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A 11 10 5 AS3M25

NBSIR 88-3820

Collapse of Masonry Wall Under

Construction in Pawtucket, Rl,

October 28, 1985

Felix Y. Yokel

U.S. DEPARTMENT OF COMMERCE

National Bureau of Standards

National Engineering Laboratory

Center for Building Technology

Gaithersburg, MD 20899

July 1988

Prepared for:

Office of Solicitor

national Safety and Health Administration

Q6 epartment of Labor

2qo ederal Building, Room 1803

ii cr nment Center

# 88-3820

1988

n, MA 02203

75 Years Stimulating America's Progress

1913-1988

NBSIR 88-3820

COLLAPSE OF MASONRY WALL UNDER

CONSTRUCTION IN PAWTUCKET, Rl,

OCTOBER 28, 1985

iKeaearcL information Center

National Bureau of Standards

Gaithersburg, Maryland 208^9

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(Xctoo

. [LSI*

fio. % % -J

int

Felix Y. Yokel

U.S. DEPARTMENT OF COMMERCE

National Bureau of Standards

National Engineering Laboratory

Center for Building Technology

Gaithersburg, MD 20899

July 1988

Prepared for: Office of Solicitor

Occupational Safety and Health Administration

U.S. Department of Labor

JFK Federal Building, Room 1803

Government Center

Boston, MA 02203

U.S. DEPARTMENT OF COMMERCE, C. William Verity, Secretary

NATIONAL BUREAU OF STANDARDS, Ernest Ambler, Director
V • .
ABSTRACT

Results from a study to determine the cause of the October 28,

1985 collapse of a masonry wall under construction in Pawtucket,

RI are presented. The wall was a 60 ft-3 in. long, 23 ft-6in.

high partially reinforced concrete masonry wall supported by

wooden braces. The study included: inspection of construction

plans and specifications? review of construction records and

eyewitness accounts recorded immediately after the collapse, as

well as testimony from OSHA inspectors and local building

officials who visited the site a short time after the collapse;

examination of photographs taken by OSHA inspectors and police

investigators; analysis of meteorological data? and a stability

analysis of the collapsed wall. It is concluded that the

collapse was probably caused by a gust of wind which exerted

lateral forces which exceeded the lateral-load capacity of the

wall and its supporting wooden braces. Contributing factors were

the lack of grout in the masonry cores which contained the steel

reinforcement dowels and the inadequate anchoring of the dowels

in the foundation.

Keywords: building? construction loads; construction safety?

lateral bracing? masonry construction; stability?

structural collapse; timber shoring? wind loads.

iii

TABLE OF CONTENTS

ABSTRACT iii

1. INTRODUCTION 1

2. ESTIMATED WIND FORCES ACTING ON THE WALL AT THE

TIME OF COLLAPSE , 3

3. ESTIMATED LOAD RESISTANCE PROVIDED BY THE BRACES 8

3.1 Probable Bracing Configuration .8

3.2 Estimated Load Capacity of the Bracing ........... 18

4. LOAD RESISTANCE PROVIDED BY THE WALL ....25

4.1 General Discussion ...25

4.2 Condition of Wall 25

4.3 Estimated Resisting Moment at the Base

of the Wall 33

4.4 Resisting Moment of Mortar and Grouted

Dowels at the Base of the Wall 35

5. ANOMALIES IN THE CONSTRUCTION SEQUENCE THAT

CONTRIBUTED TO THE RISK OF COLLAPSE 36

6. PROBABLE COLLAPSE MECHANISM 36

7. COMPLIANCE OF SHORING WITH EXISTING PROVISIONS 38

8. SUMMARY 42

9. REFERENCES 43

v

1 . INTRODUCTION

On October 28, 1985 at approximately 9:45 AM a partially

reinforced concrete masonry wall which was under construction

collapsed at the construction site of a Super Stop and Shop store

located near the southeasterly corner of the intersection of

Beverage Hill Avenue and Route 1A in Pawtucket, RI. The collapse

occurred during a coffee break, when workers were sitting near

the wall to get protection from wind. Three construction workers

were killed and two were injured as a result of the collapse.

This report summarizes the findings of a National Bureau of

Standards (NBS ) study conducted at the request of the

Occupational Safety and Health Administration (OSHA) . The study

was initiated in November 1986 and the findings were transmitted

to OSHA in a memorandum report dated January 16, 1987. The

purpose of the study was to analyze possible causes of the wall

collapse and reasons why the bracing which supported the wall

during construction did not prevent the collapse. Comparisons

were also made between the strength of the bracing as installed

and that required by applicable OSHA regulations and other

applicable standards and criteria for bracing.

The wall extended from east to west, was 60 ft-3 in. long and 2 3

ft-6 in. high and had a 6x2 0 ft window opening 4 ft above the

base of the wall and 25 ft-8 in. from its east end (see figure

1) . The wall was supported by wooden braces. The collapse was

apparently triggered by a wind gust from the northerly direction.

The masonry construction was carried out by a sub-contractor who

was retained by the general contractor. Data on the construction

details of the wall were obtained from plans and specifications

prepared by the consulting architect, and from engineering

inspection reports prepared by an inspection laboratory employed

by the general contractor. Evidence from the collapse site in the

1
« <E ,u,ur# future

Figure It Elevation of Wall at the Time of Collapse , Looking from

South.

2
form of Occupational Safety and Health Administration (OSHA) and

police photographs, statements by OSHA inspectors who visited the

site, and police records, including evidence by eye witnesses and

photographs, was obtained from the Office of the Solicitor,

Department of Labor, Boston, MA, the OSHA office in Providence

RI, and the Building Commissioner, State of Rhode Island. NBS

personnel visited the site in November 1986.

2. ESTIMATED WIND FORCES ACTING ON THE WALL AT THE TIME OF

COLLAPSE

The following information was obtained from the National Weather

Service (NWS) station anemometer located between runways of the

Warwick, RI airport, about 12 miles South from the site of the

accident, for October 28, 1985, 9:58 AM : Average wind speed 17

mph, gusting to 24 mph; wind from north-westerly direction; clear

day with some high clouds. Wind speed was averaged over 1 minute

and measured 2 0 ft above ground. The data were obtained by

sampling the wind speed at 1-hour intervals for 1 minute (no

attempt was made to pick out the "fastest" minute)

Other data for that day indicate the following:

Time Average 1 min. Max. gust Direction

(L.S.) wind speed (degrees)

8 AM 16 mph — 350

9 AM 15 mph 320

11 AM 15 mph 22 mph 330

12 AM 16 mph 24 mph 340

Figure 2 shows a stripchart recording of wind speeds taken at the

same NWS station. The wind speed in the chart is recorded in

knots (1 knot = 1.151 mph). Note that the wind speed increased

somewhat between 9:55 AM LST and 10:30 LST. At the accident site,

this increase in wind speed probably occurred somewhat earlier,

since the wind came from a northerly direction. From an

3
G G O G G O c o c G

Stripchart Recording of Wind Speeds in Knots, Taken on

August 28, 1985 at an Elevation of 20 ft* Above Ground

at the NWS Station at the Warwick RI Airport c

4
examination of the recording, the fastest mile wind speed is

estimated to have been 20 knots or approximately 23 mph.

