Technical Report Documentation Page

1. Report No.

FHWA-IF-02-034 

2. Government Accession No. 3. Recipient=s Catalog No.

4. Report Date

April 2002 

4. Title and Subtitle

GEOTECHNICAL ENGINEERING CIRCULAR NO. 5 

Evaluation of Soil and Rock Properties 6. Performing Organization Code: 

7. Author(s)

P.J. Sabatini, R.C. Bachus, P.W. Mayne, J.A. Schneider, T.E. Zettler 8. Performing Organization Report No.

9. Performing Organization Name and Address 10. Work Unit No.(TRAIS)

GeoSyntec Consultants 

1100 Lake Hearn Drive, NE 

Atlanta, Georgia 30342-1523 

11. Contract or Grant No.

DTFH61-94-C-00099 

13 Type of Report and Period Covered 

Technical Manual 

12. Sponsoring Agency Name and Address

U.S. Department of Transportation 

Office of Bridge Technology 

Federal Highway Administration 

400 Seventh Street, SW Washington, DC 20590 

14. Sponsoring Agency Code

15. Supplementary Notes

FHWA COTR: Chien-Tan Chang 

FHWA Technical Consultant: Jerry A. DiMaggio

16. Abstract This document presents state-of-the-practice information on the evaluation of soil and rock properties for geotechnical 

design applications. This document addresses the entire range of materials potentially encountered in highway engineering 

practice, from soft clay to intact rock and variations of materials that fall between these two extremes. 

Information is presented on parameters measured, evaluation of data quality, and interpretation of properties for conventional soil 

and rock laboratory testing, as well as in situ devices such as field vane testing, cone penetration testing, dilatometer, 

pressuremeter, and borehole jack. This document provides the design engineer with information that can be used to develop a 

rationale for accepting or rejecting data and for resolving inconsistencies between data provided by different laboratories and field 

tests. 

This document also includes information on: (1) the use of Geographical Information Systems (GIS) and Personal Data Assistance 

devices for the collection and interpretation of subsurface information; (2) quantitative measures for evaluating disturbance of 

laboratory soil samples; and (3) the use of measurements from geophysical testing techniques to obtain information on the modulus 

of soil. Also included are chapters on evaluating properties of special soil materials (e.g., loess, cemented sands, peats and organic 

soils, etc.) and the use of statistical information in evaluating anomalous data and obtaining design values for soil and rock 

properties. An appendix of three detailed soil and rock property selection examples is provided which illustrate the application of 

the methods described in the document. 

17. Key Words

Soil properties, rock properties, laboratory testing, 

in-situ testing, subsurface investigation, data quality, 

data interpretation, shear strength, consolidation, 

hydraulic conductivity, modulus 

18. Distribution Statement

No restrictions. This document is available to the public from the 

National Technical Information Service, Springfield, Virginia 

22161.

19. Security Classif. (of this report)

Unclassified

20. Security Classif. (of this page)

Unclassified

21. No. of Pages

385 

22. Price

Form DOT F 1700.7 (8-72) Reproduction of completed page authorized 

Archived Document
i 

ACKNOWLEDGEMENTS 

The authors would like to express their appreciation to Mr. Jerry A. DiMaggio, P.E. of the U.S. 

Department of Transportation Federal Highway Administration (FHWA) for providing significant 

technical assistance and review during preparation of the document. The authors would also like to 

thank Messer’s Norman Norrish, P.Eng. and Duncan Wyllie, P.Eng. of Wyllie & Norrish Rock 

Engineers for providing technical assistance on rock property evaluation. The authors would also 

like to thank the following individuals who reviewed the document and served on the Technical 

Working Group for this project: 

• Richard Cheney, P.E. – FHWA (retired); 

• Nari Abar – Geostructural Engineering, Inc. 

• David Shiells, P.E. – Virginia DOT; 

• Lawrence Pierson – Oregon DOT; 

• Sam Mansukhani – FHWA Midwestern Resource Center; and 

• Michelle Cribbs – FHWA 

The authors would also like to acknowledge Geotesting Express Inc. for providing photographs. 

Finally, the authors would like to thank Mrs. Ann Taylor and Mr. Michael Harris of GeoSyntec 

Consultants who drafted the figures and assisted in the layout of the document. 

Archived Document
ii 

PREFACE 

This document presents state-of-the-practice information on the evaluation of soil and rock 

properties for geotechnical design applications. This document was prepared to provide 

geotechnical engineers with tools to assist in the rational development of subsurface investigation 

programs, as well as in the execution of laboratory and field testing programs involving soil and 

rock, and interpretation of data from these programs. The document will be equally useful for 

structural engineers, engineering geologists, or geologists who may be responsible for field and 

laboratory testing programs. This document addresses the entire range of materials potentially 

encountered in highway engineering practice, from soft clay to intact rock and variations of 

materials that fall between these two extremes. 

