MST OF STAND & TECH

Ai.i.l.n7 536133_





National Bureau of Standards

Library, E-01 Admin. Bldg.

OCT 6 1981

191118 QC

ICO

NBS TECHNICAL NOTE 811

Evaluation of the Column

Connections Used in a Precast

Concrete Modular Housing System
NATIONAL BUREAU OF STANDARDS

The National Bureau of Standards ' was established by an act of Congress March 3, 1901. The Bureau's overall goal is to strengthen and advance the Nation's science and technology

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1 Headquarters and Laboratories at Gaithersburg, Maryland, unless otherwise noted; mailing address Washington, D.C. 20234.

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3 Located at Boulder, Colorado 80302. Part of the Center for Building Technology.
Evaluation of the Column Connections

Used in a Precast Concrete Modular

Housing System

National Bureau of Standards APR 2 9 1974 Felix Y. Yokel and Thomas W. Reichard

Center for Building Technology

Institute for Applied Technology

National Bureau of Standards

Washington, D.C. 20234

**>*<</

%fAU 0* *

U.S. DEPARTMENT OF COMMERCE, Frederick B. Dent, Secretary

NATIONAL BUREAU OF STANDARDS, Richard W. Roberts, Director

Issued March 1974
National Bureau of Standards Technical Note 811

Nat. Bur. Stand. (U.S.), Tech. Note 811, 63 pages (Mar. 1974)

CODEN: NBTNAE

Supersedes NBSIR 73-148 (PB 220-366/7)

U.S. GOVERNMENT PRINTING OFFICE

WASHINGTON: 1974

For sale by the Superintendent of Documents, U.S. Government Printing Office, Washington, D.C. 20402 (Order by SD Catalog No. 03.46:811). Price $1.00
SI Conversion Units

In view of the present accepted practice in this country for building

technology, common US units of measurement have been used throughout this

paper. In recognition of the position of the United States as a signatory

to the General Conference on Weights and Measures , which gave official

status to the metric SI system of units in 1960, assistance is given to

the reader interested in making use of the coherent system of SI units

by giving conversion factors applicable to US units used in this paper.

Length

1 in = 0.0254 meter (exactly)

1 ft - 0.3048 meter (exactly)

Force

1 lb (lbf) = 4.448 Newton (N)

1 kip = 4448 Newton

Pressure

1 psi = 6895 N/m 2

1 ksi = 6.895 x 10

6 N/m 2

in
Table of Contents

Evaluation of the Column Connections

Used in a Precast Concrete Modular Housing System

Page

List of Figures v List of Tables vi

1. Introduction 1

2. The Building System 2 2.1 Description of System 2 2.2 System Features That Required Performance Evaluation 6

3. Scope 11

4. Tests 12

4.1 Bearing Capacity of Column Connections Using Various

Types of Bearing Pads. 12

4.1.1 Objective 12

4.1.2 Test Specimens 13

4.1.3 Test Procedure 19

4.1.4 Test Results 19

4.1.5 Discussion of Test Results 19

4.2 Tests to Determine the Load-Deformation Characteristics

of the Neoprene Pads

4.2.1 Objective 27

4.2.2 Test Specimens 27

4.2.3 Test Procedure 28

4.2.4 Test Results and Findings 28

4.3 Tests to Evaluate the Bearing Capacity of Connections Using

a Steel-Neoprene-Steel Sandwich as a Bearing Pad and a Grouted Dowel 31

4.3.1 Objective 31

4.3.2 Test Specimen 31

4.3.3 Test Procedure 33

4.3.4 Test Results 33

4.3.5 Discussion of Test Results 33

4.4 Tests to Evaluate the Shear Capacity of Column Connections 39

4.4.1 Objective 39

4.4.2 Test Specimen 40

4.4.3 Test Procedure 40

4.4.4 Test Results 44

4.4.5 Discussion of Test Results 44

5. Summary 5 3 6. Acknowledgement 55

7. References 56

iv
List of Figures

Page

Figure 2.1. Proposed Building 3 Figure 2.2. Building Complex Under Construction 3 Figure 2.3. Proposed Modular Unit 4 Figure 2.4. Structural Skeleton of Module 4 Figure 2.5. Checkerboard Arrangement of modules 5 Figure 2.6. Completion of Structural Assembly 5 Figure 2.7. Schematic Drawing of Connection Tiedown

Figure 2.8. Detail of Interior Column Joint 7 Figure 2.9. Proposed Connection Detail 8 Figure 4.1. Specimen Used in Phase I Column Connection Tests 14

Figure 4.2. Specimen Used in Phase II Column Connection Tests 15

Figure 4.3. Typical Details of Test Specimens 16

Figure 4.4. Strength of Age Relationship of Concrete Used

for Column Stubs 17

Figure 4.5. Typical Failure of Connection Using 1/4-in

neoprene Pad 23

Figure 4.6. Typical Failure of a Grouted Connection 23

Figure 4.7. Typical Failure of Connection Using a SteelNeoprene-Steel Sandwich 24

Figure 4.8. Failure of Column Stub Bearing on a Steel Plate

and Bedded in High Strength Gypsum 24

Figure 4.9. Failure of Column Stub Bearing on a Steel Plate

Covered by a Low Friction Membrane 24

Figure 4.10. Test Arrangement to Determine Load-Deformation

Characteristics for the Neoprene Pads 29

Figure 4.11. Relationship Between Compressive Load and Average

Compressive Deformation of Neoprene Pads 29

Figure 4.12. Relationship Between Compressive Stress and

Tangent Modulus of Elasticity for the FullSized Neoprene Pads 30

Figure 4.13. Relationship Between Compressive Load and

Maximum Extrusion for the Full-Sized Neoprene

Pads 30

Figure 4.14. Test on Specimen With Grouted Dowel 32

Figure 4.15. Instrumentation of Test Specimen With Grouted

Dowel 34
Figure 4.16. Specimen with Grouted Dowel Vertical Deformation

of an 11-in Segment Including the Joint 35

Figure 4.17. Specimen with Grouted Dowel, Longitudinal Dowel

Strain in the Joint 35

Figure 4.18. Specimen with Grouted Dowel, Horizontal Strain

of Concrete on South Column Face 36

Figure 4.19. Specimen with Grouted Dowel, Horizontal Strain

of Concrete on West Column Face 36

Figure 4.20. Details of Shear Test Specimens 41

Figure 4.21. Specimen Arranged for Pre-Loading and Grouting

of the Dowel 41

Figure 4.22. Arrangement for Shear Test 1 42

Figure 4.23. Arrangement for Shear Tests 2 and 3 42

Figure 4.24. Shear Test 3 Specimen After Test 45

Figure 4.25. Vertical Displacements of South End of Center

Segment in Shear Test 1 45

Figure 4.26. Vertical Displacement of South End of Center

Segment in Shear Test 2. 46

Figure 4.27. Vertical Displacement of North End of Center

Segment in Shear Test 3 47

Figure 4.28. Dowel Strain Caused by Axial Loading of the

Specimen in Shear Test 3. 48

Figure 4.29. Dowel Strain in Shear Test 3 Recorded by Strain

Gage #7 48

Figure 4.30. Net Joint Displacement in Shear Test 1 49

Figure 4.31. Net Joint Displacements in Shear Test 2 49

Figure 4.32. Net Joint Displacements in Shear Test 3 50

Figure 4.33. Comparison of Net Joint Displacements in Shear

Tests 1, 2 and 3 50

List of Tables

Table 4.1.1 Joint Materials Used in Evaluating Bearing

Capacity of the Columns 18

Table 4.1.2 Phase I Test Results on Bearing Capacity of

Column Connections 20

Table 4.1.3 Phase II Test Results on Bearing Capacity of

Column Connections 21

Table 4.1.4 Measured Compressive Deformation of Phase II Rubber Pad Joints at Various Compressive Loads 22

VI
Evaluation of the Column Connections

Used in a Precast Concrete Modular Housing System*

F. Y. Yokel and T. W. Reichard

The column connections used in a housing system employing stacked

precast concrete box modules were tested to evaluate their structural

performance. The system was proposed for construction in Operation Breakthrough,

a research and demonstration program sponsored by the Department of Housing

and Urban Development. The system uses innovative structural design concepts,

which include: confinement of the concrete in the vicinity of the column

bearings by reinforcing ties in order to increase concrete compressive

strength; neoprene pads between column bearings in the upper stories,

stories; steel-neoprene-steel sandwich in the lower stories; and a grouted

dowel through the center of the columns to provide resistance to tension

and shear.

The test program included the following: tests to determine the

effect of various bearing pads on the load capacity of the connection;

tests to determine the load-deformation characteristics of the neoprene

pads; a test to determine the performance of a lower-story connection

using a steel-neoprene-steel sandwich and a grouted dowel; and tests to

evaluate the strength and ductility of the connections when subjected

to a shear force. The test results are presented and interpreted and

the findings are summarized.

Key Words : Building system; column connection: concrete triaxial

strength; ductility; neoprene bearing pad; performance test; precast concrete;

structural design; Operation Breakthrough.

1. Introduction

The introduction of industrialized building systems may lead to the

development of untried structural details which cannot be evaluated by

analysis based on our present design standards. V.'henever deviations from

design standards are proposed, the structural adequacy of the system must

be determined by performance evaluation.

This report presents the results of the performance evaluation of

innovative column connection details in a multi-story housing system employing

stacked precast reinforced concrete box modules which was one of the systems

proposed for construction under the Operation Breakthrough program, sponsored

by the Department of Housing and Urban Development.

