A111D2 7 7 4 E

NBS

PUBLICATIONS

' U—iLkUlM NBSIR 87- 3640

Investigation of

L' Ambiance Plaza Building Collapse

in Bridgeport, Connecticut

Center for Building Technology

National Engineering Laboratory

National Bureau of Standards

Gaithersburg, MD 20899

-QC

TMENT OF COMMERCE

• U56 AU OF STANDARDS

#87-3640

1987

C.2

Research Information Center

National Bureau of Standards

Gaithersburg, Maryland 20899

NBSIR 87-3640

Investigation of L' Ambiance Plaza

Building Collapse in Bridgeport,

Connecticut

N&ScL CiCfoo

/%

Charles G. Culver

Charles F. Scribner

Richard D. Marshall

Felix Y. Yokel

John L. Gross

Charles W. Yancey

Erik M. Hendrickson

Center for Building Technology

National Engineering Laboratory

National Bureau of Standards

Gaithersburg, MD 20899

Sponsored by:

Occupational Safety and Health Administration

U.S. Department of Labor

Washington, DC 20001

Issued September 1987

U.S. DEPARTMENT OF COMMERCE, Clarence J. Brown, Acting Secretary

NATIONAL BUREAU OF STANDARDS, Ernest Ambler, Director
'
ABSTRACT

Results from an investigation to determine the cause of the collapse of the

L' Ambiance Plaza building on April 23, 1987 are presented. The building was being constructed using the lift-slab method; collapse occurred during

construction. The investigation included on-site inspections immediately

following the collapse, review of eyewitness accounts of the collapse, review

of project documentation, laboratory and field tests and analyses of the

structure. Several potential failure mechanisms were investigated. The most

probable cause of the collapse was determined to be loss of support at a lifting jack in the west tower during placement of an upper level package of

three floor slabs. The loss of support was likely due to excessive deformation

of the lifting angle in a shearhead followed by a lifting nut slipping off the

lifting angle of the shearhead. The postulated failure mechanism was duplicated

in laboratory experiments. The local failure propagated as loads were

redistributed. The remaining jack rods along column line E supporting the

package of floor slabs slipped off the lifting angles and the slabs failed in

flexure and shear. These slabs fell causing the lower level slabs to fail.

Keywords: building; collapse; concrete; construction failure; lift- slab

construction; post- tensioned concrete; progressive collapse.

iii
EXECUTIVE SUMMARY

On April 23, 1987, the L' Ambiance Plaza building under construction in

Bridgeport, Connecticut, collapsed killing 28 construction workers. This was

the largest loss of life in a U.S. construction accident since 51 workmen were

killed in the collapse of a reinforced concrete cooling tower under construction

at Willow Island, West Virginia, in 1978.

On April 24, 1987, the Occupational Safety and Health Administration, Department

of Labor, requested that the National Bureau of Standards (NBS) carry out an

investigation of the failure to determine the most probable cause of the

collapse. NBS conducted an investigation under the terms of an Interagency

Agreement established between OSHA and NBS in 1973. Under this agreement, NBS

conducts scientific and technical investigations in connection with failures

under the jurisdiction of OSHA.

The NBS investigation commenced immediately upon arrival of NBS personnel on

the site on April 24, 1987. Data were collected and assessments made of the

nature of the failure of various elements of the structure during and subsequent

to the rescue efforts. In addition to observations from the on-site inspection,

the NBS team in carrying out its investigation used:

(1) information on the collapse obtained from interviews of survivors and

eyewitnesses

(2) project documentation including design specifications, plans, shop

drawings, construction records, testing laboratory reports and project

correspondence

( 3 ) laboratory tests of samples removed from the collapsed structure

( 4 ) data obtained from a subsurface investigation at the site after the

collapse , and

( 5 ) analytical studies, including computer analyses.

L' Ambiance Plaza was being constructed using the lift- slab method. The floor

and roof slabs were cast one on top of the other at ground level. The floor

slabs were two-way post- tensioned flat plates. After post- tensioning , the

slabs were lifted by hydraulic jacks and secured to the columns. The building

was to be a 13 -story structure with three levels of a five level parking

garage located under the building. It consisted of two offset rectangular

towers, designated in this report as the east tower and the west tower. At

the time of the collapse, erection of the slabs was over half complete. The

collapse occurred during placement of wedges under a package of three slabs

being parked in a temporary position during erection of the building. In the

collapse, all the floor slabs fell trapping the workmen involved in the

lifting operation and those on the lower floors engaged in other phases of the

construction.

A number of possible mechanisms which may have initiated the collapse were

considered in carrying out the investigation. These included: (1) lateral

instability (2) individual column instability, (3) floor slab failure, (4) weld

failure, (5) foundation failure, (6) failure due to lateral soil pressure and

( 7 ) loss of support of floor slab. Field observations, eyewitness accounts,

iv
laboratory tests and analytical studies were used to assess the likelihood of

each failure mechanism.

Based on the results of the investigation, NBS concludes that:

1. The most probable cause of the collapse was failure of the lifting

system in the west tower during placement of a package of three upper

level floor slabs. The failure most probably began below the most heavily

loaded jack (column E4.8) or an adjacent jack (column E3.8). Excessive

deformations occurred in the lifting angle of the shearhead at the

location of the initial failure. This was followed by one of the jack

rods in the lifting assembly slipping off the lifting angle in the

shearhead supporting the package of three slabs . This failure

mechanism was duplicated in laboratory experiments. The local failure

propagated as loads were redistributed and the remaining jack rods

along column line E slipped off the lifting angles and the package of

three slabs failed in flexure and shear. These slabs fell, causing

the lower level slabs to fail. This resulted in the collapse of the

entire west tower. Consequently, the east tower collapsed due to one or more of the following factors: (a) forces transmitted to it by the

west tower collapse, (b) damage to the post- tensioning tendons caused

by falling debris from the west tower, or (c) lateral instability

caused by falling debris from the west tower.

2. The quality of materials in the structure was generally in accordance

with the project plans and specifications and did not play a significant

role in initiating the collapse.

3. There were a number of deviations from the project plans and specifications

in the structure as built, but these did not play a significant role

in initiating the collapse.

4. It is unlikely that the horizontal jack used to plumb the structure

initiated the collapse.

5. The reserve capacity against lateral instability was small. It does not

appear, however, that lateral instability was the initial cause of the

collapse. Inadequate resistance to lateral instability may have

caused the collapse of the east tower.

6. It is unlikely that lateral earth pressure acting against the basement

wall on the north side of the structure caused the building to collapse.

7 . It is unlikely that differential foundation settlements caused the

building to collapse.

v
ABSTRACT iii

EXECUTIVE SUMMARY iv

1. INTRODUCTION 1

1.1 BACKGROUND 1

1.2 OBJECTIVE AND SCOPE OF THE INVESTIGATION 1

1.3 ORGANIZATION OF THE REPORT 2

2. CONDUCT OF THE INVESTIGATION 5

2.1 SITE INVESTIGATION 5

2.2 LABORATORY TESTS 5

2.3 WITNESS INTERVIEWS 6

2.4 REVIEW OF DOCUMENTS 6

2.5 SUBSURFACE INVESTIGATIONS 6

2.6 ANALYTICAL STUDIES 6

3. DESCRIPTION OF THE STRUCTURE AND CONSTRUCTION 8

3 . 1 INTRODUCTION 8

3.2 DESCRIPTION OF THE STRUCTURE 8

3.3 LIFT SLAB CONSTRUCTION DETAILS 9

3.4 CONSTRUCTION SCHEDULE 12

4. DESCRIPTION OF COLLAPSE 26

4.1 INTRODUCTION . 26

4.2 ACTIVITIES PRECEDING COLLAPSE 26

4.3 EYEWITNESS ACCOUNTS OF COLLAPSE 27

4.4 OBSERVATIONS OF DEBRIS 30

4.4.1 Overall Collapse ..... 30

4.4.2 Floor Slabs ... ......... 32

4.4.3 Shearheads . 33

4.4.4 Welds 34

4.4.5 Additional Observations 35

4.4.6 Detailed Observations of Stage IV Columns -West Tower ... 35

4.4.7 Summary 37

5. LABORATORY TESTS 81

5.1 INTRODUCTION 81

5.2 TESTS OF CONCRETE SPECIMENS 81

5.2.1 Sampling and Testing Procedures SI

5.2.2 Results of Tests 82

5.2.3 Comparison of Strengths of Core Samples and of Quality

Control Samples 82

5.3 TESTS OF STEEL SPECIMENS 83

5.3.1 Sampling and Testing Procedure 83

5.3.2 Results of Tests 84

5.3.3 Comparison of Results with ASTM Standards and Mill Tests . 85

5.4 TESTS OF WELDMENTS 85

5.4.1 Sampling and Testing Procedure 85

5.4.2 Results of Strength Tests 87

5.4.3 Metallography Results 88

5.4.4 Fractography Results ........ ..... 89

5.5 AUXILIARY TESTS 91

5.5.1 Introduction 91

5.5.2 Tests of Post- tensioning Strands 91

5. 5.

2.1

Sampling and Testing Procedure 91

vi
5. 5. 2. 2 Results of Tests 92

5. 5. 2.

3

Comparison of Results with ASTM A 416 93

5.5.3. Tests of Jack Rods 93

5. 5. 3.1 Sampling and Testing Procedure 93

5. 5. 3.

2

Results of Tests 94

5. 5. 3.

3

Metallography Results 94

5.5.4 Tests of Shearheads 95

5.5.5 Lifting Assembly 97

5.6 SUMMARY 102

6. SUBSURFACE EXPLORATION 147

6.1 INTRODUCTION 147

6.2 EXPLORATION PROCEDURES 148

6.3 EXPLORATION RESULTS 149

7 . STRUCTURAL ANALYSIS 154

7.1 INTRODUCTION 154

7.2 FLOOR SLABS ..... 154

7.2.1 Computer Model 154

7.2.2 Results of Analyses 155

7.3 COLUMN INSTABILITY 159

7.4 LATERAL LOAD ANALYSIS 160

7.4.1 Introduction 160

7.4.2 Frame Stability . 160

7.4.3 Horizontal Jacking Load 164

7.4.4 Earth Pressure . 165

7.4.5 Wind Loading 170

7.5 DIFFERENTIAL FOUNDATION SETTLEMENTS ........ 170

8. EVALUATION OF CONSTRUCTION PRACTICES . 201

8.1 INTRODUCTION 201

8.2 COMPARISON WITH PROJECT SPECIFICATIONS 201

8.2.1 Construction Materials . 201

8.2.2 Foundations 201

8.2.3 Column Sections . 202

8.2.4 Lateral Bracing 202

8.2.5 Welding 203

8.2.6 Post -Tensioning Tendon Detail 204

9. ANALYSIS OF COLLAPSE 206

9.1 INTRODUCTION 206

9.2 REVIEW OF POTENTIAL FAILURE MECHANISMS 206

9.3 PROBABLE CAUSE OF COLLAPSE 209

9.4 PROBABLE SEQUENCE OF COLLAPSE ..... 211

10. CONCLUSIONS 222

11. ACKNOWLEDGEMENTS 225

12. REFERENCES .226

APPENDICES

Appendix A Detailed Observations of Stage IV Columns

West Tower . 228

Appendix B Geotechnical Engineering Investigation 236

Appendix C Resistance Criteria for Columns and Floor Slabs . . 307

vii
--

.

-
1. INTRODUCTION

1.1 BACKGROUND

At approximately 1:30 p.m. on April 23, 1987, the L'Ambiance Plaza, a residential

apartment building under construction in Bridgeport, Connecticut collapsed,

killing 28 construction workers. This was the largest loss of life in a U.S.

construction accident since 51 workmen were killed in the collapse of a reinforced concrete cooling tower under construction at Willow Island, West

Virginia in 1978 [1]

.

L'Ambiance Plaza was being constructed using the lift-slab method. In liftslab construction, floor and roof slabs are cast one on top of the other at

ground level. The floors are usually two-way post- tensioned flat plates of

either regular or lightweight concrete. After post- tensioning, the slabs are lifted to their final positions by hydraulic jacks and are secured to the

columns. By casting the slabs at ground level, lift-slab construction can

eliminate 90 percent of the formwork required for cast- in-place construction

and reduces labor requirements [15] . Cost savings and speed of construction are two primary advantages claimed for lift-slab construction [3].

L'Ambiance Plaza was to be a 13 -story building with three levels of a five-level

parking garage located under the building. A sketch of the building is shown

in figure 1.1.1. At the time of the collapse, erection of the slabs was over

half complete; i.e., three levels of the parking garage and three to six

levels of the towers were in place. The collapse occurred while a group of

three slabs was being placed in temporary position. In the collapse, all the

floor slabs fell, trapping the workmen involved in the lifting operation and

those on the lower floors engaged in other phases of the construction.

An investigating team from the Occupational Safety and Health Administration

(OSHA) was on site shortly after the collapse. On April 24, 1987, OSHA

requested technical assistance from the National Bureau of Standards (NBS) in

investigating the collapse. A team of engineers from NBS arrived on the site

of the collapse at 6:00 p.m. that same day.

1.2 OBJECTIVE AND SCOPE OF THE INVESTIGATION

NBS carried out the investigation for the Occupational Safety and Health

Administration under the terms of an Interagency Agreement established between

OSHA and NBS in 1973. Under this agreement, NBS conducts scientific and

technical investigations in connection with failures under the jurisdiction of

OSHA. The objective of the NBS investigation was to determine the most

probable physical cause of the collapse. The study did not include evaluation

of the design of the completed structure.

The NBS investigation commenced immediately upon arrival of the team on the site

on April 24, 1987. Data were collected and assessments made of the nature of

the failure of various elements of the structure while the rescue efforts

proceeded. Special efforts were made to ensure that the investigation did not

interfere with these rescue efforts. In addition to the on-site inspection,

the NBS team in carrying out its investigation used: (1) information on the

1
collapse obtained from interviews of survivors and eyewitnesses, (2) project

documentation including design specifications, plans, shop drawings, construction

records, testing laboratory reports and project correspondence, (3) laboratory

tests of samples removed from the collapsed structure, (4) data obtained from

a subsurface investigation at the site after the collapse, and (5) analytical

studies

.

1.3 ORGANIZATION OF THE REPORT

This report is organized in 12 Chapters and Appendices:

Chapter 2 describes the procedure followed by the NBS team in conducting the

investigation. The procedure included on-site investigations, laboratory

tests, witness interviews, review of project documentation, subsurface

investigations and analytical studies.

Chapter 3 describes the general layout of the structure, the structural

elements involved in lift- slab construction, and the status of construction at

the time of the collapse. A description of the jacking system used to erect

the building and its operation are included.

Chapter 4 summarizes construction activities preceding the collapse. The

configuration of the structure at the time of the collapse and eyewitness

accounts of the sequence of events associated with the collapse are presented.

Detailed observations of the debris and the performance of the structural

components in the collapse are summarized.

Chapter 5 describes the NBS laboratory testing program and the test results.

The program included tests of the construction materials and tests of structural

components and subassemblies

.

Chapter 6 describes the procedures used and results obtained from subsurface

explorations conducted to explore the in- situ conditions with respect to

support of the footings and backfill conditions behind a basement wall.

Chapter 7 presents the results of analytical studies conducted to evaluate the

loads on the structure during erection. Deformations and stresses in the

floor slabs and support reactions for the lifting jacks and columns were

determined. Column stability, frame stability, effects produced by a horizontal

jack used to plumb the building, effects due to earth pressure on a basement

wall and wind loads were analyzed.

Chapter 8 presents a comparison of the results of the laboratory tests and

field tests with the project specifications.

Chapter 9 addresses the collapse of the structure. A number of possible

failure scenarios are reviewed. The probable cause of the collapse is

identified and the sequence of events in the collapse is presented.

Chapter 10 presents the conclusions reached by the NBS investigation team.

2
Chapter 11 includes acknowledgements of individuals providing assistance in

the investigation.

Chapter 12 lists the references cited in the text.

The Appendices present material used in conducting the investigation. Detailed

observations of the condition of the columns in the upper portion of the west

tower, data collected in the subsurface exploration and resistance criteria

used in analyzing the performance of the columns and floor slabs are included.

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2. CONDUCT OF THE INVESTIGATION

2.1 SITE INVESTIGATION

The initial on-site phase of the investigation was conducted from April 24

through May 1, 1987. During this period data were collected on the collapsed

structure. Photographs and video tape records were made as the rescue efforts

proceeded and as debris was removed from the site. Sketches were made of the

pattern of deformation of the columns and the orientation of the collapsed floor

slabs. Detailed measurements were also made of several columns as they were

removed from the debris. Several floor slabs at different levels in the

building were identified as they were removed from the debris in order that

core samples could be taken later.

The debris removed from the site was transported by the City of Bridgeport to

a landfill at Seaside Park in south Bridgeport. Samples of the construction

materials were collected on April 29-30 and transported to the National Bureau

of Standards on May 1 for testing. The materials included: (1) concrete core samples from the floor slabs and shear walls (taken at both the landfill and

at the building site) , (2) portions of columns and portions of columns

containing splices and weld blocks* (3) post- tensioning strand from the floor

slabs and unused strand at the site, (4) one large and one small hydraulic

jack, jack rods, nuts and end fittings, (5) several shearheads and wedges and

(6) a large portion of a floor slab with the shearhead and post- tensioning

strand intact.

Subsequent visits were made to the building site and the landfill during the

months of May, June, July and August to collect additional data and material

for testing. Elevations at the column locations were measured after the

debris was removed to determine whether settlement of the footings had

occurred. Additional measurements were made of the columns and welding

details, including both shop and field welds. Samples of several weld

fractures were also taken for laboratory analysis.

2.2 LABORATORY TESTS

Two types of laboratory tests were conducted. The first included standard

tests for evaluating material properties and welding details. Compression

tests and splitting tensile tests were conducted on the concrete cores. Tensile

tests, metallographic and chemical analyses, hardness tests and fracture

analyses were conducted on the column steel and weldments. Tensile tests were

also conducted on the post- tensioning strand.

The second type of test involved individual components and assemblies of the

lifting system. Tensile tests were carried out on the jack rods and attachments.

Load tests were conducted on the shearhead- column assembly and the lifting assembly

(jack, jack rods and attachments, shearhead).

* Terminology used in this report follows that used in the L' Ambiance

Plaza project documentation

5
2.3

WITNESS INTERVIEWS

Witness statements obtained by the Bridgeport City Police in the first few

days following the collapse and by the Connecticut State Police and OSHA staff

over the next several weeks were reviewed. Witnesses included construction

workers on the site at the time of the collapse, eyewitnesses in the vicinity

of the site at the time of the collapse and other parties with information related

to the construction. A total of 48 statements were reviewed. OSHA arranged

for the NBS staff to reinterview a number of the witnesses to clarify points

made in their original statements and to obtain additional information not

included in the initial interviews

.

2.4 REVIEW OF DOCUMENTS

Following the collapse, the State of Connecticut seized a variety of documents

from the construction site. Copies of these documents were provided to NBS by

the Connecticut Department of Public Safety. This material included: (1) the

daily construction logs, (2) testing laboratory reports, (3) project correspondence

and (4) design and construction drawings. Copies of the project specifications

and the architectural, structural and mechanical-electrical drawings were

obtained from the City of Bridgeport. Wind speed data and temperature records

were obtained from the weather station at the Bridgeport airport. Additional

material obtained by OSHA as the investigation proceeded included pre-collapse

photographs of the structure, mill test reports and the construction log of

the lifting subcontractor.

2.5 SUBSURFACE INVESTIGATIONS

Subsurface investigations were conducted at the building site between May 27

and June 17 after the majority of the debris had been removed. The objective

of these investigations was to evaluate the in- situ condition of the foundations

and the backfill behind the retaining wall on the north side of the building.

Soil borings were taken in the fill material behind the retaining wall and

core borings were taken at a select number of column footings. The core

borings penetrated directly through the column footings, the underlying soil,

weathered or fractured rock, and into underlying bedrock deposits. Test pits

were dug adjacent to a few footings in order to permit visual inspection of

the footings and the supporting soil or rock. One test pit was dug behind the

retaining wall to retrieve soil samples for testing. In-situ tests included

standard penetration tests and pressuremeter tests. Laboratory studies

included routine classification tests and direct shear tests on re-constituted

soil samples.

2.6 ANALYTICAL STUDIES

The structure was analyzed for loadings encountered during erection. Loadings

induced by a number of possible conditions that could have precipitated

collapse of the structure were also studied.

Structural analyses were performed to determine deformations, stresses and

support reactions for the floor slabs. Stability of the individual columns was analyzed. A two-dimensional analysis of the structural framing system and a

6
simplified column model were used to analyze the lateral stability of the

structure. Effects produced by a hydraulic jack used to plumb the building

during erection were also determined. Lateral displacements, internal forces,

and the ability of the structure to resist these lateral forces resulting from

earth pressure on a basement wall were evaluated. The effect of wind loads

and differential foundations settlements on the performance of the structure

were also considered.
3. DESCRIPTION OF THE STRUCTURE AND CONSTRUCTION

3 . 1 INTRODUCTION

This chapter describes the structure and summarizes the construction schedule

for erection of the structural system. A description of the jacking system

and its operation during lifting of the slabs is included.

3 . 2 DESCRIPTION OF THE STRUCTURE

A plan view showing the column layout and shearwall locations is given in

figure 3.2.1. The structural system consisted of steel columns (W and HP

shapes) and two-way unbonded post- tensioned concrete flat plates with shearwalls

at four perimeter and four interior locations. The building consisted of two

offset rectangular towers, each 112 ft (34 m) by 62 ft (19 m) in plan. These

towers will be referred to in this report as the east tower and the west

tower

.

An elevation view of the west tower is shown in figure 3.2.2. The perimeter

shearwalls are not shown. The status of construction and position of the

floor slabs at the time of collapse are discussed in Chapter 4. The floor

level designations in figure 3.2.2 are those used in the plans and will be

used throughout this report. Levels C, D and E are the parking garage. The

story heights were the same throughout the height of the building except at

level C and at ground level. The column schedule is given in table 3.2.1.

Each tower of the building was erected independently. The slabs of the two

towers were connected by cast- in-place reinforced concrete pour strips in the

center portion of the building. The pour strips were cast after the corresponding

lift slabs in the two towers of the building were secured in their final position.

The three levels of parking garage were below ground level on the north side of

the building. A basement wall extended from level E to level C. This wall

was in contact with the floor slab at these levels and transmitted lateral soil

pressure to the structure. The interior and perimeter shearwalls terminated

at level 11. The shearwalls were connected to the floor slabs by reinforcing

bars which protruded from the precast lift slabs and were embedded into the

cast- in-place shearwalls.

The floor slabs were 7-in (178 mm) thick two-way unbonded post- tensioned flat

plates. Regular weight concrete was used throughout. The location of the

post- tensioning tendons is shown schematically in figure 3.2.3. Pipe chase

openings shown are for a typical floor. At each line in the figure, there

were a number of tendons. The tendons in the north-south direction were

approximately uniformly spaced over the length of the slab. In the east-west

direction, the tendons were banded generally following the column lines. Due

to the presence of the elevator shaft, the tendons along column line E in the

west tower .*ere splayed as shown in figure 3.2.3. It should be noted that the

centroids of the banded tendons, particularly those at the exterior column

lines, do not coincide with the column centerlines. A note on the tendon

layout drawing indicated the stressing end and dead end were reversed on some of the transverse tendons (north- south direction) from that shown in figure

8
3.2.3. This would not be expected to have any influence on the performance of

the floor slabs. A nominal amount of bonded reinforcement was used in the

vicinity of the columns and shearwalls

.

The building was supported by spread footings which, in accordance with the

plans, were to rest on the underlying bedrock. A plan view of the footings

showing the footing size and elevation (referenced to the ground floor

elevation) is given in figure 3.2.4. Figure 3.2.5 shows a typical detail of an interior column footing.

The footings were designed for a 7 ton/ft^ (670 kPa) bearing pressure and

varied in size in accordance with the supported load. Along the basement

walls on the north and east side of the building the columns and walls were

supported by combined footings. The shearwalls were supported by combined

rectangular footings, which also supported adjacent columns and basement

walls, and in one case an adjacent elevator shaft.

3 . 3 LIFT SLAB CONSTRUCTION DETAILS

The lift-slab construction process used at L'Ambiance Plaza consisted of

casting floor slabs at ground level, raising those floor slabs to the desired

elevation using hydraulic jacks, and fixing the floor slabs in position

mechanically. The casting of floor slabs one on top of another on a slab on grade was straightforward. This procedure eliminated the need for shoring or other formwork underneath each slab, and required only that a side form for

the slabs be constructed.

Floor slabs were raised to the desired elevation using a system of hydraulic

jacks, threaded jack rods and attachments, and welded steel collars called

shearheads which were cast into the concrete floor slab at each column. A

sketch of a typical jack with attachments is shown in figure 3.3.1. The

details of a typical shearhead are shown in figure 3.3.2. Each jack consisted

of an upper and lower crossarm separated by a hydraulic cylinder. The

hydraulic cylinder and lower crossarm were an integral unit which sat directly

on the column top. The upper arm supported two jack rods which were attached

in turn to the floor slabs through the shearhead.

In the lifting operation, a hydraulic jack was mounted on the top of each

column (24 columns in the east tower and 25 columns in the west tower) . A

separate power unit and control console was provided for each tower and the

lifting and positioning of the floor slabs in each tower consisted of independent

operations. The nominal load capacity of each jack was 150 kips (667 kN) , but

for the four columns with the heaviest sections in each tower, "super jacks"

with a capacity of 300 kips (1334 kN) were used. These larger jacks were

fitted with a bearing plate and were installed on columns B9 , BIO, D9 and DIO

in the east tower and on columns E3 , E3 . 8

,

G3 and G4 in the west tower.

Although a 150 kip (667 kN) jack is described below, the essential features

and mode of operation apply equally to the 300 kip (1334 kN) jacks. Each jack

was connected to its console with two hoses, one for extending the jack and

one for retracting the jack. All jacks connected to a given console operated

at the same line pressure.

9
Lifting of floors was accomplished by raising the jacks and jack rods in 1/2-

in (12.7 mm) increments. At the start of each increment, or lifting cycle, the

upper crossarm of the jack was forced upward 5/8 in (15.9 mm), the maximum stroke

of the hydraulic piston. As each jack was being raised, a hydraulic actuator

(not shown in figure 3.3.1) and chain drive rotated a holding nut downward on each jack rod to keep the nut in contact with the bottom crossarm. This

provided a positive mechanical support of the jack rods to prevent the slabs from

dropping in the event of a hydraulic system malfunction. When the jack

reached its maximum travel and the bottom holding nuts had been turned snug

against the lower crossarm, the piston was retracted. The hydraulic actuator

and a second set of chain drives rotated an upper set of take-up nuts down the

jack rods to keep them in contact with the upper crossarm.

When the hydraulic piston had been fully retracted and the take-up nuts turned

snug against the upper crossarm, the lifting cycle was complete. Although the

stroke of the piston was 5/8 in, 1/8 in of travel was lost in each cycle

because of slack between the nuts and crossarms. As a result, the floor slab

was lifted only 1/2 in during each cycle of lifting. A system of electrical

interlocks prevented the jacks from being cycled if either the holding or take-up nuts had not followed and maintained contact with the crossarms. This

system prevented the floor slabs from being raised more than 1/2 in differentially

between any two jacks. The cycle of lifting was repeated until the slabs were

close to the desired elevation for parking. When operating in the automatic mode,

lifting rates of approximately 5 ft (1.52 m) per hour were normal.

Once the slabs had been placed approximately at the desired elevation for

parking (mechanical attachment to the columns), the jacks could be valved for

local control so that the shearhead and floor slab at each column location

could be raised or lowered to exactly the desired elevation. The floor slabs

were then mechanically attached, either permanently or temporarily, to the columns.

Slabs were attached to the columns through the shearheads by blocks welded to

the external faces of the column flanges and by steel wedges placed between

these blocks and the lower face of the shearheads . An isometric view of a typical permanent connection of a slab to a column and the nomenclature of the

components of the connection are shown in figure 3.3.3. The seal block shown

in figure 3.3.3 was present only at those locations where the slab was to be

attached permanently to the column.

The parking of a slab at a desired elevation was done by three workers who

moved as a team from column to column and fixed the slab at the desired

elevation. At each column, one worker observed the elevation of the lower surface

of the slab from beneath the slab. A second worker, located at the lifting

jack, responded to signals from the first worker, and raised or lowered the

jack until the floor was in the desired position. At that time, the first

worker and a helper would place a wedge between the weld block and the bottom

face of the shearhead on each side of the column. After the load was transferred

to the wedges, a welder fixed each wedge in place temporarily with two tack

welds, one between the wedge and the column at each end of the wedge.

When a pair of wedges at a column had been inserted and tack welded, the jack

rods were lowered for attachment to another package of slabs and the jacking

10
process was repeated. To facilitate lowering of the jack rods and to make

other operations easy, split- or swing-nuts were used to transfer loads from

the jack crossarms to the jack rods. Under load, the nuts were enclosed in

chain-driven sockets. Raising these sockets released the nuts, allowing the

jack rods to be moved up or down quickly. The nominal jack rod length is 28

ft (8.53 m) and for the lower packages of slabs it was necessary to add

extension rods as shown in figure 3.3.1.

Slabs could be parked either temporarily or permanently, depending on the

schedule of lifting. If the slab was to remain in its parked position only

temporarily, wedges would be held in position only by tack welds and by

friction. However, if the slab was to be parked permanently, the connection

between the slab and column was strengthened. In the latter case, fillet

welds were applied between the wedges and all adjacent surfaces of the

columns, weld blocks, and shearheads . The top faces of the shearhead were

welded to the seal blocks to further stiffen the connection, and concrete was placed in the cavity (referred to on the project as a "beam pocket" despite

the fact that no beam was present) between the shearhead and column.

Columns and floor slabs were erected in stages. The first step in the

construction of the east and west towers consisted of setting the stage I columns on their foundations. Shearheads were then lowered onto the column

sections in preparation for the casting of floor slabs.

Slabs were cast one on top of the other on the slab on grade (level E) , each

subsequent slab using the previously-cast slab as its bottom form. After they

had been finished, the top surfaces of the slabs were sprayed with a bond

breaker to prevent them from bonding to the following slab.

After the slabs had been cast and had reached the appropriate strength, they

were post- tensioned. The column sections for stage IV and above were stacked

on the top slab along with the power units, operating console and gantry.

After jacks had been placed on top of the stage I columns, lifting of slabs

was begun. The sequence of lifting required that up to three slabs be lifted

at one time. A group of two or three slabs that was to be lifted at one time

was referred to as a "package" of slabs. This report will refer to these

groups of slabs by the levels of the floor slab separated by virgules . For

example, the group of slabs that consisted of slabs for levels 9, 10, and 11

will be referred to as "9/10/11." When more than one slab was lifted at once,

jack rods and lifting nuts were attached to the bottom slab in the package.

The upper slabs in a package of slabs were supported only by the lowest slab

in the package and were not directly attached to the lifting rods.

Erection of the floor slabs began with the lifting of the package 12/roof in

the east tower on 02/10/87. With the exception of the first and last package

of two floor slabs in each tower, all packages consisted of three floor slabs.

The column height for stage I was 31' -2" (9.5 m)

.

Subsequent column extensions

of 15' -3" (4.65 m) , 17 '-4" (5.28 m) and 19' -9" (6.02 m) were added to reach

stage IV at the time of the collapse. Because of their weight, the stage II

and stage III column extensions were placed using a truck crane. The sequence

of lifting floor slabs and extending columns is shown in figure 3.3.4.

11
After the lifting was completed for a stage, the column extensions for the

next stage were installed. To accomplish this, one jack was removed from a column top and placed horizontally on the top slab using the service gantry.

A column extension was then positioned with the gantry and welded in place.

The jack on an adjacent column was then removed and placed on the extended

column and the process continued until all columns were extended and all jacks

were in place to commence another lift.

3.4 CONSTRUCTION SCHEDULE

The schedule followed by the contractor in building the structural system is

summarized in table 3.4.1. The dates and descriptions of activities listed in

the table were extracted from the general contractor's daily log. In certain

instances this information was supplemented with data obtained from the daily

log of the lift- slab subcontractor. An attempt has been made to limit the

description to activities directly related to construction of the building

structural system.

Using the two daily logs, it was possible to reconstruct with reasonable

certainty the schedule for adding column sections and for lifting floor slabs

.

Less clear was the actual schedule followed in the placement of shearwalls

,

the final welding of wedges to the columns and shearheads , and grouting of the

shearhead cavities. Table 3.4.1 begins on October 1 with erection of the 1st

stage columns in the east tower and ends with the building collapse on April

23 at approximately 1:30 p.m. The table does not include the schedule for

construction of the footings or construction of the retaining wall along the

north side of the building.

12
TABLE

3.2.1

4

13

Indicates

grade

50,

A572

steel;

all

other

columns

are

A36

steel.
RPT

NO.

075

076

077

078

080

081

082

088

096

097

098

099

100

101

103

104

106

107

110

111

112

113

114

117

118

120

121

122

123

124

126

127

128

129

131

132

133

136

137

140

141

142

143

146

TABLE 3.4,1

DATE

10/01/86

10/02/86

10/03/86

10/06/86

10/08/86

10/09/86

10/10/86

10/17/86

10/28/86

10/29/86

10/30/86

10/31/86

11/03/86

11/04/86

11/06/86

11/07/86

11/11/86

11/12/86

11/17/86

11/18/86

11/19/86

11/20/86

11/21/86

11/25/86

11/26/86

12/01/86

12 /02/86

12/03/86

12/04/86

12/05/86

12/08/86

12/09/86

12/10/86

12/11/86

12/15/86

12/16/86

12/17/86

12/ 22/86

12/23/86

12/29/86

12/30/86

12/31/86

01/02/87

01/07/87

BUILDING ERECTION SEQUENCE

ACTIVITY

Erect columns east wing

Erect columns east wing

Erect columns west wing

Plumb columns

Grout base plates east wing

Place slab on grade east wing

Grout base plates west wing, place slab on grade east wing

Place slab on grade west wing

Place "D" level east wing

Place "D" level west wing

Place "C" level east wing

Place "C" level west wing

Place "GND" level east wing

Place "GND" level west wing

Place "1st" level east wing

Place "1st" level west wing

Pour cancelled due to forecast of snow/sleet/rain

Place "2nd" level east wing

Place "2nd" level west wing

Place "3rd" level east wing

Pour cancelled due to snow

Place "3rd" level west wing

Place "4th" level east wing

Pour cancelled due to rain

Pour cancelled due to rain

Place "4th" level west wing

Place "5 th" level east wing. Meeting with rep from W.R.

Grace concerning slow set of yesterdays pour.

Pour cancelled due to rain

Place "5th" level west wing

Place "6th" level east wing

Place "6th" level west wing

Pour cancelled due to snow/sleet/rain

Place "7th" level east wing

Concrete plant breakdown, one load on job, pour cancelled

Place "7th" level west wing

Place "8th" level east wing

Place "8th" level west wing

Place "9th" level east wing

Place "9th" level west wing

Place "10th" level east wing

Place "10th" level west wing

Place "11th" level east wing

Work cancelled due to snow/rain

Place "11th" level west wing

14
147 01/08/87

148 01/09/87

150 01/13/87

152 01/15/87

156 01/21/87

164 02/02/87

165 02/03/87

170 02/10/87

171 02/11/87

172 02/12/87

173 02/13/87

174 02/16/87

175 02/17/87

176 02/18/87

177 02/19/87

178 02/20/87

179 02/23/87

180 02/24/87

181 02/25/87

182 02/26/87

183 02/27/87

184 03/02/87

185 03/03/87

186 03/04/87

187 03/05/87

188 03/06/87

189 03/09/87

190 03/10/87

191 03/11/87

192 03/12/87

194 03/16/87

195 03/17/87

196 03/18/87

197 03/19/87

198 03/20/87

199 03/23/87

200 03/24/87

201 03/25/87

Place "12th" level east wing

Place "12th" level west wing

Place "Roof" level east wing

Place "Roof" level west wing

Begin prestressing operation

Lift column steel onto roof slabs

Lift column steel onto roof slabs

STAGE I LIFT - ROOF/12 east wing

Park ROOF/12 east wing, lift 11/10/9 east wing

Park 11/10/9 east wing, lift ROOF/12 west wing

Blocking and welding on slabs east wing

Lift 8/7/6 (Notes say 7/6/5) east wing

Park 8/7/6 east wing, park ROOF/12 west wing

Lift 5/4/3 (Notes say 2/1) east wing, lift 11/10/9 west

wing

Park 5/4/3 east wing, park 11/10/9 west wing

Lift 8/7/6 west wing, lift 2/1/GND east wing

Work cancelled due to snow

Park 8/7/6 west wing, lift 5/4/3 west wing, park 2/1/GND east

wing, lift C/D east wing

Lock D east wing, park 5/4/3 west wing

Lift and lock C east wing, lift and park 2/1/GND west wing

Lock C east wing, set Stage II columns east wing (Notes say

west wing)

Lift C/D west wing, lock D west wing, weld columns east

wing

Lift and lock C west wing, weld columns east wing

Set Stage II columns west wing, weld columns east wing.

Observed "stress cable lifting thru the concrete slab" on D level

Set Stage II columns west wing, weld columns east and west wings

Weld columns east and west wings

STAGE II LIFT - ROOF/12 east wing, weld columns west wing.

"Hollow spots" reported in E level, Col. 12, garage

Lift ROOF/12 east wing. "Building being leveled off with

cables on east end.

Placed shearwalls on D and E levels at following locations:

C-3 , A&B- 8 , G&H-2 , H-4&5, F-8&9, A&B- 11

Lift and park ROOF/12 east wing. Place pour strips on D

level

.

Lift and park 11/10/9 east wing, weld columns west wing.

Large amount of hydraulic oil near Col. 2G, level D

Lift and park 8/7/6 east wing

Lift and park ROOF/12 west wing, lift and park 5/4/3 east wing

Lift and lock GND east wing, lift 11/10/9 west wing. Laying

in concrete block on C level.

Lift and park 2/1 and lock GND east wing, lift and park

11/10/9 west wing. "West end - leveled off building steel"

Set Stage III columns east wing, lift and park 8/7/6 west wing

Lift 5/4/3 west wing, weld column splices east wing

Park 5/4/3 and lift 2/1/GND west wing, weld column splices

east wing, lock GND west wing

Lock GND west wing, lift and lock 1 west wing

15
202 03/26/87

203 03/27/87

204 03/30/87

205 03/31/87

206 04/01/87

207 04/02/87

208 04/03/87

211 04/08/87

212 04/09/87

213 04/10/87

214 04/13/87

215 04/14/87

216 04/15/87

217 04/16/87

218 04/20/87

219 04/21/87

220 04/22/87

221 04/23/87

STAGE III LIFT ROOF/12 east wing. Set Stage III columns west

wing. Placed shearwalls at level D on 8 line, 11 line and F

line east wing. Placed elevator shearwall at level E.

Set Stage III columns west wing, park ROOF/12 east wing

Lift and park 11/10/9 east wing. Leveling off east wing.

Weld column splices west wing. Lift and park 8,7,6 east wing.

Place shearwall at C-3 and shearwall on A line between lines

8 and 9

Lift and park 5/4/3 east wing, complete welding of column

splices west wing.

Start STAGE III lifting, west wing

Lift 11/10/9 west wing, set STAGE IV columns east wing,

placed shearwalls at interior of building on level C, park

11/10/9 west wing

Park 11/10/9 and lift 8/7/6 west wing

Lift and park 8/7/6 west wing, finish welding column splices

east wing

Lift and park 5/4/3 and lock 3 west wing

Lift and lock 2 and set STAGE IV columns west wing

Set STAGE IV columns west wing, STAGE IV LIFT ROOF/12 east wing

Set STAGE IV columns west wing

Lift and park 11/10/9 east wing, fill column pockets on level

2 east and west wings, placed elevator shearwall from level D

to level C

Lifting 8/7/6 east wing, fill column pockets 1/GRD. Place

shearwalls GND floor to 1st floor.

STAGE IV LIFT ROOF/12 west wing, park 8/7/6 east wing

Lift 4th floor east wing. Park ROOF/12 west wing.

Lift 5th floor into permanent position on east wing. Park

11,10,9 west wing and guy west wing.

Place shearwalls from 1st floor to 2nd floor. Building

collapsed at 1:30 p.m.

16
230 '

COco

4-J COoo

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COCD SI

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CO£0

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12

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17
18
Figure

3.2.3

Location

of

post-tensioning

tendons
20

Figure

3.2.4

Plan

view

of

footings
TYPICAL

INTERIOR

COLUMN

CD

21

Figure

3.2.5

Typical

detail

of

interior

column

footing
22
Gusset

plates

23

Figure

3.3.2

Shearhead

detail
Figure 3.3.3 Detail of slab to column connectior,

24
CDOc©D

cr©©

_©© c c o o

*-* •M

Ifi c/i o o

Q. a

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CO

©c o CO Cl E E u.©a c _c

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05

Li.

25
4. DESCRIPTION OF COLLAPSE

4.1 INTRODUCTION

Construction activities preceding the collapse, eyewitness accounts of the

sequence of events associated with the collapse, and descriptions of the

debris are presented in this chapter. This information was used in determining

the most probable cause of the collapse as described in Chapter 9.

4.2 ACTIVITIES PRECEDING COLLAPSE

Several construction operations were in progress on the day of the collapse.

Some of these activities involved portions of the project unrelated to the two

building towers which actually collapsed. Level A of the parking garage south

of the east tower was being cast and finished. A retaining wall was being

backfilled at the northeast corner of the east tower.

Workmen were installing nonstructural items in both towers of the main

structure. Electricians were installing electrical services on the lower

floors of the west tower. Plumbers were installing pipe hangers and waste

water lines and carpenters were installing nonstructural steel studding in the

west tower. It is unlikely that any of these activities contributed to the

collapse of the structure.

Several structural operations were in progress in both the east and west

towers. Concrete was being placed in shearwall forms between ground level and

level 1 at the south side of both the east and west towers. Concrete was also

being placed in the shearwall at the elevator shaft between level C and the

ground level. In addition, concrete was being placed in the cavities between

the columns and shearheads on levels 2 and 3. Masons had just finished placing

pour strips to connect the floors of the east and west towers at level 2 at the

time of the collapse.

Potentially critical construction procedures being conducted on the 23rd of

April included lifting of the floor slabs and securing these slabs with wedges

in both the east and west towers. The lifting operation in the east tower had

progressed to the end of stage IV while in the west tower stage IV was about

half complete. On the morning of the 23rd of April, the level 5 floor slab of

the east tower was raised to its final position and wedges were installed.

These wedges, having been tack-welded in position as they were placed initially,

were being welded permanently to the columns and shearheads when the collapse

occurred.

Erection operations were being conducted in the west tower at the time of the

collapse. The roof and level 12 floor slabs had been lifted to their temporary

positions at the top of the stage IV column sections on the 21st of April.

The slabs for floor levels 9/10/11 had been raised to their temporary position

in the fourth stage of lifting by approximately 11:30 a.m. on the morning of April

23rd. A 12-ton (107 kN) capacity hydraulic jack had been placed between the

slabs at level 11 of the east and west towers approximately 4 ft (1.2 m) from

column line C to adjust laterally the position of the floors that were being

lifted. Figure 4.2.1 is a schematic representation of the state of construction

26
of floor slabs, shearwalls , and cavities around the columns at the time the

collapse occurred. As shown in this figure, shearwalls on the north side of

the east and west towers had been cast prior to the day of the collapse and

had developed some portion of their design strength. Shearwalls on the south

side of the east and west towers had been cast between the ground level and

level 1 on the day of the collapse and had not developed significant strength

at the time of the collapse.

At the time of the collapse, workmen were installing wedges to hold slabs

9/10/11 in this position temporarily. Wedges had been installed along column

lines G and H and an ironworker (#14 in figure 4.3.1) was installing temporary

wedges at column E4.8. A floor plan of the west tower showing the locations

and sequence for which wedges had been installed to support the group of slabs

for levels 9/10/11 is shown in figure 4.2.2.

4.3 EYEWITNESS ACCOUNTS OF COLLAPSE

Descriptions of the collapse were provided by 45 individuals who were on the

site or nearby. Fourteen of these individuals were in the structure at the

time of the collapse and 31 were at various locations around the structure.

Three other individuals who were not on the jobsite at the time of the collapse

were interviewed to obtain general information about construction procedures

used at the jobsite. Most witnesses who were in the structure at the time of

the collapse had a limited view of the structure and were not in a position to see all aspects of the collapse, but saw only the failure of members in their

immediate vicinity. This section will summarize the general observations of

witnesses and then consider in detail the descriptions of key aspects of the

event given by four eyewitnesses. The locations of fourteen individuals who were known to have been in the vicinity of the structure at the time of the

collapse and who could describe the event in some detail are shown in figure

4.3.1. Other witnesses in the structure were not in a position to observe the

initiation or progression of the collapse.

Descriptions of the collapse varied somewhat in detail. This is inevitable in

such situations because of the suddenness of the event and, particularly in

this case, because of the relatively short amount of time from the start of

the failure to its conclusion. Many witnesses could not accurately describe

what they saw because of their lack of familiarity with construction techniques

and equipment. Some witnesses, for example, referred to the boom of a concrete pumper truck as a crane

.

Witnesses disagreed as to the activities they recalled seeing in progress just

prior to the collapse. Two individuals indicated they saw cranes working at

the site and that steel column sections were being welded on top of existing

columns on the east tower. Other witnesses could not recall specifically what

workmen were doing, if anything, on the roofs of the buildings prior to the

collapse. The construction manager at the jobsite indicated that no cranes

were on the site on the day of the collapse and no column extensions were

being erected.

In general, all witnesses agreed that the collapse was extremely rapid.

Estimates of the time from initiation of failure (the point at which their

27
attention was first drawn to the structure by noise) to the time at which the

collapse was complete ranged between 2 and 10 seconds. The majority of witnesses

stated that the collapse took place within a time span of approximately 5 seconds. Note the time required for a particle to fall from a height of 81.5

ft (24.8 m) , the height of the roof slab above level E, under the force of

gravity is 2.25 sec. Apparently the lower level floor slabs offered minimal

resistance to the debris from above as the failure progressed vertically.

Almost all witnesses indicated they first noticed the failure because of a loud noise. Descriptions of the character of the noise varied from one witness to another. The majority of witnesses reported hearing a single

initial noise, which was followed by a general rumbling as slabs fell and the

collapse took place. This initial noise was variously described as a loud

snap, crack, or boom, which some likened to the bang of a hinged tailgate of a dump truck, a spring breaking on a big truck, or a piece of metal snapping

under pressure. Others were aware of only the general sound of the collapse

itself, which they described as the sound of a large jet, a rumble, or the

sound of an earthquake or thunder. They did not specifically mention hearing

a single individual noise which might have accompanied the failure of a single

component at the beginning of the failure.

Two witnesses believed the building might have been swaying prior to the

collapse, and one individual who had passed the construction site frequently

in the days prior to the collapse stated he believed he could detect a slight

swaying motion of the building in response to wind loads. Winds were light and

variable on the day of the collapse, with sustained winds of approximately 12

knots (6.2 m/s) ESE as measured by the National Weather Service Office (NWSO)

at the Sikorsky Memorial Airport, Bridgeport, Connecticut. A copy of the NWSO

stripchart for April 23 was obtained from the National Climatic Data Center.

The stripchart indicates peak winds of 20 knots (10.3 m/s) from 12 Noon to 1

p.m. (LST) . The Sikorsky Memorial Airport is approximately 5 miles (8 km)

southeast of the site of the collapse.

Almost all witnesses of the collapse indicated that they did not see the

building moving prior to the collapse. None of the workmen who had been in

the building and who escaped the collapse reported any swaying or unusual

motion of the building.

