of Column Height Due to Guardrail 

Figure 17.4.3-2: Accidental Increase in Column Fixity 

Due to Embedment into Grade 
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Figure 17.4.3-3: Accidental Reduction of Column Height at Ramp 

17.5 References 

ACI, 2005, Building Code Requirements for Structural Concrete (ACI 318), American Concrete 

Institute, Farmington Hills, MI. 

City of Los Angeles and SEAOSC (Structural Engineers Association of Southern California), 

Findings & Recommendations of the City of Los Angeles/SEAOSC Task Force on the Northridge 

Earthquake, 1994, City of Los Angeles, CA. 

EERI, 1989, Armenia Earthquake Reconnaissance Report, Earthquake Spectra Special Edition, 

Earthquake Engineering Research Institute, Oakland, CA, August. 

EERI, 1995, Guam Earthquake of August 8, 1993 Reconnaissance Report, Earthquake Spectra, 

Supplement B to Volume 11, Earthquake Engineering Research Institute, Oakland, CA, April. 

EERI, January 1996, Northridge Earthquake of January 17, 1994, Reconnaissance Report, 

Volume 2, Earthquake Spectra, Supplement C to Volume 11, Earthquake Engineering Research 

Institute, Oakland, CA. 

FEMA, 2004, NEHRP Recommended Provisions for Seismic Provisions for New Buildings and 

Other Structures, FEMA 450, Federal Emergency Management Agency, Washington, D.C. 
Techniques for the Seismic Rehabilitation of

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Ghosh, S.K., “Observations on the Performance of Structures in the Kobe Earthquake of January 

17, 1995,” PCI Journal, Vol. 40, No. 2, March/April 1995, Precast/Prestressed Concrete 

Institute, Chicago, IL. 

Iverson, J.K. and N.M. Hawkins, 1994, “Performance of Precast/ Prestressed Concrete Building 

Structures During the Northridge Earthquake,” PCI Journal, Vol. 39, No. 2, Precast / Prestressed 

Concrete Institute, Chicago, IL. 

Lyons, S., R. Bligh, J. Purlinton, and D. Beaudoin, 2003, “What about the Ramp? Seismic 

Analysis and Design of Ramped Parking Garages,” Proceedings of the 2003 SEAOC Convention, 

Structural Engineers Association of California, Sacramento, CA. 

Mooradian, D., 2005, personal communication. 

Naito, C. and L. Cao, 2004, “Precast Diaphragm Panel Joint Connector Performance,” 

Proceedings of the 13th World Conference on Earthquake Engineering, Vancouver, B.C. Canada. 

Nakaki, S., 1998, Design Guidelines: Precast and Cast-in-Place Concrete Diaphragms, 

Earthquake Engineeering Research Institute, Oakland, CA. 

Pantelides, C. P., V. A. Volnyy, J. Gergeley, and L.D. Reavely, 2003, “Seismic Retrofit of 

Precast Concrete Panel Connections with Carbon Fiber Reinforced Polymer Composites,” PCI 

Journal, Vol. 48, No. 1, January/February, Precast/Prestressed Concrete Institute, Chicago, IL. 

Priestley, M.J.N., S. Sritharan, JR. Conley, and S. Pampanin, 1999, “Preliminary Results and 

Conclusions From the PRESSS Five-Story Precast Concrete Test Building,” PCI Journal, Vol. 

44, No. 6, November/December, Precast/Prestressed Concrete Institute, Chicago, IL. 

Wan, G., R. Fleischman, J. Restrepo, R. Sause, C. Naito, S.K. Ghosh, L. Cao, and M. Schoettler, 

2004, “Integrated Analytical and Experimental Research Program to Develop a Seismic Design 

Methodology for Precast Diaphragms,” Proceedings of the 2004 SEAOC Convention, Structural 

Engineers Association of California, Sacramento, CA. 

Wood, S. L., J.F. Stanton, and N.M. Hawkins, 2000, “New Seismic Design Provisions for 

Diaphragms in Precast Concrete Parking Structures,” PCI Journal, Vol. 45, No. 1, 

January/February 2000, Precast/Prestressed Concrete Institute, Chicago, IL. 
Techniques for the Seismic Rehabilitation of

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Chapter 18 - Building Type RM1t: Reinforced Masonry 

Bearing Walls (Similar to Tilt-up Concrete Shear Walls) 

18.1 Description of the Model Building Type 

Building Type RM1 is constructed with reinforced masonry (brick cavity wall or concrete 

masonry unit) perimeter walls with a wood or metal deck flexible diaphragm. For this document, 

Building Type RM1 is separated into two categories. Chapter 19 describes RM1u, which is 

multistory, and typically has interior CMU walls and shorter diaphragm spans. This chapter 

covers RM1t, the large, typically one-story buildings with relatively open interiors that are 

similar to concrete tilt-ups. The exterior walls are commonly bearing, with an interior post and 

beam system of steel or wood. Older buildings of this type are generally small and used for a 

wide variety of occupancies. Recently, the building type has become commonly used for onestory warehouse and wholesale/retail occupancies similar to tilt-up (Building Type PC1) 

buildings. Figure 18.1-1 illustrates one example of the RM1t Building Type. 

Figure 18.1-1: Building Type RM1t: Reinforced Masonry Bearing Walls 
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Guidelines for Seismic Evaluation and Rehabilitation of Tilt-Up Buildings and Other Rigid 

Wall/Flexible Diaphragm Structures or SEAONC Guidelines (SEAONC, 2001) provides a 

substantial collection of information on flexible diaphragm / rigid wall building issues, the West 

Coast experience with earthquake performance, rehabilitation priorities, and techniques for 

rehabilitation. This document was a primary source of information for Chapter 16 (PC1

buildings), and it is also recommended for RM1t buildings. 

Walls 

Exterior reinforced masonry walls are the primary vertical elements in the lateral load-resisting 

system. Buildings with large plan areas may have interior walls providing additional lateral 

resistance; however, this is not common. Like tilt-up construction, masonry walls have recently 

transitioned from use of code-prescribed height to thickness (h/t) limits to use of much higher h/t

ratios, in combination with rational analysis of slenderness effects. Most existing construction, 

however, will have been designed using prescriptive ratios and allowable stress design methods. 

Reinforced masonry walls in RM1t buildings share many issues with PC1 buildings, discussed 

in Chapter 16. Two distinctive aspects of reinforced masonry walls require discussion: 

movement control joints and partially grouted masonry. 

The masonry industry recommends providing vertical control joints in new masonry wall 

construction to accommodate masonry wall shrinkage and thermal movement. The currently 

recommended maximum spacing of control joints is the lesser of 1.5 times the wall height or 25 

feet (CMACN, 2003). The inclusion of and spacing of control joints varies significantly, 

however, with the age and region of the building. Wall-to-diaphragm anchorage remains a 

primary concern irrespective of whether control joints are provided. Where vertical control joints 

occur, diaphragm chords and collectors will be provided by either horizontal reinforcing that 

continues across the control joint (typically provided at the diaphragm level only), or a steel 

angle or similar member on the face of the masonry. Rehabilitation of chords, collectors, and 

shear transfer for RM1t building will be much the same as PC1 buildings. Masonry wall 

construction without control joints will not have to rely as much on discreet chords and 

collectors; adequacy of shear transfer may still be a focus of rehabilitation. 

In areas of high seismic hazard, it is most common for masonry walls to be fully grouted. In 

many other areas, however, grout is only provided at required reinforcing. A typical reinforced 

masonry wall in a RM1t building might have vertical reinforcing and grout alongside window 

and door openings, and at between four and ten feet on center horizontally and vertically in the 

piers and spandrels. Partial grouting has a significant effect on the weight of the wall and 

calculated seismic forces, as well as wall strength for both in-plan and out-of-plane forces. 

Where partial grouting has been provided, grout locations need to be known in order to design 

wall to diaphragm anchorage. 

Gravity Load-Carrying Support at the Building Perimeter 

It is most common for wood girders to be supported on the building exterior walls in RM1t

buildings. As in PC1 buildings, girder connections to the exterior walls are required to resist wall 

out-of-plane loading in addition to gravity loads. Connections of girders to the exterior walls 

require evaluation and possible rehabilitation. Section 16.4.2 discusses applicable rehabilitation 
Techniques for the Seismic Rehabilitation of

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18-3 

measures. Where the existing masonry is partially grouted, it may be necessary to open up the 

masonry face and grout at new anchorage locations. Where this is done, cast-in anchors can be 

provided in lieu of adhesive anchors. 

Roof Diaphragms 

Like the PC1 building, the roof system will generally be either of wood or steel construction. In 

the western states, roof systems are almost exclusively wood sheathed. Outside of the western 

states, roof systems are almost exclusively sheathed with steel decking topped with rigid 

insulation or vermiculite concrete. For both the wood and the steel roof systems, the roof 

diaphragm in the RM1t building is almost always flexible compared to the walls. See Chapter 16 

for additional discussion. 

Wall-to-Diaphragm Connections 

Like PC1 buildings, wall-to-diaphragm connections are thought to be the aspect of RM1t

buildings most vulnerable to earthquake damage, due to the significant force and deformation 

demands imposed on this connection. The wall-to-diaphragm connections are therefore 

recommended as the first focus of rehabilitation measures for one-story RM1t buildings. See 

Chapter 16 for a discussion of past performance of these connections in tilt-up buildings. Similar 

to tilt-up buildings, even in RM1t buildings constructed or upgraded recently, it should be 

assumed that these connections require review and possible rehabilitation. 

Interior Additions 

Mezzanines and interior second stories, commonly constructed within large box-like RM1t

buildings, can restrain building movement under earthquake loading, resulting in unintended 

load paths and damage. See Chapter 16 for discussion. 

Foundations 

RM1t buildings are generally constructed on continuous perimeter footings, with dowels to the 

reinforced masonry walls at vertical reinforcing locations. 

18.2 Seismic Response Characteristics 

RM1t buildings, like PC1 buildings, are distinguished by rigid shear walls and flexible 

diaphragms. Like PC1 buildings, amplification of seismic forces near the center of the 

diaphragm is of concern for both diaphragm capacity and wall anchorage capacity. The in-place 

construction of the reinforced masonry walls may provide wall continuity that makes RM1t

buildings somewhat less vulnerable to partial collapse than PC1 buildings; however, the 

potential for significant performance problems exists. 

18.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

Unlike PC1 buildings, damage to RM1t buildings has not been noted as significant or wide 

spread. In the 1994 Northridge earthquake, little in the way of damage to RM1t buildings was 

reported (EERI, 1996; and Klingner, 1994). One item of interest was damage to masonry walls at 

building corners near the roof line, attributed to interaction between the masonry wall and 

flexible wood diaphragm. See below for general discussion and Table 18.3-1 for a detailed 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 18 – Type RM1t: Reinforc

18-4 

Table 18.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for RM1t Bui

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Dem

Global Strength 

 Steel braced frame [7.4.1] Concrete/masonry shear wall [21.4.8] 

Concrete wall overlay [21.4.5] Infill openings 

Global Stiffness 

Steel braced frame [7.4.1] Concrete/masonry shear wall [21.4.2] 

Concrete wall overlay [21.4.5] Infill openings 

Torsionally irregular plans 

Steel braced frame [7.4.1] Concrete/masonry shear wall [21.4.8] 

Concrete wall overlay [21.4.5] Infill openings 

Re-entrant corners Steel braced frame 

[7.4.1] Concrete/masonry shear wall [21.4.8] Collector [7.4.2] 

Enhance existing collector Concrete wall overlay [21.4.5] Infill openings 

Collector [7.4.2] 

Configuration 

Incidental bracing 

Inadequate or missing wall-todiaphragm tie for out-of-plane load 

Wall-to-diaphragm tension anchors plus subdiaphragms and cross-tie [16.4.1] 

Inadequate anchorage to diaphragms for inplane forces 

Wall-to-diaphragm shear anchors [21.4.2] 

Load Path 

Inadequate collectors 

Add collector [7.4.2] Improve collector 

member and connections 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 18 – Type RM1t: Reinforc

18-5 

Table 18.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for RM1t Bui

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Dem

Wall inadequate for out-of-plane bending 

Wall strongback or pilaster [21.4.3] 

Component Detailing 

Inadequate detailing of slender walls 

Steel braced frame [7.4.1] Concrete/masonry shear wall [21.4.8] 

Concrete wall overlay [21.4.5] Infill openings Add backup vertical supports where bearing might be lost [21.4.11] 

Inadequate in-plane strength and/or stiffness 

Steel braced frame [7.4.1] 

Concrete/masonry shear wall [21.4.8] 

Enhance existing diaphragm [22.2.1] Horizontal braced frame [21.2.10] 

Inadequate chord capacity 

Enhance chord 

 [22.2.2] 

Enhance chord [22.2.2] 

Excessive stresses at openings and irregularities 

Enhance diaphragm detailing 

Diaphragms 

Re-entrant corners Enhance diaphragm 

detailing 

Foundations See Chapter 23 [ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation measures. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 18 – Type RM1t: Reinforced Masonry Bearing Walls

18-6 

compilation of common seismic deficiencies and rehabilitation techniques for Building Type

RM1t. See also Chapter 16 for similar issues in PC1 buildings. 

Global Strength and Stiffness 

Global strength and stiffness are rarely a concern for large box-like RM1t buildings, but can be 

for smaller buildings that have very short walls along a street-front side. Rehabilitation of global 

strength and stiffness deficiencies will commonly involve adding new vertical elements, 

enhancing existing elements, or infilling openings in existing walls. 

Configuration 

Poor distribution of shear walls can result in torsionally irregular behavior of RM1t buildings. 

Common occurrences include street-front walls in commercial buildings. The most direct 

approach to rehabilitation of this condition is the addition of strength and stiffness in line with 

the perforated wall. This can be accomplished through addition of new shear walls, enhancing 

existing shear walls, or addition of steel braced frames. 

Rehabilitation at re-entrant corners requires the provision of adequate chords and collectors, 

shear transfer to the in-set wall panels, and possibly the strengthening of the wall panels and 

connections to the foundation. The SEAONC Guidelines suggest that there may be diaphragm 

continuity over this interior diaphragm support, increasing the diaphragm reaction to the in-set 

wall line. See Chapter 16 for illustration and additional discussion. 

Load Path 

As previously mentioned, load path connections between the masonry walls and the flexible 

diaphragm are suggested as the first focus of rehabilitation in RM1t buildings. Diaphragm crossties, as addressed in Chapter 22, are a required continuation of the wall anchorage system. 

Section 16.4.1 discusses applicable rehabilitation measures. Where the existing masonry is 

partially grouted, it may be necessary to open up the masonry face and grout at new anchorage 

locations. Where this is done, cast-in anchors can be provided in lieu of adhesive anchors. 

Connection between the wall and diaphragm may also be inadequate for in-plane shear loads. 

The addition of or enhancement of existing collectors may be required in order to transmit 

diaphragm forces to the resisting shear walls. This is particularly of concern when a limited 

length of shear wall intended to carry a significant portion of the building shear. Although not as 

common, there is also significant concern when vertical offsets in the roof diaphragm result in 

incomplete chords or collectors. Any breaks or offsets in chords or collectors need to be carefully 

evaluated. 

Component Detailing 

Component detailing deficiencies include inadequate out-of-plane wall capacity. Where existing 

walls are partially grouted concrete masonry, it may be possible to place additional vertical 

reinforcing and grout in ungrouted cells, accessed by cutting open face shells. Where this 

approach is taken, doweling of the vertically reinforcing at the top and bottom of the grouted 

cells would commonly be provided. In addition, the increase in wall weight should be considered 

in building seismic forces. Where existing masonry wall construction is solid, it is seldom 

practical to address wall capacity by adding reinforcing and concrete thickness to individual wall 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 18 – Type RM1t: Reinforced Masonry Bearing Walls

18-7 

sections, so addition of wall pilasters or strongbacks is common. Where pilasters are added to 

masonry walls, pilaster-to-roof diaphragm anchorage must be provided to accommodate the 

concentration of wall out-of-plane force. 

Diaphragm Deficiencies 

Due to changes in building code requirements, it is very common for diaphragms in areas of high 

seismic hazard to have inadequate in-plane shear capacity. These diaphragms may also have 

inadequate in-plane stiffness due to high unit shear stresses. Regardless of this, the SEAONC 

Guidelines indicate that diaphragm overstresses have rarely been associated with significant 

earthquake damage. Diaphragm strength and stiffness deficiencies are most often rehabilitated by 

enhancing the existing diaphragm. 

Other diaphragm deficiencies include inadequate chord capacity and stress concentrations at 

large diaphragm openings and re-entrant corners. Rehabilitation at re-entrant corners primarily 

involves the provision of adequate chords and collectors. The same is true at large diaphragm 

openings. 

18.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

No techniques have been developed for this building type. See other chapters for detailed 

descriptions of relevant rehabilitation techniques. 

18.5 References 

CMACN, 2003, “Movement Control Joints,” Masonry Chronicles Winter 02-03, Concrete 

Masonry Association of California and Nevada, Citrus Heights, CA. 

EERI, January 1996, Northridge Earthquake of January 17, 1994, Reconnaissance Report, 

Volume 2, Earthquake Spectra, Supplement C to Volume 11, Earthquake Engineering Research 

Institute, Oakland, CA. 

Klingner, R., 1994, Performance of Masonry Structures in the Northridge California Earthquake 

of January 17, 1994 (Technical Report 301-94), The Masonry Society, Boulder, CO. 

SEAONC (Structural Engineers Association of Northern California), 2001, Guidelines for 

Seismic Evaluation and Rehabilitation of Tilt-Up Buildings and Other Rigid Wall/Flexible 

Diaphragm Structures, Structural Engineers Association of Northern California, San Francisco, 

CA. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 19 – Type RM1u: Reinforced Masonry Bearing Walls

19-1 

Chapter 19 - Building Type RM1u: Reinforced Masonry 

Bearing Walls (Similar to Unreinforced Masonry Bearing 

Walls) 

19.1 Description of the Model Building Type 

Building Type RM1 takes a variety of configurations, but they are characterized by reinforced 

masonry walls with flexible diaphragms such as wood or metal deck. The walls are commonly 

bearing, but the gravity system also contains post and beam construction of wood or steel in 

interior or some façade locations. For this document, Building Type RM1 is separated into two 

categories. Chapter 18 describes RM1t, the large, typically one-story buildings with relatively 

open interiors that are similar to concrete tilt-ups. This chapter covers RM1u, which is 

multistory, and typically has interior CMU walls and shorter diaphragm spans. It is similar to 

Building Type URM (Chapter 21) and has many of the same deficiencies. Figure 19.1-1 shows 

an example of this building type. 

Figure 19.1-1: RM1u Building Type: Reinforced Masonry Bearing Walls 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 19 – Type RM1u: Reinforced Masonry Bearing Walls

19-2 

Masonry Wall Materials 

FEMA 306 (FEMA, 1999) identifies several common reinforced masonry wall types. These are: 

Fully-grouted hollow concrete block 

Partially-grouted hollow concrete block 

Fully-grouted hollow clay brick 

Partially-grouted hollow clay crick 

Grouted-cavity wall masonry (two wythes of clay brick or hollow units with a reinforced 

grouted cavity) 

Brick veneer facing may be placed on the exterior façade with the above walls used as backing 

walls. 

Floor and Roof Diaphragm 

Floor and roof diaphragm construction is similar to those of Building Type URM, although 

unfilled metal deck diaphragms can be found at the roof and occasionally at floors. See Chapter 

21. 

Foundations 

Foundations for Building Type RM1u are typically spread footings at interior columns and strip 

footings under masonry bearing walls. Footings are typically concrete. 

19.2 Seismic Response Characteristics 

As a flexible diaphragm, stiff wall structure Building Type RM1u is expected to have dynamic 

behavior similar to that described for Building Type URM. See Chapter 21. Since the walls are 

reinforced, however, in-plane and out-of-plane wall behavior modes are substantially different 

from those of unreinforced masonry walls. They are instead more similar to those of reinforced 

concrete. FEMA 306 provides the most comprehensive categorization of reinforced masonry 

wall behavior modes. 

19.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

RM1u buildings, while similar to URM buildings, are generally considered to be less hazardous. 

In-plane damage to reinforced masonry walls is much less likely to reach levels compromising 

life safety. Parapets can still be overstressed, but the risk to life safety is much less than those of 

unreinforced parapets, and out-of-plane failures of the walls spanning between diaphragms are 

relatively unlikely. The most significant risk to loss of life is due to inadequate connections 

between the walls and diaphragms. See below for general discussion and Table 19.3-1 for a 

detailed compilation of common seismic deficiencies and rehabilitation techniques for Building 

Type RM1u. 

Global Strength 

As shear wall buildings, global strength in RM1u buildings is dependent on the in-plane shear 

capacity of the walls. Relatively large seismic forces are needed to lead to life safety concerns 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 19 – Type RM1u: Reinforc

19-3 

Table 19.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for RM1u Bui

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce D

Global Strength 

Insufficient inplane wall strength 

Wood structural panel shear wall [5.4.1], [6.4.2] Concrete/masonry shear wall [21.4.8] Steel braced frame [7.4.1] Steel moment frame [21.4.9] 

Concrete wall overlay [21.4.5] Fiber composite wall overlay [21.4.6] Grouting Infill openings [21.4.7] 

Seismic [24.3] 

Global Stiffness Configuration Soft story, weak 

story, excessive torsion 

Wood structural panel shear wall [6.4.2] Concrete/masonry shear wall [21.4.8] Steel braced frame [7.4.1] Steel moment frame [21.4.9] 

Load Path Inadequate or 

missing wall-todiaphragm tie 

Tension anchors [16.4.1] 

Shear anchors [21.4.2] 

Subdiaphragms and cross-ties [22.2.3] 

 Missing collector Add collector [7.4.2]  Inadequate girderto-column 

connection 

Improve connection Supplemental vertical supports [21.4.11] 

 Inadequate wallfoundation dowels Wall-to-foundation improvements 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 19 – Type RM1u: Reinforc

19-4 

Table 19.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for RM1u Bui

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce D

Component Detailing 

Wall inadequate for out-of-plane 

Exposed interfloor wall supports [21.4.3] Reinforced cores [21.4.4] Concrete wall overlays [21.4.5] Fiber composite overlays [21.4.6] 

 Poorly anchored 

veneer or appendages 

Add ties 

Diaphragms Inadequate 

in-plane strength and/or stiffness 

Wood structural panel shear wall [6.4.2] Steel braced frame [7.4.1] Concrete/masonry shear wall [21.4.8] Add wood structural panel or moment frame crosswall [21.4.10] 

Horizontal braced frame [22.2.9] 

Enhance existing diaphragm [22.2.1] Enhance crosswall [21.4.10] 

 Inadequate chord 

capacity 

Add steel strap or angle 

 Excessive stresses 

at openings and irregularities 

Add wood or steel strap reinforcement 

 Re-entrant corner Wood structural panel 

shear wall [6.4.2] Concrete/masonry shear wall [21.4.8] Steel braced frame [7.4.1] Steel moment frame [21.4.9] 

Collector [7.4.2] 

Foundations See Chapter 23 [ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 19 – Type RM1u: Reinforced Masonry Bearing Walls

19-5 

with in-plane wall behavior, though cracking damage will occur in relatively moderate events. 

When walls are found to be deficient, new vertical lateral force-resisting elements can be added 

at interior locations or existing walls can be enhanced. 

Global Stiffness 

Reinforced masonry walls are generally quite rigid, even if punctured with window openings, so 

global stiffness deficiencies are relatively uncommon. 

Configuration 

Many commercial RM1u buildings will have a fairly open street façade at the ground level, 

leading to a weak and soft first story and torsional irregularities. This is usually addressed by the 

addition of a moment frame at the façade or another vertical lateral force-resisting element at 

some distance back from the façade. 

Load Path 

As noted above, it is the lack of adequate ties between the walls and diaphragms that is the single 

most significant deficiency in RM1u buildings. Rehabilitation measures include tension ties for 

out-of-plane forces and shear ties for in-plane forces along the typical wall-diaphragm interface. 

In many reinforced masonry walls, pilasters are formed by thickening the wall in order to 

provide support of key girder lines. They may not have adequate bearing length for the girder 

seat or sufficient reinforcement at the top of the pilaster. Anchor bolts from the ledger to the 

wall are likely to be present, but cross-grain bending under out-of-plane tension loading will be a 

common deficiency. Rehabilitation measures include supplemental column supports and 

connection enhancements. 

Component Detailing 

When brick veneer is present, it may not be adequately anchored back to the backing masonry, 

creating a falling hazard. Veneer ties can be added. 

Diaphragm Deficiencies 

Wood diaphragms may lack both strength and stiffness in RM1u buildings. This can be 

addressed by adding new interior elements to cut the diaphragm span or by enhancing the 

diaphragm itself with wood structural panel overlays. 

Foundation Deficiencies 

Foundation deficiencies for existing elements are relatively uncommon in RM1u buildings. 

Foundation rehabilitation work usually is focused on the support for new lateral force-resisting 

elements that are added to the superstructure. 

19.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

Rehabilitation techniques for Building Type RM1u are typically similar to those used in URM

buildings. See Chapter 21 for detailed descriptions of techniques. When metal deck floors are 

present, techniques in Chapter 16 provide examples of rehabilitation methods for connecting the 

metal deck to the tilt-up concrete walls; details for connecting to reinforced masonry walls are 

similar. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 19 – Type RM1u: Reinforced Masonry Bearing Walls

19-6 

19.5 References 

FEMA, 1999, Evaluation of Earthquake Damaged Concrete and Masonry Wall Buildings: Basic 

Procedures Manual, FEMA 306, May. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 20 – Type RM2: Reinforced Masonry Bearing Walls

20-1 

Chapter 20 - Building Type RM2: Reinforced Masonry 

Bearing Walls (Similar to Concrete Shear Walls with Bearing 

Walls) 

20.1 Description of the Model Building Type 

This building consists of reinforced masonry walls and concrete slab floors that may be either 

cast-in-place or precast. In this type of building, all walls usually act as both bearing and shear 

walls. The building type is similar and often used in the same occupancies as Building Type 

C2b, namely in mid- and low-rise hotels and motels. This system is also commonly used in 

residential apartment/condominium type buildings. 

Figure 20.1-1: RM2 Building Type: Reinforced Masonry Bearing Walls 

Variations Within the Building Type 

In order for this framing system to be efficient, a regular and repeating pattern of bearing walls 

are required to provide support points for the floor framing. In addition, since it is difficult and 

expensive to make significant changes in the plan during the life of the building, planning 

flexibility is not normally an important characteristic when this structural system is employed. 

The occupancy types that most often fit these characteristics are residential buildings, including 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 20 – Type RM2: Reinforced Masonry Bearing Walls

20-2 

dormitories, apartments, motels, and hotels. These buildings will often be configured with 

reinforced masonry bearing walls between rooms which act as shear walls in the transverse 

direction, and reinforced masonry walls on the interior corridor which act as shear walls in the 

longitudinal direction. Sometimes the longitudinal lateral system includes the exterior wall 

system, although this wall is normally made as open as possible. In any case, the wide variation 

in structural layouts and occupancies that is included in other shear wall buildings such as C2f is 

not seen in RM2. 

It is seldom possible to plan a building layout that provides complete gravity support with walls, 

and often local areas are supported with isolated columns, sometimes with beams and girders. 

However, story heights in these buildings are usually small, and added depth in the floor framing 

system is difficult to obtain. The extent of such beam and column framing often causes 

confusion as to the classification of the structure as a bearing wall system. However, if 

significant plan area is supported solely by walls, the structures are normally classified as RM2. 

There are important variations in floor framing systems employed in this building type, and their 

adequacy to act as a diaphragm is an important characteristic of this building type as discussed 

below. 

Floor and Roof Diaphragms 

The parallel layouts of supporting walls and the need to minimize story heights normally leads to 

the use of one-way uniform-depth concrete floor systems. Cast-in-place and precast systems, 

both conventionally reinforced and prestressed, have been employed. The precast systems are 

often built up of narrow planks, which may not provide an adequate diaphragm unless a cast-inplace topping is provided. In addition, the precast systems may be placed with only a very 

narrow bearing area on the supporting walls, almost always on the outer masonry wythe or the 

CMU shell. When prestressed, the planks may be connected into the wall system only with the 

tail of the stressing tendon, and this connection may be inadequate to provide vertical support 

during seismic movements. The adequacy of the shear connection between slab and walls is also 

often an issue for both cast-in-place and precast systems. 

Foundations 

The bearing walls obviously require some kind of starter beam at grade for construction 

purposes, and this often leads to a simple continuous grade beam system. In poor soils, piles or 

drilled piers may be added below the grade beam. A continuous mat foundation may also be 

employed due to the short spans and total length of bearing points in this building type. 

20.2 Seismic Response Characteristics 

Due to the extent of wall, bearing wall buildings will be quite stiff. Elastic and early post-elastic 

response will therefore be characterized with lower-than average drifts and higher-than-average 

floor accelerations. Damage in this range of response should be minimal. 

Overall post-elastic response may often include rocking at the foundation level. If rocking does 

not occur, the height-to-length ratio of shear walls in these buildings may force shear yielding 

near the base, which may lead to strength and stiffness degradation. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 20 – Type RM2: Reinforced Masonry Bearing Walls

20-3 

Global stability may also be compromised by poor connections between floor slab construction 

and bearing walls. 

Shear Wall Behavior 

When subjected to every increasing lateral load, individual shear walls or piers will first often 

force yielding in spandrels, slabs, or other horizontal components restricting their drift, and 

eventually walls and piers either rock on their foundations, suffer shear cracking and yielding, or 

form a flexural hinge near the base. Shear and flexural behavior are quite different, and 

estimates of the controlling action are affected by the distribution of lateral loads over the height 

of the structure. 

Yielding of spandrels, slabs, or other coupling beams can cause a significant loss of stiffness in 

the structure. Flexural yielding will tend to maintain the strength of the system, but shear 

yielding, unless well detailed, will degrade the strength of the coupling component and the 

individual shear wall or pier will begin to act as a cantilever from its base. In this building type, 

the coupling elements are often slabs, and their lack of bending stiffness may reduce or eliminate 

significant coupling action. 

Rocking is often beneficial, limiting the response of the superstructure. However, the amplified 

drift in the superstructure from rocking must be considered. In addition, if varying wall lengths 

or different foundation conditions lead to isolated or sequencing rocking, the transfer of load 

from rocking walls must be investigated. In buildings with basements, the couple created from 

horizontal restraint at the ground floor diaphragm and the basement floor/foundation (often 

termed the “backstay” effect) may be stiffer and stronger than the rocking restraint at the 

foundation and should be considered in those configurations. 

Shear cracking and yielding of the wall itself are generally considered undesirable, because the 

strength and stiffness will quickly degrade, increasing drifts in general, as well as potentially 

creating a soft story or torsional response. However, in accordance with FEMA 356 (FEMA, 

2000), shear yielding walls or systems can be shown to be adequate for small target 

displacements. Type RM2 buildings will often fall into this category. 

Flexural hinging is considered ductile in FEMA 356 and will degrade the strength of the wall 

only for larger drifts. Similar to rocking, the global effect of the loss of stiffness of a hinging 

wall must be investigated. 

20.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

See Table 20.3-1 for deficiencies and potential rehabilitation techniques particular to this system. 

Selected deficiencies are further discussed below by category. 

Global Strength 

Due to the extensive use of walls, buildings of this type seldom have deficiencies in this 

category, unless significant degradation of strength occurs due to shear failures. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 20 – Type RM2: Reinforc

20-4 

Table 20.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for RM2 Build

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Dem

Global Strength 

Insufficient inplane wall shear strength 

Concrete/masonry shear wall [12.4.2] 

Concrete wall overlay [21.4.5] Fiber composite wall overlay [13.4.1] Steel wall overlay 

Seismic isolation [2Reduce flexcapacity [13.4.4] 

 Insufficient 

flexural capacity 

Concrete/masonry shear wall [12.4.2] 

Add chords [12.4.3] 

 Inadequate 

capacity of coupling beams 

Concrete/masonry shear wall [12.4.2] 

Strengthen beams [13.4.2] Improve ductility of beams [13.4.2] 

Global Stiffness 

Excess drift (normally near the top of the building) 

Concrete/masonry shear wall [12.4.2] 

Concrete/steel column jackets [12.4.5] Provide detailing of all other elements to accept drifts 

Concrete wall overlay [21.4.5] 

Supplemendamping [2

Configuration Discontinuous 

walls 

Add wall or adequate columns beneath [12.4.2] 

Fiber composite wrap of supporting columns [12.4.4] Concrete/steel jacket of supporting columns [12.4.5] 

Improve connection to diaphragm [13.4.3] 

 Soft story or weak 

story 

Add strength or stiffness in story to match balance of floors 

 Re-entrant corner Add floor area to 

minimize effect of corner 

Provide chords in diaphragm [12.4.3] 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 20 – Type RM2: Reinforc

20-5 

Table 20.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for RM2 Build

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce Dem

Configuration (continued) 

Torsional layout Add balancing walls 

[12.4.2] 

Load Path Inadequate 

collector 

Add steel or concrete collector [12.4.3] 

 Inadequate slab 

bearing on walls 

Add diagonal dowels [13.4.3] Add steel ledger [13.4.3] 

Component Detailing 

Wall inadequate for out-of-plane bending 

Add strongbacks [21.4.3] 

Concrete wall overlay [21.4.5] 

 Wall shear critical Concrete wall overlay 

[21.4.5] Fiber composite wall overlay [13.4.1] 

Reduce flexcapacity [13

Diaphragms Precast 

components without topping 

Improve interconnection [22.2.11] Add topping 

 Inadequate inplane shear 

capacity 

Concrete slab overlay Fiber composite overlays [22.2.5] 

 Inadequate shear 

transfer to walls 

Add diagonal drilled dowels [13.4.3] Add steel angle ledger [13.4.3] 

 Inadequate chord 

capacity 

New concrete or steel chord member [12.4.3] 

 Excessive stresses 

at openings and irregularities 

Add chords [12.4.3] 

Foundation See Chapter 23 

[ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 20 – Type RM2: Reinforced Masonry Bearing Walls

20-6 

Global Stiffness 

Similar to strength, global stiffness is seldom a problem in this building type. However the 

effect of coupling slabs on initial stiffness and the potential change in stiffness due to yielding of 

these coupling slabs or wall-beams over doors should be investigated. 

Configuration 

The most common configuration deficiencies in this building type are weak or soft stories 

created by walls that change configuration or are eliminated at the lower floors. It is difficult to 

provide the needed ductility at the weak story and often strength must be added. Completely 

discontinuous walls also create a load transfer deficiency for both overturning and shear. In such 

cases, collectors are often needed in the floor diaphragm, and supporting columns need axial 

strengthening. 