* Report prepared for Office of Research and Technology, Department of

Housing and Urban Development, U'asington, D. C. 20410.
2. The Building System

2.1 Description o f System

The proposed system consisted of precast reinforced light-weight

concrete box modules, stacked in a checkerboard pattern. Figure 2.1 shows

one of the buildings proposed for construction, a 19-story structure.

Figure 2.2 shows the model of a complex of buildings which is under construction

using the structural concepts and connection details evaluated in this

report, however, the modules differ in size and configuration from those

proposed , for the structure shown in figure 2.1.

A sketch of a proposed module is shown in figure 2.3. The outside

dimensions of the module are 44 ft x 14 ft x 8 ft-11 in. The concrete

used in the modules consists of a regular-weight sand and light-weight

coarse aggregate mix with a unit weight of 120 lb/ft and a 28-day strength

2 ranging from 4000 psi (lb per in ) to 5000 psi as specified in the design.

An individual module of this type weighs 81.3 kip (kil'opounds) . Some of

the boxes in the proposed system have balconies cantilevering from one,

or both ends. A box with balconies on both ends, which is the largest

module used in the system, is 52 ft long and weighs 92.3 kip.

The structural skeleton of the modules is shown schematically in

figure 2.4. A module contains four bents (ribs) which are rigid frames

consisting of 12 x 15 1/4-in— columns and 12 x 12-in— beams near the

module ends, and 12 x 12-in— columns, with a 8-in thick slab section

providing a common beam to both ribs at the center of the module. The

bents are connected by a 4-in thick top slab, 3-in thick sidewalls, and

a 4 in thick bottom slab. The bents, top slab, and sidewalls are cast

monolithically in a single operation using a steel form creating a "hull".

The bottom slab, which is cast in a second casting operation is attached

to the hull by bent dowels protruding from the bottom of the sidewalls

and by special steel inserts welded to the four center columns. The modules

are erected in a checkerboard pattern as shown in figure 2,5. Part of

the living space is contained within the modules and part is space created

between adjacent modules. The space within modules provides a clear height

of 8 ft-3 in and a clear width between concrete wall surfaces of 12 ft6 1/2 in. The created space has the same clear height and a clear width

of 12 ft-11 1/2 in between the concrete surfaces of the sidewalls. The

structure is completed as shown in figure 2.6 by inserting special end

wall, roof, and floor panels to complete enclosure of the created spaces.

— Since column and beam sections have flared sides (they are not exactly

rectangular or square) , average dimensions are given which represent

equivalent rectangles or squares.
Figure 2.1. Proposed Building

%J3

Figure 2.2. Building Complex Under Construction
Figure 2.3. Proposed Modular Unit

Figure 2.4. Structural Skeleton of Module
Figure 2.5. Checkerboard Arrangement of Modules

Figure 2.6. Completion of Structural Assembly
The method of connecting modules is illustrated schematically in

figure 2.7. The entire load is transmitted through the columns. Transmission

of load between the sidewalls in consecutive modules is prevented by suitable

structural detailing. Anchorage through the column connections is provided

by grouted dowels consisting of consecutive lengths of #8 (1 in diameter)

deformed steel reinforcing bars, with a 60 ksi minimum yield strength,

extending from midheight of one module to the midheight of the module

above

.

Details of the proposed column to column connection are shown in

figures 2.8 and 2.9. Figure 2.8 shows the detail of an interior column

joint. The heavy lines show the column and part of the floor slab of

Module A, and column and part of the 8-in thick slab connecting the interior

column bent in Module B. The dashed lines show the location of the roof

and wall slabs. Note that the bottom of the floor slab extends 2 3/8

in below the bottom of the column and that the column connection is located

in a 2 3/8-in deep recessed pocket. The top of the ceiling in Module

B is thereby located approximately at the same elevation as the top of

the slab in Module A, thus eliminating the need for a stepped transition.

It can also be seen that the wall of Module A does not bear on the wall

of Module B. Thus all compressive and shear forces are transmitted through

the bearing pad in the column connection. The joint between the boxes

at locations other than the column connection contains a compressible

filler material. Proposed connection details are shown in figure 2,9.

Figure 2.9(a) shows the reinforcement near the column end where additional

confinement ties are provided in the vicinity of the joint. Figure 2.9(b)

shows a section through a connection.

Erection of the system proceeds in the following manner. Modules

are precast in a plant, not too distant from the erection site, and transported

to the erection site on specially-designed trailers. A 50-ton crane unloads

the modules and lifts them to their desired position. Tolerances in fit

to insure proper load distribution between the eight bearings are within

limits that can be accommodated by the compressible bearing pads. Leveling

during erection is achieved by inserting steel shims to bring all bearing

pads to a level position before a module is placed. At the end wall of

the building grouting proceeds continuously during construction to prevent

instability of the sidewall panels enclosing the created spaces. All

interior joints are grouted only after completion of the erection sequence.

2. 2 System Features That Required Performance Evaluation

The proposed system differs in several ways from conventional castin-place reinforced concrete structures: (1) Reinforcement details differ

from those normally used in column-beam connections. The unconventional
Figure 2. 7 . Schematic Drawing of Connection Tiedown

WALL BOX A

COLUMN (BOX A)

BEARING PAD BOTTOM SLAB

COLUMN (BOX B)

WALL

Figure 2.8. Detail of Interior Column Joint
COLUMN TIES

MAIN REINFORCEMENT

(a) COLUMN REINFORCEMENT AT

CONNECTION

GROUT

STEEL DOWEL

GROUT CHANNEL

NEOPRENE PAD OR

NEOPRENE STEEL SANDWICH

(b) CONNECTION DETAIL

Figure 2.9. Proposed Connection Detail
details were required because of discontinuities between modules in the

structure, the thin concrete sections and the rather heavy reinforcement,

together with the need for an arrangement of reinforcing bars that would

not impede efficiency in the mass production of modules, (2) All the

vertical and horizontal forces are transmitted across the column connections

which were planned to be "dry" connections to facilitate cold-weather

construction and which include compressible bearing pads to accommodate

construction tolerances.

The specific objective of this evaluation was to determine the load

capacity of the column connection which had several unique details. The

various parameters considered in the evaluation of these details are discussed

below:

1. Except for the No. 8 dowel, vertical column reinforcement is

interrupted at the connection. Thus, the forces resisted by columns containing

up to 3 percent vertical reinforcement, are transmitted through a plain

concrete bearing with a single grouted dowel in its center. The maximum

2/ load permitted in such a bearing by ACI-318.71 [1]— would be exceeded

in the bearing. It has been shown [2,3,4,5] that the compressive strength

of concrete under triaxial confinement is greater than the unconfined

compressive strength and the design therefore provides for additional

ties in the vicinity of the joint as shown in figure 2.9(a). To permit

a bearing stress in excess of that permitted by ACI-318 the load capacity

of bearings was evaluated by testing.

2. In order to accommodate erection and fabrication tolerances,

a 1/4-in thick neoprene pad is inserted between the columns. Available

data on neoprene bridge bearings [6] are limited to stress levels less

than 1000 psi, which are substantially lower than the 6000 psi stress

developed in the proposed bearing at the failure load and are based on tests of neoprene pads bonded to steel plates. At the stress levels expected

to occur when the required load capacity is reached the modulus of elasticity

of neoprene is lower than that of concrete, and Poisson's Ratio is higher.

The neoprene, therefore, exerts a splitting force. on the concrete column

faces, reducing the load capacity of the bearing [7]. Thus while the concrete

capacity is increased by the confinement ties, as previously discussed,

the beneficial effect of the confinement may be offset by the splitting

force exerted at the concrete-neoprene interface. This effect also had

to be evaluated by testing.

3. Since it was recognized that the neoprene would reduce the capacity

of the bearing, another type of bearing pad was proposed for the more

2/— Numbers in brackets designate literature references
heavily loaded bearings. This pad consists of a steel -neoprene-steel

sandwich, using + 1/8-in thick (11 ga.) stainless steel plates and a 1/4-

in thick neoprene pad. In this case it was assumed that the steel plate

would act to resist the radially tangential shear exerted by the neoprene

and to provide additional triaxial confinement at the steel-neoprene interface.

This type of bearing also had to be evaluated by testing.

4. In addition to the information on the load capacity of the connection

in compression, the deformation characteristics of the connection under

compressive load had to be determined in order to compute the added horizontal

drift under wind load caused by the compressible bearings.

5. The load capacity of the joints containing the grouted dowel

had to be determined. The sequence of grouting in the construction had

to be considered in this determination. As previously noted, the dowels

in the columns near the side (gable) walls of the building are continuously

grouted during construction (whenever a story is complete, the dowel is

grouted) . This procedure differs from the grouting procedure proposed

for the other columns which is to grout in a single lift after completion

of the erection. As a result of the continuous grouting, most of the

load in the sidewall connections would be transfered to the dowel and

surrounding grout, rather than to the compressible pad. This may cause crushing of the grout and yielding of the dowel in the lower stories of

the building. These effects were not judged to be detrimental, since

they would be confined to the vicinity of the compressible pad which is

only 1/4 in thick in the uncompressed state. However, this loading condition

had to be simulated in the performance evaluation.

6. In addition to the load capacity in compression, the shear capacity

of the connection had to be evaluated. The shear capacity is critical

in the resistance of the building to wind and earthquake load and for

the specific buildings under consideration, shear capacity was critical

in the prevention of progressive collapse under abnormal loads. In the

determination of shear capacity the previously-discussed effect of continuous

grouting had to be simulated. In view of the use of compressible pads,

displacements across the joint (slippage) had to be evaluated in addition

to shear capacity. Not only the capacity of the joint in shear, but also

the ductility of the joint was evaluated. Ductile behavior would prevent

the premature failure of a single joint in a group of joints, thus increasing

the load capacity of the building, and would also be an important consideration

in determining the resistance to seismic loading. Special consideration

was given to load reversals that may be caused by seismic events.