The majority of witnesses indioated they believed the collapse started in the

west tower. Only one witness indicated that the east tower began to collapse

first. As a testimony to the speed of the collapse and to the shock it

instilled in observers, this witness related in one interview that the west

tower collapsed first and in a subsequent interview that the east tower failed

first. Of the workmen who were in the building at the time of the collapse,

all who could determine the location of the first sound of the failure said

the failure appeared to start high in the structure in the west tower.

Additional eyewitness accounts of the early stages of the collapse will be

discussed in Chapter 9. The testimony of one of the survivors of the collapse

was considered particularly valuable. Just prior to the collapse, an ironworker

(#14, figure 4.3.1) was installing wedges at column E4.8 underneath slabs

9/10/11, the package of slabs which had been raised to its temporary parking

position in stage IV of the lifting sequence on the morning of April 23rd.

28
The witness stated that he "heard a loud noise... like steel breaking under

pressure," which he believed came from within 25 ft (7.6 m) of where he was working, either directly above him or toward the center of the slab in the

vicinity of column E3 . 8

.

He then noticed that the floor slab directly over

his head "was cracking just like ice breaking." The floor slabs above him

then collapsed, pushing him between the uprights of the scaffolding on which

he was standing. Protected somewhat by the scaffolding, he was carried down

with the collapsing structure and later rescued. He was the only survivor of

the men working in that area at the time of the collapse.

Three descriptions of the overall collapse were essentially similar. One of these

accounts was given by the operator of a concrete pump truck parked at the southeast

corner of the west tower (#3 in figure 4.3.1). The attention of the operator

was directed toward the building because of his need to respond to signals of

workmen who were removing a hose used for placing concrete in shearwall forms

within the building. The operator stated that he was first made aware of some problem in the structure by a loud, metallic noise that he described as being

similar to the noise which could be made by the breaking of a leaf spring on a truck. The noise seemed to come from the southwest corner of the west

building near the top of the slabs which were located in their temporary

positions (levels 9/10/11 and 12/roof)

.

The operator reported that he turned

immediately toward the sound and saw the corners of the two uppermost slabs

start to move downward until they struck the slabs immediately below them, at

which time the fall of the slabs slowed momentarily. The collapse then spread

eastward and northeastward throughout the uppermost slabs of the west tower

and then the entire building collapsed vertically. He reported that the east

tower collapsed as a result of being struck by sections of the collapsing west

tower and as a result of being linked to the west tower at several locations.

A second description of the collapse seemed to agree in general with the

report given by the operator of the pump truck. A construction contractor who

had stopped on a nearby roadway because of a minor motor vehicle accident

witnessed the event from a distance of approximately 100 yards (91.4 m) (#4 in

figure 4.3.1). He indicated that he glanced at the building several times to

observe the operations being conducted as he was waiting for police to deal

with the accident. His attention was then drawn to the building by a loud, sudden

noise which he described as either a crack, boom, or snap, stressing that the

sound was sharp rather than muffled. He related that he then saw the west

region of the top slabs of the west tower drop down to the lower floors. The

failure then appeared to progress from the west side of the west tower

eastward, enveloping all of the west tower to the extent that the slabs

appeared to be dropping vertically at some point in the failure. At some point that he could not describe definitely, the east tower became involved in

the collapse as a result of either impacts from debris falling from the west

tower or from other factors unknown. The east tower then collapsed completely.

The witness stated that the total amount of time required for the collapse

seemed to him to be no more than 5 seconds

.

The Connecticut state policeman (#11 in figure 4.3.1) who was controlling

traffic at the scene of the motor vehicle accident mentioned above also

witnessed the collapse. This witness stated that he first heard three loud

booms like explosions. As he turned to look in the direction of the noises,

29
he saw the west tower collapsing straight down. The west tower then keeled

over eastward into the east tower, causing the top floors of the east tower to

snap and fall onto the floors below. The east tower then fell slightly

eastward and collapsed completely.

4.4 OBSERVATIONS OF DEBRIS

Observations of the debris provide insight into the pattern of collapse of the

structure. This insight is helpful in identifying the sequence of collapse

and possibly the point of initiation of the failure. The information presented

in this section will be used in subsequent sections of the report in connection

with analysis of the structure and the explanation of the most probable cause

of the failure.

4.4.1 Overall Collapse

An overview of the collapsed structure is shown in figure 4.4.1. The photograph

was taken approximately two hours after the collapse and prior to any significant

amount of debris removal in the rescue operation. The undamaged columns in

the foreground are for the parking garage which had not been erected at the

time of the collapse. Two distinct heaps of debris, one the east tower and one the west tower, are clearly visible. The clear area between the heaps of

debris in the photograph is the service core between the two towers . In each

tower, the columns are draped over the pile of concrete floor slabs. It

appears that the slabs collapsed near the center of each tower, pulling the

columns inward.

Column Bending

The collapse toward the center of each tower produced inelastic bending of the

upper portions of the perimeter columns. Column C3, a perimeter column on the

north side of the west tower, is shown in figure 4.4.2. Note the column

bending at level 1 in this case is about the strong axis of the column. The

deformation of column El, a perimeter column on the west side of the same tower is shown in figure 4.4.3. Significant inelastic action and formation of

a plastic hinge occurred in the upper portions of the column near level 1.

Bending occurred over the lower three stories about an axis approximately

midway between the principal axes of the column.

Lower Story Lateral Displacement

The lower stories of each tower did not exhibit any significant overall

lateral displacement although there was some localized lateral displacement.

The lower two levels (level E and level D) of the columns on the east edge of

the east tower (column line 12) shown in figure 4.4.4, for example, are essentially vertical. The shearwalls in place at the lower levels provided some lateral support. The north face of the shearwall between column lines 8 and 9

and oriented in the east-west direction is shown in figure 4.4.5. The

orientation of the cracks indicates the wall was subjected to a shear force

directed from west to east.

30
Torsional Deformations

Several of the columns twisted significantly. Column C3 . 8

,

for example, shown

in figure 4.4.6 twisted approximately 25 degrees. The pattern of twisting of

the columns is shown in figure 4.4.7. The twists given are approximate and

represent visual estimates made at the site prior to removal of the columns.

Several columns were removed before estimates could be obtained.

Figure 4.4.7 shows twists of as much as 60 degrees occurred. For some columns, twists in opposite directions occurred at different levels. Column

C6 , for example, twisted counterclockwise up to level D, then clockwise with

increasing magnitude at levels 1, 2, 3 and 4. At the lower levels in some

cases, there was no column twist. Column C3 , for example, did not twist up to

the ground level. As indicated, some of the twisting of column B12 was elastic rather than inelastic. Observations of a series of photographs of

this column showed that some of the initial counterclockwise twist between

levels D and C was recovered as the debris was removed.

In general there does not appear to be a consistent pattern to the twists in

figure 4.4.7. Observed twists in the lower levels (level E to level C) of the

east tower, however, are all counterclockwise. Also there are no significant

differences between twists in the east versus the west tower. For example,

columns El and G1 on the west edge of the building twisted counterclockwise,

whereas column C2 on the north edge twisted clockwise. A similar situation

occurred in the east tower with columns A10 , B12 , and D12 . This lack of any

pattern consistent with rigid body twisting of the slabs indicates the slabs

generally did not remain intact during the collapse of the building.

Floor Slab Orientation

The orientation of floor slab segments in the debris provides an indication of

the manner in which the building collapsed. Figure 4.4.8 shows the orientation

of the lower level floor slabs at the northwest corner of the west tower. The

photograph was taken after the upper level slabs had been removed. Approximately

seven slabs are visible. Note the slab segments to the left of the center of

the photograph (between column lines C and E) are essentially horizontal. The

slab segments on the right are at an angle of approximately 60 degrees with the

horizontal. Clearly these slabs did not remain intact as the building

collapsed. The slabs broke and the individual segments fell separately. The

pattern in figure 4.4.8 would indicate the lower seven or eight levels of

slabs broke in the vicinity of column line E during the collapse.

The upper level floor slabs at the northwest corner of the west tower are shown in figure 4.4.9. Eight slabs are visible in the photograph. The slabs

are oriented approximately vertically. The plastic hinge in column El (the kink

in the center of the photograph) is at level 1, which is located at approximately

mid-height of the column in place at the time of collapse (column up to level

6). Apparently during the collapse, portions of the upper level slabs fell

with the columns, rotating as the columns deformed and bent toward the center

of the building.

31
The orientation of the lower level slabs in the center portion of the west

tower near the perimeter shearwall on the south side is shown in figure

4.4.10. Again the eight lower level slabs are horizontal and the remaining eight

slabs are essentially vertical.

The horizontal orientation of the lower level slabs and the more nearly

vertical orientation of the upper level slabs was also apparent in the center

portion of the east tower as shown in figure 4.4.11.

The upper level slabs at the southeast corner of the east tower fell in a fashion similar to that of the slabs on the northwest corner of the west

tower. This is apparent in figure 4.4.12. In this case some of the slabs

actually rotated beyond the vertical position and landed upside down. The two

slabs in the lower right portion of the figure are in this position.

Collapse Pattern

To more clearly visualize the overall collapse pattern of the structure, the

composite sketches of the deformed shapes of the columns in each tower in

figures 4.4.13 and 4.4.14 were prepared. Three views are shown for each

tower. Photographs of the columns taken prior to removal from the collapsed

structure were used to produce these sketches. The columns shown are not

complete in every case since portions were removed as the rescue operation

proceeded. Displacements were estimated at the various floor levels. Column

segments between floor levels and/or plastic hinges are indicated as undeformed

elements. Column A7 , for example, in the northwest corner of the east tower

(figure 4.4.14) has nine elements representing level E through level 6.

Referring to figure 4.4.13, for the west tower, note that there is no indication

of any consistent translation or rotation in which the structure deformed as a unit. It appears that each column deformed independently and the floor slabs

broke early in the collapse sequence. The upper level slabs caused the

columns to deflect toward the center of the building as these slabs fell.

Figure 4.4.1 clearly illustrates this behavior.

The lower levels of the columns remained relatively straight and in most cases

vertical. Significant deformations or buckling of the columns in general

occurred at the ground level or level 1. This was the case for columns -Cl,

C2 , C3 , C.5, C6

,

E3

,

Gl, HI, H2 and H5

.

This would be expected since the floor

slabs at level D, level C and ground level were fully welded in place. Note

that column C3 at the shearwall on the north side of the west tower remained

vertical up to level 1. There was some twisting of the columns around the

perimeter of the building at the lower levels.

The deformed shapes of the columns in the east tower shown in figure 4.4.14

indicate the mode of collapse in the two towers was similar.

4.4.2 Floor Slabs

The condition of the collapsed floor slabs in the vicinity of the shearheads

near the top of a column is shown in figure 4.4.15. The behavior noted in

Section 4.4.1 in which the upper level slabs fell with the columns, rotating as

32
the columns deformed, is apparent in the photograph. The slabs are completely

separated from the shearheads and are aligned in a nearly vertical position.

The post- tensioning strands are clearly visible.

Typical crack patterns and broken floor slabs are shown in figures 4.4.16 and

4.4.17. Note the slabs broke into a number of small pieces. This would be

expected once the post- tensioning is lost due to the lack of conventional

reinforcing steel except in the vicinity of the columns.

4.4.3 Shearheads

The deformations of the shearheads that occurred at various levels in the

structure during the collapse are shown in the stack of shearheads in figure

4.4.18. The position of the shearheads in the photograph was not changed as the

debris was removed. Note that sixteen shearheads came to rest at level E. The ten shearheads which had been located in the upper levels of the structure

were relatively undamaged. Similar conditions were observed at most locations

throughout the structure. This indicates that the shearheads in the upper

levels of the building were not subjected to forces large enough to induce

substantial inelastic deformations during the collapse. Two possible explanations are: (a) the post- tensioning in the slabs could have been lost early in

the collapse and the slabs broke away from the shearheads, and/or (b) the wedges

supporting the upper level slabs could have been dislodged or the weld blocks

on the columns could have sheared off.

Although welds at the weld blocks failed, this occured only in a few cases. These were confined to the lower levels of the structure. The crack pattern

in the slabs discussed in Section 4.4.2 suggests that the post- tensioning was

lost. The deformed shapes of the columns in figures 4.4.13 and 4.4.14

discussed in Section 4.4.1 support this conclusion. The wedges had to be

dislodged for the shearheads to come together as shown in figure 4.4.18. One

would have expected to see some concrete remaining partially intact around the

shearheads due to the presence of the reinforcing steel in the vicinity of the

columns. If only the wedges had been dislodged and the post- tensioning not

lost, one would have expected to see considerable scoring along the length of

the columns as the floor slabs fell and the shearheads scraped the column.

This was generally not the case. Scoring due to movement of the shearheads

down the columns was observed in only a limited number of cases. Clearly, the

wedges supporting the slabs in the upper levels were dislodged and the posttensioning was lost early in the collapse.

The deformations of the lower six shearheads in figure 4.4.18 indicate they

were subjected to considerable force during the collapse. It would appear

that the slabs in the lower levels remained whole or portions of the slabs

remained attached to the shearheads because of the conventional bonded

reinforcing. Enough force was transmitted to the shearheads to cause the

shearhead to fail, to cause a punching shear type failure of the slab or to

fail the welds between the weld block and the column or the welds between the

wedges and the weld blocks. Each of these conditions occurred. Figure 4.4.19

illustrates a situation where the shearhead failed. Although the welds to the

wedge and the column held, the shearhead was literally torn apart. Failure of

the weld block welds and the welds of the wedges are discussed in Section 4.4.4.

33
The locations of the shearheads along the height of the columns throughout the

structure are shown in figure 4.4.20. In general, shearheads on the perimeter

columns remained in place at the floor levels as the building collapsed. On

the interior columns, however, the shearheads were dislodged and came together

as shown in figure 4.4.18. Apparently the concrete slabs broke away from the

shearheads around the perimeter leaving the shearheads in place. At the

interior columns, the slabs exerted sufficient force to fail the shearheadcolumn connection. There was more conventional reinforcing in the slab at the

interior columns than at the perimeter columns and it was located on all four

sides of the column.

The shearhead detail at what would appear to be a corner column is shown in

figure 4.4.21. Only one such shearhead detail, however, was found. Note the

post- tensioning tendons were run through the channel sections comprising the

shearhead rather than over the shearhead. This was probably due to the need to

locate the post- tensioning tendons at mid- depth of the slab.

4.4.4 Welds

The performance of the welds varied throughout the structure. In some cases the

welds remained intact without signs of distress while in other cases the welds

failed. The performance of a column splice weld is shown in figure 4.4.22.

Despite the apparently high forces exerted on the column section and the

severe deformations illustrated in figure 4.4.22, the column splice weld

remained intact.

The variation in performance of the welds at the shearhead - column connection

is illustrated in figures 4.4.23-27. In some cases, as shown in figure

4.4.23, the welds remained intact. There were no apparent deformations of the

elements of the connection and it is not clear to what force the connection was subjected. The welds on the weld block shown in figure 4.4.19 remained

intact, although the failure of the shearhead suggests that considerable force

was applied to the weld block.

Some welds failed on the weld blocks as shown in figures 4.4.24, 4.4.25 and

4.4.26. The wedges were fully welded and these connections were obviously at

slabs in the building which were in their final locations rather than parked

temporarily. In the latter case, the wedges were only tack welded. Note the

rotation of the wedge in figure 4.4.25. Note also the small shim on the

bottom of the wedge. In all three figures, the wedge had pulled away from the

column flange and the weld block. The welds do not appear to have fractured

but rather the wedge separated from the weld, probably due to the lack of

fusion.

Failure of a weld on a weld block is shown in figure 4.4.27. A crack in the

column flange can be seen at the top of the weld block (to the right in the

photograph) . This crack propagated under the weld block and through the weld

block weld about half way down the weld block. Note the separation of the upper

portion of the weld block from the column flange. Again there appears to be a lack of fusion between the weld and weld block at this location.

34
A weld for a wedge under one of the upper level slabs which was parked

temporarily prior to being lifted to its final position is shown in figure

4.4.28. Note in this case the wedge was welded to the column at each end with

a small tack weld. The wedge was not welded to the weld block or shearhead.

This was done for ease of removal of the wedge prior to lifting the slab to

its final position. In several cases these tack welds were extremely small.

They were difficult to see and barely perceptible to the touch as a raised

portion on the flange.

4.4.5 Additional Observations

Additional observations during the on-site investigation included the performance

of the perimeter walls at the garage levels and the performance of the jack

rods used to lift the slabs.

The perimeter wall on the north side of the building was below ground level.

This basement wall extended from level E to level C in the parking garage

.

The wall was in contact with the floor slabs at these levels. The retaining

wall on the west side of the buildings was not in contact with the structure.

A photograph taken on May 1, 1987 of the basement wall on the north side of

the east tower, viewed from the east end, is shown in figure 4.4.29. Note

that the wall appears straight. Figure 4.4.30 presents a similar view taken

on July 22, 1987 after the rescue operations were complete and much of the

debris removed. There is a noticeable lateral displacement or bowing along

the length of wall. Evidence of movement of the retaining wall was also

observed on the surface of the backfill, which exhibited two parallel tension

cracks in the direction of the wall extending over the entire length of the

wall (figure 4.4.31). Either the wall was damaged during the collapse or by

surcharge loads and vibrations during removal of the debris

.

Many of the jack rods used to lift the slabs were cut with an acetylene torch

during removal of the debris. Three fractured jack rods were found in the debris

at the landfill and one fractured rod was removed from the jack at column A9

in the east tower. The jack locations of the rods from the debris could not

be determined. These three rods were approximately 3 to 5 ft (0.9 to 1.5 m)

in length. The fractures indicated the rods were subjected to bending and

axial load. Results from metallographic tests of the fracture surface are presented in Section 5.5.3.

4.4.6 Detailed Observations of Stage IV Columns - West Tower

As debris was removed from the construction site, an effort was made to

identify column sections at the Seaside Park landfill. Specifically, this

effort focused on the upper sections of the columns and the shearheads that

were either in the process of being adjusted or had been lifted just prior to

the collapse. In certain cases the entire upper column stage (stage TV) was found

intact and the location of the column could be positively identified by the

shop number. In most cases, however, the upper column stage had been flame

-

cut during the rescue operation. It was therefore necessary to determine the

location of the column by measuring the section size, by carefully examining

the upper end for the presence of tack welds near the weld plates, and by

classifying the types of shearheads whenever they were present. With few

35
exceptions, the specific location of the upper column stages removed from the

debris could be established.

As the investigation progressed, column sections from the west tower assumed

the greatest importance. These sections were examined and described in

considerable detail. Descriptions of the stage IV columns from the west tower

are included in this report as Appendix A. In some cases only segments of the

lower column stages could be positively identified. Where these lower segments

provide valuable information, such as the remains of the upper-level shearheads

,

their characteristics are included as a part of the general description. A

summary of the tack welds and marks found on the stage IV columns from the

west tower is presented in table 4.4.1.

At the time of the collapse, wedges had been placed under the shearheads of

slab 9 on column lines G and H and at column E.5. The remaining columns

picked up the loads from the package of slabs 9/10/11 through the jacks and,

during the collapse, most of these column tops were indented by the jacks.

Similarly, the underside of the shearheads at slab 9 were indented by the

lifting nuts. A typical imprint caused by the base of a 150-kip (667 kN) jack

is shown in figure 4.4.32. Figure 4.4.33 shows a typical indentation in the

underside of a shearhead lifting angle. In some cases the lifting nut slid

off or "kicked out" from under the lifting angle, leaving a clear track in the

bottom face of the angle. This usually was followed by the lifting nut

impacting the column web and causing an indentation in the web at approximately

52 in (1.32 m) below the column top.

In general, the columns along line C experienced the most damage in the

collapse. Much of this damage was caused by rotation under cantilever action

of the shearheads in slabs 9/10/11. Column C4.8, shown in figure 4.4.34, is

typical of the observed damage patterns. Most of the columns in this line

retained the upper shearheads close to their pre-collapse positions. In at

least three cases the top (roof) shearhead slid off the top of the column.

Gouges in the column flanges due to interference with the shearheads during

the lifting operation were observed on some stage IV column sections. An

example of this is column C3 shown in figure 4.4.35.

The columns along line E experienced less damage than did those in line C and

the shearheads tended to slide downward to levels 1, 2 or 3. Most or all of

the wedges at these three levels had been fully welded at the time of the

collapse. With the exception of the columns on line 6 and columns HI and H2

,

stage IV of all the columns in lines G and H suffered very little damage. Most

remarkable was the condition of the upper weld blocks on these columns . Even

though the wedges had been installed under slabs 9 and 12, there were many

instances where the wedges were dislodged without any visible damage to the

contact surface on the weld blocks.

36
4.4.7 Summary

The following summarizes the eyewitness accounts of the collapse and NBS

observations of the collapsed structure:

1. Wedges were being placed under slabs 9/10/11 in the west tower at the

time of the collapse.

2. There was no perceptible displacement or motion of the building

immediately prior to the collapse.

3. The collapse was preceded by a loud noise heard as far as 100 yards

(91.4 m) from the site.

4. The collapse began in the upper levels of the west tower.

5. The collapse was extremely rapid. The total time involved from

initiation to complete collapse was estimated to be from 2 to 10 seconds.

6. Each tower collapsed toward its center; there was no indication of any

consistent translation or rotation in which the structure deformed as

a unit.

7. The wedges supporting the slabs in the upper levels of the west tower

were dislodged and the post- tensioning in the slabs was lost early in

the collapse.

8 . Many of the shearheads in the upper levels of the structure were

undamaged; shearheads in the lower levels were significantly deformed.

9. Displacements of the basement wall on the north side of the building

occurred after removal of the collapsed structure.

37
TABLE 4,4.1

SURVEY OF STAGE IV COLUMNS FROM WEST TOWER

COLUMN SECTION COLUMN TACK WELD GOUGES COL. TOP LIFTING NUT

NO. FOUND ROOF/12 11/10/9 INDENTED IMPACT

Cl HP10X4

2

Cl Y N Y Y

C2 HP 12X5

3

C2 Y N Y *

C3 W12X65 C3 Y N Y

C3 .

8

W12X65 C3.8 Y N Y

C4 .

8

HP123X53 C4.8 Y N Y Y

C6 W8X35 MISSING

C. 5 W12X65 C . 5 Y N Y

El W10X60 El Y N Y Y • Y

E2 W12X106 E2 Y N Y Y

E3 W1 2X120 E3 Y N Y

E3.8 W12X120 MISSING

E4 .

8

W12X106 E4 .

8

Y N Y Y

E . 5 W12X72 E . 5 Y Y

G1 W10X60 G1 Y Y

G2 W12X106 G2 Y Y

G3 W12X136 G3/G4 Y Y

G4 W12X136 G3/G4 Y Y

G5 W12X106 G5 Y Y

G6 HP12X5

3

G6 Y Y

HI HP10X4

2

HI Y Y

H2 W12X65 H2 Y Y

H3 W12X79 H3 Y Y

H4 W12X79 H4 Y Y

H5 W12X65 H5 Y Y

H6 HP 10X4

2

H6 Y Y

* Some scraping visible but no clear indentation

Note

:

No entry indicates not observed or not relevant to failure condition

38
12,

roof

T3I?ctJ *o z: *d = c 0 •- © *o g 1

2

c 5 ®

E jc

a,

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<B o c 3 «

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- 5

51o a> 0- -Q

00

Q.

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O

0

E

0

C

o

">4—

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h.0

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o

o

o

0*—

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CO

cvi 0u.3

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6 — . 6 UAl 30VIS ~ -.V-.Zl- -.e-.si- -.2-.LE- 30VIS II 30VIS SNWniOO I 30V1S

39

viewed

from

the

South
© 0

C/5 Q_

03OOO n&

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05 o

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05

03 T30EE

c/5 0

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40
41

Figure

4.3.1

Location

of

key

eyewitnesses
42

Figure

4.4.1

Overview

of

collapsed

structure

-

View

to

the

Northeast
Figure 4.4.2 Column C3 bending about the strong axis - View to the East, shearwall on left

43
Figure 4.4.3 Deflected shape of column El - View to the North

44
Figure 4.4.4 East perimeter columns at Levels E and D. East towerView to the North; photo taken after rescue

operations complete

45
Figure .4.5 Diagonal tension crack in East tower shearwall - View to the South

m M

SS h

8

, "'*V'

46
Figure 4.4.6 Twisting deformation of column in West tower

47
Level E

toD

20°CCW

o

<© ©4—

©

48

Figure

4.4.7

Pattern

of

perimeter

column

twists
Figure 4.4.8 Lower level floor slabs at Northwest corner of

West tower - View to the North; column El

49
Figure 4.4.9 Upper floor slabs on Northwest corner of

West tower - View to the North; column El

50
Figure 4.4.10 Lower level floor slabs in center portion of

West tower - View to the Northeast

51
Floor slabs in the central portion of East tower - View to the Southwest

Figure 4.4.11

52

©

54

Figure

4.4.13

Pattern

of

column

deformations

-

West

tower
L'

Ambiance

Plaza

West

Tower

-

South

View

55

Figure

4.4.13

Continued
L'

Ambiance

Plaza

West

Tower -

Plan

View

56

Figure

4.4.13

Continued
L'

Ambiance

Plaza

East

Tower

Viewed

from

SW

Corner

57

Figure

4.4.14

Pattern

of

column

deformations

-

East

tower
L'

Ambiance

Plaza

East

Tower

-

South

View

58

Figure

4.4.14

Continued
L'

Ambiance

Plaza

East

Tower

-

Plan

View

59

Figure

4.4.14

Continued
Figure 4.4.15 Separation of floor slabs from shearheads

60

Figure 4.4.17 Broken floor slabs - View to the

Southeast; column line H

62
Figure 4.4.18 Deformation pattern of shearheads

- column DIO

63
Figure 4.4.19 Failed shearheads

64

Figure 4.4.21 Corner column shearhead detail

66
67
Figure 4.4.23 Intact shearhead support

68

Figure 4.4.25 Weld failure at weld block with shim

70
Figure 4.4.26 Weld failure at weld block

71
Figure 4.4.27 Weld block weld failure

72
73
Figure 4.4.29 Basement wall on North side of East Tower - viewed from the East, May 1, 1987

74
Figure 4.4.30 Basement wall on North side of

East Tower - viewed from the

West, July 22, 1987

75
Figure 4.4.31 Tension cracks in backfill behind basement wall - North side of building

—Am

76
Figure 4.4.32 Typical indentation in top of column due to lifting jack

77
Figure 4.4.33 Typical indentation in underside of lifting angle

78

Figure 4.4.35 Gouge caused by shearhead during

lifting operation

80
5. LABORATORY TESTS

5.1 INTRODUCTION

The material tested during the investigation included cores cut from randomly

selected sections of the concrete slabs and shearwalls , coupon specimens cut

from flanges and webs of the columns, coupon specimens of the welds used to

splice column sections, and coupons cut from the prestressing strand used in

construction of the floor slabs. Major structural components tested included

shearheads , jack rods, and a shearhead- column connection similar to that used

for temporary parking of slabs. The objective of these tests was to determine

the strength of the various materials and components and obtain information on

the performance of key subassemblies . This information was used in the

structural analysis and in determining the most probable cause of the collapse.

Results of the NBS laboratory tests are presented in this chapter. These

results are compared with data provided by the material suppliers and results

obtained by an independent testing laboratory during construction of the

building. Comparisons between the laboratory tests results and values in the

project specifications are included in Chapter 8.

5.2 TESTS OF CONCRETE SPECIMENS

5.2.1 Sampling and Testing Procedures

The strength of concrete was determined from cores cut from various elements

of the structure following the collapse. Cores were cut using a diamond core

bit and tested in accordance with ASTM C 42, ("Standard Method of Obtaining and

Testing Drilled Cores and Sawed Beams of Concrete"). All cores were 3 3/4 in

(95 mm) in diameter. In some instances, cores were cut to a suitable length with

a diamond masonry saw prior to testing. Cores were weighed prior to being

capped or cut and the length of cores was measured both before and after capping.

Variability of the strength of the concrete throughout the structure was also

evaluated approximately using a rebound hammer (ASTM C 805) at the construction

site and at the landfill where debris from the collapse was stored. Although

the rebound hammer cannot be used to measure concrete strength accurately

enough for use in analysis, it can be used to determine if the strength varies

widely throughout a structural member. The rebound hammer tests indicated

that the quality of the concrete was relatively uniform throughout the structure.

Forty- three core samples were taken from various portions of the slabs and

shearwalls. The characteristics and compressive strengths of the cores are listed in tables 5.2.1 and 5.2.2. Field sample numbers designate the locations

from which the samples were obtained. S denoted samples taken from a slab.

RS denoted samples taken at random and for which the location of the concrete

in the structure was not known. SWS denoted samples taken from a shearwall.

The floor and column location was used in the designation when the location of

the sampled specimen was known. Sample No. 25 in table 5.2.1, for example,

came from the level 7 floor slab (FL7) in the vicinity of column G5 . Three

specimens, indicated by asterisks in tables 5.2.1 and 5.2.2, were determined

81
to be unacceptable for testing because of damage sustained prior to being

tested or because they contained amounts of reinforcing steel which could

significantly affect the performance of the sample. Eight randomly- selected

specimens were instrumented with a compressometer to measure axial deformations

of the cylinders as they were tested. Elastic moduli for these eight specimens

were determined in accordance with ASTM C 469, ("Standard Test Method for

Static Modulus of Elasticity and Poisson's Ratio of Concrete in Compression").

Five cores were tested under diametric compression in accordance with ASTM C

496, ("Standard Test Method for Splitting Tensile Strength of Cylindrical

Concrete Specimens"). Four of the specimens that had been subjected to axial

compression were also split diametrically so the interior structure of the

concrete could be examined.

5„2.2 Results of Tests

The density of the cores given in table 5.2.1 was relatively constant. The

mean value was 144 lb/ft3 (2306 kg/m3

) , with a standard deviation of 3 lb/ft 3 (48 .

1

kg/m3

)

.

A summary of the results of the tests of the core samples is given in table

5.2.2. The average compressive strength was 5420 psi (37.4 MPa) for cores

taken from slabs and 4570 psi (31.5 MPa) for cores taken from shearwalls. The

standard deviation was 544 psi (3.75 MPa) for the cores taken from slabs and

257 psi (1.77 MPa) for cores cut from shearwalls. The compressive strength of

the samples obtained from levels 7, 9 and 11 (Sample No. 25, 26, 30) varied by

less than five percent. The roof slab core strength (Sample 34) was about

twenty percent less than these three samples. The average static modulus of

elasticity was 2.89 x 10^ psi (1.99 x 10^ MPa) for cores taken from the slabs

and 3.37 x 10^ psi (2.32 x 10^ MPa) for cores taken from shearwalls. The

average splitting tensile strength of the cores was 565 psi (3.90 MPa). The

splitting tensile strength of these cylinders is in reasonable agreement with

values of splitting strength calculated in accordance with ACI recommendations

[2] (7.5,/f'c - 552 psi or other proposed procedures (1.15 (f£) ^ - 515 psi)

for the mean core compressive strength of 5420 psi [2,3].

Failure surfaces for the four axially loaded cores that were split diametrically

were examined under 22 power magnification. This examination showed no indication of formation of ice lenses: i.e., all voids seen in the matrix of

the paste were essentially spherical.

5.2.3 Comparison of Strengths of Core Samples with Quality Control Samples

Concrete was sampled at the jobsite during construction by a commercial testing

laboratory. Standard 6 in by 12 in cylinders cast at the time of placement of

the concrete were tested in compression at ages of 7 days and 28 days. Field

records indicate that approximately four samples were taken for each 50 cubic

yards of concrete placed in the floor slabs. The rate varied at which samples

were taken from other elements of the structure such as retaining walls and

shearwalls

.

82
The 28 -day strengths of groups of cylinders for each floor level tested by the

testing laboratory are given in table 5.2.4. The average strength of these

cylinders was 4650 psi (32.1 MPa) and the standard deviation was 216 psi (1.49

MPa) . The strengths of samples taken from the collapsed structure were higher than

the strengths of the standard cylinders tested by the testing laboratory. The

standard deviation of strengths was also slightly greater. Both these results

are to be expected. The core samples taken following the collapse were much

older than the standard cylinders, and the conditions of placement of the

concrete in the structure were more variable than those used to cast the

standard cylinders.

Two questions have been raised in regard to the concrete present in the

structure. The first involves speculation that the concrete used in the floor

slabs might have frozen, causing the concrete to be weakened. Although freezing

temperatures did exist during some of the days on which the floor slabs were

cast, the slab contractor covered the slabs with electric heating blankets to

prevent damage to concrete. No indication of damage due to freezing was found

in the samples tested by NBS . Furthermore the concrete was apparently not

weakened as a result of extremes of temperature. The fact that no concrete

was found which had a strength less than that specified on the plans is a good

indication that no freezing of concrete occurred, or, if it did occur, that it

did not damage the concrete.

A second question has been raised about the use of a set-accelerating admixture

in some of the concrete placed in the floor slabs during cold weather. It has

been speculated that the admixture weakened the concrete. Records of the

commercial testing laboratory indicate that the admixture was used in slabs of

both buildings in the second through sixth floors. The average 28 -day

strength of cylinders cast in conjunction with the placement of these floor

slabs was 4830 psi (33.3 MPa). The use of the admixture, therefore, does not

appear to have reduced the strength of the concrete. In addition, literature

obtained from the manufacturer of the admixture notes that tests by a private

materials testing laboratory indicated concrete containing the admixture

conformed with required strength provisions of ASTM C 494 ("Standard Specifications

for Chemical Admixtures for Concrete") for a Type C admixture.

5.3 TESTS OF STEEL SPECIMENS

5.3.1 Sampling and Testing Procedure

Tensile coupon tests were conducted to evaluate the mechanical properties of

steel in the columns. The objectives of the tests were: to determine the

degree of variability of the mechanical properties within a column sample; to

determine the conformance of the test results to the requirements of ASTM A 36

("Standard Specification for Structural Steel") and ASTM A 572 ("High-Strength

Low-Alloy Columbium-Vanadium Steels of Structural Quality"); and to compare

the results with those of mill and laboratory tests conducted by the steel

manufacturer. Three column segments, two obtained from the landfill and one obtained from the construction site, were the sources of the specimens.

83
Twelve flat test coupons were machined in accordance with ASTM E 8-85b ("Standard

Methods of Tension Testing of Metallic Materials") from three column segments

identified as NBS 21-1, 21-2, and 21-3. Four coupons, three cut longitudinal

with respect to the rolling direction and one cut transverse to the rolling

direction, were machined from each of the column segments. The source of the

individual coupons and the pertinent dimensions are presented in table 5.3.1.

The six coupons cut from the webs were 8 in (203 mm) long and had a 2.000 +

0.005 in (50.80 + 0.13 mm) gage length. Based on the original thicknesses of

the flanges of the column samples, the six coupons cut from the flanges were

standard 18 -in (457 -mm) long except for specimen (21-3A-1). There was insufficient

parent material in flange 21-3A from which to extract an 18 -in long specimen.

Specimen 21-3A-1 was a standard 8 -in coupon. All of the flange coupons had a gage length of 2.000 + 0.005 in (50.80 + 0.13 mm).

Specimens were tested to failure in tension according to ASTM procedures in one of two testing machines. A 60,000 lbf (267 kN) capacity Baldwin testing

machine, calibrated June 8, 1987, was used for the 8 -in long coupons. A

400,000 lbf (1779 kN) capacity Tinius-Olsen testing machine, calibrated June

9, 1987, was used to test the 18 -in coupons. Estimated load accuracy was + 1

percent. On the Baldwin machine the loading rate was maintained at 5000 lb/min

(22 kN/min) up to the yield point. On the Tinius-Olsen machine the crosshead

speed was maintained at 0.02 in (0.5 mm) per minute up to the yield point.

After reaching the yield strengths of the coupons, the crosshead speeds were

adjusted to fall in the range of 0.15 to 0.25 in (3.8 to 6.4 mm) per minute.

A static value for the yield stress, a lower bound value, was also obtained by

reducing the crosshead speed to zero at several points along the yield

plateau.

The elongations of the 8 -in and 18 -in coupons were measured using a modified

Tinius-Olsen LVDT extensometer attached to the reduced section. As a result of

the modification, the original AC LVDT was replaced with a DC LVDT, thereby

providing for the plotting of load- deformation curves on an X-Y recorder. The

extensometer was calibrated on May 21, 1987 and had a gage length of 2.000 in

(50.80 mm)

.

5.3.2 Results of Tests

The static yield stress, ultimate tensile strength, elastic modulus, and

percent elongation were calculated for each coupon. The individual test

results are summarized in table 5.3.2, the specimens being grouped according

to the column segment from which they were obtained. Combining the coupons

from column segments 21-1 and 21-3, the results are: 1) the average static

yield stress was 38.4 ksi (265 MPa); 2) the average ultimate tensile strength

was 67.9 ksi (468 MPa); 3) the standard deviation was 1.16 and 0.73 ksi (8.0

and 5.0 MPa) for the static yield stress and ultimate tensile strength

respectively; 4) the average percentage elongation in a 2 -in gage length was

38.2, 5) the average percentage elongation in an 8-in gage length was 29.4; and

6) the average value for the elastic modulus was 29.8 ksi (205 MPa). The

average static yield stress and ultimate tensile strength of column segment

21-2 were 52.9 ksi (365 MPa) and 79.3 ksi (547 MPa). The corresponding

standard deviations were 1.29 ksi (8.9 MPa) and 0.91 (6.3 MPa); respectively.

The average value for the elastic modulus was 29.2 ksi (201 MPa). The average

84
percentage elongation in a 2- in gage length was 39.1. The single percentage

elongation recorded for an 8 -in gage length was 30.6.

As can be confirmed from table 5.3.2, there was no significant difference

between the mechanical properties of the coupons obtained from the webs and

those obtained from the flanges. Moreover, the longitudinal and transverse

web coupons yielded tensile properties that were not significantly different.

With the exception of elongation measurements, the tensile property measurements

do not indicate any coupon size effects.

5.3.3 Comparison of Results with ASTM Standards and Mill Tests

The project specifications [5] stipulated that the column steel conform to the

requirements for ASTM A 36 or A 572, Grade 50 structural steel. The pertinent

tensile properties for these two ASTM standards are listed in table 5.3.3 and

serve as the basis of comparison for the coupon test results. It was concluded

that all of the mechanical properties measured satisfied the minimum ASTM

requirements for structural steel.

Mechanical properties obtained from the steel supplier are summarized in table

5.3.4. The yield point, ultimate tensile strength, and percentage elongation

in an 8 -in gage length are listed for some of the structural shapes used at

the construction site. The American Institute of Steel Construction (AISC)

designation is given in the first column. In the second column are shown the

heat numbers obtained from the mill report forms. Adjacent to each heat

number, in parenthesis, is shown the applicable ASTM standard. Based on the

NBS measurements of the cross-sections of Column Segments 21-1, 21-2, and 21-

3, it was concluded that all three specimens were obtained from a W12x72

column. Based on the project column schedule, the column segment conforming

to ASTM A 572/50 was probably erected during the stage III operations. The

column schedule indicates the use of W12x72 columns conforming to ASTM A 36 in

stages IV and V. The tensile properties shown in table 5.3.4 adjacent to the

W12x72 entry serve as a second basis of comparison for the NBS test results.

The average values for yield stress in the NBS tests are lower than those

listed for the mill tests. This is expected because the NBS results indicate

the average static yield stress, which tends to be lower than the yield stress

reported in the mill test reports. The average value for the ultimate tensile

strength in the NBS tests was higher for the A 36 steel and lower for the A

572 steel than the corresponding mill test results. The percentage elongation

measured after the NBS tests was higher than that reported in mill tests for both

grades of steel.

Results of the NBS analysis of the chemical composition of the column steel are presented in table 5.3.5. These results indicate that the column steel and

the weld block attached to the column satisfied ASTM A 36.

5.4 TESTS OF WELDMENTS

5.4.1 Sampling and Testing Procedure

For the first part of the weldment test series, flat coupons were machined

from three column segments containing field splices. On two of the segments,

85
columns with the same cross-sectional dimensions were joined together. The

third segment contained a transition joint from a thick flange section to a thin flange section. Joint preparation was also a variable in that two of the

specimens contained column extensions whose lower ends were double -beveled

while the third specimen contained a column extension with a single -beveled

end. Fifteen flat transverse -weld coupons were machined in accordance with

ASTM E 8 -85b from three column segments identified as NBS 26, 27, and 28.

Column segment 26 was cut from a splice between two W12xl36 members. According

to the project column schedule, such a splice occurred between stages II and

III and between stages III and IV. Column segment 27 includes a W12xl20 upper

column and a W12xl52 lower column. This splice configuration existed only

between stages I and II. Column segment 28 consisted of portions of two

W12x65 members and was cut from a column marked "12B." This splice configuration

existed between stages I and II and between stages II and III. Based on the

project column schedule, it is concluded that column segments 26, 27 and 28

consisted of ASTM A 572, Grade 50 steel.

A description of the coupons is given in table 5.4.1. Five coupons were

machined from each segment, two from each flange and one from each web. All

coupons were centered on the weld lines. The specimens were prepared by first

milling the flat surfaces of the rectangular strips to obtain uniform thicknesses

and then grinding the surfaces to remove milling marks and to ensure that the

faces were parallel. The sketches in figure 5.4.1 illustrate the orientation

and relative locations of the flange and web coupons. The bottom sketch in

figure 5.4.1 also shows the weld (single bevel) orientation on the coupons

obtained from the second flange cut. The bevel joint was across the width of

the second-cut coupons in contrast to being across the thickness for the

first-cut coupons. The three web coupons and the six first-cut flange coupons

were 18 in (457 mm) long, 1.500 in (38.10 mm) wide (except for Specimen 28W1)

at the reduced section and had a 2.000 in (50.80 mm) gage length. The second

coupons cut from column segments 26 and 27 were 8 in (203 mm) long, 0.500 in

(12.70 mm) wide at the reduced section and had a gage length of 2.000 in

(50.80 mm). The remaining two coupons, obtained from column segment 28,

conformed to the ASTM subsize specimen specifications and measured 4 in (102

mm) long, 0.250 in (6.35 mm) at the reduced section and had a gage length of 1

in (25 . 40 mm)

.

For the second part of the weldment test series, one flat coupon was cut from

each of column segments 26, 27 and 28. The three coupons were 18 in (457 mm)

long and 1.5 in (38.10 mm) wide. In preparing these coupons, only the side

surfaces were machined. The face surfaces were left unmachined, thereby

giving each coupon the same thickness as the flange of the column segments

from which it was cut. None of the weld metal was machined off as in the

first part of the weldment test series. The coupons were labeled 26F5, 27F5

and 28F5. The average width in the reduced section and the average thickness

of each coupon are presented in table 5.4.1.

Specimens were tested to failure in tension in one of two testing machines.

The 60,000 lbf (267 kN) capacity Baldwin testing machine described in section

5.3.1 was used for the 4- in and 8 -in long coupons. The 400,000 lbf (1779 kN)

capacity Tinius-Olsen testing machine described in section 5.3.1 was used for

the 18 -in long coupons.

86
The elongations of the 8 -in and 18 -in coupons were measured using a modified

Tinius -Olsen LVDT extensometer attached to the reduced section. For the 4- in

coupons, strain gages were attached to the two flat surfaces. In addition, a dial gage with a least reading of 0.001 in (0.025 mm) was attached to the lower

crosshead of the testing machine to measure the elongation.

5.4.2 Results of Strength Tests

The individual test results are summarized in table 5.4.2. Only the ultimate

tensile strength is presented in the table because the transverse butt weld

tests do not generally provide reliable information regarding yield strength

and elongation, particularly when the tests indicate that the weld metal

strength is less than that of the base metal. In this instance, most of the

strain is localized in the weld region rather than over the entire gage

length. The elongation measurements will therefore be lower than if the

strain had been uniform over the gage length as is implied in elongation results

for uniform base metal. Likewise for yield strength measurements of base

metal, implicit in the standard specifications is that the yield strength

values assume uniform strain within the specified gage length. Consistent

with expected behavior, the load-elongation plots obtained with the X-Y

recorder generally did not indicate a yield plateau.

Of the fifteen tests run on machined coupons, only specimen 27W1 failed in the

base metal. The other fourteen coupons failed due to fractures in the weld

metal

.

Observations of the fracture surfaces of specimens 26F1, 26F3 and 27F3

revealed porosity and incomplete joint penetration. All three fracture

surfaces showed evidence of incomplete penetration; i.e., the two column

sections were not completely fused together over the entire face of the

groove. The portions of the joints that were not fused were visible as dark

planar regions on the fracture surfaces. The dark planar regions consisted of

the original column ends (identifiable by parallel lines with a spacing of

about 3 to 4 per mm that appear to be artifacts produced when the columns were

cut to length) and weld metal that flowed into the narrow groove but did not

have sufficient superheat to fuse to the base metal. These planar regions were

darker on the micrographs because they were coated with oxide, while the

remainder of the fracture surface was newly fractured metal. These planar

(unfused) regions comprise roughly 15 percent of the cross-sectional area of

27F3 , 60 percent of 26F1 and 50 percent of 26F3. The amount of unfused region

in these three specimens is greater than the value considered acceptable by

the workmanship standards of American Welding Society Specification AWS Dl.l83 "Structural Welding Code - Steel" [26], a commonly used guide to the design

and construction of welded steel structures in the U.S. Reports of nondestructive

evaluation tests of the welds by a commercial testing laboratory did not identify

any flaws in the column splices. These unfused regions help explain why the

tensile strengths for these welds in table 5.4.2 are less than the values

expected for the weld metal (near 75 ksi) . The fracture surface of 27F3

contained an amount of porosity comprising as much as 3 percent of the

fracture surface area and brittle cleavage fracture comprising about 10

percent of the fracture surface area.

87
With the exception of the results for Specimen 27F4, the ultimate tensile

strength results for coupons cut from column segment 27 compare favorably with

the minimum specified ultimate tensile strength for ASTM A 572, Grade 50 steel

(refer to table 5.3.3). The ultimate tensile strength for specimen 27W1 (81.1

ksi) compares favorably with the average strength of the column steel in

column segment 21-2 (refer to table 5.3.2), which was also concluded to be A

572, Grade 50 steel.

The five ultimate tensile strength results presented in table 5.4.2 for column

segment 26 were all considerably lower than the minimum specified ultimate

tensile strength of 65 ksi (448 MPa) for A 572, Grade 50 steel.

The results for coupons cut from column segment 28 are also compared to the minimum

ultimate tensile strength of 65 ksi (448 MPa) for ASTM A 572, Grade 50 steel.

Specimen 28F3 failed at an ultimate tensile strength that exceeded the minimum

requirements for A 572, Grade 50 steel, although it failed in the weld joint.

The ultimate tensile strength of specimens 28F1, 28F2, 28F4 and 28W1 was lower

than the minimal requirements

.

The results from testing the three unmachined coupons are presented in the

bottom portion of table 5.4.2. Specimen 27F5 failed by fracture in the

thinner base metal portion. The specimen exhibited a necking down in the

reduced section prior to failure and the fracture was a symmetrical cup -cone one with a fibrous texture. Both specimens 26F5 and 28F5 failed due to

fracture in the weld joint. Their failures were less ductile than that of

specimen 27F5. The ultimate tensile strength of 27F5 exceeded the minimum required

tensile strength of 65 ksi (448 MPa) for ASTM grade 50 steel, and the tensile

strength of 26F5 fell below the required minimum. These results are consistent

with those obtained from the machined coupons. As was generally the case for

the machined coupons, specimen 28F5 had an ultimate tensile strength which

exceeded the requirements for A 36 steel, although the fracture occurred in the

weld joint.