Load Path 

A common deficiency in this building is weakness in the load path from floor to walls, either 

collector weaknesses or shear transfer weakness immediately at the floor wall interface. Local 

transfer can be strengthened by adding concrete or steel corbel elements, dowels, or 

combinations of these components. As indicated above, discontinuous walls also often create 

load path deficiencies from the wall into the diaphragm at the discontinuity. 

Component Detailing 

The most common detailing problem in this building type is an imbalance of shear and flexural 

strength in the walls, leading to pre-emptive shear failure. This condition may be shown to be 

acceptable with small displacement demands. Walls can be strengthened in shear with overlays 

of concrete, steel, or FRP. 

The layout of walls often forces coupling between walls through the slab system or across 

headers of vertically aligned doors. These coupling components are seldom designed for the 

coupling distortions that they will undergo, particularly in older buildings. Short lengths of slabs 

between adjacent walls receive damage by coupling action that could compromise the gravity 

capacity. It is difficult to add strength or ductility to these slab areas, but vertical support at 

support points can be supplemented by corbels of steel or concrete. Damage to headers over 

doors often does not contribute to deterioration of overall response and can sometimes be 

acceptable. Local areas of wall can also be strengthened by overlays of concrete, steel, or FRP. 

Diaphragm Deficiencies 

Precast floor systems used in this building type often provide inadequate diaphragm behavior 

that could lead to bearing failures at the floor wall interface, particularly when no topping slab is 

present. Some topping slabs used primarily for leveling and smoothing the floor are 

inadequately tied to the precast elements or the walls, and are too thin or poorly reinforced to act 

as diaphragms on their own. See Chapter 22. 
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Foundation Deficiencies 

This building type often places large demands on the foundation system. If rocking is shown to 

be a controlling displacement fuse for the building, the foundations must be investigated to 

assure that these displacements can safely occur. See Chapter 23. 

20.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

Most significant recommendations listed in Table 20.3-1 are the same as techniques used in the 

similar building type, C2b, Concrete Shear Walls (Bearing Wall Systems) or general techniques 

applied to concrete diaphragms. Details concerning these techniques can be found in other 

chapters. 

20.5 References 

FEMA, 2000, Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 

356, Federal Emergency Management Agency, Washington, D.C., November. 
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Chapter 21 - Building Type URM: Unreinforced Masonry 

Bearing Walls 

21.1 Description of the Model Building Type 

Building Type URM consists of unreinforced masonry bearing walls, usually at the perimeter 

and usually brick masonry. The floors are typically of wood joists and wood sheathing 

supported on the walls and on interior post and beam construction. This building type is 

common throughout the United States and was built for a wide variety of uses, from one-story 

commercial or industrial occupancies to multistory warehouses to mid-rise hotels. It has 

consistently performed poorly in earthquakes. The most common failure is an outward collapse 

of the exterior walls caused by loss of lateral support due to separation of the walls from the floor 

and roof diaphragms. Figure 21.1-1 shows an example of this building type. 

Building Type URMA is similar to the Building Type URM, but the floors and roof are 

constructed of materials that form a rigid diaphragm, usually concrete slabs or steel joists with 

flat-arched unreinforced masonry spanning between the joists. Building Type URMA is not 

covered by this document. 

Figure 21.1-1: Building Type URM: Unreinforced Masonry Bearing Walls 

Masonry Wall Materials 

FEMA 306 (FEMA, 1999a) provides an overview of masonry wall material variables. It is 

paraphrased here. Unreinforced masonry is one of the oldest and most diverse building 
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materials. Important material variables include masonry unit type, wall construction type, and 

material properties of various constituents. 

Solid clay-brick unit masonry is the most common type of masonry unit, but there are a number 

of other common types, such as hollow clay brick, structural clay tile, concrete masonry, stone 

masonry, and adobe. Hollow clay tile (HCT) is a more common term for some types of 

structural clay tile. Concrete masonry units (CMU) can be ungrouted, partially grouted, or fully 

grouted. Stone masonry can be made from any type of stone, but sandstone, limestone, and 

granite are common. Other stones common in a local area are used as well. Sometimes 

materials are combined, such as brick facing over CMU backing, or stone facing over a brick 

backing. 

Wall construction patterns also vary widely, with bond patterns ranging from common running 

bond in brick to random ashlar patterns in stone masonry to stacked bond in CMU buildings. 

The variety of solid brick bond patterns is extensive. Key differences include the extent of 

header courses, whether collar joints are filled, whether cavity-wall construction was used, and 

the nature of the ties between the facing and backing wythes. In the United States, for example, 

typical running-bond brick masonry includes header courses interspersed by about five to six 

stretcher courses. Header courses help tie the wall together and allow it to behave in a more 

monolithic fashion for both in-plane and out-of-plane demands. The 1997 UCBC (ICBO, 1997) 

and 2003 IEBC (ICC, 2003) have specific prescriptive requirements on the percentage, spacing, 

and depth of headers. Facing wythes not meeting these requirements must be considered as 

veneer and are therefore not used to determine the effective thickness of the wall. Veneer 

wythes must be tied back to the backing to help prevent out-of-plane separation and falling 

hazards. Although bed and head joints are routinely filled with mortar, the extent of collar-joint 

fill varies widely. Completely filled collar joints with metal ties between wythes help the wall to 

behave in a more monolithic fashion for out-of-plane demands. One form of construction where 

interior vertical joints are deliberately not filled is cavity-wall construction. Used in many 

northeastern United States buildings, the cavity helps provide an insulating layer and a means of 

dissipating moisture. The cavity, however, reduces the out-of-plane seismic capacity of the wall. 

Material properties—such as compressive, tensile and shear strengths and compressive, and 

tensile and shear moduli—vary widely among masonry units, brick and mortar. An important 

issue for in-plane capacity is the relative strength of masonry and mortar. Older mortars 

typically used a lime/sand mix and are usually weaker than the masonry units. With time, 

cement was added to the mix and mortars became stronger. When mortars are stronger than the 

masonry, strength may be enhanced, but brittle cracking through the masonry units may be more 

likely to occur, resulting in lower deformation capacity. 

Given the wide range of masonry units, construction and material properties, developing 

comprehensive mitigation techniques for all permutations is not practical. The rehabilitation 

measures in this document are most directly relevant to solid clay brick masonry laid in running 

bond with a typical spacing of header courses. 

For additional general background on URM materials, see ABK (1981a), FEMA 274 (FEMA, 

1997b), FEMA 307 (FEMA, 1999b), and Rutherford and Chekene (1997). 
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Floor and Roof Diaphragms 

Building Type URM, by definition, is built with diaphragms that are considered, relative to the 

masonry walls, to be flexible. Typically, wood sheathing is attached to wood joists. Several 

types of wood diaphragm construction are common, including: 

Roofs with straight sheathing and roofing applied directly to the sheathing 

Roofs with diagonal sheathing and roofing applied directly to the sheathing 

Floors with straight tongue-and-groove flooring 

Floors with straight tongue-and-groove flooring over straight sheathing 

Floors with finished flooring over diagonal sheathing 

Wood structural panel overlays may have been added as part of past renovation work, or there 

may be additional layers of sheathing materials. In some buildings with heavy live loads, like 

warehouses, 2x or 3x decking may have been used to span between joists. 

Foundations 

Foundations for URM buildings typically are spread footings at interior columns and strip 

footings under masonry bearing walls. Footings are typically either brick or concrete, though 

stone might be found under older walls, particularly if stone masonry was used in the walls. 

21.2 Seismic Response Characteristics 

In many building types, the horizontal diaphragms are more rigid than the vertical elements of 

the lateral force-resisting system. Such buildings are often thought of as lumped mass systems 

with the weight tributary to each diaphragm level lumped along a vertical cantilever with 

dynamic properties dependent on the stiffness of the vertical lateral force-resisting elements. 

Ground motion input at the base is dynamically amplified up the cantilever, increasing at each 

floor level. Each point within a floor has a similar acceleration. 

Building Type URM, by contrast, has flexible diaphragms and stiff walls. Beginning with the 

ABK research program in the 1980s (see ABK, 1981a,b,c; and ABK, 1984), a different dynamic 

model was formulated for URM buildings. The ABK model assumes that there is relatively little 

dynamic amplification between the base and the top of the URM walls in the direction parallel to 

input motion. Significant amplification instead occurs at the midspan of the flexible diaphragms 

as they are driven by in-plane motion of the end walls. This generates large out-of-plane forces 

on the connections between the diaphragm and the coupled masonry walls. In some cases, the 

diaphragm may yield, limiting the forces that can be transmitted to the in-plane walls. If interior 

partitions are connected to the partitions, the deformation, cracking damage and resulting energy 

dissipation can help limit the movement of the diaphragms. Such existing partitions or newly 

added partitions or moment frames are termed “crosswalls”. 

The ABK program identified two modes of behavior for in-plane loading on the piers in the 

unreinforced masonry walls: shear-critical behavior and rocking-critical behavior. Each of these 

modes of behavior could be found acceptable if demands were below the capacity. These two 

modes are included in the UCBC as well. FEMA 273 (FEMA, 1997a) and then FEMA 356 

(FEMA, 2000) expanded the characterization of in-plane behavioral modes into the more brittle 

force-controlled modes (toe crushing and diagonal tension) and the more ductile deformation-
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controlled modes (bed joint sliding and rocking). FEMA 306 and FEMA 307 identify a number 

of other in-plane modes and sequences of modes. 

The ABK program also established criteria for determining the acceptability of out-of-plane 

resistance of the unreinforced masonry walls. Important variables are the height-to-thickness 

(h/t) ratio, presence of crosswalls, and overburden (axial compression) pressure on the walls. In 

short, a stocky lower story wall with overburden pressure from floors and walls above that is 

driven by a diaphragm damped by crosswalls is less likely to buckle and fail out-of-plane. 

It is important to recognize that URM buildings vary substantially in structural layout and 

characteristics, and this can have a significant effect on seismic response. Fairly rectangular 

multistory residential, office, and commercial buildings often have relatively low story heights 

and many partitions that can serve as crosswalls. Walls adjacent to other buildings will usually 

be relatively solid. These buildings typically perform much better than structures like churches, 

which can have irregular plans, re-entrant corners, tall story heights, heavy walls, offset roofs, 

few partitions, and many windows. Churches can also be some of the most expensive structures 

to rehabilitate. 

21.3 Common Seismic Deficiencies and Applicable Rehabilitation 

Techniques 

URM buildings are generally considered to be one of the most hazardous building types. 

Significant property damage and loss of life have occurred in URM buildings during earthquakes 

around the world and in the United States. The primary deficiencies are due to unbraced 

parapets which can fall on adjacent pedestrian thoroughfares and poorly connected walls and 

diaphragms which can lead wall failure and loss of vertical support for diaphragms. See below 

for general discussion and Table 21.3-1 for a more detailed compilation of common seismic 

deficiencies and rehabilitation techniques for Building Type URM. 

Global Strength 

As shear wall buildings, global strength in URM buildings is dependent on the in-plane shear 

capacity of the walls. Relatively large seismic forces are needed to lead to life safety concerns 

with in-plane wall behavior, though cracking damage will occur in relatively moderate events. 

When walls are found to be deficient, new vertical lateral force-resisting elements can be added 

at interior locations or existing walls can be enhanced. At interior locations, new elements 

include wood structural panel shear walls, concrete shear walls, reinforced masonry shear walls, 

braced frames, and moment frames. At exterior locations, care must be taken to address relative 

rigidity concerns. Typically, concrete or shotcrete overlays are used to enhance the URM wall 

capacity. When the wall is highly punctured, braced frames or moment frames may be a viable 

option. The use of wood structural panel shear walls in buildings with masonry walls is 

permitted in new construction only in limited situations, such as one-story or two-story buildings 

with low story heights and no use of diaphragm rotation to resist loads, due to concerns about 

wood flexibility. Rehabilitation standards such as the 1997 UCBC and 2003 IEBC relax these 

restrictions significantly, though they do not permit the wood shear walls to resist lateral forces 

with other materials along the same line of resistance or when there are rigid diaphragms. Use of 

wood structural panel shear walls in rehabilitating masonry buildings should be carefully 

considered. 
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Table 21.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for URM Bearing W

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce D

Global Strength 

Insufficient inplane wall strength 

Wood structural panel shear wall [6.4.2], [5.4.1] Concrete/masonry shear wall [21.4.8] Steel braced frame [7.4.1] Steel moment frame [21.4.9] 

Concrete wall overlay [21.4.5] Fiber composite wall overlay [21.4.6] Grouting Infill openings [21.4.7] 

Seismic isolation 

Global Stiffness Configuration Soft story, weak 

story, excessive torsion 

Wood structural panel shear wall [6.4.2] Concrete/masonry shear wall [21.4.8] Steel braced frame [7.4.1] Steel moment frame [21.4.9] 

Load Path Inadequate or 

missing wall-todiaphragm tie 

Tension anchors [21.4.2] 

Shear anchors [21.4.2] 

Cross-ties and subdiaphragms [22.2.3] 

Supplemental vertical supports [21.4.11] 

 Missing collector Add collector [7.4.2] 
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Table 21.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for URM Bearing W

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce D

Component Detailing 

Wall inadequate for out-of-plane bending 

Exposed interfloor wall supports [21.4.3] Reinforced cores [21.4.4] Concrete wall overlay [21.4.5] Fiber composite wall overlay [21.4.6] 

 Undesirable wall 

in-plane behavior mode 

Sawcutting to change shear mode to rocking mode 

 Unbraced parapet Brace parapet [21.4.1]  Unbraced 

chimney 

Brace chimney [5.4.6] Infill chimney [5.4.6] 

Reduce chimney [5.4.6] 

 Poorly anchored 

veneer or appendages 

Add ties [21.4.12] 

Diaphragms Inadequate 

in-plane strength and/or stiffness 

Add horizontal braced frame [22.2.9] Wood structural panel shear wall [6.4.2] Concrete/masonry shear wall [21.4.8] Steel braced frame [7.4.1] Steel moment frame [21.4.9] Wood structural panel or steel moment frame crosswall [21.4.10] 

Enhance existing diaphragm [22.2.1] Enhance woodframe crosswall [21.4.10] 
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Table 21.3-1: Seismic Deficiencies and Potential Rehabilitation Techniques for URM Bearing W

Deficiency Rehabilitation Technique 

Category Deficiency Add New Elements 

Enhance Existing 

Elements 

Improve Connections 

Between Elements Reduce D

 Inadequate chord 

capacity 

Add steel strap or angle 

Diaphragms (continued) 

Excessive stresses at openings and irregularities 

Add wood or steel strap reinforcement 

 Re-entrant corner Wood structural panel 

shear wall [6.4.2] Concrete/masonry shear wall [21.4.8] Steel braced frame [7.4.1] Steel moment frame [21.4.9] 

Collector [7.4.2] 

Foundation See Chapter 23 [ ] Numbers noted in brackets refer to sections containing detailed descriptions of rehabilitation techniques. 
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Global Stiffness 

URM bearing walls are generally quite rigid. When walls are solid or lightly punctured with 

window openings, global stiffness deficiencies are typically not an issue. In some buildings, 

though, facades facing the street can be highly punctured with relatively narrow piers between 

openings. In addition to lacking adequate strength, these wall lines may also be too flexible as 

well. 

Configuration 

Many commercial URM buildings will have a fairly open street façade at the ground level, 

leading to a weak and soft first story and torsional irregularities. This is usually addressed by the 

addition of a moment frame at the façade or another vertical lateral force-resisting element at 

some distance back from the façade. 

Load Path 

As noted above, it is the lack of adequate ties between the walls and diaphragms that is the single 

most significant deficiency in URM buildings. Rehabilitation measures include tension ties for 

out-of-plane forces and shear ties for in-plane forces. Bond beams are often employed for 

connections where the roof runs over the top of the walls. As a back-up vertical support system, 

supplemental vertical supports are added under trusses or girders where large gravity loads are 

concentrated on the wall in case the masonry is damaged locally. 

Component Detailing 

Since the masonry elements in Building Type URM are unreinforced by definition, they do not 

comply with modern ductile detailing requirements. Walls deemed susceptible to out-of-plane 

bending failures can be strengthened by strongbacks placed against them either on the outside or 

more commonly on the interior face. When preservation of finishes is critical, reinforced cores 

can be drilled and installed within the wall. Parapet bracing and chimney bracing are common. 

In some buildings, the exterior brick wythe will not be anchored back to the backing wall with 

mechanical ties or sufficient headers, and veneer ties are installed. 

Diaphragm Deficiencies 

Wood diaphragms may lack both strength and stiffness in URM buildings. This can be 

addressed by adding new interior elements to cut the diaphragm span or by enhancing the 

diaphragm itself with wood structural panel overlays. In some cases, such as sloped roofs, new 

horizontal braced frame diaphragms are added, in lieu of strengthening the existing diaphragm. 

Foundation Deficiencies 

Foundation deficiencies for existing elements are relatively uncommon in URM buildings. 

Foundation rehabilitation work usually is focused on the support for new lateral force-resisting 

elements that are added to the superstructure. 
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21.4 Detailed Description of Techniques Primarily Associated with 

This Building Type 

21.4.1 Brace or Remove URM Parapet 

Deficiency Addressed by Rehabilitation Technique 

Past earthquakes have consistently shown that unreinforced masonry chimneys and parapets are 

the first elements to fail in earthquakes due to inadequate bending strength and ductility. 

Parapets tend to have greater damage at midspan of diaphragms due to higher accelerations and 

displacements from the oscillating diaphragm. 

Description of the Rehabilitation Technique 

URM parapets can be braced or removed to minimize the falling hazard risk. Bracing is usually 

done with a steel angle brace. The brace is anchored near the top of the parapet and to the roof. 

The existing roof framing may need localized strengthening to take the reaction from the brace. 

Roof-to-wall tension anchors are typically part of parapet bracing. See Figure 21.4.1-1 for an 

example of parapet bracing. If the top of the parapet is removed, the vertical compressive stress 

on roof-to-wall anchors is reduced, so removing the parapet is often combined with adding a 

concrete cap or bond beam as part of the roof-to-wall anchorage. See Figure 21.4.1-2 for an 

example of parapet removal and addition of a concrete cap beam. See Section 21.4.2 for more 

details on wall-to-diaphragm anchorage. 

Design Considerations 

Research basis: No references directly addressing testing of parapet bracing have been 

identified.

Parapet height: Codes such as the 2003 IEBC and 1997 UCBC provide maximum allowable 

height-to-thickness (h/t) ratios for parapets. The height is taken from the lower of the either the 

tension anchors or the roof sheathing. Requirements are more stringent in higher seismic zones. 

With the 1997 UCBC, for example, the h/t ratio in Seismic Zone 4 is 1.5, so for a typical 13” 

thick, three-wythe wall, parapets taller than 19.5” above the roof-to-wall anchors require removal 

or bracing. In the 2003 IEBC, locations with SD1 > 0.4g have the same requirements. 

Fire protection: The original purpose of extending the masonry wall up to form a parapet was to 

help limit the spread of fire between the wood roofs of adjacent buildings. Removing a parapet 

must be coordinated with local building code fire safety requirements. 

Guardrails: Parapets often serve as a guardrail around a roof. Removing a parapet must be 

coordinated with local building code life safety requirements. 

Load path: When parapet bracing/roof-to-wall tension anchorage is the only rehabilitation 

technique, the out-of-plane load path can be incomplete, particularly when the roof joists are 

perpendicular to the brace. A more complete rehabilitation strategy includes developing the 

parapet and tension anchorage forces back into the roof diaphragm through the use of 

subdiaphragms and cross-ties. 
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Figure 21.4.1-1: Parapet Bracing 

Detailing and Construction Considerations 

Parapet anchorage types: Drilled dowels connecting the top of the bracing to the masonry can 

be with through bolts or adhesive anchors. See Section 21.4.2 for detailed discussion of drilled 

dowels. 

Top angle: Figure 21.4.1-1 shows a continuous angle running between braces in the roof. This 

angle can be used to span between braces to reduce the number of bracing points. It also 

increases redundancy over a localized connection of the brace to the parapet. 

Load in the roof framing: The vertical reactions at the base of the brace are typically resisted by 

roof framing. In Figure 21.4.1-1 the added blocking beneath the base of brace workpoint helps 

to engage three joists in resisting vertical loads. Tall parapets can generate substantial brace 

forces that existing wood roof joists may not be able to resist. Additional joists can be added, or 

more braces can be used to distribute the load. Horizontal loads from the brace are distributed by 

the blocking and new wood structural panel. 
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Figure 21.4.1-2: Parapet Removal and Concrete Cap Beam 

Waterproofing at the roof: The brace anchor at the roof needs to attach to the structural framing 

members, so a penetration through the roof membrane will occur that needs proper 

waterproofing design by a qualified contractor or design professional. Often, parapet bracing 
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and roof-to-wall ties and even roof diaphragm sheathing rehabilitation activities are combined 

with roofing replacement given the cost effectiveness of combining the work. 

Cost and Disruption Considerations 

Adding parapet bracing and roof-to-wall tension anchors provide some of the most effective 

seismic rehabilitation for reducing life safety risks. As a result, some communities—such as San 

Francisco—passed parapet safety ordinances requiring mandatory mitigation many years ago. 

Disruption is typically relatively low since occupants can remain in place. Combining parapet 

bracing and roof-to-wall ties and even roof diaphragm sheathing rehabilitation activities with 

roofing replacement can significantly reduce the total cost of the work. Disruption can increase 

noticeably if the roof has to be removed for installation. 

Proprietary Issues 

There are no proprietary concerns with parapet bracing, other than use of proprietary anchors as 

part of the assemblage. See Section 21.4.2.

21.4.2 Add Wall-to-Diaphragm Ties 

Deficiency Addressed by Rehabilitation Technique 

Inadequate or missing shear and tension connections between the unreinforced masonry bearing 

wall and the wood floor or roof. 

Description of the Rehabilitation Technique 

The most significant deficiency in URM bearing wall buildings is the lack of an adequate 

positive (i.e. mechanical) tie between the masonry walls and the floor and roof diaphragms. Ties 

are usually separated into two categories: tension ties and shear ties. Tension ties transfer out-ofplane inertial loads perpendicular to the face of the masonry back into the diaphragm. Shear ties 

transfer loads from the diaphragm into the wall where they are resisted by in-plane action of the 

wall. Tension ties help keep the walls from falling away from the diaphragms; shear ties help 

keep the diaphragm from sliding along parallel to the wall. Ties are assemblages that consist of 

both the anchorage to the wall (shown in detail in Figures 21.4.2-1 and 21.4.2-2) and the 

anchorage back into the diaphragm (shown in the subsequent figures). 

Design Considerations 

Research basis: The focus of wall-to-diaphragm testing to date has been on the anchorage to the 

masonry and has been done primarily by manufacturers. Paquette, Bruneau and Brzev (2003) 

tested a specimen of a small full-scale one-story building with roof-to-wall ties, but the focus of 

the work was on wall and diaphragm response. 

Anchor types and capacities: The 1997 UCBC and 2003 IEBC provide prescriptive values for 

tension and shear bolts meeting certain requirements. These are for a 2-1/2” diameter hole filled 

with nonshrink grout approach that is typically no longer used. The ICBO and now ICC 

evaluation report process has standardized procedures for vendors supplying adhesive ties for use 

in brick masonry. Three installations are included in most vendors’ ICC Evaluation Service 

reports, and they have standardized installation techniques and capacities. Adapted versions of 

these installations are shown in Figures 21.4.2-1 and 21.4.2-2. Figure 21.4.2-1A shows a 
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Figure 21.4.2-1: Drilled Dowels 
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Figure 21.4.2-2: Through Bolt Anchor 

“combination” drilled dowel that can be used for resisting both tension and shear forces. It is 

drilled into the wall at a 22.5 degree angle from horizontal at least 13” into the wall. The angle 

allows the dowel to engage more courses of brick, theoretically improving the reliability. At the 

allowable stress design (ASD) force level, it is good for 1200 lbs in tension and 1000 lbs in 

shear. Figure 21.4.2-1B shows a drilled dowel used only for resisting shear forces. It goes in 8” 

deep into the masonry and is good for 1000 lbs at the ASD level. Figure 21.4.2-2 shows a 

special through bolt anchor using a steel sleeve in the first 8” that can take tension and shear and 

has the same values as the combination anchor. These ICC capacities are typically used in 

design; they come with a number of restrictions and requirements such as quality of masonry. 

When higher values are needed, proof testing can be undertaken. In the ICC standards for both 

shear and tension testing (ICC-ES, 2005) of adhesive anchors that manufacturers must use to 

obtain ICC qualification, allowable stress design capacities are the lower of prescriptive values 

and the average ultimate test value divided by a safety factors of 5. 

It should be appreciated that the prescriptive values in the UCBC, the IEBC, and ICC Evaluation 

Service reports are based on tests of the drilled dowel itself, not the full elements of the detail. 
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Capacities for nails, wood structural panels, bolts in wood and straps come from typical code 

provisions. 

Detailing and Construction Considerations 

There are many issues to consider in detailing for tension and shear ties. These include the 

following: 

Aesthetics: Anchors that go all the way through the wall have a visible bearing plate on the 

exterior face, such as shown in Figure 21.4.2-2. There are simple circular or octagonal plates 

that can be purchased or fabricated. Some manufacturers make plates with a countersunk hole 

and use flathead bolt heads to reduce the surface projection. When the exterior face is stucco, a 

plate with a countersunk hole can be recessed into the stucco or just into the masonry and 

refinished with stucco so it is hidden. Special cast anchors can be made if there is a desire to 

match an historic exposed cast iron anchor. When the anchor plate approach cannot be used, 

drilled dowels are used such as those shown in Figure 21.4.2-1. 

Nonshrink grout vs. chemical adhesive: Early ties used cementitious nonshrink grout. They 

required larger diameter holes (such as 2-1/2”) to be cored in the masonry to place the grout. A 

number of vendors have now created special chemical adhesives and tools that have optimized 

the process. Standard details use 3⁄4” diameter threaded rods in 1” diameter holes, though other 

sizes can be used, depending on manufacturer requirements. The typical installation approach is 

to drill the hole; clean it with a brush and compressed air; fill a nylon, carbon, or stainless steel 

screen tube (which looks like a test tube made out of wire mesh) with adhesive; place the screen 

tube into the hole; and then push the rod into the screen tube forcing the adhesive out of the tube 

into the annulus between the tube and the masonry. Figures 21.4.2-1 and 21.4.2-2 show the 

anchorage using chemical adhesives and screen tubes. 

Chemical adhesive types: There are many different types of chemical adhesives, though most 

are epoxy. Epoxy products have the longest track record. Some vendors have begun to produce 

other types of chemicals. Key issues when considering an adhesive are the length of time the 

adhesive has been in use, the extent and quality of the testing, the ability to bond to damp or 

water filled surfaces, setting time, cost, the heat deflection temperature (an ASTM test method 

for quantifying the loss of strength as ambient temperature rises), and the capacities shown by 

test results. Most modern adhesives use two-component pre-packaged assemblies, rather than 

bulk products used in the past. This reduces the risk of improper mixing and not developing the 

adhesive to its proper strength. When adhesives are curing, the off-gassing can be unpleasant, 

and proper ventilation procedures are necessary. 

Dowel material type: Threaded rod is commonly specified as ATSM A36 all-thread rod. It is a 

relatively ductile material, with a minimum yield strength of 36 ksi and ultimate strength of 60 

ksi. When higher strength material is needed (which is rare), ASTM A193, Grade B7 threaded 

rod can be used with a minimum yield strength of 105 ksi and ultimate strength of 125 ksi. 

Rebar can be used as well, but this is not typically done in ties that connect to the wood 

diaphragms since the threaded connection is needed. Threaded rod is sometimes supplied with 

oil on it. This must be solvent cleaned, so that proper bonding with adhesives can occur. 
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Access: Installation of ties can be done either from below the diaphragm or above. Figure 

21.4.2-3 shows installation of floor-to-wall tension ties from below. Figure 21.4.2-4 shows 

installation occurring from above the floor. Figure 21.4.2-5 shows installation of floor-to-wall 

shear dowels from above. Similar details are contained in Rutherford & Chekene (1990), 

SEAOSC (1982) and SEAOSC (1986). The choice of whether to install from above or below 

depends on whether there are finishes that need to be avoided, whether diaphragm strengthening 

is being done, and what type of diaphragm strengthening is planned. If there is a special plaster 

ceiling to be avoided, then access and installation would proceed from above. If there is no 

plaster ceiling and the floor or roof diaphragm is not being modified or is being enhanced by 

adding a wood structural panel overlay from above, then access and installation for walldiaphragm ties would be from below. Angled dowels (see section below) installed from below 

can be angled upwards rather than the typical downward angle, provided non-sag adhesives are 

used. 

Joist direction: Framing in most buildings is orthogonal so that joists or rafters are either 

perpendicular or parallel to the in-plane direction of the wall. Installations where the joists are 

perpendicular to the wall are easier to make; installations where the joists are parallel involve 

blocking and more complicated details. Figures 21.4.2-3 to 21.4.2-5 show variations for joist 

orientation. 

Special issues at the top of the wall: In most URM buildings, the wall continues up past the roof 

forming a parapet that provides fire protection and serves as a guardrail during roof maintenance, 

as described in Section 24.4.1. In some buildings, though, the roof continues over the top of the 

wall. In these situations, the roof might be relatively flat or sloped. As a result, special issues 

arise. First, there is reduced overburden pressure at the top of the wall, reducing the reliability of 

drilled dowels. Second, eccentricities become more significant, such as the vertical eccentricity 

between the roof diaphragm and the top of the masonry. Making reliable connections between 

walls in these situations can be particularly challenging and is usually dependent on the specific 

geometry and characteristics of the existing details. A common strategy is to employ a concrete 

bond beam at the top of the wall. This ties the wall together, serves as a collector and chord, 

increases redundancy and often simplifies details. Figure 21.4.2-6 shows a bond beam placed on 

top of an existing wall under the roof framing. This is possible when the wall is wide, and there 

is sufficient distance between the masonry and rafter. Figure 21.4.2-7 shows an alternative when 

there is insufficient clearance between the rafter and top of wall that involves removing the top 

two courses of masonry to gain room for the bond beam. 

Eccentricity: It is desirable to minimize the eccentricities in a connection. Figure 21.4.2-8 

illustrates the issue and some alternate approaches with floor-to-wall tension ties. Figure 21.4.2-

8A shows a common tension tie detail in plan view where a tie-down anchor is connected to the 

side of an existing joist. The plan offset between the drilled dowel at the center of the tie-down 

where load is applied and the center of the joist where it is resisted times the force is a moment 

that must be resisted by the joist in weak way bending. This stress can be quite significant. 

Figure 21.4.2-8B shows an alternative where two tie-downs are used to make a connection that is 
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Figure 21.4.2-3: Tension Anchors Installed from Below the Floor 
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Figure 21.4.2-4: Tension Anchors Installed from Above the Floor 
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Figure 21.4.2-5: Floor-to-Wall Shear Anchors 
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Figure 21.4.2-6: Bond Beam at a Sloping Roof 

Figure 21.4.2-7: Bond Beam at a Sloping Roof with Limited Clearance 
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Figure 21.4.2-8: Tension Tie Connection Issues 
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more concentric. This detail, however, puts a large number of screws into the existing joist, so a 

sistered joist is shown. Adding the sister also permits the nailing into the diaphragm to be into 

each joist, reducing the nailing demand on the joists. Bolted tie-downs, instead of tie-downs 

with screws, can be used with through bolts placed in double shear. Traditional bolted tie-downs 

have greater slip than the more recent tie-downs using screws. There are proprietary connectors 

using tubes as tie-downs on each side without oversize holes that bolt eliminate eccentricity and 

reduce bolt slip. Both Figure 21.4.2-8A and 21.4.2-8B have dowels adjacent to the joist. This 

means the dowel will enter the wall next to or in the weakened area of joist pocket and at the end 

of new blocking used for shear transfer, where there is insufficient end distance to use the dowel 

as a shear tie. Figure 21.4.2-8C shows a V-strap detail where the drilled dowel is placed between 

joists, away from the joist pocket and with plenty of end distance. When the strap is in tension, 

forces perpendicular to the joists are produced that are resisted by the added blocking and plate 

washers. 

Truss anchorage: In some URM buildings, there will be large gravity elements that bear on the 

wall, such as girders or trusses. These also become concentrated points of stiffness in the 

diaphragm. Since the relative rigidity of the elements cannot be easily quantified, it is usually 

prudent to use an enveloping or “belt and suspenders” approach of assigning demand, so that 

typical anchors between trusses take the uniform load and the ties connecting the wall and 

trusses take additional load. 

New ties vs. reuse of existing ties: In many older URM buildings, there are existing ties called 

government or “dog” anchors. These anchors typically only occur in the direction where the 

joists are perpendicular to the face of the wall, and they may not be at sufficient spacing. The 

1997 UCBC and 2003 IEBC permit use of these anchors as wall-to-diaphragm tension anchors if 

tested in accordance with certain standards and capacities are sufficient. 

Dowel spacing and edge distance: The 1997 UCBC and 2003 IEBC have maximum spacing 

requirements on shear and tension dowels. When walls become thick, the out-of-plane demands 

and the relatively low ICC Evaluation Service report capacity values can lead to fairly tight 

spacing of dowels. The UCBC and IEBC do not have minimum spacing requirements. From a 

practical point of view, dowels should not be placed closer than 12” o.c. Some ICC reports 

provide minimum spacing limits as well, like those commonly employed for drilled dowels in 

concrete. For one vendor, these spacing limits are 16” o.c. in the horizontal and vertical 

direction, and there is 16”minimum for edge distance as well. 

Corrosion considerations: Drilled dowels are typically installed from the interior. The masonry 

cover and epoxy serve as corrosion protection, so mild steel anchors are typically considered 

sufficient. For increased corrosion protection, stainless steel dowels and screen tubes can be 

used. When through bolted connections are installed, there is a more direct path for moisture 

intrusion. The anchor plate can be painted with exterior grade paint, galvanized or be made from 

stainless steel, and the through bolt can be made from stainless steel as well. 
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Screen tubes: The purpose of the screen tube is to prevent loss of epoxy into cracks or unfilled 

collar joint voids within the wall. Screen tubes vary somewhat from vendor to vendor and 

should be considered part of the manufacturer’s assembly. Nylon screen tubes have begun to be 

supplied by many vendors as they are more economical than stainless steel and more corrosion 

resistant than carbon steel. They do have a much larger coefficient of thermal expansion than 

both steel screen tubes and masonry. 