10
3. Scope

The results and conclusions from four different test programs are presented in this report. A slightly scaled-down column cross section

(10 in x 12 in) was used in order not to exceed the capacity of the testing

equipment.

(1) Tests to evaluate the bearing capacity of column

connections using various types of bearing pads

(specimens did not have grouted dowels)

.

This program had two phases: In Phase I, two specimens

consisting of two 10 x 12 x 31 1/2-in long column stubs

with 1/4-in thick neoprene pads were tested in compression.

Subsequently, one of these stubs which did not fail in the

initial test was bedded in high-strength plaster on a steel

bearing plate and retested in compression.

In Phase II nine specimens, consisting of two 10 x 12

x 51 1/2-in long columns, were tested in compression. Eight

specimens with various proposed types of bearing material

inserted between the two stub columns and one specimen

using a steel -neoprene-steel sandwich with low-friction

membranes inserted between the neoprene and the steel

plates

.

Subsequently, one of the stub columns which did not fail in

the initial test was tested bearing on a steel plate with

a low-friction membrane inserted between the column

face and the steel plate.

(2) Tests to determine the stress -strain characteristics

of neoprene pads.

In this test three specimens consisting of two 1/4-in thick

neoprene pads each were tested in compression between steel

plates and their load-deformation characteristics were recorded up to a stress level of 4000 psi.

(3) Tests to evaluate the bearing capacity of connections using a steel -neoprene-steel sandwich as

a bearing pad and a grouted dowel.

11
In this test one pair of 10 x 12 x 51 1/2-in long

column stubs, using the connection details planned for

the lower stories, was tested in compression using the most

critical preloading condition before grouting of the

dowel (simulating continuous grouting).

(4) Tests to evaluate the shear capacity of the column

connection.

Under this program three tests were conducted on threesegment specimens, using If) x 12 x 51 1/2-in long columns

segments with the proposed connection detail between them.

Specimens were tested by applying a shear force to push

out the middle segment.

Two specimens were tested in shear while subjected to

low axial load, simulating upper-story conditions.

One of these was tested in one direction while the other

was subjected to several reverse cycles of loading.

The third specimen was tested in shear while subjected

to high axial load, simulating lower-story conditions.

This specimen was subjected to several reverse cycles of

loading

.

4. Tests

4. 1 Bearing Capacity of Column Connections Using Various Types of

Bearing Pads

4.1.1 Objective

The problems associated with the load transmission from column to

column are discussed in Section 2.2. The objective of this particular

test program was to determine the failure load of bearings using additional

column ties in the vicinity of the joint in order to provide triaxial

confinement for the concrete and using various types of bearing pads proposed

or considered as alternates. This test was necessary, since:

(1) the design concrete stress at the bearing surface, used in the

lower-story joint, exceeded that permitted in ACI-318;

(2) the effect of the radially tangential shear force exerted by

the neoprene on a neoprene-concrete interface could not be determined

by analysis.

12
(3) there was a need to substantiate the effect of the steel plates

in the proposed steel-neoprene-steel sandwich in triaxially confining

the concrete and resisting the radially tangential shear force exerted

by the neoprene.

The tests were conducted during the design stage of the system and

served as exploratory (pilot) tests. The grouted dowel was omitted in

these tests to avoid introduction of an added variable. After sufficient

preliminary information was available from these tests, another test was conducted on a connection which included the grouted dowel.

4.1.2 Test Specimens

The typical specimen for these tests consisted of two column stubs

placed end to end with the joint materials separating the two concrete

bearing surfaces. For two of the tests single column stubs were used

as the specimen.

Two types of stub columns were used. The first type, 10 in x 12

in x 31 1/2 in high, was used in the series of tests identified as Phase

I tests. The second type, 10 in x 12 in x 51 1/2 in high, was used for

the Phase II tests. The Phase I stub column is shown in figure 4.1 and

the Phase II column in figure 4.2. The main reinforcement was welded

to a 1 1/2-in thick steel plate at one end. Load was applied through

the steel plate and the welded reinforcement connection insured that the

load would be transferred simultaneously to the concrete and the main

reinforcement. The Phase II columns exclusive of the steel bearing plate

were 1/2 the length of the column in the proposed structure.

Typical details of the test specimens are shown in figure 4.3. Main

vertical column reinforcement consisted of deformed steel bars meeting

ASTM designation A615-68, Grade 60 [8] and had an average measured yield

strength of 67.4 ksi (tensile tests were conducted in the laboratory).

Ties were deformed bars meeting ASTM designation A615-68, Grade 40 and

had an average measured yield strength of 51 ksi.

The welded wire mesh cage shown in the figure corresponds to the

detail in the system used to prevent concrete breakage caused by form

stripping. The welded wire mesh had an average yield strength of 63 ksi

according to information obtained from the manufacturer.

The concrete in all specimens was a 4000 psi mix using sanded lightweight

aggregate of 3/4 in maximum size. Figure 4.4 shows the age-strength

relationship for these concretes.

The types of jointing materials used in these tests are described

in table 4.1.1.

13
31-1/2 in

NO. 9 BARS

CONFINEMENT ^TIES

NO. 3 BARS

1-1/2 in STEEL PLATE

Figure 4.1. Specimen Used in Phase I Column Connection Tests

14
NO. 9 BARS

NO. 3 BARS

Figure 4.2. Specimen Used in Phase II Column Connection Tests

15
SECTION AA

10

II 1/2

Notes Diameter of *3 Bar is 0.375 in. Diameter of fl 9 Bar is 1.125 in. Diameter of WO. 5 wire is 0.080 in.

SECTION BB

3/4" COVER FW"?

9 o a

6-* 9 BARS

Figure 4.3. Typical Details of Test Specimens

16
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I l 12 16 AGE IN DAYS

20 24

Figure 4.4. Strength of Age Relationship of Concrete Used for Column Stubs

17
TABLE 4.1.1 Joint Materials Used in Evaluating Bearing Capacity

of the Columns

.

Type Material Used in Joint When Used

1 1/4-in thick, 70 durometer (Shore A) neoprene

rubber pad, either 10 in x 12 in or 9 1/2 in x 11 1/2 in

Phase I, Test 1, and

2, Phase II, Test 1

2 1-in thick steel plate bonded to the concrete

with high strength gypsum plaster.

Phase I, Test 3

3 3/4- in thick non-shrink grout, made of Type

III portland cement, silica sand and special

aggregate. Mix: 1 lb of water to 5 lb of

dry ingredients

Phase II, Test 2 and 3

4 Two - 11 ga. (0.12 in) type 304 stainless

steel plates (10 in x 12 in) with neoprene

pad (Type 1) between the plates.

Phase II, Test 4,

5 , an d 6

5 Same as Type 4 except that the plates were

16 ga. (0.06 in) stainless steel

Phase II, Test 8

6 1/4 in-thick 90 durometer (Shore A)

rubber reinforced with many layers of

8 oz cotton duck.

Phase II, Test 7

7 Same as Type 5 with a friction breaker

of wax paper between neoprene and plates

Phase II, Test 9

8 Concrete bearing surface capped to a flat,

smooth surface with a thin layer of high

strength gypsum plaster and set on a friction breaker of 2 sheets of 6 mil

plastic film placed between the capped

surface and a steel bearing block.

Phase II, Test 10

18
4.1.3 Test Procedure

All specimens were loaded in a 600 kip hydraulic testing machine.

The specimens were preloaded to 200 kip (1670 psi) unloaded and subsequently

loaded at 75 kip (62.5 psi) per minute until failure. The failure was

not sudden, in that considerable splitting and spalling occurred prior

to the decrease in load resistance. Application of load was continued

past the point of maximum resistance until the load fell off to 80 - 90

percent of the maximum.

The method of load application is shown in figure 4.5 The specimen

was positioned concentrically with respect to the applied load, with its

base set on the testing machine platen in a bed of high strength gypsum

plaster. The load was applied by the spherically-seated head of the testing

machine which was fixed against rotation after applying a compressive

load of 1 kip. A 1/2-in thick fiberboard plate was inserted between the

steel plate at the surface of the specimen and the head of the testing

machine. This setup resulted in approximately concentric loading.

In the Phase II tests the vertical deformation of the joint material

was monitored by means of a 0.001-in dial gage. Lateral deformations were measured over a 9-in gage length using linear variable differential

transducers. These deformations were monitored by X - Y recorders with

a resolution of 0.001 in to aid in the determination of the onset of spallinj

(see figure 4.1).

4.1.4 Test Results

The results for all the tests are presented in tables 4.1.2, 4.1.3

and 4.1.4. Typical failure modes are illustrated in figures 4.5 through

4.9.

4.1.5 Discussion of Test Results

In tables 4.1.2 and 4.1.3 the results are normalized by dividing

the ultimate load by 0.85 f'A. where:

c

'

f* is the average concrete compressive strength of

6 x 12 -in cylinders taken during the casting of the

specimens and tested at the time the specimens were

tested.

A is the cross sectional area of the column.

0.85 f'A is the failure load of a plain concrete bearing

with no reinforcement ties to provide confinement, computed

in accordance with ACI 318-71. In the case of Test 6

19
Table 4.1.2 Phase I Test Results on Bearing Capacity of Column Connections

Stub Column Size Test

Number

£- »/

cpsi

Type of Joint Max. Load

kip

P s/

u 0.85f A

c

10 x 12 x 31 1/2

inch

I?