Since this was a clear, windy day with no thunderstorms (which

may be localized over a limited area) it is appropriate to assume

that wind conditions at the location of the accident were similar

to those at the weather station, differing only according to the

characteristics of the terrain. It is assumed that the terrain at

the weather station corresponds to "exposure category C" (open

terrain with scattered obstructions having heights generally less

than 30 ft) (refer to ref.l) and the terrain at the construction

site is also in exposure category C, even though it is somewhat

less exposed than the terrain at the weather station.

In accordance with reference 1, the total force exerted on a

structure by wind can be calculated by the following equation:

F = qz Cf Af (Eq.l)

where F is the force in lb, qz is the velocity pressure evaluated

at height z above ground, in psf, is a gust response factor

evaluated at a height z = H , where H is the height of the

structure, Cf is a force (drag) coefficient, and Af is the area

of the structure projected in a plane normal to the wind.

The velocity pressure qz in Eq.l can be calculated as follows:

qz = 0.00256 V2 (Eq.2)

where V is the "fastest mile" wind speed, which is used as a

reference speed in (1) , measured at height z above ground.

The gust response factor for buildings in Ref.l is given in Table

8, Ref.l. However in this instance, because of the size and

aspect ratio of this wall, a gust response factor of 1 is more

5
appropriate (refer to Fig. 3, Ref.l). The value of 1 is less, but

considered more realistic than the more conservative gust

response factors recommended in Table 8. The gust response factor

Gjj = 1 is used to estimate the windloads that acted on the wall.

The force (drag) coefficient Cf for rectangular flat plates which

have aspect ratios similar to that of the wall under discussion

and which have one edge in contact with the ground is

approximately 1.2 for wind normal to the plate. Data from wind

tunnel tests on flat plates which are not in contact with the

ground indicate that as the direction of the wind changes from

normal to 45°, (calculated with respect to the full plate

area, rather than the area projection normal to the wind

direction) first slightly increases and then returns to

approximately 1.2 for a wind direction which is inclined at about

45° with respect to the plate (Ref.2, ppl56, 157, fig. 4. 6. 3). The

wind direction was apparently somewhere between normal to the

wall and 45° inclination (the records indicate that the wind

direction was somewhere between North and North-West). The 1.2

drag coefficient is in agreement with Cf stipulated in Ref.l for

"solid signs at ground level" (Table 13, Ref.l).

Wind speed in mph as a function of height above ground can be

approximated by the equation:

V = K z 1/ 7 for "exposure C" (Ref.l flat open terrain) .... (Eq. 3

)

where K is a coefficient which depends on wind speed and z is the

height above ground in ft.

The estimated wind load acting on the wall can be reasonably

represented by a uniformly distributed pressure, qe , acting

normal to the wall surface from the windward direction (actually

it consists of pressure on the windward side and suction on the

leeward side). Pressure qe can be derived from Eq.l, using a

6
value of qz calculated for a z value of z = H, where H is the

height of the wall (23.5 ft.), a value of 1, and a Cf value of

1.2. Thus, from Eqs. 1, 2, and 3:

qe = 0. 00256x1. 2xK2 xH2 / 7 (Eq.4)

where qe is the equivalent uniformly distributed pressure

representing the wind load in psf, and K is the wind speed

coefficient in Eq.3.

To obtain the estimated fastest mile wind speed at the construction site, it is necessary to account for the difference between

the terrain at the weather station and that at the construction

site. In this instance it is estimated that the fastest mile wind

speed at the construction site was close to that recorded at the

weather station, which was 23 mph.

In accordance with Eq.3 the K coefficient for the 23 mph wind

speed is 14.99. The equivalent wind pressure on the wall,

calculated by equation 4 is therefore:

qe = 0.00256x1. 2x14. 99 2 x23.5 2 / 7 = 1.701 psf.

The following wind loads are calculated for the wall as a whole,

using the 2 3 mph fastest mile wind speed derived from from the

stripchart recording:

Mw = 26,870 ft-lb; Vw = 2,204 lb

where Mw is the total overturning moment at the base of the wall

caused by the wind load, and Vw is the total resultant shear

force at the base of the wall.

High winds were also recorded on the three days prior to the

accident. On Sunday, October 27, the recorded fastest mile wind

7
was 25 mph from a direction of 300°, on Saturday October 26 , the

fastest mile wind was 18 mph from a direction of 320° , and on

Friday October 25, the fastest mile wind was 21 mph from a

direction of 310°. These wind conditions could have contributed

to the the wall failure by rupturing the mortar bond or otherwise

weakening the resisting moment of the wall prior to the accident.

3. ESTIMATED LOAD RESISTANCE PROVIDED BY THE BRACES

3 . 1 Probable Bracing Configuration

Figure 3 shows braces supporting walls on the same construction

site that did not collapse in the accident. The photographs were

taken on the day of the accidento The braces shown in figure 3(a)

consisted of two 16 ft long 2x10 planks » The upright plank had a

2x4 cleat nailed to it which resisted the upward thrust of the

diagonal strut. The diagonal strut rested on the soil with no

noticeable embedment and no footing to resist downward thrust,

and no stakes were driven into the soil to resist horizontal

forces and displacements. The diagonal struts are restrained by

one concrete masonry block placed at their lower end and weighted

down by two concrete masonry blocks, apparently to hold them in

place. A closeup photograph of the lower end of the diagonal

strut is shown in figure 3(b). Figure 3(c) shows a strut which

was inclined at a very steep angle, apparently to brace a higher

wall. The upper portion of that latter brace is shown in figure

3(d). The vertical bracing members examined in the field were not

nailed or otherwise attached to the masonry wall and were held in

place by the horizontal thrust exerted by the diagonal member

(aided by the weight of the two masonry blocks resting on the

diagonal member)

.

The evidence shown in figure 3 indicates for the braces shown

that: (1) the two 16 ft long planks were used even when the wall

was very high (in the latter case the angle of inclination of the

8
Figure 3: Braces Supporting Other Walls on Construction Site

(photograph taken on day of collapse)

(a) Typical braces.