5.4.3 Metallography Results

Four specimens were obtained from weld block/column sample NBS 21-3A shown in

figure 5.4.2. A metallographic examination of the welds confirmed evaluations

made on the basis of visual examinations. There was an apparent region of

lack of fusion in specimen NBS 21-3A-2 as shown in figure 5.4.3. There were

several flaws in the weld of specimen NBS-21-3A-3, including lack of penetration,

slag inclusion, and a void. These flaws can be seen in figure 5.4.4. Some

lack of penetration was also noted in specimen NBS 21-3A-1.

The results of Knoop microhardness measurements on the 4 specimens from sample

NBS-21-3A are given in table 5.4.3. In each case, the hardness of the weld is

significantly greater than that of either the column flange or the weld block.

What appeared to be the harder regions of the heat affected zones (HAZ's) were

either slightly harder than the weld material or were not as hard as the weld

material

.

Several column butt weld joints were examined. The location of the section

from sample NBS 8-T1 is shown in figure 5.4.5. An etched cross section

88
showing the weld appears in figure 5.4.6. At the upper part of the joint as shown in the figure, the weld is displaced laterally from the joint. The

extent of the unwelded part of the joint is shown at higher magnification in

figure 5.4.7. The unwelded joint may have extended as a crack into the weld.

The microstructure here consists primarily of granular bainite. There was incomplete penetration of the weld in the other column welded joints examined

as well. The joint in specimen NBS 6 -FI is shown in figure 5.4.8. There are both voids and lack of penetration in this weld. There may also be some slag

inclusions. The weld in sample NBS 6-W1 exhibited a lack of penetration near

the center, and a crack in the weld emanated from the region of no penetration.

A cross section through this weld is shown in figure 5.4.9. A lack of

penetration was also found in sample NBS 4-F1, as shown in figure 5.4.10. The

weld in sample NBS 4-W1, shown in figure 5.4.11, exhibits lack of penetration,

lack of fusion, and cracking through the weld metal. There was a small region

of no penetration in the weld in specimen NBS 8 -FI. Sections through specimens

5-W1, 5 -FI and 8-W1 were also examined. No significant lack of penetration,

lack of fusion or inclusions /voids were detected in these samples. Some

martensite was found in the HAZ associated with the last weld pass in specimen

NBS 4-F1

.

The chemical composition of the weld and both column components of sample 4-

F1 satisfied the requirements of ASTM A 36. The results of the chemical

analyses are given in table 5.4.4. Knoop microhardness measurements were made

on most of the specimens that were examined metallographically . The results

of these measurements are given in table 5.4.5. The hardness measurements

indicated a hard spot in the HAZ of specimen 4-F1, where an HK500 value of 553

was measured.

5.4.4 Fractography Results

Fractography was performed on two specimens, NBS 33B and NBS 33A. Specimen

33B shown in figure 5.4.12 was a wedge found loose in the debris after the

collapse. The wedge had been fully welded to the shearhead, weld block, and

column flange prior to failure. Specimen 33A, shown in figure 5.4.13, was a wedge found loose in the debris. The specimens were sectioned to facilitate

macroscopic inspection and microscopic inspection in a scanning electron

microscope (SEM)

.

Macroscopic examination of the wedge, specimen 33B, revealed substantial

deformation. Both the point and heel of the wedge were bent downward (toward

the ground when oriented in the structure) , indicating that substantial plastic

deformation occurred during failure. There were angled gouges on the surface

of the wedge where it contacted the weld block. These gouges indicated the

wedge slid in both the wedging direction (along the flange surface on the

column) and normal to the flange (away from the flange surface) during

failure. The ratio of the movement is about 3 times in the wedging direction

to that normal to the flange. There were indications that the wedge also

rotated about the edge of the weld block as it failed.

The appearance of the welds joining the wedge to the column flange, weld

block, and shearhead were all consistent with the use of the shielded metal

arc welding process. The compositional analysis in Section 5.4.3 was consistent

89
with the use of American Welding Society type E7018 electrode, the shielded

metal arc (SMA) welding electrode used for field fabrication. These fillet

welds were of poor quality, showing evidence of undesirable features such as arc strikes, undercut, overlap, excessive convexity, and general unevenness of

the bead profiles. The undercutting of the bead is evident in figures 5.4.14

and 5.4.15, cross sections through the welds at the locations shown in figure

5.4.12.

The project specifications [5] required visual inspection of all shop and

field welds. The specifications did not describe the nature of the visual

inspection, the specification they were to satisfy, or the way any deficiencies

were to be repaired. In view of this, in the NBS investigation the welds were

compared to AWS Dl.1-83. This specification states that all arc strikes

outside the area of permanent welds are to be removed (paragraph 3.10) and .the

convexity of the weld bead is not to exceed 0.07 inch times the actual face

width, plus 0.06 inch (1.52 mm) (paragraph 3.6.1). The arc strikes and convexity

observed on the wedge exceeded these levels in several areas.

The welds on the wedge fractured primarily on a 45 -degree angle from the root

to the center of the face of the welds, the expected failure mode for fillet

welds. The fracture surfaces were examined in the SEM and found to consist of

fine dimples. The dimples indicate the failure should be classified as a ductile transgranular mode; however, the small depth of the dimples indicates

a relatively low fracture energy. A low fracture energy correlates with the

smoothness of the fracture surface when examined with the naked eye.

Macroscopic examination of specimen 33A (figure 5.4.13), the wedge joined to

the weld block, revealed deformation of the wedge ends similar to that

observed in specimen 33B. Both the narrow and wide ends of the wedge were

bent downward, but there were no angled gouges on the surface. This is attributed

to different failure modes for the two locations. The weld fracture surfaces

on specimen 33A have ridges aligned along the column axis indicating the weld

block and wedge failed by shearing from the column. In contrast for 33B the

gouges are consistent with expulsion of the wedge from between the shearhead and

the weld block.

The location of the fracture plane in the fillet welds was also different in

specimen 33A. About 60 percent of the fillet weld joining the weld block to

the column failed along the 45 -degree angle from the root to the center of the

face of the welds. The other 40 percent of the weld failed near the interface

between the weld and the column. The precise failure location, whether at the

weld fusion line or within the weld metal, could not be uniquely identified

because the matching surface on the column was not available. Failure along

the interface is unusual in fillet welds because the fracture path tends to be

shorter at some angle to the interface (usually near 45 degrees) . Failure at

the interface indicates a lower energy fracture path here. Testimony obtained

by the Connecticut Department of Public Safety from the steel fabricator indicated

the welds joining the weld block to the column used a different electrode

(E70T-4) and process (flux cored arc welding-FCA) than those used to join the

wedges to the weld block and shearhead. These differences explain the difference

in crack path. A cross section through one of the welds that failed along the

weld- column interface is shown in figure 5.4.16. The weld has substantial

90
convexity and exceeds the limits of AWS Dl.1-83 paragraph 3,6.1. Micrographs

of the fracture showed a smooth surface which suggested a low-energy fracture.

The surface was essentially free of the scratches that would have occurred if

the weld block had slid down the column. This indicated that the weld block

stayed free of the column surface after the initial shear failure.

The welds that sheared near the weld-column interface had a macroscopic

texture similar to that of a weld fractured at the weld fusion line, the

interface between the weld and the base metal heat-affected zone. Fracture

along a weld fusion surface is unusual and indicates the presence of a low

toughness fracture path, which would be consistent with the very smooth

fracture surfaces

.

The SMA welds in specimen 33A joining the weld block to the wedge had a smoother profile with less undercut and overlap than those on specimen 33B.

The better appearance of these welds may explain why the wedge did not

separate from the weld block at this location.

In summary, the SMA welds joining the wedge to the weld block and shearhead

varied in quality between specimens 33B and 33A. Some of these welds had

undesirable features such as arc strikes, undercut, overlap, excessive

convexity, and general unevenness of the bead profiles. The FCA welds joining

the weld block to the column in specimen 33A exhibited fracture near the weldcolumn interface. Both the field (SMA) and shop (FCA) welds did not meet some

of the requirements of the commonly used welding code AWS Dl.1-83. The

smoothness of the fracture surfaces indicated a low energy fracture.

5.5 TESTS OF COMPONENTS

5.5.1 Introduction

Tests were conducted to: 1) determine the tensile strength of the posttensioning strand, 2) determine the tensile strength of the smaller jack rods

and evaluate fractures observed in these rods, and 3) evaluate the performance

and strength of the shearhead, wedge, weld block, and seal block assembly.

5.5.2 Tests of Post- tensioning Strands

5. 5. 2.1 Sampling and Testing Procedure

Two series of tension tests, four tests per series, were conducted on specimens

of the post- tensioning strands. The first series was run on 50- in (1.27-m)

long segments of unused strand obtained from the city landfill. The second

series was run on 50 -in long segments of strands taken from debris at the

construction site. The objective of the tests was to determine the degree of

conformance of the specimens to the project specifications for prestressing

steel. The specifications called for the use of nominal 1/2-in (12.7-mm),

Grade 270, seven-wire, Low-Lax strand satisfying ASTM A 416 ("Standard

Specification for Uncoated Seven-Wire Stress -Relieved Steel Strand for

Prestressed Concrete"). ASTM A 416 specifies a minimum breaking strength of

41,300 lbf (184 kN) and minimum yield strength of 37,170 lbf (165 kN) (i.e. 90

percent of the specified minimum breaking strength) for such strand.

91
The tests were conducted in accordance with Supplement VII ("Method of Testing

Uncoated Seven-Wire Stress -Relieved Strand for Prestressed Concrete") of

ASTM A 370-86 ("Standard Test Methods and Definitions for Mechanical Testing of

Steel Products"). Standard V-Grips and special strand holders were used in a 400,000 Ibf (1779 kN) Tinius -Olsen testing machine to grip the specimens.

Between each set of V-Grips was placed a 5- in (127-mm) long aluminum alloy

block, split lengthwise. The block was prepared by drilling a 1/2 -in (12.7-

mm) diameter hole lengthwise through a 2-in (50.8-mm) square block of aluminum

,

and then sawing the block in half lengthwise. Thus, a semi-circular trough in

the block conformed to the perimeter of the strands. Bearing against the top

of the upper block and the bottom of the lower block was a 4- in (101. 6 -mm)

long commercial post- tensioning strand chuck. The chucks served to anchor the

strand against the gripping blocks.

The elongations of the strands were measured with two instruments. An

extensometer attached to the center of the strand specimen measured elongation

up to the yield strength at which point it was removed. A deflectometer was attached to the lower crosshead to measure crosshead separation throughout the

test. The DC output of the LVDT extensometer was recorded by a XY recorder.

The AC output of the deflectometer was recorded by a drum recorder incorporated

in the testing machine console. In addition, the crosshead separation

occurring from the time of extensometer removal to the point of fracture of

the strand was measured by a steel tape. The LVDT extensometer had a 2. 000 -in

(50.80-mm) gage length and was verified to comply with the requirements of a Class B-l extensometer as specified in ASTM E 83 ("Standard Practice for

Verification and Classification of Extensometers" ) . The extensometer was attached to the strand specimens after the application of a preload of 4,100

Ibf (18.2 kN) as specified in ASTM A 370. After making the specified adjustment

in the extensometer reading, the load was increased until the extensometer

indicated an extension of 1 percent (i.e. 0.020 in or 0.51 mm). The load

concurrent with this specified extension was recorded as the yield strength.

After removal of the extensometer , the load was increased until the first wire

fractured. The load at first wire fracture is defined as the breaking

strength of the strand.

5. 5. 2.

2

Results of Tests

The yield strength, breaking strength, elongation at first wire break and

location of the fracture are presented in table 5.5.1 for each of the tensile

tests. The first four tests listed in the table constitute Series 1. The

average yield strength for Series 1 was 38,200 Ibf (170 kN) and the standard

deviation was 693 Ibf (3 kN) . The average breaking strength was 41,075 Ibf

(183 kN) and the standard deviation was 690 Ibf (3 kN) . The average elongation

at first wire fracture was 2.78 percent with a standard deviation of 0.86

percent. All of the Series 1 strands fractured within the area of one of the

aluminum blocks.

The last four tests listed in the table comprised Series 2. The average yield

strength and breaking strength of the Series 2 strands were 39,325 Ibf (175

kN) and 42,213 Ibf (188 kN) , respectively. The corresponding standard deviations

were 442 Ibf (2 kN) and 399 Ibf (2 kN) . The average elongation at first wire

fracture was 5.48 percent with a standard deviation of 1.0 percent. The first

92
two strands in Series 2 fractured within the chuck area, although with no apparent

reduction in the breaking strengths of the specimens and the fractures were

sufficiently ductile. After the fracture of specimen 6, it was observed that

the surfaces of the aluminum blocks that encased the strands had been indented

by the harder strand material, thereby diminishing the gripping ability of the

aluminum blocks. To offset the wearing effect of the strands, a mixture of

carborundum and o?l was pasted on the contact surfaces of the aluminum blocks.

This technique was apparently effective in that the remaining two strands had

fractures either within the aluminum block area or between the blocks, rather

than in the chuck area.

5. 5. 2. 3 Comparison of Results with ASTM A416

As shown in table 5.5.1, all eight of the yield strength values exceeded the

minimum value of 37170 lbf (165 kN) specified by ASTM A 416. The table also

indicates that in Series 1 only one of the strand specimens had a breaking strength

above the specified minimum (41,300 lbf-184 kN) . The average breaking strength

(41,075 lbf -183 kN) is slightly (less than 1 percent) less than the specified

minimum. For the Series 2 specimens, all four of the strand specimens had breaking

strengths in excess of the specified minimum. ASTM A 416 specifies that the

total elongation beginning from the point of application of the preload to the

point of failure shall not be less than 3.5 percent measured over a gage

length of at least 24 in (610 mm) . Establishing the gage length as the

crosshead separation at the point of extensometer removal, the test gage

lengths ranged from 27 1/16 in (687 mm) to 28 7/8 in (733 mm) . Comparing the

results in table 5.5.1 with the A 416 minimum requirements, it is observed

that two of the elongations measured for the Series 1 strands fell significantly

below the specified minimum, while all four of the Series 2 strand specimens

exceeded the minimum requirements by at least 25 percent.

5.5.3. Tests of Jack Rods

5 . 5 . 3 .

1

Sampling and Testing Procedure

Four 40-in (1.02 m) long segments of the 1 3/4-in (44.5 mm) diameter Acmethreaded jack rods used with the standard jacks were tested to failure in

tension. The test specimens were obtained from the debris at the landfill.

Although the rated capacity of this proprietary component was not known, the

capacity should have satisfied the recommendations given by the National

Safety Council of a safety factor of 2.5 [22]. The objective of the tests was

to compare the ultimate breaking strength of the rod specimens to this

recommended capacity, [ 150 , 000/2 ]x2 . 5 - 187,500 lbf (834 kN) of the rods. The

rods were tested in a 400,000 lbf (1779 kN) capacity Tinius -Olsen testing

machine. The ends of each specimen were held with the same hardware used

during the erection procedure. One end was held by a swing nut, a nut

retainer and a bearing block with a spherical seat. The other end was held by

an anchorage consisting of a lifting nut, a cylindrical sleeve and a rectangular

block. These two end assemblies bore against the upper and lower crossheads

of the testing machine. A sketch of the test configuration is presented in

figure 5.5.1.

93
The total elongation of a specimen between the end assemblies was measured by

a deflectometer held in contact with the lower crosshead of the testing

machine. The deflectometer had a full-scale range of 4 in (101.6 mm). The

rod specimens were loaded monotonically to failure. The tensile load and

crosshead separation were recorded on the drum recorder attached to the

testing machine console.

In addition to testing the full-scale rod specimens, a standard round tensile

coupon, machined from a segment of a fractured jack rod recovered from the

debris at the landfill was tested in accordance with ASTM E 8 -85b ("Standard

Methods of Tension Testing of Metallic Materials"). The elongation of the 5 -in

(127 -mm) long coupon was measured with a Tinius -Olsen LVDT extensometer with a gage length of 2.000 in (50.8 mm). Results from metallographic analyses of

this fractured rod are presented in Section 5 . 5 . 3 . 3

.

5. 5. 3.

2

Results of Tests

The results of the tension tests of the four rod specimens are presented in

table 5.5.2. The average failure load was 185,500 lbf (825 kN) , which

translates to an average ultimate tensile strength of 102.4 ksi (706 MPa).

The ultimate tensile strength was computed by dividing the failure load by the

net area of the threaded rod [7] (1.811 in^-1168 mm^) . The standard deviation

for the failure load is 4,143 lb (18 kN) which yields a coefficient of variation

of 2.23 percent. The average failure load exceeded the rated load of 75 kips

(334 kN) by of factor of 2.5.

The load- elongation plot for rod specimen 1 is presented in figure 5.5.2. The

failure load was 190,000 lbf (845 kN) and the elongation measured over the 40

in gage length was 3.44 in. The percent elongation for each specimen is given

in table 5.5.2. The total elongations for specimens 3 and 4 were obtained by

matching the fracture surfaces of the two sections and measuring the distance

between pre-test gage marks with a metal tape. This procedure was followed

because the deflectometer was damaged after the test of specimen 2.

All four of the jack rod specimens fractured outside the holding nuts.

Specimens 1, 2 and 4 did not neck down near the fracture surface. Specimen 3 necked down slightly near its center. It is noted that specimen 4 was tested

after having been used in one of the lifting assembly tests described in

Section 5.5.5. The rod had failed in a ductile manner in combined tension and

bending in the lifting assembly test.

The tensile coupon failed at an ultimate tensile strength of 92.5 ksi (638 MPa),

which is lower than any of the four values obtained in the full-scale rod

tests. The elongation of the coupon in the 2 -in gage length was 17.2 percent.

5. 5. 3.

3

Metallography Results

The fracture surface of one of the rods removed from the debris was analyzed

in detail. The sample analyzed, designated NBS 11-2, is shown in figure

5.5.3. There was some mechanical damage evident on the surface of the jack

rod. The threads appeared to have been rolled rather than machined. A seam

from the threading operation was evident at the crest and root of each thread

94
along the entire length of the sample. The seam was particularly evident on

the side of the rod that had been deformed in tension in the vicinity of the

fracture. The seam can be seen in figure 5.5.4, which shows part of the rod

surface on the tension side.

The jack rod was sectioned transversely adjacent to the fracture so the

fracture could be examined with a scanning electron microscope. The fracture

surface is shown at low magnification in figure 5.5.5. The apparent fracture

origin is at the root of a thread at the darker region in the figure. The

fracture in the region adjacent to the apparent origin exhibited primarily

dimpled rupture, which indicates ductile fracture. Outside the darker region,

the fracture mode is primarily cleavage, indicating a much less ductile

fracture. A representative fractograph indicated the fracture appeared to

have occurred in overload, probably as a single event rather than as a result

of any repeated or fatigue loading.

Two longitudinal sections through the jack rod were examined metallographically

.

One of these sections was in the deformed region adjacent to the fracture and

included the fracture profile. Part of this section is shown in figure 5.5.6.

The section was taken through the apparent fracture origin which is at the

upper left in the figure. The lap seam at the top and root of the thread

adjacent to the fracture is evident in this figure. The lap seam at the root

of the first thread removed from the fracture on the tension side of the screw

is shown at higher magnification in figure 5.5.7. The etchant solution used

to prepare the fracture surfaces was a 2 percent Nital compound. The tip of

the seam appears to have a small crack emanating from it. The lap seam at the

root of the second thread away from the fracture on the tension side of the

rod is shown in figure 5.5.8. A crack appears to have grown from the base of

the lap seam at this location. The crack tip is relatively sharp indicating

possible recent crack growth. The lap seams on the compression side of the

jack rod are not as open as they are on the tension side.

A sample of the jack rod was analyzed for chemical composition. The results

are given in table 5.5.3. The material is similar to AISI 4137 grade steel.

Knoop microhardness measurements at a load of 500 grams force were also made

on a longitudinal section through the jack rod. Two traverses were made, each

starting at the thread top and continuing across the diameter of the rod. The

results are given in table 5.5.4. The average hardness was 252 HK5 qo- There

appeared to be no hardness gradient across the jack rod.

5.5.4 Tests of Shearheads

Tests were conducted to develop information on the behavior of the shearheadwedge assembly under conditions simulating those that existed when the floor

slabs were parked in a temporary position. The test setup is shown in figure

5.5.9. The shearhead, column and wedges were samples obtained from the debris

at the landfill. The column was from an unused section, the shearhead had

been used but there was no visible damage due to the collapse. A plate was welded to the web of each channel under the loading points to prevent web

crippling. The shearhead was positioned off the center of the column; i.e.,

clearance between the shearhead and column was provided on one side only. This

95
produced the maximum eccentricity of the shearhead on one wedge and simulated

the worst case that could occur in the field.

Two tests were conducted. In the first test the wedges were tack welded to

the column. A 3/8 in (10 mm) tack weld was placed on the point of the wedge

and a 1 in (25 mm) tack weld on the heel. In the second test, no tack welds were used.

Load was applied to the shearhead using a Bliss 12 million pound universal

testing machine. Load resolution was 300 pounds (1.3 kN) on the 600 kip (2669

kN) load range used for the tests. A loading rate of approximately 2 kips/min

(8.9 kN/min) was used. Load was applied in increments of 10 kips (44 kN) . The behavior of the specimen was noted after each load increment.

For the tack welded specimen, the first perceptible deformation occurred at

120 kips (534 kN) ; a very slight rotation of the wedge with the maximum

eccentric load was observed. At 250 kips (1112 kN) the tack welds failed.

The maximum load applied to this specimen was 300 kips (1334 kN) . There was

no apparent distress of the specimen at this load. The test was stopped at

this point in order to retest the specimen after grinding the fractured tack

welds from the wedges.

The maximum load applied to the assembly without tack welds was also 300 kips

(1334 kN) . There was also no distress of the specimen at this load. Due to

irregularities of flame cut edges of the wedge, the face of the wedge was not

in full contact with the column flange at the beginning of the test. The very

small clearance between the wedge and the column did not increase, however, up

to the maximum load. The tendency for the wedge to "seat" itself in this

position was apparently the cause of the tack weld fracture in the previous

test. Since the wedge had been clamped flush with the column flange prior to

application of the tack weld, the weld had to fracture before the wedge could

seat itself.

The lack of distress in the channels of the shearhead, the wedge and the weld

block in these tests did not correspond with observations of the debris.

Considerable deformations of the wedge and shearhead occurred in the collapse

of the structure. It was therefore decided to run an additional test in which

no plate was welded to the web of the channel under the loading points. In

this third test the shearhead was positioned in the center of the column;

i.e., it was loaded symmetrically with each wedge carrying half of the applied

load. The wedges were tack welded. Other details of the test set up (figure

5.5.9) and the loading procedure were the same as in the first two tests.

The maximum load in this third test was also 300 kips (1334 kN) ; however the

behavior was considerably different. At a load of approximately 100 kips (445

kN) the tack welds on one wedge broke. As the load increased, the wedge

rotated about the outer edge of the weld block; i.e., the wedge separated from

the column. This rotation increased and the wedge deformed considerably as the 300 kip (1334 kN) load was applied. The test was stopped at this point

because the specimen was not carrying additional load, but the channel and

wedge were simply continuing to deform at this relatively constant load.

96
Deformations of the wedge and weld block are shown in figures 5.5.10 and

5.5.11. Note the substantial rounding of the edge of the weld block. The

shearhead channel also deformed substantially. The bottom flange and web

rotated and bent outward as the wedge rotated. The deformation of the

shearhead channel is shown in figure 5.5.12. These observed deformations of

the shearhead wedge and edges of the weld block were similar to those observed

in the collapsed structure. The similarity between the shearhead deformations

in this test and in the collapsed structure indicates the concrete floor slabs

offered little resistance to local deformation of the shearhead channels

produced by forces resulting from the tendency for the wedges to rotate on the

edge of the weld block.

The three tests indicated the shearhead-wedge assembly subjected to direct

shear loads was able to carry a load of 300 kips (1334 kN) with the wedge tack

welded or not tack welded.

5.5.5 Lifting Assembly

Tests were conducted to evaluate the performance and capacity of the lifting

assembly used to raise the floor slabs. The assembly consisted of the lifting

jack, jack rods and attachments, and the shearhead. The test setup is shown

schematically in figure 5.5.13.

A 150 kip (667 kN) jack recovered from the landfill was used in the test. The

shearheads and jack rods used were obtained from the debris at the landfill.

Components that exhibited a minimum amount of damage were selected for these

tests. The jack rod length of 53 in (1.35 m) from the underside of the

hydraulic jack to the bottom of the shearhead in figure 5.5.13 corresponds to

the distance from the top of the column to the underside of the floor slab at

level 9, as illustrated in figure 4.2.1, i.e., the length of jack rods being

used to position the slabs at levels 9/10/11 in the west tower at the time of

the collapse.

The lifting jack obtained from the landfill was not used to apply loads during

the test, but merely served to support the ends of the jack rods in a manner

similar to that present in the structure. Load was applied using two 100 ton

(890 kN) capacity hydraulic rams positioned between the base of the lifting

jack and a spacer that had been fabricated from structural tubing.

The initial distance between the centerline of the jack rods and the sides of

the shearhead was considered to be an important parameter in the tests. Figure

5.5.14 shows the dimensions that were varied for these tests. The gusset

plates or stiffners on the lifting angle are omitted from the figure for

clarity. The values of these dimensions used in each of the tests are shown

in table 5.5.5.

It was intended that the shearheads in the test assembly be supported uniformly

around their top perimeter to simulate the conditions at a shearhead at the

bottom of a package of slabs. Bare shearheads (not confined by concrete) were

used in three of these tests. Simulated floor loads were transferred to these

shearheads through four square steel bars positioned over the flanges of the

channel sections of the shearheads. Fiberboard was placed between the

97
shearheads and the steel bars to improve the bearing between the bars and the arms of the shearheads

.

The shearhead used for the third test was specially stiffened in an attempt to

determine the sensitivity of the failure mode to lateral restraint of the top

flange of the shearhead arm channels,, For this specimen, two 2 1/2 in x 1/2

in (63.5 mm x 12.7 mm) steel bars were welded between the tops of the legs of

the lifting angles as shown in figure 5.5.15, effectively preventing any lateral

movement of the top of the lifting angles and the top flanges of the shearhead

arm channels, as shown in figure 5.5.15. This represented an effective upper

bound of lateral and torsional restraint.

In tests 4 and 5, shearheads were confined by a concrete slab that was intended to simulate the lateral restraint conditions present in the actual

structure. No post- tensioning was used in these specimens. The concrete was

confined by steel channel sections that surrounded the specimen and by mild

steel bonded reinforcement placed around the shearhead. The configuration of

these specimens is shown in figure 5.5.16. Load was transferred to the

concrete slab through wooden timbers positioned around the perimeter of the

shearhead opening.

Because the overall mode of failure was considered the most important aspect

of these tests, little instrumentation was necessary. Small plastic scales

were attached magnetically to the specimens so motion of the lifting nuts

relative to the bottom surface of the lifting angles could be monitored

visually. In four of the five tests, linear variable differential transformers

(LVDT's) were attached to the top and bottom flanges of the arm channels of

the shearhead to measure the displacement of those flanges. All tests were

recorded on videotape.

Load was applied at a rate of approximately 5 kips (22 kN) per minute in 10

kip (44 kN) increments. After each increment of load had been applied, the

load was maintained at a constant magnitude so the shearheads and other parts

of the specimens could be examined for signs of distress.

Failure of the lifting assembly took place in one of two ways, each of which

centered around the connection of the jack rod to the shearhead. In each type

of failure, the lifting angle and the arm channel of the shearhead deformed in

response to applied loads. This deformation consisted of rotation of the

lifting angles and twisting of the arm channels of the shearheads as well as local deformations of the lifting angles near the point at which the lifting

nuts applied loads. The first type of failure occurred when the rotation of

the lifting angle allowed the lifting nut to slip out from under the lifting

angle. In the second type of failure, the rotation of the lifting angle and

arm channel of the shearhead caused the jack rod to fracture as a result of

combined flexure and axial load.

Deformation of the lifting angles was first observed when the load on the

system reached approximately 120 kips (534 kN) . At this load, mill scale

began to flake off the lifting angles. When load was increased to 130 kips

(578 kN) , the lifting angles continued to yield and to twist the arm channels

of the shearhead. This also caused the web and bottom flange of the shearhead

98
arm channel to deform inward toward the center of the shearhead as the top flange

of the channel deflected outward, or away from the center of the shearhead.

When load was increased to 140 kips (623 kN) and above, further yielding of

the lifting angles, stiffeners, and arm channels of the shearheads allowed the

end of the jack rods to move away from the side of the shearhead (toward the

center of the shearhead) . This pattern of lifting angle deformation and

accompanying jack rod movement continued until the failure load was reached.

Failure occurred when the jack rod and lifting nut slipped off the lifting

angle or when the jack rod broke. Both types of failure were very sudden and

were accompanied by a loud bang. A summary of failure loads for the lifting

assembly tests is given in table 5.5.6. Figure 5.5.17 shows a deformed lifting

angle. Deformed lifting angles were observed on several shearheads in the

collapsed structure.

Two general observations of behavior of the lifting assembly may be made on

the basis of behavior of these specimens. First, because failure of the

lifting assembly depended primarily on deformation of the arm channels of the

shearhead, confinement of the shearhead significantly affected behavior of the

arm channels and the ability of the shearhead to accept loads from the jack rods.

A second observation is that the amount of load a shearhead could accept from

a jack rod depended on the distance between the face of the lifting angle and

the center of the jack rod, a distance which may be defined as eccentricity of

loading of the lifting angle. In each test of the lifting assembly, failure

took place in the jack rod having the highest eccentricity of loading. Large

eccentricity of loading of a lifting angle produced large twisting moments on

the shearhead arm channel and lifting angle. Because failures of the jack rod

and slippage of the jack rods was directly related to twist of the lifting

angle and arm channels of the shearheads, large eccentricities of loading of

the lifting angles gave rise to failure of the entire assembly at lower loads

.

This result is significant with respect to the configuration of the lifting

assemblies used in the L' Ambiance Plaza construction. Because the lifting

nuts used with the super jacks were of slightly larger diameter than those

used with the small jacks (5 1/2 in vs 4 7/8 in-140 vs 124 mm), it is possible

that a lifting assembly employing a large jack could actually have had a somewhat smaller capacity than such an assembly employing a small jack,

assuming failure by lifting nuts slipping off in both cases, simply because of

the slightly larger eccentricity of loading inherent in the use of large jack

rods and large lifting nuts. The arm channel and lifting angles for the

shearheads used with the large capacity jacks were the same size as those on the

shearheads tested. However, • the shearheads designed for use with the super

jacks were provided with a reinforcing plate welded to the inside face of the

header channel. Typical shearheads used with small jacks and with the super

jacks are shown in figure 5.5.21 and 5.5.22.

These two observations are confirmed by the data obtained from LVDT's that

measured lateral motions of the flanges of the arm channels of the shearheads.

Plots of the relationship of flange motion to load for tests 2 through 5 are shown in figures 5.5.18 and 5.5.19. The effect of very stiff lateral confinement

at the top flange of the arm channel can be seen by comparing the plots of top

flange motion for tests 2 and 3 in figure 5.5.18. The largest failure load

99
recorded in any test was that for test 3, in which top flanges of the arm channels

were essentially prevented from moving laterally. Even though a large

eccentricity of loading was used on one side of the shearhead in test 3 (a = 7/8 in- 22 mm) as compared to the small eccentricity of loading used on both

sides of the shearhead in test 2 (a - 1/8 in-3 mm), very stiff confinement of

the top flange of the arm channel in test 3 reduced rotation of the lifting

angle in that test and allowed the jack rod to fail at a higher load than in

other tests.

The plots in figure 5.5.18 and 5.5.19, in conjunction with other data from the

tests, also illustrate the interaction between eccentricity of loading of jack

rods, shearhead confinement, and load capacity of the lifting assembly.

Figure 5.5.18 shows that the lateral deflection of the top of the arm channel

of the shearhead in test 2 was similar to that observed in tests 4 and 5. Two

things must, be recalled to put this in perspective. First, minimum eccentricity

of loading of the lifting angle was used in test 2. That the side channels of

the shearhead in test 2 twisted less than those in tests 3 and 4 is shown by

the fact that the deflection of the lower flange of the shearhead in test 2 was very small in comparison to the deflections of the shearheads used in tests 4

and 5. Second, the shearheads in tests 4 and 5 were confined by concrete,

whereas the shearhead in test 2 was unconfined. This indicates that the

concrete did not provide as much confinement as might be expected.

The support conditions for the shearheads tested here did not duplicate exactly

the support conditions that existed for the shearheads in the actual structure.

For the first three tests, the square bars used to apply load to the top of the

shearhead torsionally restrained the top flanges of the channels of the

shearhead. For tests 4 and 5, the 2x4 wooden timbers placed between the

loading plate and the top of the concrete slab applied shear but could not

simulate the moment condition which existed around the shearhead in the real

structure. The tests did, however, define the bounds of shearhead strength,

within the limits of eccentricity of loading for the lifting angles used in

the tests. The shearhead in test 1 represented a lower bound for capacity of

the lifting assembly, as the shearhead was essentially unconfined. The

capacity of the shearhead in test 3 represented an upper bound of the capacity

of the lifting assembly, as the bars welded between the top legs of the

lifting angles provided a lateral restraint much larger than could have been

provided by confinement by concrete. The tests in which the shearheads were

confined by concrete (tests 4 and 5) yielded lifting assembly capacities

between these two extremes. In addition, it can be argued that the load capacities

of the assembly indicated by these tests were good approximations of the

capacities of the lifting assemblies in the actual structure, even though the

concrete slab confining the shearheads was small relative to the dimensions of

a typical interior panel of the real floor slab and even though the test slabs

contained no prestressing tendons. The lateral stresses in the top of the

slabs used in tests 4 and 5 were zero. As an analysis of the floor slabs will

show in Chapter 7, the lateral stresses in the top of the slabs near column

locations E3 . 8 and E4.8 were also near zero. Therefore, conditions of lateral

restraint of the top flanges of the side channels of the shearheads in tests 4

and 5 could be considered approximately equal to the lateral restraint of the

comparable sections of the shearheads in the real structure at columns E3.8 and

E4.8.

100
Because they were fabricated by welding rolled shapes , nominally identical

shearheads varied slightly from one to another. It is inevitable that some Mark P50 shearheads were somewhat stiffer or more flexible than those used for

these tests. Sufficient tests were not conducted to determine statistically

the mean stiffness of these shearheads. In addition, the capacity of shearheads

other then those of the type tested could have been weaker or stronger.

Twenty- five different configurations of shearheads were used in the structure.

The sizes of the arm channels and lifting angles were uniform for all but two

of these configurations of shearheads. On the basis of the tests conducted at

NBS, the flexural and torsional flexibility of the arm channels and lifting

angles can be considered to be the primary factor influencing the stiffness

and strength of the shearhead and therefore of the lifting assembly, regardless

of the size of the jack used to lift the shearhead. This fact made it

particularly important that a shearhead of the type used in conjunction with

the large jacks (Type 1, Mark X51, Y51, or Z51) used on the project be tested.

A Mark X51 shearhead and P50 shearhead were tested to determine the relative

stiffness of the arm channels and lifting angles of the two types of shearheads.

In these tests, the shearheads were inverted and supported by steel bars

placed between their header channels and the table of a universal testing

machine. Load was applied to the lifting angles of the shearheads through a roller

placed one inch (25.4 mm) from the inner edge of the lifting angle as illustrated

schematically in figure 5.5.20. The objective of these tests was to determine

the relative flexibility of the lifting angles under load rather than to

duplicate exactly the condition of loading that existed in the structure.

A schematic representation of the test condition and the load vs. deflection

relationships determined for these two shearheads is illustrated in figure

5.5.20. As illustrated in this figure, the shearhead used in conjunction with

the super jack was slightly stiffer than the P50 shearhead. However, as noted

previously, large jack rods applied loads to the lifting angles at a slightly

larger eccentricity than did the small jack rods. With this difference in

loading taken into consideration, the effective stiffness of the two shearheads

is essentially identical. On the basis of these tests, it can be concluded

that the effective capacity of a lifting assembly which included large jacks

was not significantly greater than that of an assembly in which small jacks

were used, even though the large jacks had the ability to apply much larger

loads to the jack rods.

A final factor noticed in tests of the shearheads and evident in the plotted

data is that the arm channels of the shearheads began to twist plastically

when loads larger than 150 kips (667 kN) were applied. The rate at which

plastic twist took place was dependent on the amount of load applied. The

system of loading of the lifting assembly was designed to maintain load as deformations within the lifting assembly took place, similar to the way load

would have been applied in the real structure. However, no attempt was made

during the tests to simulate the rate of load application that would have

existed in the real structure, in which design loads were applied rapidly and

allowed to remain on the shearheads for the time it took to raise the floor

slabs as high as necessary to park them at a new location and to install

temporary wedges. No test in the laboratory took longer than two hours to

complete. The lifting of slabs 9/10/11 to their temporary position in the

101
fourth stage of lifting of the west building took several hours, and the floor

slabs remained supported by the jack rods for approximately an hour on the

23rd of April while the workmen were at lunch. If the capacity of the lifting

assembly had been marginal at any column in the west building, any delay in

parking the slabs and wedging them in place could certainly have increased the

likelihood of a failure.

The actual capacity of the lifting assembly at every column in the structure

will never be known exactly. However, the tests performed in the laboratory

serve to establish a range of possible capacities. On the basis of these tests,

and considering the variability of shearhead construction and loading conditions

discussed previously, the minimum capacity of the weakest lifting assembly

could have been expected to be approximately 170 kips (756 kN) . The largest

capacity of any lifting assembly could not have been as large as the load

measured in test 3, as the lateral confinement of that shearhead was unrealistically stiff. This would suggest a maximum capacity of approximately

200 kips (934 kN) for the lifting assembly under conditions comparable to

those in these tests.

5.6 SUMMARY

The following summarizes the results of the laboratory tests conducted by NBS

on the materials, components and subassemblies.

1. The average compressive strength of the concrete in the floor slabs

and shearwalls exceeded the value required by the project specifications.

2. There was no evidence that the concrete in the floor slabs or shearwalls

had frozen during curing.

3 . There was no evidence an admixture had weakened the concrete in the

floor slabs or shearwalls.

4. The mechanical properties and chemical composition of the steel used

in the columns satisfy the ASTM requirements for the grades of steel

in the project specifications.

5. There was evidence of inadequate joint penetration and porosity in the

welds. Some of the welds did not meet the workmanship standards of

the American Welding Society.

6. The ultimate tensile strength of some of the column splice specimens

was less than the strength of the base metal. Failure occurred in the

weld metal in most of the column splice specimens.

7. The yield strength of the post- tensioning tendons exceeded the ASTM

requirement for the steel in the project specifications.

8. Failure of the lifting assembly (jack rod/shearhead) occurred by

fracture of the jack rod or by the lifting nut sliding off the lifting

angle. Both types of failure were accompanied by a loud bang.

102
9 . Deformations of the shearhead lifting angles and scrape marks due to

the lifting nut sliding off observed in the lifting assembly tests

were observed on several shearheads in the collapsed structure.

10. The strength of the lifting assembly is affected by the confinement of

the shearhead and the eccentricity of loading of the lifting angle

.

103
TABLE 5.2.1

SUMMARY OF CONCRETE CORE SAMPLE CHARACTERISTICS

NBS Field Length, Weight

,

Density

,

Sample No. Sample No. In. Lb. Lb/Ft^ Comments

1 RS-34 6.75 6.40 148.3

2 RS-30 6.06 6.62 146.6

3 RS-25 6.88 6.37 145.0

4 RS-27 6.81 6.28 144.2 rough end

5 RS-23 6.75 6.11 141.6

6 RS-22 6.75 6.09 141.2 end sloped

7 RS-35 6.56 6.17 147.1

8 RS-29 7.00 6.68 149.3 contains rebar

9 RS-33 6.75 5.99 138.8 sides chipped

10 RS-31 6.81 6.14 141.0

11 RS-9 7.00 6.49 145.1

12 RS-28 6.75 6.35 147.2

13 RS-1 7.31 6.81 145.7

14 RS-3 6.94 6.39 144.1

15 RS-10 7.06 6.35 140.7

16 RS-2 7.00 6.32 141.3

*17 RS-4 6.63 6.60 155.9 cracks, tendon,

rebar, chair

18 RS-21 6.81 6.25 143.5

19 RS-15 6,81 6.57 150.9

20 SWS - 8 8.25 7.51 142.4

*21 SWS - 7 8.19 7.50 143.3 cracks, voids

22 SWS = 6 8.31 7.73 145.5

23 RSW-1 8.25 7.65 145.1

24 S83E 6.69 6.19 144.8

25 S12G5FL7 6.69 6.13 143.4

26 S16C2FL9 6.56 6.33 150.9 chipped

*27 S17Y5 6.5 6.00 144.4 cracked, chair

28 S5E 7.06 6.25 138.5

29 S63E 6.75 5.92 137.2 contains chair

30 S13G5FL11 7.19 6.54 142.4

31 S73E 6.81 6.40 147.0 contains chair

32 S18Y5 6.56 6.01 143.3 contains chair

33 S19Y5 6.75 6.20 143.7 contains chair

34 S11FL12 6.81 6.20 142.4 light-colored agg

35 SWS -3 12.31 11.47 146.2 large chip at end

36 SWS -2 12.31 11.60 147.4

37 SWS -5 12.69 11.72 144.5

38 SWS -4 12.19 11.41 146.5

* Specimens unsuitable for testing due to damage or presence of large amounts

of reinforcement

104
TABLE 5.2.2

SUMMARY OF CONCRETE SAMPLE TESTS RESULTS COMPRESSION TESTS

NBS Field Capped Failure Compressive Ec

,

Sample No. Sample No. Length, In. Load, Lb. Stress, psi psi x

1 RS-34 6.97 64,150 5750

2 RS-30 7.25 56,800 5140

3 RS-25 7.12 62,900 5670 3.07

4 RS-27 6.56 64,000 5680

5 RS-23 6.94 72,000 6460

6 RS-22 6.88 68,300 6120

7 RS-35 6.75 65,900 5880

8 RS-29 7.19 57,250 5170

9 RS-33 6.75 61,700 5500

10 RS - 31 7.00 56,100 5040 2.69

11 RS-9 7.13 54,600 4920

12 RS-28 7.03 66,800 6010

13 RS-1 7.50 64,000 5800

14 RS-3 7.13 56,600 5100

15 RS-10 7.31 48,900 4430

16 RS-2 7.13 59,300 5350

*17 RS-4 6.81 Specimen damaged

18 RS-21 7.00 51,400 4620 2.75

19 RS-15 6.94 73,900 6630

20 SWS-8 7.75 49,700 4500

*21 SWS-7 7.75 Specimen damaged

22 SWS-6 7.69 45,100 4080 3.48

23 RSW-1 7.75 49,700 4500

24 S83E 6.88 58,800 5260 3.11

25 S12G5FL7 6.88 59,000 5280

26 S16C2FL9 7.00 61,700 5550

*27 S17Y5 6.75 Specimen damaged

28 S5E 7.25 52,000 4710

29 S63E 5.19 59,000 5280

30 S13G5FL11 7.41 60,800 5500 2.70

31 S73E 7.06 61,300 5520 2.99

32 S18Y5 6.81 60,100 5370

33 S19Y5 6.88 63,400 5680

34 S11FL12 7.00 48,700 4380

35 SWS-3 7.75 51,600 4670

36 SWS-2 7.75 53,500 4840

37 SWS-5 7.69 53,200 4820

38 SWS-4 7.72 50,300 4550 3.26

Where appropriate, strengths have been modified by correction factors given in ASTM

C 42 to account for nonstandard length- to-diameter ratios of cores.

* Specimens unsuitable for testing due to damage or presence of large amounts of

reinforcement

.

105
TABLE 5.2.3

SUMMARY OF CONCRETE SAMPLE TEST RESULTS - SPLITTING TESTS

NBS Field Length

,

Failure Tensile

nple No

.

Sample No. In. Load, Lb. Strength,

SI RS-S24 6.81 25,100 626

S2 RS-32 6.75 19,750 497

S3 RS -20 7.06 20,350 489

S4 RS-14 7.00 26,000 631

S5 RS-26 6.94 23,700 580

Average of all cylinders : 565 psi

106
TABLE 5.2.4

COMPRESSIVE STRENGTH OF CONCRETE MEASURED BY AN INDEPENDENT TESTING LABORATORY

Mean Compressive Strength, psi (Std. Deviation, psi)

Floor Level West Building East Building

Roof 4570 (98.2) 4560 (68.9)

12 4670 (99.5) 4680 (81.9)

11 4670 (84.8) 4600 (72.4)

10 4600 (78.0) 4580 (106)

9 4560 (65.8) 4650 (74.7)

8 4550 (30. 1) 4580 (70.0)

7 4590 (99.3) 4680 (105)

6 4950 (203) 4810 (85.1)

5 4980 (82.2) 4560 (87.2)

4 4980 (120) 5290 (86.2)

3 4940 (107) 4720 (133)

2 4450 (88.6) 4650 (84.0)

1 4530 (111) 4420 (83.5)

Ground Level 4540 (75.2) 4670 (116)

C Level 4480 (121) 4380 (81.0)

D Level 4550 (57.7) 4380 (74.0)

Summary - all floors, both buildings: 4650 (216)

Calculated compressive strengths and standard deviations of strength for each

floor based on results of nine cylinders.

107
TABLE 5.3.1

DESCRIPTION OF TENSILE TEST COUPONS

Specimen Specimen Average Average

. o SDecimen Location Orientation Thickness . in Width . in Area . m— Length . in

COLUMN SEGMENT 21-1

21-1C-1 web longtd. 0.388 0.501 0 . 1944 8 21-1C-2 web transverse 0.388 0.502 0.1948 8 21-1A-1 flange longtd. 0.578 1.504 0.8693 18

21-1B-1 flange longtd. 0.578 1.504 0.8693 18

COLUMN SEGMENT 21-2

21-2C-1 web longtd. 0.388 0.501 0.1944 8 21-2C-2 web transverse 0.388 0.500 0.1940 8 21-2A-1 flange longtd. 0.578 1.503 0.8687 18

21-2B-1 flange longtd. 0.578 1.503 0.8687 18

COLUMN SEGMENT 21-3

21-3C-1 web longtd. 0.388 0.501 0.1944 8

21-3C-2 web transverse 0.388 0.502 0.1948 8 21-3A-1 flange longtd. 0.677 0.501 0.3390 8 21-3B-1 flange longtd. 0.578 / 1.504 0.8693 18

108
TABLE 5.3.2

I

TENSILE TEST RESULTS - COLUMN STEEL

Specimen Static Ultimate Elastic Percent

Specimen Location Yield Stress. ksi Tensile Strength. ksi Modulus . ksi Elongation

COLUMN SEGMENT 21-1 (ASTM A36)

21-1C-1 web 37.5 68.5 31.0 36.0(a)

21-1C-2 web 40.0 68.5 29.5 37.6(a)

21-1A-1 flange 37.0 66.5 27.5 29.0(b)

21-1B-1 flange 38.0 67.5 27.5 29.8(b)

COLUMN SEGMENT 21-2 (ASTM A572/50)

21-2C-1 web 52.5 79.0 29.0 32.8(a)

21-2C-2 web 53.0 79.5 28.5 31.4(a)

21-2A-1 flange 54.5 80.5 29.5 25.4(b)

21-2B-1 flange 51.0 78.0 25.4(b)

COLUMN SEGMENT 21-3 (ASTM A36)

21-3C-1 web 37.5 68.5 29.5 39.2(a)

21-3C-2 web 40.0 68.0 28.5 37.6(a)

21-3A-1 flange 39.0 68.5 40.6(a)

21-3B-1 flange 38.0 67.5 29.5 30.6(b)

(a) in a 2 -inch gage length

(b) in an 8 -inch gage length

TABLE 5.3.3

ASTM REQUIREMENTS FOR MECHANICAL PROPERTIES

Minimum Ultimate Minimum

ASTM Yield Strength Tensile Strength Elongation. %

S tandard

ksi MPa ksi MPa

in 8 in

or 200 mm

in 2 in or 50 mm A 36 36 248 58 - 80 400 - 552 20 23

A 572(50) 50 345 > 65 450 18 21
TABLE 5.3.4

MILL TEST RESULTS

Yield Tensile

Designation Heat No. Point ,psi Strength

HP 12x53 170J528 (A36) 37100 67000

712 (A36) 44300 72500

542 (A36) 40600 72000

530 (A36) 39600 71500

713 (A36) 38800 66500

W12x72 170J515 (A36) 39200 65500

517 (A572) 55500 86500

W12x79 170J514 (A3 6) 41400 75500

517 (A572) 55500 86500

W12x87 172H103 (A36) 41200 77500

170J513 (A572) 60500 89500

514 (A36) 41400 75500

W12x96 170J514 (A36) 41600 76000

513 (A572) 57000 86000

513 (A572) 60500 89500

W12xl06 170J125 (A36) 45400 71000

180H271 (A36) 44500 77500

170J513 (A572) 60500 89500

514 (A3 6) 40900 75500

699 (A36) 49200 69000

W12xl20 170J512 (A36) 46100 79000

513 (A572) 57000 86000

698 (A572) 60000 85000

699 (A36) 47300 77500

W12xl36 170J513 (A572) 60500 89500

513 (A572) 57000 86000

512 (A36) 46600 78000

1/ in 8 inches

172H111 (A36) 47800 78000 .