1

2

3

3660

3660

3660

1/4 x 10 x 12 inch

neoprene

1/4 x 10 x 12 inch

neoprene

Single stub set in

plaster on 1 inch

thick plate

350

354

585

0.94

0.95

1.56

a/ Symbols: f<L = Concrete cylinder (6" x 12") strength at time of test.

P = Maximum load

A = Bearing area = 120 sq. in.

20
Table 4.1.3 Phase II Test Results on Bearing Capacity of Column Connections

(Column Stub Size 10 x 12 x 51 1/2 in)

Test

Number

fI a/

c psi

Type of Joint

Max.

Load

Kip

pi/

u 0.85 A

1 4340 1/4 x 10 x 12 in neoprene 477 1.08

2 4160 3/4 thick grout^/ 535 1.26

3 4510 3/4 thick grout£/ 519 1.12

4 4155 2-St.St. plates and 1/4 x

10 x 12 in neoprene

580 1.37

5 4342 2-St.St. plates and 1/4 x

9 x 11 in neoprene

586 1.32

62/ 4881 2-St.St. plates and 1/4 x

9 x 11 in neoprene

420®/ 1.262/

7 4590 1/4 x 10 x 12 in reinforced

rubber

493^ 1.05

8 4881 2-St.St. plates and 1/4 x 9 x 11 in neoprene

556 1.11

9 4881 2-St.St. plates, wax paper

and 1/4 x 9 x 11 in

neoprene

598 1.20

10 4881 1 stub set in plaster + 2 sheets of plastic film

between plaster and steel

600^ 1.20

bearing plate

a/— Symbols: f' = Concrete cylinder strength at time of test

P^ = Maximum Load

A = Bearing area = 120 in

Grout compressive strength (2 in cube) at time of test = 5210 psi

6200 psi

Smaller size (9 x 11 in) neoprene pad filled bearing area at failure

Smaller size was selected to prevent unsightly extrusion from joint.

Loaded with 2-in, major-axis eccentricity at the joint. Bearing area, A, was computed using the Whitney stress block concept.

g/ Held 600 kips for 5 min. before failure initiated,

21
a/ Table 4.1.4 Measured Compressive Deformation of Phase II—

Rubber Pad Joints at Various Compressive Loads

Test Number Type of Joint Deformation at Stated Load

100 kip

in

200 kip

in

300 Kip

in

400 Kip

in

500 Kip

in

1 Neoprene .034 .037 .041

4 11 ga. plate and

neoprene

.034 .042 .049 .057 .068

5

it .029 .036 .041 .049 .062

7 Reinf. rubber .026 .033 .041 .045

8 16 ga. plate and

neoprene

.022 .038 .062 .088 .118

9 16 ga. plate, wax

paper § neoprene

.014^/ .024 .046 .068 .086

Test Results

Section 4.3

Neoprene and steel .027 .040 .052 .060

a/—All specimens were preloaded to about 200 kip prior to test run, Deformations do not include "permanent set" from preload.

b/ During preloading to 200 kip pad was compressed approximately

0.10 in. Recovered only to 0.09 in. Initial compression was significantly lower and recovery greater for others.

22
Figure 4.5. Tyoical Failure of Connection Using 1/4-in Neoprene Pad

^*S

//i

in —

; 1'

*•

•*

' 9

1 ^ -

^1 < -.4 1 v1

j < £< .; Y i

" 9*'.

II

Figure 4.6. Typical Failure of a Grouted Connection

23
Figure 4.7 Typical Failure of Connection

Using a Steel-Neoprene-Steel

Sandwich

Figure 4.8. Failure of Column Stub

Bearing on a Steel Plate

and Bedded in High Strength

Gypsum

Figure 4.9. Failure of Column Stub

Bearing on a Steel Plate

Covered by a Low Friction

Membrane

24
2 of Phase II of this testing program, A was taken as 80 in

accounting for the 2-in eccentricity from the major axis.

By normalizing the loads in this manner a strength coefficient (k)

is computed which represents the ratio of the strength of the test specimens

to the strength of a plain concrete bearing with no confinement and no splitting effects exerted by neoprene pads. Examination of the strength

coefficients in the last columns of Tables 4.1 and 4.2 leads to the following

observations

(1) In spite of the confining reinforcement, the average strength

of the plain neoprene bearing tended to be less than that estimated for

an unconfined concrete bearing (Tests 1 and 2, Phase I; Test 1, Phase

II).

(2) The strength of the connection using the steel-neoprene-steel

sandwich with the 11 ga. (gage) steel plate tended to exceed the strength

estimated for a plain concrete bearing by approximately 30 percent (Phase

II, Tests 4, 5, and 6). With the 16 ga. steel plates the strength was reduced to 11 percent over that estimated for a plain concrete bearing

(Phase II, Test 8); however, when low-friction membranes were inserted

between the neoprene and the steel plates, the strength increased to 20

percent overstrength (Phase II, Test 9).

(3) The two grouted specimens exceeded the estimated strength of

an unconfined plain concrete bearing by 12 to 26 percent (Phase II, Tests

2 and 3)

.

(4) The column stub with the low-friction membrane exceeded the

estimated strength of an unconfined plain concrete bearing by 20 percent,

while the stub set in high-strength plaster on a 1-in steel plate exceeded

the estimated strength by 56 percent.

Note that in Tests 5, 6, 8, and 9 the size of the neoprene pad was smaller than that of the bearing area. This size reduction was to prevent

excessive extrusion of the neoprene from the sides of the bearing. Since

under pressure the lateral dimensions of the neoprene pad increased and

extended over the entire bearing area, there was no apparent adverse effect

from the decrease in the size of the neoprene pad.

It can be concluded from these observations that the radially tangential

force exerted on the concrete bearing surface has a major effect on the

capacity of the joint. The specimens ranged from those where the joint

exerts a splitting force on the surface to those where the joint is capable

of developing a significant confining force.

25
At one extreme is the neoprene bearing pad which exerts a splitting

force on the concrete counteracting the confining effects of the reinforcement

ties

.

The specimens bearing on low friction surfaces, as well as the grouted

specimens probably did not develop significant radially tangential shear

forces on the concrete bearing surfaces (in the latter case Poisson's

Ratio and the Modulus of Elasticity were similar for concrete and grout).

Thus, the k-coeff icient for these specimens which averages approximately

1.2, probably represents the effect of the confinement provided by the

reinforcement ties.

The steel-neoprene-steel sandwich using the 11-ga. steel plates,

where the k-coeff icient averaged 1.32 exerted a confining force which

prevented concrete splitting and was transmitted by the friction between

the concrete and the steel plate and resisted by the steel plate. The

steel plate also resisted the additional splitting force exerted by the

neoprene pad. The 16-ga. steel plates which yielded during the test did

not have sufficient strength to resist the splitting force.

The greatest confining force was provided by the steel plate set

in the plaster where a confining force greater than that provided by the

frictional resistance of the steel/concrete interface could be developed.

This additional confining force increased the k-coeff icient to 1.56.

The previous conclusions are based on both Phase I and Phase II testing.

Originally the longer column stubs used in Phase II testing were used

because there was concern that the short length of the Phase I columns

might be insufficient to develop the bond between the concrete and the

main column reinforcement. The tests, however, do not provide evidence

that the change in column length had a significant effect on the mode

of failure or on the load capacity. The conclusions are also based on

the assumption that, except for Test 6 in Phase II, the applied load was approximately concentric.

The effect of the type of bearing pad on the mode of failure can

be observed in figures 4.5 to 4.9.

Figure 4.5 shows a typical failure of a connection using a neoprene

pad. The spalling between the core inside the ties and the shell outside

the ties is an indication that at ultimate load the core supported most

of the load. The same observation can be made in figure 4.6 which shows

the typical failure of a grouted connection.

On the other hand, it can be seen from figure 4.7 that in the specimens

using the steel-neoprene-steel sandwich there was no spalling of the shell,

and thus, the entire cross -sectional area contributed to the load resistance.

It can also be seen in this figure that the concrete bearing surface is

26
smooth and locally deformed with protruding imprints of the main reinforcing

bars. The deformation is probably attributable to plastic flow of the

concrete before failure which is taken as an indication of triaxial confinement

The imprint of the bars was caused by bond breakage near the end of the

bars where the strain in the concrete could not be transmitted to the

steel bars.

The failure of the column stub that was bedded in high-strength gypsum

plaster, which is shown in figure 4.8, occurred in the middle of the column

above the region of the additional confining ties. This is taken as an indication that in this case the effect of the confinement was even greater

than that provided by the steel -neoprene-steel sandwich. On the other

hand, the failure mode of the column stub which was tested with a lowfriction membrane at the column face and is shown in figure 4.9, gives

clear evidence of spalling of the shell.

Thus, in summary, it is concluded that the radially tangential shear

force exerted on the concrete face at the interface with the bearing pad

had a significant effect on the bearing capacity of the column connection.

The coefficients shown in the last column of Tables 4.1.2 and 4.1.3 give an indication of the magnitude of this effect.

4 . 2 Tests to Determine the Load-Deformation Characteristics of the

Neoprene Pads

4.2.1 Objective

The modulus of elasticity of neoprene generally tends to increase

as compressive stresses in the neoprene are increased [6] . The loaddeformation characteristics of a neoprene bearing pad depend on its shape

and on the properties of interfacing materials.

The objective of these tests was to determine the load-deformation

characteristics of the proposed neoprene pads. This information could

be used in improved analytical models for the prediction of joint deformation

characteristics and for the assessments of the effect of dimensional

inaccuracies on the distribution of loads between columns. A secondary

objective was to arrive at an estimate of the magnitude of the lateral

expansion of neoprene bearing pads when subjected to compressive load.