(b) Strut support at ground level.

(c) Brace with steeply sloped strut.

(d) Cleat for brace with steeply sloped strut.

9
diagonal strut to the horizontal was increased in order to

support the wall at a higher level) ; (2) no intermediate support

was provided to reduce the unbraced length of the diagonal strut;

(3) no attempt was made to anchor the lower end of the diagonal

strut in the soil or otherwise prevent it from moving

horizontally (except for the restraint provided by the weight of

one concrete block) or to increase the soil bearing area in order

to prevent excessive settlement of the strut.

Figure 4 shows the evidence of bracing found on the windward

(north) side of the collapsed wall. Figure 4(a) shows a view from

east to west. It shows collapsed braces at the east end of the

wall (column B9 in plan drawing A 1.1, dated July 15, 1985) and

near the center of the wall (to the east of the window

opening). It can be seen that, like the braces shown in figure 3,

each brace consisted of two 16-ft long 2x10 planks, one of which

had a cleat nailed to it. Figure 4(b) shows an end view of the

collapsed brace in the center of the wall together with the three

concrete blocks which were used to restrain the diagonal

strut. Figure 4(c) shows the brace at the west end of the wall

(col B8.1) and figure 4(d) shows the restraining concrete blocks

for the brace in figure 4(a). The western diagonal was restrained

by a pipe (not shown) . The distance from the lower end of the

vertical plank to the underside of the cleat was 9 ft 9 in for

the brace at the east and west end of the wall and 10 ft

-4 in. for the brace at the center.

It is concluded from figure 4 that: (1) the braces used on the

windward side of the collapsed wall were similar to those shown

in figure 3; (2) the wall was supported by three braces on the

windward side; (3) the height at which the diagonal struts

supported the wall equaled or exceeded the distance from the

bottom of the vertical plank to the underside of the cleat, but

could not be less than that distance (for instance in figure 3(c)

the bottom of the vertical plank is higher than the ground, but

10
Figure 4: Collapsed Braces on the Windward Side of the Wall.

(a) View from East to West showing collapsed braces

near East end and near center of wall.

(b) End view of collapsed brace near center of wall.

(c) Collapsed brace near West end of wall.

(d) Restraining concrete masonry units for brace near

East end of wall.

11
it is unlikely that the vertical plank was pushed into the

ground)

.

Figure 5 shows some of the evidence of bracing found on the

leeward (south) side of the collapsed wall. In this instance,

bracing members were damaged either by the wind load prior to the

collapse or by the falling debris from the wall. Figure 5(a)

shows a vertically split 2x10 plank at the east end of the wall

(col.B9). This was a vertical bracing member and the cleat that

has been nailed to it near the upper end can be seen in the

picture. Figure 5(b) shows a shattered 2x10 plank on the west end

of the wall. Figure 5(c) shows a broken 2x10 plank, covered by

debris, near the west end of the wall, looking south. Examination

of the field evidence which started within about 60 minutes after

the collapse and continued through the removal of the debris

turned up components of two braces similar to those shown in

figure 3. There is no evidence that there was a third brace near

the center of the wall. The location of the cleats in the two

braces found was 16 inch from the top of the vertical plank to

the underside of the cleat for the brace at the east end of the

wall, and 24 inches from the top of the plank to the underside of

the cleat for the brace at the west end of the wall.

It is also important to note that the ground elevation at the

leeward side of the wall was below the base of the wall (part of

the foundation was exposed-see figure 1) . At the east end of the

wall the ground elevation was approximately 20 inches below the

base of the wall and at the west end the ground elevation was

approximately 10 inches below the base of the wall.

It is deduced from figure 5 and observations reported by OSHA

inspectors that, on the leeward side, the wall was probably

supported by two braces located near the two ends of the wall,

which were similar to the braces shown in figure 3, and that near

the east end of the wall (B9 in plan drawing A 1.1) the point of

12
Figure 5: Bracing Members Found on Leeward Side of Wall.

(a) 2x10 plank with cleat near East end of wall.

(b) Shattered 2x10 plank near West end of wall.

(c) Broken 2x10 plank near West end of wall/ looking

South

13
support provided by the diagonal brace was 13 ft or more above

the base of the masonry wall (the top of the footing) [16 ft - 16

inches (top of plank to bottom of cleat) - 20 inches (base of

wall to ground) ]

.

Near the west end of the wall (B8ol) the point of support

provided by the diagonal brace was 13 ft-2 inches or more above

the base of the masonry wall (16 ft - 24 inches - 10

inches) . This implies that, even if the vertical members rested

on the ground, the inclination of the diagonal member was steep,

similar to that of the brace shown in figure 3(c). On the east

end the inclination was 14.7 vertical to 6.4 horizontal or

approximately 67° to the horizontal. On the west end the inclination was 14 vertical to 7.8 horizontal or approximately 61° to

the horizontal.

Figure 6 shows a plan view of the wall and the probable location

of the braces on both sides of the wall, based on measurements

and observations by OSHA inspectors who visited the site after

the collapse. The cross section of the collapsed wall and the

probable geometry of the braces on the leeward side which was

deduced from the available evidence is shown in figure 7. There

was 8 in. block to an elevation 10 ft-8 in. above the base of the

wall and 12 inch block above that elevation. It is therefore

logical, that the support was provided above the 10 ft-8 in. elevation from the base of the wall. Evidence from field observations supplied by OSHA inspectors indicates that during construction the overhanging portion of the 12 in wall was supported by

vertical 2x4 shores to prevent uneven settlement of the mortar

bed. No traces of these 2x4 shores were found in the debris of

the collapsed wall, indicating that these support members were

probably removed before the collapse. Figure 8 shows the photograph of another wall on the construction site from which the 2x4

shores were removed, which was taken one day after the collapse.

14
N

Figure 6: Top View

Location

of the Collapsed Wall Showing Probable Bracing

15
Figure 7: Probable Bracing Configuration on Leeward Side of

Collapsed Wall.

16
Figure 8: Wall on the Same Construction Site from which the 2x4

Planks Supporting the Overhanging Wall Were Removed.