Hollow masonry: Anchorage of hollow clay tile, ungrouted concrete masonry units and other 

hollow masonry systems to diaphragms is particularly challenging. When forces are large, 

grouting in the region of the anchor is usually required. When forces are small, use of screen 

tubes may be acceptable. The screen tube is filled with adhesive, inserted into the wall and as 

the dowel is pushed into it, the adhesive seeps through the screen tube forming a key behind the 

face shell of the masonry. Capacities are small and the connection is nonductile. This type of 

connection may be viable for out-of-plane wall strengthening (see Section 21.4.3) where the 

demands are lower, but it is not recommended for wall-to-diaphragm connections. Figure 

21.4.2-9 shows a method of connecting a floor to an ungrouted CMU wall. Even in ungrouted 

CMU, a grouted bond beam is usually found beneath the floor, and it helps provide bearing 

support for the floor joists. Figure 21.4.2-9 involves locally grouting the courses at and just 

above the floor to install a new anchor. Figure 21.4.2-10 shows an alternative that avoids 

working from above and uses the existing bond beam. Sistering and a nailer help get the new 

anchor to the proper elevation. If a grouted bond beam is not present, it may be necessary to 

create one to make the proper anchorage, similar to the top courses in Figure 21.4.2-9. 

Drilling: Holes need to be drilled with a rotary drill or a rotohammer drill with the percussion 

setting turned off to limit vibration into the wall. This can slow drilling significantly. In some 

cases, coring with a diamond tipped blade is more efficient. This may be the only way some 

hard masonry, like granite, can be drilled. Sometimes water is used to cool the bit, and the slurry 

produced by the water, mortar and masonry can stain the face of the wall. 

Cost/Disruption 

Considerations for cost depend on the number, type and depth of dowels; the difficulty of access; 

and the extent of finishes that are impacted. Through bolts are usually less expensive than 

adhesive anchors. 

Drilling is loud and can be disruptive to occupants. Typically, either the floor or ceiling has to 

be removed to install the dowels. Thus, it is usually not practical to install dowels in occupied 

rooms, though the work can be phased by building area so disruption is minimized.

Proprietary Issues 

Values for anchor capacity come from individual vendors, but there are no known concerns with 

use of a properly procured product. 
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Figure 21.4.2-9: Wall-to-Floor Tension Tie in Hollow Masonry 
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Figure 21.4.2-10: Wall-to-Floor Tension Tie in Hollow Masonry Alternate 

21.4.3 Add Out-of-Plane Bracing for URM Walls 

Deficiency Addressed by Rehabilitation Technique 

Inadequate out-of-plane bending resistance of an unreinforced masonry wall. 

Description of the Rehabilitation Technique 

Two types of bracing can be used: diagonal braces that reduce the effective height of the 

masonry wall (Figure 21.4.3-1A) and vertical braces or strongbacks that span the full height of 

the inside face of the wall (Figure 21.4.3-1B). Vertical braces can be surface mounted or, when 

aesthetic considerations are paramount, recessed into the wall; see Figure 21.4.3-2. 

Design Considerations 

Research basis: The most comprehensive set of testing done to date on out-of-plane response of 

URM walls was part of the ABK research program in the 1980s, and it is documented in ABK 

(1981c). Full-scale, dynamic testing of 20 wall specimens was conducted. Specimens were 6’ 

wide, 10’ to 16’ tall, and had height-to-thickness (h/t) ratios that varied from 14 to 25. 

Superimposed axial loads were varied; and materials included brick, grouted CMU, and 

ungrouted CMU.

H/t limits: It is tall, narrow walls that have been found to be susceptible to out-of-plane wall 

demands. The 1997 UCBC and 2003 IEBC provide maximum h/t requirements. Walls with 

larger h/t ratios must be braced.

Spacing: For strongbacks, such as shown in Figures 21.4.3-1B and 21.4.3-2A, the maximum 

spacing requirements are set by the 1997 UCBC or 2003 IEBC at the minimum of 10 feet or half 

the unsupported height of the wall. For diagonal braces, the maximum spacing is set at 6 feet. 
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Figure 21.4.3-1: Exposed Out-of-Plane Wall Bracing 

Stiffness: For strongbacks, such as shown in Figures 21.4.3-1B and 21.4.3-2A, the 1997 UCBC 

limits deflection of the wall at ASD demands to one tenth of the wall thickness. This is not a 

particularly stringent requirement. Say that the first story of a multistory building in Seismic 

Zone 4 is 13” thick and 18’ tall and its resulting h/t ratio of 16.7 exceeds the h/t limit of 16 in the 

UCBC. Bracing would be need to be stiff enough to keep deflections down to 10% of 13” or 

1.3”. This is L/166, which is comparatively low to most masonry design requirements, which 

are typically L/360 or higher, up to even L/600. Kariotis (1982) notes that the goal of a flexible 

vertical brace is to keep the brace elastic and provide a predictable restoring force during cracked 

excursions of the masonry wall. For diagonal braces, the UCBC encourages detailing to 

minimize vertical deflections. 

Diagonal braces loading vs. bracing the wall: If the roof deflects downward on a diagonal brace, 

a horizontal reaction is imparted to the wall. One concern with diagonal braces is that vertical 

vibration of the roof in an earthquake can contribute to the out-of-plane inertial forces on the 

wall. This concern, combined with the difficulty of making the roof stiff enough for against 

vertical deflections, makes vertical bracing a preferred engineering choice over diagonal bracing. 
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Figure 21.4.3-2: Vertical Bracing Alternatives 

Recessed steel and concrete and surface-mounted concrete: Provisions in the 1997 UCBC and 

2003 IEBC do not explicitly consider the approaches shown in Figures 21.4.3-2B, 21.4.3-2C and 

21.4.3-2D. These approaches are unusual, but they can be used when a more sensitive aesthetic 

approach or higher loads are needed. 
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Detailing and Construction Considerations 

Materials: Braces are typically done with steel as shown in Figures 21.4.3-1 and 21.4.3-2A, but 

strongbacks can also be done with wood posts or with concrete pilasters (Figure 21.4.3-2C). 

Aesthetics: Figure 21.4.3-1 shows exposed braces. This is the least expensive approach and is 

appropriate for certain occupancies. When there is architectural desire to hide the steel, the 

bracing can be furred at added cost and impact on the usable space. To minimize the impact on 

the space, the vertical brace can recessed into a cavity cut in the wall with either a steel or a 

concrete member. See Figure 21.4.3-2. Recessing the steel or concrete requires significantly 

more work and raises the potential for cracking to propagate from the inside of the recess to the 

masonry face. 

Strongback anchor spacing: Figure 21.4.3-1B shows only a central anchor at midheight of the 

wall. Often demand/capacity ratios for anchorage to the wall with through bolts or drilled 

dowels (see Section 21.4.2) will dictate a tighter spacing of anchors. 

Floor/roof framing capacity: Figure 21.4.3-1 shows anchorage to joists oriented perpendicular 

to the wall. When joists are parallel to the wall, the horizontal anchorage force must be 

developed out into the diaphragm. In Figure 21.4.3-1A, the existing roof beams may need to be 

strengthened to provide adequate strength to resist downward loading. 

Hollow masonry: Figures 21.4.3-1 and 21.4.3-2 apply to solid masonry. When the existing 

masonry is hollow, alternative connection methods are needed. Figure 21.4.3-3 shows use of 

vertical concrete ribs. A chase is created by removing the face shell on one side of the wall. 

Reinforcing steel is added and then grout or concrete fill. There is typically insufficient space 

for ties. This approach is messy and noisy. Figure 21.4.3-4 shows an alternative where steel 

strongbacks are bolted to the wall with either drilled dowels or through bolts. The screen tube 

anchor of Figure 21.4.3-4A relies on mechanical keying action from the spreading adhesive to 

engage the face shell. The capacity is limited to the face shell of the masonry and can be quite 

low, in the low hundreds of pounds at allowable stress design levels. It is also nonductile as the 

failure mechanism is spalling of the face shell. The through bolt in Figure 21.4.3-4B provides 

increased capacity and locally grouting in the anchor provides additional capacity. 

Cost/Disruption 

Diagonal bracing is usually less expensive, but is considered less reliable than vertical bracing. 

Furring can be used to cover the braces at added cost. Exposed braces are typically less 

expensive than more architecturally sensitive alternatives like recessed vertical braces or 

reinforced cores (See Section 21.4.4). Installation of bracing is fairly disruptive since it must 

occur around the entire perimeter; and it involves drilled dowels, and accessing and connecting 

to horizontal diaphragms. 

Proprietary Issues 

There are no known proprietary concerns with bracing of URM walls. 
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Figure 21.4.3-3: Concrete Ribs in Hollow Masonry 
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Figure 21.4.3-4: Connection of Strongback to Hollow Masonry 
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21.4.4 Add Reinforced Cores to URM Walls 

Deficiencies Addressed by Rehabilitation Technique 

Inadequate out-of-plane capacity and in-plane capacity of unreinforced masonry wall. 

Description of the Rehabilitation Technique 

Installing the reinforced core involves drilling a core from the roof down the inside of an 

unreinforced masonry wall. A steel reinforcing bar and grout are placed inside the hole to 

increase the wall strength. See Figures 21.4.4-1A and 21.4.4-1B. This process is used to avoid 

the aesthetic impact of exposed bracing described in Section 21.4.3. 

Design Considerations 

Research basis: The original research at CSU Long Beach and North Carolina State University 

for reinforced cores is summarized in (Plecnik, Cousins, and O’Conner, 1986) and Plecnik 

(1988). It covered both out-of-plane and in-plane loading. Subsequent vendor tests for in-plane 

loading were done at UC Irvine but have not been published. More recent in-plane testing is 

summarized in Abrams and Lynch (2001). 

Out-of-plane capacity: When reinforced cores are used for enhancing out-of-plane bending 

capacity, the wall is analyzed as a reinforced masonry element. Some engineers have used a 

traditional allowable stress design code format for masonry. Another common approach is to use 

factored design methods. Plecnik, Cousins and O’Conner (1986) provided an ultimate strength 

design formulation. As with concrete design or typical reinforced masonry design, the 

compressive strength of the masonry, f’m, is needed. Default values are available in FEMA 356, 

but it is important not to over-reinforce the masonry section and cause a brittle failure of the 

masonry, so it is often prudent to obtain the masonry strength. 

In-plane capacity: Plecnik, Cousins and O’Conner (1986) tests showed significant increase in inplane loading from the reinforced cores, but a design methodology was not provided. Breiholtz 

(1987) suggested using the test results as well, but did not provide a complete design 

methodology. One design approach is simply to extrapolate the test results on a per lineal foot 

basis. Another is to consider the vertical bars as the tie element in a strut-and-tie methodology, 

with diagonal struts in the masonry connecting the cores. 

Post-tensioned masonry: Post-tensioned masonry has been used in a few instances. The goal of 

post-tensioning the bars can be to add compressive stress to the masonry wall to increase the 

effective shear stress, since unreinforced masonry shear capacity formulations, such as those in 

FEMA 356 or the ICBC, provided increased shear strength with higher compressive strength. 

Recent research by Rosenboom and Kowalsky (2004) provides some results from cyclic testing, 

but the specimens have central cavities filled with grout rather than solid multi-wythe brick, and 

design equations are not provided. 

Detailing and Construction Considerations 

Detailing and construction considerations for reinforced cores include the following. 
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Figure 21.4.4-1A: Reinforced Cores 
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Figure 21.4.4-1B: Plan Detail of Reinforced Core in Masonry Wall 

Wet vs. dry drilling: Traditionally, holes cored in masonry were done similarly to those in 

concrete, using diamond tipped coring bits cooled by water. The slurry created by the water and 

brick dust can lead to staining of sensitive surfaces. Reinforced cores gained popularity when 

drilling companies developed specialized drilling equipment that did not need water to cool the 

bit. In many cases, this involves coring drills that rotate quite slowly compared to traditional 

coring equipment. The material within the core comes out in cylindrical chunks or in small 

pieces of debris or dust that are vacuumed into debris containers. 

Reducing leaks: To minimize leaks during wet drilling and during grouting, loose masonry 

should be repaired and cracks repointed. To limit the extent of repointing, consideration can be 

given to monitoring the location of dust clouds that escape cracks during dry drilling and 

repointing those locations. 

Obstructions: Drilling progresses most rapidly if the masonry is neither too hard nor too soft and 

is relatively homogeneous. Encountering wood debris inside walls or metal veneer ties can slow 

or stop the drilling. 
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Angled drilling: Drilling is typically done from the top and straight down. Occasionally, special 

situations arise where angled drilling might be necessary. This can be done, but requires much 

greater skill from the driller. 

Hole diameter: Hole diameters are typically about four inches, but can range from three to six 

inches. 

Drilling tolerances: The wider and shorter the wall, the easier it is to drill because it provides 

better tolerance against drilling inaccuracy. Tolerances of about ±2” in reasonably tall walls are 

usually achievable. A simple way to check this is as follows. Say the hole is four inches in 

diameter. Attach a small penlight flashlight to the end of a string, creating a lighted plumb bob. 

Drop the plumb bob down from the center of the hole at the top. If it does not hit the side of the 

cored hole at any point on the way down, a two-inch tolerance has been met. 

Bar material type: The reinforcing bars used in reinforced cores are typically regular ASTM 

A615 mild steel, as they are protected by the grout. For increased corrosion protection, stainless 

steel or epoxy coated rebar can be used. If post-tensioned center coring is done, high strength 

ASTM A722 threaded bars can be used.

Bar size: Bar size depends on the demand/capacity ratios, but typically ranges from #5 to #8. 

Centralizers: In order to keep the rebar centered in the hole, a plastic centralizing wheel is used 

by some engineers. Others consider this an obstruction limiting the flow of grout. 

Grout type: The original research by Plecnik, Cousins and O’Conner (1986) evaluated several 

grouts and concluded that a formulation using polyester grout provided the best dispersion into 

the masonry. Polyester has some offgassing concerns from styrene vapors and requires the hole 

to be dry before installing the grout. It also does not have the long-term track record of other 

more widely used grout materials. Some engineers, as a result, use a high quality nonshrink 

cementitious grout. With cementitious grouts, the hole needs to be prewetted prior to grout 

placement. 

Grouting process: Center coring can be done in multistory buildings, so the depth of holes can 

get quite large. A tremie grouting technique can be used to assure placement of grout. A grout 

tube can be tied loosely to the bar and/or centralizer(s) as the bar is lowered into the hole. As 

grout is pumped into the hole, the tube is slowly withdrawn as the level of grout rises. 

Verification ports: Grout will leak into voids in the masonry. To confirm that the grout is rising 

in the core, horizontal holes can be drilled into the wall. When the grout reaches the port, the 

port is plugged, and the grout is allowed to continue to rise to the top of the core. 

Bottom of holes: Coring usually goes into the foundation. In a concrete foundation, there will 

not be any place for water used to cool the bit in wet coring to escape. The hole can be 

vacuumed out, or the core can be continued all the way down to the bottom of the foundation. 
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Post-tensioning: When reinforced cores are post-tensioned, several additional issues come into 

play. First, grouting is done in two stages. The first stage is in the foundation where the posttensioning is anchored. After the grout cures and the bar is stressed, the second stage of grouting 

occurs up to the top of the hole. Post-tensioning vendors provide proprietary anchorage 

hardware for the bar at the top of the wall. A concrete cap or bond beam may be necessary or 

desirable to distribute the load on the top of the wall to reduce the stress on the masonry. 

Access to top of wall: The top of the wall must be accessible to drilling equipment. Temporary 

scaffolding or work platforms will usually need to be erected adjacent to the hole. Bracing of 

drilling equipment back to the wall and a point in the roof is necessary to keep the drill plumb. 

Spacing: Reinforced core spacing will depend on demand/capacity ratios, but a minimum 

spacing of six to ten feet is desirable. 

Cost/Disruption 

Adding reinforced cores can be considerably more expensive than exposed bracing, so it is 

usually only performed in historically and architecturally sensitive buildings. Since the core is 

placed inside the wall, the disruption to interior and exterior faces is limited to sealing cracks, 

access to the roof to place the drilling equipment, and drilling noise and vibration. 

Proprietary Issues 

Research for reinforced cores is in the public domain. Some drilling contractors reportedly have 

patents on certain types of proprietary drills. Some of the terms used with the process have been 

trademarked by some of the first engineers to implement the technique. As a result, the generic 

term “reinforced core” is used in this document. 

21.4.5 Add Concrete Overlay to Masonry Wall 

Deficiency Addressed by Rehabilitation Technique 

Improving inadequate in-plane wall capacity is the primary purpose of a new concrete overlay, 

but the concrete can also improve inadequate out-of-plane bending capacity. 

Description of the Rehabilitation Technique 

New concrete is applied against an existing unreinforced masonry wall to increase the shear 

capacity of the wall. The new concrete is attached to the old wall with adhesive anchors and can 

either be cast-in-place concrete or sprayed-in-place. In rehabilitation work, sprayed concrete, 

known as shotcrete, is more commonly employed than cast-in-place construction, since the 

existing wall provides the back-side form. The thickness of the new concrete varies with 

strength requirements, but it is usually from four to 12 inches. See Figure 21.4.5-1. 

Design Considerations 

Research basis: A fair amount of research in the use of shotcrete overlays on masonry has been 

done. A summary is given in El Gawady, Lestuzzi and Badoux (2004). Early diagonal tension 

testing was done by Kahn (1984), and more recently static cyclic tests were done by Abrams and 

Lynch (2001). Kahn (1984) showed significant increases in strength from the shotcrete and that 

adding drilled dowels between the shotcrete and masonry or an epoxy bonding agent did not lead 
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Figure 21.4.5-1: Concrete or Shotcrete Wall Overlay 

to significant improvements. A saturated masonry surface was recommended. Testing by 

Abrams and Lynch (2001) aimed at increasing the shear capacity to lead to flexural yielding of 

the tension bars in the shotcrete. Strength increased by about a factor of 3, but displacement 

capacity did not increase. 

Design criteria: When a concrete overlay is used, there are several common force-based design 

approaches for the wall, due to the relatively high strength of the concrete compared to the 

masonry. One is to take 100% of the demand tributary to the strengthened wall line in the 

concrete overlay itself and ignore the masonry. While this may sound conservative, it can mean 

that the masonry will be significantly damaged before the concrete ever sees the majority of its 

design load. Another approach is to share the load, by relative rigidity, between the masonry and 

the concrete. When this is done, both the masonry and the concrete must be checked to confirm 

they are not overstressed. The most conservative approach is to use the overlay to resist 100% of 

the tributary load, but to also check that the masonry can resist the loads it will actually attract. 

Displacement-based design approaches inherently consider the relative rigidity of the concrete 

and masonry, but they are less commonly employed. 
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Discretely applied overlays: The URM walls to which the overlay is applied are typically 

punctured with window and door openings. It can be tempting to apply the concrete to wide 

piers. The comparatively high strength of the concrete means it can take high loads, but it is 

unlikely to have sufficient stiffness to actually attract the load it was intended to take. 

Eventually, though, when the masonry cracks, the load will find the concrete, but this can lead to 

significant cracking at the ends of the concrete in the masonry spandrels. See Figure 21.4.5-2A. 

This can be addressed by spreading out the influence of the overlay by using top and/or bottom 

spandrels or grade beams, such as shown in Figure 21.4.5-2B. Alternatively, drilled piers can be 

placed at the ends of the new walls to add stiffness, as in Figure 21.4.5-2C. Or most simply, a 

continuous overlay can be used, enabling a reduced thickness and consistent finish surface, as 

shown in Figure 21.4.5-2C. 

Collector load transfer pathways: With discretely applied overlays in the field of the masonry 

wall, the question arises of how the load in the floors and the masonry wall will reach the 

overlay. Some engineers ignore this issue and assume the masonry wall will serve as the 

collector. Others provide an explicit steel collector in the edge of the diaphragm or a bond beam 

on top of the wall. 

Out-of-plane load resistance: When the overlay is added to the wall, its additional inertial load 

must be considered in the seismic weight of the structure. Out-of-plane anchorage requirements 

at the overlays are thus larger, and wall-to-diaphragm tie spacing often decreases there. 

Detailing and Construction Considerations 

Detailing and construction considerations for concrete overlays include the following. 

Drilled dowel spacing: It is common practice to connect the overlay to the masonry wall with 

drilled dowels. The drilled dowels transfer shear between the two materials, and they also serve 

as out-of-plane ties for the masonry. Spacing of two feet to three feet on center is typical. Good 

detailing involves drawing an elevation and showing the location of dowels around openings 

since the nominal spacing will typically change there. 

New foundation at the base of wall: The base of the shotcrete can be set on the ledge of the 

existing footing if conditions permit or a new footing can be provided. The added load from the 

shotcrete and distribution of stresses on the existing footing must be considered. See Section 

23.6.2 for issues involved in adding a new footing next to an existing footing. 

Interface between the wall and diaphragms: Where the new wall meets the existing floors is 

usually the location where special consideration must be given to detailing. Conditions where 

the floor joists are parallel to the wall are the easiest to address. The first and/or second joists in 

from the masonry are removed to install the overlay, a ledger is placed back, and the floor 

sheathing run up to the ledger. If shotcrete is used, sufficient clearance must be provided to 

avoid shadowing during spraying. This should be checked as part of the preconstruction test 

panel. 
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Figure 21.4.5-2A: Potential Damage at Ends of Narrow Concrete Overlay 
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Figure 21.4.5-2B: Alternatives to Distribute Overturning Loads 

in Concrete Overlay 
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Figure 21.4.5-2C: Alternatives to Distribute Overturning Loads 

in Concrete Overlay 
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When the joists are perpendicular to the wall, they can be cast into the wall, but when this is 

done, the joists should be treated with preservative, an air gap provided on the top and sides of 

the joists and building paper placed on the bottom to minimize moisture entering the lumber. 

Special rebar detailing will be needed to transfer shear from above the floor to below the floor 

through the weakened area of the joists. Given these issues, the perpendicular joists are often 

headed off and supported off a ledger on the face of the new wall. This requires shoring, but 

simplifies the remaining construction and provides for a better wall. This approach is shown in 

Figure 21.4.5-1. 

Pre-wetting the masonry wall: Cast-in-place walls and, to a lesser extent, shotcrete walls have 

moisture in them during placement that will be absorbed by the masonry. Some engineers 

require wetting the masonry wall, just prior to placement. 

Curing considerations in existing building: Curing concrete emits moisture. If the building has 

finishes that are sensitive to moisture emission, precautions will need to be taken to protect the 

finishes. This can be particularly critical if shotcrete is used. Curing of the face of the concrete 

is either done with curing compound, continuous spraying or a moisture-retaining cover. 

Finding a curing compound that will later be acceptable for certain adhered finishes can be 

difficult. Continuous spraying, however, adds substantial moisture to the interior space. 

Efflorescence concerns from additives and moisture: Like all concrete, overlays can be 

susceptible to alkali salts leaching to the surface, usually leading to white streaks or spots. These 

stains can come from additives within the concrete or from salts within the masonry wall. Use of 

low-alkali concrete is recommended, and additives should be limited to those known not to lead 

to efflorescence. 

Protection of existing masonry substrate: The main advantage of shotcrete is that the existing 

masonry serves as the backside form and forming the front is unnecessary. In typical stone 

masonry and brick masonry situations, the wall will be adequate to serve as the backside form. 

When the wall is particularly thin or in poor condition, the contractor will have to take care to 

brace the masonry to resist the force of the applied shotcrete. 

Cost/Disruption 

Adding new concrete, particularly with shotcrete, can be quite disruptive. Where access is 

sufficient, shotcrete is typically chosen as it is less expensive than cast-in-place work which 

requires front-side formwork. 

Shotcrete: Placing shotcrete requires access for the hose and concrete truck and sufficient room 

(several feet) to spray the concrete. It is desirable to shoot downward, so scaffolding is needed at 

upper portions of walls to achieve the necessary angle. Spraying is noisy and very dusty. If an 

indoor wall is being shot, the room will usually be sealed off with plastic sheeting to control 

dust. During shooting, residue—known as rebound—forms at the base of the shoot and must be 

cleaned away so that it does not become part of the overlay. Protection against rebound on 

existing floor and wall surfaces is needed. 
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Cast-in-place concrete: Placing cast-in-place concrete also requires access for the hose and 

concrete truck. Less front-side access is needed than shotcrete, but a front-side form is required 

with the associated sawing and hammering noise of construction. Concrete placement is noisy, 

and in addition to workmen and concrete truck noise, there is the vibrator used to consolidate the 

concrete.

Proprietary Issues 

There are no known proprietary concerns with shotcrete or cast-in-place overlays on existing 

masonry walls. 

21.4.6 Add Fiber-Reinforced Polymer Overlay to Masonry Wall 

Deficiency Addressed by Rehabilitation Technique 

Improving inadequate in-plane wall strength is the primary purpose of a new fiber-reinforced 

polymer (FRP) overlay, but the overlay can also improve out-of-plane bending capacity. 

Description of the Rehabilitation Technique 

An FRP overlay, typically made of glass or carbon fibers in an adhesive matrix, is applied 

against an existing unreinforced masonry wall to increase the shear strength of the wall. The 

existing wall surface must be prepared to receive the new material, and after application the fiber 

composite must be protected against ultraviolet rays. 

Design Considerations 

Research basis: Research in fiber composites is extensive, but has primarily been focused on 

enhancement of concrete elements or reinforced concrete masonry. There is, nonetheless, a 

growing body of research for unreinforced masonry strengthening. A partial listing of some 

papers is given here. 

For out-of-plane strengthening of unreinforced masonry, tests include Reinhorn and Madan 

(1995a) on clay brick; Portland State University (1998) on hollow clay tile; Vandergrift, 

Gergely, and Young (2002) on hollow concrete masonry; Tumialan, Galati, Namboorimadathil, 

and Nanni (2002) on surface applied fiber reinforced bars to hollow concrete masonry; 

Tumialan, et al. (2002) on glass and aramid fiber reinforced polymer composites on both 

concrete and clay brick; Tumialan, Galati, and Nanni (2002) on situ field tests in an infill frame 

building being demolished of brick and clay tile walls strengthened using glass fiber strips; 

Ehsani, Saadatmanesh, and Velazquez-Dimas (1999); and Velazquez-Dimas, Ehsani, and 

Saadtmanesh (2000) on half-scale clay masonry. 

In-plane testing for clay brick masonry includes Reinhorn and Madan (1995b); Ehsani and 

Saadatmanesh (1996); Ehsani, Saadatmanesh, and Al-Saidy (1997); Haroun and Mosallam 

(2002); Senescu and Mosalam (2004). Elgwady, Lestuzzi and Badoux (2003) performed 

dynamic tests on slender and squat hollow clay masonry piers with and without aramid, glass, 

and carbon fiber composite overlays. Vandergrift, Gergely, and Young (2002) performed tests on 

half-scale hollow concrete masonry. Schwegler and Ketterborn (1996) discusses in-plane 

masonry strengthening with carbon fiber, including straps at various angles of orientation. 

Reinhorn and Madan (1995b) found about a 120% increase in strength from the composite and 
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some additional displacement capacity in a one-cycle reversed cyclic test. After the fiber 

ruptured, however, masonry cracks immediately widened in a brittle manner. Haroun and 

Mosallam (2002) found a minor increase (about 20%) in strength capacity and about a 30% 

increase in displacement capacity. Cracking in the masonry was the ultimate limit which 

occurred after the composite debonded. Senescu and Mosalam (2004) used monotonic diagonal 

tension tests with mixed results, some showing improvements in strength and displacement 

capacity, others actually showing reductions in strength and displacement capacity. Paquette, 

Bruneau, and Brzev (2004) and Paquette and Bruneau (2004) investigated strengthening or 

repairing a one-story building with rocking-critical piers using fiber composite chord strips at 

ends of piers. 

Design basis: There are no code guidelines or FEMA 356 provisions explicitly addressing FRP 

overlays on unreinforced masonry. Information can be obtained from manufacturer literature 

and can be used in conjunction with criteria in the ICC-ES interim standard (ICC-ES, 2003). 

They focus only on the design of the fiber itself, not the fiber and masonry combined 

performance. Velazquez-Dimas and Ehsani (2000) provide modeling and design 

recommendations for out-of-plane strengthening. 

Behavioral mode: It is important to understand the underlying governing behavioral mode of 

both the unstrengthened unreinforced masonry wall and the strengthened wall. Fiber composites 

have little ductility. Adding an FRP overlay to a rocking critical wall pier may be able to reduce 

cyclic degradation of the pier, but will not change the strength or behavior mode if the fiber is 

only applied to the pier. If the fiber crosses the top and bottom of the pier into the spandrel, in 

can inhibit or prevent formation of a rocking mode. Strength will be increased, but ductility will 

be reduced from that of a rocking-critical mode to that of a shear critical mode. 

Detailing and Construction Considerations 

Detailing and construction considerations for FRP overlays include the following. 

Surface preparation: The surface of the masonry needs to be cleaned of loose material and 

finishes that prevent proper adhesion. Sandblasting of the masonry is not usually necessary; a 

wire brush is used instead. See Figure 24.4.6-1. 

Complete overlay vs. strips: Both in research and in practice, both complete overlays over the 

full surface of the wall and use of strips are found. Vertical strips are used when only 

improvement to out-of-plane resistance is needed. Diagonal strips have been used to resist 

diagonal tension stresses from in-plane shear. 

One side or both sides of wall: Applying fiber to both sides of the wall improves performance, 

particularly for out-of-plane resistance, but testing has been performed with fiber on only one 

side. 

Continuity of fiber at top and bottom: Providing load transfer with the fiber can challenging, 

particularly at floor-to-wall interfaces. If the fiber is being used to resist out-of-plane loads and 

is transferring these loads back into the floor diaphragm, special details may be needed to turn 

the vertical fiber overlay into the horizontal diaphragm. Fiber cannot be bent at 90 degrees; 
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Figure 21.4.6-1: Fiber Composite Wall Overlay on URM Wall 
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rather a radius is needed. Sometimes, steel reinforcing plates are used to stiffen the turns. If the 

fiber is being used to transfer in-plane loads from one story to the next, continuity past the floor 

is needed and will require special details as does shear transfer out of the diaphragm into the wall 

Moisture barrier: Fiber composites are impermeable. If continuous overlays are used, moisture 

transmission through the masonry wall will be stopped at the fiber. Eventually, the concern 

would be the moisture would build up and begin to delaminate the fiber bond and lead to general 

building concerns with excessive moisture. 

Additional information: See Section 13.4.1 for more detailed discussion on FRP issues including: 

Composite makeup and application 

Mechanical properties 

Fiber and mechanical anchors 

Durability 

Constructability 

Cost/Disruption 

Fiber composites are relatively expensive as an application for masonry wall strengthening and 

have not seen significant use. Disruption comes from the sandblasting and surface preparation of 

the masonry wall, the fumes from the adhesives used in application of the fiber composite and 

removal and access requirements where the fiber transitions from story to story at the floor levels. 

Proprietary Issues 

Fiber composite materials are supplied by vendors. Capacities and design methods vary 

depending on the vendors. 

21.4.7 Infill Opening in a URM Wall 

Deficiency Addressed by Rehabilitation Technique 

Inadequate unreinforced masonry in-plane wall strength. 

Description of the Rehabilitation Technique 

Window and door openings are filled to increase the shear capacity and reduce the shear stresses 

on the unreinforced masonry wall. The opening is typically filled with concrete, reinforced 

concrete masonry units, or reinforced clay brick, rather than with unreinforced masonry due to 

code concerns with adding unreinforced masonry. To provide adequate shear transfer between 

the existing wall and the new infill, the interface can be toothed, but more typically, drilled 

dowels are used. See Figure 21.4.7-1. 
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Figure 21.4.7-1: Infilling an Opening in a URM Wall 

Design Considerations 

Research basis: Research that directly addresses testing of infilled openings has not been 

identified.

Capacity of infill: The typical infill materials are stronger and potentially stiffer than the 

surrounding masonry. It is typical, however, to consider the infilled wall as solid unreinforced 

masonry in determining shear capacity of the composite wall. Since the increase in shear 

capacity from the infill is often not substantial, it is done when only moderate increases in 

capacity are needed. 

Behavioral mode: It is important to understand the underlying governing behavioral mode of 

both the unstrengthened unreinforced masonry wall and the infilled wall. Infilling openings 

could change a rocking-critical wall line to a shear-critical wall line, which may be less 

desirable. 
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Detailing and Construction Considerations 

Aethestics: Infilling openings can obviously have a significant visual impact. Sometimes new 

concrete is used and is set back from the exterior face so that a window can be placed. Lighting 

can be added between the glazing and infill to mitigate the opacity of the infill. 

Cost/Disruption 

Infilling an opening is relatively inexpensive if no architectural treatment is done to the face. 

Disruption is also more localized compared to other in-plane wall strengthening methods like 

concrete and fiber reinforced polymer overlays. 

Noise will occur during drilling holes for drilled dowels and placing the infill.

Proprietary Issues 

There are no proprietary concerns with infilling masonry wall openings. 

21.4.8 Add Concrete or Masonry Shear Wall (Connected to a Wood Diaphragm) 

Deficiency Addressed by Rehabilitation Technique 

A new concrete or masonry wall provides additional global strength and stiffness, reduces 

demands on existing masonry walls and can reduce demands on diaphragms by cutting tributary 

spans. 

Description of the Rehabilitation Technique 

The new wall should be properly designed to meet current code detailing provisions. This 

section focuses on detailing at the interface between the new wall and the existing wood 

diaphragm. Figure 21.4.8-1 shows sample concepts both for joists parallel to the wall and joists 

perpendicular to the wall at an interior floor location; Figure 21.4.8-2 shows similar concepts at a 

roof. 

Design Considerations 

Research basis: Research specific to ledger connections between wood diaphragms and concrete 

or masonry walls has not been identified. 

Capacity: The connection can be designed either for the code level demands or to develop the 

diaphragm, depending on where inelastic action is intended to occur. 

Detailing and Construction Considerations 

Detailing and construction considerations for connecting a new wall to an existing wood 

diaphragm include the following. 

Masonry vs. concrete: Masonry is usually considered quicker to install and less expensive; 

concrete (or shotcrete) is stronger and stiffer and usually considered to have better earthquake 

performance. Making connections from concrete to wood diaphragms can be easier than with 

masonry. 
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Figure 21.4.8-1: Connecting a New Concrete Wall to an Existing Wood Floor 
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Figure 21.4.8-2: Connecting a Concrete Wall to an Existing Wood Roof at Interior 
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In-plane shear transfer: In Figure 21.4.8-1, shear transfer from the diaphragm to the wall goes 

from the diaphragm boundary nailing to the ledger and through the threaded rod into the wall. A 

tight fit on the rod and ledger is needed. The ledger should be dry dimensional lumber or glulam 

material to minimize vertical shrinkage of the ledger. When the wall is not as long as the 

diaphragm (a very common occurrence), a collector attachment into the wall will be needed. 

Figure 21.4.8-1A shows a steel angle with headed studs cast into the wall and diaphragm-tocollector connections using lag screws. The steel could go above or below the floor. When 

loads are relatively low, wood members such as the ledger can be used as the collector. 