4.2.2 Test Specimen

All specimens were 1/4 in thick, and made of 70 durometer neoprene

pad materials. Four 9 x 11 x 1/4 in specimens were used in two tests.

Each of these four full-sized specimens was tested with a 2-in diameter

hole in the center. In an additional test two half-sized specimens (9 x 5 1/2 x 1/4 in) were tested without center holes.

27
4.2.3 Test Procedure

All tests were made in a hydraulic testing machine by loading the

specimens through a 2 1/2-in thick bearing block. In each test, two pads were sandwiched between three 2 1/2-in thick steel blocks as shown in

figure 4.10. Thus, two pads were tested in each test. The size of the

steel blocks used was as shown in figure 4.10. The bearing surfaces of

the steel blocks were ground flat in a tub-grinder. Deformation measurements

were made on the four sides of each pad using dial gages with a 0.001-

in resolution and were referenced to the platen of the testing machine.

Compressive deformations were measured in Tests 1 and 2 and lateral expansion

deformations were measured in Test 3. Full-sized pads were used for Tests

1 and 3 and the half-sized pads were used for Test 2. Prior to each test the specimens were preloaded to the maximum test

load for 10 minutes in order to reduce creep effects during the test.

The specimens were tested immediately after removal of the preload and

loaded at a slow average rate because of the interruptions necessary for

dial gage reading. Gage readings were taken for each load increment after

a compression was reached that could be maintained in a stable position

while the readings were taken.

4.2.4 Test Results and Findings

Figure 4.11 presents the data from Tests 1 and 2. The deformations

are the average for the eight measurements on the two specimens in each

test. The effect of the shape factor of the pad* on the compressive deformation

is quite apparent from this figure. For load levels above 100 kip the

deformation of the full-sized pad is only 58 percent of that for the halfsized pad. It is also apparent from figure 4.11 that the modulus of elasticity

for these pads increases with increasing stress.

Figure 4.12 is a plot of compressive stress (based on the 9 x 11-

in area) versus tangent modulus of elasticity for the full-sized pad.

This plot shows that the tangent modulus at 4,000 psi is over 6 times

the initial tangent modulus.

In all of the tests the lateral expansion of the pad was non-uniform;

i.e. the pad was extruded from the steel blocks significantly less at

the corners than at mid-length and width. At high stress levels the pad

assumed a roughly elliptical shape. In Test 3 an attempt was made to measure the lateral extrusion of the full-sized pad using dial gages.

Because of the out of plane distortion of the pads the measurements only

approximately convey the magnitude of this effect. Figure 4.13 shows

the extrusions measured for various load levels. The extrusions plotted

are the arithmetic means of 8 measurements made on 2 specimens. The measurements

were made at the center of each side so that the extrusions are maximum

values

.

28
LOAD LOAD

STEEL BLOCKS

NEOPRENE PADS

TESTING MACHINE PLATEN

ALL TESTS

END VIEW

2 1/2

FULL-SIZE PAD

SIDE VIEW

Figure 4.10. Test Arrangement to Determine Load-Deformation Characteristics

for the Neoprene Pads

400

9" x 11" x 1/4" © TEST NO. I 9" x 5 1/2" x 1/4" H TEST NO 2

777777777777

02 04 06 08 10 .12

AVERAGE COMPRESSIVE DEFORMATION (IN 1/4 in), in

Figure 4.11. Relationship Between Compressive Load and Average Compressive

Deformation of Neoprene Pads

29
5000

1 J 1 LUii

1 r~

y>

9 XII X 1/4 in neoprene pad-^—

Test No. 1

>iii nil 1 /

1

IIIIIIITTT, 4000

'toQ.

</> {2 3000

orH

V)

- %/ -

z

JE 2000 <

111 CD

— -

1000

n -L^l 1 1 1 i

Figure 4.12.

10000 20000 30000 TANGENT MODULUS OF ELASTICITY, psi

Relationship Between Compressive Stress and Tangent Modulus

of Elasticity for the Full-Sized Neoprene Pads

500 Test No 3

O Average from 8 measurements

on two 9" X ll" pads.

0.2 4 0.6 0.8 1.0 AVERAGE LATERAL EXPANSION (2H),in

1.2

Figure 4.13. Relationship Between Compressive Load and Maximum Extrusion

for the Full-Sized Neoprene Pads

30
As stated above, all test results were obtained after applying a preloading cycle immediately before the load application for which deformations

are recorded. Since the visco-elastic properties of the material may

delay load recovery it is reasonable to expect somewhat greater deformations

in the initial load cycle. Thus the data presented would more accurately

predict load-deformation relationships for interior column connections

which are preloaded before grouting of the dowel.

In Table 4.1.4 data obtained in this test are compared with measurements

of compressive deformations of pads in Phase II of the bearing tests on column connections. Even though in the Phase II tests preloading was only 200 kip, there is reasonable agreement between the test results discussed

herein and the measurements in the Phase II tests.

4. 3 Tests to Evaluate the Bearing Capacity of Connections Using a Steel-Neoprene-Steel Sandwich as a Bearing Pad and a Grouted Dowel

4.3.1 Objective

The objective of this test was to evaluate the capacity of the connection

as proposed for the lower story containing the grouted dowel and the steelneoprene-steel sandwich. The most critical condition with respect to

the dowel was simulated, namely, continuous grouting. This was achieved

by preloading with an axial load of 10 kip before grouting which simulated

the weight of a single story. This preloading condition could cause the

dowel to yield before occurrance of a compression failure in the concrete.

The effect of such premature yielding, together with crushing of the

grout and bond breakage between the dowel and the grout in the vicinity

of the neoprene pad, could not be predicted.

4.3.2 Test Specimen

The test specimen shown in figure 4.14, consisted of two 51 1/2-

in high stub columns similar to those described in section 4.1.1. However,

these stub columns, as shown in the figure, had five confinement ties

instead of the three used previously. This change was made because of

a design change in the proposed system. In addition, a 2-in round flexible

conduit was cast in the center of each stub. The hole formed in the column

by the conduit was for the #8 deformed reinforcing bar (ASTM A615-68,

Grade 60 Steel) used as the dowel.

The two stubs were placed end to end with the steel-neoprene-steel

sandwich joint separating the two concrete bearing surfaces. The length

of the dowel inserted through the holes in the two stubs was slightly

longer than the overall length of the two stubs (103 in). The six externally

threaded 1-in diameter steel bars, which can be seen in figure 4.14, were used to apply a 10 kip axial preload to the specimen before grouting and

to sustain the preload until testing.

31
Figure 4.14. Test on Specimen With Grouted Dowel

32
The grout was type III portland cement and water mixed at a watercement ratio of 0.4. The concrete in the column stubs was as described

in section 4.1.2.

4.3.3 Test Procedure

The test was performed in a 600-kip hydraulic testing machine. Load

was applied in 25-kip increments and maintained after each increment for

electronic recording of measurements. Loading was continued to the 600

kip level. The 600-kip load (maximum testing machine capacity) was maintained

for 12 minutes. Then the specimen was unloaded, reloaded to 600 kip

in three 200-kip increments, maintained at 600 kip for two minutes and

unloaded.

The load was applied concentrically as shown in figure 4.14. The

steel plates at the ends of the specimens as well as the loading plates

had a 2-in hole in their center to prevent direct load transfer to the

steel dowel. The preloading bars were left on the specimen during the

test. Displacement transducers as shown in figure 4.15, were used to measure the deformation of the joint during the test. Bonded-foil strain

gages were used to measure the strain on two opposite sides of the dowel

in the center of the joint and also the strain in the concrete directly

over each of the five confinement ties on two adjacent faces of one column

stub (see figure 4.15).

4.3.4 Test Results

The test results are given below:

Maxium Load: 600 kip

Loading Condition: Concentric (figure 4.14)

Concrete Compressive Strength: (6 x 12 in cylinders) 3930 psi

Grout Compressive Strength: (2 x 2 in cubes) 11,000 psi

Condition After Testing: Minor cracking

Joint deformation data are shown in figure 4.16. Strain data for the

dowel and typical strain data for the concrete over the confinement ties

are shown on figures 4.17, 4.18 and 4.19.