Elongation , %^

27.0

26.0

27.0

29.0

27.0

26.0

22.0

27.0

22.0

26.0

22.0

27.0

24.0

20.0

22.0

25.0

26.0

22.0

25.0

23.0

26.0

20.0

22.0

28.0

22.0

20.0

26.0

25.0

110
TABLE 5.3.5

CHEMICAL ANALYSIS OF COLUMN STEEL

("Percent')

Sample C P S Mn Si Ni Cr V Mo Cu

NBS 3A-4 1 .16 .012 .032 .79 .24 .091 .13 .001 .017 .20

3 .23 .008 .029 .75 .054 .088 .071 .001 .018 .040

TABLE 5.4.1

DESCRIPTION OF WELDMENT TENSILE TEST COUPONS

Specimen Average Average

SDecimen Location Thickness . in Width. in Area, in-4 Length . in

COLUMN SEGMENT 26

26W1 web 0.565 1.503 0.849 18

26F1 flange 0.956 1.503 1.437 18

26F2 flange 0.956 1.503 1.437 18

26F3 flange 1.443 0.498 0.719 8 26F4 flange 1.446 0.500 0.723 8 COLUMN SEGMENT 27

27W1 web 0.506 1.503 0.761 18

27F1 flange 0.956 1.509 1.442 18

27F2 flange 0.956 1.507 1.441 18

27F3 flange 1.446 0.500 0.723 8 27F4 flange 1.446 0.500 0.723 8 COLUMN SEGMENT 28

28W1 web 0.252 1.253 1/ 0.316 18

28F1 flange 0.412 1.503 0.619 18

28F2 flange 0.412 1.503 0.619 18

28F3 flange 1.462 0.248 0.362 4

28F4 flange 1.463 0.250 0.366 4

UNMACHINED COUPONS

26F5 flange 1.3342/ 1.499 2.000 18

27F5 flange 1.052 3 / 1.499 1.577 18

28F5 flange 0.722 2 / 1.499 1.082 18

V The coupon had to be machined to a lesser width than that specified in

ASTM E8 due to the> presence of a surface flaw within the reduced section

2/ Average thickness across weld joint

3/ Average thickness of thinner flange

111
TABLE 5.4.2

TENSILE TESTS RESULTS - COLUMN SPLICE WELDMENTS

Specimen Ultimate Ultimate

Specimen Location Area. in— Tensile Load.lbf Tensile Strength

- COLUMN SEGMENT 26

26F1 flange 1.437 61,500 42.8

26F2 flange 1.437 60,500 42.1

26W1 web 0.849 34,500 40.6

26F3 flange 0.719 12,000 16.7

26F4 flange 0.723 26,250 36.3

COLUMN SEGMENT 27

27F1 flange 1.442 107,000 74.2

27F2 flange 1.441 91,000 63.2

27W1 web 0.761 61,700 81.1

27F3 flange 0.723 53,050 73.4

27F4 flange 0.723 15,200 21.0

COLUMN SEGMENT 28

28F1 flange 0.619 33,400 54.8

28F2 flange 0.619 39,500 63.8

28W1 web 0.316 8,800 27.8

28F3 flange 0.362 27,050 74.7

28F4 flange 0.366 23,200 63.4

UNMACHINED COUPONS

26F5 flange 2.000 117,000 58.5

27F5 flange 1.577 131,000 83.1

28F5 flange 1.082 71,600 66.2

112
TABLE 5.4.3

KNOOP MICROHARDNESS MEASUREMENTS OF WELD BLOCK /COLUMN SAMPLE

(NBS - 21-3A)

Specimen

Location 3A1 3A2 3A3(1) 3A3 (2) 3A4 X

Plate - 181 168 195 192 185 178

& HAZ - 254 260 230 225 250 244

Hard Spot - 276 333 287 264 262

Weld - 320 307 300 285 310 304

HAZ - 244 294 244 234 263 256

Block - 202 210 168 160 212 197

TABLE 5.4.4

CHEMICAL ANALYSIS OF COLUMN JOINT WELD METAL

NBS 4 -FI

(Percent')

Sample C P S Mn Si Ni Cr V Mo Cu

Plate 1 .21 .010 .027 1.19 .034 .058 .071 .057 .011 .20

Weld 2 .097 .010 .012 1.00 .50 .040 .029 .005 <.005 .015

Plate 3 .22 .005 .016 1.25 .026 .124 .038 .058 .022 .060

TABLE 5.4.5

KNOOP MICROHARDNESS MEASUREMENTS OF COLUMN JOINT WELD SPECIMENS

fHK 500 gml

NBS 8 -FI HK500 X NBS 8-T1 HK500 X

7/16" Plate - 197 199 198 5/8" Plate - 160 165 163

& HAZ - 226 280 253 6c HAZ - 208 190 199

Weld Back - 205 202 204 Weld Back - 184 204 194

Weld V - 280 207 244 Weld V - 180 165 173

Hard spot — 320 294 Hard Spot — 294 246

HAZ - 223 226 225 HAZ - 210 206 208

7/16" Plate - 212 198 205 1/2" Plate - 166 166 166

NBS 5-W1 NBS 6 -FI

3/8" Plate 2 188 195 192 1 5/6" Plate - 210 220 215

6c HAZ - 196 203 200 6c HAZ - 280 295 288

Weld Back — 193 212 203 Weld Back — 237 360 299

Weld V - 208 200 204 Weld V - 237 240 239

Hard spot - 264 236 Hard Spot — 491 553

HAZ - 230 216 223 HAZ - 266 329 298

3/8" Plate - 169 172 170 1 5/6" Plate - 225 254 240

113
TABLE 5.5.1

TENSILE TESTS RESULTS - POST-TENSIONING STRANDS

Specimen

Number

Yield

Strength

Ibf

Breaking

Strength

Ibf

Elongation @ 1st

Wire Fracture

%

Location of

Fracture

Unused Strand from Landfill Area

17-1 38,000 40,500 1.9 Alum block area

17-2 39,200 42,000 3.7 Alum block area

17-3 37,600 40,600 2.2 Alum block area

17-4 38,000 41,200 3.3 Alum block area

S trand Obtained from Rubble at Construction Site

#5 38,900 42,150 6.8 In chuck area

#6 39,600 42,650 5.5 In chuck area

#7 a 39,800 42,350 4.4 Alum block area

#8 39,000 41,700 5.2 Between alum blocks

a/ In light of the two immediately previous strands failing in the grip(i.e.

chuck) area, a mixture of carborundum powder and oil was pasted along the

semi-circular trough of the aluminum blocks prior to testing specimen #7.

114
TABLE 5.5.2

TENSILE TEST RESULTS - JACK RODS

Specimen

Designation

Failure Ultimate Percent elongation

Load, lbf Tensile Strength, ksi in 40 in.

#1

#2

#4

#4

190.000 104.9 8.6

180.500 99.7 8.2

184.000 101.6 7.7*

187.500 103.5 8.0*

* Elongation measured by metal tape instead of by deflectometer

115
TABLE 5.5.3

Sample

NBS 11-2

C P

.38 .008

CHEMICAL ANALYSIS OF JACK ROD

(Percent’)

S Mn Si Ni Cr V Mo Cu

.007 .70 .26 .073 .83 .001 .17 .035

TABLE 5.5.4

KNOOP MICHROHARDNESS MEASUREMENTS OF JACKROD (NBS 11-2^

[TWO HK TRAVERS E ' S 500 gm/20x obi 1

TRAVERSE HK1 HK2

THREAD

TIP

260 281

1/10" 302 275

2/10" 253 250

3/10" 280 228

4/10" 248 279

5/10 269 293

6/10" 245 218

7/10" 228 267

8/10" 201 223

9/10" 201 226

X 249 254

116
Table 5.5.5

LOCATION OF JACK RODS IN LIFTING ASSEMBLY TESTS

Jack rod location dimensions - inches—

Test No. a b c d e Shearhead details

1 5/8 5/8 3/16 0 13 3/4 bare shearhead

2 1/8 1/8 1/8 1/8 14 1/2 bare shearhead

3 7/8 5/8 0 1/8 14 1/8 bare shearhead - top

flanges of lifting

angles joined by

steel bars

4 5/8 3/4 1/8 0 14 3/8 shearhead confined

by concrete

5 3/4 5/8 5/8 5/8 13 5/8 shearhead confined

by concrete

1 See Figure 5.5.18 for locations of dimensions

TABLE 5.5.6

SUMMARY OF LIFTING ASSEMBLY TEST RESULTS

Test No. Failure Load, Kips Failure Location and Type

1 165 Bottom nut slipped off lifting angle

2 196 Jack rod broke above lifting nut due to

flexure and axial load

3 227 Jack rod broke above lifting nut due to

flexure and axial load

4 201 Bottom nut slipped off lifting angle - East side

5 198 Bottom nut slipped off lifting angle - West side

117
Weld

joint

View looking at top of flange View looking at web

L = 4’ or 8'

S \

Section A-A. Specimen obtained from 2nd cut

Figure 5.4.1 Orientation and location of Weldment specimens

118
1,1

Figure 5.4.2 Weld block/column sample NBS 21-3A

I

I

I
Figure 5.4.3 Cross section showing weld in sample

NBS 21-3A-2 (x 100)

Figure 5.4.4 Cross section

NBS-3A-3 (x

showing weld in sample

4)

120
Figure 5.4.5 Location of metallographic specimen

from sample NBS 8-T1

Figure 5.4.6 Cross section showing weld in specimen

NBS 8-T1 (x 4)

121
ure 5.4.7 Cross section showing weld in specimen NBS 8-T1

at higher magnification than in Figure 5.4.6

Figure 5.4.8 Cross section showing the weld

NBS 6-F1 (x 4)

in specimen

122
[l

p

Figure 5.4.9 Cross section showing the weld in specimen

NBS 6-W1 (x 4)

/

Figure 5.4.10 Cross section showing the weld in specimen

NBS 4-F1 (x 4)

123
Figure 5.4.11 Cross section showing the weld in specimen

NBS 4-W1 (x 4)

124
Cross section 2 Cross section 1

Figure 5.4.12 Fracture of fillet weld on wedge - specimen NBS 33B

Figure 5.4.13 NBS specimen 33A, weld block/wedge

sample from debris

125
Figure 5.4.14 Metallographic cross section #1 - specimen NBS 33B

126
Figure 5.4.15 Metallographic cross section #2 - specimen NBS 33B

Figure 5.4.16 Metallographic cross section of failed weld

along weld/column interface - specimen

NBS 33

A

127
128

Figure

5.5.1

Test

setup

for

jack

rod

specimen
CM -rsd!>i ‘avoi

129

ELONGATION,

in.

Figure

5.5.2

Load-elongation

results

for

jack

rod

specimen
47 -18 44 <V> SI M t < >4 •>»• ,> ’ mH Figure 5.5.3 Jack rod sample NBS 11-2

Figure 5.5.4 Jack rod on tension side showing lap seam

from thread rolling operation (x 3)

130
2

Figure 5.5.5 Fracture surface of jack rod, sample

NBS 11-2 (x 1)

Figure 5.5.6 Fracture profile of jack rod, left, first thread

adjacent to fracture (Sample NBS 11-2) (x 9)

131
Figure 5.5.8 Lap seam at root of second thread away from

fracture of jack rod. As polished (x 40)

Figure 5.5.7 Lap seam at root of first thread removed from fracture

of jack rod (Sample NBS 11-2) (x 500)

132
133

Figure

5.5.9

Shearhead-wedge

assembly

test

setup
Figure 5.5.10 Deformation of wedges in test of shearhead

134
Figure 5.5.11 Weld block deformation in test of shearhead

135
Figure 5.5.12 Deformation of shearhead channel in test of shearhead

136
137
East

Section X-X

a, d: Clearance between lifting nut and shearhead arm channel

b, c: Clearance between jack rod spacer and lifting angle

e: Center-center spacing of jack rods

Figure 5.5.14 Critical dimensions of shearhead loading

138
Figure 5.5.15 Shearhead No. 3 stiffened by plates welded between

legs of lifting angles

139
33'

44'

r

4-#4 bars

each way

Mark P50

Shearhead

LiA

T

Figure 5.5.16 Shearhead configuration used in lifting

assembly tests 4 and 5

140
Figure 5.5.17 Shearhead No. 2 after failure showing rotation of

lifting angles

141
LOAD

(kips)

LOAD

(kips)

(a) Top flange of channel - failure side

(b) Bottom flange of channel - failure side

Figure 5.5.18 Lateral deformation of shearhead - failure side

142
LOAD

(kips)

LOAD

(kips)

(a) Top flange of channel

(b) Bottom flange of channel - unfailed side

Figure 5.5.19 Lateral deformation of shearhead - unfailed side

143
Support

under

header

channels

only

t 1

1 r

LO

o

d

cod

CM

d

o

o

co oLO

o o

co

(sd(>i) avcn

O O o

CM T"

144

DEFLECTION

UNDER

LINE

LOAD

(in)

Figure

5.5.20

Applied

load

vs.

deflection

of

lifting

angle

for

X-51,

P-50

shearheads
10

’ 10

"

Arm

channel —

7-1/2" 7-1/2’

<£ lifting rods

Header

channel

A

T

_L___ _ nz

r

tzzt

1C

Lifting angle

-z.5/8'x3-1/2’

H x5'

1 L

f"“ L

JT~TL

/Header

channel

1

co

00

CO

I in

CO

of shearhead

P50 SHEARHEAD - PLAN VIEW

Figure 5.5.21 Plan view of typical shearhead used with

smaller jacks

145
Stiffener

plate

Arm

channel

Stiffener

plate

X51 SHEARHEAD - PLAN VIEW

Figure 5.5.22 Plan view of shearhead used with super jacks

146
6 . SUBSURFACE EXPLORATION

6.1 INTRODUCTION

The purpose of the subsurface investigation reported in this chapter was to

explore in situ conditions of the footing support and basement wall backfill

in order to ascertain whether these conditions could have contributed to the

collapse

.

The plans for L' Ambiance Plaza specify the following for the tower foundations

in the general notes (Drawing S201, Note 2): "All footings shall rest on undisturbed rock. Rock shall have a minimum bearing capacity of 14,000 pounds

per square foot." Nineteen reports on the inspection of footing bottoms in the

tower area, prepared by a commercial testing laboratory between August 15 and

October 1, 1985, contain the following statement: "The material is of broken

rock and earth mixed, was placed and compacted with backhoe bucket and

vibratory compactor. The basic allowable bearing value for these locations

(identified in the report) exceeds the 7 tons job specification."

It is not clear how the commercial testing laboratory reports can be

interpreted. The reports indicate that in at least some instances the footings

were not placed directly on rock, and that a mixture of broken rock and earth

was compacted into place before pouring the concrete. However, neither the

thickness, nor the composition or degree of compaction of the material that was placed under the footings is reported. It is also not clear what criteria

were used to determine that "the basic allowable bearing value exceeds the 7

tons job specification." Since the possibility that the condition of the

footings contributed to the collapse could not be ruled out in advance, and

the available information did not clearly indicate what that condition was , it was decided to explore the condition at the base of several randomly selected

footings

.

Similarly, the general contractor's daily construction reports, starting with

Report 211, dated April 8, 1987 and continuing to the date of the collapse,

indicate that backfilling behind the basement wall on the north side of the

tower was taking place. Figure 6.1.1, which is a reproduction of detail drawing

3/S203 in the plans for L' Ambiance Plaza shows that the lateral earth pressure

acting on the basement wall is resisted by the floor slabs at levels D and C,

which bear directly against the wall. There was no evidence following the

collapse that the keyways shown in the slabs at levels D and C in figure 6.1.1

were present. A field survey of the basement wall of the east tower showed

that the wall was cracked, and that the top of the wall displaced as much as 3 in (76.2 mm) toward the tower. Further observations and telltale signs

associated with the basement wall rotation and the resulting displacements of

the backfill are discussed in Section 4.4.5. It was reasoned that the observed

wall displacements resulted from the collapse, which deprived the wall of the

support provided by the floor slabs at levels C and D. This in turn was taken

as an indication that the slabs on levels C and D and the connected shearwalls

were subjected to lateral soil pressures prior to the building collapse. The

condition of the backfill behind these walls was therefore explored in order

to estimate the magnitude of these pressures and their potential effect on the

building during construction.

147
6.2 EXPLORATION PROCEDURES

The locations of the borings and test pits that were used to explore the site

are shown in Figure 6.2.1.

The condition below the bottom of the foundations was explored by soil

borings, test pits, and in situ tests where feasible. The soil borings were

taken through the basement slab (level E) , the underlying compacted fill, the

footing, the underlying fill or fractured rock and cored into the underlying

bedrock formation. In the borings, a HW [4-in (102 mm) I.D.] casing was advanced to within 1 to 2 in (25 to 51 mm) above the bottom of the footing

(cutting through the reinforcement) . The concrete within the casing was then

removed by a 3 -in (76 mm) diameter roller bit, and samples of the fill

material were taken with a 1 3/8 -in (35 mm) I.D. split spoon sampler. The

roller bit removed the concrete to the bottom of the footing, and standard

penetration testing (SPT) was commenced within 1.5 to 2 minutes, advancing the

borehole by continuous sampling through any loose material. The SPT was performed in accordance with ASTM D 1586 "Standard Method for Penetration Test

and Split Barrel Sampling of Soils," using a CME 55, 1986 model rig with a 140

lb (63.5 kg) trip hammer dropped from a 30 in (762 mm) fall height. A split

spoon sampler [2 in (51 mm) O.D., 1 3/8 in (35 mm) I.D.] without a plastic

liner was used. This procedure was designed to minimize the effect of water

penetration on the SPT blow count obtained when loose material at the bottom

of the footing was penetrated (water had to be circulated during the concrete

coring and it was suspected that this water could affect the blow count) . Soil samples were saved for visual examination and laboratory classification

tests

.

NX double tube core barrels were used to take 3 in (76 mm) diameter cores

through decomposed or overblasted rock and 3 to 5 ft (0.9 to 1.5 m) into what

was judged to be undisturbed rock. The rock cores were saved, and core

recovery as well as rock quality designation (RQD) (16) was determined.

The test pits next to the footings were dug by a backhoe along the side of the

footing. Excavation continued to a point where no more loose material could be

removed. The pits were then pumped out if groundwater accumulated and the

condition below the footing was carefully examined. The pits were at least 2 ft (0.6 m) wide at the bottom.

A pressuremeter test was performed under footing F10 using a Roctest Model

G-AM^- pressuremeter. The specifications for, and interpretation of, the

pressuremeter tests are presented in Appendix B.

The conditions of the backfill behind the basement walls were explored by two

borings to the north of the wall, one test pit, in situ density tests and

direct shear tests in the laboratory.

Specification of commercial equipment or products does not imply

recommendation or endorsement by the National Bureau of Standards

148
The borings (B1 and B2) were advanced using a 3 1/4 in (83 mm) I.D. hollow

stem auger and SPT sampling. A plug was used during the auger advance to

prevent penetration of loose material into the core of the auger. The procedure

for SPT testing was the same as that used below the bottom of the footings

,

except that sampling was performed at 2-ft (0.6 m) intervals.

Two in situ density tests were performed in the test pit at location B2 using

the sand cone method (ASTM D 1556 "Standard Test Method for Density of Soils

in Place by the Sand Ccne Method") , and one pressuremeter test was performed

in boring B2

.

Laboratory soil classification included visual-manual classification of the

samples retrieved and laboratory soil classification in accordance with ASTM D

2487 "Standard Test Method for Classification of Soils for Engineering

Purposes" (unified soil classification)

.

Direct shear tests were performed on samples retrieved from the pit at B2

which were re-constituted to the in situ density. The tests were performed in

accordance with ASTM D 3080 "Direct Shear Test of Soils Under Consolidated

Drained Conditions .

6.3 EXPLORATION RESULTS

The results of the soil exploration are presented in detail in Appendix B and

summarized and interpreted in this section.

1. Footings

In the following discussion settlements are estimated in accordance with

references 17 and 18, which utilize empirical data that were developed for

soil and rock deposits similar to those encountered on the site. The following

findings and estimates are derived from the available subsurface exploration data:

(1) The rock cores taken below the footings indicate that, except in the

case of footing DIO, the rock underlying the footings was competent to

safely support the 7 ton/ft^ (670 kPa) design pressure. In the case of

footing DIO, the poor RQD recorded is attributed to the coring

procedure (a single tube core barrel) rather than the rock

quality. Estimated total and differential settlements of the footings

due to bedrock compression during construction were probably very

small (17), and therefore could not have contributed to the collapse.

(2) In all borings a layer of disintegrated rock, approximately 1 ft (0.3

m) thick, was encountered on top of the rock that was cored. It was difficult to confirm the presence of this layer by visual examination

of the pits which were on the perimeter of the footings, but it is

assumed it was present. Using an estimated compression modulus

ranging from 250 ton/ft^ to 1000 ton/ft^ (24-96 MPa) (18) footing

settlements during construction, attributable to a 1 ft (0.3 m) thick

layer of disintegrated rock under a 10x10 ft (3x3 m) footing subjected

to a 4.5 ton/ft^ (431 kPa) load would have ranged from 0.02 to 0.1 in

(0.5 to 2.5 mm). The condition described corresponds to that at

149
footing E4.8, which was the most heavily loaded during construction.

Some of these settlements could have shown up as differential settlements

if one footing rested on disintegrated rock, and the adjacent footing

on solid rock.

(3) In Boring B-3E there was a 1-ft + (305 mm) thick layer of fill,

followed by a 11.5 in (292 mm) thick layer of disintegrated rock.

Using an estimated compression modulus of 70 ton/ft^ (6.7 MPa),

derived from the SPT blow count (18, 19), the resulting settlement

under construction loads is estimated to have been on the order of 0.3

in (8 mm). This settlement could have shown up as a differential

settlement with respect to adjacent footings. Similarly, Footing F10

,

where the compression modulus of the fill is estimated to be on the

order of 80 ton/ft^ (7.7 MPa), based on SPT and pressuremeter tests

(Appendix B and 19), could have settled an average of 0.35 in (9 mm)

and rotated 1/300 th of a radian (causing a differential settlement of

1/300 th of its width) as a result of the backfill which varied in

thickness

.

In summary, it is estimated that differential settlements not exceeding 3/8 in

(10 mm) could have occurred between some footings after the full construction

load was applied. However,, in most instances, the differential settlements

between footings were probably on the order of 0.1 in (3 mm) or less. Footing

F10 probably rotated, as well as settled under the applied load.

When creep criteria from reference (18) and a two months time span between the

fixing of the first slab in its permanent position and the collapse are assumed, it is estimated that about 95 percent of the settlement occurred

before the first slab was fixed in its permanent position and the remaining 5 percent were creep settlements which gradually increased until the time of the

collapse

.

Residual footing settlements after the debris was removed from the site are difficult to determine from available data which are confined to single

points. The only reference point encountered was the top of the footings

which probably originally deviated from their specified elevation by more than

the anticipated residual settlements. The data indicate that residual

differential settlements are small and do not exceed 1/2 in (13 mm) between

adjacent footings.

2. Basement Wall

The backfill behind the basement wall was explored by two borings with SPT

samples, one test pit, a pressuremeter test, two in-place density tests,

several density tests taken from two tube samples and three laboratory direct

shear tests, one considered drained and one undrained. The material consisted

of silty sand fill with rock fragments and some construction debris and was classified as silty sand (SM) . The material in boring B-l had a natural

moisture content of 9.4 percent and 17 percent of fines (passing No. 200

sieve). The material in boring B-2 had a natural moisture content of 10.7

percent and 25.6 percent of fines. The in-place dry densities of the two sand

150
cone test samples taken were 90 lb/ft^ (1.44 ton/m^) and 110 lb/ft^ ( 1.76

ton/m^) , and their moisture contents 24 percent and 9 percent, respectively.

Dry densities determined from two "undisturbed" tube samples varied from 84 to

102 lb/ft^ (1.35 to 1.63 ton/m^) and corresponded to 70 to 80 percent of

maximum dry density per ASTM D 1557 "Test Method for Moisture-Density Relations

of Soils and Soil-Aggregate Mixtures Using 10-lb Rammer and 18-in Drop." No

ground water was encountered. Angles of internal friction determined by

direct shear tests of samples re-constituted to the densities of 90 and 110

lb/ft^ were of the order of 37 degrees. These results seem high when compared

with the low blow counts and densities in borings B-l and B-2 (20, 21) and

probably resulted from particle interlocking associated with this constant

volume test. At small shear displacements (0.2 in (5 mm) or less), the shear

stress obtained corresponded to an angle of internal friction of 33 degrees or

less. On the basis of all the available data, including interpretation of

blow counts using references [20, 21] it is estimated that, on the average,

representative values for lateral soil pressures should be based on a moist

unit weight of * 110 lb/ft^ (1.76 ton/m^) and an angle of internal friction

between 28 and 30 degrees. Since there was no significant precipitation

between April 19 and April 23, it is assumed that no hydrostatic pressures

acted on the basement wall.

Differential foundation settlements and lateral soil pressure on the basement

wall based on the subsurface exploration data will be presented in Chapter 7.

151
Post-tensioned slab

Ground level

Dry pack grout

Water stop

Porous fill

Dowels

to match

Drain

Note: 1. Lap vert, bars 24 " min.

2. Wall shall be cast after

slabs are lifted

Figure 6.1.1 Typical cross section of basement wall - North side
153

Figure

6.2.1

Location

of

soil

borings

and

test

pits

for

subsurface

exploration
7. STRUCTURAL ANALYSIS

7.1 INTRODUCTION

The structure was analyzed for loadings encountered during erection. Loadings

induced by a number of possible conditions that could have precipitated

collapse of the structure were also studied. This latter case served as a sensitivity study and was used in determining the most probable cause of

collapse discussed in Chapter 9.

Two types of analysis were performed. The first considered only the floor

slabs. The floor slab was modeled with plate bending elements with the columns

treated as pinned supports. Deformations and stresses in the slab and support

reactions were calculated. The second type of analysis considered the structural

frame. A two-dimensional nonlinear analysis was performed for the structural

framing system. Internal member forces and deformations of the structure were

calculated in the elastic and inelastic range. The resistance of the structure

to lateral earth pressure on the basement wall at the lower levels and the

effect of differential foundation settlements were also considered.

Results of the analyses and comparisons with the strength of the structural

components determined using resistance values specified in current design

standards are included in this chapter. Resistance values which are compared

with estimated construction loads are not multiplied by the capacity reduction

factors specified in the applicable design standards. Evaluation of these results

to determine the most probable cause of collapse of the structure are presented

in Chapter 9.

7.2 FLOOR STABS

7.2.1 Computer Model

A two-dimensional finite element computer model was developed to analyze the

floor slabs [9], The model was based on small deflection theory and linear

elastic behavior. Isotropic, homogeneous four-node plate bending elements

with six degrees of freedom per node (three translations and three rotations)

were used. A total of 768 elements was used to model a slab to ensure

accurate representation of the loading conditions and stress distribution.

Openings in the slab at the column locations (shearheads) and pipe chases were

not included in the model. The opening for the shearwall along column line 2 between lines G and H was included. Reactions at the lifting points were

considered point loads at the column centers.

The effects of post- tensioning were included in the model. Normal stress due

to the post- tensioning force in the plane of the slab was calculated. Secondary

moments induced by the drape of the post- tensioning cables were evaluated

using the procedure developed by Lin [10], Assuming the drape of the cable

could be approximated by parabolic segments, the uniformly distributed line

154
load normal to the plane of the slab required to produce drape- induced moments

would be:

8hF

where

:

w - equivalent uniformly distributed line load

h - total drape of cable for parabolic segment

F - force in post- tensioning cable

L - span of parabolic segment

The location of the post- tensioning tendons in each slab was shown schematically

in figure 3.2.3. A line load along the centerline of the columns was used to

represent the load for the banded tendons oriented in the east -west direction.

This was done for computational convenience. It should be noted that the

centroids of the banded tendons do not necessarily coincide with the column

centerlines. Any effects due to this simplification would be more significant

along exterior column lines . The displacement of the tendons from the column

centerline in the vicinity of column E4.8 was modeled. In view of the uniform

spacing of the tendons in the north- south direction, the loads for these

tendons were treated as uniformly distributed over the entire element.

Material properties (unit weight of concrete, elastic modulus) used in the

analysis were based on the test results presented in Chapter 5.

The models were generated using the interactive graphics computer facility at

the N3S Center for 3uilding Technology. The models were solved using a VAX

11/750 computer. Pre-processing and post-processing of the data were done

using PATRAN *11]. The analysis program was the ANSYS stress module.

The majority of the analyses deal with the west tower. Results for the east

tower would be similar except for the effects of displacing the post -tensioning

tendons around column E4.8 and the variable column spacing around the service

core. Several load cases were considered. Support reactions and rotations at

each column location and stresses and displacements throughout the slab were

determined for the slab subjected to its own dead weight. The influence of

loss of support at various columns on the reactions and stresses was determined.

Support reactions due to jacking the slab at one location and due to differences

in elevations of the slab supports were also determined. Support reactions in

the east and west towers were determined.

7.2.2 Results of Analyses

The results of the analysis for a single floor slab in the west tower are shown in figures 7.2.1 through 7.2.4. The loading case considered represents

a slab subjected to its own dead weight and s imply supported at each column

location. The columns and the jack rods are approximately ten times stiffer

tr.an the slab and the assumption of simple supports is reasonable. The

results obtained for a single slab may be extrapolated to the case where

multiple slabs are lifted. Since a bond breaking agent was applied to the
slabs during casting, it is reasonable to assume there is no significant

horizontal shear transfer at the interface between slabs when multiple slabs

were lifted. In this case, the series of slabs behaves as a laminated plate

with each plate bending independently. The stresses and deformations of each

slab, therefore, are the same. The support reaction at a column equals the

sum of the reactions for the slabs at that column. The column extensions for

stages V, VI, and VII were stored on the roof slab. The total weight of the

column extensions for the three stages was 81.5 kips (362 kN ) or 14 percent of

the weight of one floor slab. They were positioned at the various column locations

and this total weight was essentially uniformly distributed over the roof

slab. Support reactions for the package consisting of the roof and level 12,

therefore, can be obtained by multiplying the results for one slab by 2.14.

The support reactions for one slab are shown in figure 7.2.1a. The reactions

in the center of the slab (columns E3 , E3 . 8 , G3 and G4) are higher than the

reactions around the perimeter. Larger jacks or super jacks were used in

these locations to lift the slabs. Note, however, that the reactions at

columns E2 , G2 , E4.8 and G5 are about the same magnitude with the reaction at

E4.8 being the largest in the west building. If the reactions in figure 7.2.1

are multiplied by three for the case where three slabs are lifted, the loads

for columns E3 , E3 . 8 , G3 and G4 are well below the 300 kip (1334 kN) capacity

of the super jacks at these column locations. For most of the other columns

the loads are below the 150 kip (667 kN) capacity of the regular jacks. For

column E4.8, however, this three -slab load is equal to the nominal 150 kip

(667 kN) capacity of the jack.

The ratio of the load required to cause a punching shear failure of the slab

to the support reactions is given in figure 7.2.1b and is denoted as "reserve

capacity." The punching shear failure loads were computed according to the ACI

values [2] given in Appendix C. These values are given in table 7.2.1; they

are a function of the shear area (perimeter of shearhead) and post- tensioning

stress at each column. The reserve capacity values range from 4.0 to 23.1. A

large reserve capacity is to be expected since the reactions are due only to

the dead weight of the slab. When a package of three slabs is lifted, it is

possible there may be lack of intimate contact of the slabs at some points.

In this case, the shearhead at the bottom of the package could be subjected to

the shear force for three slabs. Even for this extreme case, the reserve

capacities obtained by dividing the reserve capacities in figure 7.2.1 b by

three range from 1.3 to 7.7.

The distribution of normal stress at the top surface of the slab is shown in

figure 7.2.2. Compressive stress is denoted as minus, tensile stress as plus.

The large dots represent the column locations. Stresses in the longitudinal

(east-west direction) are shown in figure 7.2.2a and in the transverse

direction (north- south) in figure 7.2.2b. At some locations in this figure

and subsequent stress contour plots, e.g., at the shearwall opening, the

stress on the free boundary is obviously zero. The stress gradient at these

locations is very steep. Because of the size of the finite element mesh and

the contouring algorithm used to generate the figures, the gradient, i.e., the

reduction to zero stress, is not shown. The compressive stresses are relatively

low since the moments induced by the post- tensioning effectively balance the

dead load moments. In most areas these stresses are 400 psi (2.76 MPa) or

156
less. These stresses are well within those allowed by ACI [2] for service

load conditions for post- tensioned structures. Localized stresses as high as 800 psi (5.52 MPa) occur in the vicinity of the service core and around column

H2 and the adjacent shear wall opening. These higher stresses result from the

positioning of the post- tensioning cables (see Figure 3.2.3). In several

locations, most notably near columns H2 , E5 , H4, H5 , F8 , and F9 , either the

majority or all of the east-west banded tendons are placed on one side of the

column.

The transverse stresses in figure 7.2.2b are similar in magnitude to the

longitudinal stresses although the distribution is different.

The distribution of the normal stresses at the bottom surface of the slab is

shown in figure 7.2.3. The magnitude of the compressive stresses is similar

to that at the top surface. Again the post- tensioning effectively balances

the dead load moments. Note that tensile stresses do occur in the vicinity of

the service core. Rotations of the slab about its longitudinal and transverse axes in its plane

at each column location are given in figure 7.2.4. Rotations about the

longitudinal axis (east-west) are similar in magnitude along all the column

lines. An unusual combination of rotations occurs in the northeast corner at

column C6 . Large rotations occur at HI and H6. The largest rotations occur at

columns Cl and C6 about the transverse axis (north- south) . These rotations

would likely have a negligible effect on the behavior of the structure.

The computer model was used to investigate the sensitivity of the floor slab

to loading conditions associated with loss of support. Such loss of support

could be due to a jack rod failure, failure of a shearhead, failure of a weld

block or loss of a wedge. Two illustrative examples will be discussed in this

section, loss of support at a perimeter column (column Cl) and at an interior

column (column E4.8).

Stresses and deflections for the case of loss of support at column Cl are shown in figure 7.2.5 and 7.2.6. The deflections are referenced to the "as

cast" condition. The normal stresses at the top surface of the slab in figure

7.2.5 are essentially the same as those in figure 7.2.2. The situation was similar for the stresses at the bottom surface. Loss of column Cl apparently

has a minor effect on the slab stresses. This is not surprising in view of

the small load (6.3 kips - 28 kN) in column Cl.

Deflections due to loss of support at column Cl shown in figure 7.2.6 are also

small (approximately 0.6 in - 15 mm) at the column.

Stresses and deflections for loss of support at column E4.8 are shown in

figures 7.2.7, 7.2.8 and 7.2.9. The stress distribution throughout the slab

is affected. There is a significant change in the stresses in the vicinity of

column E4.8. This is as expected in view of the large load the column was carrying. Compressive stresses over a large area of the top surface of the slab

around the column reach 800 psi (5.52 MPa). Tensile stresses at the bottom

surface in this vicinity reach 1000 psi (6.90 MPa). These tensile stresses

exceed the tensile strength of the concrete (7.5 Jf^ reference 2).

157
Deflections due to the loss of support at column E4.8 are shown in figure

7.2.9. The deflection at the column location is 1.8 in (46 mm). Deflections

in excess of 1 in (25 mm) occur over a large area.

Loss of support at column E4.8 causes a redistribution of forces to the other

supports. The magnitude of this redistribution is shown in figure 7.2.10.

Tne support reactions due to a package of three slabs with all supports in

place are given in figure 7.2.10a. Reactions due to loss of support at column

E4.8 are given in figure 7.2.10b. Note the considerable increase in reactions

at the surrounding columns, C4.8, E3 . 8 , G5 and E.5. With the loss of column

E4.8, column E3 . 8 becomes the most heavily loaded column, experiencing a fortyeight percent increase in load.

The influence on the support reactions of jacking the slab at one column

location was also studied. This loading condition corresponds to the situation

encountered during the placement of the wedges at a column. In this case the

slab is raised to its approximate final position (for parking or permanent

placement) by operating all the jacks. The wedges are then positioned at each

column by raising or lowering the slab at that column as necessary. The

amount of the displacement required at each column varies. To evaluate the

magnitude of this effect on the support reactions, localized jacking of the

slab at column E4.8 was considered. Results are shown in figure 7.2.11.

Support reactions for a three slab package are given in figure 7.2.11a. The

reactions due to raising the slab 1/2 in (12.7 mm) are given in figure

7. 2. lib. One -half in (12.7 mm) corresponds to one full stroke of the jack.

The load on the jack at column E4.8 must be increased from 151.8 kips (675 kN)

to 201.1 kips (894 kN) , an increase of thirty- three percent. Reactions at the

surrounding jacks change, in general, to a lesser degree. A forty- four percent

decrease occurs in the reaction at column C.5. Roughness of the surface of the

slabs could cause some interaction between the slabs. In this case the force

required to raise the slabs at one point would be greater than the value

given.

The preceding results are based on the assumption that the supports are all at

the same elevation; i.e., the slab is level. During lifting, the slab will

not remain perfectly level since the jacks are not perfectly synchronized.

There will be some differences in elevation between the various support points

prior to placing the slab at its final position on the columns. The lifting

specifications permit a tolerance of 1/2 in (12.7 mm) difference in elevation

between adjacent supports. This will affect the reactions on the jacks.

Results for two conditions are presented in figure 7.2.12 to illustrate this

effect

.

Figure 7.2.12a gives the reactions on the jacks for a package of three slabs

in which the support at location E3 . 8 is 1/2 in (12.7 mm) high. This could

have occurred since there was a super jack on this column and it would have

reached its upper limit earlier than the small jacks. Figure 7.2.12b gives

the reactions for the case in which the support at location C4.8 is low by 1/2

in (12.7 mm). This could have occurred if there were any binding between the

column and slab during lifting. Comparing the results in figure 7.2.12 with

those in figure 7.2.11a illustrates the influence of difference in elevation

at the supports. The effects are somewhat localized. The magnitude of the

158
effect is a function of the load in the jack at the point at which the slab is

high or low. A 1/2 in (12.7 mm) increase in elevation at E3 , 8 affects

primarily the reactions at E3 . 8 , C3 . 8 , and E4.8. The reaction at E4.8, for

example, is reduced by 13 percent. Similarly, if the slab is low by 1/2 in

(12.7 mm) at C4.8 the reactions at C3 . 8 , C4.8, C6 and E4.8 are primarily

affected. An increase of 12.5 percent in the reaction at E4.8 occurs in this

case

.

It is of interest to compare the support reactions or loads on the lifting

jacks in the east and west towers. A comparison for the case of a single slab

is given in figure 7.2.13. Support reactions for the west tower are given in

figure 7.2.13a and for the east tower in figure 7.2.13b. Note that the east tower

slab has a different configuration than the west tower slab and is slightly heavier

(622 kips vs. 601 kips, 2767 kN vs. 2673 kN) . For the east tower, the largest

reactions occur at columns B9 , BIO, D9 , and DIO. These were the locations of

the super j acks

.

7.3 COLUMN INSTABILITY

Instability of the columns in stage IV of the west tower was investigated to

determine whether this mode of failure could have caused collapse of the

building. Critical loads were calculated for each column in accordance with

the AISC LRFD procedure [8] . The unbraced length of each column was assumed

to be 30 ft- 5 in (9.3 m) . Referring to figure 4.2.1, this corresponds to the

length of column from level 3 to the top of stage IV, or the location of the

jacks. Level 3 was selected as one end since the wedges had been welded in

place at this level. Level 12 rather than the top of stage IV could have been

selected as the other end since wedges had been tack welded at this level.

Using the top of stage IV as one end point for the effective column length

increased the unbraced length slightly and provided conservative results. The

slabs at level 3 up to and including the roof were assumed to offer no lateral

restraint in view of the clearance between the columns and the shearheads at

the floor levels. Pinned end conditions were assumed. Rotational restraint

provided by the column below level 3 was neglected and an effective length

factor of K - 1 was used.

Results of the analysis are given in table 7.3.1. Critical loads according to

LRFD are listed in the fourth column of the table. A value of 1.0 was used

for the reduction factor for compression in calculating these loads. The

calculated critical loads therefore represent upper bounds. The reserve

capacity or the ratio of the calculated critical loads to the loads on the

columns due to the weight of the roof slab and levels 12 through 9 is also

given for each column. This reserve capacity ranges from 2.15 for column

E4 . 8 to 7.77 for column C.5. Clearly, the columns in this case have considerable

reserve capacity and stability of individual columns was not a problem.

During the initial stages of collapse of the building, it is possible the

intermediate floor slabs could have collapsed while the upper level slabs

remained temporarily in place. In this case the unbraced lengths of the

columns would have increased considerably and column buckling could have occurred

after the initial failure but during collapse of the building. The unbraced

lengths required for this to occur were calculated for two conditions

:

condition 1 with the roof and levels 12 through 9 in place, and condition 2 with

159
only the roof and level 12 in place. These unbraced lengths are also given in

table 7.3.1. Note the values are quite large, in some cases exceeding the

total length of the column. Buckling could have occurred in those cases where

the critical length is less than 79 ft 1 in (24 m) , the distance between level

E and level 9 in figure 4.2.1. This was possible for many of the columns

under condition 1 in table 7.3.1. This will be discussed further in Chapter 9 in considering the sequence of the collapse.

The possibility of frame instability or sidesway buckling will be considered

in Section 7.4.

7.4 LATERAL LOAD ANALYSIS

7.4.1 Introduction

Three cases of lateral loading were considered in this investigation. The

first case involved loads applied by a hydraulic jack placed between the west

and east towers at the upper levels in order to plumb the building. The second

involved lateral soil pressure on the basement wall on the north side of the

building at levels D and C. The third involved wind loads. Frame stability

or sidesway buckling was also analyzed. These are considered in this section.

7.4.2 Frame Stability

The possibility of sidesway buckling of the west tower is considered in this

section. It is necessary to check this mode of failure since there was no evidence that the west tower, at the time of collapse, had effective lateral

bracing above the shear wall at level 2 as indicated in figure 4.2.1.

To check the capacity of the partially complete structure against sidesway

buckling, several analyses were conducted. First, inelastic stability in the

east-west direction was determined by a two-dimensional, linear, inelastic

analysis of each column line. A schematic of column line C of the west tower

and diagram of the model used for the stability analyses are shown in figure

7.4.1. The steel columns in the structure were modeled as line elements, 83

ft 6 in (26.4 m) in length (measured from level E to the top of stage IV) and

were fixed at the base (level E) . The shear wall was assumed to provide

restraint against lateral movement. The columns were, therefore, restrained

against lateral displacement at level D through level 2 (top of shear wall).

The concrete slabs were assumed to be infinitely rigid in their plane and the

attachment between the floor slabs and columns was assumed to be incapable of

transmitting a moment. Consequently, the floor slabs were modeled as rigid

links pinned at the columns.

Geometric properties of the steel sections were obtained from the AISC Manual

of Steel Construction [7]. Material properties used in the computer model for

the 36 ksi and 50 ksi steel were based on the test results reported in Section

5.3.

The loads on the west tower at the time of collapse were the gravity loads of

the slabs supported by the columns. The magnitude of these loads was determined

from the linear analysis of a single slab supported at each column at the same

160
elevation and loaded with self weight. The column reactions found from this

analysis are shown in figure 7.2.1.

At the time of collapse, wedges were being placed beneath slabs 9/10/11. At

those columns where the wedges had not yet been installed, the slabs were

supported by the jacks and the slab loads, therefore, were applied at the top

of the column. For the purpose of this analysis, this was neglected and the

loads were assumed to be applied through the wedges in all cases.

The computer program used for this study [12] is capable of analyzing elastoplastic steel frames subjected to non-proportional static loads. It uses an incremental, nonlinear analysis procedure and includes post-collapse behavior.

Both large displacement and member stability effects are included. The beamcolumn element consists of an elastic component and an elastic-perfectly

plastic component in parallel. The element may yield through the formation of

localized plastic hinges. The member yield criterion is defined by a twodimensional yield surface which accounts for the interaction between axial force

and bending moment. The effects of strain hardening may be accounted for in an approximate way by assigning a non- zero elastic component. Geometrical

nonlinearities are accounted for by including the frame member geometric

stiffness matrix and formulating the equilibrium equations on the basis of the

displaced structure.

Results of the inelastic stability analysis for each column line are given in

table 7.4.1. The results are reported as the ratio of column line capacity to

total load on the column line. A value less than 1.0 indicates that the

column line by itself is not capable of carrying the gravity load of the

slabs. Similarly, a value greater than 1.0 indicates that the column line can

not only carry its share of the slab loads, but that it has some reserve

capacity for bracing the other column lines.

Since the floor slabs are effectively rigid in their plane, no single column

line can fail in a sidesway mode without all column lines failing in this

manner. Consequently, one needs to consider the capacity of the entire west

tower (all four column lines) . The capacity of the building may be determined

from the individual column line capacities by noting that the building will

not fail until the sum of the capacities is exceeded by the total applied

load. The capacity of an individual column line is determined by multiplying

the load on the column line by the ratio reported in the first column of table

7.4.1. The column line loads and the resulting capacities are given in the

second and third columns of table 7.4.2, respectively. The sum of the four

column line capacities is computed to be 10265 kips (45659 kN) while the sum

of the loads on all column lines is 9621 kips (42794 kN) . The ratio of

capacity- to- load is, then, 10265/9621 - 1.07 which can be interpreted as a seven percent reserve capacity against inelastic sidesway buckling.

161
From the above analyses it was found that inelastic behavior (yielding of the

steel columns) did not occur until the elastic buckling capacity was just

about reached. This was determined by observing the load vs. top of frame

displacement curves and noting that the slope was almost flat at the time of first

yielding. This means that, for practical purposes, an elastic stability

analysis for sidesway is adequate. The four column lines were analyzed again

ignoring inelastic behavior. The results of these analyses are presented in

table 7.4.2. It can be seen that the capacities obtained are only slightly

greater than those found from an inelastic analysis. Using the procedure

described above for determining the stability of the entire building, one obtains a capacity- to -load ratio of 1.08, or an eight percent reserve capacity

against elastic sidesway buckling.