In reviewing texts and course materials that are currently available to the practicing engineer, it is 

recognized that two important areas have not been sufficiently addressed. These are: (1) the use and 

role of in-situ testing; and (2) the interpretation of conflicting, contradicting, and inconsistent data. 

Regarding the first point, it is recognized that over the past 20 years, several in situ testing 

techniques have moved from the arena of university research to routine engineering practice. In 

2002, in situ testing plays a critical role in assessing soil properties and, to a lesser extent, rock 

properties, particularly by complementing laboratory-derived data. In this document detailed 

information on parameters measured, evaluation of data quality, and interpretation of properties are 

provided for conventional soil and rock laboratory testing, as well as in situ devices such as field 

vane testing, cone penetration testing, dilatometer, pressuremeter, and borehole jack. Regarding the 

second point, data resulting from the range of laboratory and in situ tests are often not completely 

consistent with other data obtained for the project and/or soil deposit. This document provides the 

design engineer with information that can be used to develop a rationale for accepting or rejecting 

data and for resolving inconsistencies between data provided by different laboratories and field tests. 

This document relies on previous good practice in the evaluation of soil and rock properties. This 

good practice is extended by more recent developments in the areas of engineering property 

evaluation methods by including: (1) use of Geographical Information Systems (GIS) and Personal 

Data Assistance (i.e., handheld computer) devices for the collection and interpretation of subsurface 

information; (2) quantitative measures for evaluating disturbance of laboratory soil samples; and (3) 

use of measurements from seismic and geophysical testing techniques to obtain information on the 

modulus of soils for static deformation analyses. Other features of this document include a chapter 

on evaluating properties of special soil materials (e.g., loess, cemented sands, peats and organic 

soils), a chapter on the use of statistical information in evaluating anomalous data and obtaining 

design values for soil and rock properties, and an appendix of three detailed soil and rock property 

selection examples which illustrate the application of the methods described in the document for 

property evaluation. 

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iii

TABLE OF CONTENTS 

CHAPTER 1 ....................................................................................................... 1

1.1 INTRODUCTION................................................................................................................... 1

1.2 BACKGROUND ..................................................................................................................... 1

1.3 DOCUMENT ORGANIZATION.......................................................................................... 2

CHAPTER 2 ...................................................................................................... 4

2.1 INTRODUCTION................................................................................................................... 4

2.2 PROCESS OF SOIL AND ROCK PROPERTY SELECTION ......................................... 4

2.3 USE OF CORRELATIONS TO ASSIST PROPERTY SELECTION .............................. 7

2.4 USE OF OBSERVATIONAL METHOD ............................................................................. 9

CHAPTER 3 ..................................................................................................... 10

3.1 INTRODUCTION................................................................................................................. 10

3.2 PLANNING THE SUBSURFACE INVESTIGATION AND LABORATORY TESTING 

PROGRAM............................................................................................................................ 10

3.2.1 General.............................................................................................................................10 

3.2.2 Identify Data Needs .........................................................................................................10 

3.2.3 Gather and Analyze Existing Information .......................................................................11 

3.2.4 Conduct Site Visit............................................................................................................17 

3.2.5 Develop Preliminary Site Model .....................................................................................18 

3.2.6 Developing a Site Investigation Program ........................................................................20 

3.2.7 Developing a Laboratory Testing Program......................................................................22 

CHAPTER 4 ..................................................................................................... 26

4.1 INTRODUCTION................................................................................................................. 26

4.2 BORING METHODS ........................................................................................................... 27

4.3 SAMPLING METHODS...................................................................................................... 32

4.3.1 Disturbed Sampling of Soil..................................................................................................32 

4.3.2 Undisturbed Sampling of Soil..............................................................................................34 

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TABLE OF CONTENTS (continued) 

iv

4.3.2.1 General.........................................................................................................................34 

4.3.2.2 Overview of Thin-Walled Tube Sampling ..................................................................39 

4.3.3 Rock Coring.........................................................................................................................42 

4.4 STANDARD PENETRATION TEST (SPT) ...................................................................... 44

4.4.1 General.................................................................................................................................44 

4.4.2 Procedures............................................................................................................................45 

4.4.3 Parameters Measured...........................................................................................................45 

4.5 CONE PENETRATION TESTS (CPT / CPTU / SCPTU)................................................ 48

4.5.1 General.................................................................................................................................48 

4.5.2 Equipment............................................................................................................................50 

4.5.3 Procedures............................................................................................................................52 

4.5.4 Parameters Measured...........................................................................................................54 

4.6 FLAT DILATOMETER TEST (DMT) .............................................................................. 56

4.6.1 General.................................................................................................................................56 

4.6.2 Equipment............................................................................................................................56 

4.6.3 Procedures............................................................................................................................56 