17
Evidence from photographs and reports by OSHA inspectors from the

collapse site indicates that in addition to the braces described

above there were scaffolds adjacent to the wall. These consisted

of scaffolds under the window opening and scaffolds on the

leeward side at the two ends of the wall. The scaffolds under the

window opening were originally used to provide support to the

concrete forms for the lintel. The scaffolds were adjusted by

screw jacks at the bottom. A.fter the lintel was poured, the screw

jacks were relaxed to remove the forms. Evidence provided by the

OSHA inspectors indicates that after removal of the forms these

screw jacks were tightened lightly, but no upward thrust was

exerted against the wall. The scaffolds at the two ends of the

wall were left in place for future use to finish the connection

of the ends of the wall to the columns which were to be erected

at both ends of the wall. These scaffolds did not bear against

the wall. The remains of the scaffold that supported the lintel

form are shown in figure 9(a). Figure 9(b) shows a scaffold

recovered from the wall (on side) and an undamaged scaffold. The

probable position and configuration of the scaffolds prior to the

collapse is shown in figures 10 and 11. There is no indication

that these scaffolds could provide significant lateral support to

the wall.

3 . 2 Estimated Load Capacity of the Bracing

Several possible failure mechanisms could have led to the

collapse of the braces shown in figure 7. Possible failure

mechanisms considered are: (1) structural failure of diagonal

plank; (2) upward sliding of vertical plank; (3) kickout or

settlement of diagonal plank at ground level; (4) shortening of

diagonal plank as a result of sag. These failure mechanisms are

discussed below.

18
Figure 9: Scaffolding that was Used Along the Collapsed Wall.

(a) Remain of Scaffold that Supported Lintel Forms.

(b) Scaffold Recovered from Collapsed Wall (on side)

19
Tubular welded frame scaffolds, approx. 18' high

remain at the concrete block wall ends

East

Bond beam

LA Foundation

B.8.1 Screw jacks

Ground level

B.9'

-Tubular weided frame scaffolds approx. 9’ high

with screw jacks, erected through window opening

to support lintel, formwork

Figure 10.° Front View of Scaffolds Looking From South

.

20
Bond beam

3 6’ tubular welded

frame sections erected on top of each other

with an outrigger

12

' block wail erected on top

of an 8* block wall to an approximate height of 23'6‘

Bond beam

A 3' on top of a 6' tubular

welded frame scaffold erected

through the window opening

used to support the formwork

for the poured concrete lintel

Screw jack

North side of wall

Ground level

Foundation wall

Screw jack (relaxed before collapse

to remove lintel forms)

Figure 11. Side View of Position of Scaffolds

21
(1) Structural failure of diagonal strut:

The actual dimensions of the 2x10 strut are 1.5 in. x 9.25

in. The 1.5 in. narrow dimension, combined with the 16 ft length

make this strut a very slender member. Neither of the ends of the

strut has any significant restraint against rotation, so that the

strut will act like a column which is pin ended at both ends. Its

propensity for buckling is enhanced by the initial downward sag,

caused by its own weight. The upper bound for the ultimate

buckling load can be calculated by the "Euler equation":

P = (3.1416 2 El ) / (L) 2 ........... (Eq.7)

where: E is Young's Modulus for the wood, estimated at 10 6 psi

I is the moment of inertia of the cross section

L is the length of the plank = 16x12 inches.

I = ( 9 . 25x1 .

5

3

) /12 = 2.6010 in4

P = ( 3 . 14 16 2 x10 6 x2 .6016)/(16x12) 2 = 697 lb.

The actual ultimate load capacity of the strut, assuming that

flexural failure would occur prior to buckling as a result of the

amplification of the initial downward sag by the axial load, at

an approximate ultimate bending stress of 2,500 psi, would be 660

lb. Only the horizontal component of this load can effectively

resist the wind pressure. The following resisting moments would

therefore be exerted by the struts on the base of the wall at the

instant of strut failure:

At B9 (east side): Mr = 660x13x6.39/16 = 3,427 ft-lb

At B8.1 (west side): Mr = 660x13.17x7.75/16 = 4,210 ft-lb

22
Since the window opening is nearer to the west end, the east end

brace which has a lower load capacity was probably more heavily

loaded. But even if both struts failed simultaneously, the

maximum overturning moment over the base of the wall that could

be resisted by the braces was 7,600 ft-lb. This compares with an

estimated wind induced overturning moment of 26,900 ft-lb. Thus

if the braces alone had to resist the wind load, their failure

under the prevailing wind condition was a probable event.

(2) Upward Sliding of Vertical Plank

The vertical component of the axial strut load must be resisted

by frictional forces between the vertical plank and the masonry

wall. Since the horizontal component of the axial strut force is

less than 50% of the vertical component (refer to figure 6) , a

friction coefficient of more than 2 is required to develop the

required frictional resistance. Upward sliding of the vertical

plank would cause the brace to fail. No reliable data on the

friction coefficient between an aged timber plank and a concrete

masonry wall could be located and no laboratory experiments were

conducted to obtain information. However the required friction

coefficient is high when compared with accepted engineering

practice. For instance ref . (3) recommends a range from 0.5 to 0.6

for the static coefficient of friction of masonry on wood for the

purpose of engineering design (Table 3.1). Sliding failures

could have been prevented by nailing the vertical board to the

wall or by a cleat under the 12 in. wall at the transition point

between the 8 in. and 12 in. wall. No evidence was found to

indicate that either of these precautionary measures was taken.

(3) Kickout or Settlement of Diagonal Plank '

It is deduced from figures 2 and 3 that the diagonal plank simply

rested on the ground, was restrained by one concrete block and

weighted down by two blocks. In accordance with field density

23
tests performed on the construction site the soil on which the

planks rested was compacted fill, which is considered competent

soil for support of proper footings. However the arrangement as

installed entirely relied on the surface condition of the soil

(the upper 2 to 5 inches) , which is not predictable. Failure by

horizontal displacement is not considered likely in this case,

because of the steep inclination of the diagonal strut. However

there was little resistance to settlement (pushing of the strut

into the ground) , at least for the initial 1 to 2 inches until

the soil under the strut is compressed by the downward thrust. It

is important to note that each 1 in. of downward displacement of

the strut in the inclined direction of its axis would cause a

horizontal displacement of 2 to 2.5 inches at the point where the

strut supports the wall. Good practice would require, as a

minimum, to restrain the diagonal struts by stakes, driven into

the ground (ref. 4), or preferably to provide an adequate footing

to prevent displacement at the ground level (ref. 5).

(4) Shortening of Diagonal Plank

The diagonal strut is very slender and flexible. A large lateral

displacement at the cleat is required to develop the load capacity of the strut. By the time the diagonal strut is loaded to

630 lb it experiences a deflection (sag) of approximately 8 in.

in its center which in turn will cause the distance between the

cleat and the ground support of the strut to shorten by more than

1 inch (measured along the initial undeflected axis of the

diagonal strut) . This shortening is associated with a horizontal

displacement of the wall at the cleat of more than 2.5 inches. At

this displacement the resisting moment provided by the gravity

load of the wall itself would be zero. The displacement caused by

the sag of the plank could be further increased by settlement of

the strut at its ground support. This "softness" of the strut

will cause an initial failure at the base of the wall and a tilt

in the wall before the brace resistance is effectively mobi24
lized. Had the struts been adequately braced against buckling,

their shortening at the 630-ib load level would have been about

0.01 inches.