Since an elastic stability analysis was shown to give good results for this

structure and loading, a single column model was used for determining the

sidesway buckling capacity in the north- south direction. The geometry was the

same as that used for the column line analyses except that there was only a single column. The moments of inertia and cross sectional areas were computed

by summing the contributions from all columns at each level. Similarly, the

gravity loads were found by adding all the loads at a given elevation. An

elastic stability analysis in the east-west direction was first performed

using this model to compare with the calculations made above. The results

indicated a capacity- to- load ratio of 1.08 which agrees with the results

considering the four column lines individually. Next, an analysis in the

north- south direction was conducted and a capacity- to- load ratio of 1.16 was obtained. From these analyses, it can be concluded that the west tower is

more susceptible to sidesway buckling in the east-west direction than in the

north-south direction.

Whether the computed reserve capacity is eight percent or even 16 percent,

such a low value is cause for concern. The assumptions under which the stability

analyses were performed should be challenged. First, the slab- to -column

connections where the wedges and seal blocks were fully welded and the

cavities around the columns filled (level D through level 1) provide a rotational

restraint which was ignored in the above analysis. Omitting this restraint

should not affect the results significantly since very little rotation of the

columns occurs below level 2. However, it would tend to increase the stiffness

of the structure and thereby increase the lateral buckling capacity. To

verify this, the model was laterally restrained at level 2 and completely

restrained at level 1 to simulate the condition of infinite slab rigidity

below level 2. Results indicated a capacity- to- load ratio of 1.10, only

slightly greater than the value of 1.08 obtained assuming the slabs do not

provide any rotational restraint.

Next, the assumption of full lateral restraint at level 2, the top of the

freshly cast shear wall, could be questioned. To test the extreme case, the

single column model was analyzed again with no lateral restraint at level 2.

Results showed a capacity- to- load ratio of 0.65. This analysis serves to

point out that the capacity- to- load ratio of 1.08 should be regarded as an upper bound and that the true capacity for the idealized structure analyzed is

somewhat lower than this value.

162
Construction of the east tower was slightly ahead of the west tower at the

time of collapse (see figure 4.2.1). In the east tower, slabs 6/7/8 had been

parked at elevation 76 ft 8 in (23.4 m) and slabs 3, 4 and 5 had been lifted

to their final positions. Rather than analyze the east tower, the west tower

was analyzed for the configuration of slabs in the east tower under the

assumption that the behavior of the two towers is essentially the same. For

this analysis, no rotational restraint was assumed for any of the slabs, but

level 2 was laterally restrained as assumed in previous analyses. Results

indicated a capacity- to -load ratio of 0.72. A more severe condition existed

in the east tower during lifting of slabs 4 and 5 since the weight of these

two slabs would have been supported by the jacks on top of the columns. This

case was analyzed also and a capacity- to- load ratio of 0.59 was obtained. It

remains, however, that the east tower survived the lift to the configuration

shown in figure 4.2.1. There was obviously some stabilizing effect which has

not been considered in the analyses so far. A discussion of this effect follows.

In the above analyses, no rotational restraint was assumed for those levels

where the seal blocks were not welded nor the cavities around the columns

filled. This condition existed from level 2 to the top of Stage IV in the

west tower as indicated in figure 4.2.1. while the building remains undeflected,

the slab load, W at each column is shared by the two wedges with each carrying

approximately the same load as shown in figure 7.4.2 (a). However, if the

building were to sway, the load in those columns oriented with their web in

the plane of sway would transfer to one wedge. Because the wedges are separated

by a distance approximately equal to the column depth, this shift in load

would produce a moment which acts to resist the sway deflection as indicated

in figure 7.4.2 (b) . When the slab load is completely transferred to one wedge, the slab would lift off the other wedge as shown in figure 7.4.2 (c)

and the moment due to the eccentricity of the load would remain constant. The

rotational restraint produced by this restoring moment was not considered in

the above analyses and is felt to be important.

Prior to lift off, the slab and column rotate together (compatible deformations)

and behave as if they are rigidly connected. The slab resists rotation since

it is continuous at the column support and has a stiffness, k greater than

zero. After lift off, however, the moment due to the eccentricity of the slab

load remains constant with increasing rotation. Thus, the slab offers no rotational restraint, or k-0 . The slab may be considered as providing linear

elastic rotational restraint up to the point of lift off and no rotational

restraint from that point on. This is analogous to the elastic-plastic

behavior depicted in figure 7.4.3. The limiting moment (moment at lift off) is

Ml - Wx(d+t)/2 in which W is the slab load, d is the column depth, and t is

the thickness of a wedge.

The influence of this rotational restraint on sidesway buckling capacity is

illustrated in the following simple example. Consider an axially loaded

column of length L which is fixed at the base as shown in figure 7.4.4. If

this column is rotationally unrestrained at the top, the theoretical K-value

is 2. If, on the other hand, the column is fully rotationally restrained at

the top, the theoretical K-value is 1 which increases the buckling capacity to

four times that of the unrestrained case. If there is a rotational restraint

at the top of the column with a stiffness greater than zero and less than

163
infinity, then the buckling capacity will be greater than that of the unrestrained

case and less than that of the fully rotationally restrained case, i.e., 1 < K < 2. However, if sufficient rotation occurs to cause lift off (<£ > in figure

7.4.3), then the rotational spring stiffness becomes zero and the buckling

capacity becomes that of the unrestrained column. This means that a load

greater than the sidesway buckling load can be carried by a column with such a rotational spring until such time that the rotation is sufficient to cause the

spring stiffness to go to zero (lift off) at which time the buckling capacity

crops and the column buckles

.

To understand the effect of this restraining force on the capacity of the west

tower, an analysis was performed using the computer program cited above. A

single column model of the entire building was used as shown in figure 7.4.5.

The base elevation was chosen to be level 1 where full fixity was assumed and

lateral restraint was provided at level 2. The slabs were assumed to be one half the average bay spacing or approximately 10 ft (3.05 m)

.

The effective

width of the slabs on either side of a column line was taken to be 8*h where h

is the slab thickness. The ends of the slabs (centerline of a bay) were not

fixed against rotation but were constrained to move vertically with the slab

to -column joint to eliminate slab bending due to axial shortening of the

column.

It was assumed that 30 slab panels act to resist moment where each panel has

an effective width of 56 in (1.42 m) . This was determined by noting that the

slab resists rotation only at those columns which are oriented with the plane

of their web in the east-west direction. Since 45 percent of the slab self

weight was carried by columns oriented with their webs in the east-west

direction, the lift off moments were computed on that basis. Additionally, the

weight of the column sections for stages V through VII, which were stockpiled

on the roof, and the weight of the lifting jacks were included.

Results indicated that the sidesway buckling capacity- to-load ratio was increased to 5.6. Therefore, this effect explains the apparent stability of

the structure and how slabs in the east tower were lifted into position

despite a capacity- to- load ratio of 0.59 computed earlier. The lateral

deflection at which instability occurred, however, was computed to be only

0.56 in (14.2 nun). This indicates that the structure was extremely vulnerable

to lateral displacements and that a displacement at the top of the structure

of only about one half inch was capable of causing lift off and subsequent

instability

.

7.4.3 Horizontal Jacking Load

A 12 ton (107 kN) hydraulic jack was used on the day of the collapse to plumb

the building during erection of the floor slabs in the west tower. The

location of this jack is shown in figure 7.4.6. The foreman for the firm

responsible for constructing the floor slabs indicated the west tower was plumbed by moving the slabs approximately 5/8 in (16 ram) laterally.

Representatives from the lifting subcontractor confirmed this in interviews on

July 9 and 10, 1987. The magnitude of the load applied was not known. This

jack was in-place and under load at the time of the collapse.

164
To determine the influence of this jacking force, the single column computer

model, including the restoring moment and lift off mechanism, was used. Two

load cases were applied. First the gravity loads were applied and then a lateral load was applied at the top of the structure to simulate the jacking

force. Results showed that a lateral load of 41 kips (182 kN) was required to

obtain first lift off which occurred at a lateral displacement of 0.54 in (13.7

mm) . All slabs had lifted off under a lateral load of 47 kips with a displacement

of 1.54 in (39.1 mm). Since the structure was just barely stable under the

boundary conditions of no rotational restraint provided by the slabs, sidesway

buckling did not occur. The lateral stiffness, however, was found to be 0.76

kips per inch. Since a lateral load of 41 kips (182 kN) is required for lift

off and the capacity of the jack was approximately 24 kips (107 kN) , it is not

likely that the use of this jack to plumb the building initiated the collapse

of the west tower.

7.4.4 Earth Pressure

Introduction

It is deduced from available evidence that the lateral pressures exerted by

the backfill placed behind the basement wall to the north of the building

prior to the building collapse were in part resisted by the floor slabs at

levels C and D. The evidence includes the observed basement wall deflections

after the collapse and the construction logs. The reaction forces acting on

the slabs were in turn transmitted by the slabs to the columns and shearwalls

,

which carried the forces to the foundations. This section presents an estimate

of the effect of these forces on the building under construction.

Loads

Data on the backfill behind the basement walls are presented in Chapter 6 and

Appendix B. It is estimated on the basis of these data that the average unit

weight of the backfill at the time of the collapse was approximately 110 lb/ft^

(1.76 ton/m^) , that the material was cohesionless, and that the average angle

of internal friction was between 28 and 30 degrees. No significant precipitation was recorded during the five-day period preceding the collapse and it is thus

assumed that no hydrostatic pressure acted on the wall. Since the slabs

restrained the wall from rotating over its base it is assumed that a reasonable

upper limit for the soil pressure acting on the wall was the at-rest pressure,

which is estimated to have been approximately 55 to 58 times the depth of the

backfill, where the depth of the backfill is in ft and the pressure is in psf.

Actually, as the backfilling behind the wall proceeded, some wall movement had

to occur, somewhat relaxing the pressure against the wall. A lower limit for the

backpressure would be the active pressure, which is estimated to be approximately

37 to 40 times the depth of the backfill in psf. The pressure would drop to

the active pressure only after a rigid body rotation producing displacement on

the order of 1/4 to 1/2 in (6-13 mm) at the top of the wall, which is more

than the displacement that is estimated to have occurred during construction. In

the following calculation a lateral soil pressure of 50 times the depth of the

backfill is used.

165
Figure 7.4.7 shows the assumed soil pressure diagram together with the

calculated reaction forces. It was conservatively assumed that the wall was hinged at the top of the foundation. Even though the dowels to the footing

provided moment resistance, the footing itself was narrow. Calculated forces

per linear foot of wall are 251 lb (3.66 kN/m) at level C and 5610 lb (25.0 kN)

at level D.

Resistance Mechanism

The lateral forces are resisted by the shearwalls and the columns. If it is

assumed that the in-plane stiffness of the slabs is large, an assumption which

reasonably represents actual conditions , then the portion of the lateral load

acting on each load resisting element will be approximately proportional to

its stiffness. The following element stiffnesses were calculated for the west

tower in the north- south direction (those for the east tower do not differ

significantly)

:

N-S shearwalls: 2 I - 7.593x10^ in4 (3.16 m

4

) 2 El - 2 . 580xl0 13 lb-in2 (7.404xl04 MN-m2

) columns

:

2 I - 1.515xl04 in4 (6.306xl0 9

nun4 ) 2 El — 4.395X10 11 lb-in2 (1.261xlCH MN-ra2

) E-W shearwalls: 2

2

I - 2 . 070xl04 in4 (8.616xl0 9

nun4

) El - 7 . 050xl0 10 lb-in2 (2.02xl0 2 MN-m2

) where I - moment of inertia, in the north-south direction, E - Young's

modulus, and 2 stands for the algebraic sum of the quantities indicated for

the west tower building.

These stiffnesses show that, unless a shearwall displacement by structural

failure, slippage, or tilting occurs, only approximately 2 percent of the load

resulting from lateral soil pressures is transferred to the columns and eastwest shearwalls, and the remaining 98 percent of the load is resisted by the

shearwalls in the north- south direction. A similar situation occurs in the

east-west direction, even though the longitudinal shearwalls in that direction

are somewhat shorter. Thus it is reasonable to assume that the lateral loads

attributable to the soil pressures against the basement walls are resisted by

the shearwalls in the north-south direction, and torsional effects are resisted by the shearwalls in the north-south direction and the shearwalls in

the east-west direction.

Resistance to Shear Forces

The following material properties are used in the assessment (the properties

were derived from the NBS laboratory test results and the project specifications):

Concrete cylinder strength (f' c

): 4,570 psi (31.2 MPa)

Concrete modulus (E) : 3,400,000 psi (23443 MPa)

Re-bar strength (fy): 60 ksi (414 MPa)

166
If it is conservatively assumed that the shear forces shown in figure 7.4.3 acted on the slabs over the entire 112 ft (34.14 m) length of the building (the

resistance provided by the cross wall to the east end of the west tower

basement wall is disregarded) , then the total shear force acting on the two

north- south shear walls between levels D and E was 656 kips (2918 kN) . The

shear force acted with an eccentricity of 5.2 ft (1.59 m) with respect to the

center of rigidity of the building. The resulting torsional moment was resisted by the shearwalls in both directions, inducing a shear force of 27 kips

(120 kN) in each wall in the north south direction and a force of 18 kips (80

kN) in each wall in the east-west direction. The resulting shear forces are

355 kips (1579 kN) for wall H-G2 and 301 kips (1339 kN) for wall E.5-C.5 between

lines 5 and 6. The forces induced in the shearwalls in the east-west direction

had a negligible effect.

The shear capacity of the 13 -ft (3.96 m) long and 1-ft (305 mm) wide shearwalls

is calculated as follows (2)

:

Vu - V

where

:

s

us - [3.3 (f' c 1/2) hd + Avfyd/s ] - 499 kips (2220 kN)

.

wall thickness - 12 in (305 mm)

effective wall depth - 0.8x13x12 - 124.8 in(3.17 m)

reinforcement area for 2 #4 horizontal reinforcing bars

0.4 in^ (258 mm^)

vertical spacing between horizontal reinforcing bars

18 in (457 mm)

.

Thus the ultimate shear resistance of the most critical shearwalls is 40 percent

greater than a conservative estimate of the horizontal soil pressure.

A failure could also occur between the slab at level D and the shearwalls to

which it transmits the lateral load. Details of the slab connection to the

shearwall are shown in structural drawings S302, S303 and S304. The slab is

connected to the shearwalls by #4 dowels spaced at 12 in (305 mm) on center.

In accordance with the plans, the slab surface was roughened to improve frictional

resistance. The load capacity of the connection for wall H-G2, where the slab

to wall connection is subjected to double shear, is calculated in accordance

with reference [2], using the shear- friction method as follows:

Vn - Avffy - 2(13x0.2x60) - 312 kips (1388 kN) (with a friction coefficient of

1 for roughened concrete)

where: Avf

is the area of the shear friction reinforcement.

Thus the ties were not quite adequate to transmit the entire shear force of

340 kips (1512 kN) transmitted by the level D slab. However, for a shear

failure to occur, the slab bearing against the north side of the shearwall

would also have to fail. Assuming that the contact area of the slab with the

12 in (305 mm) wall face was 84 in^ and the concrete strength 4,500 psi (31.0

MPa) , it would take a force of approximately 320 kips (1423 kN) to crush the

slab. It is conceivable that the slab could have deflected excessively or buckled before reaching its crushing strength. Such a failure mode is resisted

by the transverse post tensioning. Since the difference between the shear load

and the capacity of the ties was only 28 kips (125 kN) , this type of failure is

167
considered unlikely. The reserve capacity against failure of the connection

between the slab and shearwall is therefore 1.9 [(312 + 320)/340].

For wall E.5-C.5 the shear force transmitted by the slab at level D is 287 kips

(1277 kN) . For this wall which is adjacent to the elevator shaft this force

is transmitted in single shear. Thus the load capacity of the connections is

1/2 x 312 - 156 kips (694 kN) and the difference between the capacity of the

ties and the shear load is 131 kips (583 kN) . The reserve capacity of this

latter connection is estimated to be 1.65 [(156 + 320)/287] . Since the

shearwall was cast after the slab was in place there was probably a tight fit

between the vertical face of the slab and the shearwall. It is therefore

unlikely that the simultaneous mobilization of the shear- friction mechanism and

the bearing of the slab against the wall resulted in any horizontal displacements

.

The question also arises whether sliding or overturning of the shearwalls

could have occurred. Sliding resistance at the foundation level is examined

for the two north- south shearwalls in the west tower building.

Footing H-G2 rests on a 1 ft thick layer of disintegrated rock underlain by

undisturbed bedrock. No soil fill material was found under the footing. The

plans do not call for a shear key into the rock, and the test pits indicate

that the footing was poured against a form by overexcavating the rock in front

of the footing. Thus no lateral resistance is provided by the rock itself,

except for the frictional resistance at the bottom of the footing. In

addition to this frictional resistance, sliding resistance can be provided by

the 4- in (102 mm) thick slab on level E which bears against the shearwall and

by the backfill below the slab on level E.

In addition to the shearwall, footing H-G2 supported columns H2 and G2 , which

in turn supported the dead weight of the lifted slabs. The shear wall itself

did not support any dead weight of slabs, because the slab -shearwall connection was poured after the slabs were in place. Because of the shrinkage of the cast

in place concrete, the shearwall may even have pulled down on the slabs.

However, this effect was probably negligible. Thus the total dead weight is

estimated as the two column reactions plus the dead weight of the footing and

the shearwall. Assuming that five levels of shearwall were in place, the

total weight of the shearwall, the footings, and the two columns is estimated

to have been 1270 kips (5649 kN) . To resist the 355 kips (1579 kN) lateral

force, a friction coefficient of 28 percent is required. The friction

coefficient at the base of the footing was controlled by the material under

the footing. The borings and test pits indicate that in general there was either fill material or disintegrated rock; however, the field data indicate

that this particular footing rested primarily on disintegrated rock. The fill

material encountered under footings in the field exploration is estimated to

have an angle of shearing resistance between 31 degrees and 34 degrees. The

disintegrated rock would have a much larger resistance; however, construction

records indicate that the footings were leveled using at least a thin layer of

soil. Using the worst case of a 31 degrees angle, the coefficient of friction

would be on the order of 60 percent. Thus sliding of this footing was unlikely.

The other potential mechanisms providing sliding resistance would require some

lateral displacement and therefore were probably not mobilized. The reserve

168
capacity against failure, relying on sliding resistance alone, was on the order

of 2.7 or more. An analysis for overturning at shearwall footing H-G2 indicates that the

resulting force acted within the middle third of the footing. Thus the

footing was stable.

Footing E.5-C.5 between lines 5 and 6 was not explored by borings or tost

pits. It is assumed that conditions were no worse than those encountered at

footings resting on soil fill. In addition to its own weight and the weight

of the supported shearwall, the footing supported columns C.5 and E.5 on line

6. The total weight of the wall, the footing and the supported columns is

estimated to have been 723 kips (3216 kN) . To resist the 300 kips (1334 kN)

lateral force by friction alone, a friction coefficient of 42 percent is

required. It is estimated that the actual friction coefficient between the

bottom of the footing and the fractured rock was at least 53 percent. The reserve capacity attributable to frictional resistance alone was therefore at least

1.26 and probably more. An analysis for overturning indicates that the resulting force acted at a distance of 6.6 ft. (2.0 m) to the south of the centerline of the footing, which

is 1.93 ft (0.59 m) outside the middle third of the footing. Thus there was contact pressure over only part of the footing area. However, the footing was stable, even though it could have rotated slightly if there was fill below the

footing, causing an increase in lateral deflections. Boring B-4.8E, which is

the closest to this wall, shows a foundation resting on disintegrated rock with

no fill. If similar conditions prevailed at the shear wall the effect of

tilting was negligible.

Thus it is concluded that it is unlikely that the lateral soil pressures

acting on the north basement wall caused Che shearwalls to fail either by

sliding or overturning over their base or by a structural failure of the

concrete, or by a failure of the connection between the shearwalls and the

lift slabs.

Lateral Displacements

Lateral displacements were conservatively calculated assuming that the floor

slabs have an infinite in-plane stiffness and no out of plane stiffness. Lateral

displacements caused by flexural and shear deformations of the shearwall were

calculated to be on the order of 1/8 in (3 mm) or less at level C and 3/100 in

(0.08 mm) or less at level D. Rotation due to foundation settlements could have

increased these displacements.

Conclusion

It is concluded that it is unlikely that shearwall failure or excessive

displacements caused by lateral soil pressure against the basement wall to the

north of the tower structure contributed to the building collapse.

169
7.4.5 Wind Loading

The maximum wind loads likely to have been acting on the structure at the time

of collapse were calculated using the procedure described in ANSI A58.1 [14]

and the wind speed data described in Section 4.3.

From the maximum recorded 10-minute mean speed of 16 knots (8.24 m/s) at the

anemometer site, the corresponding hourly mean speed at a height of 10 meters

in open terrain (Exposure Category C) was 18.7 mph (8.36 m/s). The corresponding

wind direction was approximately 120 degrees. Conversion of this speed to the

equivalent fastest-mile speed yields 21.1 mph (9.43 m/s) which is the basic

wind speed defined in ANSI A58.1 for the calculation of wind loads.

The terrain surrounding the construction site is considered to be best

described as Exposure Category B for which the exposure factor and the gust

response factor at the tops of the stage IV columns are 0.79 and 1.34,

respectively. Depending on the reference data used, drag coefficients for the

floor slabs and columns can range from 1.5 to 2.0. Making a conservative

choice of 2.0 for the drag coefficient and assuming that the local variation

in wind direction is sufficient to bring the wind normal to the E-W column

lines, the computed unit drag load at the top of the structure is 2.4 psf (115

Pa) .

If it is assumed that the unused column extensions lying on the roof slab

formed a barrier with an effective height of 3 ft (0.91 m) and the three

packages of floor slabs in the east building acted aerodynamically as a single

obstacle, the effective blockage height for drag load calculations at the top

of the building is 9 ft (2.74 m) . The corresponding drag load per unit width

is (9)*(2.4) - 21.6 Ibs/ft (315 N/m) . The nominal load on an interior N-S

column line due to wind acting on the parked slabs and column extensions is

(21.6)*(27 + 24.5)/2 - 556 lb (2.47 kN) . A similar calculation yields 454 lbs

(2.02 kN) for the nominal load on an interior E-W column line assuming the

wind direction is parallel to the column line.

Because the "project north" does not coincide with the true north, the wind

direction at the site could have ranged from 130 to 160 degrees, referenced to

"project north." As a result, the above calculations could overestimate the

wind loads by 30 percent or more. However, this analysis does not include the

drag load contribution of the vertical columns or the permanently positioned

floor slabs, nor does it account for the actual terrain features at the site.

In Section 7.4.3 the lateral load in the east-west direction required to

obtain first lift off from a supporting wedge was shown to be 41 kips (182

kN) . Since this is an order of magnitude greater than the combined drag force

on the column lines, it is concluded that wind effects did not play any significant

role in the collapse.

7 . 5 DIFFERENTIAL FOUNDATION SETTLEMENTS

As pointed out in Chapter 6, it is estimated from available subsurface data

that differential foundation settlements not exceeding 3/8 in (10 mm) could

have occurred during construction with approximately 90 percent of these

settlements occurring before any slab was fixed in its permanent position. In

170
the lift-slab method of construction, all the slabs are lifted at the beginning

of the process. Once the slabs are lifted, most of the foundation loads are applied

and subsequent load increases during construction are much smaller. Another

feature of this construction method is that slabs are leveled after they are lifted close to their final position.

In the case of the L' Ambiance Plaza project this means that, to the extent

that differential settlements occurred, a major portion of these settlements

probably occurred during the initial stages of the project, a considerable

time before the collapse. The subsequent leveling of the slabs compensated

for these initial differential settlement effects and it was well within the

tolerances of the system to compensate for the estimated differential settlements.

Since foundation settlements occurring after slabs are leveled and in a fixed

position could result in a downward displacement at column supports, and since

the post- tensioned slabs are sensitive to a downward displacement occurring at

a location where the post tensioning strands are near the upper surface of the

slabs, it is of interest to quantify possible settlements that could have

occurred after the slabs were fixed. Such settlements would be the sum of

creep settlements and settlements occurring from additional construction load.

The maximum creep settlements are estimated to have been five percent of the

maximum 3/8 in (10 mm) settlement or 0.02 in (0.5 mm). The maximum construction

load, resulting when a package of three slabs is lifted up 1/2 in (13 mm) was calculated by a finite element analysis to be on the order of 60 kips (267 kN)

.

Under the worst foundation condition encountered in the subsurface exploration,

it is estimated that such an overload could have caused a settlement on the

order of 0.025 in (0.64 mm). Thus the maximum settlement after initial lifting

of the slab is estimated to have been on the order of 0.05 in (1.3 mm) or less. It is considered unlikely that a settlement of that order of magnitude

could have caused a structural failure of a slab.

Another characteristic of the L' Ambiance Plaza footings is that, even though

there was evidence of fill in some locations which could have in some cases

permitted settlements and footing rotations, the depth of the fill layers and

underlying disintegrated rock layers was limited and there was a base of

competent bedrock usually 1 ft (305 mm), and probably not more than 3 ft (0.9

m) below the base of the footing. Thus the possibility of major settlements or stability failures is considered unlikely. This point is addressed in more

detail in Appendix B.

It is therefore concluded that it is unlikely that the building failure was triggered by differential foundation settlements.

This conclusion is based on available data, which are confined to borings and

test pits taken by NBS at several randomly-selected footings, and borings taken

by Heynen Engineering prior to the construction. Actual physical data are only available where actual borings were taken and test pits were dug.

171
TABLE 7.2.1

SHEAR CAPACITY OF FLOOR SLAB

Column Location Shearhead Type Shear capacity Vc (Kips)

Cl Mark A50 54 .

2

C2 Mark C50 106

C3 Mark W50 101

C3 .

8

Mark G50 116

C4.8 Mark C50 106

C6 Mark J50 53.8

C . 5 Mark L50 54.4

El Mark B50 106

E2 Mark P50 182

E3 Mark X51 223

E3.8 Mark X51 220

E4.8 Mark P50 202

E . 5 Mark U51 54.4

G1 Mark B50 76.7

G2 Mark P50 184

G3 Mark Y51 219

G4 Mark Y51 218

G5 Mark P50 198

G6 Mark C50 77

HI Mark A50 54.2

H2 Mark D50 112

H3 Mark N50 130

H4 Mark N50 130

H5 Mark D50 130

H6 Mark A50 54.2

172
TABLE 7.3.1

COLUMN STABILITY DATA

Colum No. Section

(Stage IV)

Axial*

Load (kips)

(1)

Critical**

Load (kips)

(2)

Reserve

Capacity

(2)/(l)

Critical Length (ft

Condition | Conditioi 1 | 2

Cl HP 10x42 31.5 135.7 4.3 63.1 99.8

C2 HP 12x53 79.5 236.5 3.0 52.7 83.4

C3 W 12x65 81.5 318.4 3.9 61.0 96.6

C3.8 W 12x65 79.5 318.4 4.0 61.8 97.8

C4.8 HP 12x53 96.5 236.5 2.5 47.9 75.7

C6 W 8x35 11.5 80.0 7.0 80.2 127

C.5 W 12x65 41.0 318.4 7.8 86.6 136

El W 10x60 68.5 219.0 3.2 54.4 86.0

E2 W 12x106 213 543.4 2.6 49.7 78.6

E3 W 12x120 223 620.5 2.8 52.0 82.2

E3.8 W 12x120 222 620.5 2.8 52.1 82.4

E4.8 W 12x106 253 543.4 2.2 45.6 72.1

E.5 W 12x72 61.0 355.3 5.8 74.6 118.0

G1 W 10x60 67 219.0 3.3 55.0 87.0

G2 W 12x106 199.5 543.4 2.7 51.4 81.2

G3 W 12x136 242 711 2.9 53.6 84.7

G4 W 12x136 243.5 711 2.9 52.9 83.6

G5 W 12x106 221.5 543.4 2.5 48.7 77.1

G6 HP 12x53 60.5 236.5 3.9 60.4 95.6

HI HP 10x42 42.5 135.7 3.2 54.4 86.0

H2 W 12x65 85.5 318.4 3.7 59.6 94.2

H3 W 12x72 119 392.6 3.3 65.2 88.9

Hi W 12x79 124 392.6 3.2 55.1 87.1

H5 W 12x65 92 318.4 3.5 57.5 90.9

H6 HP 10x42 41.5 135.7 3.3 55.0 87.1

Roof, Levels 12,11,10,9

I K-l, unsupported length - 30 ft 5 in

Jrdition 1 - Load due to Roof and Levels 12,11,10,9

bndition 2 - Load due to Roof and Level 12

^e: capacity reduction factor <£c - 1

173
Table 7.4.1

COLUMN LINE CAPACITIES FOR INELASTIC SIDESWAY BUCKLING IN THE EAST-WEST DIRECTION

Column Line Capacity- to- Load Capacity

Load Ratio (kips) (kips)

c 0.87 1216 1058

E 1.20 3466 4159

G 1.13 3325 3757

H 0.80 1614 1291

Total 9621 10265

174
TABLE 7.4.2

COLUMN LINE CAPACITIES FOR ELASTIC SIDESWAY BUCKLING IN THE EAST-WEST DIRECTION

Column Line Capacity- to- Load Capacity

Load Ratio (kips) (kips)

C 0,89 1216 1082

E 1,21 3466 4194

G 1.14 3325 3791

H 0.82 1614 1323

Total 9621 10390

175
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190
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192
Figure 7.2.13 Comparison between single slab reactions in

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196
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buckling capacity

197
Level 2

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198
Roof slab

Level 12

Level 11

Level 10

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the building

199
Level

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Figure 7.4.7 Soil pressure acting on basement wall - North side of building

200
8. EVALUATION OF CONSTRUCTION PRACTICES

8.1 INTRODUCTION

Comparisons of the laboratory tests and field tests with the project specifications are important considerations in identifying factors that may have contributed

to causing the collapse. These comparisons are included in this Chapter.

8.2 COMPARISON WITH PROJECT SPECIFICATIONS

8.2.1 Construction Materials

The project manual [5] and the structural drawings specified the following

material properties for the various components of the structure:

Column Steel - A 36 and A 572, Grade 50

Concrete Floor Slabs - 4000 psi (27.6 MPa)

Shear Wall Concrete - 4000 psi (27.6 MPa)

Post-Tensioning Strand - 7 wire strand, 270 ksi Low-lax (1862 MPa)

Welding Electrodes - E70 Series

The mill test reports and the chemical analyses and tensile coupon tests

reported in Section 5.3.2 verify that the column steel used in the structure

was A 36 and A 572 Grade 50 steel.

The strength of the concrete core samples in Section 5.2.2 exceeded the

specified 4000 psi (27.6 MPa) for both the floor slabs and the shearwalls

.

There was some variation in concrete strength between the various floor levels

tested.

The breaking strength of the 7 wire post- tensioning strand reported in Section

5.5.2 exceeded the 270 ksi (1862 MPa) value specified in only half the tests.

All the yield strength values in Section 5.5.2 exceeded the 37,170 lbf (165 kN)

value specified.

Observations of the debris at the site indicated that E7018 welding rods were

being used for the field welds. The chemical analysis of the weld material

reported in Section 5.4.3 also confirms this.

8.2.2 Foundations

Pertinent specifications for foundation construction are contained in the

Project Manual for L'Ambiance Plaza, [5] Sections 02010, "Subsurface

Investigation," 02210, "Site Grading," 02220, "Excavating, Backfilling and

Compacting," and in the General Notes: Foundations on structural drawing S201.

The general notes in structural drawing S201 address footing support in the

following notes:

The general notes in Drawing S201 require that all footings in the east and

west tower rest on undisturbed rock with a minimum bearing capacity of 14,000

psf (670 kPa)

.

The notes also make provision for the case where the material

201
below the bottom of the footing can not support the design load. Three options

are provided: replacement of the unsuitable material with approved engineered

fill; increase in the footing size; and lowering of the footing to soil of

suitable bearing capacity. These requirements, taken together with those in

the project manual, stipulate that the footings have to be poured on rock of

suitable quality, and if this cannot be accomplished at the elevation shown in

the plans, the footings either have to be lowered, or the gap between the

footings and the rock surface has to be filled with lean concrete. No other

"approved engineering fill" could be chosen by the soils engineer without

written approval from the architect. As an alternative solution, footing sizes

could be increased.

The borings and test pits indicated that most footings rested on a layer of

disintegrated rock, about 1 ft (0.3 m) thick. In addition, a layer up to 2 ft

(0.6 m) of local material was placed under some footings. Even though the

fill material at the bottom of the footings has been certified by the independent

testing laboratory to be capable of supporting 14,000 psf (670 kPa) , no record

is available to indicate whether written permission had been obtained from the

architect to deviate from the specifications, or whether the independent

testing laboratory ascertained the thickness of the disintegrated rock layer,

or otherwise performed tests or analyses to ascertain that the bearing

capacity of the footings was adequate.

Thus it is concluded on the basis of available information that support

conditions below the footings, as encountered in a number of soil borings and

test pits, do not comply with those stipulated in the specifications. This

conclusion applies to locations where boring and test pit information is

available. It does not necessarily apply to other locations. The conclusion

does not imply that the support condition of the footings in any way contributed

to the collapse, or that the footings would have settled excessively under

the conditions for which they were designed.

8.2.3 Column Sections

In studying the debris at the site and the landfill, measurements were made of

the cross sectional dimensions of a number of the column sections.

The columns measured were those from stage I (levels E, D, C and ground) and

stage IV (levels 5 and 6) where it was possible to identify their positions in

the structure. Fourteen columns from stage I in each tower were located. In

each case the measured dimensions corresponded to the specified dimensions [7]

for the sections in the column schedule in the structural drawings.

8.2.4 Lateral Bracing

A drawing by the lifting subcontractor provided to NBS by the Department of

Public Safety of the State of Connecticut (obtained from the construction

site) described lateral bracing to be used in the construction. The drawing

entitled "Concrete Deadman Layout and Details" was approved on 12/5/85. The

drawing shows the placement of deadmen in the vicinity of the exterior corners

of each tower. The stated purpose of the deadmen is:

202
"Deadmen for attaching guy cables to grade shall be provided

by the General Contractor prior to the arrival of the lifting

subcontractor. Guy cables are used to maintain plumbness on

the building frame during the lifting phase of erection."

It was further stated:

"Guy cable attachments to be provided in following slabs:

Tower: Roof, 10th fir., 8th fir., 5th fir., 3rd fir., 1st

fir., C Lvl; Parking: 4 Lvl. , B Lvl"

There was no evidence of these deadmen or guy cables at the site. One witness

interviewed following the collapse said this type of lateral bracing was not

used on this job. A review of the project manual and project correspondence

did not indicate any approval for elimination of this lateral bracing.

Shearwalls constructed as the slabs are lifted provide lateral support after

the concrete achieves a certain strength. The structural drawings note that

the shearwalls shall be 4000 psi (27.6 MPa) concrete and

"shear walls shall be cast so that no more than three equivalent

floors of height of lift slab structure shall be advanced above

cast top of shear wall"; also "Structure may advance when

shear walls have attained 75% of design strength"

This was not the case at the time of the collapse of the structure. On April

23, 1987 the shearwalls between the ground level and level 1 on the south side

of the building were being poured. The shearwalls on the north side of the

building were up to level 2 (figure 4.2.1). The lift slab erection for stage

IV had proceeded to five equivalent floors of height (level 1 to level 6)

above the cast top of the shearwall at level 1.

8.2.5 Welding

The project specifications include requirements for inspection and testing of

the shop welds and field welds under the section entitled "Structural Steel

Inspecting and Testing." Requirements for the shop welds include using

certified welders and conducting inspections and tests as required. Types and

locations of defects and work required and performed to correct deficiencies

were to be recorded. All welds were to be visually inspected. Tests for

fillet welds included liquid penetrant inspection and magnetic particle

inspection. For complete penetration welds, tests included radiographic inspection

and ultrasonic inspection. These tests were to be used at the testing

agency's option. Inspection requirements for field welds were the same as those for shop welds.

In an interview provided to the State of Connecticut Department of Public

Safety on June 24, 1987 the vice president of the steel fabricator indicated

that no testing was done on the shop welds; he noted the welders were certified,

that visual inspection was used as an ongoing process and that no ultrasonic

or other type of non -destructive tests were called for in the specifications.

It should be noted that the vice president of the structural engineering firm

203
of record for L' Ambiance Plaza waived the requirements for inspection of shop

welded steel under the following conditions:

° Welders must have current certification for types and position of weld

they are installing.

° Complete penetration welds must be shop inspected in accordance with the

specifications

.

° Fillet welds made in the shop shall be inspected when steel arrives in

the field in accordance with the specifications.

This waiver was transmitted to the general contractor by the structural

engineer of record on March 3, 1987 [13].

Information obtained by the State of Connecticut Department of Public Safety

from the independent testing laboratory providing inspection of the welds on

the job site, indicated the shop welds were visually inspected and the

complete penetration field welds were ultrasonically inspected.

The General Steel Notes listed on Drawing S301 contain the following provisions

that are relevant to the evaluation of the field-welded column splices:

4. All steel details and connections shall be in accordance with the

requirement of the AISC Specifications (latest edition) , including all

supplements and revisions.

6. Field splices shall be designed to develop the full capacity of member

at the point of splice in bending, shear and axial load (compression

and tension) unless otherwise noted.

Column splice details are shown on Drawing TCC-2. As drawn, these details do

not meet the requirements of the AISC Specification or of the AWS Structural

Welding Code for prequalified complete-penetration groove-welds.

The results from the tensile coupon tests described in Section 5.4.2 and the

fractographic analysis described in Section 5.4.4 raise questions about the

quality of some of the welds in the structure. In some cases the strength of

the welds was less than one would expect and definitely less than the strength

implied by Note 6 in the General Steel Notes (Drawing S301). Visual inspections

conducted in the course of this investigation indicated some of these welds did

not meet the requirements of the AWS Dl.1-83. It is not possible to determine

the frequency with which this occurred throughout the structure based on the

limited amount of weldment testing included in this investigation.

8.2.6 Post-Tensioning Tendon Detail

In the vicinity of column E4 . 8 , east-west tendons are splayed in such a way

they are separated from the column by over 5 ft. The American Concrete

Institute building code [2] appears to prohibit such a detail when it states

in section 18.12.4 that:

204
"A minimum of two tendons shall be provided in each direction through

the critical shear section over columns."

The report of ACI-ASCE Committee 423 [25]

,

also suggests such a detail should

be avoided. The committee states:

"Within the limits of tendon distribution that have been tested,

research indicates that the moment and shear strength of two-way

prestressed slabs is controlled by total tendon strength and by the

amount and location of nonprestressed reinforcement, rather than by

tendon distribution. While it is important that some tendons pass

within the shear perimeter over columns, distribution elsewhere is not

critical ..."

It may be inferred from this statement that a tendon layout such as that used

near column E4.8 does not satisfy the assumption that banded tendons pass

within the shear perimeter of the support.

The horizontal curvature of cables near column E4.8 is also questionable.

Committee 423 specifically discourages the use of horizontally curved, banded

monostrand tendons unless special reinforcement transverse to the axis of the

banded tendons is provided.

205
9. ANALYSIS OF COLLAPSE

9.1 INTRODUCTION

A number of possible failure mechanisms were considered prior to determining

the most probable cause of the collapse. These mechanisms are discussed and

the basis for considering them unlikely causes for triggering the collapse is

noted. The most probable cause of the collapse and the sequence of failure

events are then presented.

9.2 REVIEW OF POTENTIAL FAILURE MECHANISMS

The following failure mechanisms leading to collapse of the building were

considered: (1) lateral instability, (2) column instability, (3) floor slab

failure, (4) weld failure, (5) foundation failure (6) failure due to lateral

soil pressure and (7) loss of support of floor slab. Each of these mechanisms

are discussed in turn.

The potential contributions of problems that occurred with respect to the floor

slabs during construction and lifting as causes of the collapse were also

considered. The problems included: exposure of a post- tensioning tendon

during casting of a slab, accidental lifting of an extra slab and misplacement

of a lifting nut while lifting a slab. Information obtained from representatives

of the lifting contractor indicated that concrete was placed over the exposed

post- tensioning tendon, damaged concrete in the floor slab due to accidental

lifting of an extra slab was repaired using pressure injected epoxy and the

damage to the shearhead caused by misplacement of a lifting nut was not

serious. It was therefore considered unlikely any of these problems were the

cause of the collapse.

Lateral Instability

The lack of external lateral bracing and the fact that construction of the

shearwalls was behind the schedule called for on the erection drawings raises

the question of stability of the structure under lateral loads. The analysis

in Section 7.4.2 indicates that, by classical theory, the west tower had

essentially no reserve capacity against sidesway buckling. When finite member

size was accounted for in an approximate way, the reserve buckling capacity

increased to around 6. Lateral bracing provided by the shearwalls constructed

in accordance with the plans and specifications or other lateral supports (guy

wires, etc) would have been consistent with good practice. It was not likely,

therefore, that lateral instability was the main cause of the collapse.

Evidence supports this conclusion. Eyewitnesses did not report seeing the

building move laterally but rather collapse vertically toward the center of

each tower. They also did not report any perceptible lateral movement at the

time of collapse. Observations of the debris reported in Section 4.4 also did

not indicate any evidence of significant lateral displacement of the structure.

The displacements produced by the horizontal jack used to plumb the building

were small. The analysis in Section 7.4.3 indicated that the lateral load

required to reduce the effect of the restoring moment associated with load transfer

between the wedges exceeded the capacity of the horizontal jack. It is,

therefore, not likely that the use of this jack to plumb the building initiated

the collapse of the west tower.

206
Column Instability

The results in Section 7.3 indicate that instability of the individual columns

was not a problem. For the conditions prevailing at the time of collapse in

the west tower (figure 4.3.2), the reserve capacity or ratio of the buckling

loads for columns in stage IV (unsupported length - 30 ft 5 in) to the dead

weight of the roof and level 12 and level 9/10/11 ranged from 2.2 to 7.8.

Using the capacity reduction factor of 0.85 in the LRFD design procedure [8]

reduces these to a range of 1.9 to 6.6. The unsupported length required for

column buckling under these dead loads ranged from 45.6 ft (13.9 m) to 86.6 ft

(26.4 m) . As the failure progressed and the slabs began falling, column

buckling could have occurred as the floor slabs moved down the columns or separated from the columns, thus increasing the unsupported length. As this

occurs however, the slab dead load is removed, thereby reducing the probability

of column buckling.

Floor Slab Failure

The test results in Sections 5.2 and 5.5 indicate the concrete and posttensioning strand in the floor slabs conformed to the project specifications.

Also there was no indication that any freezing of the concrete had occurred in

the floor slabs during curing. It is unlikely, therefore, that failure of a floor slab due to inadequate material strength caused the collapse.

The post- tensioning tendon detail in the vicinity of column E4.8 was discussed

in Section 8.2.6. There are no indications that this detail led' specifically

to collapse of the structure, or to propagation of the collapse after it had

begun. Nevertheless, in light of the recommendations discussed previously, a tendon detail such as that employed near column E4.8 should be avoided until

its behavior and capacity can be established through research.

The floor slab at level C in the east tower in the vicinity of columns 9B, 9D

10B, and 10D had been damaged during lifting as noted in Section 3.4. This

damage had been repaired. It was discounted as the cause of the collapse

since eyewitness accounts clearly point out initiation of the failure in the

upper levels of the west tower. Note that one eyewitness, the concrete pump

operator (#3 in figure 4.3.3), was in the vicinity of the level C floor slab

near the center of the building and did not report noticing anything relating

to this slab at the initiation of failure of the building.

Weld Failure

The laboratory tests and metallographic examination presented in Section 5.4

raise questions regarding the quality of some of the welds in the structure,

both shop welds and field welds. A detailed inspection was made at the

landfill of the shop welds on the weld blocks and the tack welds on the wedges

supporting the roof and level 12 slabs on the west tower. Very few failures were observed of any of the welds on the weld blocks. Those that were found were

located in the lower levels and were caused by extreme loads. The tack welds

on the wedges obviously had failed since the roof and level 12 slabs failed as the building collapsed. The failure of these tack welds was considered a consequence of the overall failure of the building rather than the initial

207
cause of the collapse. Failure of a tack weld alone also would not cause the

wedge to fall out. The noise one would expect to be associated with failure

of a tack weld is much less than that reported by the eyewitness as occurring

at the initiation of the collapse. Other weld failures including failure of

the column splice at the ground level in column E4 . 8 were considered a result

of the collapse and not the cause. Foundation Failure

The results in Chapter 6 indicate there was no significant settlement or rotation of the footings supporting the columns. Representatives of the

lifting subcontractor also indicated there was no column for which the weld

blocks were consistently low when the wedges were placed under the slabs.

This would not have been the case had there been differential foundation

settlement. Foundation settlement was therefore discounted as a cause of the

collapse

.

Lateral Earth Pressure

The behavior of the basement wall on the north side of the building discussed

in Section 4.4.5 raises the question of the effects of lateral earth pressure.

The analysis in Section 7.4.4 indicates the structural elements, the connections

and structure as a whole were capable of resisting these loads.

As noted previously, there was no indication of overall lateral displacement

of the structure during collapse. Also, the displacement of the basement wall

probably did not occur until after removal of the debris. Examination of the

connection between the floor slab and the shearwall in the east tower near the

center of the building did not indicate any relative movement. Lateral earth

pressure was also discounted as the cause of the collapse.

Loss of Support of Floor Slab

Tests of the lifting assembly reported in Section 5.5.5 indicated the floor

slabs could lose support due to combined bending and tension failure of the

jack rod or due to the lifting nut slipping off the lifting angle of the

shearhead. The likely location for such a failure would be the most heavily

loaded jack. As shown in figure 7.2.10a, for a level slab the jack on column

E4.8 had the largest load (152 kips - 676 kN) . The failure load for the

lifting assembly was estimated to range from 170 kips (756 kN) to 200 kips

(873 kN) . Based on the 152 kip (676 kN) load on column E4 . 8

,

the reserve

capacity against loss of support of the slab ranges from 1.1 to 1.3. Note

that the 200 kip (873 kN) load exceeds the capacity (180 kips - 801 kN) of the

lifting jack discussed in Section 7.2.2. This possibility of loss of support

of the floor slab causing collapse of the building will be discussed further

in Section 9.3.

Summary

The reserve capacity or the ratio of ultimate load to the applied load provides

a basis for comparison of the various failure mechanisms. A summary of these

reserve capacities is given in table 9.2.1. The reserve capacities for a

208
punching shear failure of the slab are the highest and this failure mechanism

was not considered to be a likely cause of the collapse. The reserve capacities

for loss of support of the floor slab and lateral instability of the frame are the lowest. As discussed previously for each of the failure mechanisms, these

reserve capacities need to be considered together with other factors (eyewitness

accounts, orientation of debris) in determining the most probable cause of the

collapse

.

9.3 PROBABLE CAUSE OF COLLAPSE

Based on the review of possible failure mechanisms described in Section 9.2,

the most probable cause of the collapse of the L' Ambiance Plaza building was

loss of support of the package of floor slabs 9/10/11 along column line E in

the west tower. More specifically, it is considered likely that this loss of

support originated at either column E3 . 8 or E4.8.

At the time of the collapse temporary wedges were being installed under floor

slab 9 (Elev. 51 ft 9 in-17 m) at column E4.8, and it is probable that the

installation of these wedges required some adjustment of the elevation of the

slab package at that column. Although their sections were heavier than that

of column E4.8 and they were equipped with 300 -kip (1334 kN) jacks, each of

columns E3 , E3 . 8 , G3 and G4 carried less load than did E4.8. The nominal

reaction at E4.8 due to slabs 9/10/11 was 152 kips (676 kN) which, for the

erection stage, made it the most heavily loaded column in the west tower.