Out-of-plane tension transfer: In Figure 21.4.8-1, tension transfer of wall loads goes into the tiedown anchor, into the blocking, through straps in the blocking to additional blocks as required 

and eventually back into the diaphragm. Alternatively, blocking for a bay or two can be placed 

on both sides and out-of-plane resistance accomplished by compression bearing on the 

diaphragm joists. 

Joist direction: When the wall can be fit in between existing joists, the amount of labor is 

reduced. When joists are perpendicular to the wall, the joists are typically headed off on each 

side of the wall to allow the wall to pass through. This requires temporary shoring of the floor 

around the wall. At the top of the wall, the wall can stop just under the joists and be blocked up 

to the diaphragm for shear transfer. 

Shotcrete vs. cast-in-place concrete: See Section 21.4.5 for discussion of shotcrete vs. cast-inplace concrete issues. 

Cost/Disruption 

See Section 21.4.5 for discussion of cost and disruption issues. 

Proprietary Issues 

There are no proprietary concerns with connecting a concrete or masonry wall to a wood 

diaphragm. 

21.4.9 Add Steel Moment Frame (Connected to a Wood Diaphragm) 

Deficiency Addressed by Rehabilitation Technique 

A new moment frame provides additional global strength, reduces demands on existing masonry 

walls and can reduce demands on diaphragms by cutting tributary spans. 

Description of the Rehabilitation Technique 

When a moment frame is added into a URM building, it typically goes either just behind a 

highly punctured street front façade or at an interior location within the diaphragm. Figure 

21.4.9-1 shows the perimeter condition; Figure 21.4.9-2 shows interior conditions. A moment 

frame retrofit at a W1A building with a soft story is discussed in Chapter 6. 

Design Considerations 

Research basis: New steel moment frame issues are covered by FEMA 350 (FEMA, 2000). The 

CUREE woodframe project report on tuckunder building testing (Mosalam, et al., 2002) 
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documents quasistatic component testing of moment frame to wood diaphragm connections and 

full-scale testing of a three-story tuckunder apartment building rehabilitated with a ground story 

moment frame on the open front side. 

Figure 21.4.9-1: New Perimeter Steel Moment Frame to an Existing Wood Floor 

Stiffness considerations: At either the perimeter or interior condition, reasonable stiffness of the 

frame is desirable. At the perimeter, minimizing the amount of drift and resulting masonry 

façade cracking is desirable. At the interior, if the moment frame does not have sufficient 

stiffness, the diaphragm will span between the end walls with the moment frame taking out 

relatively small loads due to its flexibility. 
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Figure 21.4.9-2: New Interior Steel Moment Frame to an Existing Wood Floor 
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Design forces: The new moment frame design can be governed by either stiffness or strength. 

Strength demands can either be minimum design loads or in some cases the moment frame can 

be designed to be stronger than the diaphragm so inelastic action happens in the diaphragm. For 

connection design of the frame to the diaphragm, it is particularly desirable to make sure the 

connections are stronger than the weaker of the diaphragm or the moment frame. 

Pinned base: To minimize foundation demand requirements, new moment frames in retrofits are 

often designed with pinned bases. 

Detailing and Construction Considerations 

Detailing and construction considerations for connecting a new moment frame to an existing 

wood diaphragm include the following. 

Welding vs. bolting: Welding adjacent to wood framing poses a very real fire hazard. 

Specifications and common sense usually dictate various fire watch provisions in these 

situations. Cases of hot welding slag lost from view and later reigniting wood material after the 

welding for the day was finished have been observed and are particularly troublesome. Where 

possible, detailing with shop welded connections, and then field bolting, is desirable. See 

Chapter 8 for additional comments on welding. 

Connecting directly to the masonry: In Figure 21.4.9-1, the moment frame is connected to both 

the masonry façade and the diaphragm to take out load from the punctured wall into the frame 

and from the diaphragm into the frame. In alternative details, the load can be taken from the wall 

into the diaphragm and then through the diaphragm to the frame. 

Cost/Disruption 

Installation of a new moment frame can be fairly disruptive, though it is usually less disruptive 

than a new wall. The frame is chosen when existing window or door openings need to be 

preserved, but head height and visual issues must be considered. Adding new structural steel 

members can be comparatively expensive, but if the choice is to provide a wood structural panel 

overlay on a floor or add a new moment frame, the new moment frame can often be less 

expensive. 

Proprietary Issues 

There are no proprietary concerns with connecting a steel moment frame to a wood diaphragm. 

Certain moment frame beam-to-column connections may have proprietary considerations. See 

Chapter 8. 

21.4.10 Add or Enhance Crosswalls 

Deficiency Addressed by Rehabilitation Technique 

Inadequate diaphragm strength and/or excessive diaphragm displacement. 

Description of the Rehabilitation Technique 

The ABK research program (ABK, 1984) showed that partition walls, called crosswalls, serve as 

energy-absorbing, displacement-limiting damping elements during seismic loading. The 2003 
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IEBC and 1997 UCBC permit certain qualifying buildings to use the “Special Procedure” with 

crosswalls as an integral part of the procedure. There are three basic types of crosswalls: 

existing partitions with various sheathing materials, new partitions, and new steel moment 

frames. Existing partitions may be adequate without any rehabilitation or they may need 

strengthening if they are not connected to the diaphragms or have insufficient capacity. New 

wood structural panel partitions must also be connected to the diaphragm and meet certain 

minimum capacities. New moment frames must meet minimum strength and stiffness criteria. 

See Figures 21.4.10-1 to 21.4.10-3. 

Figure 21.4.10-1: Strengthen Existing Crosswalls 

Design Considerations 

Research basis: ABK (1984) provides background for the basis of the crosswall concept.

Qualifying buildings: In the 2003 ICBC, the Special Procedure can be used only with buildings 

have flexible diaphragms at all levels and meet certain requirements regarding open fronts and 

number of wall lines in each direction. For the 1997 UCBC, these requirements apply, and 
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buildings must be a maximum of six stories, and they cannot be essential or hazardous facilities. 

The crosswall concept and the Special Procedure were not adopted in FEMA 356. 

Figure 21.4.10-2: Add New Crosswalls 

Crosswall requirements: The 1997 UCBC and the 2003 IEBC have a number of requirements 

on crosswall locations, aspect ratios, connection strength, spacing limits that need to be satisfied. 

New crosswalls: New crosswalls will typically be done with structural wood panels and are 

similar to adding new wood structural panel shear walls. See Chapters 5 and 6 for additional 

information. 

Steel moment frames: Adding moment frames for use as crosswalls is very similar to adding 

moment frames for use as new lateral force-resisting elements. See Section 21.4.9 for additional 

information. 
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Figure 21.4.10-3: Add New Moment Frame as Crosswall 

Detailing and Construction Considerations, and Cost/Disruption 

Adding or enhancing woodframe crosswalls is similar to adding or enhancing woodframe shear 

walls; see Chapters 5 and 6. Adding a moment frame as a crosswall is similar to adding a 

moment frame as a lateral force-resisting element; see Section 21.4.9. 

Proprietary Issues 

There are no proprietary concerns with adding crosswalls. Certain steel moment frame beam-tocolumn connections are proprietary. 

21.4.11 Add Supplemental Vertical Support for Truss or Girder 

Deficiency Addressed by Rehabilitation Technique 

Supplemental vertical supports provide a secondary load path for concentrated gravity loads on 

unreinforced masonry walls. 

Description of the Rehabilitation Technique 

A steel or wood post is added under existing trusses and girders. 
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Design Considerations 

Research basis: There are no known tests of supplemental supports. 

Purpose of the support: To some engineers, the goal of adding a supplemental support is to 

provide a back-up gravity load path if there is local deterioration of the masonry underneath a 

concentrated load like a truss bearing point. To others, it provides support if more wholesale 

failure of the wall occurs. 

Independence of the support: The 1997 UCBC and 2003 IEBC both use the term “independent 

secondary columns” when referring to supplemental vertical supports. To some engineers, 

“independent” means separated from the wall. A post an inch or two away from the wall 

satisfies this requirement. To other engineers, “independent” simply means an alternative 

support, so that a ledger or pilaster on the wall is sufficient. In this scenario, the wall just 

beyond the damaged area is assumed to remain intact enough that gravity load resistance is not 

compromised. In Figure 21.4.11-1, a gap is shown. 

Use as out-of-plane brace also: Some engineers like to take advantage of supplemental vertical 

support posts to serve as out-of-plane braces for the walls as well. Combined bending and axial 

demands must be considered. 

Continuity of support: The 1997 UCBC and 2003 IEBC both do not specify whether the posts 

need to continue down to the next story. It is common to transfer loads at the base of upper 

story supplemental supports back to existing framing. This framing must be adequate to take the 

loads that would occur if they supplemental support began to take load. 

Foundation support for posts: The 1997 UCBC and 2003 IEBC both do not explicitly specify 

whether the posts need a compliant new foundation or whether the posts can simply bear on an 

existing slab-on-grade. SEAOC (1992), however, states that a foundation is not required for the 

posts. Nonetheless, some engineers believe it is prudent to check the slab for bearing support 

and provide additional support if needed. 

Triggering elements: The 1997 UCBC and 2003 IEBC only indicate that triggering elements are 

“trusses and beams, others than rafters and joists”. The implication is that elements that support 

other structural members are the primary focus. 

Detailing and Construction Considerations 

Detailing and construction considerations for supplemental vertical supports include the 

following. 

Steel vs wood: Supplemental support posts can either be of steel or wood. Steel members are 

smaller; wood members are less expensive. 

Finish the new elements are leave bare: For certain architectural approaches, leaving the 

supplemental posts bare is compatible with the existing aesthetic. If it is not, the posts can be 

furred at added cost. 
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Figure 21.4.11-1: Supplemental Vertical Support 

Cost/Disruption 

The relative cost of adding supplemental supports depends on the number used, whether they 

continue down to the ground and whether a new foundation is installed. Interior occupants will 

be disrupted locally as the posts are installed, and the usable space in the vicinity of the posts 

will be reduced. 

Proprietary Considerations 

There are no known proprietary concerns with employing supplemental vertical supports. 
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21.4.12 Add Veneer Ties in a URM Wall 

Deficiency Addressed by Rehabilitation Technique 

Missing or inadequate ties between a masonry veneer wythe and the backing wythes can lead to 

delamination of the veneer and a falling hazard. 

Description of the Rehabilitation Technique 

If the front or facing wythe of brick is not integrally tied into the interior wythes with header 

courses, it is considered a veneer. If sufficient metal veneer ties are not present to anchor the 

veneer to the backing wythe, new ties can be provided. Figure 21.4.12-1 shows anchorage using 

drilled dowels to connect the wythes, either from the exterior or interior, in a typical brick wall. 

Figure 21.4.12-2 shows anchorage between stone facing and brick backing. 

Figure 21.4.12-1: Veneer Ties in Brick Masonry 

Design Considerations 

Research basis: No academic research on veneer ties has been identified. Individual vendors 

have performed internal testing of their own products. 

Veneer definition: Codes such as the 2003 ICBC and 1997 UCBC give minimum lay-up 

requirements for multi-wythe solid brick. The facing and backing wythes are to be bonded so 

than not less than 10 percent of the exposed face area is composed of solid headers extending 

less than 4 inches into the backing. The clear distance between adjacent full-length headers shall 

not exceed 24 inches vertically or horizontally. Facing wythes that do not meet these lay-up 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 21 – Type URM: Unreinforced Masonry Bearing Walls

21-60 

requirements are to be considered as veneer. Veneer wythes are not used in resisting shear 

forces in the wall and do not count in the thickness used for determining out-of-plane bending 

resistance. 

Figure 21.4.12-2: Veneer Ties for Stone Masonry Facing 

Veneer tie requirements: Codes such as the 2003 ICBC and 1997 UCBC provide acceptance 

criteria for existing ties. They are to be corrugated galvanized iron strips shown to be in good 

conditions, with dimensions no less than 1” wide, 8” long and 1/16” thick and with a maximum 

spacing of 24” on center and a maximum supported area of four square feet. Veneer ties not 

meeting these requirements are to be strengthened. 

Detailing and Construction Considerations 

Detailing and construction considerations for veneer ties include the following. 

Types of ties: The ties used in new construction are typically inappropriate for rehabilitation 

since they are installed as the wall is built up. There are, however, a fair number of proprietary 

products made by masonry accessory manufacturers that can be used in retrofit applications. 

Some involve expansion anchors in the backing wythes. Others involve helical anchors or spiral 

ties that “screw” into the backing wythes. Figures 21.4.12-1 and 21.4.12-2 show traditional 

drilled dowels. When drilled dowels are used, the tie diameter does not need to be large since 

loads are low and minimizing the size of the hole is important. 

Drilled dowels installation face: Drilled dowels can be installed from the interior or exterior 

depending on which face is more sensitive. 
Techniques for the Seismic Rehabilitation of

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21-61 

Location of drilled dowel: The most reliable location of the drilled dowel is in the center of the 

brick, but this is will cause the largest aesthetic impact. The dowel can be placed in the bed joist 

or bed and head joint intersection to minimize the impact. Recessing the tip of the dowel and 

covering the end with repointing mortar is recommended. 

Brick veneer vs. stone veneer: Anchoring thick stone is usually easier to do from the interior 

because the thickness of the stone permits greater cover on the face, reducing the likelihood of 

spalling. With random ashlar layout and interior installation, however, trying to locate the dowel 

away from the edges of the stone is not practical. 

Corrosion considerations: Any dowel installed from the exterior should be done in stainless 

steel to minimize corrosion.

Cost and Disruption Considerations 

Veneer anchorage can be relatively expensive depending on the number of new ties added. Ties 

installed from the inside are much more disruptive. See Section 21.4.2 for additional 

information on drilled dowels. 

Proprietary Considerations 

Many of the veneer tie anchorage systems are proprietary. 

21.5 References 

ABK, 1981a, Methodology for Mitigation of Seismic Hazards in Existing Unreinforced Masonry 

Buildings: Categorization of Buildings, a joint venture of Agbabian Associates, S.B. Barnes and 

Associates, and Kariotis and Associates (ABK), Topical Report 01, c/o Agbabian Associates, El 

Segundo, CA. 

ABK, 1981b, Methodology for Mitigation of Seismic Hazards in Existing Unreinforced Masonry 

Buildings: Diaphragm Testing, a joint venture of Agbabian Associates, S.B. Barnes and 

Associates, and Kariotis and Associates (ABK), Topical Report 03, c/o Agbabian Associates, El 

Segundo, CA. 

ABK, 1981c, Methodology for Mitigation of Seismic Hazards in Existing Unreinforced Masonry 

Buildings: Wall Testing (Out-of-Plane), a joint venture of Agbabian Associates, S.B. Barnes and 

Associates, and Kariotis and Associates (ABK), Topical Report 04, c/o Agbabian Associates, El 

Segundo, CA. 

ABK, 1984, Methodology for Mitigation of Seismic Hazards in Existing Unreinforced Masonry 

Buildings: The Methodology, a joint venture of Agbabian Associates, S.B. Barnes and 

Associates, and Kariotis and Associates (ABK), Topical Report 08, c/o Agbabian Associates, El 

Segundo, CA. 

Abrams, D.P. and J.M Lynch, 2001, “Flexural Behavior of Retrofitted Masonry Piers,” EERCMAE Joint Seminar on Risk Mitigation for Regions of Moderate Seismicity, IL. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 21 – Type URM: Unreinforced Masonry Bearing Walls

21-62 

Breiholtz, D., 1987, “Centercore Strengthening System for Seismic Hazard Reduction of 

Unreinforced Masonry Bearing Wall Buildings,” Proceedings of the 56th Annual Convention of 

the Structural Engineers Association of California, San Diego, CA. 

Breiholz, D.C., 1992, “Center Core Seismic Hazard Reduction System for URM Buildings,” 

Proceedings of the Tenth World Conference on Earthquake Engineering, Vol. 9, pp. 5395-5399. 

Ehsani, M.R. and H. Saadatmanesh, 1996, “Seismic Retrofit of URM Walls with Fiber 

Composites,” The Masonry Society Journal, Spring. 

Ehsani, M.R. H. Saadatmanesh, and A. Al-Saidy, 1997, “Shear Behavior of URM Retrofitted 

with FRP Overlays,” Journal of Composites for Construction, ASCE 1(1), 17-25. 

Ehsani, M.R. H. Saadatmanesh, and J.I. Velazquez-Dimas, 1999, “Behavior of Retrofitted URM 

Walls Under Simulated Earthquake Loading,” Journal of Composites for Construction, ASCE 

4(3), 134-142. 

ElGawady, M., P. Lestuzzi, and M. Badoux, 2003, “Rehabilitation of Unreinforced Brick 

Masonry Walls Using Composites,” Architectural and Structural Design of Masonry With Focus 

on Retrofitting of Masonry Structures and Earthquake Resistant Design, International Short 

Course, Dresden University of Technology, December 7-18, Dresden, Germany. 

ElGawady, M, P. Lestuzzi, and M. Badoux, 2004, “A Review of Conventional Seismic 

Retrofitting Techniques for URM, 13th International Brick and Block Masonry Conference, 

Amsterdam, July 4-7. 

FEMA, 1997a, NEHRP Guidelines of the Seismic Rehabilitation of Buildings, FEMA 273, 

Federal Emergency Management Agency, Washington, D.C., October. 

FEMA, 1997b, NEHRP Commentary on the Guidelines of the Seismic Rehabilitation of 

Buildings, FEMA 274, Federal Emergency Management Agency, Washington, D.C., October. 

FEMA, 1999a, Evaluation of Earthquake Damaged Concrete and Masonry Wall Buildings: 

Basic Procedures Manual, FEMA 306, May. 

FEMA, 1999b, Evaluation of Earthquake Damaged Concrete and Masonry Wall Buildings: 

Technical Resources, FEMA 307, May. 

FEMA, 2000, Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 

356, Federal Emergency Management Agency, Washington, D.C., November. 

Ganz, H.R., 1993, “Strengthening of Masonry Structures with Post-Tensioning,” Proceedings of 

the Sixth North American Masonry Conference, Philadelphia, Pennsylvania, June 6-9, pp. 

645-655. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 21 – Type URM: Unreinforced Masonry Bearing Walls

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Ghanem, G., M.A. Zied and A.E. Salama, 1994, “Retrofit and Strengthening of Masonry 

Assemblages Using Fiberglass,” Proceedings of the Tenth International Brick and Block 

Masonry Conference, Vol. 2, pp. 499-508. 

Haroun, M.A and A.S Mosallam, 2002, Cyclic Shear Test of Multi-Wythe Existing Brick Wall: 

Retrofitted by a Single Layer of TYFO SHE-51A Applied on One Side, University of California, 

Irvine and California State University Fresno, February. 

Hutchison, D.L., P.M.F. Yong and G.H.F. McKenzie, 1984, “Laboratory Testing of a Variety of 

Strengthening Solutions for Brick Masonry Wall Panels,” Proceedings of the Eighth World 

Conference on Earthquake Engineering, Vol. 1, pp. 575-582. 

ICBO, 1997, Uniform Code for Building Conservation, 1997 Edition, International Conference 

of Building Officials, Whittier, CA. 

ICC, 2003, International Existing Building Code, 2003 Edition, International Code Council, 

Country Club Hills, IL. 

ICC-ES, 2003, “Interim Criteria for Concrete and Reinforced and Unreinforced Masonry 

Strengthening Using Fiber-Reinforced Polymer (FRP) Composite Systems,” Acceptance Criteria 

AC-125, June, International Code Council Evaluation Service, Whittier, CA. 

ICC-ES, 2005, Acceptance Criteria for Anchors in Unreinforced Masonry Elements (AC60), 

Approved April 2005, Whittier, CA.

Jurukovski, D., L. Krstevska, R. Alessi, P.P. Diotallevi, M. Merli and F. Zarri, 1992, “Shaking 

Table Tests of Three Four-Storey Brick Masonry Models: Original and Strengthened by RC 

Core and RC Jackets, Proceedings of the Tenth World Conference on Earthquake Engineering,

July, Madrid, A A Balkema, Rotterdam, Vol. 5, pp. 2795-2800. 

Kahn, L.F., 1984, “Shotcrete Retrofit of Unreinforced Brick Masonry,” Proceedings of the 

Eighth World Conference on Earthquake Engineering, San Francisco, CA, July, pp. 583-590. 

Kariotis, J., 1982, “Bracing of Unreinforced Masonry Walls, Out-of-Plane,” Earthquake Hazard 

Mitigation of Unreinforced Masonry Buildings Built Prior to 1934, Seminar Proceedings, 

Structural Engineers Association of Southern California, April, Whittier, CA. 

Mosalam, K., et al., 2002, Seismic Evaluation of an Asymmetric Three-Story Woodframe 

Building, CUREE Publication No. W-19, Consortium of Universities for Research in Earthquake 

Engineering, Richmond, CA. 

Paquette, J., M. Bruneau, and S. Brzev, 2003, “Pseudo-Dynamic Testing of Unreinforced 

Masonry Building with Flexible Diaphragm,” Journal of Structural Engineering, Vol. 129, No. 

6, June, pp. 708-716. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 21 – Type URM: Unreinforced Masonry Bearing Walls

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Paquette, J., M. Bruneau, and S. Brzev, 2004, “Seismic Testing of Repaired Unreinforced 

Masonry Building Having Flexible Diaphragm,” Journal of Structural Engineering, Vol. 130, 

No. 10, October, pp. 1487-1496. 

Paquette, J. and M. Bruneau, 2004, “Pseudo-Dynamic Testing of Unreinforced Masonry 

Building with Flexible Diaphragm, Proceedings of the 13th World Conference on Earthquake 

Engineering, Vancouver, B.C., Paper 2609. 

Plecnik, J., T. Cousins and E. O’Conner, 1986, “Strengthening of Unreinforced Masonry 

Buildings,” Journal of Structural Engineering, American Society of Civil Engineers, Vol. 112, 

No. 5, May, pp. 1070-1087. 

Plecnik, J.M., 1988, “Preliminary Report on “Summary of the Second Meeting of the Advisory 

Committee for NSF Research Project on the Centercore Rehabilitation Technique,” Structures 

Laboratory Report 88-8-18, California State University, Long Beach, August 17. 

Portland State University, 1998, Full Scale Tests of Retrofitted Hollow Clay Tile Walls, Final 

Report Prepared for Contech, Department of Civil Engineering, June. 

Reinhorn, A.M. and A. Madan, 1995a, Evaluation of Tyfo-W Fiber Wrap System for Out of 

Plane Strengthening of Masonry Walls, Report No. AMR 95-2001, Department of Civil 

Engineering, State University of New York at Buffalo, Buffalo, NY, March. 

Reinhorn, A.M. and A. Madan, 1995b, Evaluation of Tyfo-W Fiber Wrap System for In Plane 

Strengthening of Masonry Walls, Report No. AMR 95-2002, Department of Civil Engineering, 

State University of New York at Buffalo, Buffalo, NY, August. 

Rosenboom, O.A. and M.J. Kowalsky, 2004, “Reversed In-Plane Cyclic Behavior of 

Posttensioned Clay Brick Masonry Walls,” Journal of Structural Engineering, Vol. 130, No. 5, 

May, pp. 787-798. 

Rutherford & Chekene, 1990, Seismic Retrofitting Alternatives for San Francisco’s Unreinforced 

Masonry Buildings: Estimates of Construction Cost & Seismic Damage, for the Department of 

City Planning of the City and County of San Francisco, Oakland, CA, May. 

Rutherford & Chekene, 1997, Development of Procedures to Enhance the Performance of 

Rehabilitated Buildings, prepared by Rutherford & Chekene Consulting Engineers, published by 

the National Institute of Standards and Technology as Reports NIST GCR 97-724-1 and NIST 

97-724-2. 

Schwegler, G. and P. Kelterborn, 1996, “Earthquake Resistance of Masonry Structures 

Strengthened with Fiber Composites,” Proceedings of Eleventh World Conference on 

Earthquake Engineering, Acapulco, June, Elsevier Science Ltd., Paper No. 1460. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 21 – Type URM: Unreinforced Masonry Bearing Walls

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Senescu, R. and K.M. Mosalam, 2004, Retrofitting of Unreinforced Masonry Walls Using Fiber 

Reinforced Polymer Laminates, Report UCB/SEMM 2004/3, University of California, Berkeley, 

CA, May. 

SEAOC (Structural Engineers Association of California), 1992, Commentary on Appendix 

Chapter 1 of the Uniform Code for Building Conservation, Seismic Strengthening Provisions for 

Unreinforced Masonry Bearing Wall Buildings, June 20. 

SEAOSC (Structural Engineers Association of Southern California),1982, Earthquake Hazard 

Mitigation of Unreinforced Masonry Buildings Built Prior to 1934, Seminar Proceedings, April, 

Whittier, CA. 

SEAOSC (Structural Engineers Association of Southern California), 1986, Earthquake Hazard 

Mitigation of Unreinforced Masonry Buildings Pre-1933 Buildings, Seminar Proceedings, 

October, Whittier, CA. 

Tumialan, J.G., N. Galati, and A. Nanni, 2002, “Field Assessment of URM Walls Strengthened 

with FRP Laminates,” Journal of Structural Engineering, American Society of Civil Engineers, 

Vol. 129, No. 8, August. 

Tumialan, J.G. N. Galati, S. Namboorimadathil, and A. Nanni, 2002, “Strengthening of Masonry 

with FRP Bars,” Proceedings of the Third International Conference on Composites in 

Infrastructure ICCI’02, June 10-12, San Francisco. 

Tumialan, J.G., A. Morbin, F. Micelli, and A. Nanni, 2002, “Flexural Strengthening of URM 

Walls with FRP Laminates, Proceedings of the Third International Conference on Composites in 

Infrastructure ICCI’02, June 10-12, San Francisco, CA. 

Vandergrift, Gergely, and Young, 2002, “CFRP Retrofit of Masonry Walls,” Proceedings of the 

Third International Conference on Composites in Infrastructure ICCI’02, June 10-12, San 

Francisco, CA. 

Velazquez-Dimas, J.I. and M.R. Ehsani, 2000, “Modeling Out-of-Plane Behavior of URM Walls 

Retrofitted with Fiber Composites,” Journal of Composites for Construction, ASCE 4(4), pp. 

172-181. 

Velazquez-Dimas, J.I., M.R. Ehsani, and H. Saadatmanesh, 2000, “Out-of-Plane Behavior of 

Brick Masonry Walls Strengthened with Fiber Composites,” ACI Structural Journal, 97(3), 

American Concrete Institute, Farmington Hills, MI, pp. 377-387. 
Techniques for the Seismic Rehabilitation of

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22-1 

Chapter 22 - Diaphragm Rehabilitation Techniques 

22.1 Overview 

Diaphragm failures are less commonly observed in earthquakes, and the disruption caused by 

strengthening the diaphragm can be quite significant, so diaphragm rehabilitation is less 

commonly employed than adding global strength and stiffness, or improving connection paths. 

Some diaphragms are inherently less likely to be an issue, such as cast-in-place concrete flat 

slabs or waffle slabs; others like straight sheathed wood or poorly connected precast floors are of 

greater concern. This chapter provides examples of various diaphragm systems and their 

strengthening techniques. They are organized here in a single chapter for convenience and 

because many of the diaphragms can be found in different building types. For discussion of 

diaphragm-to-wall connection issues, see individual building type chapters. 

22.2 Detailed Description of Diaphragm Rehabilitation Techniques 

22.2.1 Wood Diaphragm Strengthening 

Deficiency Addressed by Rehabilitation Technique 

Inadequate diaphragm strength and/or stiffness 

Description of the Rehabilitation Techniques 

The addition of new wood structural panel sheathing is a traditional and common approach to 

diaphragm strengthening. Adding fastening and blocking to existing wood structural panel 

sheathing can also be done. Specifically, this section covers: 

Replacing existing sheathing with new wood structural panel sheathing 

Wood structural panel sheathing overlays with new blocking 

Wood structural panel sheathing overlays without new blocking 

Improving strength and stiffness of an existing wood structural panel sheathed diaphragm 

Each of these techniques aims to improve the shear strength and lateral stiffness of the existing 

diaphragm. Figure 22.2.1-1 shows the replacement of existing sheathing with new sheathing 

directly onto the existing joists. Figure 22.2.1-2 shows a wood structural panel overlay on 

existing straight sheathing floors or roofs when new blocking is added below the existing 

sheathing. Figure 22.2.1-3 shows an overlay when blocking is not added, and Figure 22.2.1-4 

shows a similar overlay to use when the bottom of the existing sheathing is to remain exposed to 

view and penetrations through it would not be acceptable. Figure 22.2.1-5 shows how shear 

transfer can be made to get past an existing partition sill that is to remain in place. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 22 – Diaphragm Rehabilitation Techniques

22-2 

Figure 22.2.1-1: Remove and Replace Existing Wood Sheathing 

with Wood Structural Panel at a Roof 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 22 – Diaphragm Rehabilitation Techniques

22-3 

Figure 22.2.1-2: Wood Panel Overlay with Blocking Over Existing Sheathing 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 22 – Diaphragm Rehabilitation Techniques

22-4 

Figure 22.2.1-3: Wood Panel Overlay without Blocking Over Existing Sheathing 
Techniques for the Seismic Rehabilitation of

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22-5 

Figure 22.2.1-4: Wood Panel Overlay without Blocking Over Existing Sheathing 

When the Bottom of the Existing Sheathing is Visible 
Techniques for the Seismic Rehabilitation of

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Figure 22.2.1-5: Shear Transfer in New Overlay at Existing Partitions 

Design Considerations 

Research basis: When new wood structural panel sheathing replaces existing sheathing, then the 

basic research for panel sheathing used to develop diaphragm capacities is applicable, and values 

would be taken from the relevant building code. When structural panel sheathing is used as an 

overlay, there is less research available. Values that have made it into model codes such as the 

UCBC (ICBO, 1997) and IEBC (ICC, 2003b) are based in part on the ABK research program for 

URM bearing wall strengthening, including ABK (1981). In this program, a series of 14 fullscale, 20’x60’ horizontal diaphragm specimens were subjected to quasi-static, cyclic, in-plane 

displacements and dynamic, in-plane earthquake shaking. Specimens include filled and unfilled 

steel deck, blocked and unblocked plywood, and straight and diagonal sheathing with and 

without plywood overlays and with roofing material. More recent tests include Peralta, Bracci, 

and Hueste (2004) where a series of twelve 12’x24’ horizontal diaphragm specimens were 

subjected to quasi-static, reversed cyclic in-plane displacements. Specimens included tongue 

groove sheathing retrofit with strapping and with an underlying steel truss, straight sheathing 

with and without openings retrofit with a steel truss and with blocked and unblocked plywood 

overlays. Results were compared with both FEMA 273 (1997a) and FEMA 356 (2000). 
Techniques for the Seismic Rehabilitation of

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Types of diaphragms: Approaches to diaphragm rehabilitation can be categorized as follows: 

Structural wood panel sheathing where the existing sheathing is replaced: This is the 

approach typically used when high capacities are needed. 

o “High load” diaphragms where 3x and 4x blocking is added and multiple lines of 

nailing are used: This may be done in accordance with provisions in the IBC; 

additional detailing information in ICC-ES Legacy Report 1952 (ICC-ES, 2004) 

is highly recommended. See APA (2000) for testing results. 

o Traditional diaphragms with 3x and 2x blocking and various panel layouts: The 

relevant building code capacities are used. An issue that often arises is whether 

existing joists, which are typically thicker than the code assumed 1-1/2”, can 

count as 3x blocking. Some engineers ratio values between 2x and 3x code 

capacities. 

o Unblocked diaphragms: It is relatively unusual to remove existing sheathing only 

to replace it with unblocked wood structural panels as the capacities are not 

substantially different. 

Wood structural panel sheathing overlays over existing 1x nominal sheathing: In the 

1997 UCBC, there are values given for the following three approaches. The 2003 IEBC 

only lists the first type. Inherent in these approaches is the assumption that existing 

lumber sheathing is one-inch nominal (commonly 5/8-inch to 7/8-inch actual) thickness. 

o Wood structural panel overlays nailed directly over existing straight sheathing 

with ends of the panels bearing on joists or rafters and edges of the panels located 

on center of individual sheathing boards: The lack of blocking makes this a 

relatively weak diaphragm. 

o Wood structural panel overlays nailed directly over existing diagonal sheathing 

with ends of wood structural panel sheets bearing on joists or rafters: Diagonal 

sheathing provides increased strength compared to the overlay of straight 

sheathing. 

o Wood structural panel overlays nailed directly over existing straight or diagonal 

sheathing with ends of panels bearing on joists or rafters with edges of panels 

located over new blocking and nailed to provide a minimum nail penetration into 

framing and blocking of 1-5/8”: The 1997 UCBC limits this to 75% of code 

values for wood structural panel overlays without the existing sheathing, due in 

part to the potential for bending of the nail in the existing sheathing before it 

reaches the main member blocking and the risk of the nailing being near the edges 

of the existing sheathing. 

Wood structural panel sheathing overlays over existing lumber planking (2-inch nominal 

or thicker) or laminated decking; the IBC (ICC, 2003a) and the AF&PA (2005) permit 

wood structural panel diaphragm sheathing to be fastened over solid lumber planking or 

laminated decking using full tabulated values for new construction. Inherent is the 

assumption that the sheathing nail will have a penetration of not less than 10 diameters 

(1-3/8 inches for 8d common and 1-1/2” for 10d common) into the planking or decking. 

Special attention is needed at all diaphragm boundaries to ensure shear transfer from the 

sheathing, through the planking or decking to the boundary members below. 
Techniques for the Seismic Rehabilitation of

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22-8 

Wood structural panel sheathing overlays over existing spaced (or skip) sheathing: A 

common roof framing system is to span 1x nominal boards across rafters. Building paper 

is placed on top of the boards and under the final roofing layer such as shakes or shingles. 

Wide spaces of several inches are left between the 1x boards both to save sheathing 

material and to permit air flow to help dry the roofing sandwich. This construction is the 

most flexible and the weakest type of existing wood diaphragm and has no code values. 

Wood structural panel overlays can be placed across the skip sheathing. Care should be 

taken to align the panel edges atop the spaced sheathing. Due to the 1x thickness of the 

spaced sheathing, full development of the nail will not be achieved. With the gaps 

between sheathing boards, two edges of the wood structural panels will not be blocked. 

1x sheathing or wood structural panel nailing strips with matching thicknesses can be 

placed atop the rafters in the gap to serve as “blocking” at these edges. Direct code 

values for these overlays are not available, though some engineers use code values 

reduced down by the amount of actual vs. full nail development length. Alternatively, 

staples can be used to help address the shallow sheathing depth. 

Wood structural panel sheathing overlays over existing wood structural panel sheathing: 

Two layers of wood structural panel sheathing have been tested and documented in APA 

(2000). The tested configuration used overlays at panel ends in high-load regions. 