4. LOAD RESISTANCE PROVIDED BY THE WALL

4 . 1 General Discussion

In order to overturn, a wall must rotate over the leeward edge of

its base. In addition to the shoring, the wall itself, if

properly constructed, provides resistance to overturning, which

consists of the resisting moment provided by the weight of the

wall and the resisting moment provided by the tensile strength of

the mortar beds and the tensile resistance of the grouted dowels

connecting the wall reinforcement to the foundation, which is

small immediately after construction but increases rapidly with

time. Since it took time to construct the reinforced masonry wall

to its 23 ft-6 in. height, the grouted dowels and the mortar, had

they been constructed in accordance with plans and specifications, would have provided significant resistance to the wind

load. It is therefore important to establish the probable

conditions of the mortar and the grouted dowels at the time of

the collapse.

4 . 2 Condition of Wall

In accordance with the plans and specifications, this was to be a

partially reinforced, grouted masonry wall, whose lower 10 ft-8

in. part is of composite construction having an 8-in. concrete

block wythe and a 4-in. clay brick wythe, connected by metal

ties, and whose upper 12 ft-10 in. part consisted of 12-in. concrete block. The wall was to be reinforced vertically with #5

bars placed 2 ft O.C. in grouted cores and with 2 #5 bars each

placed in horizontal bond beams at its mid-height and at its

top. The connection between the wall and the strip footing which

25
supported it was to be accomplished by 5 ft long #5 dowels,

spaced 2 ft O.C. and embedded 2 ft-6 in. in the foundation. These

dowels extended from both ends of the wall to a distance of 4

ft~8 in. from either side of the 6x20 ft. window opening. From

these points to the edges of the window opening the plans

specified #6 dowels spaced 8 in. O.C. The plans show flashing on

the south side of the wall, to be installed between the brick and

the block wythes and to extend from the underside of the bottom

brick course to the top of the second 8-in. masonry

course. Masonry block were to be units with a 1500 psi gross-area

compressive strength, which are specially fabricated for

reinforced masonry and shaped so that the cores would line up

vertically for easier grouting. Specified grout was to be ASTM

C476 (2,500 psi) and specified mortar ASTM C270 Type S mortar.

Grout with coarse aggregate was to be used to grout large cores

with reinforcement and grout with sand only to grout narrow

cores. Grouting was to be low lift, with a lift (pour) height of

4 ft.

In accordance with the specifications masonry work was to .be

continually inspected in the field under the supervision of a

professional engineer.

Actual field evidence indicates that the wall that collapsed was

not constructed in accordance with the plans and specifications

summarized above. The following deviations are noted:

(1) The dowels were not engaged by the foundation

Available records indicate that the dowels were omitted when the

original footing was poured. To remedy this situation, the

contractor drilled holes into the foundation and grouted the #5

dowels into these holes. No #6 dowels were found at the collapse

site even though they were called for in the plans. The depth to

which the holes were drilled was 5 inches, and sometimes somewhat

26
less. Thus the dowels were embedded in the foundation a maximum

of 5 inches (the plans called for 2 ft-6 in. embedment) . However

even the 5 in. embedment was not fully effective. Apparently the

top of the foundation, as originally poured, was about 2

in. below its intended elevation. To remedy this situation, a 2

in. concrete cap was poured. As a result, the dowels were

embedded 3 inches in the foundation and 2 inches in the cap,

which was not bonded to the foundation.

The above-discussed situation is illustrated in figure 12. Figure 12(a) shows a dowel which was pulled out of the foundation

by an OSHA inspector. The part of the dowel that was embedded in

the foundation can be seen by its white coloration (the rest of

the dowel was not embedded in concrete at all). Figure 12(b)

shows the 2-in. concrete cap, figure 12(c) shows the foundation

with the cap and a solitary dowel, and figure 12(d) shows the

depth of the hole from which a dowel was pulled out. In accordance with oral statements to the author by the OSHA inspectors

who visited the site one man could extract the dowels from the

foundation with ease, by pushing them to the side in two opposite

directions and then pulling up. Minimum embedment for #5 dowels

that would permit development of their tensile strength would

have to be 24 dowel diameters or 15 inches. If the 2 inch cap is

added to this, the dowel would have to be embedded at least 17

inches to be fully effective. It should also be noted, that even

with the 3 inch actual embedment it would not have been as easy

to extract the dowels, had they been effectively grouted (it

would take about 1,100 lb to pull out the bars if the grout

achieved 1/2 its 28-day strength)

(2) The grouting of the dowels and the other vertical

reinforcement into the masonry units was not sufficient

to attain the required structural performance.

The dowel shown in figure 12(a) was not grouted into the masonry

cores. This is further illustrated in figure 13(a) which shows

27
Figure 12: Pulled Cut Dowel and Unbonded 2" Concrete Cap on

Footing of Collapsed Wall.

(a) Dowel pulled out by inspector.

(b) 2" unbonded concrete cap.

(c) Footing with concrete cap and remaining dowel.

(d) Depth of hole from which dowel was Pulled.

28
Figure 13: Grouting of Dowels and Other Masonry Reinforcement

into Masonry Cores.

(a) Failure surface with protruding ungrouted dowel.

(b) End view of collapsed wall with ungrouted

reinforcement bars.

(c) Holes opened in wall that did not collapse. Cores

were found to be ungrouted, and in some instances

lacked specified reinforcing bars.

29
dowels and other reinforcement protruding from masonry units. The

bars were rust colored, an indication that they were not embedded

in grout. This is further illustrated in figure 13(b). Figure

13(c) shows holes which were opened to inspect other masonry

walls on the same project. In many cores which were intended to

be reinforced no grout was found, and in some cores the

reinforcing bars were also missing.

The masonry units actually used in the construction were standard

units rather than the special units specified in the plans, which

are specifically designed so that the cores of the units line

up. The resulting problem is illustrated in figure 14. Figure

14(a) shows stacked special units where the cores line up. Figure

14(b) shows stacked standard units, where part of the core is

blocked by successive courses of masonry. This results in cores

which are more difficult to fill with grout in a satisfactory

manner. While present standards (6,7) permit the use of such

units, they do not recommend it. There is also no evidence that

cleanout windows were used at the bottom of the grouted cores to

permit removal of debris from the cores before they were grouted.