Using the computer model described in Section 7.2.1, it was determined that

increasing the elevation of the slab package at column E4.8 by one full jack

stroke of 0.50 in (12.7 mm) from its "as cast" position would involve a jack

load of 201 kips (894 kN) . Employees of the lifting subcontractor testified

that the line pressure normally supplied by the system console ranged from

2300 to 2500 psi (15.9 to 17.2 MPa) while lifting a package of 3 slabs and

that the maximum line pressure attainable was approximately 3000 psi (20.7

MPa). Allowing for a friction loss of 5 percent, the corresponding jack

forces at column E4 . 8 normally would be 137 to 149 kips (609 to 663 kN) and

the maximum force would be 179 kips (796 kN) . Thus the jack at column E4.8

would have had slightly less than the required lifting capacity when operating

at the upper end of the normal pressure range. It would not have developed the

force required for an upward adjustment of 0.50 in (12.7 mm) from the "as

cast" position, even when operating at the maximum possible line pressure.

Note that the limiting jack load of 179 kips (796 kN) falls within the range

of failure loads of 170 to 200 kips (756 to 889 kN) for Type P50 shearheads as discussed in Section 7.2.

The top of column E4.8 is shown in figure 9.3.1. The form of the indentations

suggests that one of the jack rods lost its load, allowing the jack to roll

off the column top and chamfering the edge of the web. An examination of the

column showed the existence of lifting nut impact marks on each side of the

web at approximately 52 in (1.32 m) below the top of the column. These impact

marks are an indication that each of the lifting nuts "kicked out" from under

the lifting angles on the shearhead at slab 9. The bottom surfaces of the

lifting angles are shown in figure 9.3.2. The bearing surface on the left

clearly shows the lifting nut contact area. The contact area on the right is

209
not so sharply defined, but there is evidence that the lifting nut kicked out

from under the lifting angle.

Whether loss of the jack at column E4.8 was the cause of the collapse or simply one of its effects cannot be established with certainty, but there is

at least one other sequence of events that could have produced the same result. If an upward adjustment exceeding the capability of the jack at

column E4.8 was required, it is possible that the 300-kip (1334 kN) jack at

column E3 . 8 was used to effect this adjustment. Again assuming an operating

line pressure of 2500 psi (17.2 MPa), the jack at column E3 . 8 could have

developed a lifting force of approximately 2x149 - 298 kips (1.33 MN) . A

full -stroke upward displacement of 0.50 in (12.7 mm) at column E3 . 8 would have

developed a reaction of about 190 kips (845 kN) and, therefore, the jack could

easily have produced this force within the limit of its stroke.

As has been noted in the discussion of test results in Section 5.5.5, a jack

load of 190 kips (845 kN) acting on the type X51 shearheads of column E3 .

8

would produce very significant rotations of the lifting angles, possibly

leading to kick out of one or both of the lifting nuts. The bottom surfaces

of the lifting angles of the shearhead at slab 9, column E3 . 8

,

are shown in

figure 9.3.3. The marks on these surfaces show that kick out did indeed

occur. This sudden release of energy would be expected to raise dust from the

surfaces of the concrete slabs and the noise associated with the kick out

experienced in the lifting assembly tests described in Section 5.5.5 could

readily be described as steel breaking under stress. This is consistent with

the eyewitness accounts (#3 and #4 - figure 4.3.1). The surviving witness

working at column E4.8 at the time of the collapse (#14 - figure 4.3.1) noted

that the first indication of a problem was a loud noise either right above his

head (E4.8) or more towards the center of the slab (E3.8). Due to load

redistribution, loss of a jack at either E3 . 8 or E4.8 would lead to failure of

the other jack. As floor slabs 9/10/11 moved downward, the shears and moments

in the slabs adjusted to the new support conditions. Because the posttensioning tendons now were located in the compression side of the slabs, the

tendons adjacent to columns E3 .

8

and/or E4 .

8

became ineffective and flexural

cracks developed in the bottom face of the slabs. This explains the sudden

appearance of cracks in the slab directly above the workmen who were installing

wedges at column E4.8. The underside of shearhead 10, column E3 . 8

,

is shown

in figure 9.3.4 and that of shearhead 11 is shown in figure 9.3.5. It appears

that one of the lifting nuts glanced off of the web and became partially lodged

under shearhead 10. As the slabs moved downward due to loss of support at

column E3 . 8

,

the lifting nut was again kicked out from under a lifting angle.

The imprint on the underside of shearhead 11 suggests that this process was repeated for a third time. It was not possible to confirm this sequence of

events with lifting nut impact points on the web because the top section of

column E3 .

8

could not be positively identified. The likely progression of the

collapse through the upper levels of the west tower is described in the

following paragraphs

.

In order for load redistribution to have taken place, the slab would have had

to redistribute forces after the loss of support at a single column. To determine

if the slab capacity was sufficient to allow redistribution to take place

following loss of support at a single column, yield line analyses were

210
performed considering loss of support at either column E3 . 8 or E4.8. For each

analysis, it was conservatively assumed that force in all post- tensioning

tendons would be ten percent less than the tendon force used for design of

slabs. Positive moment capacity of the slabs was neglected for each analysis.

Simple yield mechanisms and a virtual work solution method were used to

determine distributed load capacity of the slabs. These analyses showed

clearly that the floor slabs would be able to distribute loads much larger

than the dead weights of the slabs, regardless of whether support was lost

initially either at column E3 . 8 or column E4.8.

9.4 PROBABLE SEQUENCE OF COLLAPSE

It is difficult to accurately predict how the initial failure propagated

through the structure because of the unknown dynamic loads and load redistribution

that occurred during the collapse. The failure sequence postulated is a best

estimate based on eyewitness accounts and observations of the debris.

The probable sequence of collapse is indicated in figure 9.4.1. With the loss

of support of slabs 9/10/11 at column E3 . 8 and/or E4.8, it is likely that the

remaining jacks on this line shed their load by kick out of one or both

lifting nuts as the loads were redistributed. Calculations of the ultimate

strength of the floor slab with the loss of support at either column E4.8 or

E3 . 8 using a yield line approach indicated the slab could support its own

weight if only one of these supports were lost. For the initial failure to

progress, therefore, it was necessary for additional supports to be lost.

This loss of support along the E line is in agreement with observed marks on the

stage IV column extensions and on the shearhead lifting angles (See Appendix

A, columns El, E2 and E3) . It is also likely that some loss of prestress in

the tendons along and normal to E line accompanied this load redistribution.

Calculations indicate, for example, that loss of the drape in the tendons in

the east-west direction would decrease the post- tensioning by 10 to 15

percent. At approximately this same time a negative moment crack would have

developed along the length of the west tower just to the north of G line due

to the clear span between lines C and G. With loss of support along the E

line, the span of the slab in the north- south direction increases by 140

percent (C-E-G-H to C-G-H) . The negative moment at the G line about an axis

parallel to the G line in the east-west direction increases by a factor of

approximately 3.6. Analysis of a unit width of slab indicates this negative

moment exceeds the ultimate moment capacity of the slab (14.6 kip -ft/ft vs 9.6

kip-ft/ft) (64.9 vs 42.7 kN-m/m) . On the north side of E line, slabs 9/10/11

caused bending in the C-line columns with portions of the floor slabs being

held in place by the conventional reinforcing around the columns. This is clearly

illustrated by the distortion of shearheads 9/10/11 at column C4.8 as shown in

figure 4.4.34. Also, it is apparent from figure 4.4.34 that during the

progression of failure of slabs 9/10/11 along column lines C and E, the two

slabs at the top of the building (12/roof) remained intact. This is the

reason for the reverse curvature at the very top of column C4.8. Along column

line E the falling debris from floor slabs 9/10/11 caused the progressive

failure of floor slabs at the lower levels with the result that shearheads on

the interior columns along this line ended up closely stacked above level 1.

211
A different mechanism is believed to have been involved with the progression

of the failure along G line and south to column line H. Neither column lines

G or H involved the failure of jacks as slabs 9/10/11 had been temporarily

wedged and the lifting nuts had been removed from the jack rods in preparation

for the next lift. An inspection of the column tops (stage IV) from these two

lines revealed that very little scraping or gouging occurred during the

collapse and, with the exception of the columns in lines 1 and 6, the shearheads

released cleanly and slid straight down. At columns G6 , HI and H6 the

shearheads at the upper levels did not release cleanly and some local bending

of columns resulted. To a lesser degree, this also happened at column H2

.

The bending moment about G line, and eventually about H line, would have led

to progressive loss of the wedges under floor slab 9. This action would have

been augmented by the racking forces produced by the transverse (north- south)

tendons with the effect that wedges also were lost under floor slab 12 along

column lines G and H. The analysis in Section 7.4.2 also indicated that' the

slab would lift off one wedge at each column at small lateral displacements

.

This lift off could be followed by the wedges being lost.

Laboratory tests conducted on a column segment with a pair of weld blocks as described in Section 5.5.4 showed that the tack welds fail at small increases

in load and that a wedge can be "rolled out" from under a shearhead at a load

on the wedge of from 100 to 150 kips (445 to 667 kN) with very little distortion

of the weld block. This is consistent with the marks on the stage IV columns

recovered for column lines G and H between lines 2 and 5. It is also consistent

with the orientation of the floor slabs in this same region as can be seen from

photographs taken shortly after the collapse and prior to any substantial

removal of debris (Section 4.4.1). The absence of scrapes and gouges on the

stage IV column extensions along these two lines and the extensive damage to

the stage IV extensions along C line strongly suggest that failure of the

upper two floor slabs (12/roof) initiated in the south sector of the west

tower

.

The failure sequence described is consistent with several of the eyewitness

accounts. The surviving witness working at column E4.8 at the time of the collapse

noted that after he heard the loud bang, the ceiling directly over his head

(underside of slab 9) was cracking just like ice breaking. Loss of support at

E3 .

8

and/or E4.8 would account for this. Another witness (#11 in figure

4.3.1) located approximately 100 yd (91 m) south of the site heard several

loud noises in rapid succession prior to the collapse. Other witnesses in the

building offered similar testimony. This succession of noises is consistent

with the progression of the lifting assembly failures along E line.

Cracking of floor slabs 9/10/11 between the E line columns and cracking along

the G line involves failure in the center portion of the building. This

agrees with the discussion in Section 4.4.1 in which the debris formed a heap

in the center of each tower. Note the post- tensioning cables also tend to pull

the columns inward in this fashion. The following witness observations also

relate to failure of the center portion:

212
1. Witness No. 10.- located on the southside of the building on the

parking garage

0 "The roof floors caved into the center of the slabs and they

all fell on top of the other."

2. Witness No. 13 - jumped from ground floor and observed collapse

while running from building

0 Observed two columns, the column at the extreme west corner

(Cl) and the column next to it (C2) on the Washington Ave

.

side bend towards the center of the building. He then

observed the center of the slab drop and the edges of the

slab raise up.

3. Witness No. 15 - welding wedges on level 3 or 4 at southeast

comer of the east tower.

0 Heard a loud cracking sound like a rifle shot magnified

0 Observed large pieces of flooring coming down in the west

tower

.

The concrete pump truck operator's observations given in Section 4.3 relate to

the collapse of the upper floor slabs south of the G line. He noted the two

uppermost slabs in the vicinity of the southwest comer (column HI) started

downward and then slowed momentarily (struck slabs 9/10/11) . The collapse then

spread eastward and northeastward throughout the uppermost slabs and then the

building collapsed vertically.

The most probable cause of the initiation of the collapse of the east tower is

less clear. Several possibilities exist. The possibility that debris from

the west tower attacked the columns along line 7 was considered. Observation

of the columns did not indicate this occurred. Three other mechanisms appear

equally possible. First, the pour strips in place at the time of collapse

could have transmitted destructive forces from the west tower into the east

tower. Counterclockwise twisting of several of the columns in the lower

levels of the east tower and a force directed from west to east on the

exterior shear wall on the south face of the east tower were noted in Section

4.4.1. These observations are consistent with this possibility. Second,

falling debris near the west end of column line D could have damaged the

tendon anchorages in the east tower floor slabs at this location. With loss

of anchorage in one or more floor slabs, the progression of failure would have

been very similar to that described for column line E in the west tower. One

eyewitness, the concrete pump truck operator located near the southeast corner

of the west tower, reported the east tower collapsed as a result of being

struck by sections of the collapsing west tower. Third, forces exerted by

falling debris from the west tower could have caused lateral displacements of

the east tower large enough to overcome the restoring moments discussed in

Section 7.4.2. Loss of wedges supporting the floor slabs would follow. This

loss of support could propagate through the floor slabs in the east tower in a manner similar to that discussed for the slabs at level 9/10/11 in the west tower.

213
Collapse of the east tower could have involved a combination of these three

possibilities. Clearly it becomes more difficult to identify the exact

sequence of events as failure progresses and force redistribution within the

structure becomes more complex. The probable cause of the collapse discussed

in Section 9.3 and the eyewitness accounts make it clear, however, that the

collapse started in the west tower and progressed to the east tower.

214
TABLE 9.2.1

SUMMARY OF RESERVE CAPACITIES FOR VARIOUS FAILURE MECHANISMS

Condition Investigated Failure Mechanism Reserve Capacity Rang

Floor slab

dead weight

Punching shear

(Section 7.2.2)

4.0-23.1

Lateral Earth Pressure

on Basement Wall

Shear failure of

Shear wall

(Section 7.4.4)

Connection between

floor slab and shear

wall (Section 7.4.4)

1.4

1.7 - 1.9

Sliding of Shear

wall footing

(Section 7.4.4)

1.3-2.

7

Stability

Buckling of

Individual Column

(Section 7.3)

2. 2-7.

8

Lateral Stability 1.

of Frame

(Section 7.4.2)

, 1 (5.6)** (East- West)

1.2 (North- South)

Loss of Support

of floor slab

Jack rod failure,

bottom nut sliding

off (Section 5.5.5)

1.1-1.

3

* Calculated or measured ultimate load

Applied load on structure

** based on restoring moment effect

215
it#,,.

Figure 9.3.1 Indentation in top of column E4.8

* r.

*2! 31 5' . . t

.

216
217
*4. m ; - 'VvV* «Jfc

' « -..' £ft f 'r£u$M& S'*

«"*>, >- * . -r

.

218

Figure

9.3.3

Indentations

in

underside

of

lifting

angles,

column

E3.8,

level 9
Figure 9.3.4 Indentation in underside of lifting angle,

column E3.8, level 10

219
Figure 9.3.5 Indentation in underside of lifting angle,

column E3.8, level 11

220
221

Figure

9.4.1

Probable

failure

sequence
10. CONCLUSIONS

The following conclusions are based on the laboratory and field tests,

computer analyses, the witness interviews and review of the project documentation

conducted by the National Bureau of Standards during the course of the

L' Ambiance Plaza building collapse investigation:

1. The most probable cause of the collapse was failure of the lifting

system in the west tower during placement of a package of three upper

level floor slabs. The failure most probably began below the most heavily

loaded jack (column E4.8) or an adjacent jack (column E3.8). Excessive

deformations occurred in the lifting angle of the shearhead at the

location of the initial failure. This was followed by one of the jack

rods in the lifting assembly slipping off the lifting angle in the

shearhead supporting the package of three slabs. This failure

mechanism was duplicated in laboratory experiments. The local failure

propagated as loads were redistributed and the remaining jack rods

along column line E slipped off the lifting angles and the package of

three slabs failed in flexure and shear. These slabs fell, causing

the lower level slabs to fail. This resulted in the collapse of the

entire west tower. Consequently, the east tower collapsed due to one or more of the following factors: (a) forces transmitted to it by the

west tower collapse, (b) damage to the unbonded post- tensioning

tendons caused by falling debris from the west tower, or (c) lateral

instability caused by falling debris from the west tower.

(a) Calculations showed that these supports (hydraulic jacks at the time

of collapse) were the most heavily loaded in the west tower.

(b) An eyewitness account of the slab behavior at this location at

the time of failure is consistent with loss of support.

(c) This failure mechanism occurred in laboratory tests simulating

the conditions existing in the structure at the time of collapse.

The failure resulted in a loud noise in the laboratory tests

which is consistent with the eyewitness reports.

(d) The computed reserve capacity for this failure mechanism is one of the lowest of the possible mechanisms analyzed.

2. The quality of materials in the structure was generally in accordance

with the project plans and specifications and did not play a significant

role in initiating the collapse.

(a) Tensile properties and chemical composition of the column steel

satisfied the ASTM requirements for A 36 and A 572 Grade 50 steel.

(b) The results of tests of cores from the floor slabs and shear

walls indicated the compressive strengths of the concrete met the

project plans and specifications.

222
(c) Yield strength and breaking strength values of the post -tensioning

strand generally satisfied the ASTM requirements for steel

strand.

3. There were a number of deviations from the project plans and specifications

in the structure as built, but the investigation indicated these

deviations did not play a significant role in initiating the collapse.

(a) Data from borings and test pits indicated a number of footings

rest on layers of fill and disintegrated rock rather than "rock of

suitable quality" or lean concrete.

(b) The strength of some of the welds was less than the strength

implied in the structural drawings.

(c) Column splice details shown on the plans do not meet the requirements

of the AISC Specification or of the AWS Structural Welding Code

for prequalified complete-penetration groove-welds.

(d) Lack of joint penetration and large amounts of porosity were

observed in some of the field welds.

4. It is unlikely that the horizontal jack used to plumb the structure

initiated the collapse.

(a) Displacements produced by the jack in plumbing the building were

small

.

(b) The maximum load the jack could apply to the structure was not

sufficient to cause inelastic action.

5. The reserve capacity against lateral instability was small. It does not

appear, however, that lateral instability was the initial cause of the

collapse. Inadequate resistance to lateral instability may have

caused the collapse of the east tower.

(a) Deadmen and guy cables whose purpose would have been to plumb the

building and provide lateral bracing were not used in the

construction.

(b) The height of the structure above the cast top of the shear walls

at the time of collapse was greater than three equivalent floors

of height.

(c) Eyewitnesses did not report seeing or feeling any lateral

movement at initiation of the collapse.

(d) Observations of the debris did not indicate any overall lateral

displacements or drift in the collapse.

223
6. It is unlikely that lateral earth pressure acting against the basement

wall on the north side of the structure caused the building to collapse.

(a) Calculations indicated that the capacities of: (1) the connections

between the floor slabs and shearwalls, (2) the shearwalls and

(3) the resistance of the shearwalls to sliding over their base

were sufficient to resist the lateral soil pressure.

7. It is unlikely that differential foundation settlements caused the

building to collapse.

(a) Data from the subsurface explorations indicate that differential

foundation settlements not exceeding 3/4 in (19 mm) could have

occurred during construction. These are not considered excessive

for this type of structure. The greater part of these settlements

occurred early in the erection process.

224
11. ACKNOWLEDGEMENTS

Regional and area officers of the Occupational Safety and Health Administration

provided continued support throughout this investigation. The OSHA team

posted in Bridgeport, Connecticut following the collapse were very helpful in

handling the logistics for the NBS investigation, obtaining documents relating

to the collapse and dealing with legal issues. The support of John Miles

Regional Administrator, Ronald Morin and John Stanton, Area Directors and the

following OSHA staff: Joseph Normand, Edward Eagan, William Freeman, Romas

Bossone, Thomas Guilmartin and Kang Yi is gratefully acknowledged.

The Department of Public Safety of the State of Connecticut provided copies of

the documents relating to the construction including construction logs

,

testing reports, design drawings and shop drawings. Lieutenant Edmund

Dunstone, Executive Officer, Bureau of State Fire Marshal and Safety Services

and Leo Belval, State Building Inspector supplied this information and provided

copies of video tapes of the collapse site taken during the rescue efforts and

interviews of witnesses taken by the State Police during the subsequent

investigation.

Morris Schupack, Mario G. Suarez, Dr. Randall W. Poston and Daniel J. McCarthy

of Schupack Suarez Engineers Inc. in Norwalk, Connecticut participated with NBS

in conducting the investigation. They provided field support in data collection,

interpretation of design and construction practices relating to lift slab

construction and analysis of the post- tensioned concrete slabs.

Clive DeVault from Schnabel Engineering Assoicates conducted the subsurface

exploration, and together with Raymond deStephen, prepared the geotechnical

engineering report in Appendix B. Connecticut Test Borings, Inc. performed

the test borings and J. E. Barrett and Sons dug the test pits.

Neil Moreton, N.S.M. Co. Inc., provided an assessment of the welding practices

and quality of the welding details.

T. Robert Shives , Dr. Richard J. Fields, Charles E. Brady, Leonard C. Smith

and Dr. Thomas Siewert of the NBS Institute for Materials Science and Engineering

and John Norris of the National Measurement Laboratory provided assessments of

the material properties of welding details. Dr. William Stone assisted in

developing the computer models and in conducting the structural analyses. Stuart

Foltz prepared the drawings of the column deformations in Chapter 4. Frank

Rankin, Jim Little and Herb Wechsler assisted in performing the structural

laboratory tests. Dr. Nicholas J. Carino and Dr. H. S. Lew provided a comprehensive review of the report. Nancy Fleegle typed the report with great

care

.

225
12 . REFERENCES

1. Lew, H.S., Fattal, S. G., Shaver, J. R. , Re inhold, T. A., Hunt, B. J.,

"Investigation of Construction Failure of Reinforced Concrete Cooling

Tower at Willow Island," West Virginia, NBSIR 78-1578, National Bureau of

Standards, Washington, D.C., November, 1979.

2. "Building Code Requirements for Reinforced Concrete," ACI 318-83, American

Concrete Institute, Detroit, MI, 1983.

3. "Lift Slab - The Modern Construction Method, Quality- Speed- Economy ,

Texstar Construction Corporation Publication, undated.

4. Carino , N. J., and Lew, H. S., "Re-examination of the Relation Between

Splitting Tensile and Compression Strength of Normal Weight Concrete ," .ACI

Journal, Proceedings Vol. 79, No. 3, May- June 1982, pp. 214-219.

5. "L'Ambiance Plaza Residential Apartment Building, Project Manual," TPM

International Inc., Darien, CT, undated.

6. "Annual Book of ASTM Standards," Vols. 01.01, 01.04 AND 04.02, American

Society for Testing and Materials, Philadelphia, PA, 1986, 1987.

7. Manual of Steel Construction . American Institute of Steel Construction,

8th Edition, Chicago, IL, 1981.

8 . Load and Resistance Factor Design Specification for Structural Steel

Buildings . American Institute of Steel Construction, 1st Eddition,

Chicago, IL. , September 1, 1986.

9. "Stress Analysis Module Primer," Publication #D005 , PDA Engineering, 1560

Brookhollow Drive, Santa Ana, CA, Jan. 1985.

10. Lin, T. Y. , "Load-Balancing Method for Design and Analysis of Prestressed

Concrete Structures," Journal of the American Concrete Institute,

Proceedings, V. 60, No. 6, Detroit, MI, June 1963, pp . 719-742.

11. "PATRAN User's Guide," PDA Engineering Software Products Division, 1560

Brookhollow Drive, Santa Ana, CA, undated.

12. Gross, J. L. , Hambacher, D. G., "Analysis of Partially Erected Structurers"

,

Proceedings of the 8th Conference on Electronic Computation, ASCE, Houston,

Texas, American Soc. of Civil Eng., New York, NY, Feb. 1983.

13. Correspondence from structural engineer to general contractor dealing with

inspection of shop welded steel, dated March 3, 1987.

14. American National Standard A58. 1-1982, "Minimum Design Loads for Buildings

and Other Structures," American National Standards Institute, 1430

Broadway, New York, NY 10018.

226
15. Russillo, Michael A., "Lift Slab Construction: Its History, Methodology,

Economics and Applications", unpublished draft report.

16. Deere, D. U. and Miller, R. P. , "Classification and Index Properties of

Intact Rock," Tech. Special Report AFWL-TR-65, AF Special Weapons Center,

Kirtland AFB, 1966.

17. Gardner, W.S., "Design of Drilled Piers in the Atlantic Piedmont,

Foundations and Excavations in Decomposed Rock of the Piedmont Province,"

Geotechnical Special Publication No. 9, ASCE, New York, NY, April 1987.

18. Martin, R. E. , "Settlement of Residual Soils, Foundations and Excavations

in Decomposed Rock of the Piedmont Province," Geotechnical Special

Publication No. 9, ASCE, New York, NY, April 1987.

19. Bowles, J. E. , "Elastic Foundation Settlements on Sand Deposits," J. of

Geotechnical Engineering, ASCE, Vol. 113, No. 8, August, 1987, p. 854.

20. deMello, V., "The Standard Penetration Test - A State-of- the -Art Report,"

4th Pan Am. Conference on Soil Mechanics and Foundation Engineering,

Puerto Rico, 1971.

21. Meyerhof, C. G. , "Penetration and Bearing Capacity in Cohesive Soils," J.

Soil Mechanics and Foundation Div. ASCE, Vol. 82, No. SMI, Jan. 1956.

22. "Lift-Slab Concrete Construction," Data Sheet 514, Revision A, National

Safety Council, 444 North Michigan Avenue, Chicago, IL, 1974.

23. Jaky, J., "Pressure In Silos," 2nd Conference of ICSMFE, Vol. 2, pp 103-

107, 1948.

24. Bowles, J. E. , Foundation Analysis and Design . 3rd Edition, McGraw Hill,

1982.

25. "Recommendations for Concrete Members Prestressed with Unbonded Tendons,"

ACI-ASCE Committee 423, ACI 423.3R-83, 1983, 16 pp

.

26. ANSI/AWS Dl.1-83, Structural Welding Code-Steel, American Welding Society,

Miami, Fla.

227
APPENDIX A

Detailed Observations of Stage IV Columns - West Tower

228
APPENDIX A

DETAILED OBSERVATIONS OF STAGE IV COLUMNS - WEST TOWER

COLUMN NO

SECTION

SHEARHEAD TYPE

LENGTH

MARK

TACK WELDS

SHEARHEADS

COMMENTS

:

=- Cl - HP 10X42 - A50

- 19' -9” (FRACTURED AT STAGE III/IV SPLICE) - 1C - 1 SET ABOVE TOP WELD BLOCK

- NONE

GRADUAL BEND TO SOUTH ABOUT WEAK AXIS AT LEVEL OF TOP SEAL BLOCK.

SHEARHEADS 9-R APPEAR TO HAVE SLID OFF TOP OF COLUMN. UPWARD

(NORTH SIDE) AND DOWNWARD (SOUTH SIDE) BITE MARKS ON FLANGE EDGES

NEAR TOP OF COLUMN DUE TO SHEARHEAD RACKING. COLUMN BUCKLED

APPROX 3 FT ABOVE LEVEL 1. LIFTING NUT INDENTED WEB ON SOUTH

SIDE (NORTH SIDE WAS NOT INSPECTED) . THREAD MARKS FROM JACK ROD

ON NORTH SIDE OF WEB AT COLUMN TIP AND JACK FOOTPRINT ON TOP OF

COLUMN INDICATE JACK ROLLED OFF OF COLUMN TOP TO THE SOUTH. NO

VISIBLE DAMAGE TO WEDGE CONTACT SURFACES ON UPPER WELD BLOCKS.

COLUMN NO

SECTION

SHEARHEAD

LENGTH

MARK

TACK WELDS

SHEARHEADS

COMMENTS

:

- C2 - HP 12X53

TYPE - C50 - 19' -2" - 2C - 1 SET ABOVE TOP WELD BLOCK

- 5 IN PLACE (9-R)

TWO SHEARHEADS ARE LOCATED BETWEEN THE TACK WELDS AND COLUMN TOP.

THREE SHEARHEADS ARE LOCATED DIRECTLY BELOW TACK WELDS. COLUMN

BUCKLED APPROX 3 FT ABOVE LEVEL 2. SLIGHTLY DEFORMED HEADER

CHANNEL OF SHEARHEAD AT LEVEL 11 SUGGESTS CANTILEVER ACTION OF

FLOOR SLAB. MARKS FROM LIFTING NUTS ARE VISIBLE ON UNDERSIDE OF

SHEARHEAD AT LEVEL 9. EAST SIDE HAS ONLY A ROUND MARK (NO

INDENTATION) WHILE THE WEST SIDE IS INDENTED SOMEWHAT LESS THAN

1/16 INCH. EAST SIDE OF WEB INDICATES SOME SCRAPING BY LIFTING

NUT, BUT CONTAINS NO CLEARLY DEFINED INDENTATION DUE TO IMPACT.

A JACK FOOTPRINT IS VISIBLE ON THE TOP SURFACE, ALONG WITH SMALL

GOUGES ON WHAT IS BELIEVED TO BE THE NORTH FLANGE.

COLUMN NO

SECTION

SHEARHEAD

LENGTH

MARK

TACK WELDS

SHEARHEADS

COMMENTS

:

- C3 - W 12X65

TYPE - W50

- 20 '

-

8

" - 3C-4L-11

- 1 SET ABOVE TOP WELD BLOCK

- 3 IN PLACE (9-11)

COLUMN IS BENT ABOUT STRONG AXIS APPROX 1 FT ABOVE THE STAGE III/IV

SPLICE. GRADUAL BEND TO EAST ABOUT WEAK AXIS NEAR TOP OF SEAL

BLOCK. THE THREE SHEARHEADS (9, 10, 11) ARE CLUSTERED BETWEEN

TOP WELD BLOCK AND LEVEL 6. SHEARHEADS 12 & R APPEAR TO HAVE

SLID OFF TOP OF COLUMN. GOUGES IN NORTH FLANGE START AT TOP OF

229
SEAL BLOCK AND EXTEND UPWARD 7 3/8 INCHES. THESE APPEAR TO HAVE

BEEN CAUSED BY THE LEVEL 10 SLAB DURING LIFTING. THE HEADER

CHANNEL ON THE NORTH SIDE OF ALL THREE SHEARHEADS IS BENT

OUTWARD, INDICATING CANTILEVER ACTION OF THE FLOOR SLABS. THE

THIRD SHEARHEAD FROM THE TOP (LEVEL 9) SHOWS IMPRINTS OF THE LIFTING

NUTS IN UNDERSIDE OF THE LIFTING ANGLES. THE EAST SIDE OF THE

SHEARHEAD SHOWS A DOUBLE NUT IMPACT AS THE LIFTING NUT WORKED

TOWARD THE EDGE OF THE LIFTING ANGLE. THE WEST SIDE SHOWS A

CONTINUOUS GROOVE CAUSED AS THE LIFTING NUT SLID OFF THE LIFTING

ANGLE. THE TIPS OF THE WEST ROD SLOT ARE BENT UPWARD APPROX 1/2

INCH. ALL THREE SHEAR HEADS HAVE A TEAR IN THE WELD BETWEEN THE

LIFTING ANGLE AND THE HEADER CHANNEL. JACK ROD INDENTATION IS

VISIBLE AT THE JUNCTURE OF THE WEST SIDE OF THE WEB AND THE NORTH

FLANGE . FOUR UPWARD GOUGES AND FOUR DOWNWARD GOUGES ON EDGES OF

FLANGES BETWEEN TOP WELD PLATE AND TOP OF COLUMN DUE TO SHEARHEAD

RACKING. COLUMN BUCKLED JUST BELOW LEVEL 2.

COLUMN NO

SECTION

SHEARHEAD TYPE

LENGTH

MARK

TACK WELDS

SHEARHEADS

COMMENTS

:

- C3.8

- W 12X65

- G50

- 6

'

2

"

- 1 SET ABOVE TOP WELD BLOCK

- 3 IN PLACE

DEFORMED SHAPE OF COLUMN SUGGESTS CANTILEVER ACTION BY SIABS 9-11

WHILE SLABS 12 -R REMAINED IN PLACE. THE TOP OF THE COLUMN

CONTAINS A JACK FOOTPRINT, DEEPER ON THE EAST SIDE THAN ON THE

WEST. THERE ARE NO VISIBLE INDENTATIONS IN THE WEB DUE TO

LIFTING NUT IMPACT.

COLUMN NO

SECTION

SHEARHEAD TYPE

LENGTH

C4 .

8

HP 12X53

C50 (ROOF - L50)

17' -6"

MARK

TACK WELDS

SHEARHEADS

COMMENTS

:

- C - 1 SET ABOVE TOP WELD BLOCK

- 4 IN PLACE (9-12)

TWO SETS OF GOUGE MARKS ON NORTH FLANGE AT SEAL BLOCK AND JUST ABOVE

SEAL BLOCK ARE ASSUMED TO HAVE BEEN CAUSED BY SLABS 9 AND 10

DURING JACKING OPERATION. FOUR SHEARHEADS REMAIN ON THIS

SECTION, TWO BELOW THE SEAL BLOCK AND TWO ABOVE. LOCAL BENDING

OF COLUMN FLANGES AND HEADER CHANNEL DUE TO CANTILEVER ACTION OF

SLABS 9-11. THIS RESULTED IN GRADUAL BEND IN COLUMN ABOUT THE

STRONG AXIS AT LEVEL 6. TOP SECTION WAS RESTRAINED BY SLABS 12-

R, PRODUCING A REVERSED CURVATURE ABOVE LEVEL 6. THE FOURTH

SHEARHEAD FROM TOP SHOWS MARKS FROM LIFTING NUTS AND IS BELIEVED

TO BE LEVEL 9. NUT IMPRINT ON BOTTOM OF SHEARHEAD 9 ON EAST SIDE

IS CLEAR. NUT DOES NOT APPEAR TO HAVE SLIPPED OFF. NUT ON WEST

SIDE APPEARS TO HAVE SLIPPED OFF IN DIRECTION OF WEB. SHEARHEAD

FROM LEVEL 11 BIT INTO EDGE OF FLANGE DURING RACKING . FOUR FT

SECTION OF JACK ROD RUNS FROM TOP OF COLUMN DOWN TO SEAL BLOCK AND

IS COMPLETE WITH LIFTING NUT AND SLEEVE. UPPER END OF ROD HAS

230
COLUMN NO

SECTION

SHEARHEAD

LENGTH

BEEN FLAME -CUT. TOP OF COLUMN SHOWS INDENTATION FROM JACK

ROTATING ABOUT WEAK AXIS OF COLUMN. MARKS INDICATE JACK ROLLED

TO EAST AS IT LEFT TOP OF COLUMN. THERE ARE NO VISIBLE INDENTATIONS

IN THE WEB DUE TO LIFTING NUT IMPACT. - C6 - W 8X35

TYPE - J50

- COLUMN TOP NOT IDENTIFIED

COLUMN NO

SECTION

SHEARHEAD

LENGTH

MARK

TACK WELDS

SHEARHEADS

COMMENTS

:

- C.5 ' r

• - W 12X65

TYPE - L50 - 5 FT - V

- 1 SET ABOVE WELD BLOCK

- NONE

JACK FOOTPRINT IS CLEARLY VISIBLE IN TOP OF COLUMN. THERE ARE NO

VISIBLE INDENTATIONS IN THE WEB DUE TO LIFTING NUT IMPACT.

COLUMN NO - El

SECTION - W 10X60

SHEARHEAD TYPE - B50

LENGTH - 23 FT

MARK - IE

TACK WELDS - 1 SET ABOVE TOP WELD BLOCK

SHEARHEADS - 5 IN PLACE (9-R)

COMMENTS: GOUGE MARK LOCATED ON WEST FLANGE BEGINS AT TOP OF UPPER SEAL BLOCK

AND APPEARS TO HAVE BEEN CAUSED BY SHEARHEAD 10 DURING LIFTING.

COLUMN HAS GRADUAL BEND TO THE NORTH ABOUT WEAK AXIS APPROX 11 FT

BELOW COLUMN TOP. WEST FLANGE HAS LOCAL DAMAGE THAT APPEARS TO HAVE

BEEN CAUSED BY CANTILEVER ACTION OF SLABS 10 AND 11. SOUTH JACK

ROD MADE CHAMFER MARK AT TOP OF THE WEB AND THE LIFTING NUT LEFT

INDENTATIONS AT THE TIPS OF THE ROD SLOT IN SHEARHEAD 9. THIS

SAME LIFTING NUT LEFT IMPACT MARK ON SOUTH SIDE OF WEB. NORTH

LIFTING NUT INDENTED THE UNDERSIDE OF THE SHEARHEAD BUT LEFT NO

MARKS INDICATING KICKOUT OF NUT. FOOTPRINT OF JACK VISIBLE ON

TOP OF COLUMN INDICATES BOTH JACK RODS WERE HEAVILY LOADED.

COLUMN BUCKLED TO EAST APPROX 4 FT BELOW THE STAGE III/IV SPLICE.

COLUMN NO

SECTION

SHEARHEAD TYPE

LENGTH

MARK

TACK WELDS

SHEARHEADS

COMMENTS

:

- E2 - W 12X106

- P50 (ROOF - AA51)

- 14' -1" - U

- 1 SET ABOVE TOP WELD BLOCK

- NONE

THIS SECTION IS VERY STRAIGHT WITH NO GOUGE MARKS ON FLANGES AND

THERE IS NO CLEAR SIGN OF A JACK FOOTPRINT ON THE COLUMN TOP.

IMPACT OF THE LIFTING NUTS CAUSED INDENTATIONS ON EACH SIDE OF

THE WEB. THE UPPER SEAL BLOCKS WERE DAMAGED BY THE SHEARHEADS AS

THEY SLID DOWN THE COLUMN. THE NEXT LOWER COLUMN SEGMENT WAS

IDENTIFIED BY MATCHING THE CUTOFF SURFACES. THIS SEGMENT IS 12'-

231
10" LONG ( 5 ' - 7

FROM STAGE IV AND 7 '-3" FROM STAGE III) AND

CONTAINS FOUR SHEARHEADS IDENTIFIED AS LEVELS 10,11,12, R.

COLUMN NO - E3

SECTION - W 12X120

SHEARHEAD TYPE - X51

LENGTH - 11' -3"

MARK - 3E

TACK WELDS - 1 SET ABOVE TOP WELD BLOCK

SHEARHEADS - NONE

COMMENTS: THIS SECTION IS BENT SLIGHTLY TO THE SOUTH ABOUT THE WEAK AXIS.

NO UPWARD SCRAPING MARKS OR GOUGES ARE VISIBLE. ANOTHER SECTION

OF THIS COLUMN IS LOCATED ADJACENT TO THE TOP SECTION. IT ALSO

IS BENT ABOUT THE WEAK AXIS AND CONTAINS 14 SHEARHEADS. THE 14

SHEARHEADS ARE STACKED AT LEVEL C AND INCLUDE LEVEL 12. THE ‘ UNDERSIDE OF SHEARHEAD 9 HAS A DEEP TRACK ON THE SOUTH LIFTING

ANGLE WHERE THE LIFTING NUT SLID OFF AND THE SOUTH SIDE OF THE

WEB CONTAINS AN IMPACT MARK FROM THE LIFTING NUT. THE LIFTING

ANGLE ON THE NORTH SIDE EXHIBITS A DOUBLE SET OF LIFTING NUT

INDENTATIONS AND THE WEB EXHIBITS 3 NUT IMPACT MARKS AT 50 TO 53

INCHES FROM THE TOP OF THE COLUMN.

COLUMN NO

SECTION

SHEARHEAD

LENGTH

COMMENTS

:

- E3.8

- W 12X120

TYPE - X51

- COLUMN TOP NOT IDENTIFIED

A COLUMN SEGMENT FROM STAGE II OF COLUMN E3 .

8

WAS LOCATED. ITS

LENGTH IS 9' -3" AND THIS SECTION CONTAINS 11 SHEARHEADS (LEVELS

1-11). THE LOWER THREE LEVELS WERE FULLY WELDED. NUMBER 9

SHEARHEAD CONTAINS MARKS FROM THE LIFTING NUTS AND EVIDENCE THAT

BOTH NUTS KICKED OUT FROM UNDER THE LIFTING ANGLES . IT APPEARS

THAT ONE NUT GLANCED OFF THE WEB AND CAUGHT BOTH THE NUMBER 10

AND NUMBER 11 SHEARHEADS AS THEY WERE MOVING DOWNWARD.

COLUMN NO

SECTION

SHEARHEAD

LENGTH

MARK

TACK WELDS

SHEARHEADS

COMMENTS

:

- E4.8

- W 12X106

TYPE - P50 (ROOF - R51)

- 14' -10" - Y

- 1 SET ABOVE TOP WELD BLOCK

- NONE

INDENTATIONS IN THE TOP OF THIS COLUMN SHOW SIGNS OF THE JACK

ROTATING ABOUT THE WEAK AXIS OF THE COLUMN. ONE EDGE OF THE

FLANGE NEAR UPPER END OF THE COLUMN HAS BEEN INDENTED BY THE JACK

ROD THREADS. ON THIS SAME SIDE OF THE COLUMN THE WEB HAS BEEN

CHAMFERED BY THE JACK ROD AND THE WEB CONTAINS A LIFTING NUT

IMPACT MARK. A SIMILAR MARK, BUT LESS PRONOUNCED, IS LOCATED ON

THE OTHER SIDE OF THE WEB. A SEGMENT OF STAGE III OF THIS COLUMN

WAS IDENTIFIED BY ITS COLUMN MARKINGS, ITS SECTION SIZE AND ITS

COMBINATION OF MARK P50 AND MARK R51 SHEARHEADS. THE SEGMENT IS

12' -18" LONG AND CONTAINS SIX SHEARHEADS IDENTIFIED WITH LEVELS

8-R.

232
COLUMN NO - E . 5 SECTION - W 12X72

SHEARHEAD TYPE - U51

LENGTH - 6 ' -7"

MARK - T

TACK WELDS - 2 LEVELS AT UPPER WELD BLOCKS

SHEARHEADS - 1

COMMENTS

:

THE WEST FLANGE SHOWS SIGNS OF LOCAL BENDING DUE TO CANTILEVER ACTION

OF FLOOR SLABS 9 TO R.

COLUMN NO - G1

SECTION - W 10X60

SHEARHEAD TYPE - B50

LENGTH - 11 ' -4"

MARK - 5-7

TACK WELDS - 2 LEVELS AT UPPER WELD BLOCKS

SHEARHEADS - NONE

COMMENTS

:

THIS COLUMN SECTION IS VERY STRAIGHT WITH NO SIGNS OF SCRAPING OR

GOUGING. COLUMN BUCKLED DIRECTLY ABOVE LEVEL 1. A LOWER SECTION

OF THIS COLUMN IS A W 10X60 AND CONTAINS 10 B50 SHEARHEADS.

COLUMN NO - G2

SECTION - W 12X106

SHEARHEAD TYPE - P50 (ROOF - AA51)

LENGTH - 19 ' -9"

MARK - x TACK WELDS - 2 LEVELS AT UPPER WELD BLOCKS

SHEARHEADS - NONE

COMMENTS: COLUMN SECTION IS STRAIGHT AND CLEAN WITH NO SIGNIFICANT SCPAPES

OR GOUGES. SHEARHEADS APPEAR TO HAVE SLID STRAIGHT DOWN WITHOUT

DAMAGING THE UPPER LEVEL WELD BLOCKS OR THE WEDGE CONTACT SURFACES

.

COL NO - G3 OR G4

SECTION - W 12X136

SHEARHEAD TYPE - Y51

LENGTH I VO 1 H-4 O

3 MARK - 1-1 P

TACK WELDS - 2 LEVELS AT UPPER WELD BLOCKS

SHEARHEADS - NONE

COMMENTS : LITTLE OR NO DAMAGE TO FLANGES NEAR TOP. NOTCH OBSERVED IN THE TOP

OF ONE UPPER WELD BLOCK.

COLUMN NO - G3 OR G4

SECTION - W 12X136

SHEARHEAD TYPE - Y51

LENGTH - 10 FT

MARK - X

TACK WELDS = 2 LEVELS AT UPPER WELD BLOCKS

SHEARHEADS - NONE

COMMENTS

:

INDENTATION (SCRAPING) IN ONE FLANGE JUST ABOVE TOP WELD BLOCK.

233
COLUMN NO

SECTION

- G5 - W 12X106

SHEARHEAD TYPE - P50 (ROOF - AA51)

LENGTH

MARK

TACK WELDS

SHEARHEADS

COMMENTS

:

- 18 FT

- XX

- 2 LEVELS AT UPPER WELD BLOCKS

- NONE

SECTION IS PRACTICALLY STRAIGHT WITH NO SIGNIFICANT SCRAPES OR

MARKS. NO VISIBLE DAMAGE TO UPPER WELD BLOCKS OR WEDGE CONTACT

SURFACES. A LOWER SEGMENT OF THIS COLUMN CONTAINS SHEARHEADS

FROM LEVELS 2 TO R.

COLUMN NO

SECTION

- G6 - HP 12X53

SHEARHEAD TYPE - C50

LENGTH

MARK

TACK WELDS

SHEARHEADS

COMMENTS

:

- 15 ' -10” - ww

- 2 LEVELS AT UPPER WELD BLOCKS

- 4 IN PLACE (9-12)

THE FOUR SHEARHEADS ON THIS COLUMN DID NOT GET PAST THE WEDGES UNDER

SHEARHEAD 9. ROOF SHEARHEAD APPEARS TO HAVE SLID OFF TOP OF

COLUMN. JACK RODS ARE STILL IN PLACE. FLANGES HAVE LOCAL DAMAGE

AT SHEARHEAD POSITIONS AND COLUMN SEGMENT HAS BEEN TWISTED

CLOCKWISE APPROXIMATELY 45 DEGREES.

COLUMN NO

SECTION

- HI

- HP 10X42

SHEARHEAD TYPE - A50

LENGTH - 7 ' -10"

MARK

TACK WELDS

SHEARHEADS

COMMENTS

:

- K

- 2 LEVELS AT UPPER WELD BLOCKS

- 4 IN PLACE (9-12)

COLUMN BUCKLED DIRECTLY ABOVE GND LEVEL. THERE ARE BITE MARKS IN

THE FLANGE EDGES CAUSED BY RACKING OF THE SHEARHEADS. ALSO, THE

FLANGES IN THE VICINITY OF THE SHEARHEADS HAVE UNDERGONE LOCAL

BENDING WHICH INDICATES CANTILEVER ACTION BY THE FLOOR SLABS.

COLUMN NO

SECTION

- H2

- W 12X65

SHEARHEAD TYPE - D50

LENGTH

MARK

TACK WELDS

SHEARHEADS

COMMENTS

:

- 13 FT

- G - PROBABLY 2 LEVELS AT UPPER WELD BLOCKS

- 4 IN PLACE (9-12)

TOP SHEARHEAD APPEARS TO HAVE SLID OFF TOP OF COLUMN. THERE IS

SIGNIFICANT LOCAL BENDING OF THE FLANGES NEAR THE TOP OF THE

COLUMN DUE TO RACKING OF SHEARHEADS 9-12. A SEGMENT OF ONE OF

THE JACK RODS RUNS DOWN THROUGH THE SHEARHEADS. THE COLUMN

BUCKLED APPROX 2 FT ABOVE THE BOTTOM END OF THIS SEGMENT.

ANOTHER SEGMENT OF THIS COLUMN FROM STAGE III CONTAINS SHEARHEADS

FROM LEVELS 3 TO 8 WITH LEVEL 3 BEING FULLY WELDED.

234
COLUMN NO - H3

SECTION - W 12X79

SHEARHEAD TYPE - N50 (ROOF - E50)

LENGTH - 30 FT

MARK - ZZ

TACK WELDS - 2 LEVELS AT UPPER WELD BLOCKS

SHEARHEADS - NONE

COMMENTS: THIS COLUMN SEGMENT HAS A GRADUAL AND UNIFORM BEND ABOUT WEAK

AXIS. SHEARHEADS APPARENTLY SLID DOWN COLUMN WITHOUT DOING

SIGNIFICANT DAMAGE.