4.6.4 Parameters Measured...........................................................................................................58 

4.7 PRESSUREMETER TEST (PMT) ..................................................................................... 58

4.7.1 General.................................................................................................................................58 

4.7.2 Equipment............................................................................................................................59 

4.7.3 Procedures............................................................................................................................59 

4.7.4 Parameters Measured...........................................................................................................60 

4.8 VANE SHEAR TEST (VST) ................................................................................................ 62

4.8.1 General.................................................................................................................................62 

4.8.2 Equipment............................................................................................................................62 

4.8.3 Procedures............................................................................................................................63 

4.8.4 Parameters Measured...........................................................................................................64 

4.9 USE OF DRILL RIGS TO PERFORM IN-SITU TESTS ................................................ 65

4.10. IN-SITU TESTING IN ROCK ............................................................................................ 65

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v

4.10.1 General.............................................................................................................................65 

4.10.2 Borehole Dilatometer.......................................................................................................66 

4.10.3 Borehole Jack...................................................................................................................67 

4.10.4 In-situ Direct Shear Testing .............................................................................................67 

4.11 GEOPHYSICAL TESTING ................................................................................................ 68

4.12 LABORATORY SOIL TESTING....................................................................................... 74

4.12.1 Introduction......................................................................................................................74 

4.12.2 Quality Assurance for Laboratory Testing ......................................................................74 

4.12.2.1 Sample Tracking ..........................................................................................................74 

4.12.2.2 Sample Storage ............................................................................................................78 

4.12.2.3 Sample Handling..........................................................................................................78 

4.12.2.4 Specimen Selection......................................................................................................78 

4.12.3 Effects of Sample Disturbance.........................................................................................79

4.12.4 Laboratory Index Tests for Soils......................................................................................82

4.12.4.1 General.........................................................................................................................82 

4.12.4.2 Moisture Content .........................................................................................................82 

4.12.4.3 Unit Weight..................................................................................................................83 

4.12.4.4 Atterberg Limits...........................................................................................................83 

4.12.4.5 Particle Size Distribution .............................................................................................86 

4.12.4.6 Laboratory Classification.............................................................................................87 

4.12.4.7 Specific Gravity ...........................................................................................................87 

4.12.4.8 Organic Content...........................................................................................................87 

4.12.4.9 Electro Chemical Classification Tests .........................................................................88 

4.12.5 Laboratory Performance Tests for Soils ..........................................................................88 

4.12.5.1 General.........................................................................................................................88 

4.12.5.2 Consolidation ...............................................................................................................89 

4.12.5.3 Soil Strength.................................................................................................................92 

4.12.5.4 Permeability .................................................................................................................99 

4.13 LABORATORY ROCK TESTS........................................................................................ 102

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vi

4.13.1 Introduction....................................................................................................................102 

4.13.2 Laboratory Testing of Rock ...........................................................................................104

4.13.2.1 Point-Load Strength Test ...........................................................................................104

4.13.2.2 Unconfined Compressive Strength of Intact Rock Core............................................105 

4.13.2.3 Elastic Moduli of Intact Rock Core ...........................................................................106 

4.13.2.4 Laboratory Direct Shear Test.....................................................................................106 

CHAPTER 5 ................................................................................................... 108

5.1 INTRODUCTION............................................................................................................... 108

5.2 INTERPRETATION OF SUBSURFACE STRATIGRAPHY....................................... 108

5.2.1 General...............................................................................................................................108 

5.2.2 Soil Classification by Soil Sampling and Drilling.............................................................108 

5.2.3 Soil Classification by Cone Penetration Testing ...............................................................112 

5.2.4 Soil Classification using the Flat-Plate Dilatometer..........................................................117 

5.2.5 Generating a Subsurface Profile ........................................................................................118 

5.3 IN-SITU STRESS STATE.................................................................................................. 123

5.3.1 General...............................................................................................................................123 

5.3.2 Overburden Stresses...........................................................................................................123 

5.3.3 Horizontal Stresses.............................................................................................................124 

5.4 CONSOLIDATION PROPERTIES OF SOIL................................................................. 125

5.4.1 General...............................................................................................................................125 

5.4.2 Laboratory Consolidation Tests.........................................................................................125 

5.4.2.1 General.......................................................................................................................125 

5.4.2.2 Soil Parameters from Laboratory Consolidation Tests..............................................126 

5.4.2.3 Selection of Samples for Laboratory Consolidation Testing.....................................126 

5.4.2.4 Evaluation of σp′ from Laboratory Consolidation Tests............................................128 

5.4.2.5 Evaluation of Cc and Cr..............................................................................................130 

5.4.2.6 Laboratory Evaluation of cv .......................................................................................131 

5.4.2.7 Evaluation of Cαε .......................................................................................................134 