The flashing, which consisted of a vinyl strip, was put over the

top of the first course of masonry rather than the second

course. This resulted in a situation where a substantial portion

of the top of the first-course unit was covered by flashing, the

bottoms of the cores to be grouted were partially blocked by

flashing, and one mortar bed was laid on top of flashing thus

interrupting the tensile bond. A slot was cut in the flashing

where there were dowels. This permitted placement of the flashing

around the dowel, but blocked the rest of the core. Thus even if

some of the grout reached this level (much of it did not) , the

flashing prevented it from reaching the top of the

foundation. Figure 15 shows the flashing strip with the slot to

accomodate the dowel, pealed away from the top of the first

course of masonry.

30
Figure 14: Core Alignment in Stacked Masonry Units

(a) Stacked Ivony Type block (cores aligned)

(b) Stacked standard CMU units (cores not aligned)

31
Figure 15s Flashing Strip Peeled Away from Top of

Course of Masonry.

First

32
(3) The mortar bond was partially interrupted on top of the first

course of masonry and the effectiveness of the mortar bond at

the foundation level was impaired by the presence of an

unbonded 2 -in. concrete cap.

As previously noted, the flashing, which was placed at the

leeward side of the wall covered the top surface of the first

masonry course. This is well illustrated in figure 15. Note also

that the top side of the first course units shows no sign of any

mortar bonding. The geometry of the flashing was such that it

interrupted the mortar bond on the leeward side of the

wall. Different parts of the failure surface are shown in figures

16(a) to (d) . It appears that the combination of the effects of

the flashing and the 2 in. unbonded concrete cap on top of the

foundation was to lower the tensile resistance of the mortar bond

at the base of the wall below that normally expected even from an

unreinforced masonry wall.

4 . 3 Estimated Resisting Moment at the Base of the Wall

The resisting moment of the wall to overturning over the leeward

edge of the base of the masonry wall is attributable to two

effects: the effect of the gravity load and the combined effect

of the mortar bond and grouted dowels.

The effect of gravity loads is evaluated, assuming that the 2x4

vertical shores supporting the overhanging 12 in. wall were

removed prior to the collapse. Since the outer edge of the mortar

courses is somewhat inward from the face of the units and the

wall cannot rotate without some crushing failure of the edge over

which it rotates, it is assumed that the center of rotation is

1/4 in. inward from the outer edge of the 8-in. masonry

units. The weight of the wall attributable to the masonry units

alone was 3.8 psf for the 8 in. wall and 55 psf for the 12

in. wall. With an allowance for the grout the weight is estimated

33
Figure 16: Failure Surface? Almost No Evidence of Effective

Mortar Bonding Can Be Detected*

34
at 50 psf and 75 psf for the 8 in. and 12 in. walls,

respectively. The center of gravity of the wall is calculated to

be 2.44 inches horizontally from the leeward center of rotation

and its position could vary ±1/4 in. due to construction

tolerances. The resulting calculated resisting moment is

approximately between the limits of 17,070 ft. -lb. and 20,980

ft. -lb. (with an allowance for the weight of the lintel). The

upper bound of the calculated resisting moment is less than the

26,900 ft. -lb. estimated moment attributable to the wind load.

4 . 4 Resisting Moment of Mortar and Grouted Dowels at the Base

of the Wall

In accordance with available records, the mortar and grout at the

base of the wall at the time of collapse was more than one week

old and should have achieved at least half its design

strength. Had the wall been constructed in accordance with the

plans and specifications, it is estimated that the resistance of

the mortar bed, based on 23 psi tensile strength (the NCMA

allowable tensile strength, which is estimated to be 1/2 the

ultimate strength) and an approximate 1.5 in. width of the mortar

beds would have been approximately 14,000 ft-lb.

Since the specified embedment of the dowels in both, the

foundation and the grouted core was 2.5 ft, which is about twice

the embedment depth required for the development of their full

tensile strengths, it is reasonable to assume that at the time of

the collapse the dowels could have resisted a moment of the order

of 140,000 ft-lb which corresponds to their specified minimum

yield strength of 40,000 psi. Thus the combined resisting moment

attributable to the mortar bed and the grouted dowels would have

been on the order of 154,000 ft-lb, which is more than five times

the estimated maximum moment caused by the wind. This leads to

the conclusion that the wall collapse would have been highly

improbable, had the wall been constructed in accordance with the

35
plans and specifications. As it was, the evidence indicates that

the dowels were ineffective and the mortar bond was at best

partially effective.

5. ANOMALIES IN THE CONSTRUCTION SEQUENCE THAT CONTRIBUTED TO

THE RISK OF COLLAPSE

In addition to the deficiencies in the bracing and the as-built

condition of the wall, there were two aspects of the construction

sequence which contributed to the risk of collapse: the construction of the concrete masonry portion of a composite brick-block

wall ahead of the brick portion; and the construction of the 60

ft wide and 2 3 ft high wall in advance of the steel columns on

its ends and the abutting wall at its east end which could have

provided lateral support. Had the brick and block wythes of the

wall been built simultaneously, the base of the wall would have

increased from 7 5/8 to 11 5/8 inches and the resisting moment to

overturning provided by the gravity load would have more than

doubled and probably prevented the collapse from happening.

Likewise, had the wall been tied to the steel columns at its

ends, and the steel columns secured in accordance with the plans,

the collapse would probably not have occurred, even though the

wall was inadequately braceed.

6. PROBABLE COLLAPSE MECHANISM

The estimated moment acting on the base of the wall as a result

of the wind load was approximately 2 6,900 ft-lb. The leeward

shores, as installed, had a load capacity which produced a moment

at the base of the wall of less than 7,600 ft-lb, and would have

failed if the shores alone had to resist the wind load.

The estimated moment resistance against overturning provided by

the weight of the wall alone would have been approximately

between 17,000 and 21,000 ft-lb, had the wall been constructed to

36
accepted tolerances. The estimated moment resistance provided by

the tensile strength of the mortar at the base of the wall would

have been approximately 14,000 ft-lb, and that of the grouted

dowels approximately 140,000 ft-lb, had the wall been constructed

in accordance with plans and specifications. However, in the

as-built wall, the moment resistance of the mortar was probably

smaller than 14,000 ft-lb. and the dowels were ineffective.

The braces were very flexible and therefore contributed little to

the load resistance when the wall initially began to tilt in the

leeward direction.

Since the only sound reported by eye witnesses before the wall

impacted on the ground was that of failing timber, it is unlikely

that any significant mortar bond failure occurred during the

collapse, since such a failure tends to be associated with a

sharp, explosive noise. Thus, probably there was no effective

mortar bond at the time of the failure.