COLUMN NO - H4

SECTION - W 12X79

SHEARHEAD TYPE - N50 (ROOF - E50)

LENGTH - 19' -9" (FRACTURE AT STAGE III/IV SPLICE)

MARK - Z

TACK WELDS - 2 LEVELS AT UPPER WELD BLOCKS

SHEARHEADS - NONE

COMMENTS: THIS COLUMN SEGMENT IS RELATIVELY FREE OF SCRAPES AND GOUGES AND

THE WELD BLOCKS AND WEDGE CONTACT SURFACES EXHIBIT VERY LITTLE

DAMAGE. STAGE III OF THIS COLUMN CONTAINS SHEARHEADS FROM LEVELS

2 TO 12. SHEARHEAD 12 SHOWS MINIMAL DAMAGE AT THE POINT OF WEDGE

CONTACT WHILE NUMBER 9 SHOWS SOME DISTORTION. THIS COLUMN SEGMENT

BUCKLED JUST ABOVE THE STAGE II/III SPLICE.

COLUMN NO

SECTION

SHEARHEAD TYPE

LENGTH

H5

W 12X65

D50

12' -7"

MARK

TACK WELDS

SHEARHEADS

COMMENTS

:

- YY

- 2 LEVELS AT UPPER WELD BLOCKS

- NONE

COLUMN H5 IS STRAIGHT AND SHOWS VERY LITTLE IN THE WAY OF MARKS OR

SCRAPES. THERE IS A SLIGHT ROUNDING OF THE EDGES OF THE WEDGE

CONTACT SURFACES AT THE UPPER WELD BLOCKS.

COLUMN NO

SECTION

SHEARHEAD TYPE

LENGTH

H6

HP 10X42

A50

8

' - 8

" MARK

TACK WELDS

SHEARHEADS

COMMENTS

:

- 5-2 D - 2 LEVELS AT UPPER WELD BLOCKS

- NONE

THIS SECTION HAS A GRADUAL BEND TO THE WEST JUST BELOW THE UPPER

SEAL BLOCKS. THERE IS SOME LOCAL BENDING OF THE WEST FLANGE THAT

APPEARS TO HAVE BEEN CAUSED BY THE SHEARHEADS. THE LOWER SECTION

OF THE STAGE IV COLUMN IS IDENTIFIED BY THE SHOP NUMBER. ITS

LENGTH ABOVE THE STAGE III/IV SPLICE IS 11' -1" AND 4' -6” BELOW

THE SPLICE. THE SECTION IS BENT ABOUT THE WEAK AXIS APPROX 3 FT

ABOVE THE STAGE III/IV SPLICE. NEITHER OF THESE TWO SECTIONS

CONTAINS SHEARHEADS.

235
APPENDIX B

Geotechnical Engineering Investigation

Prepared by: Schnabel Engineering

Consulting Engineers Associates

and Geologists,

West Chester, PA

236
GEOTECHNICAL ENGINEERING INVESTIGATION

L' AMBIANCE PLAZA BUILDING COLLAPSE

WASHINGTON AVENUE

BRIDGEPORT, CONNECTICUT

Outl ine Pag

1. Summary of Conclusions 1

2. Description of Site and Background Information 2

3. Field Investigation 4

4. Soil and Rock Material Properties 9

5. Foundation Analysis 13

6. Lateral Earth Pressure Analysis 16

Appendix A

Summary of Soil Laboratory and In Situ Tests

Gradation Test Curves (4)

Proctor Test Curve

Direct Shear Test Curves (2)

Pressuremeter Test Procedures

Pressuremeter Test Curve

Appendix B

Subsurface Exploration Data

Identification of Soil Samples

General Notes for Test Boring and Test Pit Logs

Test Boring Logs: B-l, B-2, B-2H, B-2HA, B-3E,

B-9D , B-10D , B-10F

TP-1H (east face) , TP-1H (south

TP-2H (west face) ,TP-2H (south

TP-2H (east face), TP-10D (south

Test Pit Logs

B-4.8E, B-6H,

face)

,

face)

,

face)

,

TP-10F (north face), TP-10F (west face)

TP-2

Test Boring and Test Pit Location Plan

Appendix C

Previous Test Borings by Others

Logs of PS-1, PS-2, PS-3, 1, 2,

Location Plan of Previous Borings

4, 5, 6, 7, 8, 9, 17

Appendix D

Bibliography

237
1 CONCLUSIONS

Based on the information contained in this report, the following

summary of conclusions is presented:

1. On-site soils and rock encountered in our field investigation

consisted of fill materials, residual soils formed by the in-place

weathering of parent bedrock and schist bedrock. The bedrock is

believed to be a metamorphic rock of the Prospect Formation.

2. Footings were not supported on rock in some cases, as up to

about 2.0 ft of probable fill and/or very compact residual soil

was encountered beneath footings.

3. Due to the relatively shallow depth of soil beneath footings,

we do not believe the bearing capacities of the footings examined

were exceeded.

4. Estimated settlements for the footings investigated varied

from negligible to 1/4 inch based on estimated column loads at the

time of failure.

5.

There may be a subgrade material beneath Footing 10F which

could compress more than the estimated settlement as discussed

herein

.

6« Generally loose, but variable density silty sand backfill was

encountered behind the north building

limiting earth pressure distributions

collapse based on at-rest and active conditions

Sheets 2 and 3, respectively.

wall. Upper and lower bound

on the north wall prior to

are indicated on

238
2 . DESCRIPTION OF SITE AND BACKGROUND INFORMATION

a . Site Description

The L'Ambiance Plaza* site is located south of Washington Avenue

and west of the entrance ramp to Route 8 in Bridgeport, Connecticut. It was to consist of a thirteen-story apartment tower with

three basement parking levels. The building tower was divided

into an east and west section, each about 62.5 x 112.0 ft (19 x 34

m) in plan view. A five-story parking garage was also being built

as a part of the complex south of the apartment building. The

structures were being built using the lift slab method.

The collapse occurred in the east and west sections of the

building tower. At the time of our field investigation, some of

the rubble had been cleared. Columns were cut off at varying

lengths. Concrete slab debris was piled on the basement slab of

the east and west sections. The basement slab remained intact.

This slab had an architectural drawing elevation of EL -27.33, and

this grade was used as a reference grade in this study. The

footings, some shear walls, and the east, west, and north basement

walls were also still in place during our study. The estimated

loads on columns at the time of the building collapse are noted on

Sheet 1.

b . Contract Drawings

The building tower loads were supported on spread footing foundations. Footings in the east and west sections were designed for a

7 tsf (670.3 kPa) allowable bearing pressure, as noted and

dimensioned on the foundation plans for the project. General

notes for the foundations in the plans specified an undisturbed

rock footing subgrade suitable for this bearing pressure. An

additional note on the plans indicated that if an unsuitable

subgrade material was encountered, the contractor should either:

a) remove the unsuitable material and replace with an approved

engineered fill, b) increase the footing size per structural

engineer's instructions, or c) lower footings to soil of suitable

bearing capacity.

The lower two levels of the basement wall forming the north side

of the tower had been backfilled prior to the collapse. This wall

is about 18 ft in height above Floor Level E, and plans indicate

it is supported on a 2 ft wide footing except at column locations

where wider footings have been used. The garage level floor slabs

were keyed into these basement walls on two levels, "C" and "D".

Walls below grade were also cast on the east and west sides of the

site but backfilling of these walls was incomplete. Project

plans indicate a foundation drain at the exterior base of the

basement walls. A layer of porous fill or bank run gravel was

required as backfill behind the walls per project plans.

239
SCHNABEL ENGINEERING ASSOCIATES

CONSULTING ENGINEERS

SOIL MECHANICS AND FOUNDATIONS

ESTIMATED COLUMN LOADS AT TIME

OF FAILURE y IN KIPS (IK = 454 kg)

SHEET NO OF

DATE ?2L?.§IL§_Z

CONTRACT NO. I

L' AMBIANCE PLAZA

BUILDING COLLAPSE

INVESTIGATION

BRIDGEPORT , CT

240
c . Original Geotechnical Engineering Study

A geotechnical engineering report was provided for this project by

Heynen Engineers, Clinton, Connecticut. Test boring information

obtained from this report is included in Appendix C, along with a Location Plan. In the report it was recommended that all footings

in the tower be founded on rock and dimensioned for a bearing

pressure of 7 tsf. Settlements were calculated based on elastic

solutions discussed in Schmertman (9) and Martin (5). Total

settlements for an 8.5 x 8.5 footing were calculated assuming the

7 tsf bearing pressure and were conservatively estimated to range

from < 0.5 to 1.5 inches (13 to 38 mm).

Recommendations by Heynen Engineers for the lateral earth

pressures on the building walls appear to be based on Rankine

active and passive lateral earth pressure coefficients, and an atrest coefficient. The coefficients listed in the report are K a = 0.3 (active), Kn = 3.5 (passive), and K0

= 0*5. (at-rest). The

lateral earth pressure recommended for use in design was a triangular earth pressure distribution with an ordinate at the

base of 45H psf (1 psf = 47.88 Pa). Recommended resistance to

lateral loading by footings was to be calculated from a reduced

passive pressure coefficient, K

p

= 1.9 and a coefficient of

sliding friction of 0.45.

d . Construction Observations by Others

Construction observations performed by Fairfield Testing

Laboratories, Stamford, Connecticut, were reported in letters

dated from July 10 , 1986 through March 12 , 1987 provided to us.

These reports indicate footing subgrades in the tower were

observed and consisted of bedrock, and/or a broken rock and earth

fill. The fill was noted as compacted with a backhoe bucket and

vibratory compaction, but there were no records of testing for inplace density. Backfill around the footings was tested on a spotcheck basis and results indicated compaction to 95 percent of

maximum dry density per Modified Proctor ASTM D-1557. From the

records, the backfill material appears to be a poorly graded sand

with silt and gravel, and maximum dry densities had been

estimated at 124.3 to 127.6 pcf.

Project plans provided to us are Sheets S101, S201, S202, S203,

S301, S302, Al, All, and A12 dated January 1, 1986, for L'Ambiance

Plaza by TPM Architects Inc.; and Texstar Construction Corporation

Lift Slab Plan, Sheet TCC-1 with latest revision dated February

14, 1986. Estimated building loads at the time of failure were

provided by NBS.

241
3. FIELD INVESTIGATION

Six test borings and two probe holes were drilled at footing

locations selected by NBS personnel. In addition, two borings

were performed within the north building wall backfill. All

borings were performed by Connecticut Test Borings, Inc. of

Seymour, Connecticut, under our inspection from May 28 to June 1,

1987 . Test pits were excavated adjacent to four footings, three of which

were adjacent to test boring locations. These are designated by

column location. A test pit was also excavated north of the

building wall, and is designated Test Pit TP-2. These test pits

were excavated by J.E. Barrett & Sons, Inc., on June 1 and 17,

1987. Test boring and test pit logs are included in Appendix B

along with the Test Boring Location Plan, Sheet 2. The following

is a summary of the field observations.

a . Geology

We understand the site was originally blanketed with surface

layers of existing fill placed during previous development; and

sand, gravel, cobble, and silt sedimentary deposits, probably from

a glacial outwash. These materials were removed during the

general excavation for the planned L'Ambiance Plaza construction.

The soils observed during our field exploration consisted of fill

materials placed during construction, residual soils, and bedrock.

Material described on our test boring and test pit logs as

"probable fill" appears to be composed of the on-site residual

soils and bedrock and is designated Stratum A. Backfill

materials, designated Stratum Al, are generally sandy soils, and

are probably a mixture of both residual and sedimentary soils.

The residual soils, designated Strata B and C herein, are soil

materials formed by the in-place physical and chemical weathering

of parent bedrock. The silty sands of Stratum B have undergone

more advanced weathering and typically will exhibit none or very

little of the relic structure from the parent bedrock. The

disintegrated rock of Stratum C is less weathered and may exhibit

certain rock-like qualities. In this report, disintegrated rock is

defined as a very compact density, undisturbed, naturally

occurring residual soil with Standard Penetration Test "N" values

in excess of 60 blows/ft.

We believe the parent bedrock, designated Stratum D herein, is the

Golden Hill schist of the Prospect Formation of Ordovician age.

This schist bedrock is generally medium to coarse-grained

containing quartz, mica in the form of muscovite and biotite,

plagioclase and garnet, in order of decreasing mineral quantity.

It is interlayered with fine to medium-grained gneiss with similar

mineral composition. The schist bedrock appears to have an East

North-East strike and a very steep dip to the north. These

directions are referenced to north on project plans. Core

242
drilling of rock was started after SPT "N" values in excess of 100

blows over 6 inches or less of penetration were recorded.

b. Groundwater

Water levels were not initially noted for borings at footing

locations because the water circulated during coring would have

caused a superficial level. Follow up readings are noted on the

boring logs, but indicate the borings were dry to cave depths of

about 4.0 to 4.4 ft (1.2 to 1.3 m) up to 16 days after completion.

In Borings B-l and B-2, water was not encountered and the holes

were dry upon completion to depths of caving at 10 to 12 ft.

An apparent water level was observed as noted in test pits

adjacent to footings 2H and 10F. The water levels at these

locations were comparable and relatively constant over a 16 day

period. In the excavation at Footing 2H a flow was observed

through fissures in the bedrock. The water levels indicated by

this data is about EL -32.2 to -32.7.

Water level readings which were obtained in the field

investigation are noted on the boring and test pit logs. The

water table should be expected to fluctuate with variations in

precipitation, surface runoff, leaking nearby utilities, pumping,

and evaporation.

c ./ Footing Subgrade Conditions

Bearing conditions for selected footings were observed in some

test pits and borings. The test pits were excavated to footing

subgrade, and deeper if possible, to expose a cross section

consisting of the side of the footing, the contact line between

the footing concrete and subgrade materials, and underlying

1 materials. It was apparent in the test pit excavations that

footings had been formed rather than poured neat. The lateral

over-excavation was typically a minimum 1.5 to 2 ft beyond the

footing at the bearing grade and sloped wider toward ground

surface. Test borings were cored through the footings to probe

underlying subgrade materials.

Footing subgrade materials observed consisted of disintegrated

• rock and bedrock, and also probable fill materials. Materials

labelled as probable fill were a mixture of silts, sands, mica,

| and rock fragments from on-site natural soils. No foreign matter,

| such as wood, construction debris, glass, paper, or similar was

(

observed in the material to indicate a man-made deposit. However,

the particle orientation in the probable fill stratum did not

indicate any relic structure of a residual soil. In addition,

construction records indicate similar materials were placed and

compacted beneath footings.

IK

243
Footing 1H

The entire east face of this footing, and about 3 ft of the south

face, was exposed down to bearing grade. Bearing material along

the exposed footing ‘perimeter consisted of schist bedrock. This

bedrock was hard, slightly weathered and thinly bedded as observed

in the bottom of the test pit adjacent to the footing perimeter.

Contact between the footing and bedrock appeared to be continuous.

Footing 2H-G

Generally the southern half of this 31'-10" (9.7 m) long footing

was exposed by test pit excavations. Observations in the test pit

indicated relatively continuous contact between the bottom of the

footing and schist bedrock. Bedrock, as observed, was a hard,

slightly to moderately weathered schist. However, test Boring B2HA indicated 11 inches (279 mm) of disintegrated rock immediately

underlying the footing.

A joint face was exposed along the west wall of Test Pit TP-2H

(west face). The rock material at the face of the joint is

slightly weathered and rust stained. There are discontinuous

cracks in the joint face which extend generally vertically across

the face. Rust stains were not present and these cracks were

jagged and irregular. One crack extended across the bottom of the

test pit and to the edge of the footing where access prevented , further observation. Water was flowing from the cracks as noted

on the test pit log.

Footing 3E

The test boring performed through Footing 3E indicated that the

footing may be underlain by up to 2 ft (0.6 m) of soil and

disintegrated rock. One foot of a compact density, silty sand

material was encountered immediately beneath the footing. The

silty sand was underlain by 11.5 inches (292 mm) of disintegrated

rock before hard, gray schist bedrock was encountered. We believe

the compact silty sand is a fill material because the sampled

material appeared to have been mixed.

Footing 4.8E

About 5.5 inches (140 mm) of very compact density disintegrated

rock was encountered immediately beneath Footing 4.8E in our test

boring. The soil was underlain by a fresh, hard, gray schist.

Footing 6H

The SPT test taken immediately below this footing indicated a very compact disintegrated rock material with an "N" value in

excess of 100 blows per foot. However, the 6 inch (152 mm)

244
seating interval indicated a considerably less compact layer (12

blows/6 inches), and we do not believe this lower blowcount was

due entirely to disturbance from boring procedures. The 6 inch

(152 mm) layer just underlying the footing may be silty sand fill

placed beneath the footing. Bedrock was encountered at a depth of

17 inches below the footing.

Footing 9D

This footing was cored with an Nx core barrel which could not be

removed after breaking through the bottom of the footing. In the

process of trying to remove the core barrel, it was bumped with

the 140 pound (63.6 kg) donut hammer, as controlled by a rope with

several wraps around the cathead. The core barrel penetrated the

materials beneath the footing relatively easily, to a depth of

about 6 inches (152 mm), indicating the material was not bedrock.

Footing 10D

About 6 ft (2 m) of the south face of Footing 10D was exposed by a test pit to the bearing grade. Observations in this test pit

indicated a continuous contact between the footing and the

bedrock, which was observed to be moderately hard, moderately

weathered gray schist. The test boring in the footing indicated 9

inches (228 mm) of very compact disintegrated rock between the

bottom of the footing and bedrock.

Footing 10F

Variable bearing conditions were observed beneath this footing.

The test boring indicated up to 2.2 ft (670 mm) of soil beneath

the footing. Of this 2.2 ft, less than the lowest 0.5 ft (152 mm)

may have been disintegrated rock, overlying schist bedrock.

However, in the split spoon sample recovered, we could not observe

any residual rock structure and the material has been listed as a probable fill.

The test pits excavated along the north and west faces of the

footing exposed continuous contact between the footing and bedrock

along the west face, but a bedrock surface sloping to a grade up

to about 1.5 ft (457 mm) below the footing on the north face. The

subgrade materials are described in detail on the test pit log.

In part, exposed materials on the north face appeared to be

disintegrated rock or residual soil materials; however, the

materials may have been partially disturbed from a natural

orientation. There were gaps between natural planes of rock

fragments up to 1/2 inch (13 mm). The materials could also be

easily removed by hand and the excavation was extended about 6 to

8 inches (150 to 200 mm) back under the footing.

245
d . North Building Wall

Two borings, B-l and B-2, and one test pit were performed within

the backfill placed adjacent to the north building wall. The

subsurface sampling' and testing indicated a loose to compact

density silty sand backfill to a depth of 13.5 to 18.5 ft (4.1 to

5.6 m). The variable density fill was underlain by compact to

very compact residual soils to the depths of boring penetration at

16 to 20 ft (4.5 to 6.1 m)

.

Cracks in the ground surface were observed behind the wall

adjacent to the east tower. These cracks were generally parallel

to the wall and about 11 to 15 ft north of it. The cracks were up

to about 1 inch wide.

246
4. SOIL AND ROCK MATERIAL PROPERTIES

Seven jar samples, two tube samples, and one bulk sample were

tested in the soils laboratory for plasticity characteristics and

grain size distribution. The bulk sample was also remolded and

tested for shear strength parameters. Pr essuremeter and field

density tests were performed at the site during the field

investigation. Results of the laboratory and in situ testing are

shown on the Summary and Graphs of Appendix A. All samples were

tested in accordance with applicable ASTM standards. Properties

of the soil and rock materials are discussed below by stratum.

Stratum A1 ; Backfill behind North Building Wall

Stratum A1 materials placed for floor slab support were not tested

in our soils laboratory. The backfill behind the retaining wall

was tested for plasticity indices, and the tested portion of the

material was non-plastic. The tested samples contained silt to

gravel-size particles; however, the grain-size distribution

indicates a poorly graded material. About 16.9 to 25.8 percent of

the samples were finer than the No. 200 sieve. The materials

classified as silty sand with rock fragments, SM in accordance

with ASTM D-2487.

SPT "N" values, field density tests, and dry densities performed

on tube samples indicate the backfill is generally loose. Density

tests performed by sandcone methods indicated dry densities of ’ 87.4 and 109.6 pcf (1.40 and 1.76 ton/m^) with respective moisture

contents of 23.5 and 7.8 percent. In place dry densities obtained

on five samples within two undisturbed tubes were low, varying

from 84.2 to 104.0 pcf (1.35 to 1.67 ton/m^). The above densities

correspond to 67.2 to 87.5 percent of maximum dry density per ASTM

D-1557. We believe these densities are generally representative

of the backfill.

One pressuremeter test was performed on the representative loose

density fill in Boring B-2 between 6 to 8 ft (1.8 to 2.4 m)

depths. The test results were inconclusive. The material was so

loose that the correction for probe membrane stiffness masked the

resistance being measured in the soil backfill.

The backfill material appeared to be predominantly an excavated

residual soil and rock fragment mixture. Since these materials

are remolded, we believe the relic mineral bonds, generating an

apparent cohesion within a natural residual soil, have been

broken. The materials are also non-plastic, indicating no

significant attraction between particles. Therefore, it is our

opinion that this sample has no significant cohesion.

Direct shear tests were performed to determine the internal angle

of friction for the mixture. These tests indicated a high angle

of friction 0 = 37°. We believe this angle is not representative

of the actual value for the loose density materials and may, in

part, be attributed to particle interlocking associated with the

247
constant volume direct shear test. We believe the shear stresses

which can be obtained at lower strains with each test series are

less likely to be biased by the constant volume test constraint

and are probably more representative. For shear displacements of

0.2 inches and le-ss, the angle of internal friction is

accordingly, 0 f* 33°«

Based on the low "N" values and natural densities obtained in

testing, we estimate the average friction angle in the backfill to

vary between 28° and 32° based on Meyerhof's 0-N correlation (6).

Stratum A: Probable Fill Beneath Footings

Tests indicated these materials are non-plastic. There was a predominant quantity of weathered rock fragments in these samples,

ranging from sand to gravel in particle sizes. Gravel size rock

fragments comprised about 34 to 58 percent of the tested samples.

These particles were in a relatively advanced stage of weathering

and therefore friable. In our opinion, repeated handling of the

particles would have caused significant changes in grain size

distribution, but ultimately the material would have broken down

to a silty sand or sand with silt material. Therefore, we have

classified the probable fill in accordance with plasticity and the

relative silt and sand content. Thus, materials were classified

silty sandy with rock fragments, or poorly graded sand with silt

and rock fragments, SM in accordance with ASTM D-2487.

The probable fill soils are not considered to have cohesion for

the same reason discussed previously on Stratum A1 backfill. The

internal angle of friction was estimated from the standard

penetration values obtained in the test borings based on Meyerhof

0-N correlations (6). Accordingly, a range of values from about

31° to 34° is estimated for a range of "N" values from 16 to 24.

It was possible to run only one pressur erne ter test on the probable

fill encountered beneath footings. The pressur emeter test was

performed beneath Footing 10F, where about 2 ft of probable fill

material was encountered. The test indicated a Pressur emeter

Modulus (Ep) of 80 tsf and a Limit Pressure (Pl) 7.0 tsf.

Elsewhere rock or disintegrated rock was encountered at a shallower depth below footing subgrade, and the test could not be

performed

.

We believe the rheologic factor relating the pressuremeter modulus

to the Deformation Modulus for the silty sand materials is between

1.0 and 2.0. Therefore, the estimated Deformation Modulus, E s , for

the probable fill tested is between 80 and 160 tsf. The higher

value corresponds closely with an empirical formula developed by

Bowles (2) to relate "N" values to Eg for cohesionless materials.

This formula

E s = 5 [N + 15] tsf

248
has been used to estimate the Modulus of Deformation beneath the

other footings where pressur erne ter tests could not be run. As

estimated by this formula, the probable fill beneath two other

footings, 3E and 6H,. had upper bound Deformation Moduli of E s = 155 and 195 tsf, respectively. Lower bound moduli were estimated

from these values, reduced by 50 percent to correspond to a rheologic factor of 1.0. This estimated range of moduli was used

in our settlement calculations. The "N" values must be

extrapolated for the estimate because layer thickness was less

than 18 inches and included the six inch seating interval of the

SPT test, which probably also contains less compact material than

the underlying layers.

Stratum B: Residual Soil

Residual soil of Stratum B was encountered in only one boring, B2, near the depth of penetration at 13.5 ft. The SPT "N" values

indicate a compact density material which was visually classified

as a silty sand, SM in accordance with ASTM D-2487. This material

was not encountered elsewhere and was not examined further.

Stratum C: Disintegrated Rock

The very compact density residual materials of Stratum C are

characterized by SPT "N" values greater than 60 and less than » 100/6". One jar sample of materials was classified using a sieve

analysis, plasticity indices, and a rationale similar to that

applied to Stratum Al. This material was similarly non-plastic

and contained a predominant portion of weathered rock fragments,

classified as a poorly graded sand, with rock fragments (SP-SM).

The disintegrated rock material has not undergone the in-place

physical and chemical weathering to the extent that all relic

bonds are broken. Published data on strength and d e f o r mab i 1 i t

y

characteristics of similarly weathered micaceous schist and gneiss

has been related to "N" values between 60 and 100/2". The

effective cohesion (c) has been measured at 0.2 to 2.5 tsf with

the lower bound corresponding to the lower SPT values and cohesion

increasing with density as indicated by higher "N" values, Gardner

(3). The friction angle (0) similarly was found to vary between

29 and 36°. Def ormab il i ty of the disintegrated rock has also been

related to a similar range of SPT "N" values from a data base

developed in piedmont residual materials. It ranges from about

300 tsf to 1200 tsf, Martin (5). We believe these strength and

deformation parameters are reasonably representative of the onsite disintegrated rock material for purposes of this study.

Straight line interpolation has been used for estimating strength

parameters over the range of "N" values. These parameters have

been used in our investigation calculations.

249
Stratum D; Schist (Prospect Formation)

The schist bedrock cored in the test borings for this

investigation was generally slightly weathered and moderately hard

or better. Joint faces did not exhibit much more extensive

weathering and were rust stained, but exposed moderately hard

rock. Published data indicates a peak friction angle 0 between

43° and 50° for this type of rock, Gardner (3). Similarly, the

cohesion can be correlated to the friction angle by

c =

2 tan (45 + 0/2)

where q u is the compressive strength of the rock. We have used

the compressive strengths observed by Heynen Engineers to estimate

the cohesion of on-site rock. Accordingly, the range in cohesion

varies considerably from about 150 to 2100 psi.

The Deformation Modulus for intact rock will be greater than the

upper values discussed for Stratum C materials because less

weathering has occurred. For the purpose of this investigation,

such a modulus indicates that, as a matter of practical

consideration, the intact rock is not deformable within the stress

range being considered. Any deformation would occur along joint

faces with more advanced stages of weathering. However, our

investigation did not indicate the presence of highly weathered

material to significant depths below footings.

250
5. FOUNDATION ANALYSIS

The plan dimensions, estimated axial loads, and resulting bearing

pressures of those footings examined during our field

investigation are tabulated below.

Column Loads and Bearing Pressures

at Time of Failure for Tested Footings

Plan Estimated Load (1) Bearing Pressure

^

Footing Dimens ions at Failure (kips) at Failure (ksf)

1H 4' 6" x

4' 6" x 1* 9"

2H-2G 31* 10" x 10' 0" x

5' 0"

3E 10' 0" x 10' 0" x

3' 6"

4.8E 10' 0" x 10' 0" x

3' 6"

6H 5' 6" x

5' 6" x

2 ' 1

"

9D 10’ 0" x 10' 0" x

3 ' 6

"

10F 7' 6" x

7' 6" x

2 ' 8

"

10D 10' 0" x 10' 0" x

3' 6"

146

282 + 652 + 338

(combined load)

782

877

150

890

413

839

7.2

3.99

(average stress)

7.82

8.77

4.95

8.9

7.33

8.39

These values were used in our analysis as discussed below. Other

footings were not considered in this analysis.

(1) 1 kip = 4.448 kN

(2) 1 ksf = 47.88 kPa

251
a e Bearing Capacity of Footings

Based on the estimated loads at the time of failure, footing

contact pressures were about 4.0 to 8.9 ksf for the footings

investigated during* this study. Under these loads a bearing

capacity failure is considered very unlikely because of the high

shear strength of the rock which was encountered in most cases at

bottom of footing grades. Where soil was encountered beneath

footings, its relatively shallow depth and granular nature

precluded general bearing capacity failure. Similarly, punching

or local bearing capacity failure is not considered likely as the

footing/rock geometry does not lend itself to this failure mode.

We have conservatively estimated the bearing capacity for footings

observed during our field investigation using factors for local

shear, Bowles (1). In accordance with standard practice, a safety

factor of three is generally used for calculating the allowable

capacity. The lowest safety factors against local shear were

calculated for Footings 10D, 3E and 10F, and were between 1.3 and

2.0 based on contact pressures listed in the preceding table.

However, as noted, these calculated factors of safety are

considered to be unrealistically low.

b . Footing Settlement

Settlement of the footings has been estimated using elastic

solutions. We have considered the approach developed by

Schmertmann (9), as modified by Martin (5) for residual soils, to

be the most applicable. This method can account for subsoil

layers, the effect of time, and strain distribution. The

concentration of loads in a relatively narrow, compressible zone

overlying bedrock has also been considered as discussed by Martin

(5). We have assumed the bedrock is incompressible for the stress

range considered. Additional methods based on elastic theory

would have been considered in more detail to evaluate the relative

effects of various parameters, had the estimated settlements been

significant. However, our estimate for the worst case conditions

observed during our field investigation, at Footing 10F, indicates

less than 1/4 inch total settlement could have been anticipated.

We note that this is much less than the settlement estimated in

the design report calculations. However those design calculations

were based on higher loads and, more significantly, on a .lower

quality designation for the underlying bedrock than that

determined for the cores taken in our investigation. There was,

at most, a 2 ft thick depth of relatively compressible material

observed beneath the investigated footings.

The estimated settlement beneath Footing 10F was based on the 2 ft

layer of compressible materials sampled in the test boring. We

note that the test pit exposed bedrock on one side of the footing

and soil on the other. Thus, an angular distortion of between

0.001 and 0.00 3 has been estimated for this footing based on the

differential settlement across the footing.

252
However, beneath Footing 10F, the highly weathered rock observed

on the east side of the test pit had open joints and gaps along

foliation planes. In addition to the advanced stage of

weathering, these voids could have allowed for appreciable

deformation. This ‘deformation would be impossible to quantify

with any accuracy, because of the random geometry and unknown

extent. A better indication of the extent of this condition could

be obtained by removing the footing or examining the other two

sides from adjacent test pits.

Settlement estimated at the footings examined in this study range

between negligible for Footings 1H and 2H, and less than 1/4 inch

for Footing 10F. Settlement beneath Footings 4.8E, 3E, 10D and

6H were estimated to be less than 1/10 inch. We believe a major

portion of the estimated settlement for the applied loads listed

herein had occurred prior to building collapse.

253
6 LATERAL EARTH PRESSURE ANALYSIS

Estimates of the lateral earth pressure acting on the north

building wall below grade must take into account the soil behind

the wall, and the deflections which the wall had likely undergone

prior to collapse. The highly variable fill conditions behind the

wall complicate estimates of soil parameters. It is estimated on

the basis of the data obtained on the backfill that, in general,

tne backfill is in a loose state representative of a low angle of

internal friction, 0 = 28 to 30°. An average moist unit soil

weight of 110 pcf is considered to be applicable. Back

calculation of a failure wedge corresponding to the tension cracks

observed in the backfill assuming a Rankine active failure state

also indicates 0 values of about 30° and less.

We believe the cracks visible in the soil behind the north

building wall resulted from a displacement of the wall, but this

displacement could have occurred after the collapse, possibly

during rescue efforts. If the partially constructed building wall

afforded enough rigidity to prevent wall movement prior to the

collapse, then an at-rest pressure condition would have been

approached. However, we have noted from project plans that this

wall would have been supported by transmitting the loads through

the building shear walls to spread footing foundations. For this

support to have been mobilized in the foundations, some strain

must have occurred. Therefore, an estimate of the pressure

distribution would be limited by the lower bound of an active , pressure distribution and an upper bound of an at-rest pressure

d istr ibut ion . An active earth .pressure coefficient K a = 0.33 to 0.36 is

considered applicable for the wall. The estimated range of

equivalent fluid pressures acting on the wall assuming active

conditions is shown on Sheet 2.

A similar equivalent fluid pressure distribution for the at-rest

condition has been estimated using Jaky's approximation

(Winterkorn and Fang, 11) of the at-rest coefficient estimated as

K a

= 0.5 to 0.53. The estimated range of equivalent fluid

pressures acting on the wall assuming at-rest condtions is shown

on Sheet 3.

254
SCHNABEL ENGINEERING ASSOCIATES

CONSULTING ENGINEERS

SOIL MECHANICS AND FOUNDATIONS

SHEET NO. OF

DATE 8~2J5~8_7 „ CONTRACT NO. CT870587

ESTIMATED LATERAL' EARTH PRESSURES

ASSUMING ACTIVE CONDITIONS

L' AMBIANCE PLAZA

BUILDING COLLAPSE

INVESTIGATION

BRIDGEPORT , CT

H (ft.)

SURCHARGE

HORIZONTAL PRESSURE

FROM SURCHARGE

(0.33 to 0.36 x VERTICAL

SURCHARGE)

255
SCHNABEL ENGINEERING ASSOCIATES

CONSULTING ENGINEERS

SOIL MECHANICS AND FOUNDATIONS

SHEET NO._JL .OF

date. 8-25-87 . CONTRACT NO-5lT-?.7_9A?_Z

L* AMBIANCE PLAZA

BUILDING COLLAPSE

ESTIMATED LATERAL EARTH PRESSURES INVESTIGATION

ASSUMING AT-REST CONDITIONS BRIDGEPORT , CT

H (ft.)

SURCHARGE

HORIZONTAL PRESSURE

FROM SURCHARGE

(0.50 to 0.53 x VERTICAL

SURCHARGE)

256
Appendix A

SOIL LABORATORY AND IN SITU TESTS

257
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7 8 9 10 11 12

MOISTURE CONTENT, %

SPECIFIC

GRAVITY 2.65

specification: astm D-1557

METHOO:

DESCRIPTION OF SAMPLE

SILTY SAND WITH WEATHERED POCK FRAGMENTS,

BRCWN

SCHNABEL ENGINEERING

ASSOCIATES

classification: sm

MOISTURE DENSITY

RELATION

SAMPLE NO.: TP-2 / 5.0* LIQUID limit: np

,L* AMBIANCE PLAZA

BRIDGEPORT, CONNECTICUT source: on-site plasticity index: np

1 % PASSING 3/4" SIEVE MAX DRY DENSITY 125.3 PC

F

DATE

8-25-87

CONTRACT NO. CT870587 % PASSING NO. 4 SIEVE OPT. MOISTURE 9.5 %

265
SHEET OF __

l.Q

NORMAL STRESS, KSF

Lo

DESCRIPTION OF SOIL SAMPLE TESTED

KEY BORING DEPTH NORMAL MOIST. CONT. % DENSITY, PCF

NO. FT. STRESS INITIAL FINAL DRY WET

o TP-Z <3-0 0-5

t'S.S

G TP'L S.Q \.o rs-s 1^7 m. s

A 19-1 S\o US V&.7 »*-7 SVl '» 0

*7

SCHNABEL ENGINEERING ASSOCIATES

DIRECT SHEAR TEST

TYPE OF TEST: Cov*Sov.\Cv*?Tt& <-**» RATE OF SHEAR: q.or^.Q.;^ wv*< s /V*

PROJECT:

<

Byr »-« DATE: 3ZS-37 CONTR. NO. 266

PRESSUREMETER TESTING

SCHNABEL

ENGINEERING

ASSOCIATES

Introduction

The pressuremeter developed by Louis Menard of France has been in use by

Schnabel Engineering Associates since 1967. We were first to use this instrument in the eastern United States and offer pressuremeter testing services.

The pressuremeter has developed into one of the more successful in-situ testing

methods in the past several years and is now widely used. The following short

description of the instrument and the outline of its use and its limitations is

intended to familiarize our clients with the advantages of field testing by the

pressuremeter.

Brief description of the pressuremeter test : The test is performed in a borehole by a cylindrical metal probe covered with rubber membranes. The probe is

inflated by water under pressure from a surface control apparatus. (See Fig.-l.)

Pressure is increased in steps and deformations are recorded and thus the procedure represents a load test on the walls of

the borehole. Volume changes for one particular loading step are recorded at 15 seconds,

30 seconds and one minute after load application. The probe may be lowered and tests be

performed up to 100 ft depth. Tests are generally made in test borings and some special

equipment and techniques are required to prepare a boring for pressuremeter tests.

Areas of application

The pressuremeter test may be considered most

useful in residual soils, granular soils with

some cohesion, very stiff to hard clays and

soft rock. Methods are also available for

conducting the test in granular soils below

groundwater level.

Results of pressuremeter tests

The test furnishes information as to the un- drained strength and deformation characteristics of the material. Results provide a basis

to predict bearing capacity and settlement of

foundations, slope stability and other soil

mechanics problems.

Figure 1

The basic result of the test is the pressuremeter curve which indicates volume increase of

the probe versus the pressure applied considering readings at the end of each loading step.

This curve also represents the deformation of

the soil under lateral radial stresses. (See

Fig. 2.) The initial portion represents the

adjustments of the probe to the bore hole and

further the restoration of the original horizontal pressures. Then a straight

line portion of the curve follows which is the elastic deformation of the subsoil

and can be measured by the slope of the line, resulting in the pressuremeter

modulus Ep. This modulus is evaluated for each test and is shown in the units of

tons per square foot (kg/cm^). The pressuremeter modulus is similar to the

modulus of elasticity except it is measured in the horizontal direction. Corrections for anisotropy are necessary in most soils to obtain elasticity in the
800

700

600

500

400(

300

200

100

SCHNABEL ENGINEERING ASSOCIATES

CONSULTING ENGINEERS

SOIL MECHANICS AND FOUNDATIONS

SHEET OF

date. ,_8_T_25_"8_7____

„K1TDArTM . CT870587 CONTRACT NO .

PRESSUREMETER TEST RESULTS

L* AMBIANCE PLAZA

BUILDING COLLAPSE

INVESTIGATION

BRIDGEPORT , CT

BORING NUMBER: B-10F

PROBE DEPTH: 3.9 to 5.9 FT

ESTIMATED LIMIT PRESSURE: 7 tsf

PRESSUREMETER MODULUS: 80 tsf

LEGEND

O TEST VOLUME

& CREEP VOLUME

2.0 4.0 6.0

PRESSURE

(kg/cm2

)

8.0 10.0

269
Appendix B

SUBSURFACE EXPLORATION DATA

270
Appendix B

SUBSURFACE EXPLORATION DATA

General Notes for Test Boring Logs

Identification of Soil Samples

Test Boring Logs: B-l, B-2, B-2H, B-2HA, B-3E, B-4.8E, F-6H,

B-9D, B-l OD f B-10F

Test Pit Logs: TP-1H (east face) , TP-1H (south face) , TP-2H

(west face) ,TP-2H (south face),TP-2H (east face),

TP-10D (south face) , TP-10F (north face) , TP-10F

(west face) , TP-2

Test Boring and Test Pit Location Plan, Sheet 2

Test 3orings

Test borings at footing locations were advanced through the

footings to allow sampling and/or testing of actual subgrade

materials. Borings were advanced by coring into the floor slab,

footing, and intermediate backfill for floor slab support with an

HW core barrel. The core barrel was advanced to within 2 inches

of plan footing bottom and through all reinforcing steel

encountered. A three inch tri-cone roller bit was used to break

apart the concrete and flush the concrete cuttings until the

bottom of the footing was penetrated. Reinforcing steel was

removed with a magnet before breaking through the bottom of the

footing. The roller bit was immediately removed and Standard

Penetration Testing was used to sample and test materials to the

rock surface.

To sample the underlying bedrock once encountered, the HW core

barrel was advanced to the depth of split spoon penetration. The

tri-cone roller bit and circulated water was used to cut to this

depth. At Footing 10-F the roller bit was advanced about 0.2 ft

further to seat in the rock surface. Rock was cored using double

tube NX core barrels, except at Footing 10-D where a single tube

NX core barrel was used because it was the only core barrel

available at the time the test coring was performed. Percentages

of recovery and RQD are indicated on the boring logs.

Test borings north of the building wall were drilled using hollowstem augers. The augers were advanced with a plug inserted.

Continuous sampling using the driven split spoon was performed

following removal of the plug. The sampling was performed in 2 ft

intervals as indicated on the Test Boring Logs. Augers were

advanced following two successive samplings.

271
The driven split spoon was used to obtain the Standard Penetration

Test resistance "N" values in soil materials. "N" values were

determined as the number of blows required to drive a 2 inch O.D.,

1-3/8 inch I.D. sampling spoon one foot using an automatic SPT

hammer system. The' driving force- is provided by a 140 pound

hammer falling 30 inches. In accordance with ASTM D-1556, the "N”

value is taken after seating the sampler 6 inches in the bottom of

the hole. The driving force provided by the automatic hammer has

been studied by Riggs, Mathes and Rassieur (8), and results

indicate "N" values in the lower range of those produced by using

the standard cathead and rope method, with two wraps.

Test Pits

Most test pits were excavated with a crawler mounted, hydraulic

backhoe. Some less accessible areas of the test pits and TP-2H

(south face) were excavated with a rubber-tire backhoe/loader

Test 3oring/Pit Location and Elevation Survey

Test borings and test pits were located by taping from existing

building features as shown on Sheet 2. These locations should be

considered accurate to 1 ft + . Elevations at footing locations

were measured from the top of* the footing and were referenced to

plan top of footing elevations. Test pit and boring elevations

north of the building wall were referenced to plan top of wall

elevations. The elevations should be considered accurate to the

nearest 0.5 ft + from the referenced feature.

272
GENERAL NOTES FOR TEST BORING LOGS

1. NUMBERS IN "SAMPLE SPOON" COLUMN INDICATE BLOWS REQUIRED TO DRIVE A 2 INCH

O.D., 1-3/3 INCH I.D. SAMPLING SPOON 6 INCHES USING A 140 POUND HAMMER FALLING

30 INCHES ACCORDING TO ASTM D-1586.

2. VISUAL CLASSIFICATION OF SOIL IS IN ACCORDANCE WITH TERMINOLOGY SET FORTH IN

"IDENTIFICATION OF SOIL." THE UNIFIED SOIL CLASSIFICATION SYMBOLS SHOWN IN

PARENTHESES ARE BASED ON VISUAL INSPECTION.

3. ESTIMATED GROUNDWATER LEVELS INDICATED BY THESE LEVELS ARE ONLY ESTIMATES

FROM AVAILABLE DATA AND MAY VARY WITH PRECIPITATION, POROSITY OF THE SOIL, SITE

TOPOGRAPHY, ETC.

4. REFUSAL AT THE SURFACE OF ROCK, BOULDER, OR OBSTRUCTION IS DEFINED AS A

PENETRATION RESISTANCE OF 100 BLOWS FOR 2 INCHES PENETRATION OR LESS.

5. THE BORING LOGS AND RELATED INFORMATION DEPICT SUBSURFACE CONDITIONS ONLY AT THE

SPECIFIC LOCATIONS AND AT THE PARTICULAR TIME WHEN DRILLED. SOIL CONDITIONS AT

OTHER LOCATIONS MAY DIFFER FROM CONDITIONS OCCURRING AT THESE BORING LOCATIONS.

ALSO, THE PASSAGE OF TIME MAY RESULT IN A CHANGE IN THE SUBSURFACE SOIL AND

GROUNDWATER CONDITIONS AT THESE BORING LOCATIONS.

6. THE STRATIFICATION LINES REPRESENT THE APPROXIMATE BOUNDARY BETWEEN SOIL AND

ROCK TYPES AS DETERMINED FROM THE DRILLING AND SAMPLING OPERATION. SOME VARIATION

MAY ALSO BE EXPECTED VERTICALLY BETWEEN SAMPLES TAKEN. THE SOIL PROFILE, WATER

LEVEL OBSERVATIONS AND PENETRATION RESISTANCES PRESENTED ON THESE BORING LOGS

HAVE BEEN MADE WITH REASONABLE CARE AND ACCURACY AND MUST BE CONSIDERED ONLY AN

APPROXIMATE REPRESENTATION OF SUBSURFACE CONDITIONS TO BE ENCOUNTERED AT THE

PARTICULAR LOCATION.

7. BORING LOG VERTICAL SCALE: 1/6 INCH = 1 FT.

8. TEST BORINGS DRILLED BY CONNECTICUT TEST BORINGS, INC., SEYMOUR, CONNECTICUT

UNDER INSPECTION OF SCHNABEL ENGINEERING ASSOCIATES.

9. KEY TO SYMBOLS AND ABBREVIATIONS:

STANDARD PENETRATION TEST

2" or 3" UNDISTURBED TUBE SAMPLE

(RECOVERY SHOWN IN REMARKS COLUMN)

PRESSUREMETER TEST

VANE SHEAR TEST

STATIC CONE PENETRATION TEST

NX OR 2 INCH O.D. ROCK CORE RUN

(RECOVERY SHOWN IN REMARKS COLUMN)

*, NO SAMPLE RECOVERY

do, DITTO

RQD, ROCK QUALITY DESIGNATION

v, NATURAL MOISTURE CONTENT

273
SCHNABEL ENGINEERING ASSOCIATES

Consulting Geotechnical Engineers

IDENTIFICATION OF SOIL

I. DEFINITION OF SOIL CROUP NAMES ASTM D-2487-83 Symbol Group Name

Coarse-Grained Soils More than 50% retained on No. 200 sieve

Gravels —

More than 50% of coarse fraction retained on No. 4 sieve Coarse, 3/4' to 3'

Fine, No. 4 to 3/4'

Clean Gravels

Less than 5% fines GW Weil graded gravel GP Poorly graded gravel Gravels with Fines More than 12% fines CM Silty gravel CC Gayey gravel Sands — 50% or more of coarse

fraction passes No. 4 sieve Coarse, No. 10 to No. 4 Medium, No. 40 to No 10

Fine, No. 200 to No. 40

Clean Sands

Less than 5% fines SW Well-graded sand SP Poorly graded sand Sands with Fines More than 12% fines SM Silty sand SC Clayey sand

Fine-Grained Soils 50% or more passes

the No. 200 sieve

Silts and Clays —

Liquid Limit less than

50Low to medium plasticity

Inorganic CL Lean clav ML Silt Organic

OL

Organic day

Organic silt

Silts and Clays —

Liquid Limit 50 or more Medium to high plasticity

Inorganic CH Fat clav MH Elastic silt Organic OH

Organic day

Organic silt Highly Organic Soils Primarily organic matter, dark in color, and organic odor PT Peat

II. DEFINITION OF MINOR COMPONENT PROPORTIONS

Minor Component

Adjective Form

Gravelly, Sandy

With

Sand, Gravel

Silt. Clay

Trace Sand, Gravel

Silt. Clay

Approximate Percentage of Fraction by Weight

30% or more coarse grained

15% or more coarse grained

5% to 12% fine grained

Less than 15% coarse grained

Less than 5% fine grained

III. GLOSSARY OF MISCELLANEOUS TERMS

SYMBOLS — Unified Soil Classification Symbols are shown above as group symbols. Use A Line Chart for laboratory identification. Dual svmbols

are used for borderline classifications. BOULDERS & COBBLES — Boulders are considered rounded pieces of rock larger than 12 inches, while cobbles range from 3 to 12 inch size. DISINTEGRATED ROCK — Residual rock material with a standard penetration resistance (SPT) of more than 60 blows per foot, and less than

refusal. Refusal is defined as a SPT of 100 blows for 2

" or less penetration.

ROCK FRAGMENTS — Angular pieces of rock, distinguished from transported gravel, which have separated from original vein or strata and are present in a soil matrix. QUARTZ — A hard silica mineral often found in residual soils IRONITE — Iron oxide deposited within a soil layer forming cemented deposits

CEMENTED SAND — Usually localized rock-like deposits within a soil stratum composed of sand grains cemented by calcium carbonate or other

materials. MICA— A soft plate of silica mineral found in many rocks, and in residual or transported soil derived therefrom. ORGANIC MATERIALS (Excluding Peat): Topsoil - Surface soils that support plant life and which contain considerable amounts of organic matter: Organic Matter - Soil containing organic colloids throughout its structure;

Lignite - Hard, brittle decomposed organic matter with low fixed carbon content (a low grade of coal). FILL — Man made deposit containing soil, rock and often foreign matter. PROBABLE FILL — Soils which contain no visually detected foreign matter but which are suspect with regard to origin LENSES — 0 to 1/2 inch seam of minor soil component.