Existing wood structural panel diaphragm enhancement without overlays: A wide variety 

of rehabilitation measures are available for existing wood structural panel diaphragms 

that do not involve new overlays. These include: 

o Addition of 2x wood blocking to an unblocked diaphragm (Dolan et al., 2003) 

o Addition of sheet steel blocking to an unblocked diaphragm (APA, 2000) 

o Addition of nailing to existing blocked diaphragm (allows limited improvement 

because framing member requirements change at closer nail spacing) 

o Adding staples to existing wood structural panel diaphragm. Staples are designed 

to carry entire seismic unit shear 

o Stapling of tongue and groove sheathing joints (APA, 2000). 

o Addition of a wood structural panel soffit in local areas of high diaphragm shear 

(see Section 22.2.2) 

Existing diaphragms without overlays: In the 1997 UCBC and 2003 IEBC, there are 

values for the following existing materials: 

o Roofs with straight sheathing and roofing applied directly to the sheathing 

o Roofs with diagonal sheathing and roofing applied directly to the sheathing 

o Floors with straight tongue-and-groove sheathing 

o Floors with straight sheathing and finished wood flooring with board edges offset 

or perpendicular: Values are relatively high for this combination 

o Floors with diagonal sheathing and finish wood flooring: Values are also 

relatively high for this combination 

FEMA 356 has its own extensive listing of diaphragm types, and there are examples and even 

tests in the literature exploring the influence of glue, double layers of panel sheathing, 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 22 – Diaphragm Rehabilitation Techniques

22-9 

herringbone panel overlays. IBC, APA (2000), and ICC-ES (2004) provide techniques for 

calculating code level values, including stapled diaphragms. 

In order to select and properly detail diaphragm rehabilitation measures, it is important to 

determine the layout and thickness of existing sheathing and framing. Significant attention is 

needed to transfer of shear at all diaphragm boundaries. This includes diaphragm chords (Section 

22.2.2), subdiaphragms and cross-ties for flexible diaphragm/rigid wall buildings (Section 

22.2.3), and collectors (Sections 6.4.5 and 7.4.2). 

Condition assessment of the existing roof structure is important. It is common to find decay 

damage to existing framing and sheathing in the vicinity of roof drains. 

Detailing and Construction Considerations 

Detailing and construction considerations for wood diaphragm strengthening include the 

following. 

Aligning panel edges: When the existing sheathing is removed, the joists or rafters typically 

remain in place. Their spacing will vary. To align the edges of new 4’x8’ sheets of structural 

wood panels on top of the supporting framing requires field measuring and cutting the sheets. 

Alternatively, new blocking can be added between existing framing to reduce the need to cut the 

structural wood panels. See Figure 22.2.1-1 for examples of each approach. 

Missing sheathing edges: To reduce the risk of splitting during installation or later during the 

earthquake, nailing through the center of existing joists is desirable. This can take considerable 

field effort, however, due to the need to field measure and cut the structural wood panels. See 

Figure 22.2.1-2 and 22.2.1-3 for examples. 

Staples, short nails, regular length nails: When the existing sheathing is removed and the 

structural wood panel is placed directly on the framing, regular length nails are commonly used. 

When the structural wood panel is applied to the existing lumber sheathing without blocking, 8d 

and 10d nails will go well through the underside of the sheathing. “Short” or “diaphragm” nails 

can be used to reduce the amount of nail protrusion. See Figure 22.2.1-3. When the overlay is 

on a diaphragm that is architecturally exposed from below, nail penetrations are not desirable. 

Staples can also be used, such as shown in Figure 22.2.1-4; per IBC, 16 gage staples require oneinch penetration into framing for tabulated values. 

The nail penetration into diaphragm framing members required to achieve code and standard 

tabulated allowable shear values has changed recently. In the past, a nail penetration of 1-5/8 

inches was required to obtain full diaphragm capacity. As a result, allowable shear reductions 

were applied when only 1-1/2 inch penetration was provided, as commonly occurs with 2x flat 

blocking in diaphragms or engineered joist top chords. The 2003 IBC only requires 1-3/8-inch 

penetration for 8d common nails and 1-1/2-inch penetration for 10d common nails. APA T98-22 

(APA, 1998) provides one explanation, based on calculation using yield-mode equations. The 

nail penetration requirements are stated specifically in the diaphragm tables, and methods to 

adjust for reduced penetration are not suggested. Reduction in penetration below the IBC 

minimums is not recommended; because considerable slip can occur between sheathing and 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 22 – Diaphragm Rehabilitation Techniques

22-10 

framing as a diaphragm takes up load, reduced embedment may lead to premature withdrawal 

failure. These APA and IBC penetration requirements are applicable to sheathing-to-framing 

fastening. 

Nail penetration requirements have also been changing in the NDS (AF&PA, 2005), where a nail 

penetration of 10 diameters is now adequate to develop tabulated nail capacities. This number 

has been 12 and 11 diameters in previous provisions. Nails with a penetration of less than six 

diameters are not permitted to be used. These NDS penetration requirements are applicable to 

framing-to-framing fastening. 

Gluing of diaphragms: Adding glue between a wood structural panel and supporting framing in a 

diaphragm or shear wall assembly where inelastic behavior is anticipated is strongly 

recommended against, as glued sheathing has limited ductility or energy dissipation capacity. 

This applies whether or not nailing is provided in addition to the glue. Dolan et al. (2003) 

evaluated the effect of diaphragm gluing on strength and stiffness. 

Partitions: A diaphragm that is continuous between walls provides the stiffest and most direct 

load path. In an existing building, however, there are almost always existing partitions on the 

floor. If they are to remain during the rehabilitation, Figure 22.2.1-5 shows a detail for shear 

transfer from one side to the other of the partition sill in an overlay. This approach is adequate 

when the value of the load transfer is relatively low; when higher capacities are needed such as 

for boundary nailing or double rows of nails, alternative details will need to be developed and 

typically include blocking down and around the partition. 

Weight: Adding structural wood panel sheathing over existing sheathing adds weight to 

diaphragm. This rarely poses a problem, but the engineer should consider the issue. 

Location of diaphragm: Figures 22.2.1-1 through 22.2.1-5 all show the structural wood panel 

added to the top of the floor. In many situations, due to finishes on the top of the floor or usage 

of a particular story, enhancing the underside of the diaphragm is a less disruptive approach. 

Cost/Disruption 

Adding structural wood panel overlays can be a significant disruption to occupants, just from the 

need for access to either the top or underside of the floor, as well as from the noise of sawing and 

hammering. If the building is to remain occupied during rehabilitation, work is sometimes 

phased by floor or wing to minimize the number of impacted occupants at any one time. Many 

existing buildings have had roof strengthening done from above with the occupants in place. 

Sometimes the work is limited to certain hours that are considered less disruptive. When 

improvements or overlays are installed on top of the roof, it may be necessary to develop 

detailing to allow work around existing roof top equipment platforms and curbs, skylights, etc. 

Proprietary Issues 

There are typically no proprietary concerns with wood diaphragm strengthening. 
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22.2.2 Add or Enhance Chord in Existing Wood Diaphragm 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses inadequate, incomplete or missing chords in buildings 

with reinforced concrete or masonry shear walls; also addressed is inadequate shear transfer into 

chord members. Provision of chord members is specifically not required for diaphragms in 

unreinforced masonry buildings, where wall bed joint shear is thought to provide some chord 

member capacity. See Chapter 21 for additional discussion of URM buildings. 

Rehabilitation approaches discussed may also be applicable to collectors and detailing at reentrant corners. While systematic evaluation may identify the need for chord enhancement, it is 

also often provided in conjunction with diaphragm enhancement, as discussed in Section 22.2.1. 

Description of the Rehabilitation Technique 

The purpose of a diaphragm chord is to act as a tension or compression member resisting 

diaphragm flexural forces; this requires both an adequate member and adequate transfer of shear 

from the diaphragm to the chord member along the full member length. In buildings with wood 

diaphragms and reinforced concrete or masonry walls, the most common chord members are 

reinforcing steel placed in the wall at or near the roof diaphragm elevation and a structural steel 

angle bolted to the wall. 

Where the existing chord member is adequate, rehabilitation may be limited to enhancing shear 

transfer. Figures 22.2.2-1A and 1B show added fastening at the roof diaphragm boundary and 

added adhesive anchors to the concrete or masonry wall, where the existing reinforcing steel is 

adequate. The reader is cautioned to check the adequacy of the reinforcing as-built conditions at 

tilt-up concrete walls and reinforced masonry walls with movement joints. See Chapters 16 and 

19 for further discussion. 

Where additional chord capacity is needed, it is most practical to add a new steel angle on the 

surface of the existing concrete wall, as shown in Figures 22.2.2-2A and 22.2.2-2B. 

Diaphragm chords may be incomplete when vertical offsets occur in the roof diaphragm. When 

this occurs, it may be possible to use a tilt-up panel to resolve the vertical offset, as shown in 

Figure 22.2.2-3. Where chords are not occurring at the roof diaphragm level, care should be 

taken in assessing the unsupported length for compression design. 

Design Considerations 

Research basis: No research applicable to this rehabilitation measure has been identified. 

Enhancement to an existing chord member must be compatible with existing chord behavior. It is 

unlikely that any chord enhancement applied to the wall face can be compatible with an existing 

reinforcing steel chord, because of fastener slip required to develop forces in the new chord 

member. Where an existing reinforcing steel chord is being enhanced, it is suggested that the 

capacity of the existing reinforcing be neglected. 
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Figure 22.2.2-1: Enhanced Chord Member Fastening at Wood Diaphragm 
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Figure 22.2.2-2: Enhanced Chord Member and Fastening at Wood Diaphragm 
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Figure 22.2.2-3: Elevation of Wall Panels with Incomplete Chord 

Due to Vertical Offset in Roof Diaphragm 

The new or enhanced chord member must be anchored into the diaphragm for unit shear transfer. 

Anchorage for shear transfer is also discussed in Section 22.2.3. As a wood sheathed diaphragm 

is loaded, slip will occur between the perimeter framing member and the sheathing. Fastening of 

the chord or chord enhancement should not inhibit this slip. If the slip is not permitted, premature 

failure at the opposite side of the sheathing panel could occur. This is not a concern with a 

welded steel deck diaphragm, which has limited slip. 

Chord stresses due to shrinkage and temperature change have been identified as a concern for 

connections between tilt-up panels (SEAOSC, 1979), as discussed in Chapter 16, and these 

stresses should be considered in chord design. 

Detailing Considerations 

It is desirable to keep the chord elevation as close as possible to the elevation of the diaphragm in 

order to minimize secondary stresses and additional deformation. At the edge of the diaphragm 

this is most easily accomplished by putting a new chord member on the top of the diaphragm 

(shown as an alternate location in Figures 22.2.2-2A and 22.2.2-2B). This is only possible when 

re-roofing will occur at the time of rehabilitation work. Otherwise added chord members must be 

located below existing perimeter members and connections. Splicing of the new or enhanced 

chord member needs to be specifically detailed. 

Diaphragm boundary fastening: Based on observed shear wall test behavior (Gatto and Uang, 

2002), providing extra nailing at the diaphragm boundary will likely not provide extra diaphragm 
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capacity, and it may result in premature failure at the first interior joint due to shifting of the 

center of the fastener group. As a result, sheathing fasteners should be placed symmetrically 

around the panel edge where possible, and care should be taken to not arbitrarily put extra rows 

of fasteners at the boundary chord and collector members. 

It is preferable to use the same type and size of sheathing fastener at the diaphragm boundary as 

at the diaphragm interior; however, this may be difficult where new steel chord members are 

being added on top of the diaphragm, as shown in the alternate location in Figures 22.2.2-2A and 

22.2.2-2B. Although graphically shown as a nailed connection from the steel angle chord 

member to the diaphragm, it may become necessary to use wood screws or lag screws for higherload diaphragms. Testing of this mix of fasteners has not been identified, so behavior is not 

known. Behavior of cut-thread wood screws in sheathing to framing fastening has been observed 

to be problematic, as discussed in Section 6.4.2. 

Partially grouted masonry walls: Where shear transfer is being provided into partially grouted 

masonry walls, it is necessary to verify that the existing wall is grouted at the anchorage location. 

It is generally anticipated that the existing masonry will be grouted and reinforced at the existing 

roof ledger location. If, however, anchorage to the wall needs to occur above or below this 

location, presence of grout will need to be verified. Although methods of anchoring only to the 

face shell are available, these have very low capacities and should never be mixed with anchors 

to grouted masonry. So, it is recommended that anchorage to grouted cells be provided. It may 

be possible to grout cells at desired anchor locations, particularly if just above the roof line and 

accessible from at the parapet. Care should be taken so that the anchor force in a grouted cell 

does not exceed the force that can be transferred by the unit bed joint. 

Collector connections: Where possible, it is desirable for the collector member to be located at 

the face of the shear wall and extend the full length of the shear wall, matching the chord 

detailing shown in Figures 22.2.2-1A and 22.2.2-1B and Figures 22.2.2-2A and 22.2.2-2B. This 

detailing approach is often but not always possible. Great care should be taken when a significant 

collector load needs to be transferred into the very end of a concrete or masonry wall. The load 

needs to be transferred far enough into the wall that wall reinforcing can develop adequate 

capacity. Edge and center to center spacing requirements need to be met for anchorage to the 

wall. 

Cost, Disruption and Construction Considerations 

When rehabilitation work is undertaken on the roof diaphragm, it is important that the cost and 

the preferred location for work take into account the combination of work, rather than 

considering one portion at a time. If several diaphragm measures will be undertaken, it will 

quickly become cost-effective to remove the roof and allow work from the top. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique other than the use of 

proprietary connectors and adhesives as part of the assemblage. 
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22.2.3 Add or Enhance Diaphragm Cross-ties for Out-of-Plane Wall-toDiaphragm Loads in Flexible Wood and Steel Diaphragms 

Deficiency Addressed by Rehabilitation Technique 

This rehabilitation technique addresses inadequate or missing diaphragm cross-tie systems, as 

part of wall anchorage requirements for flexible diaphragm / rigid wall buildings. This 

rehabilitation technique is used when diaphragm cross-tie systems have not been provided, or do 

not provide adequate strength. Both wood and steel flexible diaphragms are addressed. The 

diaphragm cross-tie system is an extension of wall to diaphragm anchorage for out-of-plane 

loads, as addressed in Chapter 16 for PC1 buildings, Chapter 18 for RM1t buildings, and 

Chapter 21 for URM buildings. 

The addition or enhancement of the diaphragm cross-tie system is recommended as a high 

priority for rehabilitation for wood diaphragm PC1, RM1t, and for URM buildings. Due to 

limited earthquake experience to date, the vulnerability of and need to rehabilitate cross-tie 

systems in flexible steel diaphragms is not known; however, vulnerabilities similar to wood 

diaphragms buildings might occur. This section illustrates the basic rehabilitation concepts. 

SEAONC Guidelines (SEAONC, 2001) provides exhaustive treatment of detailing for PC1

buildings. 

Description of the Rehabilitation Technique 

A system of continuous ties between exterior walls of flexible diaphragm / rigid wall buildings is 

now a requirement for new construction in areas of high seismic hazard. The concept is to tie all 

the way across the diaphragm to opposing walls. The wall anchorage will generally occur at four, 

six or eight feet on center. 

Cross-ties at each wall anchor location can be fairly easily accommodated in new steel deck 

diaphragm buildings. The steel deck is permitted to be used as the cross-tie in the direction of its 

span, provided it can be shown to be adequate for tension and compression forces. See Chapter 

16 for further discussion. Perpendicular to the decking span, with relatively long-span steel joist 

members it is practical to provide diaphragm cross-ties at each joist. The number of cross-tie 

splices required is not excessive, and wall anchorage forces do not greatly change the open web 

joist design. This is also the preferred approach for rehabilitation of cross-ties in steel deck 

construction, where the forces can be accommodated by decking and joists. 

Cross-ties at each wall anchor location are not as easily accommodated in wood diaphragm 

systems, particularly in panelized wood diaphragm systems with eight foot subpurlins spans, due 

to the number of breaks in framing members across which connectors would have to be 

provided. A cross-tie system using subdiaphragms has been developed for wood diaphragm 

buildings. This same approach can be used in steel diaphragm buildings. Rather than 

representing anticipated building behavior, subdiaphragms need to be viewed as a computational 

tool. Unit shears from subdiaphragm design are not intended to be added to main diaphragm 

shears. Design in each area of the diaphragm needs to be for the more critical of subdiaphragm or 

main diaphragm seismic forces. 
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Figure 22.2.3-1A illustrates a roof plan for a wood diaphragm that uses subdiaphragms as part of 

the cross-tie system. For loading in the east-west direction, subdiaphragms are provided between 

Lines A and B and Lines G and H. Similarly, for loading in the north-south direction, 

subdiaphragms are provided between Lines 1 and 2 and Lines 3 and 4. The depth of the 

subdiaphragm is selected based on the unit shear at the subdiaphragm reaction, as well as having 

a member available to act as a subdiaphragm chord. The wall anchor force is transferred into the 

subdiaphragm over the full subdiaphragm depth. For east-west loads subdiaphragms span 

between Lines 1 and 2, 2 and 3, and 3 and 4. Subdiaphragm reactions are resisted at the exterior 

walls at Lines 1 and 4, and interior cross-ties are provided on Lines 2 and 3. Boundary nailing 

must be provided for each subdiaphragm on Lines 1, 2, 3, 4, A and B. The cross-tie provides a 

continuous tie between exterior walls with a capacity not less than the subdiaphragm reaction. 

This pattern is repeated for subdiaphragms between Lines G and H, 1 and 2, and 3 and 4. 

Figure 22.2.3-1A: Roof Plan with Diaphragm Cross-Tie System 

Using Subdiaphragms, Shown for Wood Diaphragm 

Figures 22.2.3-1B, 22.2.3-C, and 22.2.3-D depict sections through the subdiaphragm extending 

between Lines A and B. Figure 22.2.3-1B shows the assumed subdiaphragm where existing roof 

sheathing is not being modified. The subdiaphragm depth will be controlled by the capacity of 

the existing sheathing. The wall anchor engages each wall purlin across the subdiaphragm depth. 
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Existing subpurlin-to-sheathing nailing must be adequate to transfer the wall anchor force to the 

subdiaphragm. In Figure 22.2.3-1B the added wall anchor is located between existing subpurlins 

in order to engage more existing sheathing nailing. Sheathing fastening to subpurlins must be 

assumed to be field nailing unless edge nailing has been confirmed. 

Figure 22.2.3-1B: Subdiaphragm for Flexible Wood Diaphragm – 

Roofing Not Removed 
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Figure 22.2.3-1C: Subdiaphragm for Flexible Wood Diaphragm – 

Roofing Removed 
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Figure 22.2.3-1D: Enhanced Wood Subdiaphragm with 

Added Wood Structural Panel Soffit 

Figure 22.2.3-1C depicts a subdiaphragm where access from the top is assumed, and the 

subdiaphragm can be renailed to meet required demands. A new member is provided at the wall 

anchor. Tie-downs are added to carry the wall anchorage force across the entire subdiaphragm 

width. Figure 22.2.3-1D illustrates a third subdiaphragm alternative where new subdiaphragm 

sheathing is provided as a soffit at the underside of the roof framing. Wall out-of-plane 

anchorage is not shown, but would be similar to Figure 22.2.3-1C. Attention is needed to 

providing shear transfer into the main diaphragm at all subdiaphragm boundaries. See other 

chapters for additional discussion of wall anchorage. 
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Figure 22.2.3-1E: Subdiaphragm for Flexible Wood Diaphragm at Purlins 

Work can be conducted either from the underside or the top of the diaphragm. Location of access 

needs to be decided early on in the design process and will drive both calculations and detailing 

of the rehabilitation work. Where the roofing is not going to be removed, it is possible to 

strengthen the diaphragm in local areas by sheathing the underside of the roof subpurlins, as 

shown in Figure 22.2.3-1D. This is expensive and tedious work that should not occur over large 

areas, but may be advantageous for reinforcing of subdiaphragms in combination with wall 

anchorage. 

Figure 22.2.3-1E illustrates anchorage of the north and south walls into subdiaphragms extending 

between Lines 1-2 and 3-4. 
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Figure 22.2.3-1F: Cross-Tie for Flexible Wood Diaphragm at Glulam Beams 

Figure 22.2.3-2A illustrates a similar roof plan with a steel diaphragm. Figures 22.2.3-2B 

through 22.2.3-2D provide details. Instead of using subdiaphragms, direct ties are provided. 

Alternative connections locations for field welded connections between joists (Figures 22.2.3-2C 

and 22.2.3-2D) include the joist top chord, vertical and horizontal legs. The alignment of joists at 

support locations will greatly affect the connection detail used, so field determination of detail 

and alignment should be made. See Section16.4.1 for additional discussion. 
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Figure 22.2.3-1G: Cross-Tie for Flexible Wood Diaphragm at Purlins 

Design and Detailing Considerations 

Research basis: No research relating to the performance or adequacy of enhanced anchorage 

methods has been identified; however, the demands created in flexible diaphragms have been 

studied by Fonseca, Wood and Hawkins (1996); Hamburger and McCormick (1994); and Ghosh 

and Dowty (2000). 

The reader is referred to the extensive discussion in the SEAONC Guidelines for design and 

detailing considerations for the wood diaphragm. 
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Figure 22.2.3-2A: Roof Plan with Diaphragm Cross-Tie System Using Direct Ties, 

Shown for Steel Diaphragm 

Cost, Disruption and Construction Considerations 

When rehabilitation work is undertaken on the roof diaphragm, it is important that the cost and 

the preferred location for work take into account the combination of work, rather than 

considering one piece at a time. If several diaphragm measures will be undertaken, it will quickly 

become cost-effective to remove the roof and allow work from the top. This is particularly true if 

a steel deck requires several rehabilitation measures. 

Proprietary Concerns 

There are no proprietary concerns with this rehabilitation technique other than the use of 

proprietary connectors and adhesives as part of the assemblage. 
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Figure 22.2.3-2B: Cross-Tie for Flexible Steel Diaphragm 

Figure 22.2.3-2C: Steel Open Web Joist Connection 

for Diaphragm Cross-Ties 
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Figure 22.2.3-2D: Steel Open Web Joist Connection 

for Diaphragm Cross-Ties 

22.2.4 Infill Opening in a Concrete Diaphragm 

Deficiencies Addressed by the Rehabilitation Technique

Inadequate diaphragm shear or chord capacity at existing opening. 

Description of the Rehabilitation Technique 

Addition of a structural infill to close an existing opening is a relatively simple method of 

correcting this type of local diaphragm deficiency in a concrete diaphragm. The new infill will 

reduce concentrated shear and chord force demand in the surrounding diaphragm and eliminate 

the need for often nonexistent local chords around the edges of the opening. In almost all cases, 

the new infill will be made with cast-in-place reinforced concrete or shotcrete. While it is 

conceivable, and perhaps possible in some unusual cases, to close the opening with steel plate or 

a precast concrete “plug,” the connections to the surrounding slab are very problematic, and their 

effectiveness as a mitigation measure is doubtful. 

Design Considerations 

Gravity load support: In addition to diaphragm shear demand, a new infill of an existing 

opening will create new floor or roof area which must be designed to support its self weight and 

the associated live load. In addition, the surrounding floor or roof system must be capable of 

supporting the gravity loads delivered from the newly infilled area. For larger infills, new beams 

may be required, both in the infill area and at the affected surrounding slabs, to provide this 

capacity. 
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Detailing Considerations 

Connection to existing concrete floor and roof diaphragms: Typical details of a reinforced 

concrete (cast-in-place or shotcrete) infill are indicated in Figure 22.2.4-1. Sufficient dowels 

must be placed into the existing diaphragm slab on all sides of the opening to transfer the 

required shear demand to and from the infill section. Forms may be supported from the floor 

below or suspended from the surrounding floor or roof. This latter option is much more common 

for smaller openings or for openings surrounded by waffle ribs, pan joists or beams. Since the 

concrete infill will shrink relative to the surrounding slab, some care should be given to use 

shrinkage compensated mix. 

Figure 22.2.4-1: Typical Infill Opening in a Concrete Diaphragm 

Cost/Disruption Considerations 

The cost of this type of infill is very modest and will generally be a very small component in the 

overall retrofit project. Except for the noise and vibration associated with the dowel drilling, 

disruptions associated with this type of infill will be very localized, affecting only the immediate 

surrounding floor area and the area on the floor below. 
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Construction Considerations 

The existing concrete surfaces around the entire perimeter of the existing opening to be in 

contact with the new concrete infill should be thoroughly cleaned of all finishes, paint, dirt, or 

other substances and then be roughened to provide 1⁄4” minimum amplitude aggregate interlock at 

joints and bonded surfaces. Alternatively, a lower μ-factor and more dowels can be used with 

less roughening. 

For shotcrete applications, separate test “panels” should be made to represent the slab infill work 

in addition to the normal test panels for shear walls. Nozzle operators should have several years 

experience with similar structural seismic improvement applications. 

22.2.5 Add Fiber-Reinforced Polymer Composite Overlay to a Concrete 

Diaphragm 

Deficiencies Addressed by the Rehabilitation Technique 

Inadequate shear capacity in a slab 

Description of the Rehabilitation Technique 

The use of an FRP overlay with slabs for in-plane shear strength (diaphragm shear) enhancement 

is a very new technique that has had limited implementation. For shear enhancement of 

monolithic slab construction, the fibers are oriented parallel to the applied shear direction. The 

technique is also used for precast floor systems, where the shear plane is the joint between 

panels. Joint strengthening usually employs bi-directional fibers orientated at 45 degrees to the 

shear plane. 

Design Considerations 

Research basis: Although there has been a significant amount of research conducted on flexural 

strengthening of concrete slabs or strengthening of bridge decks using FRP overlays, published 

research focused specifically on strengthening of concrete diaphragms using FRP overlays has 

not been identified. Designers have typically considered results of tests performed on FRP 

composite strengthened shear walls relevant for diaphragm strengthening applications. 

Chord and collector considerations: The diaphragm usually resists seismic loads in both 

directions, which requires bi-directional fiber orientation. 

While shear transfer between two concrete elements has been tested and proved to be reliable, 

there are diaphragm internal forces termed chord and collector forces. This rehabilitation 

technique, which may have been intended solely as a shear enhancement may, in fact, have chord 

and collector force demands. 

Chord actions, which develop from in-plane flexing of the full diaphragm depth, are developed 

in boundary elements gradually over the span length. These forces can be very high. The 

limited bond capacity and difficulty of anchoring the FRP composite may prohibit development 

of such large forces. Further, the strain limitations of the FRP composite prevent significant 

yielding; hence, the diaphragm chord forces should be based on the diaphragm forces required to 

yield the vertically-oriented elements of the lateral force-resisting system. This force level 
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would be similar to a code level force multiplied by omega, an over-strength factor, which is the 

same force level used to design diaphragm collectors. 

The use of FRP composite overlay to provide collector type load transfer is more difficult than 

that for chords. The collector force is usually being transferred from the diaphragm to a 

concentrated location, such as a brace frame or shear wall element. Strain compatibility and 

anchorage issues discussed with the bond-critical application (see Section 13.4.1, “Enhance 

Shear Wall with Fiber-Reinforced Polymer Composite Overlay, Fiber-Reinforced Polymer 

Overview, Requirements at the FRP-to-Substrate Interface”) prevent reliable transfer of the 

collector force to the frame of wall element. 

If, however, this technique must be used, then the bond, load transfer, strain compatibility, 

uncertainty in diaphragm demand forces, etc. must be carefully considered and reflected in the 

design and details. 

See Section 13.4.1, “Enhance Shear Wall with Fiber-Reinforced Polymer Composite Overlay, 

Fiber-Reinforced Polymer Composite Overview,” for background information. 

Detailing Considerations 

Given the high dependence on the bond strength of the FRP overlay to the substrate, in situ bond 

testing is recommended as part of the contract documents. A testing program will verify the 

design assumptions and assist in providing quality assurance. The vertical offset between the 

two slabs should be minimized. This can be achieved by removing surface projections and 

applying leveling compound to ensure that the FRP composite overlay does not exceed the 1-2% 

out-of-plane angle. Offsets exceeding this limit or lack of bond between the leveling compound 

or substrate and the polymer may cause premature delamination. 

In many situations, improvement in shear transfer capacity at the edge of the diaphragm will be 

needed in addition to enhancement of the capacity of the diaphragm itself. Transfer details from 

the slab to the wall using FRP need careful consideration. See Figure 22.2.5-1. Typically, the 

fiber is lapped from the slab to the wall, and fibers are oriented at 45 degrees (in plan view) to 

the length of the wall. The 90 degree bend in the fiber at the turn to the wall creates several 

issues. First, preparation of the existing sharp corner with resin putty is needed to allow a 

reasonable radius for the fiber. Second, when shear forces develop, they create tensile forces in 

the fiber. Because of the bend in the fiber, a substantial out-of-plane component is developed 

which must be resisted. The bond stress of the fiber has limited capability to take this force, 

usually leading to the need to reinforce the bend with mechanical means. A cut pipe placed 

against the corner, matching the radius of the curve, can be anchored with drilled dowels through 

the fiber to the wall or slab. Finally, testing to date of slab-to-wall shear transfer details is 

limited, necessitating increased caution. 

Construction Considerations 

Should underside of slab strengthening be used, the utilities at this location may need to be 

removed and reinstalled. This could impact building function during the construction period, 

and will add to the construction cost. For above slab strengthening architectural finishes, 

thresholds, and slopes will need to be considered. 
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Proprietary Concerns 

See Section 13.4.1 for brief discussion of proprietary concerns. 

Figure 22.2.5-1: Shear Strengthening of Concrete Diaphragm Using FRP Composite 

22.2.6 Infill Opening in a Concrete Fill On Metal Deck Diaphragm 

Deficiency Addressed by Rehabilitation Technique 

Increase diaphragm shear and/or chord capacity by infilling opening. 
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Description of the Rehabilitation Technique 

Adding infill to an existing opening is a simple method of reducing local stresses around the 

opening as well as the demand on the diaphragm. However, this technique can only be 

employed if an existing opening is no longer necessary for the function of the building. Thus, it 

would likely have to coincide with other building renovations that eliminate the function of the 

opening. The opening may have been used for stairs, an elevator shaft, a pipe and conduit shaft, 

or an atrium. The infill should be constructed in a similar manner as the existing diaphragm 

when possible, using similar types of metal deck and concrete as well as reinforcing steel layout. 

This ensures that the infill matches the strength and stiffness of the surrounding diaphragm. The 

new metal deck can be connected to the existing deck with welds or fasteners while the new 

reinforcing steel bars are doweled into the edges of the opening. The edges of the opening 

should be roughened to ensure adequate bond between the new and existing concrete. For 

smaller openings, it may be acceptable to span the opening with a flat piece of gauge steel 

instead of metal deck, provided that proper measures are taken to fill the openings between the 

deck flutes. 

Design and Detailing Considerations 

Gravity loads: The infill has to support its self-weight and additional dead and live loads. The 

surrounding floor system should also be evaluated for these new loads. At larger infills, new 

steel framing may be required either directly below or at the edge of the infill.

Metal deck attachment: The new metal deck should overlap the existing metal deck around the 

perimeter of the opening. The deck can be attached to one another with puddle or seam welds, or 

mechanical fasteners, which may include expansion anchors, screws, or shot pins. 

Bar development: Details of the reinforcement are similar to that for infilling an opening in a 

concrete diaphragm shown in Figure 22.2.4-1. Development lengths for the same size 

reinforcing bar will vary depending on the grout or adhesive product used to dowel the bar into 

the existing concrete. Bars on opposite sides of the openings should be spliced inside the 

opening. At smaller openings, the splice lengths will be limited by the size of the opening. The 

bars can be hooked in these cases for development. Adding bars to thin slabs will be difficult, 

particularly in the direction perpendicular to the metal deck flutes. Existing bars that are parallel 

to the flutes may be damaged while drilling holes for the new dowels. As an alternative, it may 

be easier to use welded wire fabric (WWF) instead of reinforcing steel. The slab would have to 

be chipped back around the opening to allow for development of the WWF. 

Cost/Disruption 

The cost associated with this technique is minimal compared to other diaphragm strengthening 

techniques, such as adding concrete overlays or horizontal braced frames. Since the infilling of 

an opening is likely related to other changes to a building, the disruption caused by the other 

changes are often more significant. 

Construction Considerations 

See Section 8.4.1 for general discussions of welding issues, removal of existing nonstructural 

and structural elements, and construction loads. 
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Proprietary Concerns 

Many grout and adhesive products are available. 

22.2.7 Increase Shear Capacity of Unfilled Metal Deck Diaphragm 

Deficiency Addressed by Rehabilitation Technique 

Strengthen inadequate bare metal deck diaphragm. 

Description of the Rehabilitation Technique 

Metal deck diaphragms are governed by either the capacity of the deck or its connection to other 

components of the lateral force-resisting system. Connection capacity is limited by the strength 

of the welds or other mechanical fasteners. At locations where welds or fasteners cannot be 

directly added, such as concrete walls, the addition of a steel angle connected with expansion 

anchors or adhesive dowels to a wall and diaphragm is often feasible. The capacity of a 

longitudinal joint between deck units is limited by the strength of the crimps or seam welds. 

These connections should be upgraded to the strength of the metal deck to achieve ductile 

diaphragm behavior during an earthquake. If the connections can develop the metal deck 

capacity, but the deck is found to be inadequate, significant increases in capacity may be 

obtained by adding a reinforced concrete fill or horizontal braced frame (Section 22.2.9). 

Design and Detailing Considerations 

Connections: In order to enforce deformation compatibility, new connections should be 

constructed similarly to the existing connections. Thus, puddle welds should be used if the 

existing diaphragm is welded to the steel framing. Similarly, the same types of mechanical 

fasteners should be used to match the existing fasteners when screws, shot pins, or expansion 

anchors are found at the connections. 

Deck stiffeners: Some deck manufacturers fabricate stiffeners specifically intended for use with 

unfilled metal decks. The stiffeners are constructed to match the profile of the decks, which 

provide additional stiffness at the supports and in turn, increase the strength of the diaphragm. 

The stiffeners are typically welded to the deck and the steel beams. 

Concrete fill: When reinforced concrete is added over metal deck, a shear transfer mechanism 

from the concrete to the lateral force-resisting system is required, e.g. welded shear studs at steel 

beams and drilled dowels at concrete members. Since the addition of a concrete overlay will 

increase the dead weight of the structure, the existing forces, members, connections, and 

foundation must be checked to determine whether they are capable of resisting the added loads. 

Cost/Disruption 

Diaphragm connection upgrades can be performed efficiently to minimize disruption and are cost 

effective if upgrades to other parts of the lateral force-resisting system are not required. If 

concrete fill is added, cost and disruption could increase significantly if upgrades are required to 

other parts of the lateral force-resisting system. Also, nonstructural elements such as insulation 

fill, roofing, and partitions would all require temporary removal. 
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Construction Considerations 

See Section 8.4.1 for general discussions of welding issues, removal of existing nonstructural 

and structural elements, and construction loads. 