4.3.5 Discussion of Test Results

(1) Load Capacity

It is noted in section 4.3.4 that the test load was limited by the

600-kip capacity of the testing machine. The maximum test load was maintained

for a 12-minute period and, after an unloading and reloading cycle, for

another 2-minute period. The specimen experienced some cracking as shown

33
Figure 4.15. Instrumentation of Test Specimen With Grouted Dowel

34
GAGE LENGTH- II i

< 300 —

D;0-A—<j>A- - - '"7 ^'-V f

ji y

crp' j£" i*i 1

' /' P''

,

°' S E'l /

/

*

* ay / iii', 1

' /' p'/ ,j /],. i // ' /

Py /'/' // / ' / pV ,

I'll

1 / /

1 ' i P?' Aj* /ttvi/ 16

j # y ///'l/ 1/ 1

I ,0/ /*

/'/;/ / /

i I P^jF <ii I i

i rip' jf i

-l*' /

I ? / ill \ //

I — <? Jf% N Mia 1

// ^-'7 LVDT

i 'if i/i i , / 4^y A

/ * /' /

' * / * /^ w °«

>

>Af

LVOT // /

>' '/

//,/ , / ii * , V SECT A A

t 1 1 y> • i * » i

1 /Xa< i ^ a i

i .10 .15 .20 .25 COMPRESSIVE DEFORMATION, in Figure 4.16. Specimen with Grouted Dowel Vertical Deformations of an 11-in Segment Including the Joint

600 P- -O-Og^Cl—•

*-f ?' ^7 I 1 t^ ' >' / // 500 ^ 400 / d '<</ V>'' /// // / ' / ' ill III III ,ii 1 / 1

i ' i III i i; III

300 p' ii< if/

,J*

\

i // / / /

vrr

' X

3 / '/' / / /

200 -

< X

X (9< / ^ //

/

/

13

<

tz

XX

X

<

a. ^

/// 100 -// « KX

i

/ z If

i / / if

II

1

1

il

i /

i / i

i l/l

J\ 1 \1 n * 1 A A

.0030 0045 .0060 COMPRESSIVE STRAIN

.0075 .0090

Figure 4.17. Specimen with Grouted Dowel, Longitudinal Dowel Strain in

the Joint

35
©

~ /rf- ' I /

/ "I

I l/l

I 'ii

I i '/ / J If STRAIN GA6E RUPTURED A T ill

I liJ

i ill

I i!/

it

<&

SOUTH FACE

.002 STRAIN

.004

Figure 4.18. Specimen with Grouted Dowel, Horizontal Strain of Concrete

on South Column Face

. 300

^

-1- -1-

//

r; STRAIN GAGE RUPTURED

i WEST FACE

002 STRAiN

Figure 4.19. Specimen with Grouted Dowel, Horizontal Strain of Concrete

on West Column Face

36
in figure 4.15, but did not lose its ability to support the 600-kip load.

On the basis of this evidence it may be assumed that the ultimate load

would have exceeded 600 kip.

Strain data from the gages on the column faces are plotted in figures

4.18 and 4.19. On the basis of the characteristics of the load-strain

relationship and the magnitude of strain it appears that significant cracking

3/ in the concrete occurred above the 350-kip load level. Very large apparentstrain developed at load levels above 500 kip. Three of the gages failed

at 475, 525 and 600 kip respectively, probably as a result of severe cracking;

gage 18, the closest to the joint, survived the entire load cycle. While

the 600-kip load was sustained this gage showed a substantial increase

in apparent strain (from 0.0059 to 0.0107). After removal of the load

the gage returned to a residual strain of 0.0082. A similar trend is

shown by the strain gage data on the other instrumented column face which

are shown in figure 4.19. In this case all the gages failed between

the 500 and 600 kip load levels; Thus, it can be concluded from the strain

gage readings that substantial cracking occurred above the 500-kip load

level and during the time the 600-kip load was maintained.

Figure 4.16 shows the length change in a 11-in vertical column segment

which includes the joint at its center. It can be seen in this figure

that substantial ,T creep" occurred while the 600-kip load was maintained.

As noted in the subsequent discussion, much of this length change was attributable to concrete deformation rather than joint deformation.

The previously discussed trends in measured deformations are taken

as an indication of incipient failure at the applied test load of 600

kip. Thus while the load capacity of the specimen exceeded the 600-

kip test load, it probably did not exceed that load by a substantial margin.

Using the concrete strength of 3930 psi, a dowel capacity at yield of

53 kip (based on the 67.4 psi yield strength), and a k-coeff icient of

1.32 for the confining effect of the steel-neoprene-steel sandwich (see

section 4.1.4, Table 4.1.3, average k-coeff icient for Tests 4, 5, and

6), a load capacity of 580 kip is predicted for this specimen.

It can be concluded on the basis of the previous discussion that

the load capacity of the specimen exceeded the capacity predicted on the

basis of the confining effect of the steel-neoprene-steel sandwich. It can also be concluded that yielding of the dowel, bond slippage, and local

grout crushing apparently did not adversely affect the load capacity.

— The term apparent is used since cracking was mostly responsible for

the observed deformation.

37
(2) Load Response Characteristics

Figure 4.16 shows a plot of load versus the compressive deformation

of an 11-in high segment of the specimen including the joint at its center.

The deformations were measured at four column faces and it can be seen

from the plot that there is reasonable consistency in the trend of these

four measurements.

A definite break in the load-deformation curve occurred between the

loads of 100 and 150 kip. Since the specimen was preloaded to 10 kip

before grouting it is assumed that the initial 10 kip load is supported

by the neoprene pad. Subsequently, however, because of the low neoprene

stiffness, most of the load was transferred to the steel dowel and the

2-in diameter grout column surrounding it. This mechanism of load transfer

had to eventually break down by breaking of the bond between the dowel

and the grout in the vicinity of the joint and by crushing of the grout

column surrounding the dowel in the region of the 1/4-in neoprene pad.

The break in the load-deformation curves is probably attributable to this

change in the load transfer mechanism. Dowel yielding is ruled out on the basis of the strain measurements in the dowel at the 100-kip load

level

.

At the 600-kip load level there was substantial additional deformation

during the period when the 600-kip load was sustained. Some of this deformation

may be attributable to creep of the neoprene pad; however, in view of

the large increase in transverse concrete strain shown in figures 4.18

and 4.19, it is reasonable to assume that most of this deformation was attributable to cracking of the concrete. This contention is supported

by the large residual deformation after removal of the load. This residual

deformation was approximately 0.21 in. The total deformation of the neoprene

pad, when subjected to a 550-kip load, (50 kip would be carried by the

dowel) is not expected to exceed 0.1 in (extrapolation from Table 4.19

and figure 4.11); and even though recovery of some of this deformation

may have been prevented by a permanent set in the dowel and by viscoelastic time lag, most of the 0.21-in residual deflection should be attributto the concrete compressive deformation.

Figure 4.17 shows the strain measured in the dowel over a 1/4-in

length centered in the center of the joint. Very little strain was recorded

during the first 20 kip of loading. This is in part attributable to the

10-kip preload which reduced the rate of load transfer to the column for

initial load increments. The subsequent relationship between strain and

compressive load leads to the conclusion that up to the 100-kip load level

the dowel transmitted only part of the load. At the 100-kip load level

the approximate load transmitted by the various components of the system

is estimated as follows on the basis of the data in figures 4.17

38
through 4.19: The average strain in the steel was 0.009 which corresponds

to a 22-kip load acting on the dowel. The neoprene pad is estimated

to have supported approximately 30 kip. This leaves a 48-kip load to

be supported by the 1/2-in higli. grout column. Such a load would be associated

with a 20,000 psi stress in the grout which probably could be sustained

because of triaxial confinement (cube strength of the grout was 11,000

psi). It is evident from figure 4.16 that some change occurred at the

100 kip level. Since dowel yielding is ruled out on the basis of the

average dowel strain this change was probably caused by crushing of the

grout and load transfer to the neoprene pad.

Between the load levels of 170 and 270 kip the dowel strain did not

increase. A possible explanation for this observation may be provided

by breakage of bond between the dowel and the surrounding grout in the

vicinity of the connection which caused any additional load to be transferred

to the neoprene pad. Dowel yielding may have contributed to this process,

however the strain level at which dowel yielding is expected was reached

at .the slightly higher load of approximately 300 kip. No large vertical

deformation was associated with dowel yielding since at the 300-kip load

level most of the load was carried by the neoprene. After unloading the

permanent set in the dowel corresponded to an average strain of 0.0048.

In the re-loading cycle the dowel yielded at approximately 400-kip and

the final permanent set in the dowel corresponded to an average strain

of 0.0055.

The conclusion that can be drawn from the preceding discussion is

that load transfer to the neoprene pads occurs at a load level of approximately

100 kip and is probably attributable to crushing of the grout column

and limited breakage of bond between the dowel and the grout; and that

dowel yielding occurs at approximately 300 kip and was not associated

with a significant decrease in the vertical stiffness of the joint.

4. 4 Test to Evaluate the Shear Capacity of Column Connections

4.4.1 Objective

Since all loads are transmitted through the column connection the

shear capacity of the connection may be critical with respect to the resistance

of the system to horizontal wind or seismic load. Because of the complex

nature of the connection, the relatively low resistance of the neoprene

pad to distortion under in-plane shear, and the effect of dowel yielding

during construction, testing was used to determine load-deformation

characteristics, load capacity and ductility of the connection.

The load-deformation charteristics were to provide information that

could be used to evaluate the contribution of the joints to drift of the

structure under wind and seismic load. Determination of ductility, including

39
application of several cycles of reversed loading, was to provide important

information with respect to required safety margins (a higher margin would

be required if a single joint could fail and lose its load resistance

before the other joints in the same story level are loaded to capacity)

and with respect to structural response to seismic loading which is related

to the ability of the joints to absorb energy.

4.4.2 Test Specimen

The test specimen for these tests consisted of three 1/2-story high

(51 1/2 in long) stub columns which were connected end to end by a continuous

grouted dowel using connection details previously described. The stub

columns were similar to those described in section 4.3.2 except that there were 5 confinement ties at both ends of each stub column replacing the

heavy steel plate used at one end of the stubs in the compressive tests.

Figure 4.20 shows stub reinforcement and several column segments under

construction.

A specimen arranged for preloading and grouting of the dowel is shown

in figure 4.21. The three stub columns shown aligned in the figure were preloaded to 10 kip using an external bar system similar to that described

in Section 4.3 before grouting a full-length dowel. This 10 kip preload

was sustained until after the test for Tests 1 and 2. For Test 3 the

external preload was increased to 90 kip after the grout had attained

the specified compressive strength (5000 psi)

.

The joint for Tests 1 and 2 was a 1/4-in thick neoprene pad. For

Test 3 the 11 ga. stainless steel neoprene sandwich was used.