LAYERS — 1/2 to 12 inch seam of minor soil component.

POCKET — Discontinuous body of minor soil component

COLOR SHADES — Light to dark to indicate substantial difference in color. MOISTURE CONDITIONS — Wet, moist, or dry to indicate visual appearance of specimen.
SCHNABEL ENGINEERING ASSOCIATES

CONSULTING ENGINEERS 1 TEST BORING LOG BORING NO! b-i

PROJECTS. 1 AMBIANCE PLAZA r ’RTTTTDTNT4 (TIT ,T APSE INVESTIGATION SHEET NO : 1 OF 1

CLIENT! national bureau of standards job no: CT87Q587 J BORING_ CONTRACTOR roKrMPrmrrTrr it

tp<?t pdpint:^

.

i\rQRILL'. rvESS

WATER LEVEL DATA

elevation: -9.3+

ORIVE SAMPLER CASING size:

DATE TIME DEPTH CAVED TYPE as. DATE START: 5-30-87

ENCOUNTERED 5-3 Q. 2EC DIA. g'O.D. DATE FINISHED: 5-30-87

AFTER CASING PULLED 5-3 Q DEL 121 WT. 140* driller: J. DeANGELISS

HR. READING 3a inspector:

23

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UJ

-JUJ

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Isis

-J2

2>- IDENTIFICATION REMARKS

o

-Q. 7 ±

rn ^ CO 4-3- SILTY SAND FILL WITH CONCRETE FRAGMENTS IN NOS E 9-1 s OF SPOON FROM 0 TO 2 FT, MOIST - BROWN (SM)

1-1-

1-1 s WOR-1-

1-2 $

A

1

6-9 - 10-16 s DO, WITH ROCK FRAGMENTS AND BRICK BACKFILL

ir 1O' i

_=2D_ 3-7 s DO, WITH CINDERS

7—4 - o-in s 4-6-

i n— 1

1

3 DO, WITH ROCK FRAGMENTS IN NOSE OF SPOON

4-7-

2-4 ,8 DO, WITH MICA - GRAY

3-2-

4-3 S

1R.5 5-7- -'DISINTEGRATED ROCK WITH QUARTZ AND MICA,N C 20.

0

27-82 s MOIST - GRAY RESIDUAL

BORING TERMINATED AT 2 O'

275
SCHNABEL ENGINEERING ASSOCIATES

CONSULTING ENGINEERS TEST BORING LOG BORING NO: b-2 ]

PROJECT!/ AMBIANCE PLAZA, BUILDING COLLAPSE INVESTIGATION SHEET NO : 1 OF 1

client: national bureau of standards JOB NO! CT870587

BORING CONTRACTOR OONNFCTTCT1T TEST BORINGS, INCORILL*. CME55 elevation: -io.3±

WATE R LEVEL.DATA DRIVE SAMPLER CASING SIZE:

DATE TIME DEPTH CAVED TYPE as. DATE START: 6-1-877

ENCOUNTERED 6-1 DRY — DIA. 2"0.D. DATE FINISHED: 6-1-87

AFTER CASING PULLED 6-1 DRY O

*

—

WT. 140# DRILLER: J. DeANGELISS

.HR. READING RAOEETr.r.m ttpon oompUFTIQN FALL 30

" inspector: c . drvault

2-

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_lUJ IDENTIFICATION REMARKS

5-3-

3=1. 6-4 -

4-4

A 1 2-2

2-1

SILTY SAND FILL WITH ROCK FRAGMENTS

MOIST-BROWN (SM)

DO, WITH BRICK

BACKFILL

1-2

- 1=2- =23. ,2-1-.

1=2. 1=1: iltl 3-12

3 12=20. 16.0 22.- 42. SILTY SAND WTIH MICA, MOIST - GRAY AND BRCWN

iSU RESIDUAL

BORING TERMINATED AT 16.0

276
SCHNABEL ENGINEERING ASSOCIATES

CONSULTING ENGINEERS 1 TEST BORING LOG BORING NO! b-2h

PROJECTY'amrtantf PLAZA, BUILDING COLLAPSE INVESTIGATION SHEET NO : 1 OF 1 _ CLIENT! NATIONAL BIIRFATT OF STANDARDS JOB NO! CT870587

BORING CONTRACTOR rnMMTrrTTrrrr tp.ft RDR-mns. tmrQRILL*. rMPq^ elevation: -97 . 3+ WATER LEVEL DATA DRIVE SAMPLER CASING size:

DATE TIME DEPTH CAVED TYPE as. DATE START: 5-30—37

ENCOUNTERED 5-30 3EE NOT 3 BELOW DIA 2"O.D. DATE FINISHED: 5-30-87

AFTER CASING PULLED .5-30 WT. 140# DRILLER: J. DeANGELISS

HR. READING PFP WATPP TFT/FT. POP TF..PT FALL 30" inspector: r. n»vfttTr.T

STRATUM DEPTH FT. >

HI

-I

HI -27.3:

BLOWS ONSAMPLE SPOON,PER 6

Al -30

2.3

5.0

_

H

u

5

PIT TP-2H EAST

IDENTIFICATION

d" rn^rp-RTR

FINE TO COARSE SILTY SAND FILL WITH

GRAVEL - BROWN (SM)

ROCK FRAGMENTS FILL CUTTINGS TO 5 FT

BORING TERMINATED AT 5'

NOTE .—WATER USED DURING CORING. THEREFORE

WATER LEVEL IN BORING UNNATURALLY

HIGH DURING AND UPON BORING COMPLETION

NOTE: -NOT IN FOOTING AREA, BORING DONE

ADJACENT TO FOOTING

REMARKS

FLOOR SLAB AND

BACKFILL

CORE AND ROLLER

BIT BELOW 2

f 4"

277
SCHNABEL ENGINEERING ASSOCIATES I CONSULTING ENGINEERS I TEST BORING LOG BORING NO iB- 2HA

PROJECTIL ' AMBIANCE PLAZA, BUILDING COLLAPSE INVESTIGATION SHEET NO ; 1 OF

client: national bureau of standards job no: CT87Q587

BORING CONTRACTOR rnNrNFTTTrriT tfot borings. tntDRILL: gmfss elevation: -27. 3t

WATER LEVEL DATA DRIVE SAMPLER CASING size:

DATE TIME DEPTH CAVED TYPE SS. DATE START: 5-28-87

ENCOUNTERED 5-3 Q SEE NOTE BELP DIA. 2 O.D. DATE FINISHED: 5-28-87

AFTER CASING PULLED 5-

33. WT. 140# driller: J. DeANGELISS

HR. READING fi-17 OAVm ANn DRY AT 4fl'1 FALL W INSPECTOR: c. DeVAULT

STRATUM DEPTH FT.

Al 1.0

6.1

C 7.0

D

7.5

11.5

-

>

LLi -I

LlI -27.3

-30

103-125/5tl

REC=100%

RQD= 73%

s

V BROWN (SM)/

~

IDENTIFICATION

5—3/4" CONCRETE

FINE TO COARSE SILTY SAND FILL WITH GRAVEL

CONCRETE

DISINTEGRATED ROCK WITH MICA, MOIST - BROWN

TAND GRAY A

SLIGHTLY WEATHERED, HARD, GRAY SCHIST,

MODERATELY FRACTURED

BORING TERMINATED AT 11.5’

NOTE: -WATER USED DURING CORING. THEREFORE

WATER LEVEL IN BORING UNNATURALLY

HIGH DURING AND UPON BORING COMPLETION

NOTE:- FINAL WATER LEVEL READING OBTAINED

FROM TEST PIT CN EAST SIDE OF FOOTING

2H, LEVEL AT EL -32.2

NCTE:-WATER OBSERVED FLOWING INTO TEST PIT

EXCAVATION ON THE WEST SIDE OF FOOTING

2H. WATER FLOWING IN THROUGH FISSURES

IN ROCK AT OR UP TO ABOUT 6 INCHES

BELOW BOTTOM OF FOOTING

REMARKS

FLOOR SLAB AND

BACKFILL

FOOTING

RESIDUAL

PROSPECT

FORMATION

a. ROLLER BIT

7.5 FT TO

REMOVE

REINFORCING

STEEL

TO

278
SCHNABEL ENGINEERING ASSOCIATES I CONSULTING ENGINEERS I TEST BORING LOG BORING NO i b-3e

PROJECT!/ AMBIANCE PLAZA, BUILDING COLLAPSE INVESTIGATION SHEET NO : 1 OF 1

client: national bureau of standards

BORING CONTRACTOR rnNNFrTTrriT test springs, tncPRILL'.

job no: CT870587

elevation: -27. 0±

WATER LEVEL DATA DRIVE SAMPLER CASING size:

DATE TIME DEPTH CAVED TYPE SS. DATE START: 5-30-87

ENCOUNTERED 5z3.CL 3EE NOT S BELQn DIA 2 O.D. DATE FINISHED: 5-30-87

AFTER CASING PULLED I 5-30 WT. 140* driller: J. DeANGELISS

HR. READING 30 inspector:

23

<

<T X

fcfc

UJ

ELEV.

UJ - 3gi|5

_l

2

2

>

COMPLETION

IDENTIFICATION REMARKS

o

-27.0 . flD cO

OT q- CO 6.73" CONCRETE FLOOR SLAB AND

AI FINE TO COARSE SILTY SAND, FILL BACKFILL

-30 CONCRETE (FOOTING) FOOTING

4.8 /SILTY SAND WITH ROCK FRAGMENTS, MOIST -N A 8.8 8-8 - s RRTWN (PM) PROBABLE FILL

r fi.fl 88-100/3. c DISINTEGRATED RfTK f MOIST-BROWN AND GRAY RESIDUAL

\ REC=92% NX SLIGHTLY WEATHERED, HARD, GRAY SCHIST, PROSPECT

D RCD=75% \

SLIGHTLY TO MODERATELY FRACTURED FORMATION

11.7 \

\

j BORING TERMINATED AT 11.7’

NOTE: -WATER USED DURING CORING. THEREFORE * WATER LEVEL IN BORING UNNATURALLY

HIGH DURING AND UPON BORING COMPLETION

1

!

j

j€

||

?79
SCHNABEL ENGINEERING ASSOCIATES!

CONSULTING ENGINEERS TEST BORING LOG BORING NO!b-4.8e

PROJECTJ,’ amrtanjce plaza, building collapse investigation SHEET NO I 1 OF 1

client: national bureau of standards JOB NO! CT870587

BORING CONTRACTOR /-nNTMPrTTrrrr tfot RORTNnq. tnrDRILL: cmfrr elevation: -27.2+

WATE R LEVEL DATA DRIVE SAMPLER casing size:

DATE TIME DEPTH CAVED TYPE as. DATE START: 5-29-87

ENCOUNTERED 5-29 3EE NOTE BELOi DIA.

_ri _ _ 2 O.D. DATE FINISHED: 5-30-87

AFTER CASING PULLED 5-30 — 1 — WT. 140# DRILLER: J. DeANGELISS

« ry-v READING a-i i cawn awn nRv at q?. =FALL 30" INSPECTOR: r. n**vArrr,T 2

r><

<r A1

D

x

UJo

hi.

UJ-j

UJ -27.2

-30

cd

100/5.5"

-B£C=3.£1.

PQD=60%

s

>cc IDENTIFICATION

4" CONCRETE

FINE TO COARSE SILTY SAND, FILL

1WITO GRAVEL - BROWN (SM)|

CONCRETE

NX

DlSlNTHiKATEL) ~RJCK 7 MOIST - GRAY

FRESH TO SLIGHTLY WEATHERED, HARD,

GRAY SCHIST, MODERATELY FRACTURED

BORING TERMINATED AT 9.3'

NOTE: -WATER USED DURING CORING. THEREFORE

WATER LEVEL IN BORING UNNATURALLY

HIGH DURING AND UPON BORING COMPLETION

REMARKS

FLOOR SLAB AND

BACKFILL

FOOTING

RESIDUAL"

PROSPECT

FORMATION

280
281
282
SCHNABEL ENGINEERING ASSOCIATES 1 CONSULTING ENGINEERS ! TEST BORING LOG BORING NO 1 b-iof

PROJECTY.' AMRTANTF PLAZA, BUTTDING COLLAPSE INVESTIGATION SHEET NO ! 1 OF 1

CLIENT: mattonat. btipfatt of STANDARDS JOB NO! CT87D587 _

BORING CONTRACTOR ^MNTPr'Trrrrrr 'iwp RnoTNr:^ . t*jrORILL’. rMTTpt; elevation: -27. 4±

WATEIR LEVEL DATA DRIVE SAMPLER CASING size:

DATE TIME DEPTH CAVED TYPE as. date start: 5-28-37

ENCOUNTERED 5-38 3RF! BELOW DIA. 2"O.D. date finished: 5-29-37

AFTER CASING PULLED 5-29 * - T- WT. 140# DRILLER: J. DeANOELISS

_9_DAY READING fi-l 7 rwnm skti nay at aa" FALL 30" INSPECTOR: r. np\7AriT ,t 2

5

X

£L

x

fcfc

UJQ

0.9

>

UJ-iUJ -27.4::

-30

UJ - isils <D 5^°“

5-6-

10-39

jrop-m

FINE TO COARSE SILTY SAND FILL - {BROWN (SM)f

3

IDENTIFICATION

7" CONCRETE

CONCRETE

POORLX GRADED SAND WTIH SILT AND ROCK FRAGMENTS, TRACE MICA, MOIST - BRCWN (SP-SM)

SLIGHTLY WEATHERED, MODERATELY HARD TO

HARD GRAY SCHIST, MODERATELY FRACTURED

MODERATELY FRACTURED

REMARKS

FLOOR SLAB AND

BACKFILL

FOOTING

PROBABLE FILL

PROSPECT

FORMATION

BORING TERMINATED AT 11.1' a. ROLLER BIT TO

6'1" PRIOR TO

CORING

NOTE .‘-WATER USED DURING CORING. THEREFORE

WATER LEVEL IN BORING UNNATURALLY

HIGH DURING AND UPON BORING COMPLETION

NOTE: -WATER LEVEL MEASURED AT 62 INCHES

BELOW THE FLOOR SIAB IN THE TEST PIT

BETWEEN FOOTINGS 10F AND 10D ON 6-1-87

AND 6-17-87, WATER LEVEL AT EL -32.5

283
SCHNABEL ENGINEERING ASSOCIATES

GEOTECHNICAL CONSULTANTS

WEST CHESTER , PA.

TEST PIT LOG

Contract No. CT870587 Date Started : 6-1-87 Date Ended : 6-1-87

Test Pit NO. TP-1H (EAST FACE) Surface Elevation: -27.3

Project: L* AMBIANCE PLAZA BLDG. Groundwater Elevation: DRY

COLLAPSE INVESTIGATION

Location : WASHINGTON AND Equipment for Excavation : RUBBERTIRE BACKHOE/

COURTLAND, BRIDGEPORT, CT SEA Representative:

LOADER

C. DeVAULT

1 Depth Elev.

-27 .3

Stratum

Description of Soil and Observations — Remarks

_ l'_

“2.75

5" CONCRETE

FINE TO COARSE SILTY SAND FILL WITH GRAVEL, MOIST - FLOOR SLAB AND

BACKFILL

”1 BRCWN (SM) —

CONCRETE FOOTING

BACKHOE REFUSAL ON ROCK AT FOOTING SUBGRADE @ 2.75’ Entire east face

of footing exposed

(4.5 ft N to S)

Notes on materials at bottom of test pit:

1) Gray, hard, slightly weathered schist rock, thinly bedded.

2) ENE strike, very steep dip (appears to be greater than 70° north).

3) Backhoe barely able to gouge rock surface.

4) Contact between footing concrete and bedrock continuous.

284
SCHNABEL ENGINEERING ASSOCIATES

GEOTECHNICAL CONSULTANTS

WEST CHESTER, PA.

TEST PIT LOG

Contract No. CT870587 Date Started: 6-1-87

Test Pit No. TP-1H (SOUTH FACE) Surface Elevation:

Project: L' AMBIANCE PLAZA BLDG.

COLLAPSE INVESTIGATION

Location: WASHINGTON AND

COURTLAND, BRIDGEPORT, CT

Groundwater Elevation:

Date Ended: 6-1-87

-27.3

DRY

Equipment for Excavation: RUBBERTIRE BACKHOE/

LOADER

SEA Representative: C. DeVAULT

Depth Elev.

-27.3

Stratum

Description of Soil and Observations Remarks

_ l'_

~2.75

5" CONCRETE

FINE TO COARSE SILTY SAND FILL WITH GRAVEL, MOIST - FLOOR SLAB ANDBACKFILL

[BROWN (SM) I CONCRETE

1 FOOTING

I

! BACKHOE REFUSAL ON ROCK AT FOOTING SUBGRADE @ 2.75' About 3 ft of

south face exposed

extending front SE

corner

Notes on materials at bottom of test pit:

1) Gray, hard, slightly weathered schist rock, thinly bedded.

2) ENE strike, very steep dip (appears to be greater than 70° north).

3) Contact between footing concrete and bedrock continuous.

285
SCHNABEL ENGINEERING ASSOCIATES

GEOTECHNICAL CONSULTANTS

WEST CHESTER , PA.

TEST PIT LOG

Contract No. CT870587

Test Pit No. TP-2H (WEST FACE) Surface Elevation:

Date Started: 6-17-87 Date Ended: 6-17-87

-27.3

Project: L

1 AMBIANCE PLAZA BLDG.

COLLAPSE INVESTIGATION

Location: WASHINGTON AND

COURTLAND, BRIDGEPORT ,CT

Groundwater Elevation: -32.2 (SEE NOTES)

Equipment for Excavation: CRAWLER MOUNTED

HYDRAULIC BACKHOE

SEA Representative: C. DeVAULT

Depth Elev.

-27.3

Stratum

Description of Soil and Observations Remarks

l

1

,

5-6" CONCRETE

FINE TO COARSE SILTY SAND FILL WITH GRAVEL, MOIST - FLOOR SLAB AND

BACKFILL

BROWN (SM) 1

'

i

— —

CONCRETE FOOTING

6’

BACKHOE REFUSAL ON ROCK AT FOOTING SUBGRADE @

6’ About 14 ft of

west face of

footing exposed

extending north

from SW corner.

Notes on materials at bottom of test pit:

1) Gray, hard, slightly to moderately weathered schist rock, thinly bedded.

Shiny, sparkling appearance from mica content.

2) ENE strike, very steep dip (appears to be about 80 to 85° north).

3) Backhoe able to gouge rock surface with difficulty.

4) Joints observed extending perpendicular to bedding planes and nearly

vertical. At about 3 ft centers. Joint face hard, with rust discoloration.

5) Up to about 1.0 inch wide cracks in joint face on west face of test pit,

extend down to footing subgrade level, traced discontinuously across bottom

of test pit but not observed extending beneath footing.

6) The cracks (Note 5) are about 4 ft north of SW corner and water was observed

flowing into test pit from the cracks in two places. One in crack on side of

test pit with flow issuing from up to about 6 inches above footing subgrade.

The other crack was in the bottom of the test pit about 1.0 ft west of the

footing

286
Boring Log TP-2H (West Face)

Page Two

7) Water being pumped from test pit during our observations. Did not reach a stable level while we were on-site.

8) See TP-2H (east face) for water level.

9) Strand of yellow wire extending about 3 inches from beneath footing at a point about 13 ft north of SW corner.

10)

Contact between footing concrete and bedrock continuous and tight except near

mentioned crack area. Along a 2 to 2.5 ft strip there were loose rock

fragments which could be pulled from beneath the footing, leaving up to 6

inches deep pit beneath footing subgrade. Pulled fragments from as far as 6

inches back from side face of footing.

287
SCHNABEL ENGINEERING ASSOCIATES,

GEOTECHNICAL CONSULTANTS

WEST CHESTER, PA.

TEST PIT LOG

Contract No. CT870587

Test Pit No. TP-2H (SOUTH FACE) Surface Elevation:

Date Started: 6-17-87 Date Ended: 6-17-87

-27.3

Project: L

1 AMBIANCE PLAZA BLDG.

COLLAPSE INVESTIGATION

Location: WASHINGTON AND

Groundwater Elevation -32.2 (SEE NOTES)

COURTLAND, BRIDGEPORT, CT

Equipment for Excavation: RUBBERTIRE BACKHOE/

LOADER

SEA Representative: C. DeVAULT

Depth Elev.

-27.3

Stratum

Description of Soil and Observations Remarks

1'

5-6" CONCRETE

FINE TO COARSE SILTY SAND FILL WITH GRAVEL, MOIST - FLOOR SLAB AND

BACKFILL

6

’

BRCWN (SM)[

CONCRETE FOOTING

BACKHOE REFUSAL ON ROCK AT FOOTING SUBGRADE @

6' About 6 to 7 ft

of south face

exposed extending

east from SW

corner

Notes on materials at bottom of test pit:

1) Gray, hard, slightly to moderately weathered schist rock, thinly bedded.

2) ENE strike, very steep dip (appears to be about 80 to 85° north).

3) Contact between footing concrete and bedrock continuous and tight.

4) Test pit dry but water being removed by sump in western portion of test pit.
SCHNABEL ENGINEERING ASSOCIATES

GEOTECHNICAL CONSULTANTS

WEST CHESTER , PA

.

TEST PIT LOG

Date Started: 6-1-87 Date Ended: 6-1-87

Surface Elevation: -27.3

Groundwater Elevation: -32.2 (SEE NOTES)

Equipment for Excavation: CRAWLER MOUNTED

HYDRAULIC BACKHOE

SEA Representative: C. DeVAULT

Depth Elev.

-27.3

Stratum

Description of Soil and Observations Remarks

_____

1'

5-6" CONCRETE

FINE TO COARSE SILTY SAND FILL WITH GRAVEL, MOIST - FLOOR SLAB AND

BACKFILL

BROWN (SM)/

— —

CONCRETE FOOTING

6' BACKHOE REFUSAL ON ROCK AT FOOTING SUBGRADE @ 6' About 12 ft of

east face of

footing exposed

extending north

from SE corner.

Contract No. CT870587

Test Pit No. TP-2H (EAST FACE)

Project: L' AMBIANCE PLAZA BLDG.

COLLAPSE INVESTIGATION

Location: WASHINGTON AND

CQURTLAND, BRIDGEPORT , CT

Notes on materials at bottan of test pit:

1) Water in test pit at time of observations. Not removed entirely by sumps

during observations.

2) Water level assumed to be perched in rock surface from coring water and

recent test boring work.

3) Follow-up observations 6-17-87, water level at 4.9 ft depth, EL -32.2. Water

noted flowing into west face of test pit through fissures.

4) With water in test pit, observations of subgrade made by probing and feeling

materials and interface between footing and rock.

5) Hard rock at bottom of test pit elevation comparable to footing subgrade.

6) Contact between footing concrete and bedrock feels continuous.
SCHNABEL ENGINEERING ASSOCIATES

GEOTECHNICAL CONSULTANTS

WEST CHESTER, PA.

TEST PIT LOG

Contract No. CT870587 Date Started: 6-1-87

Test Pit No. TP-1OP (SOUTH FACE) Surface Elevation:

Date Ended

-27.3

6-1-87

Project: L' AMBIANCE PLAZA BLDG. Groundwater Elevation:

COLLAPSE INVESTIGATION

Location: WASHINGTON AND

DRY

COURTLAND, BRIDGEPORT, CT

Equipment for Excavation: CRAWLER MOUNTED

HYDRAULIC BACKHQE

SEA Representative: C. DeVAULT

Depth Elev.

-27.3

Stratum

Description of Soil and Observations Remarks

_ l'_

~4.5“

4" CONCRETE

FINE TO COARSE SILTY SAND FILL WITH GRAVEL, MOIST - FLOOR SLAB AND

BACKFILL

BRCWN ( SM) | CONCRETE FOOTING

• BACKHOE REFUSAL ON ROCK AT FOOTING SUBGRADE AT ABOUT About 6 ft of

4.5’ south face exposed

extending from SW

corner

Notes on materials at bottom ’of test pit:

1) Gray, moderately hard, moderately weathered schist.

2) ENE strike, very steep dip (appears to be 85° + north)

3) Contact between footing concrete and rock continuous.

4) Water at 4.6 ft depth (6-1-87).

5) Subgrade dry 6-17-87.

290

ta
Legend for materials in sketch:

•Thinly laminated micaceous zone. Highly weathered zone within the

rock matrix, which is flaked easily by hand.

291
Boring Log TP-10F (North Face)

Page Two

Gray, moderately weathered, medium to moderately hard schist.

Highly fractured, both vertically and horizontally. Rock

fragmentsA>locks can easily be removed by hand.

Void spaces between rock bedding planes and fractures up to 1/2 inch

width probed to 3 inch depth.

Rock fragments appear to have been loosened from natural orientation.

Silt, sand rock fragment fill, moist - brown.

o

¥

Golf-ball size gravel piece.

Gray, slightly weathered, hard schist.

ENE strike, very steep dip (nearly vertical)

Water level measured at EL -32.5 (18 inches below bottom of footing).

292
SCHNABEL ENGINEERING ASSOCIATES

GEOTECHNICAL CONSULTANTS

WEST CHESTER, PA.

TEST PIT LOG

Contract No. CT870587 Date Started: 6-1-87

Test Pit No. TP-10F (WEST FACE) Surface Elevation:

Project: L

1 AMBIANCE PLAZA BLDG.

COLLAPSE INVESTIGATION

Location: WASHINGTON AND

COURTLAND, BRIDGEPORT, CT

Groundwater Elevation:

Date Ended: 6-1-87

-27.3

DRY

Equipment for Excavation: CRAWLER MOUNTED

HYDRAULIC BACKHOE

SEA Representative: C. DeVAULT

Depth Elev.

-27.3

Stratum

Description of Soil and Observations Remarks

-

~3.7~

6" CONCRETE

FINE TO COARSE SILTY SAND FILL WITH GRAVEL, MOIST - FLCCR SLAB AND

BACKFILL

BRCWN 1 CONCRETE FOOTING

BA2KHOE REFUSAL ON ROCK AT FOOTING SUBGRADE @ 3.7'

of west face ex- posed extending

from NW corner

.

Notes on materials at bottcm of test pit:

1) Gray, hard to moderately hard, slightly to moderately weathered schist,

thinly bedded.

2) ENE strike, very steep dip.

3) Contact between footing concrete and bedrock continuous.
SCHNABEL ENGINEERING ASSOCIATES

GEOTECHNICAL CONSULTANTS

WEST CHESTER, PA

TEST PIT LOG

Date Started: 6-17-87 Date Ended: 6-17-87

Surface Elevation: -10.3

Groundwater Elevation: DRY

Equipment for Excavation: CRAWLER MOUNTED

HYDRAULIC BACKHOE

SEA Representative: C. DeVAULT

Depth Elev.

-10.3

Stratum

Description of Soil and Observations Remarks

FINE TO COARSE SILTY SAND FILL WITH GRAVEL, ROCK Loose to compact

FRAGMENTS, BRICK, MOIST - BROWN (SM) density

— — A1

~5.5r

*

TEST PIT TERMINATED @ 5.5'

Contract No. CT870587

Test Pit No. TP

Project: L' AMBIANCE PLAZA BLDG.

COLLAPSE INVESTIGATION

Location: WASHINGTON AND

COURTLAND, BRIDGEPORT, CT

Notes

1) Two shelby tubes pushed with backhoe bucket from 5.5 to 7.0 ft depth.

2) Two field density tests performed at 5.5 ft depth using sand-cone method

(oversize material replaced in cone test hole)

294

Appendix C

PREVIOUS TEST BORINGS BY OTHERS

296
AUiATUCK, COMM. 0*770 | TEST l8IINt REPORT rufNT T P n INTERNATIONAL, INC.

BORING NO. PSm‘ 1

• BORING NO ps“2

LINE A STA. LINE A STA.

’ OFFSET OFFSET

GR. ELEV.

* GR. ELEV.

October 23, 1985 October 22, 1985

A STRATUM DESCRIPTION

0CNUT7 o«

COMtST.

•107%

HI

8 A STRATUM DESCRIPTION

MNSTT

o« COW.

KO*1

ft* •“ B

3 0.2 Bit. Conor.

Br. (W Sand 4 Grav., • Br. I*l-F Silty Sand,

CobbIs9. some C-F Grav.

0 n t/.Comp 10-21 6.0 n H. Comp 7-10

Decomposed Rock Dry 35 Decomposed Rock Dry 13

V.Comp

Dry

23-31 tt Dense

Dry

Rec:0’

50/2"

36

14.0

9 Runffl: 14.0 to 17.0

n No earn: C.mate

ile due

ial

' 12 Recovery:23"

17.0 11 Phvllites

12 Huoff^ : .17. U to 22.5

Tfi Recovery : 72" . ?g Fractured /

ft No sam lie du< i jL2 r,< to C.m iteria’ 22.5 ]\

1 4 Run#3:22.5 Nto 24.0

Recovery: 13" 0 Run#l:23.5 to ^5.0 24.0 12 • 0 13 Recovery : 12” Fracture . iia Rur>j?2: 25.0 to 2ti.5 End of Boring-24.0

16 Recovery: 54" Fractur 3d

17 Phylllte GUO—None

T 5 20

19 Run#3: 28.5 to 33.5

1

. 22 Recovery: 21" * J2SL F ractured

,5 •

End of Boring-33.5

GUO-Nrne

I

!

J 1 C0L’ A B#ow« OO C*«if»f * DRILL TIME PER FOOT

* 8 Blo«% w (I.D.) 3 HAMMER - 140 FALL 30'*

1 A SAMPLER = O. D. SPLIT SPOON

1 3 GWO - GROUND \TATER OBSERVATIONS

FIELD — % CONTEN

AND — 40 to 507

SOME — 10 to 407

TRACE — 0 to 107

T

ci

a 297
Apartment Complex, Washington Ave

Bridgeport, Connecticut

ASSOCIATES SOniNOS CO., INC.

!f« MANOAHCT OINOLC

MAUMTUCK, CONK. 0«T70

PMOMC 7I«'I4II

TEST BOHNS IEP 0 RT

PTOJ.

cuisnt T P W INTERNATIONAL, INC,

BORING NO. ps"3 BORING NO.

LINE A STA. LINE A STA._

* OFFSET OFFSET

GR. ELEV. _____ GRo ELEV.

DATE October 24, 1985 DATE„

8.0

liufl. 16.0

2n.n

24.

29.0

34.0

STRATUM DESCRIPTION

dcnsitt ON

comm.

tO'TS «4“ 8

Sit. Concr,

8r. n~F Sand, some H—

Grav.,Slag, Brick,

Glass^Cobbles,

Boulders.

Decomposed Rock

7?

15

10

UL

IX

II

TT

10

ii

12

16

IX

Run^l: 14.0 to 16.0

Recovery: 30"

Run#2: 16.0 to 2Q e Q

Recovery: 48”

Fractured PhylliteA uoff3: l^u.U to 14.

Recovery :18"

24. u co zy.il

Recovery : 57”

Fractured

RunflfO: 29.0 to 34.0

Recovery: 42”

F ractured

End of Boring—34.0

GldO-i'.’one

11

34

5

Loose

Dry

4-5

5

Dense

Dry

22-36

50/0”

1.

— ——, - l xiui run COL. B Glow* on 1 %** S«ipJ«r (I.D.) HAMMER • 140 *, FALL 30 ** SAMPLER = O. D. SPLIT SPOON GWO _ GROUND WATER OBSERVATIONS

A

otwsrrr STRATUM DESCRIPTION CCXSCT. non

«* -

B

FIELD — % CONTENT

AND — 40 to 50%

SOME — 10 to 40%

TRACE — 0 to 10%
a o c

i

atc o aaniNaa ca., inc.

If* MAR9AMCT eiMOCC

NAU9ATUCK, COMM. 04170

RHOMC TI 4 - 44 I 4 TEST BOHNS REPORT

Apartment Complex, Washington Av/s af

pwoJ Bridgeport, Connecticut

TPH INTERNATIONAL, INC.

CLIENT —

BORING NO..

LINE A STA._

OFFSET

GR. ELEV.

ATE July 18» i g85

^ STRATUM DESCRIPTION

BORING NO.

LINE A STA..

OFFSET

GR. ELEV.

OCNUTT o«

consist.

KOVt

ffl 4“ 8

DATE July 11, 1985

STRATUM DESCRIPTION

oenurr o« COM8T.

•CO*"!

Ft* 4“ B

l.o L

11.0

14.0

19.

C

Toosoil

rted or . iw SXXty

Sand, some F. Qrav.

Oeccmp^^ocIT

12

1 ?12

Gneiss

Run#l: 14.0 to 19.0

Recovery: 12"

Fractured 4 Seamy

End of 8oring-19.0

GUo-None

Comp

Dry

Oenae

Dry

16-17

21

13-10

50

25.0

1 COL. A Blow, oa DRILL TIME PER FOOT

* £OL. B Blow* o*i 1 Siipj^r (|.D.) 3 HAMMER - 140 FALL 30 ** 299

4 SAMPLER = O. D. SPLIT SPOON

5 GWO - GTOUND WATER OBSERVATIONS

i a12

15

.15

10

Br. PI—F Sand,Tr. fl-F

Grav.,Tr. Oacomp.

Rock, Rock Frags.

bneiss

Run#l:20.0 to 25.0

Racovary: 20 " Fractured 4 Seamy

End of Boring-25.0

GWO-None

ense

Dry

10-25

50

Danse

Ory

Rec:0"

50/1"

Dense

Dry

50

FIELD — % CONTENT

AND — 40 to 30%

SOME — 10 to 40%

TO A CZ rs « .
- • ».< >« ASSOCIATED lORINOa CO., INC.

»»• MA4VANCT OIMOCC

nau«ATUCKi OONN. 0*170

SMOMI '/IM4JO

TEST BORINS REf 8RT

PNOJ.

Apartment Complex(Ua9hing ton Ava.

Bridgeport. Connecticut

client T P H CONSTRUCTORS. INC.

BORING NO..

LINE A $TA._

OFFSET

GR. ELEV.

DATE Jjns 26, 1985

STRATUM DESCRIPTION

MNSJTV om

COr*«SV. RO«1

HI »" B

0.2

3.0

12.0

16.0

i 7 i Gneiss

21.0

Bit. Conor.

TOTTW band

Br. M-F Silty band, some fl-C Grav., Tr.

Mica

Br. PI-F Sand.

17

19

20

11

Run#l: 16.0 to 21.0

Recovery? 18"

F r actucsd A Seamy

End of 3oring-21.0

GVJO— None

V.Comp

Dry

17-30

21

V.Comp

Ory

R

18-25

26

Dense

Ory

50/5"

0.1

2.5

47U

570

10.0

w unrii, Tine rert

T COL. B Slow* o* 1%" S*o»pf*f (I.D.) 3 HAMMEO — 1 40 * , FALL 30 ** 4 SAMPLER = O. D. SPLIT SPOON

5 GWQ - GROUND WATER OBSERVATIONS

300

BORING NO..

LINE A STA._

OFFSET

17

GR. ELEV..

DATE Pune 26, 1985

^ STRATUM DESCRIPTION

MMOTT

o«

c«

-in

Bit. Conor.

bln . n-r 5iir y 5xne, some H-F Grav.v,

Br. M—T^TTty Sand

Ugcomposaa nocK

Gnexse

Run#l: 5.0 to 10.0

Recovery: 12"

Fractured A Seamy

End of Boring—1Q.0

GUO-None

FIELD — % CONTENT

AND — 40 to 50%

SOME — 10 to 40%

TRACE — 0 to 1C%

110*1

ra

B
ASSOCIATED BOftlNO* CO., INC.

US MAMSAACT CinOLC

NAU9ATUCK, COMM. 0*770

AMOMC 7U*»4I»

TEST BOHNS REPORT

PBOJ.

Apartment Complex, Washington Ave

Bridgeport, Connecticut

CUIHT TPP1 INTERNATIONAL, INC.

BORING NO..

LINE A STA._

OFFSET

GR. ELEV

BORING NO..

LINE A STA._

OFFSET

DATE ^uly 11, 1985

^ STRATUM DESCRIPTION

GR. ELEV.

date 3uiy 1985

14.5

10

10

19.5

12

10

10

OCNSITT OS

CONSIST.

•COSTS «*" B

ocmsht

STRATUM DESCRIPTION cosm.

non

HM •** B

Br. 1*1—F Sand,Tr. 1*1—

F

Grav.,Tr. Decornp.

Rock f’rags.

Goexsts

Run#l: 14.5 to 19.5

Recovery: 31”

Fractured 4 Seamy

End oF Boring-19.

5

GUO- None

il.Comp

Ory

10-7

7

Dense

Dry

13-20

21

/

4.0

10.0

15.0

JJL

12

JSL

Uark dr. fT—

F

iailty

Sand, some H-F Grav.

QecompT^TacJT

Gneiss

Run#l: 10.0 to 15.0

Rscovery: 22”

Fractured 4 Seamy

End of Boring-15.0

GUO-None

Comp

Ory

10-21

20

»

^ ^L. 8 0( jwi m i%" S**ipi*r (I.D.)

3 HAMMER - 140 », FALL 30** 4 SAMPLED = O. D. SPLIT SPOON

5 GWO - GROUND WATFP OfKFRVA TIOkjc

301

FIELD — % CONTENT

AN0 — 40 to 50%

SOME —10 to 40%
1 1 « mamsakct oincli

M*U«ATUCK, COMM. OA7TO

MON< 7M<I4«I

TEST BOHNS IEPQRT

P«OJ. Bridgeport, Uonnecticut

clicnt T P n INTERNATIONAL , INC.

BORING NO.

LINE A STA._

OFF$ET__

GR, ELEV.

DATE July 16, 1965

f STRATUM DESCRIPTION

DCNSIYT Ot

CONSIST.

PCO'TS

••• B

JUL

14.

19. C

12

12

JLQ

Dark Br. P1-F Sand, some N—F Grav.

Cobbles

Lt. Br. H-F Sand, aome fl-F Grav.

2SL —

Gneiss

Run#l: 14.0 to 19.0

Recovery : 8.0

F ractured 4 Seamy

with deromo. rock

•spp- End of 0oring-19.O

GUO—None

Loose

Dry

2-2

3

• •

Dense

Dry

50

•

8,0

10,0

15.0

^ COL. B Slews on Sniplnf (I.D.) 3 HAMMEfc - 140*, FALL 30“

4 SAMPLER = O. D. SPLIT SPOON

5 GWO - GROUND WATER OBSERVATIONS

302

BORING NO..

LINE A STA._

OFFSET.

GR. ELEV.

DATE July 15, 1985

A STRATUM DESCRIPTION

OtNffTT o«

CONttST.

PLO^S

«* *** B

n

JJi

-UL

ii

Pcs., Fill Area.

Uecomp^TocJT

Gneiss

Run#ls 10.0 to 15.0

Recovery: 24"

Fractured 4 Seamy

jirithJOecomo^Roci^^

End of Bering-15.0

GUQ-None

U

Loose

Dry

2-2

4

/

•

*

FIELD — % CONTENT

AND — 40 to 50%

SOME — 10 to 40%

TRACE — 0 to 10%
ASSOCIATED SOftlNOS CO., INC.

t1« MAMnAKCT OIHCLC

NAUflATUCK, COMM. 0*770

PHONI ,71 «*|4I«

TEST SBIINS REPORT

Apartment Complex, Washington A vs pwoJ Bridgeport, Connecticut _____ CLIENT T P H INTERNATIONAL, INC.

BORING NO.

LINE A STA._

OFFSET

a

GR. ELEV.

DATE July 15, 1985

STRATUM DESCRIPTION

OCNSITT o« comm

M.OV1

B

4.0

5.0

i n . n

TT

12

in.

in.

in

Br. Pl-F Sand & Grav.,

Cobbles.

Oecomp. Rock

Gneiss

Run#l: 5.0 to- 10.0

Recovery: 12”

Fractured A Seamy

End of Boring-10.0

GWQ-None !

.

3.0

7.5

12.5

1

1 COL- A <*» * DRILL TIME PER FOOT

T COL. B Blow* on 1 %’* Stapjif (I.D.)

3 HAMMER - 140*, FALL 30' 4 SAMPLER = O. D. SPLIT SPOON

303

5 _ r r ni ivjn w*rm ^o<*nj\y A r.

BORING NO..

LINE A STA._

OFFSET

GR. ELEV.

DATE July 15 » 1985

A STRATUM DESCRIPTION

DCMvrr o« comm.

rto^i

«* •’ B

13

Br. n-f Sand, some

H—F Gray.

ecompTTTociT

11

11

JA

a

Gneiss

Run#l: 7.5 to 12.5

Recovery: 42"

Fractured A Seamy

Oenee

Bry

End of . Boring-12.5

GUO-Nons

13-50

FIELD — % CONTENT

AND — 40 to 30%

SOME —10 to 40%
I l

<= -i 5 o c* pJ . -r-» jT

<T, *0' c*v i r

TJ . <?I • -i

Sj. C

i>

r~ \

t'. ^t}\ ^ -r • ‘ \ <J* <_-> J ' , •’ } \ <-«* v/-*
Appendix D

BIBLIOGRAPHY

'

305
Appendix D

BIBLIOGRAPHY

(1) Bowles, Joseph E. , (1968), Foundation Analysis and Design , McGraw Hill, New York, 659 pp

(2) Bowles, Joseph E. , (1982), Foundation Analysis and Design , 3rd Edition, Table 5, McGraw Hill, New York.

(3) Gardner, W.S., (1987), "Design of Drilled Piers in the Atlantic

Piedmont" , Foundations and Excavations in Decomposed Rock

of the Piedmont

- Pr ovine

^

T Geo technical^ Special "Publication

No. 9, ASCE, April 28, 1987, pp. 62-86.

(4) Lambe, T. William, and Whitman, Robert V. , (1969), Soil

Mechanics , John Wiley and Sons, New York, 553 pp.

(5) Martin, R.E. , (1987), "Settlement of Residual Soils", Foundations and Excavations in Decomposed Rock of the Piedmont

Prov ince , Geotechnical Special Publication No. 9, ASCE,

April 28, 1987, pp. 1-14.

(6) Meyerhof, C.G., (1956), "Penetration and Bearing Capacity in

Cohesive Soils", Journal of Soil Mechanics and Foundation

Division , ASCE , Vo IT 82 , NoT SM l".

‘

(7) NAVFAC DM- 7.1, (1982), Soil Mechanics , Department of the Navy,

Alexandria

(8) Riggs, Charles 0., Mathes, Gary M. , and Rassieur, Charles L.,

( 1984) , "A Field Study of an Automatic SPT Hammer System",

reprint from Geotechnical Testing Journal , September 1984.

(9) Schmertmann, J.H., Hartman, J.P., and Brown, P.R., "Improved

Strain Influence Factor Diagrams", Journal of the Geotechnical Division, ASCE, Vol . 104, No. GT8 , August 1978, pp . 1131-

1135.

(10) Sowers, George F. , ( 1979), Soil Mechanics and Foundations: Geotechnical Engineering , MacMillan Publish ing ’ CoTT New

- York”

621 pp.

(11) Winterkorn, Hans F., and Fang, Hsai-Yang, Editors, (1975),

Foundation Engineering Handbook , Van Nostrand Reinhold

Company, New York, 751 pp

306
APPENDIX C

Resistance Criteria for Columns and Floor Slabs

307
APPENDIX C

RESISTANCE CRITERIA FOR COLUMNS AND FLOOR SIj\BS

Columns

AISC - Reference 8

^n " Ag^cr

for A c < 1.5

2

Fcr - (0.658 c) Fy

for A c > 1.5

0.877

c

where

K1

' c *

rn

Pn - nominal axial strength

Ag - gross area of member, in.^

Fy - specified yield stress, ksi

E - modulus of elasticity, ksi

K - effective length factor

1 - unbraced length of member, in.

r = governing radius of gyration about plane of buckling, in.

Floor Slabs

ACI - Reference 2

Punching Shear

Vn

V, ( 3 . 57fi + 0 . 3fp C )b 0 d + V

p

where Vn nominal shear force

specified compressive strength of concrete, psi

average value of effective stress at centroid of section

perimeter of critical section

structural depth of slab at support

vertical component of effective prestress forces, taken as zero for this analysis.

308
Alternatively,

times the slab

slab, where:

for exterior columns having a slab edge distance less than four

thickness, shear capacity is calculated as for a nonprestressed

vc - (2 + 4//3 c )7f' b 0 d

where /? c -

bo -

d

ratio of lengths of long, short sides of reaction area

perimeter of critical section

structural depth of slab at critical section

Moment Capacity

% - %

Ap S ~

fps ~

d

Aps fps (^.-a/2)

Nominal moment capacity of unit width strip of slab

area of prestressing strand

effective stress in prestressing strand

depth from extreme compression fiber to centroid of prestressing

strand at location where moment is evaluated

a =- depth of uniform compression stress block at nominal moment

capacity based on ACI assumptions of stress block

309
NBS-114A (rev. 2-80

U„S. DEPT. OF COMM.

BIBLIOGRAPHIC DATA

SHEET fSee instructions)

1. PUBLICATION OR

REPORT NO.

NBSIR 87-3640

2. Performing Organ. Report No. 3. Publication Date

4. TITLE AND SUBTITLE

INVESTIGATION OF L' AMBIANCE PLAZA BUILDING COLLAPSE IN BRIDGEPORT, CONNECTICUT

5. auth0R(S) CharleS G. Culver, Charles F. Scribner, Richard D. Marshall, Felix Y. Yokel,

John L. Gross, Charles W. C. Yancey, Erik M. Hendrickson

6. PERFORMING ORGANIZATION (If joint or other than NBS, see instructions) 7. Contract/Grant No. NATIONAL BUREAU OF STANDARDS

U.S. DEPARTMENT OF COMMERCE 3 . Type of Report & Period Covered

GAITHERSBURG, MD 20899

9. SPONSORING ORGANIZATION NAME AND COMPLETE ADDRESS (Street, City, State, ZIP)

The Occupational Safety & Health Administration

Department of Labor

Washington, D.C. 20001

10. supplementary notes

I

1 Document describes a computer program; SF-185, FlPS Software Summary, is attached.

11. ABSTRACT (A 200-word or less factual summary of most significant information. If document includes a si gnificant

hi bl iography or literature survey, mention it here)

Results from an investigation to determine the cause of the collapse of the

L ‘Ambiance Plaza building on April 23, 1987 are presented. The building was

being constructed using the lift-slab method; collapse occurred during

construction. The investigation included on— site inspections immediately

following the collapse, review of eyewitness accounts of the collapse, review

of project documentation, laboratory and field tests and analyses of the

structure. Several potential failure mechanisms were investigated. The most

probable cause of the collapse was determined to be loss of support at a lifting jack in the west tower during placement of an upper level package of

three floor slabs. The loss of support was likely due to excessive

deformation of the lifting angle in a shearhead followed by a lifting nut

slipping off the lifting angle of the shearhead. The postulated failure

mechanism was duplicated in laboratory experiments. The local failure

propagated as loads were redistributed. The remaining jack rods along column

line E supporting the package of floor slabs slipped off the lifting angles

and the slabs failed in flexure and shear . These slabs fell causing the lower

level slabs to fail.

12. KEY WORDS (Six to twelve entries; alphabetical order; capitali ze only proper names; and separate key words by semi colon s) building; collapse, concrete; construction failure; lift-slab construction;

post-tensioned concrete; progressive collapse.

13. availability 14. NO. OF

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