Proprietary Concerns 

Metal deck stiffeners are only provided by some manufacturers for use with their decks. 

22.2.8 Enhance Masonry Flat Arch Diaphragm 

Deficiency Addressed by Rehabilitation Technique 

A relatively common type of floor in a masonry building or steel frame infill building, 

particularly outside the West Coast, uses narrowly spaced steel beams to support shallow or 

“flat” arches of masonry. The masonry can be made of hollow clay tile or brick. It is usually 

bearing on the bottom flange of the steel beam and supports nonstructural and acoustic fill above 

it. The horizontal kick from the base of the arch is balanced in the diaphragm interior by the 

adjacent arch. At the exterior, this kick either goes into the wall, or a tension tie of steel is 

provided at the bottom of the beams. In some cases, the steel strapping or bars run the full width 

of the diaphragm. When a tension tie is missing at the base of the arch and the diaphragm 

vibrates and expands, localized gravity failure can result when loss of arching action occurs. At 

the exterior of the diaphragm, the unbalanced kick of the arch can add to out-of-plane demands 

on the wall, contributing to out-of-plane wall failure and loss of vertical support. See Figure 

22.2.8-1 for examples of failure scenarios. 

Description of the Rehabilitation Techniques 

There are several rehabilitation techniques for masonry flat arches floors. They can be combined 

for economy of scale. 

Wall-to-diaphragm tension ties: Figure 22.2.8-2 shows the addition of tension ties from the wall 

to the steel beams for conditions when the beams are perpendicular to the wall and when they are 

parallel. When beams are perpendicular, an angle and drilled dowel is sufficient. When beams 

are parallel, strapping back to joists inside the floor is necessary. Figure 22.2.8-3 shows an 

example of placing the strapping on top of the beams, in case this is the preferred location for 

work. 

Wall-to-diaphragm shear ties: The drilled dowels in Figure 22.2.8-2 also serve as ties for 

transferring shear forces from the edge of the diaphragm into the wall. 

Chord: If the angle in Figure 22.2.8-2 is continuous, it can serve as a diaphragm chord. 

Interior tension: While providing a tension tie for the case when the beams are parallel to the 

wall next to the wall is the most critical priority, it is desirable as well to continue the strapping 

all the way across the floor so local interior failure does not occur. Figure 22.2.8-4 shows the 

straps, plus notes the tension and shear ties and the chord. 
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Figure 22.2.8-1: Failure Scenarios for Masonry Flat Arch Floors 
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Figure 22.2.8-2: Add Wall-to-Diaphragm Ties and Chord for Masonry Flat Arch Floor - 

Access from Below the Floor 
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Figure 22.2.8-3: Add Wall-to-Diaphragm Ties and Chord for Masonry Flat Arch Floor - 

Access from Above the Floor 

Diaphragm strengthening: Figure 22.2.8-4 also shows how adding diagonal bracing can be 

combined with existing beams and straight to create a horizontal braced frame diaphragm. 

Topping slab: Theoretically, part of the flooring substrate can be replaced with a reinforced 

concrete diaphragm, though the vertical capacity of the floor would need to be sufficient and the 

weight of the new concrete adds to the inertial weight of the building. 

Design Considerations 

Research basis: No research specific to seismic rehabilitation of flat arch floors has been 

identified. There is also very limited information about how the floors have performed in actual 

earthquakes. There was some damage in the 1906 San Francisco Earthquake reported for these 

floors (Himmelwright, 1906) though much of the damage was due to fire. There are photos of 

the flat arch roof failures and reports of significant damage in Iranian earthquakes when tension 

ties are not present (Alimoradi, 2005). 
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Figure 22.2.8-4: Masonry Flat Arch Floor Strengthening 

Shear capacity: This type of floor has not been addressed by recent evaluation publications like 

FEMA 273 (FEMA, 1997a), FEMA 274 (FEMA, 1997b), FEMA 356 (FEMA, 2000), or ASCE 

31-03 (ASCE, 2003), so capacity evaluations are from first principles. One strategy is to take all 

of the lateral force resistance in the new diaphragm strengthening due to the lack of 

interconnections in the diaphragm. Another approach is to develop strut-and-tie models in the 

diaphragm with the new and existing steel as ties and the masonry as a strut. 

Stiffness: Although this floor lacks interconnections, it is likely to be quite stiff, as well as 

extremely heavy. 

Detailing and Construction Considerations 

Floor types: Lavicka (1980) is a reprint of an 1899 textbook on turn-of-the-century construction 

techniques and has an excellent summary of masonry flat arch variations. The system was 

intended to provide improved fireproofing and acoustic benefits. Flat tile arches were popular 

and had flat top and bottom surfaces to the tile, but beveled edges to create internal arching 
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action. Side method arches had the voids in the hollow clay tile parallel to the beams; end 

method arches oriented the voids perpendicular to the beams. There were combinations of the 

orientations as well. The tile at the steel beam was usually notched around the bottom flange to 

provide masonry cover of the bottom of the bottom flange. Tile depths range from 6” to 12” 

with beams spaced from 3’6” to 7’6”. Segmental tile arches had shallow arches of several inches 

at the crown, the voids were parallel to the beams, and the end tile would bear on top of the 

bottom flange. Other systems have been observed to include clay bricks oriented with the long 

direction of the brick perpendicular and parallel to the beams. The masonry arches often 

supported a fill of cinders, sometimes mixed with mortar. This in turn would support wood 

sleepers spanning over the top of the steel beams and a wood floor. Tension ties were 

recommended; they were to be 3⁄4” diameter rods placed near the bottom of the steel beam web 

and at about a spacing of 7’-8’. 

Bottom cover: Figure 22.2.8-2 shows clay tile floors covering the bottom of the bottom flange. 

There is typically plaster adhering to the masonry. To install steel strapping, the plaster and 

masonry must be notched. Figure 22.2.8-3 shows an alternative to avoid damaging the 

underside by adding steel plate or straps, but working from the top. Of course, this is quite 

disruptive to occupants as well. In some arch types, though, the bottom flange is not covered and 

adding steel from below is much less disruptive. 

Cost/Disruption 

Rehabilitation of a masonry flat arch floor can be quite disruptive and expensive, particularly 

when ties are necessary at the building interior and if plaster ceilings and masonry or floors must 

be temporarily removed and patched. 

Proprietary Issues 

There are no proprietary concerns with diaphragm improvements in masonry flat arch floors. 

22.2.9 Add Horizontal Braced Frame as a Diaphragm 

Deficiency Addressed by Rehabilitation Technique 

Strengthen inadequate diaphragm. 

Description of the Rehabilitation Technique 

Providing a horizontal braced frame as a diaphragm strengthening technique is useful if the 

existing floor cannot be disturbed for functional reasons or the cost of replacing the existing 

diaphragm is more expensive (e.g., a sloped roof). This is also an alternative when concrete 

overlays add too much mass or lead to other construction complications. The existing diaphragm 

could be constructed of concrete filled or unfilled metal deck, or wood. The new horizontal 

bracing is added under the existing diaphragm, in which the existing framing with new diagonal 

members forms the horizontal bracing system. The diaphragm shears are shared with the 

existing diaphragm in proportion to the relative rigidity of the two systems. The design 

philosophy is generally to have the diaphragm remain essentially elastic, with the goal of 

achieving ductile inelastic behavior in the vertical lateral force-resisting elements. See Chapter 9 

for a general discussion of braced frames. 
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Design Considerations 

Force distribution: The diaphragm strength could be evaluated by considering boundary 

solutions. First, its capacity including both the existing diaphragm and the horizontal braced 

frame is determined based on their relative rigidities. This alternative may not be always be fully 

effective if the existing diaphragm has much greater rigidity of that of the bracing system, such 

as metal deck with heavily reinforced concrete fill. Thus, an evaluation should also be 

performed assuming failure of the concrete fill. The diaphragm strength would only include that 

of the braced frame with minimal contribution from the metal deck without the concrete fill. If 

the latter solution yields a greater value, extensive cracking of the concrete fill and greater 

diaphragm displacements would be assumed to be acceptable. 

Sloped roofs: The horizontal braced frames could be sloped to match the roof slopes, which 

would require proper consideration of the slopes and their effects on the diaphragm forces. 

Alternatively, the braced frames could have a flat layout, but this may affect the functional space 

as well as aesthetics. 

Brace members: Similar to the selection of members in braced frames, compact and non-slender 

sections are preferred for their ductility. Installation of the braces should be factored into their 

selection due to the logistics associated with delivering and attaching the braces to their final 

locations. Note the self-weight of the braces adds a component to the flexural forces that may be 

reduced by adding hanger rods. 

Chords and collectors: The new horizontal bracing system requires continuous chord and 

collector members to receive the brace forces and transfer these forces to the lateral forceresisting elements. The existing members that serve this purpose should be used when possible, 

as shown in Figure 22.2.9-1. 

Detailing Considerations 

Connections: For steel structures, the braces can be welded or bolted with or without gusset 

plates to the existing framing. An example of a welded connection is shown in Figure 22.2.9-2. 

Bolting eliminates welding issues that include space restrictions and venting weld fumes while 

welding may permit smaller and more compact connections. In concrete structures, connection 

of the new horizontal bracing system to the existing vertical system is accomplished by welding 

braces to plates that connect to the walls or frames with mechanical fasteners, such as threaded 

dowels and expansion anchors. 

Cost/Disruption 

These costs of adding horizontal bracing must be weighed against that of a concrete overlay. 

Temporary removal or relocation of nonstructural elements such as piping and partition walls are 

required and should be included in the cost evaluation for both options. The horizontal braced 

frame requires connection modifications, which are locally very disruptive. 

Construction Considerations 

The engineer’s involvement during the construction phase is critical during a seismic 

rehabilitation. The design of the retrofit scheme must not neglect the construction phase and 

should consider these issues at a minimum: 
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Figure 22.2.9-1: Diaphragm Strengthening using Horizontal Braced Frame 

Welding/bolting issues: See general discussion in Section 8.4.1. Primary issues associated with 

bolting consist of typical field bolting issues such as set up, fit-up, and alignment. 

Removal of existing nonstructural elements: This technique requires access to the underside of 

the floor or roof framing and may require relocation of piping, ducts, or electrical conduits as 

well as difficult and awkward connections to the existing framing. See Section 8.4.1 for 

discussions of fireproofing, asbestos, and concrete encasement. 

Removal of existing structural elements: Existing structural elements do not typically have to be 

removed to add horizontal steel bracing. However, if required, shoring and temporary bracing 

may be necessary. 

Construction loads: See general discussion in Section 8.4.1. 

Proprietary Concerns 

There are no known proprietary concerns with this technique. 
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Figure 22.2.9-2: Horizontal Braced Frame Connection 

22.2.10 Improve Tension Rod Horizontal Steel Bracing 

Deficiency Addressed by Rehabilitation Technique 

Repair nonductile tension rod bracing and/or connections 

Description of the Rehabilitation Technique 

Tension rod bracing consist of rods that are spliced together by turnbuckles and connected to 

clevis pins at the ends. The clevis pins are bolted to typical gusset plates. Tension rods that are 

inadequate for the seismic demands should be replaced entirely since it would probably be more 

complicated to upgrade existing rods. Increasing the rod size also requires replacing the 

turnbuckles and clevis pins. Connections that are inadequate can be upgraded similarly as 

typical braced frame connections. An example of a typical rod connection to a concrete or CMU 

wall is shown in Figure 22.2.10-1. The connection to the wall should develop the strength of 

the rod. 
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Figure 22.2.10-1: Tension Rod Connection at Wall 

Design and Detailing Considerations 

Tension rod bracing is used in applications where seismic forces are relatively low. It would be 

most appropriate for unfilled metal deck or wood diaphragms. The rod upgrades may increase 

the stiffness of the existing diaphragm and the total diaphragm force. Thus, all other elements of 

the lateral force-resisting system—connections, chords, collectors, frames or walls, and 

foundations—should be evaluated and upgraded accordingly. 

Cost/Disruption 

Replacing tension rods is fairly efficient on both a cost and time basis compared to other types of 

diaphragm upgrades. Connection modifications will only be locally disruptive and can be 

performed rapidly. 

Construction Considerations 

See Section 8.4.1 for general discussions of welding issues, removal of existing nonstructural 

and structural elements, and construction loads. Also see Section 22.2.9 for a discussion of 

construction issues related to modification of horizontal steel bracing. 

Proprietary Concerns 

There are no known proprietary concerns with this technique. 

22.2.11 Improve Shear Transfer in Precast Concrete Diaphragm 

Deficiency Addressed by Rehabilitation Technique 

Inadequate diaphragm strength and/or stiffness 
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Precast diaphragm deficiencies and observed behavior have been discussed in some detail in 

Chapter 17; the reader is referred to these discussions. To date, construction of precast buildings 

in areas of high seismic hazard in the U.S. has been of limited quantity, resulting in limited 

opportunities to observe earthquake performance. The poor performance of some long-span 

precast diaphragms in parking structures in the 1994 Northridge earthquake has raised questions 

about shear capacity in diaphragms with topping slabs, excessive diaphragm deformation due to 

performance of chords and collectors and the interaction of shear and flexure. The complete lack 

of connection between hollow core floor planks within diaphragms appears to have been a 

primary contributor to collapse of nine-story residential precast concrete frame buildings in the 

1988 Armenia earthquake (EERI, 1989). 

Description of the Rehabilitation Techniques 

There are three types of precast concrete diaphragms commonly used: topped precast tee-beam, 

untopped precast tee-beam, and untopped precast hollow-core plank. Topped precast hollow-core 

may be used on occasion, but is not as common. 

To date, very little rehabilitation of precast diaphragms has occurred in the U.S. As a result, the 

following discussion of rehabilitation measures draws from limited available research, suggested 

details for new precast construction, and application of rehabilitation techniques for concrete 

buildings to the specific configurations of precast elements. 

Fiber-reinforced polymer (FRP) composite overlays provide one possible approach to shear 

connections between adjacent precast diaphragm members, and overlays could be used for any of 

the three common systems noted above. For parking structures, attention to both ultraviolet (UV) 

ray exposure and wearing under vehicle loads would be important to performance. The FRP 

overlay could be applied continuously over the area of high diaphragm shear and then used to 

transfer loads into supporting shear walls or frames, or applied locally at each member joint. 

Research by Pantelides, Volnyy, Gergeley, and Reaveley, (2003) on FRP composite connection 

between wall panels may be of interest; however, the reader is cautioned to consider the effects 

of simultaneous shear and tension at joints. See Section 17.4.2. 

For untopped hollow-core diaphragms it may be possible to add construction roughly equivalent 

to that used for new construction. In new design, where diaphragm shear stresses exceed those 

allowed for grout key shear transfer, the cast concrete beams at the diaphragm perimeter or 

interior are used as flexural elements, resisting horizontal diaphragm forces. New beams could 

be added to serve this purpose. Connection between the precast sections and the new beams, 

either by bearing or mechanical connection is required. Attention to adequate strength and 

stiffness is also required. 

Bolted steel plate connections providing shear connections from panel to panel are another 

possible approach. This involves use of a continuous plate or series of plates crossing the precast 

panel joint, with adhesive or expansion anchors on each side of the joint. Steel plate thickness 

must be selected in order to avoid plate bucking between connections. See Chapter 17 for 

discussion of anchors. 
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Design Considerations 

Research basis: No research applicable to rehabilitation of precast diaphragm strength and 

stiffness has been identified; however, the following research for new construction may provide 

some guidance for rehabilitation: 

A significant integrated analytical and experimental research program is currently 

underway to develop a comprehensive design methodology for precast concrete 

diaphragm systems. The project intends to address the discrepancy between current 

design practice, based on inelastic behavior concentrating in vertical elements, and 

observed performance in which substantial inelastic behavior has occurred in diaphragms 

(Wan et al., 2004; and Naito and Cao, 2004). The project proposes to determine force and 

deformation demands required for design, connection details to support the performance, 

and address deformation relative to the gravity load-carrying system. This information 

will be invaluable for both new design and rehabilitation. Testing will include individual 

connections, joints, and half-size components. Analytical modeling of full buildings is 

being used to identify critical demands. Of particular interest is the simultaneous 

occurrence of shear and tension or compression on connections normally considered to 

carry only shear. Published information to date (Naito and Cao, 2004) provides a 

database of connector properties from existing literature and suggests a simplified 

analysis model based on initial finite element testing. Additional information should be 

available over the next several years. 

Shear Diaphragm Capacity of Untopped Hollow-Core Floor Systems (Concrete 

Technology Associates, 1981) describes testing of grouted hollow-core joints. Note that 

issues raised by the Northridge earthquake might imply modification of testing approach. 

Research by Pantelides, Volnyy, Gergeley, and Reaveley, (2003) on FRP composite 

connection between wall panels. 

Research by K.S. Elliott, University of Nottingham, on untopped hollow-core 

diaphragms. 

Basic design approach: The PCI Handbook (PCI, 1999) and Design and Typical Details of 

Connections for Precast and Prestressed Concrete (PCI, 1988) are basic references for design 

and detailing of precast concrete structures. These documents discuss the use of grouted keys for 

diaphragm-to-diaphragm connections, and they also recognize use of friction connections, 

without positive anchorage for wall-to-diaphragm connections. Use of these mechanisms must be 

given very careful consideration for possible inelastic seismic demands. 

Shear and flexure interaction: One of the issues identified from the performance of parking 

structures in the Northridge earthquake is the interaction of shear and flexural deformations. 

Diaphragm deformations will result in tension and compression forces between adjacent 

diaphragm members. As a result, shear connections between members will need to accommodate 

simultaneous tension or compression plus shear. It is recommended that the diaphragm chord and 

collector members also be evaluated, and rehabilitated if necessary to control tension forces. It is 

also recommended that the rehabilitation measure chosen be capable of withstanding anticipated 

simultaneous forces. Methods of estimating diaphragm demands are proposed in Nakaki (1998) 

and Naito and Cao (2004). 
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Transfer into and out of diaphragm reinforcing: Transfer of loads is key to the use of fiber 

composites or steel plate for connecting between precast diaphragm segments. Where panel-topanel connections are made, it is necessary to transfer the full design load in and out at each 

connection. In some cases it may become more practical to provide reinforcing over the entire 

diaphragm or highly loaded sections of the diaphragm; the load portions of load carried in the 

existing diaphragm and the reinforcing would need to be determined by deflection compatibility. 

Topping slabs: The addition of a topping slab is seldom a practical approach because of the 

added weight for vertical loads and mass for seismic loads. In rare cases where additional 

vertical load capacity has been provided, this may be possible. The additional capacity is needed 

not only in the diaphragm slab and beam system, but in all of the vertical support system through 

the foundation. The removal and replacement of a topping slab could permit the addition of 

reinforcing and connections without increasing gravity or seismic loads. This, however, is a 

costly process. 

Proprietary Concerns 

Fiber composite materials and adhesive and mechanical anchors are proprietary and must be 

used in accordance with manufacturer and ICC-ES requirements. 

22.3 References 

ABK, 1981, Methodology for Mitigation of Seismic Hazards in Existing Unreinforced Masonry 

Buildings: Diaphragm Testing, A Joint Venture of Agbabian Associates, S.B. Barnes and 

Associates, and Kariotis and Associates (ABK), Topical Report 04, c/o Agbabian Associates, El 

Segundo, CA. 

ASCE, 2003, Standard for the Seismic Evaluation of Buildings, ASCE 31-03, Structural 

Engineering Institute of the American Society of Structural Engineers, Reston, VA. 

AF&PA, 2005, National Design Specification for Wood Construction, ASD/LRFD, American 

Forest and Paper Association, Washington, D.C. 

Alimoradi, A., 2005, “Steel Frame with Semi-Rigid ‘Khorjini’ Connections and Jack Arch Roof 

‘Taagh-e-Zarbi’,” EERI World Housing Encyclopedia, www.world-housing.net; Report 25. 

APA, 2000, Research Report 138, Plywood Diaphragms, APA The Engineered Wood 

Association, Tacoma, WA. 

Concrete Technologies Associates, 1981, Shear Diaphragm Capacity of Untopped Hollow-Core 

Floor Systems (TCA 38), Concrete Technology Associates, Tacoma, WA. 

Dolan, J.D., D. Carradine, J. Bott, and W. Easterling, 2003, Design Methodology of Diaphragms, 

CUREE Publication No. W-27, CUREE, Richmond, CA. 

EERI, August 1989, Armenia Earthquake Reconnaissance Report, Earthquake Spectra Special 

Edition, Earthquake Engineering Research Institute, Oakland, CA. 
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FEMA, 1997a, NEHRP Guidelines for the Seismic Rehabilitation of Buildings, FEMA 273. 

Federal Emergency Management Agency, Washington, D.C. 

FEMA, 1997b, NEHRP Commentary on the Guidelines for Seismic Rehabilitation of Buildings, 

FEMA 274, Federal Emergency Management Agency, Washington, D.C. 

FEMA, 2000, Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 

356, Federal Emergency Management Agency, Washington, D.C. 

Fonseca, F., S. Wood and N. Hawkins, 1996, “Measured Response of Roof Diaphragms and 

Wall Panels in Tilt-Up Systems Subject to Cyclic Loading,” Earthquake Spectra, Volume 12, 

Number 4, Earthquake Engineering Research Institute, Oakland, CA. 

Ghosh, S.K. and S. Dowty, 2000, Anchorage of Concrete or Masonry Walls to Diaphragms 

Providing Lateral Support, Draft 2000, Not Published. 

Hamburger, R.O. and D. McCormick, 1994, “Implications of the January 17, 1994 Northridge 

Earthquake on Tilt-up Wall and Masonry Wall Buildings with Wood Roofs,” Proceedings of the 

1994 Convention of the Structural Engineers Association of California, Sacramento, CA. 

Himmelwright, A., 1906, The San Francisco Earthquake and Fire, A Brief History of the 

Disaster, A Presentation of Facts and Resulting Phenomena, with Special Reference to the 

Efficiency of Building Materials, Lessons of the Disaster, The Roebling Construction Company, 

New York, NY. 

ICBO, 1997, Uniform Code for Building Conservation, 1997 Edition, International Conference 

of Building Officials, Whittier, California. 

ICC, 2003a, International Building Code, International Code Council, Country Club Hills, IL. 

ICC, 2003b, International Existing Building Code, International Code Council, Country Club 

Hills, IL. 

ICC-ES, 2004a, 303 Siding and High-Load Diaphragms, ICC-ES Legacy Report ER-1952, ICC 

Evaluation Service, Inc, Whittier, CA. 

Lavicka, W., 1980, Masonry, Carpentry, Joinery, reissued version of 1899 textbook of the same 

name, Chicago Review Press: Chicago, IL. 

Naito, C. and L. Cao, 2004, “Precast Diaphragm Panel Joint Connector Performance,” 

Proceedings of the 13th World Conference on Earthquake Engineering, Vancouver, B.C. Canada. 

Nakaki, S., 1998, Design Guidelines: Precast and Cast-in-Place Concrete Diaphragms, 

Earthquake Engineeering Research Institute, Oakland, CA. 

Pantelides, C. P., V. A. Volnyy, J. Gergeley, and L.D. Reavely, Seismic Retrofit of Precast 

Concrete Panel Connections with Carbon Fiber Reinforced Polymer Composites, PCI Journal, 

Vol. 48, No. 1, January/February 2003, Precast/Prestressed Concrete Institute, Chicago, IL. 
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PCI, 1988, Design and Typical Details of Connections For Precast and Prestressed Concrete, 

Prestressed Concrete Institute, Chicago, IL. 

PCI, 1999, PCI Design Handbook, Precast and Prestressed Concrete, Fifth Edition, Prestressed 

Concrete Institute, Chicago, IL. 

Peralta, David, Joseph Bracci and Mary Beth Hueste, 2004, “Seismic Behavior of Wood 

Diaphragms in Pre-1950s Unreinforced Masonry Buildings,” Journal of Structural Engineering, 

ASCE: Reston, VA, Volume 130, Number 12, December, pp. 2040-2050. 

SEAOSC (Structural Engineers Association of Southern California), 1979, Recommended Tilt-up 

Wall Design (Yellow Book), SEAOSC, Los Angeles, CA. 

SEAONC (Structural Engineers Association of Northern California), 2001, Guidelines for 

Seismic Evaluation and Rehabilitation of Tilt-Up Buildings and Other Rigid Wall/Flexible 

Diaphragm Structures, SEAONC, San Francisco, CA. 

Wan, G., R. Fleishchman, C. Naito, J. Restrepo, R. Sause, L. Cao, M. Schoettler, and S.K. 

Ghosh, 2002, “Integrated Analytical and Experimental Research Program to Develop a Seismic 

Methodology for Precast Diaphragms,” SEAOC 2004 Convention Proceedings, Structural 

Engineers Association of California, Sacramento, CA. 
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Chapter 23 - Foundation Rehabilitation Techniques 

23.1 Overview 

While the need to add or supplement existing foundations for new superstructure elements such 

as shear walls and braced frames is relatively common in seismic rehabilitation, rehabilitation of 

existing foundation deficiencies is comparatively less common. There are two basic reasons for 

this: foundation work in existing buildings is quite expensive, and there has been relatively little 

note in earthquake reconnaissance reports of life loss and property damage resulting from 

foundation failures in buildings. 

Foundation analysis can be one of the most challenging areas of seismic rehabilitation. Different 

assumptions regarding base conditions of restraint, soil properties, and locations and types of 

potential nonlinearity can lead to widely varying results. For many buildings, it can take 

significant analytical effort in modeling and evaluating interim results to understand how the 

foundation interacts with the superstructure and surrounding soil under earthquake loading. 

Often, the weakest link or governing mechanism may be a foundation element or soil yielding, 

but it is only after looking at the substructure and superstructure as a whole that the sequence and 

nature of element behavior can be determined. 

In the past, force-based analytical techniques placed emphasis on strength capacity and whether 

the foundation and underlying soils were “overstressed”. With the advent of displacement-based 

analytical techniques, the extent of soil movement is acknowledged as more critical. Due to the 

cost and disruption of foundation rehabilitation work, the consequences of foundation deflection 

should be carefully evaluated to determine if there are actually going to be unacceptable 

movements. Large soil movements from rigid body rotation of a shear wall, for example, may 

have minimal consequences if the entire structure rotates, but they may have significant 

consequences to attached adjacent elements which are not rotating in phase or at all. 

When careful analysis reveals that new foundations must be added or that existing foundations 

must be enhanced, the structural engineer must have a good understanding of soil engineering 

issues; rehabilitation goals, performance criteria, and assumptions; and construction techniques 

and limitations. Obviously, it is usually much more difficult to perform work inside an existing 

structure than it is in a new building when the site is open. Because of the cost of foundation 

rehabilitation, other options should be fully explored, and the need for foundation modification 

should be thoroughly investigated. 

There are relatively few, if any, proprietary issues associated with foundation rehabilitation, 

though some equipment used to install new elements in limited access areas may have been 

developed by a specialty contractor and thus not widely available. 

This chapter provides a short discussion of general goals for foundation rehabilitation, brief 

mention of some key analytical considerations, and general construction issues; then provides 

discussion of structural rehabilitation techniques for foundations; reviews common ground 

improvement techniques; and ends with a short discussion of other ground hazards such as fault 

rupture, lateral spreading, and seismic-induced landsliding. 
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23.2 General Goals for Seismic Rehabilitation of Foundations 

The goal of any seismic evaluation is to identify deficiencies, their relative likelihood of 

occurrence, and the hazards they pose. The foundation must not be ignored during the 

evaluation, and foundation behavior response must be placed in the context of the overall 

performance of the building. If the foundation is identified as the weak link, the type of 

foundation mechanism needs to be identified A shear wall might be overstressed in shear or 

bending if assumed to have a fixed base, but when its small foundation is considered, rocking or 

overturning might be the governing mechanism. A braced frame might have adequate strength 

and stiffness, but the pile caps its columns sit on may not have any reinforcing to take uplift 

forces that occur beyond code level forces. Existing drilled piers may lack adequate confining 

ties in the top of the pier or insufficient lateral resistance in general or their connections to the 

pier cap may be insufficient. 

Consideration of the foundation is an integral part of the overall rehabilitation strategy for the 

structure. It may be possible to change the building behavior response by superstructure 

rehabilitation to preclude undesirable foundation modes. When foundation work is necessary, 

goals for rehabilitation design include providing sufficient strength, stiffness, and ductility for 

compression, tension, and lateral loading; identifying a defined and ductile mechanism of energy 

dissipation; and minimizing gravity stress redistribution within the existing foundation system. 

New foundations should not undermine existing foundations, either during construction or over 

the long-term. Moreover, the relative lower stiffness of unconsolidated soil under new 

foundations versus the higher stiffness under existing older foundations needs to be considered. 

23.3 Construction Issues 

Construction issues are quite critical during foundation work in existing buildings and will often 

drive the systems and techniques being considered. Issues include: 

Access and height restrictions: Installing shallow foundations, such as spread footings or 

grade beams, is usually done with hand methods or small excavation equipment and will 

rarely be a problem, though it will take longer than it would in a new building. Installing 

deep foundations, however, can run into several construction limitations. Drill rigs for 

piers, for example, are much more efficient when they are larger. Getting a drill rig into a 

building may require enlarging existing openings. Once inside, story heights will usually 

significantly limit the size of the drill rig that can be used. Special drills have been 

developed for use in existing buildings, but they often require at least 9 feet to 12 feet of 

vertical clearance. Drilling next to adjacent walls may limit the size of the pier or lead to 

shifting it inboard of the wall creating a horizontal eccentricity to be addressed.

Noise and vibrations limits: Pile driving imparts significant noise and vibration. Even if 

there were clearance outside the building for a pile driving rig, the vibration is usually too 

significant. Drilled piers impart less vibration, though the noise requires consideration. 

Micropiles have even less vibration and noise, so they are a common rehabilitation 

technique.

Restrictions imposed by existing utilities: Most buildings will have utilities beneath the 

existing ground level suspended floor or slab-on-grade. The locations and depths may 
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not be fully known. Excavating below grade requires careful effort, often with hand 

methods, so that utilities are not damaged.

Restrictions associated with ongoing operations: As with any rehabilitation work in the 

superstructure, if the building is occupied with people or equipment, foundation 

demolition, drilling, and excavation work will have to be coordinated.

Contaminated soil: There can be contaminated soil underneath the existing building, 

particularly if it has or had industrial uses. Removal of contaminated soil requires special 

techniques and must be taken to special landfills, increasing costs.

23.4 Analytical Issues 

This document’s focus is on detailing of rehabilitation techniques, not on analyzing the existing 

or rehabilitated structure, but it still worth pointing out a few analytical considerations in 

foundation modeling that often arise in seismic rehabilitation since codes and design guidelines 

provide limited guidance. 

Modeling the base of the building: The most basic question to be established is: Where is 

the dynamic base of the building? If there is no basement, this is straightforward. When 

there is a basement, partial basement, or sloped site, this is not a simple issue. Say that 

the building is four stories above grade and has a one-story full basement. Figure 23.4-1 

shows several possible modeling approaches. Model A is probably the most common 

approach—to stop the model of the superstructure at grade on a fixed base and take the 

results and impart them separately to the foundation walls and other elements. Model B 

is to ignore the ground entirely and put the base of the building at the bottom of the 

basement. When this is done, the inertial loads of the ground floor are usually not 

included. Model C is the same as Model B, except the ground floor loads are 

conservatively included. Model D changes the base conditions to account for vertical 

flexibility of the soil under the building. Significant modeling effort and variability have 

to be considered when springs are used. None of these models captures the “backstay” 

effect caused by the embedded foundation and the potential for shear reversals in the 

basement shear walls from soil pressures. To evaluate this effect, horizontal springs must 

be added to simulate the strength and stiffness of the surrounding soil, as shown in Model 

E. Note that this type of effect is similar but not the same as the backstay effect resulting 

from upper levels landing on larger, stiffer podium bases, which distribute local 

overturning loads out to other resisting elements using the diaphragms at the top and 

bottom of the podium. Nonlinearity can be added to the superstructure, substructure, and 

soil springs in these models as well. See below. 

Modeling soil stiffness: With displacement-based analytical seismic rehabilitation 

methodologies, understanding and quantifying displacements has becoming increasingly 

necessary. In the past, when displacement was considered, it usually was in the form of 

construction and long-term differential settlements between columns or the modulus of 

subgrade reaction for gravity loading under a mat or grade beam on soft soil. During 

seismic loading, we need stiffness values relevant to the short-term nature of earthquake 

demands. ATC-40 and FEMA 356 provide detailed advice on these issues, but there 
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Figure 23.4-1: Modeling Approaches for Buildings with Basements 
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remains relatively limited data on short-term stiffnesses, particularly under high loads, 

and wide ranges of potential properties must be considered. These documents 

recommend taking half and twice the target stiffness estimates (i.e., a factor of four on the 

range). Key issues include whether to model soils springs with initial high stiffness 

relevant before yielding, a lower secant stiffness for some larger displacement, or to use 

nonlinear models that account for the expected nonlinear force-displacement curve of the 

soil. While this is the most accurate, it can take a significant analytical effort in any 

moderate to large building. Quantitative information on soil nonlinearity at high strains 

is limited. Some geotechnical engineers continue to use linear models, even in soils like 

clay. Significantly different results can occur if strains are sufficient to reach the point of 

nonlinearity. In fact, nonlinearity in the soil can lead to the accumulation of permanent 

deformation. 

Damping, basement embedment and base slab averaging: Soil-structure interaction 

generally tends to reduce the input motion to the building as does an embedded basement 

and a slab or other foundation system that can distribute or average peak motions over the 

site. The input motion reduction is higher for buildings with a fundamental period below 

0.7-1.0 seconds and not that significant for longer period buildings. These effects are 

now being considered in seismic evaluation and rehabilitation, and they are the subject of 

FEMA 440 (FEMA, 2005).

Second opinions: In some situations, the lower bound and higher bound of geotechnical 

strength and stiffness properties that are being provided can lead to significantly different 

results. Alternative opinions or geotechnical peer review can be advantageous in 

identifying alternative sources of information, narrowing the range of assumptions, or 

increasing the strength and displacement capability.

For detailed information on evaluation, analytical and design for foundation elements see 

ATC 40 (ATC, 1996), FEMA 274 (FEMA, 1997), FEMA 356 (2000), and ASCE 31-03 

(ASCE, 2003). 

23.5 Increasing Estimates of Capacity by In-Situ Testing 

Existing shallow and deep foundations might have as-built capacities that exceed their design 

capacities. If these higher capacities can be confirmed, additional loads can be imposed on these 

foundations without any modifications to the existing foundations. Alternatively, the estimated 

capacity of new micropiles or drilled piers installed as part of a rehabilitation project can be 

verified or increased by performing in-situ load tests on them. The most common direct method 

for confirming these higher capacities is by performing in-situ load tests of the foundation 

elements. 