Once the center of gravity of the wall moved approximately 2.5

inches in the leeward direction due to tilting of the wall, the

resisting moment provided by the weight of the wall

disappeared. At this latter tilt displacement the estimated force

in the struts of the braces was 63 0 lb and the resisting moment

of the bracing, which was very flexible, was 7,300 ft-lb. Thus

at this tilt displacement, the resisting moment of the wall

decreased from approximately 19,000 ft-lb or more at the onset of

tilting to zero, while at the same time the resisting moment

provided by the braces increased from zero to approximately 7,300

ft-lb. This indicates that if there was a wind gust strong

enough to overcome the initial resisting moment provided by the

dead weight of the wall (and possibly the mortar bond) , the

resisting moment decreased as the wall tilted in the leeward

direction causing the wall to overturn. It should also be noted

that, as the center of gravity of the wall moved beyond the

37
leeward edge of its base, the weight of the wall itself produced

an overturning moment. Thus even if the wind gust subsided at

this point, the weight of the wall alone could have collapsed the

braces

.

The calculations presented indicate that, in the absence of

effective dowel connections, the collapse of the wall under the

prevailing wind conditions was a probable event.

7. COMPLIANCE OF BRACING WITH EXISTING PROVISIONS

The question also should be asked whether the shoring satisfied

existing standards and regulations. The relevant regulations and

standards are summarized below:

20 CFR 1926.700 (a) (Applicable OSHA regulation)

General

.

All equipment and material used in concrete construction and

masonry work shall meet the applicable requirements for design,

construction, inspection, testing, maintenance and operation as

prescribed in ANSI A 10.9-1970, Safety Requirements for Concrete

Construction and Masonry Work.

ANSI A10 . 9-197 0 , Safety Requirements for Concrete Construction

and Masonry Work:

12 c 5 Shoring and Bracing

Masonry walls shall be temporarily shored and braced until the

designed lateral strength is reached, to prevent collapse due to

wind or other forces.

ANSI A10. 9-1983

11.5 Shoring and Bracing

Masonry walls shall be shored and/or braced until the designed

lateral strength is reached, or the top supporting members are in

place to prevent collapse due to wind or other forces. The

support of bracing shall be designed by or under the supervision

of a qualified person to withstand a minimum of 15 pounds per

square foot. Braces or shores shall be secured in position.

38
ANSI A41.1 American Standard Code Requirements for Masonry

11.9 Precautions During Erection

11.9.1 Bracing to Resist Lateral Loads.

Masonry walls in locations where they may be exposed to high

winds during erection shall not be built higher than 10 times the

thickness, unless adequately braced or until provision is made

for the prompt installation of permanent bracing at the floor or

roof level immediately above the story under construction.

National Concrete Masonry Association Standard, April 1985:

4.8.1 ... Adequate precautions shall be taken to prevent damage

to walls during erection by high winds or other causes. . .

ANSI A58. 1-1982, Minimum Design Loads for Buildings and Other

Structures

6.1.1 Wind Loads During Erection and Construction Phases.

Adequate temporary bracing shall be provided to resist wind

loading on structural components and structural assemblages

during the erection and construction phases

6.4.2.

Minimum Design Wind Loading

The wind load used in the design of the main force resisting

system for buildings and other structures shall be not less than

10 lbf/ft2 multiplied by the area of the building or structure

projected in a vertical plane normal to the wind direction.

In the calculation of design wind loads for components or

cladding of buildings, the pressure difference between opposite

faces shall be taken into consideration. The combined design

pressure shall not be less than 10 lbf/ft2 acting in either

direction normal to the surface.

It can be seen that explicit provisions are provided in ANSI

A10. 9-1983, which requires bracing to resist a 15 psf horizontal

pressure and in ANSI A58. 1-1982 which requires a minimum design

pressure of 10 psf. Other standards leave the determination of

what is adequate shoring to the judgment of engineers.

Opinions by various experts on lateral pressures that temporary

shoring of masonry walls should be designed to resist were

presented at a June 17, 1986 public hearing on a draft revision

39
of 20 CFR Part 1926-"Concrete and Masonry Safety Standards",

which was presented in a "Notice of proposed rulemaking" published in the Federal Register on September 18, 1985:

Section (a) (2) of the draft revision states that :" The lateral

supports for masonry walls shall be capable to withstand a load

of 15 pounds per square foot (73 .

2

kg/m2

) applied to the

wall". The commentary notes that the 15 psf requirement is based

on the ANSI A10. 9-1983 safety requirements.

In the dicussion on the proposed revision, a professional

engineer testified on behalf of the masonry industry committee

and suggested to delete the 15 psf requirement "to avoid unnecessary economic penalties on masonry construction." He also stated

that: " Average wind speeds in selected cities seldom exceeds 10

miles per hour for construction located in cities. If construction is in the suburbs, or on flat terrain, the average speed

would increase to 10 to 15 mph. If we consider gusting effects,

the wind might reach a speed of 35 miles per hour. This, however,

would result in a horizontal pressure of approximately 5 psf, a

situation taken into account by reputable contractors."

In the same hearing, a private consultant stated that: " While it

may be true that lack of bracing rather than improper design of

the bracing is the primary cause of wall collapses, at the same

time I regard lateral loading of 15 psf as being a nominal

requirement and not one that should be difficult to meet."

Another opinion was presented by a university professor: "... it

is necessary to arrive at an acceptable lateral design load. It

is not reasonable to prescribe loads corresponding to the 50 year

recurrent wind speed applicable to permanent buildings for a wall

standing temporarily unbraced for a fraction of a year. If a one

quarter of a year long unbraced condition is acceptable, a 12.5

year recurrent wind speed would give the same annual probability

of 0.02 as used for permanent buildings".

40
In the case of Pawtucket, RI, the fastest mile wind speed 3 0

ft above ground, with a 12.5 year mean recurrence interval would

be approximately 66 mph, resulting in a design wind pressure of

approximately 12 psf.

Thus it can be seen that required design pressures in standard

provisions range from 10 psf in ANSI A58 to 15 psf in ANSI A10.9,

and that experts who testified in the June 1986 public hearing,

while not always supporting inclusion of specific design pressure

requirements in standards, suggested that design lateral pressures ranging from 5 psf to 15 psf should be used.

It is also of interest to note, that data on the effects of wind

on human activities imply that fastest mile winds with speeds

somewhere between 35 and 50 mph, measured 6 ft above ground,

would cause the work to be stopped (derived from data in

reference 2, pp 454 and 455). These wind speeds are associated

with pressures between approximately 6 and 12 psf. Winds with

lower speeds would probably not cause an interruption of work and

thus workers would be exposed to their effect.