The concrete and grout were similar to those described in section

4.3.2

4.4.3 Test Procedure

The test setup is shown in figures 4.22 and 4.23 for Test 1, and

Tests 2 and 3, respectively. Shear load was applied simultaneously to

both connections by applying lateral load to the center stub while the

two end stubs were restrained from translation and rotation. Load was applied in the direction of the narrow dimensions of the columns. The

tests were performed in a 600-kip hydraulic testing machine.

Joint displacement (slip) was measured electronically by vertical

displacement transducers with a resolution 0.001 in and referenced to

the laboratory floor. The displacement of the center segment was measured

in Test 1 by two transducers mounted near each joint on either side of

the segment and in Tests 2 and 3 by three tranducers mounted in its centerline,

two near the joints, and one in the center of the segment. In all three

tests the two end segments were each instrumented by two transducers on

40
Figure 4.20. Details of Shear Test Specimens

Figure 4.21. Specimen Arranged for Pre-Loading and Grouting of the Dowel

41
10 KIPS

• < < |\ - -V

* V:'.| I

Figure 4.22. Arrangement for Shear Test 1

Figure 4.23. Arrangement for Shear Tests 2 and 3

42
either side of the segment near the joint and by one transducer in the

centerline of the segment near the end of the specimen. Even though an attempt was made to keep the end segments in a fixed position, displacements

up to 0.03 in were measured near the -joints. Thus net joint slip had

to be computed by compensating for these displacements.

In Test 3, axial dowel strain was measured by bonded-foil strain

gages on two sides of the dowel in the center of each joint. The gages were arranged on the two sides of the dowel in a horizontal plane normal

to the direction of loading to eliminate the effect of dowel bending.

Dowel strain was also monitored during axial preloading. Loads were monitored

electronically by pressure transducers connected to the hydraulic systems

of the rams and the testing machine.

All data were electronically recorded after load increments varying

from 5 to 20 kip. The displacements at one of the transducers of the

center segment was also graphically monitored by an X-Y recorder to provide

information on the behavior of the specimen and to aid in the determination

of load increments. In Tests 2 and 3 the information from the X-Y plotter

was used to determine the magnitude of the displacement excursion in each

load cycle.

In Test 1 the lateral load was applied as shown in figure 4.22 in

only one direction by the testing machine. In Tests 2 and 3 the lateral

load was applied alternately in two directions by using the testing machine

in one direction and hydraulic rams in the opposite direction. This second

test set-up is illustrated in figure 4.23.

In Test 1 the center segment was loaded in increments until the

+_ 1-in clearance prevented continuation of the test.

In Tests 2 and 3 reverse cycles of load were applied as shown in

the following schedule.

Loading Schedule of Tests 2 and 3:

a/ Cycle 1: 35— kip total lateral load in both directions

(down and up)

Cycle 2: Excursion of 2 Ay— in both direction (magnitude

of load 50 kip maximum was determined by the

requirement to produce the desired excursion)

.

a/— 35 kip (17.5 kip per connection) was considered the highest load

level at which the load-displacement relationship was still linear.

— Ay is the deflection at yield as determined from the X-Y plot. Each

excursion was measured from the unloaded position at the completion of

the previous load cycle.

43
Cycles 3 through 7: Excursions of 5Ay in both directions

(maximum load approximately 70 kip)

Cycles 8 and 9: 35 kip load in both directions

Cycles 10 through 14: Excursions of 5Ay in both directions

Cycles 15: Downward load to failure

The 90 kip axial preload in Test 3 was monitored and adjusted at the

beginning of each load cycle.

4.4.4 Test Results

Figure 4.24 shows the Test 3 specimen after failure. Note the large

slip at the. joints, the extrusion of the neoprene and the cracked concrete

near the upper left corner of the center segment. The two inset photographs

show the appearance of the fractured dowel. The low-cycle fatigue fracture

of the dowel originated at a defect (see inset in lower left corner) and was similar in appearance in both Test 2 and Test 3. Test results in the form of load versus displacement at one side

of the center segment relative to the laboratory floor are shown in figures

4.25, 4.26 and 4.27. "Load" is the load on each connection which is 1/2

of the total applied load. Dowel strain caused by the axial loading of

the Test 3 specimen is shown in figure 4.28. Strain gage 8 did not function.

Figure 4.29 shows the dowel strain during Test 3 (including that caused

by the preloading to 90 kip). Figures 4.30, 4.31, and 4.32 show the net

joint displacement (slip) in Test 1 and the first half of the second cycle

in Tests 2 and 3, corrected for the measured displacements of the end

segments near the joints.

4.4.5 Discussion of Test Results

(1) Load Capacity

Test results for the connections subjected to low axial load are shown in figures 4.25 and 4.26. In figure 4.25 (Test 1) a change in

stiffness similar to yield occurred at approximately 19 kip at a joint

displacement of approximately 0.05 in. The test was discontinued at a 40-

kip load when the joint displacement reached approximately 1 in . In

Test 2 the load was applied in reverse cycles. In the first reverse load

cycle the load was carried to 17.5 kip in either direction and there was substantial deflection recovery (over 80 percent) . In the subsequent cycle

yield occurred at approximately 25 kip. The envelope of the reverse load

cycles in the direction of initial displacement (downward) is similar

to the load-displacement curve in figure 4.25. At the failure load of

+ 37 kip the displacement was approximately 1.2 in. Thus the joint was

able to resist 10 reversed load cycles with excursions equal to 5 times

44
Figure 4.24. Shear Test 3 Specimen After Test

40

Q.

Cl.

02

TEST DISCONTINUED

SHEAR TEST 1

1.0

Figure 4.25.

0.4 0.6 0.8

DISPLACEMENT in.

Vertical Displacements of South End of Center Segment

in Shear Test 1

45
-.6 -4

Figure 4.26. Vertical Displacement t)f South End of Center Segment in

Shear Test 2

46
DISPLACEMENT, in

SHEAR TEST 3 NORTH END DEFLECTION

90 kip AXIAL LOAD

Figure 4.27. Vertical Displacement of North End of Center Segment in

Shear Test 3

47
00002 00004 0.0006 00008 00010 00012 00014

STRAIN Figure 4.28. Dowel Strain Caused by Axial Loading of the Specimen in

Shear Test 3

PROBABLE YIELD

S /

S.G. Ri

_L _L -0014 -001 .0005 .001 .0015 002

STRAIN A^ i

.0025 .003

CYCLE I CYCLE 2- CYCLE 3-

30 —

Figure 4.29. Dowel Strain in Shear Test 3 Recorded by Strain Gage #7

48
40

*--N0RTH JOINT

#- SOUTH JOINT

TEST I

0.2 0.3 0.4 0.5 DISPLACEMENT, in

Figure 4.30. Net Joint Displacement in Shear Test 1

0.7

.ENVELOPE OF CYCLIC TEST MAXIMA

-±- NORTH JOINT

-#— SOUTH JOINT

TEST 2 CYCLE 2

0.2 0.3 0.4 0.5 DISPLACEMENT, in0.6 0.7

Figure 4.31. Net Joint Displacements in Shear Test 2

49
ENVELOPE OF CYCLIC TEST MAXIMA

-A-- NORTH JOINT —•— SOUTH JOINT

TEST 3 CYCLE 2

0.2 0.3 0.4 0.5 DISPLACEMENT, in

_J L

0.6 0.7

Figure 4.32. Net Joint Displacements in Shear Test 3

TEST 3 ENVELOPE _ TEST 3 CYCLE 2 (90 kip preload)

TEST 2 ENVELOPE

-TEST I SOUTH JOINT (10 kip preload)

TEST 2 CYCLE 2 SOUTH JOINT (10 kip preload)

0.2 0.3 0.4 0.5 0.6 0.7 DISPLACEMENT, in Figure 4.33. Comparison of Net Joint Displacements in Shear Tests 1, 2 and 3

50
the initial yield deflection (initial yield deflection was ^0.1 in in

Test 2) and failed at a final joint displacement equal to approximately

12 times the yield deflection.

Since Tests 1 and 2 were upper story simulations plain neoprene pads

without steel plates were used. Figure 4.26 shows the result of shear

tests on a connection using the steel-neoprene-steel sandwich and subjected

to a 90-kip axial load. In this case the preload before grouting was

10 kip and the total axial load was increased to 90 kip before the shear

force was applied. Even though it was anticipated that frictional resistance

would affect the strength of this connection, yield occurred at approximately

20 kip and thus the yield strength did not differ from that of the specimens

in the previously-disucssed tests. Dowel fracture, however, occurred

at a higher load level (62 kip) than in Test 2. As in Test 2 the response

in the initial load cycle, carried to 17.5 kip in both directions, was substantially elastic (80 percent displacement recovery)

In both Test 2 and Test 3 the initial loading was in a downward direction.

After the second load cycle a residual deflection in the downward direction

remained after unloading. This downward drift increased progressively

as additional load cycles were applied.

From the previous discussion the conclusion is drawn that the axial

load applied during the shear test had no significant effect on the yield

load in shear which was approximately 20 kip. The compressive strain

in the dowel caused by the 90-kip axial load applied in Test 3 was approximately

0.0011 as shown in figure 4.28 which is within the range of strains recorded

for a 90-kip axial load in figure 4.16 for the compressive test on the

column connection with the grouted dowel. It has been noted in Section

4.3.5 that, in the compressive load test, load transfer to the neoprene

occurred at compressive load levels above 100 kip. Thus it is reasonable

to assume that the applied axial load was supported by the dowel and the

surrounding grout column. Frictional resistance to sliding was , therefore,

probably mobilized only after the axial load was transferred to the neoprene,

and did not affect the load resistance of the specimen up to the yield

point. Since in the proposed structure only columns in the side walls

are continuously grouted, and since these side-xvall columns support only

1/2 the load acting on interior columns, it is anticipated that all joints

in side-wall columns would develop a shear capacity of approximately 20

kip. Interior lower-story column joints, where the entire dead load is

transferred to the neoprene before grouting, could be expected to develop

greater yield strength when subjected to lateral load. The loading condition

of interior joints was not simulated since it was not considered critical.