The plate bearing test is probably the most common direct in-situ test for estimating the capacity 

of an existing shallow foundation. It involves the determination of the load-deformation 

characteristics of the soil directly below the shallow foundation. It is worth noting that indirect 

methods involving in-situ (instead of laboratory determination) of the settlement characteristics 

of foundation soils are sometimes employed. These indirect methods, which are not covered in 

this document, allow one to more accurately estimate settlement associated with additional loads 
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to be imposed on the existing foundation. Examples of such methods include 1) the use of 

dilatometer tests to define the in-situ deformation characteristics of sandy and clayey soils and 2) 

the use of pore pressure dissipation techniques during cone penetration tests to estimate in-situ 

settlement characteristics of soft clays. 

The most common direct method for estimating the load-deformation characteristics of deep 

foundation elements is static load tests in tension or compression. 

23.5.1 Plate Bearing Tests 

Preparation 

Plate bearing tests are generally performed on existing shallow foundations to determine their 

capacities. Access to the bottom of the existing foundation, which is used as a reaction element, 

is required for the test to be performed. Access to the bottom of the foundation is facilitated via 

an access pit, which is at least 3 feet by 3 feet in plan view and extends at least 18 inches below 

the bottom of the foundation. The access pit is located in such a way that the exterior edge of the 

foundation is exposed in the pit. This access pit can be dug with a backhoe. Depending on the 

depth of the pit and the materials that are exposed in the pit, shoring may or may not be required. 

From the bottom of the pit, a rectangular mini-tunnel that extends from the exposed to the 

opposite edge of the foundation is dug underneath the foundation using handmining techniques. 

The tunnel has to be at least 18 inches wide in cross section to facilitate the placement of bearing 

plates and hydraulic jacks for the test. Sometimes, it is necessary to chip off excess concrete 

from the bottom of the foundation to create a flat surface for the placement of the upper bearing 

plate. Also, the bottom of the mini-tunnel must be prepared to create a flat surface for the lower 

bearing plate. The access pit and the mini-tunnel are depicted in Figure 23.5.1-1. It must be 

noted that sometimes it is more economical to dig the access pit from the crawl space side of the 

existing foundation in Figure 23.5.1-1. 

Equipment, Set-Up and Testing 

The minimum required equipment includes the following: 

A hydraulic ram that is capable of imposing load exceeding the design capacity of the 

existing foundation. The pressure gage of the ram must be calibrated to allow the load 

imposed by the ram to be estimated. A load cell can be used in addition to the pressure 

gage for more accurate determination of imposed loads. 

One-inch thick steel 12-inch square bearing plate. 

Minimum of four dial gages for measuring soil deformation. Linear variable 

displacement transducers or transformers (LVDT)s could be used in lieu of dial gages for 

measurement of deformations. 

The hydraulic ram and bearing plates are set up as depicted in Figure 23.5.1-1. The test is 

performed by imposing load incrementally on the soil below the lower bearing plate and 

measuring the corresponding deformations. The procedure has been standardized as ASTM 

D1194. 
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Figure 23.5.1-1: Plate Bearing Tests for In-Situ Bearing Capacity Determination 

After the test is completed, the mini-tunnel and access pit are usually partially backfilled with 

lean concrete or controlled density fill to the top of the foundation, and the balance of the pit is 

backfilled with either the same material or compacted native soil. 

Test Results 

The load-deformation data that are recorded are applicable to the 12-inch square lower bearing 

plate. The data must be corrected for scale effects to apply them to the prototype foundation. 

Reasonable results are usually obtained when plate bearing tests are performed on very stiff clays 

or sandy or gravelly soil. Poor results are usually obtained when tests are performed on soft to 

stiff clays. Refer to Bowles (1996) for discussion on extrapolating test results. 

Cost/Disruption/Challenges 

Digging the access pit and mini-tunnel can be somewhat disruptive and costly, though much less 

expensive than the cost of foundation rehabilitation. If the pit and mini-tunnel are dug from the 
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crawl space side, usually only handmining techniques can be employed, and the hauling of 

excavated spoils becomes time-consuming. 

If groundwater is encountered, the conditions in the pit are mucky. Even without groundwater, 

the conditions in the pit are damp and cramped for the individual who has to set up the plates, 

hydraulic ram and dial gages in the mini-tunnel as well as for the person who has to crouch in the 

pit to read the dial gages while a test partner applies the load and records readings from a 

position near the edge of the access pit. It is often necessary to provide a plastic covering at the 

bottom of the pit. 

Plate bearing tests on clayey soils can be time consuming because it takes a longer time for the 

deflection under each load increment to level off. 

23.5.2 Static Tests on New Deep Foundations 

Preparation 

Static load tests are usually performed on new micropiles to determine their axial capacities 

because of the potential effects of installation procedures on the capacities. The tests can be 

performed in compression or tension. For a compressive type load test, reaction micropiles must 

be installed at a distance of at least three times the diameter of the test or reaction micropile, 

whichever is greater, to minimize the potential for group effects between the test pile and the 

reaction elements. For tension tests, timber mats could be used as reaction elements in lieu of 

reaction micropiles. The test and reaction micropiles should be allowed to cure for at least seven 

days after installation before the load test is performed. 

Static load tests can also be performed on new drilled piers installed as part of a rehabilitation 

project as a means of increasing their estimated axial capacities. The preparatory work described 

above for micropiles also applies to drilled piers. Because of the size of drilled piers in 

comparison to micropiles, the spacing between the test pier and the reaction piers is much larger. 

This implies that a much larger reaction beam is required for tests on drilled piers. 

Equipment, Set-Up and Testing 

The minimum required equipment includes the following: 

A reaction beam spanning between the reaction elements and capable of sustaining the 

maximum test load without excessive deflection. 

A hydraulic ram that is capable of imposing load exceeding the design capacity of the 

existing foundation. The pressure gage of the ram must be calibrated to allow the load 

imposed by the ram to be estimated. Usually, a load cell is used in addition to the 

pressure gage for more accurate determination of imposed loads. 

An independent reference beam with supports that are located away from the test or 

reaction micropiles. 

Minimum two dial gages for measuring the deflection of the pile head. LVDTs could be 

used in lieu of dial gages for measurement of deformations. Whether LVDTs or dial 

gages are used, a secondary system of deflection measurement is required as a back-up. 
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The reaction beam, the reference beam, hydraulic ram and dial gages are set-up as depicted in 

Figure 23.5.2-1 for compression tests and in Figure 23.5.2-2 for tension tests. The test is 

performed by imposing load incrementally on the soil below the lower bearing plate and 

measuring the corresponding deformations. The maximum test load is usually about 1-1/2 to 2 

times the design load. The test can be performed in accordance with ASTM D1143 (for 

compression tests) and ASTM D3689 (for tension tests). 

Figure 23.5.2-1: Static Pile Load Test in Compression 
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Figure 23.5.2-2: Static Pile Load Test in Tension 

Test Results and Interpretation 

The load-deformation data that are recorded must be interpreted in two phases. The first phase 

involves the determination of the axial capacity in tension or compression of the test foundation 

element. The next phase involves interpreting the axial capacity relative to the known 

foundation conditions, such as the applicability of the axial capacity to a group of deep 

foundation elements or the applicability of the observed settlement from the load tests, given its 

short duration, to a production deep foundation element bearing in clayey soil that could 

consolidate. 
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Cost/Disruption/Challenges 

Setting up and performing a load test can be quite costly. Setting up can be time consuming 

because there is more manual labor involved in transporting test equipment from one test 

location to another in cramped situations. In the case of drilled piers, it might be impossible 

because of access related issues to set up an adequate reaction beam for a static load test. It may 

be possible, however, to find locations on site that are not within the building, such as parking 

lots or landscaped areas, with similar underlying soils and perform the test on elements that will 

not be used under the building. 

23.5.3 Static Load Tests on Existing Deep Foundations 

While theoretically possible, this approach is so disruptive and costly that it is generally not 

implemented in practice except where the existing foundation consists of timber piles or where 

the existing pier is located on the exterior of the building. This kind of testing would require 

temporary shoring of the column supported by the pier or pile to be tested and removal by 

cutting of the structural connection between the deep foundation element and the column. The 

top of the pier or pile must be accessible to allow for a load test set up. For a compressive type 

load test, reaction micropiles or piers must also be installed. 

23.6 New Foundations 

23.6.1 Types of New Foundations Commonly Used in Seismic Rehabilitation 

Foundation elements can be broadly classified into two basic categories: shallow and deep 

foundations. Shallow foundations include continuous strip footings, isolated spread footings, 

grade beams, and mats. Deep foundations include drilled piers and micropiles. Driven piles are 

rarely used in existing construction due to access and vibration limitations. Figure 23.6.1-1 

shows examples of these foundation types. Several excavation approaches are shown in the 

figures. In cohesive soils, the soil may be able to be cut without it sloughing into the hole. 

Metal stayforms (expanded metal lath forms) are sometimes used when there is some risk of the 

soil sloughing after the intial excavation. The stayforms are left in place when the concrete is 

poured. When the excavation gets to a certain depth, however, shoring can be required due to 

safety regulations or an open cut excavation can be used. In cohensionless soils, like sand, an 

open cut excavation will be necessary. A form can be placed, the concrete poured, the form 

removed, and then soil backfilled into the remaining open cut. Alternatively, the form can be left 

out and the concrete for the footing “overpoured” in the full open cut. The eccentricity of the 

overpour should be evaluated. 

23.6.2 Add Shallow Foundation Next to Existing Shallow Foundation 

Description of the Rehabilitation Technique 

When a concrete overlay is placed against an existing wall, a new footing is typically needed. A 

common situation is the existing footing is a continuous strip footing and the new footing is 

either a strip footing or a grade beam. Figure 23.6.2-1 shows an example of a new concrete wall 

and footing against and existing unreinforced masonry wall and concrete strip footing. 
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Figure 23.6.1-1: Types of New Foundations Commonly Used in Seismic Rehabilitation 

Design Considerations 

Effective footing width: Several approaches to footing design are used. One is to assume only 

the new footing resists the loads under the new overlay. Another is to share loads between the 

new and existing footing simply on the basis of area. The most sophisticated approach is to 

recognize the potentially different stiffness between the soil under the existing footing which has 

been consolidated already and the soil under the new footing which is likely to be more flexible 

since loading is likely to be lighter and only newly applied. Sometimes jacking is employed to 

transfer loads to new foundations. 
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Figure 23.6.2-1: New Concrete Strip Footing Next to Existing Strip Footing 

Shear transfer: It is standard practice to connect the new and existing footings with drilled 

dowels, though it useful to consider whether the dowels are actually necessary elements. Dowels 

in the footing and wall above should be designed to be sufficient to transfer the force intended to 

be resisted under the existing footing. 

Unreinforced existing footings: The existing footing may be unreinforced masonry or poorly 

reinforced concrete. If the footing is wide enough so that so beam action will result under 

bearing pressure, the bottom drilled dowels can be extended deep into the existing footing near 

the base of the footing to serve as positive reinforcing. 

Detailing and Construction Considerations 

New footing is deeper than existing footing: A key goal when adding a new footing is not to 

surcharge or undermine an existing footing. The best approach, then, is to match the new and 

old footing depths. This is, of course, not always possible. Figure 23.6.2-2 shows the situation 

when the new footing needs to be deeper than an existing footing. If excavation proceeds 

without underpinning, particularly in soils with minimal cohesion, soil can slough away from 

under the existing footing into the new excavation leading to damaging footing movement. 

Underpinning is used to address this situation. Underpinning means digging a series of short 

length pits separated by a sufficient distance, digging under the existing footing adjacent to the 

pit, adding concrete to the base of the final excavation depth, and then going back and 

completing the underpinning in between the initial pits. An alternative underpinning approach 

is to place long underpinning piers intermittently beneath the new footing to derive support at 

depth so that the typical new footing need not be deeper than the existing footing. 
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Figure 23.6.2-2: New Shallow Footing is Deeper than Existing Shallow Footing 

New footing does not need to be a deep as existing footing: Figure 23.6.2-3 shows the situation 

when the new footing does not need to be as deep. If the excavation is kept shallow, the new 

footing when loaded can impart additional and eccentric loads into the existing footing that may 

not be desirable. As a result, it is common to extend the bottom of the new footing down to 

match the depth of the existing footing. The extension is often lightly reinforced. 

Existing footing is in the way: As Figures 23.6.2-1 to 23.6.2-3 show, the existing footing will 

often extend inboard from the existing wall underneath the new wall. To place the new footing, 

the existing footing often must be chipped away to develop a properly reinforced footing. This 

can be done with jackhammering or sawcutting. The capacity of the existing footing during the 

temporary condition where it is smaller and eccentrically loaded should be verified as adequate. 
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Figure 23.6.2-3: New Shallow Footing Does Not Need to be as Deep as 

Existing Shallow Footing 
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Cost/Disruption 

Adding a new footing is quite disruptive and costly. The existing slab-on-grade must be sawcut 

and removed, then the trench excavated, drilled dowels installed, rebar laid, debris in the footing 

removed, and concrete placed. This is all time-consuming, messy, and noisy. 

23.6.3 Add Shallow Foundation Next to Existing Deep Foundation 

Adding a new shallow foundation next to an existing deep foundation is relatively rare for two 

reasons. First, the existing foundation was deep because soil or structural loading conditions 

would not permit a shallow foundation. Without ground remediation, a new foundation would 

have the same issue. Second, as noted in Section 23.6.2, if the new foundation is higher than the 

existing foundation, the new foundation will impart gravity and earthquake loads to the existing 

foundation which is usually undesirable. With careful study of relative rigidity considerations, 

there can be situations where a new shallow foundation can be added adjacent to an existing deep 

foundation, such as a mat next to drilled piers. See Section 23.8.2. 

There can be cases, though, when adding new deep foundations are very disruptive or not 

economical practicable due to existing access limitations. Sometimes a shallower foundation is 

added, such as a new mat next to an existing drilled pier foundation. The relative stiffness of 

each foundation then becomes the key consideration. 

23.6.4 Add Deep Foundation Next to Existing Shallow Foundation 

Adding a new deep foundation next to an existing foundation is occasionally done, such as 

drilled piers under a new wall next to an existing strip footing. Figure 23.6.4-1 shows an 

example of this technique. Drilling limitations can be significant, and they include access 

requirements for the drill rig, height restrictions for the drill rig, the offset needed to get the edge 

of the drill up against the existing wall, vibration during drilling, and utilities in the way of the 

drilling. Sometimes when the exterior face of the building is accessible, slanted drilling is done 

under the existing footing. Usually, the drilled piers are spaced at a sufficient distance that the 

existing footing and walls can span around or over the open hole. After the pier and new wall 

are installed and dowelled into the existing wall and footing, a composite system has been 

created. While many engineers simply take gravity in the existing spread footing, and 

overturning in the piers, live loads and earthquake loads are of course actually distributed 

throughout the system by relative rigidity. 

23.6.5 Add Deep Foundation Next to Existing Deep Foundation 

There will also be situations where new deep foundations are added next to existing deep 

foundations. New deep foundations include drilled piers and micropiles. See Section 23.8 for 

examples. 

23.7 Structural Rehabilitation for Existing Shallow Foundations 

23.7.1 Goals 

Typical structural improvements to existing shallow foundations can be simplified into two basic 

categories: enhancing compression capacity and enhancing tension capacity. 
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Figure 23.6.4-1: New Drilled Pier Next to Existing Strip Foundation 
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General techniques for improving inadequate compression capacity: Compression strength 

capacity of existing spread and strip footings can be addressed by widening the footing base; 

replacing the footing with an enlarged foundation; adding micropiles, screw anchors or drilled 

piers adjacent to the existing footing; adding micropiles through the existing footing; or adding 

grade beams to connect isolated spread footings together. 

General techniques for improving inadequate tension capacity: Improving inadequate tension 

capacity of existing spread and strip footings uses similar techniques to those for improving 

compression capacity, including widening the footing base to increase the dead load; replacing 

the footing with an enlarged foundation; adding micropiles, screw anchors, or tie-downs adjacent 

to or through the existing footing; or adding grade beams to adjacent footings and columns to 

pick of dead load to resist uplift. 

The following sections provide some examples of rehabilitation techniques for existing shallow 

footings. 

23.7.2 Add Micropiles Adjacent to an Existing Strip Footing 

Deficiencies Addressed by Rehabilitation Technique 

Inadequate compression capacity at the toe of strip footing beneath a wall 

Inadequate tension capacity at the heel of a strip footing beneath a wall 

Description of the Rehabilitation Technique 

To improve the compression and/or tension capacity of the existing footing, the footing is 

widened and micropiles, also known as pin piles, are added. Figure 23.7.2-1 provides an 

example. 

Design Considerations 

Research basis: FHWA (2000) provides guidelines for the design and construction of micropiles.

Compression strength and stiffness: When micropiles are added together with the strip footing, 

resistance is shared between the two different elements, depending on their relative rigidity. 

Micropile strength and stiffness are given in the geotechnical report. Governing strength 

depends on both the soil capacity and the structural capacity of the pile, including the pipe, grout, 

and reinforcing bar. Compression stiffness considers the pile elements and surrounding soil 

movement. 

Tension strength and stiffness: Uplift resistance is taken by the micropiles. Structural tension 

strength is lower than compression strength in the micropiles and is usually based on just the 

reinforcing bar, unless special details are used to engage the top of the casing in tension. 

Tension stiffness is also usually lower; tension flexibility comes from the reinforcing bar 

elongation and surrounding soil movement. 

Corrosion effects: Permanent casing associated with micropiles is typically uncoated. 

Depending on the corrosivity of the soil, corrosion of the permanent casing can occur over time. 

Techniques are available for estimating the extent of thickness of the steel pipe lost to corrosion; 
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with the estimates, a reduced thickness and reduced lateral and buckling capacities of the pile can 

be calculated. 

Figure 23.7.2.-1: Micropile Enhancement to Existing Strip Footing 

Testing: Performance and proof load testing are performed at the start of and periodically during 

construction to verify that specified design capacities will be achieved. During performance 

testing, the test piles are usually loaded to 2.0 to 2.5 times the design load. Proof testing, on the 

other hand, involves testing the pile to 1.33 to 1.67 times the design load. Proof testing is 

usually limited to a percentage of the production piles. Creep tests are typically performed as 

part of the performance and proof tests, especially if the micropiles are to be bonded in clayey 

soils that are susceptible to creep. PTI (1996) provides guidelines on performing and evaluating 

performance, proof, and creep tests on foundation elements. 
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End bearing vs. friction: Because of its small size, micropiles generally derive most of their 

capacities from friction. The geotechnical compression capacity of the micropile is therefore 

generally equal to the axial tensile capacity. 

Filling the annulus with grout: Where the cutting tool of micropile drilling equipment creates a 

hole slightly larger than the permanent casing, an annulus is created around the casing. 

Typically, this annulus is not grouted. 

Detailing and Construction Considerations 

Detailing and construction considerations for adding micropiles to an existing footing include the 

following. 

Connecting to the new footing: Figure 23.7.2-1 shows bars drilled all the way through the 

existing footing. If this not done, the existing capacity of the footing for bending and the center 

where the moment is largest must be checked; it is unlikely to be acceptable. In the figure, the 

through dowels are installed from the right and coupled on the left. Headed bars are shown for 

ease in installation. Hooked bars could be used, but they would trigger a position coupler (one 

that eliminates the need to rotate the bar), at least for the bottom row of bars. To install the 

longer dowels, over excavation of the adjacent soil is needed. This needs to be understood 

during detailing, as their may be existing elements on top of that portion of the slab. 

Access and height limitations: Adequate clearance must be available for the equipment used to 

install micropiles inside existing buildings. 

Anchorage to the footing: Figure 23.7.2-2 shows a micropile and some of its details. In this 

figure, tension is taken by threaded rod and the plate at the top of the rod. Sufficient embedment 

of the plate above the base of the footing is needed to develop the strength of the rod. Similarly, 

the bottom plate is designed to take the compression and deliver it to the pipe. In some cases, the 

bottom plate may not be necessary as the grout diameter or top plate can be sufficient. If the top 

plate is used, it must be sufficiently deep below the top of the footing so it is not the weak link. 

Bar types and size: Bar types include ASTM A722 high strength threadbar, with Fy = 150 ksi, 

with common sizes of 1”, 1-1/4”, 1-3/8” and 1-3/4” diameter. CALTRANS has typical details 

using #18 bars in ASTM A615 steel, where ends needing nuts or couplers are threaded. 

Pipe types and sizes: API casing with Fy = 80 ksi is commonly used, with 7” diameter and 9-5/8” 

diameter pipes being common. 

Depth of pipe and grouting: The pipe typically goes down into the bearing layer the requisite 

depth. The reinforcing bar usually continues deeper. Grouting fills up the hole at the base, the 

annulus around the pipe and the inside of the pipe. Post-grouting or secondary grouting can be 

used at the base to increase the bar capacity. 

Strain limits at the top: To increase the length over which the bar is strained in tension, the top 

of bar below the anchorage plates are sometimes debonded with a greased PVC pipe. 
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Figure 23.7.2.-2: Micropile Details 
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Cost/Disruption 

Micropiles are typically less expensive than drilled piers, unless very large capacities are 

required. They require less headroom and smaller footings to receive the bars and pipe. 

Excavation noise and dust, and drilling and grouting noise must be considered as part of the 

rehabilitation strategy. 

Proprietary Issues 

Micropile specifications are often written like tiebacks, so that the contactor must design and 

build the micropile to meet performance requirements. Figure 23.7.2-2 shows a generic type of 

pile. There are other proprietary piles that use a pointed pipe casing as the drill. 

23.7.3 Enlarge or Replace an Existing Spread Footing 

Deficiency Addressed by Rehabilitation Technique 

Inadequate compression capacity at a spread footing 

Inadequate tension capacity at a spread footing 

Description of the Rehabilitation Technique 

An existing spread footing may be under a braced frame, moment frame or a concrete column 

below a discontinuous shear wall and be subjected to compression or tension forces that exceed 

the footing capacity. The existing footing can be enlarged or replaced to increase compression 

capacity or the dead load for resisting tension. 

Design Considerations 

Research basis: No research specific to enlarging or replacing existing footings has been 

identified. 

Bending moment and shear checks in enlarged footing: Gaining large increases in compression 

capacity by enlarging an existing footing is often difficult given the limits of the existing footing. 

In Figure 23.7.3-1, reinforcing is drilled in from the sides, but does not go through to the other 

side. The shear capacity of the footing is not increased. The bending capacity has to be checked 

at critical locations “A” and “B”. Location A will typically govern. If sufficient capacity 

cannot be achieved, the footing can be replaced as shown in Figure 23.7.3-2. 

Tension capacity: Tension capacity can be quite limited if the existing spread footing only has 

bottom reinforcing bars which would be typical. Drilled dowels can be added to the top of the 

footing and top steel added in the slab-on-grade level. See Section 23.9 for a similar example in 

a pile cap. 

Detailing and Construction Considerations 

Detailing and construction considerations for enlarging or replacing an existing spread footing 

include the following. 

Existing reinforcing: Existing reinforcing should be preserved in the footing. This will typically 

require placing new drilled dowels at a higher elevation, with a resulting lower moment capacity. 
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Figure 23.7.3-1: Enlarge Existing Spread Footing 
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Figure 23.7.3-2: Replace Existing Spread Footing 

Installing drilled dowels: Figure 23.7.3-1 shows two approaches for installing bars. On the left a 

coupler is used, permitting a smaller overexcavation past the footing, but triggering a hole large 

enough to accommodate the coupler. This is likely to force pressure grouting with nonshrink 

grout as the annulus will be too large for most adhesives like epoxy. On the right, a larger overexcavation is used and a single piece bar is installed. 

Lapping with the existing slab-on-grade: The existing slab is likely to have wire mesh. To 

minimize vertical offsets the new and existing slabs should be dowelled together. Either the 

mesh in the existing slab-on-grade can be preserved when the slab is demolished or drilled 

dowels can be drilled into the edge of the existing slab. 
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Shear transfer between the new and existing footings: Transferring shear between the existing 

footings is necessary. This can be accomplished by roughening the existing footing face. Some 

engineers bevel the existing face as well with the top wider than the bottom, so outward pressure 

is exerted under compression loading. Some engineers dig the new footing slightly deeper than 

the existing footing and undercut the soil at the edge of the existing footing, so that the new 

footing acts as a corbel to resist downward pressure from the existing footing. 

Shoring: If the existing footing is replaced, shoring will be needed. It is critical that the base of 

footing be properly compacted and the new concrete be tightly placed beneath the existing 

column to minimize or eliminate any settlement when the shores are removed. 

Cost/Disruption 

Enlarging or replacing an existing footing is a localized but disruptive process, involving 

excavation, dust, mud, drilling/jackhammering noise and concrete placement. Protection of 

existing finishes in the vicinity and in the working path is necessary. 

23.8 Structural Rehabilitation for Existing Deep Foundations 

23.8.1 Goals 

Typical structural improvements to existing deep foundations can be simplified into several basic 

categories: enhancing the overall compression capacity, tension capacity, or lateral capacity of 

the foundation; and improving the ductility and detailing of specific elements or connections 

within the system. 

General techniques for improving overall inadequate compression, tension and lateral capacity: 

Inadequate strength and deformation capacity of existing pile and pier foundations can be 

addressed by adding new shallow adjacent shallow foundations, and new piers or micropile 

foundations, either in vertical or battered orientations. 

General techniques for improving inadequate improving ductility and detailing: Inadequate 

confinement can be improved with enlarged or replacement pier and pile caps; lack of top steel 

in pier and pile caps can be addressed with new concrete overlays on top of the cap. 

The following sections provide some examples of rehabilitation techniques for existing deep 

footings. 

23.8.2 Add a Mat Foundation, Extended Pile Cap or Grade Beam 

Deficiency Addressed by the Rehabilitation Technique 

The deficiency addressed by this technique is inadequate compression capacity of an existing 

deep foundation element. The technique involves taking advantage of the contributions of 

shallow foundation elements that are part of the overall foundation system. 

Description of the Rehabilitation Technique 

When existing piers or piles have inadequate capacity, the usual approach to increasing their 

capacity is to install new micropiles or piers connected by grade beam to the existing adjacent 
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piles or piers. Where competent bearing soil is within five feet of building ground floor, an 

alternative approach to installing new piles or piers is to widen and deepen the cap or grade 

beams atop the pier or pile or connecting adjacent piles or piers. Figure 23.8.2-1 shows an 

example of existing piers whose capacities are augmented by installing a mat between the piers. 

Figure 23.8.2-1: New Mat Foundation Between Existing Drilled Piers 
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Design Considerations 

Analysis: Several approaches to design are used. One is to assume that only the new mat 

foundation resists the new loads imposed by the retrofit scheme. This assumption is inaccurate. 

The most sophisticated approach is to model the soil under the new mat as a spring with stiffness 

that is different from that of the spring representing the existing piers. The analysis would show 

that new loads are supported by both the new mat foundation and the existing piers based on the 

relative stiffnesses of the two sets of foundation elements. 

Shear transfer: It is standard practice to connect the new mat or cap or grade beam and existing 

piers with drilled dowels. Dowels in the piers and grade beams above should be designed to be 

sufficient to transfer the force intended to be resisted under the new arrangement. 

Detailing and Construction Considerations 

The best approach is to match the new mat and old pier cap and grade beam depths. This is, of 

course, not always beneficial especially if the soils that the caps or grade beams are bearing on 

have low bearing characteristics. Sometimes the new mat needs to be deeper than the existing 

grade beams or pier. 

Cost/Disruption 

Adding a new mat or cap or grade beam is quite disruptive and costly. The existing slab-ongrade must be sawcut and removed, then the foundation excavation completed, drilled dowels 

installed, rebar laid, debris in the excavation removed, and concrete placed. This is all timeconsuming, messy and noisy. 

23.8.3 Add Drilled Piers to an Existing Drilled Pier Foundation 

Deficiency Addressed by Rehabilitation Technique 

Inadequate compression capacity of a drilled pier foundation 

Inadequate tension capacity of a drilled pier foundation 

Inadequate lateral capacity or ductility of a drilled pier foundation 

Description of the Rehabilitation Technique 

An existing drilled pier footing beneath a shear wall may lack sufficient compression capacity at 

the toe, tension capacity at the heel, or the existing pier reinforcing may be inadequate for lateral 

demands. Adding new, well detailed drilled piers provides supplemental capacity to reduce the 

demands on existing elements or increase the overall capacity and ductility. Figure 23.8.3-1 

shows an example where the “web” or center of the footing is widened or replaced and new 

drilled piers are added. 

Design Considerations 

Research basis: No research specific to supplementing existing drilled pier footings has been 

identified. 

Relative rigidity: In Figure 23.8.3-1, all of the piers—new and existing—will participate in 

resisting axial and lateral demands and should be considered in modeling efforts. Demands in 

existing piers should be confirmed as adequate. 
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Figure 23.8.3.-1: Adding Drilled Piers to an Existing Drilled Pier Foundation 
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Detailing and Construction Considerations 

Detailing and construction considerations for adding new drilled piers to an existing drilled pier 

footing include the following. 

Access and height limitations: All of the drilling limitations noted in Section 23.8 for drilled 

piers apply here as well. 

Spacing: Drilled piers typically have spacing limits of three times the pier diameter to avoid 

group effect reductions. This can limit the number of piers that can be installed. 

Pier cap/thickened footing: The concrete above the piers will likely require widening as shown 

in Figure 23.8.3-1. Either drilled dowels can be installed in the existing footing, or the footing 

can be demolished and replaced. With a shear wall above, the wall may be able to bridge across 

to the belled ends of the footing without any shoring. 

Cost/Disruption 

Adding new drilled piers in an existing building is very disruptive, involving excavation, dust, 

mud, drilling/jackhammering noise and concrete placement. Protection of existing finishes in the 

vicinity and in the working path is necessary. 

23.8.4 Add Micropiles to an Existing Drilled Pier Foundation 

Deficiencies Addressed by Rehabilitation Technique 

Inadequate compression capacity of a drilled pier foundation 

Inadequate tension capacity of a drilled pier foundation 

Inadequate lateral capacity or ductility of a drilled pier foundation 

Description of the Rehabilitation Technique 

An existing isolated drilled pier footing supporting a braced frame column, moment frame 

column or concrete column under a discontinuous shear wall may lack sufficient compression 

capacity, tension capacity, or the existing pier reinforcing may be inadequate for lateral demands. 

Adding new micropiles provides supplemental capacity to reduce the demands on the existing 

drilled pier. Figure 23.8.4-1 shows an example. 

Design Considerations 

Research basis: No research specific to supplementing existing drilled pier footings with 

adjacent micropiles has been identified. 

Relative rigidity: In Figure 23.8.4-1, both the new micropiles and existing drilled pier will 

participate in resisting axial and lateral demands and should be considered in modeling efforts. 

Demands in the existing piers should be confirmed as adequate. While the axial strength of the 

new micropiles may be comparable to the drilled pier, they will much lower lateral stiffness. 

Detailing and Construction Considerations 

Detailing and construction considerations for adding new micropiles to an existing drilled pier 

footing include the following. 
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Figure 23.8.4.-1: Micropile Enhancement of an Existing Drilled Pier Footing 
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Access and height limitations: All of the drilling limitations noted in Section 23.7 for micropiles 

apply here as well. 

Spacing: Spacing limits between the drilled pier and micropile to avoid group effects should be 

addressed. 

Collar around drilled pier: In Figure 23.8.4-1, a concrete collar wraps the top of the drilled pier 

and provides the termination point for the pile anchors. It can also provide the starter bar 

location for the concrete jacket used to wrap the concrete column above. 

Cost/Disruption 

Adding new micropiles in an existing building is less disruptive than new drilled piers, but it still 

involves excavation, dust, mud, drilling/jackhammering noise and concrete placement. 

Protection of existing finishes in the vicinity and in the working path is necessary. 

23.8.5 Add Top Bars to an Existing Pile Cap 

Deficiency Addressed by Rehabilitation Technique 

Inadequate bending capacity of the top of the pile cap to resist uplift forces 

Description of the Rehabilitation Technique 

An existing pile cap or pier cap may lack top reinforcing bars because the original design showed 

no net uplift. When a capacity design approach to evaluation or a pushover is conducted, it is 

likely that uplift will occur at some point and trigger the need for top bars in the pile cap to resist 

bending. Sometimes there is sufficient capacity in the reinforcing of the slab and its nominal 

connection to the top of the pile cap that it can serve the function of top steel. If not, top bars can 

be added as shown in Figure 23.8.5-1. 

Design Considerations 

Research basis: No research specific to adding top bars to existing pier or pile caps has been 

identified. 

Anchorage of the pile to the pile cap: In Figure 23.8.5-1, it is assumed that the anchorage of the 

pile to the pile cap is adequate for uplift. If the foundation was not originally designed for uplift, 

only nominal anchorage between the pile and pile cap is likely to be found. This could be a 

single large bar or a bundle of two bars placed in a grouted hole in the top of the pile. It could be 

the pile reinforcing extended up into the pile cap. This would be more likely in end bearing piles 

where refusal is hit early and the top of the pile must be chipped down to the right elevation. It is 

important to realize that even a pile designed for a pin top with a central bar will resist moment 

unless special design considerations such as neoprene pads are added on top of the pile. This is 

highly unlikely in an older building. The tension in the pile anchorage under lateral loading has 

to be added to the uplift from the superstructure. 

Anchorage of the column to the pile cap: In Figure 23.8.5-1, it is assumed that the anchorage of 

the column to the pile cap is adequate for uplift. If the foundation was not originally designed 

for uplift, anchor bolt embedments and diameters may not be sufficient. 
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Figure 23.8.5-1: Adding Top Bars to an Existing Pile Cap 

Pile cap modeling: The pile cap is typically fairly deep and may behave more as a deep beam. It 

can also be analyzed using strut-and-tie models. 

Detailing and Construction Considerations 

Detailing and construction considerations for adding top bars to an existing pile cap include the 

following. 

Existing reinforcing: Existing reinforcing should be preserved in the footing. This will typically 

require placing new drilled dowels inboard of the edges of the existing footing. Added edge 

distance is also desirable if there or no or minimal side bars. 
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U bars or lapped L bars: Figure 23.8.5-1 shows new U bars, so that each leg of the U must be 

lowered simultaneously into the pile cap. L bars lapping over the top of the cap may make 

installation easier. 

Confinement: The existing slab is likely to have wire mesh. To minimize vertical offsets the 

new and existing slabs should be dowelled together. Either the mesh in the existing slab-ongrade can be preserved when the slab is demolished or drilled dowels can be drilled into the edge 

of the existing slab. 

Cost/Disruption 

Adding top bars to the pile cap is a localized and less disruptive process than many foundation 

retrofits. It does involving dust, drilling/jackhammering noise and concrete placement. 