The shoring that was actually in place could only withstand

horizontal pressures on the order of 0.5 psf, and additionally

lacked the stiffness needed to prevent initial tilting of the

wall when the wind-induced moment exceeded the resisting moment

provided by the wall itself. The shoring also did not satisfy the

provisions of the "National Design Specifications for Wood

Construction" (8) which limit the length-to-thickness ratio of

timber compression members to 50 (no load capacity could be

assigned to these struts in accepted engineering practice)

41
8 . SUMMARY

(1) It is estimated on the basis of available metereological

information that the wind load acting on the wall at the time

of the collapse produced overturning moments at the base of

the wall on the order of 26,900 ft-lb.

(2) From an examination of the evidence it is deduced that at the

time of the collapse two braces supported the leeward side of

the wall, one near the east end of the wall and one near the

west end of the wall. Each of these braces consisted of two

16-ft long 2x10 planks, one vertical and one diagonal. It is

estimated that the two braces as installed had the combined

capacity of resisting an overturning moment at the base of

the wall of less than 7,600 ft-lb. Thus if the wind load had

to be -resisted by the braces alone, the failure of the braces

under the prevailing wind conditions was a probable event.

The braces were very flexible and were inadequately supported

at the ground level, and therefore provided little support to

the wall when tilting was initiated. The diagonal struts of

the braces were installed at a steep angle to the horizontal,

creating the risk of failure of the braces, caused by upward

sliding of the restraining vertical shoring plank.

(3) The collapse occurred as a result of the combined effect of

inadequate bracing, ineffective dowel connections to the

foundation, reduced effectiveness of mortar bonding near the

base of the wall resulting from vinyl flashing which covered

part of the top surface of the bottom course of masonry units

and an unbonded 2 in. concrete cap on the foundation, and

anomalies in the construction sequence which increased the

risk of collapse. The collapse was a probable event under the

prevailing wind conditions.

42
(4) Existing regulations and standards require bracing of masonry

walls during construction against the effects of high

winds. Most of the standards leave the decision of what is

adequate bracing to engineers, however ANSI A10. 9-1983

requires bracing to be designed for uniform lateral pressures

of 15 psf and ANSI A58. 1-1982 stipulates minimum lateral

pressures of 10 psf. Experts testifying in a June 1986 public

hearing on a draft revision of the applicable OSHA

regulations, while not necessarily favoring inclusion of

explicit values for design lateral pressures in the

regulations, suggested design lateral pressures ranging from

5 to 15 psf. Work at construction sites would probably

proceed at wind speeds causing lateral pressures of less than

6 psf.

It is deduced from available information, that the bracing

that supported the leeward side of the wall at the time of

the collapse had the capacity to support an ultimate

horizontal pressure of not more than 0.5 psf, acting normal

to the surface of the wall.

9 . REFERENCES

(1) ANSI Standard A58. 1-1982 , "Minimum Design Loads for

Buildings and Other Structures", American National

Standards Institute, New York, NY, 1982.

(2) Simiu, E., Scanlan, R.H.

Wind Effects on Structures

Second

Edition, John Wiley & Sons, New York, NY, 1986.

(3) Amrhein, J . E ..

Reinforced Masonry Engineering Handbook".

4th. Ed. , Masonry Institute of America, Los Angeles, CA, 1983.

(4) National Concrete Masonry Association, NCMA-TEK 72, "Bracing

Concrete Masonry Walls During Construction", Herrndon, VA,

1975.

(5) Carr, R.I. et al., "Guidelines for Temporary Masonry Wall

Support Systems", Masonry, Research, Application and

Problems, American Society for Testing and Materials,

Philadelphia, PA, April, 1985.

43
(6) National Concrete Masonry Association (NCMA),

"Specifications for the Design and Construction of Load

Bearing Concrete Masonry" , Herndon, Va, 1985.

(7) American Concrete Institute, ACI 531-79, and 531R-79,

"Building Code Requirements for Concrete Masonry

Structures", (revised 1983), Detroit, MI, 1983.

(8) National Forest Products Association, "Design Values for

Wood", National Design Specifications for Wood Construction,

Washington, DC, 1986.
NBS-114A (REV. 2-80

U.S. DEPT. OF COMM.

BIBLIOGRAPHIC DATA

SHEET (See instructions)

1. PUBLICATION OR

REPORT NO.

NBSIR 88-3820

2. Performing Organ. Report No. 3. Publication Date

JULY 1988

4. TITLE AND SUBTITLE

COLLAPSE OF MASONRY WALL UNDER CONSTRUCTION IN PAWTUCKET, RI, OCTOBER 28, 1985

5. AUTHOR(S)

Felix Y. Yokel

6. PERFORMING ORGANIZATION (If joint or other than NBS. see instructions) 7. Contract/Grant No.

NATIONAL BUREAU OF STANDARDS

8. Type of Report & Period Covered

U.S. DEPARTMENT OF COMMERCE

GAITHERSBURG, MD 20899

9. SPONSORING ORGANIZATION NAME AND COMPLETE ADDRESS (Street. City. State, ZIP)

Occupational Safety & Health Administration

Department of Labor

Government Center

Boston, MA 02203

10. SUPPLEMENTARY NOTES

[ Document describes a computer program; SF-185, FIPS Software Summary, is attached.

11. ABSTRACT (A 200-word or less factual summary of most significant information. I f document includes a significant

bibliography or literature survey, mention it here)

Results from a study to determine the cause of the October 28, 1985 collapse of a masonry wall under construction in Pawtucket, RI are presented. The wall was a 60 ft3 in. long, 23 ft-6 in. high partially reinforced concrete masonry wall supported by

wooden braces. The study included: inspection of construction plans and specifications; review of construction records and eyewitness accounts recorded immediately

after the collapse, as well as testimony from OSHA inspectors and local building

officials who visited the site a short time after the collapse; examination of photographs taken by OSHA inspectors and police investigators; analysis of meteorological

data; and a stability analysis of the collapsed wall. It is concluded that the

collapse was probably caused by a gust of wind which exerted lateral forces which

exceeded the lateral-load capacity of the wall and its supporting wooden braces.

Contributing factors were the lack of grout in the masonry cores which contained the

steel reinforcement dowels and the inadequate anchoring of the dowels in the

foundation.

12. KEY WORDS (Six to twelve entries ; alphabetical order; capitalize only proper names; and separate key words by semicolon s) building; construction loads; construction safety; lateral bracing; masonry

construction; stability; structural collapse; timber shoring; wind loads.

13. AVAILABILITY 14. NO. OF

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