51
In summary it is concluded that the lateral load capacity at yield

of upper-story joints and lower story side-wall joints is approximately

20 kip and that all joints failed in a ductile manner when subjected to reverse cycles of lateral load. The load capactiy of interior joints

is probably higher.

(2) Response to Lateral Load

Load-displacement characteristics are shown in figures 4.30, 4.31

and 4.32. Displacement at each joint was computed by compensating for

the translation of the exterior segments of the specimen near the joint.

Figure 4.30 shows the data for Test 1. Since there was considerable rotation

in the north joint the measurement at the south joint is considered more

reliable. The test data for Test 2 are shown in figure 4.31. In this case the transducer measuring the north joint displacement may have been

obstructed since south joint displacements agree reasonably with the

displacements at the center of the center section. Thus, in this case,

the south joint measurements are taken as representative. In Test 3,

shown in figure 4.32, there is good agreement between the displacement

in both joints. The broken lines shown in Tests 2 and 3 approximately

follow the envelope connecting the peak loads in the various cycles of

downward load. In figure 4.33 the three load-displacement curves are superimposed. Test specimens 1 and 3 had a similar load-displacement

relationship up to the 20-kip load level. The Test 2 specimen had lesser

stiffness, probably because it was subjected to a previous load cycle.

Thus there is no indication that the magnitude of the axial load had any

effect on the stiffness in the elastic- range. After yielding occurred

the Test 3 specimen developed greater stiffness than those in Tests 1 and

2. Figure 4.29 shows the average axial dowel strain measured in Test

3. Only strain gage 7 is shown, since the data from the other strain

gages were erratic. The erratic readings were probably caused by dowel

bending and other disturbances in the connection. It appears that average

axial yield strain was reached during the unloading cycle of cycle 2,

which is after the completion of an excursion equal to twice the yield

displacement. Thus, "yielding" in the load-displacement curve may be

attributable to crushing of the grout column in the joint, and not necessarily

dowel-yielding. This observation, however, should be qualified since

the strain gage was located in the neutral axis of the dowel with respect

to bending and, therefore, approximately measures net axial tensile strain

rather than maximum strain, while initial dowel yielding may have been

caused by bending rather than tension.

52
It can be concluded from the previous discussion that no correlation

between axial load and stiffness in shear was observed prior to yielding.

After yielding occurred the specimen subjected to the high axial load

developed greater stiffness and strength. This observation is attributed

to the effect of frictional shear which was not mobilized prior to compressiveload transfer to the neoprene.

5. Summary

(1) The compressive -load capacity of the column connection is affected

by the radially tangential shear exerted by the bearing pad on the column

face. The effect of the various types of bearing pads on the joint capacity

is computed in the last column of Tables 4.1.2 and 4.1.3.

(2) A connection using a steel-neoprene-steel sandwich and a grouted

dowel similar to that proposed to be used in the lower stories of the

building systems had a load capacity which exceeded that predicted using

the strength coefficient computed on the basis of column connection tests

and tabulated in the last column of table 4.1.3.

(3) In a test specimen simulating the continuously-grouted column

used in the plane of the side wall:

1. the load transfer from the dowel and the surrounding

grout column to the neoprene pad occurred at the load

level of approximately 100 kip. The load transfer

mechanism was probably associated with crushing of

the grout column in the connection space together

with bond breakage between the dowel and the grout

in the vicinity of the connection;

2. dowel yielding occurred at the approximate load level

of 300 kip;

3. neither the initial load transfer to the neoprene nor

the dowel yielding caused a sudden large increment

of compressive deformation (a deformation which is

not associated with an increase in load)

;

(4) The load-deformation characteristics observed in the tests on the neoprene pads presented in figure 4.11 are reasonably consistent with

joint compression measured in the bearing tests on column connections

and presented in Table 4.1.4.

(5) For three pairs of column joints subjected to combined axial

compression and shear:

1. Yield in shear occurred at a lateral load level

of approximately 20 kip per connection and the

yield load was independent of the applied axial

compression.

53
2. Up to the yield load in shear no correlation was observed between the applied axial load and the

stiffness in shear (magnitude of slip displacement).

After yielding, the connections subjected to

high axial load developed greater stiffness in

shear and their ultimate load capacity was greater

than that 'of the connections subjected to low

axial load.

3. The connections failed in a ductile manner under

monotonic loading as well as under reversed

cycles of loading.

The findings under 1 and 2, as related to connections subjected to

high axial load, apply to the continuously-grouted columns in the plane

of the side wall. Interior column connections where the entire dead load

is supported by the neoprene pads could develop greater stiffness up to

the yield load and yield at a higher shear load.

54
6. Acknowledgment

The authors wish to acknowledge the cooperation of the following

persons

Dr. Norman F. Somes; Mr. Frank A. Rankin; Mr. James F. N. Seiler; Mr. Charles Bulik; Mr. Randolph Williams;

Mr. Thomas E. Ruschell; and Miss Cathy M. Glaze, Structures

Section, Center for Building Technology, National Bureau

of Standards

55
7. References

[1] American Concrete Institute, Building Code Requirements

for Reinforced Concrete (ACI 318-71) American Concrete

Institute, (Detroit Mich. , 1970).

[2] Richart, F. E., Brantzaeg, A., and Brown, R. L. , A Study

of the Failure of Concrete Under Combined Compressive

Stresses, Bulletin 185, University of Illinois Engineering

Experimental Station, (Urbana, 111., 1928).

[3] Richart, F. E., Brantzaeg, A., and Brown, R. L., The

Failure of Plain and Spirally Reinforced Columns in

Compression Bulletin 190, University of Illinois Experimental

Station, (Urbana 111., 1929).

[4] Blume , J. A., Newmark , N. M. , and Corning, L. H., Design

of Multistory Reinforced Concrete Buildings for Earthquake

Motion, Ch . 5, Portland Cement Association, (Chicago , 111 .

1961).

[5] Somes, N. F. , Compression Tests on Hoop-Reinforced Concrete,

Journal of the Structural Division Proceedings of the

American Society of Civil Engineers, (New York, N. Y.,

July 1970) .

[6] National Cooperative Highway Research Programs, "Elastomeric

Bearing Research," Report No. 109 Highway Research Board

(Washington, D. C. 1970).

[7] Newman, K. , and Lachance, L., "The Testing of Brittle

Materials Under Uniform Uniaxial Compressive Stress,"

ASTM Proceedings Vol. 64, American Society for Testing and

Materials, (Phila. Pa., 1964).

[8] ASTM A615-68, Deformed Billet Steel Bars for Concrete

Reinforcement, ASTM Standards, Part 4, American Society

for Testing and Materials, (Phila, Pa., 1968).

56
NBS-1 14A (REV. 7-73)

U.S. DEPT. OF COMM. BIBLIOGRAPHIC DATA

SHEET

1. PUBLIC A 'HON OK REPORT.NO. NBS TN-811

2. Gov't Ac

No.

3. Recipient's Accession No

4. TIT LI- AND SUBTITLE

Evaluation of the Column Connections Used in a Precast Concrete Modular Housing System

5. Publication Date March 1974

6. Performing Organization C odt

7. AL'TIIOR(S)

Felix Y. Yokel and Thomas W. Reichard

Performing Organ. Report N<

9. PERFORMING ORGANIZATION NAME AND ADDRESS

NATIONAL BUREAU OF STANDARDS

DEPARTMENT OF COMMERCE

WASHINGTON, D.C. 20234

10. Project Task Work Unit N<

II. Contract Grant No. IAA-H-16-70

2. Sponsoring Organization Name. and Complete Address (Street, City, State, ZIP)

Office of Research and TechnologyDepartment of Housing and Urban Development

Washington, D.C. 20410

13. Type of Report & Peril Covered Final

14. Sponsoring Agency (..

15. SUPPLEMENTARY NOTES

Supersedes NBSIR 73-148 (PB 220-366/7]

16. ABSTRAC1 (A 200-word or less (actual summary of most significant information. If document includes a significant

bibliography or literature survey, mention it here.) The column connections used in a housing system employing stacked precast concrete

box modules were tested to evaluate their structural performance. The system was proposed for construction in Operation Br eakthrough, a research and demonstration program

sponsored by the Department of Housing and Urban Development. The system uses innovative

structural design concepts, which include: confinement of the concrete in the vicinity

of the column bearings by reinforcing ties in order to increase concrete compressive

strength; neoprene pads between column bearings in the upper stories, stories; steelneoprene-steel sandwich in the lower stories; and a grouted dowel through the center of- the columns to provide resistance to tension and shear.

The test program included the following: tests to determine the effect of various

bearing pads on the load capacity of the connection; tests to determine the loaddeformation characteristics of the neoprene pads; a test to determine the performance of

a lower-story connection using a steel-neoprene-steel sandwich and a grouted dowel; and

tests to evaluate the strength and ductility of the connections when subjected to a shear force. The test results are presented and interpreted and the findings are summarized.

17. KEY Vt'ORDS (six to twelve entries; alphabetical order; capitalize only the first letter of the first key word unless a proper name; separated by semicolons)

Building system, column connection, concrete triaxial strength, ductility, neopre bearing pad, performance test, precast concrete, structural design, Operation

Breakthrough.

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