Protection of existing finishes in the vicinity is necessary. 

23.9 Ground Improvement for Existing Shallow and Deep 

Foundations 

23.9.1 Goals 

Typical goals for ground improvement under existing shallow and deep foundations can be 

classified into two categories: mitigating the potential impacts of an identified geologic hazard, 

and enhancing the capacity of the foundation by changing the load-deformation characteristics of 

the foundation soil. The general techniques used to achieve these two goals separately or in 

combination include compaction grouting and permeation grouting. Warner (2004) is a good 

resource for both types of grouting. 

Typical geological hazards that are mitigated using ground improvement include liquefaction 

and compaction settlement. These hazards have to be established by a Geotechnical Engineer 

who will define the recommended zone of geologic hazard mitigation. 

23.9.2 Compaction Grouting 

Description of the Rehabilitation Technique 

Compaction grouting involves the injection of a very stiff grout at a high pressure into a layer of 

soil to force the individual soil particles into a tighter packing. The resulting increase in the 

density of the soil substantially increases its resistance to liquefaction as well as its bearing 

capacity. Compaction grouting can be performed in a wider range of soil types than other 

grouting methods. It can be performed in various types of sands, and clayey materials, but has 

limited effectiveness in clean coarse sands and gravels and in high plasticity soils. 

The grout is required to have low flowability. This low flowability is necessary because the 

most important characteristic for effective densification is for the grout to form a controlled 

mass, which is columnar or tear-shaped, when injected. If it behaves instead like a fluid in the 

ground, it can create fractures in the soil, through which the grout can flow. Since the 

effectiveness of the grout is based on its ability to stay as a mass pushing soil particles together, 

that effectiveness is lost when the grout flows. 
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The grout—which consists of mostly of sand, cement, and water—is injected through grout holes 

that are drilled in a grid pattern of between 4 and 12 feet. Casing that typically has an internal 

diameter of 2 to 4 inches is usually installed in the grout holes. The injection pressure is directly 

proportional to the pumping rate, the optimal pumping rate being between 1 – 2 cubic feet per 

minute. Grout is usually injected in a strict primary-secondary pattern. Alternate primary holes 

are drilled and grouted first, followed by the secondary holes. 

Grout is usually injected in stages. Staging involves the injection of only a few feet of grout hole 

at a time. Staging can proceed from top-down or bottom-up, the latter approach being the most 

commonly used. 

The bottom-up grouting approach involves the following: 

1. A hole is drilled to the bottom of the zone to be grouted. 

2. Casing is installed to within a few feet of the bottom of the hole. 

3. Grout is injected until refusal is reached. Refusal is assumed to have been reached if 

A slight movement of the ground surface or overlying improvement occurs. 

A predetermined amount of grout is injected. 

A given maximum pressure is reached at a given pumping rate. 

4. The casing is raised one to two feet. 

5. Grout injection is resumed until refusal is reached. 

6. Steps 4 and 5 are repeated until the top of the grout zone is reached. 

The level of densification achieved is verified by performing a cone penetration test or standard 

penetration test. 

Specific Issues Relating to Grouting Under Shallow Foundations 

Compaction grouting can be performed under shallow foundations in the manner described 

above except that the grout holes tend to be vertical rather than inclined. This is because inclined 

grout holes result in large horizontal areas that increase the likelihood of surface heave. The 

grout zone usually extends from the bottom of the existing shallow foundation to a dense or very 

stiff layer below the foundation. 

Unlike compaction grouting performed in an open undeveloped area, the level of densification 

achieved in the soil below a shallow foundation cannot be verified using cone penetration or 

standard penetration tests. The level of densification is verified instead through monitoring the 

volume of grout injected in the holes. 
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To establish the relationship between the volume of grout and the level of densification, a pilot 

test program is performed in an open area adjacent to the existing building which will have 

compaction-grouted footings. The pilot test site is divided into segments where injection points 

at different spacings are laid out in a grid format as shown in Figure 23.9.2-1. In each hole, 

Figure 23.9.2-1: Compaction Grouting Under Existing Shallow Foundations – 

Pilot Test Program 
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grout is injected within the upper and lower limits of the zone to be grouted. The volume of 

grout injected in each hole is recorded. After grout injection is completed, the level of 

densification achieved in each segment of the pilot test area is verified by performing cone 

penetration or standard penetration tests at the test locations as depicted in Figure 23.9.2-1. The 

spacing and the corresponding volume of grout injected in each hole that produced the 

acceptable level of compaction is selected for production grouting underneath the shallow 

foundations. The injection point grid pattern will be similar to the pattern in a segment in Figure 

23.9.2-1 with the test location coinciding with the center of a square footing or the centerline of a 

continuous footing. Note that a separation gap equal to three times the minimum spacing of four 

feet is placed between the segments to minimize the impact of one segment on the other. 

Specific Issues Relating to Grouting Under Deep Foundations 

The goal of compaction grouting around deep foundations is to enhance the skin friction 

contribution from the soils surrounding the deep foundation element. As in the case of shallow 

foundations, compaction grouting around deep foundation elements is performed using vertical 

rather than inclined grout holes. The grouting zone extends from the top to the tip of the deep 

foundation element. The injection points are usually set up at least six feet away from the center 

of the deep foundation element. A pilot test program, similar to the one described above for 

shallow foundations is performed in an open area adjacent to the existing building, which will 

have compaction grouted deep foundation elements. See Figure 23.9.2-2. Verification tests are 

performed at the location marked “T/DF”. The spacing and corresponding volume of grout 

injected in each hole that produced the acceptable level of compaction is selected for production 

grouting around the deep foundation elements. The injection point grid pattern for production 

grouting is set up similar to the pilot test program in Figure 23.9.2-2, except that the deep 

foundation element location will correspond to a location marked “T/DF.” 

Cost/Disruption 

Compaction grouting is quite disruptive and costly especially if the creation of injection holes 

includes drilling through existing pile or pier caps, grade beams, or concrete footings and slab. 

This grouting process could also be time-consuming and messy. Disruption to the current 

operations of the building is usually minimized by performing the compaction grouting at night 

and cleaning up the work area before the start of work the next morning. Compaction grouting is 

generally less costly than permeation grouting for a given scope of work. 

23.9.3 Permeation Grouting Under Existing Shallow and Deep Foundations 

Description of the Rehabilitation Technique 

Permeation grouting involves the injection of chemical or cement grout into the pore spaces of 

soils and aggregates without displacing the materials. This helps solidify the usually sandy soils 

that are amenable to this technique. The resulting increase in shear strength of the soil 

substantially increases its resistance to liquefaction as well as its bearing capacity. Permeation 

grouting can be performed in sands and sandy soils that contain minor amounts of fine particles. 

The structure and the size of voids in the soil structure dictate the type of grout that can be 

effectively used. In general, either micro-fine cement grout or a chemical grout. The use of 

chemical grouts has been diminishing for environmental reasons. 
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Figure 23.9.2-2: Compaction Grouting Under Existing Deep Foundations – 

Pilot Test Program 
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The grout is injected through grout holes that are drilled in a grid pattern of between 2 and 6 feet. 

Casing that typically has an internal diameter of 2 to 4 inches is usually installed in the grout 

holes. Grout is usually injected in a strict primary-secondary pattern. Alternate primary holes 

are drilled and grouted first followed by the secondary holes. The level of solidification 

achieved is verified by exhuming grouted soil bulbs, taking samples of the grouted soil and 

performing unconfined compression tests on the samples. 

Shallow foundations: The goal for the shallow foundation elements is to create a solidified mass 

of sandy soil below the footprint of the footing as a minimum. The solidified mass should 

extend from the bottom of the footing to the top of the dense sand layer as shown in Figures 

23.9.3-1 and 23.9.3-2. 

Figure 23.9.3-1: Permeation Grouting of Loose Sand Under Existing Shallow Foundation 

Deep foundations: The goal for the deep foundation element is to create a zone of solidified 

sand around it. The injection points can be as close as three feet to the foundation elements. The 

zone of grouting should extend from the bottom of the grade beam or cap atop the deep 

foundation element to the top of the dense sand layer shown in Figure 23.9.3-3. 
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Figure 23.9.3-2: Permeation Grouting of Liquefiable Layer 

Under Existing Shallow Foundation 
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Figure 23.9.3-3: Permeation Grouting of Liquefiable Layer 

Around Existing Deep Foundation 

Cost/Disruption 

Permeation grouting can be quite disruptive and costly especially if injection holes have to be 

drilled through existing concrete footings and slabs. If current operations in the building are to 

continue, the usual approach is to do the permeation grouting at night and clean up the work area 

before the start of work the next morning. 
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In the case of grouting under shallow foundations, there is a tendency for grout to migrate down, 

resulting in a weakly cemented lens of sand immediately below the shallow foundation elements. 

The tendency can be minimized by ensuring that grouting is performed in a strictly primarysecondary sequence. 

23.10 Mitigating the Impacts of Other Ground Hazards on Existing 

Foundations 

23.10.1 Issues to be Addressed 

The mitigation measures described in the preceding sections of this chapter deal primarily with 

individual foundation elements in a building. Even if these mitigation measures are 

implemented, their usefulness can be negated by other ground hazards that tend to have global 

stability effects on the behavior of the entire foundation system and could lead to the collapse of 

the building to be rehabilitated. These other hazards must therefore be mitigated if they exist, for 

the intent of the mitigation methods described in the previous sections to be realized. This class 

of ground hazards includes fault rupture, lateral spreading, and seismic-induced landslide. The 

potential for these hazards, the level of severity, and the necessary mitigation measures to be 

implemented must be established by a geotechnical engineer and/or an engineering geologist. 

See FEMA 274 for additional information. Mitigating the potential impacts of these hazards can 

be very costly. Disruption to the existing building, however, should be minimal since the work 

is external to the building. 

23.10.2 Fault Rupture 

The potential for fault rupture exists when an engineering geologist establishes through fault 

trenching that an active fault trace traverses the footprint of the existing building that is to be 

rehabilitated as depicted in Figure 23.10.2-1. The rupture results in displacement along the fault 

trace, which depending on the type of fault, could be lateral or vertical movement. The 

magnitude of earthquake-induced displacement in the ground along the fault trace can range 

from a few inches to several feet. Such displacements can have the effect of tearing a building 

apart when they occur under or adjacent to a building. 

It is difficult to upgrade a building straddling an active fault to accommodate such displacements 

without collapse. Options for mitigating the hazard include: 

Change the occupancy level from the current to a much lower level in an effort to 

minimize the potential for loss of life. 

Move the affected structure to a location at least 50 feet from the mapped fault trace, if 

feasible. 
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Figure 23.10.2-1: Active Fault Traversing the Footprint of Existing Building 

23.10.3 Lateral Spreading 

Definition of Hazard and Potential Impacts 

Lateral spreading is one of the phenomena associated with liquefaction. It occurs when the 

blocks of non-liquefiable surface material above a layer of liquefiable soil move laterally 

towards an open face as depicted in Figure 23.10.3-1. The magnitude of lateral movement can 

range from a few inches to several feet. 

Lateral spreading has the effect of globally moving the building laterally, or tearing it apart, if 

the building is supported on shallow foundations or on deep foundations that do not extend into 

stable material below the liquefiable layer. If the building is supported on deep foundations that 

extend into the stable material below the liquefiable layer, lateral spreading could result in loss of 

lateral capacity of the deep foundation elements. 

Mitigation 

The potential for lateral spreading can be mitigated by creating a stable mass of material near the 

open face. This can be accomplished by either densifying the layer of potentially liquefiable soil 

or solidifying the soil to prevent liquefaction. Techniques used to achieve this goal include 

compaction grouting (densifying) and permeation grouting (solidifying), which were described 

in Section 23.9. Alternately, vibrocompaction methods involving the installation of stone 

columns can be used to densify the potentially liquefiable layer as shown in Figure 23.10.3-2. 
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Figure 23.10.3-1: Lateral Spreading of Soil Layer Underlying Existing Building 

23.10.4 Seismic-Induced Landslide 

Definition of Hazard and Potential Impacts 

Seismic-induced landslides result from the failure of an existing slope under earthquake loading. 

The landslide can occur under various sscenarios as shown in Figure 23.10.4-1. In Figure 

23.10.10-1A, the landslide could undermine the building, causing it to collapse. In Figure 

23.10.4-1B, the landslide displacement along the failure surface would be much greater than the 

displacement under the building at the ground surface. Extensional fissures, however, can occur 

under the building due to the displacement at depth. The geotechnical engineer and engineering 

geologist can estimate the magnitude of these fissures. In Figure 23.10.4-1C, the landslide could 

result in debris flow impact on the building. 

Mitigation 

Various mitigation schemes developed by the geotechnical engineer and engineering geologist 

can be implemented in each of the three scenarios. In the first scenario, potential mitigation 

measures include the construction of a stabilizing berm at the toe of the slope as shown in Figure 

23.10.4-2A or the construction of a soil nail wall as shown in Figure 23.10.4-2B. In the second 

scenario, the existing foundation system can be enhanced to span or accommodate the estimated 

extensional fissures. In the third scenario, a debris wall, as shown in Figure 23.10.4-2C, can be 

built to protect the building or techniques in Figure 23.10.4-2B can be used to stabilize the slope. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 23 – Foundation Rehabilitation Techniques

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Figure 23.10.3-2: Mitigation Measure for Lateral Spreading – Stone/Gravel Column 
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Figure 23.10.4-1: Landslide Hazards
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Existing Buildings: FEMA 547 Chapter 23 – Foundation Rehabilitation Techniques

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Figure 23.10.4-2: Mitigation of Landslide Hazards 
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23.11 References 

ASCE, 2003, Standard for the Seismic Evaluation of Buildings, ASCE 31-03, Structural 

Engineering Institute of the American Society of Structural Engineers, Reston, VA. 

ASTM D1143, 1994, Standard Test Method for Piles Under Static Axial Compressive Load, 

ASTM. 

ASTM D1194, 1994, Standard Test for Bearing Capacity of Soil for Static Load and Spread 

Footings, ASTM. 

ASTM D3689, 1995, Standard Test Method for Piles Under Static Axial Tensile Load, ASTM. 

ATC, 1996, The Seismic Evaluation and Retrofit of Concrete Buildings, ATC-40, Applied 

Technology Council, Redwood City, CA. 

Bowles, J.E., 1996, Foundation Analysis and Design, McGraw-Hill, New York, NY. 

FEMA, 1997, NEHRP Commentary on the Guidelines for Seismic Rehabilitation of Buildings, 

FEMA 274, Federal Emergency Management Agency, Washington, D.C., November. 

FEMA, 2000, Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 

356, Federal Emergency Management Agency, Washington, D.C., November. 

FEMA, 2005, Improvement of Nonlinear Static Seismic Procedures, FEMA 440, Federal 

Emergency Management Agency, Washington, D.C. 

FHWA, 2000, Micropile Design and Construction Guidelines Implementation Manual, Report 

No. FHWA-SA-97-070., June. 

PTI, 1996, Post-Tensioning Institute Recommendations for Prestressed Rock and Soil Anchors, 

Post-Tensioning Institute (PTI), Phoenix, AR. 

Warner, J., 2004, Practical Handbook of Grouting, John Wiley & Sons, Hoboken, NJ. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 24 – Reducing Seismic Demand

24-1 

Chapter 24 - Reducing Seismic Demand 

24.1 Overview 

Most seismic rehabilitation projects utilize rehabilitation strategies involving adding strength, 

stiffness, ductility, and/or improvement load path details. Another approach, less commonly 

employed, is to reduce the seismic demand on the structure. This chapter covers three methods 

of reducing seismic demand on the structure: reducing the effective seismic weight, seismic 

isolation, and passive damping. 

24.2 Reduction of Seismic Weight 

Reduction of seismic weight may reduce the seismic demand on an existing structure in certain 

cases; however, the engineer must carefully evaluate the dynamic effects of such an approach 

before adopting it as part of a retrofit scheme. Techniques may include replacing heavy cladding 

with a curtain wall system, removing high permanent live loads, or removing upper stories. 

Since removing upper stories results in a loss of usable floor area, this approach is usually 

considered after an owner has built a new adjacent building that provides replacement space. 

While the reduction of seismic weight may potentially improve performance by changing a 

structure’s yielding sequence, reducing story drifts, reducing global overturning, or reducing 

base shear, these reductions in demand, particularly base shear, may not be directly proportional 

to the decrease in seismic weight. For example, the removal of a building’s upper stories will 

typically shorten the structure’s fundamental period of vibration, often leading to an increased 

spectral acceleration. If this increase in spectral acceleration is greater than the corresponding 

decrease in seismic weight, the demand base shear on the structure will increase. Buildings with 

periods in the velocity-sensitive region of the response spectrum should be evaluated for this 

effect early in the development of a rehabilitation strategy. In particular, tall buildings in which 

much of the seismic weight is concentrated in the lower stories are likely to have very limited 

benefits in base shear reduction associated with the removal of upper stories. 

The following examples in Figures 24.2-1, 24.2-2 and 24.2-3 illustrate how the removal of a 

building’s upper stories may not significantly decrease the calculated base shear demand. All 

three examples assume a concrete moment frame system and employ ASCE 7-05 (ASCE, 2005) 

base shear equations. The first examines a generic model building with uniform story heights 

and masses, using the ASCE 7-05 approximate period calculation. The second examines a 

similar model building, in which additional weight is concentrated in the lower stories. The third 

examines a real building, using fundamental periods calculated from computer analysis. In all 

three examples, the removal of upper stories decreases global overturning demands but does not 

significantly decrease base shear demands. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 24 – Reducing Seismic Demand

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Figure 24.2-1: Generic Building Example of Decreasing Base Shear with Decreasing Height 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 24 – Reducing Seismic Demand

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Figure 24.2-2: Generic Building Example of Increasing Base Shear with Decreasing Height 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 24 – Reducing Seismic Demand

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Figure 24.2-3: Real Building Example of Increasing Base Shear with Decreasing Height 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 24 – Reducing Seismic Demand

24-5 

24.3 Seismic Isolation 

Seismic or base isolation involves lengthening a building’s fundamental period of vibration to 

reduce the seismic demand transmitted from the ground to the building. It has been more 

commonly used in new building design, but it has also been employed in the United States for 

several high profile existing buildings as the key strategy in the rehabilitation design. 

Types of isolation components: Isolation components include elastomeric bearings and sliding 

bearings. Elastomeric bearings include high damped rubber, low damped rubber, and low 

damped rubber with lead cores. Sliding bearings include the friction pendulum system. 

Dampers are often part of the isolation system to limit displacements. See Section 24.4 for some 

discussion on dampers. 

Applicable buildings: The period range for isolated buildings is from about 2 seconds to 4 

seconds. As such, buildings on very soft soils and very tall, flexible buildings may not achieve 

much benefit from isolation. Seismic isolation is usually a very expensive rehabilitation strategy 

and has been primarily applied in the United States to important historic structures, usually as a 

way of minimizing the amount of superstructure strengthening an impact on the historic fabric of 

the building. Isolation displacements are highly site specific, but in high seismic zones the 

Maximum Capable Earthquake (MCE) displacements are often on the order of 30” or more. 

Sufficient clearance from adjacent structures is necessary to avoid pounding during seismic 

response 

System elements: In addition to the bearings and dampers, a complete isolation system will 

require a number of other special elements, including a moat around the building to 

accommodate the displacements. The moat has to go down past the plane of isolation. There is 

usually a complete or partial moat cover at the top of the moat for aesthetic or security 

considerations. Elevators are typically hung from the superstructure, as they cannot cross the 

isolation plane without special detailing. Utilities entering the building need to be able to 

accommodate the isolation displacements; this often triggers special vaults outside the building 

or areas under the building for joint details. A foundation is needed below the isolators to take 

the forces they impart, and a structural system is needed above the isolators as well to deliver 

forces to the isolators and resist the moments that are induced. All of these elements add to the 

cost of isolating the building. 

Analysis and design requires special expertise: The analytical and design effort for an isolated 

building is typically much more extensive than in fixed base rehabilitations. Time history 

analysis, where at least the isolators are modeled nonlinearly, is standard practice. Material 

properties must be achievable with components in the marketplace and must account for material 

variability from manufacturing, loading, temperature, velocity, wear, aging, and other effects. 

Experience with this type of work and the properties of the various vendor’s components and 

associated issues is quite useful. 

Determining the plane of isolation: Selecting the plane of isolation is a critical design choice. It 

is usually near the base of the building, though there are examples of isolation elements placed at 

the top of columns under heavy roofs to limit the forces in the columns. Isolation at the base is 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 24 – Reducing Seismic Demand

24-6 

either done either in the basement level, leading to some loss of use of the basement or under the 

basement, leading to additional excavation to place the isolators, the foundation below them and 

the superstructure assembly above them. The elements directly above and below the isolators 

are designed to take the Design Basis Earthquake (DBE) motions, without force reductions, 

which is a comparatively severe demand. The different types of isolation components have 

different sizes and means of transferring moments. With rubber bearings, P-delta moments are 

assumed to be split with half of the moment going up and half going down. With the traditional 

friction pendulum system, all of the P-delta moment goes up or down depending on which way 

the dish is oriented. How moments are resisted can lead to the selection of specific types of 

isolators. 

Reducing tension: Rubber bearings are much less stiff and have much less strength to resist 

tension forces. Lead cores in low damped rubber bearings also have limited capacity for 

resisting tension. Many engineers have concerns about sliding bearings under tension, though 

sliding bearings that resist some tension have recently come into the market. As a result, 

isolation layout and superstructure design is often aimed at minimizing tension in bearings. 

Transferring load from the existing building foundations to the new bearings: When a new 

isolated building is built, the columns and remaining elements of the superstructure can be 

erected directly on top of the isolation bearings. In an existing building, the superstructure is 

already in place. A key issue in design is developing details that delineate and facilitate the load 

transfer process of shoring the existing building, cutting the base of columns free, installing new 

foundations and new horizontal structure above the isolator, installing the isolators, and 

transferring load to the isolators without damaging movements of the superstructure. 

Proprietary/bidding considerations: Detailing for rubber and sliding bearings is quite different. 

If multiple vendors are necessary, vendors of different types of systems are usually considered. 

Often they are procured in an early package, due to the long lead time. This also permits the 

design engineer to move into final design knowing which type of system will be used. 

24.4 Energy Dissipation 

Adding damping to an existing structure, like seismic isolation, is a relatively unusual seismic 

rehabilitation strategy. The added damping reduces overall building displacement and 

acceleration response, and local interstory drifts; but it can impart additional localized forces that 

must be addressed. 

Types of damping components: FEMA 356 provides guidance for displacement-dependent 

devices or velocity-dependent devices. Displacement-dependent devices include devices that 

exhibit rigid-plastic (friction devices), bilinear (metallic yielding devices) or trilinear hysteresis. 

Velocity-dependent devices include solid and fluid viscoelastic devices and fluid viscous 

devices. There are other devices as well, including shape-memory alloys, friction-spring 

assemblies with recentering capability, and fluid restoring force-damping devices. 

Applicable buildings: Most engineers believe that adding damping is most relevant in flexible 

buildings, such as steel or concrete moment frames. Damping is also a common element in the 
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24-7 

seismic isolation system, but there it must accommodate very large displacements. See 

Section 24.3. 

System elements: Figures 24.4-1 and 24.4-2 show examples of adding damping devices in an 

existing steel moment frame building to minimize drifts and demands on the beam-column 

joints. Other dampers, such as wall dampers, are possible but not shown. The damper must be 

connected to the existing structure and potentially the foundation. Installing dampers is similar 

to installing braced frames. See Chapter 8 for detailed discussion on adding braced frames to a 

steel building and Chapter 12 for adding a braced frame to a concrete building. Some damper 

devices and orientations require out-of-plane bracing for stability, such as those shown in 

Figure 24.4-2. 

Figure 24.4-1: Damper Alternatives for Rehabilitating an Existing Moment Frame 
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Figure 24.4-2: Additional Damper Alternatives for Rehabilitating an Existing 

Moment Frame 

Analysis can require special expertise: The analytical and design effort for a rehabilitation 

design involving damping can be more extensive than in fixed base rehabilitations. Time history 

analysis, where at least the dampers are modeled nonlinearly, is common. Material properties 

must be achievable with components in the marketplace and must account for material variability 

from manufacturing, temperature, velocity, wear, aging, and other effects. Experience with this 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Chapter 24 – Reducing Seismic Demand

24-9 

type of work and the properties of the various vendor’s components and associated issues is 

useful. Hanson and Soong (2001) is a comprehensive monograph that covers analysis of 

buildings with supplemental energy dissipation devices, and it includes several design examples 

of seismic rehabilitation using damping devices. 

Aesthetic impact: Adding dampers looks very similar to adding a braced frame, with the 

resulting visual and programmatic impacts. Some dampers or their connections can be 

particularly visually obtrusive. 

Checking the existing structure: When dampers are added to the structure, the loads they impart 

locally must be considered in the design. 

Proprietary issues: Most dampers available on the market are proprietary. Material properties, 

testing histories, limitations and detailing considerations are obtained from the manufacturer. 

Like seismic isolation components, the particular category of damper such as a fluid viscous 

damper or a friction damper is usually selected early in the design because the analysis and 

detailing can be significantly different between categories. There is also a patent regarding 

certain techniques for connecting bracing and dampers to beams when sliding is employed. 

24.5 References 

ASCE, 2005, Minimum Design Loads for Buildings and Other Structures, ASCE 7-05, American 

Society of Civil Engineers, Reston, VA. 

Hanson, R.D. and T.T. Soong, 2001, Seismic Design with Supplemental Energy Dissipation 

Devices, Monograph MNO-8, Earthquake Engineering Research Institute, Oakland, CA. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Glossary

G-1 

Glossary 

BLOCKED DIAPHRAGM: A diaphragm in which all sheathing edges not occurring on framing members 

are supported on and connected to blocking. 

BOUNDARY ELEMENT: An element at the edge of an opening or at the perimeter of a shear wall or 

diaphragm. 

BOUNDARY NAILING: Nailing at the perimeter edge of a wood diaphragm to framing members and 

blocking below. 

BRACED FRAME: An essentially vertical truss, or its equivalent, of the concentric or eccentric type that is 

provided in a building frame or dual system to resist lateral forces. 

CHEVRON BRACING: Bracing where a pair of braces, located either both above or both below a beam, 

terminates at a single point within the clear beam span. 

CHORD: See DIAPHRAGM CHORD. 

COLLECTOR: A member or element provided to transfer lateral forces from a portion of a structure to 

vertical elements of the lateral-force-resisting system (also called a drag strut). 

CONCENTRICALLY BRACED FRAME (CBF): A braced frame in which the members are subjected 

primarily to axial forces. 

CONTINUITY PLATES: Steel column stiffeners at the top and bottom of the panel zone. They are also 

known as transverse stiffeners. 

CONTINUITY TIES: Structural members and connections that provide a load path between diaphragm 

chords to distribute out-of-plane wall loads. 

COUPLING BEAM: A structural element connecting adjacent shear walls. 

DAMPING: The internal energy absorption characteristic of a structural system that acts to attenuate induced 

free vibration. 

DEMAND: The prescribed design forces required to be resisted by a structural element, subsystem, or system. 

DIAPHRAGM: A horizontal, or nearly horizontal, system designed to transmit lateral forces to the vertical 

elements of the lateral-force-resisting system. The term "diaphragm" includes horizontal bracing systems. 

DIAPHRAGM CHORD: The boundary element of a diaphragm or shear wall that is assumed to take axial 

tension or compression. 

DIAPHRAGM STRUT: The element of a diaphragm parallel to the applied load that collects and transfers 

diaphragm shear to vertical-resisting elements or distributes loads within the diaphragm. Such members may 

take axial tension or compression. Also refers to drag strut, tie, or collector. 

DOUBLER PLATE: A steel plate added to a panel zone to increase panel zone strength. 

DRAG STRUT: See COLLECTOR. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Glossary

G-2 

DRIFT: See STORY DRIFT. 

DUCTILITY: The ability of a structure or element to dissipate energy inelastically when displaced beyond its 

elastic limit without a significant loss in load-carrying capacity. 

ECCENTRICALLY BRACED FRAME (EBF): A diagonal braced frame in which at least one end of each 

brace frames into a beam a short distance from a beam-column joint or from another diagonal brace. 

EDGE NAILING: Nailing at the perimeter edge of a wood structural panel in a shear wall or diaphragm to 

framing members and blocking. 

FIELD NAILING: Nailing within the interior of a wood structural panel in a shear wall or diaphragm to 

framing members. 

FUNDAMENTAL PERIOD OF VIBRATION: The time it takes the predominant mode of a structure to 

move back and forth when vibrating freely. 

HORIZONTAL BRACING SYSTEM: A horizontal truss system that serves the same function as a diaphragm. 

K-BRACING: Bracing where a pair of braces located on one side of a column terminates at a single point 

within the clear column height. 

LAMINATED VENEER LUMBER: An engineered wood product created by layering dried and graded wood 

veneers with waterproof adhesive into blocks of material. It is also known as structural composite lumber. 

LATERAL FORCE-RESISTING SYSTEM: That part of the structural system assigned to resist lateral forces. 

LINK BEAM: That part or segment of a beam in an eccentrically braced frame that is designed to yield in 

shear and/or bending so that buckling or tension failure of the diagonal brace is prevented. 

MOMENT RESISTING SPACE FRAME: A structural system with an essentially complete space frame 

providing support for vertical loads. 

NOTCH TOUGHNESS: A measure of material ductility related to the ability to resist fracture. It is typically 

measured with Charpy V-notch (CVN) test standards. 

PANEL ZONE: Area of the beam-to-column connection delineated by beam and column flanges. 

REDUNDANCY: A measure of the number of alternate load paths that exist for primary structural elements 

and/or connections such that if one element or connection fails, the capacity of alternate elements or 

connections are available to satisfactorily resist the demand loads. 

RE-ENTRANT CORNER: A corner on the exterior of a building that is directed inward such as the inside 

corner of an L-shaped building. 

SHEAR WALL: A wall, bearing or nonbearing, designed to resist lateral forces acting in the plane of the 

wall. 

SHOTCRETE: Concrete that is pneumatically placed on vertical or near vertical surfaces typically with a 

minimum use of forms. 

SOFT STORY: A story in which the lateral stiffness is less than 70 percent of the stiffness of the story above. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Glossary

G-3 

SOIL-STRUCTURE RESONANCE: The coincidence of the natural period of a structure with a dominant 

frequency in the ground motion. 

STORY DRIFT: The displacement of one level relative to the level above or below. 

STRUCTURE: An assemblage of framing members designed to support gravity loads and resist lateral forces. 

Structures may be categorized as building structures or nonbuilding structures. 

SUBDIAPHRAGM: A portion of a larger wood diaphragm designed to anchor and transfer local forces to 

primary diaphragm struts and the main diaphagm. 

SUBSYSTEMS: One of the following three principle lateral-force-resisting systems in a building: vertical 

resisting elements, diaphragms, and foundations. 

SUPPLEMENTAL ELEMENT: A new member added to an existing lateral-force-resisting subsystem that 

shares in resisting lateral loads with existing members of that subsystem. 

TIE-DOWN: A prefabricated steel element consisting of a tension rod, end brackets and bolts or lags used to 

transfer tension across wood connections. It is also known as a hold-down. 

V-BRACING: Chevron bracing that intersects a beam from above. Inverted V-bracing is that form of chevron 

bracing that intersects a beam from below. 

VERTICAL-RESISTING ELEMENTS: That part of the structural system located in a vertical or near 

vertical plane that resists lateral loads (typically a moment frame, shear wall, or braced frame). 

WEAK STORY: A story in which the lateral strength is less than 80 percent of that in the story above. 

WOOD STRUCTURAL PANEL: A wood-based panel product that satisfies the requirements of Voluntary 

National Product Standard PS-1 or PS-2 and is bonded with waterproof adhesive. Included under this 

designation are plywood, oriented strand board (OSB) and composite panels. 

X-BRACING: Bracing where a pair of diagonal braces crosses near mid-length of the bracing members. 
Techniques for the Seismic Rehabilitation of

Existing Buildings: FEMA 547 Abbreviations

A-1 

Abbreviations 

The following abbreviations are commonly used by structural engineers and have been used in 

figures and/or text throughout the document. 

B.N. Boundary nailing 

BRBF Buckling-restrained braced frame 

CIP Cast-in-place 

C.J. Construction joint 

CJP Complete joint penetration weld 

CL Centerline 

CP Complete penetration weld 

(E) Existing 

EA. Each 

E.N. Edge nailing 

F.N. Field nailing 

FTG. Footing 

MC Moisture content 

M/E/P Mechanical/electrical/plumbing 

(N) New 

PL Plate 

PP Partial penetration weld 

RBS Reduced beam section (in a beam-to-column moment frame connection) 

SPSW Steel plate shear wall 

TYP. Typical 

WSMF Welded steel moment frame 
Techniques for the Seismic Rehabilitation 

of Existing Buildings: FEMA 547 Workshop Participants

WP-1 

Workshop Participants 

A workshop was held on September 20-21, 2005 in Oakland, California to solicit 

comment on a draft of the final document. Workshop participants included the following 

individuals. 

ICSSC Members 

Krishna Banga 

Department of Veteran Affairs 

Washington, D.C. 

Cathleen Carlisle 

FEMA 

Washington, D.C. 

James A. Caulder 

Air Force Civil Engineer Support Agency 

Tyndall AFB, FL 

James Farasatpour 

Federal Aviation Administration 

Inglewood, CA 

Jack Hayes 

U.S. Army 

Champaign, IL 

Richard Kahler 

U.S. Navy 

Norfolk, VA 

H.S. Lew 

NIST 

Gaithersburg, MD 

Dai H. Oh 

U.S. Department of State 

McLean, VA 

Raymond F. Schuler 

NASA Ames Research Center 

Moffett Field, CA 

 

Academics and Industry 

Daniel Abrams 

University of Illinois 

Urbana, IL 

Melvyn Green 

Melvyn Green & Associates 

Torrance, CA Richard Howe 

Memphis, TN 

Lawrence Reaveley 

University of Utah 

Salt Lake City, UT 

Andrew W. Taylor 

KPFF 

Seattle, WA 

James Parker 

Simpson, Gumpertz & Heger 

Waltham, MA 

Barry H. Welliver 

Draper, UT 
Techniques for the Seismic Rehabilitation 

of Existing Buildings: FEMA 547 Workshop Participants

WP-2 

Project Review Panel 

Daniel Dolan 

Terry Dooley 

Kurt Gustafson 

Robert D. Hanson 

Neil Hawkins 

Jim Harris 

Bela I. Palfalvi 

Daniel Shapiro 

Technical Update Team and Contributing Staff 

Kelly Cobeen 

William T. Holmes 

Jack Hsueh 

Bret Lizundia 

James Malley 

Karl Telleen