5.4.3 Evaluation of σp′ from In-situ Test Methods.....................................................................136 

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vii

5.4.4 Evaluation of ch from CPTu Dissipation Data...................................................................140 

5.4.5 Selection of Design Values for Consolidation Analyses ...................................................142 

5.5 GENERAL STRESS-STRAIN AND STIFFNESS PROPERTIES ................................ 146

5.5.1 Background........................................................................................................................146 

5.5.2 Settlement Analysis for Soils.............................................................................................147 

5.5.3 Method to Evaluate Equivalent Elastic Modulus...............................................................148 

5.5.4 Evaluation of Shear Wave Velocity...................................................................................151 

5.5.4.1 General.......................................................................................................................151 

5.5.4.2 Field Measurements of Shear Wave Velocity ...........................................................151 

5.5.5 Correlations for Small-Strain Shear Modulus....................................................................154 

5.5.6 Evaluation of Modulus Degradation Value .......................................................................155 

5.5.7 Summary............................................................................................................................156 

5.6 SHEAR STRENGTH PROPERTIES OF SOIL .............................................................. 156

5.6.1 Introduction........................................................................................................................156 

5.6.2 Fundamental Concepts of Soil Shear Strength ..................................................................159 

5.6.2.1 Drained versus Undrained Loading ...........................................................................159 

5.6.2.2 Drained Stress-Strain-Strength Behavior...................................................................160 

5.6.2.3 Undrained Stress-Strain-Strength Behavior...............................................................162 

5.6.2.4 Effective Stress Parameters........................................................................................163 

5.6.2.5 Total Stress Parameters..............................................................................................167 

5.6.3 Relevance of Design Applications to Soil Shear Strength Evaluation ..............................167 

5.6.4 Laboratory Testing Methods for Evaluating Soil Shear Strength......................................168 

5.6.4.1 Selection of Laboratory Testing Method ...................................................................168 

5.6.4.2 Triaxial Testing..........................................................................................................169 

5.6.4.3 Direct Shear Testing ..................................................................................................175 

5.6.4.4 Unconfined Compression Testing..............................................................................178 

5.6.4.5 Relevance of Laboratory Strength Tests to Field Conditions....................................179 

5.6.5 Undrained Shear Strength from In-situ Tests ....................................................................180 

5.6.6 Drained Friction Angle of Granular Soils from In-situ Tests............................................184 

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viii

5.6.7 Selection of Total Stress Parameters (su) for Undrained Strength Design Analyses.........187 

5.6.8 Selection of Effective Stress Parameters (φ′, c′) for Design Analyses ..............................189 

5.7 HYDRAULIC CONDUCTIVITY PROPERTIES OF SOIL.......................................... 189

5.7.1 Introduction........................................................................................................................189 

5.7.2 Laboratory Output/Data Reduction....................................................................................190 

5.7.3 Correlation Methods ..........................................................................................................191 

5.7.4 Interpretation Methods.......................................................................................................193 

CHAPTER 6 ................................................................................................... 195

6.1 INTRODUCTION............................................................................................................... 195

6.2 ROCK MASS CLASSIFICATION ................................................................................... 196

6.2.1 Description of Rock Masses ..........................................................................................196 

6.2.2 Core Recovery and Rock Quality Designation..............................................................200 

6.2.3 CSIR Classification........................................................................................................202 

6.3 ROCK UNIAXIAL COMPRESSIVE STRENGTH ........................................................ 204

6.4 ROCK DEFORMATION MODULUS VALUES ............................................................ 204

6.4.1 Intact Rock Modulus......................................................................................................204 

6.4.2 Rock Mass Modulus ......................................................................................................206 

6.4.2.1 Method Based On Rock Mass Rating........................................................................206 

6.4.2.2 Method Based on RQD..............................................................................................207 

6.4.2.3 Use of In situ Tests to Evaluate Rock Mass Modulus ...............................................208 

6.4.3 Selection of Rock Deformation Modulus for Design ....................................................211 

6.5 ROCK SHEAR STRENGTH............................................................................................. 211

6.5.1 Mohr-Coulomb Materials ..............................................................................................211 

6.5.2 Shear Strength Of Discontinuities .................................................................................213 

6.5.2.1 General.......................................................................................................................213 

6.5.2.2 Friction Angle of Rock Surfaces................................................................................213 

6.5.2.3 Surface Roughness.....................................................................................................214 

6.5.2.4 Measurement of Surface Roughness..........................................................................216 

6.5.2.5 Discontinuity Infilling................................................................................................217 

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TABLE OF CONTENTS (continued) 

ix

6.5.2.6. Effect of Water on Shear Strength.............................................................................218 

6.5.2.7 Laboratory Direct Shear Testing................................................................................222 

6.5.3 Shear Strength Of Fractured Rock Masses ....................................................................223 

6.5.3.1 General.......................................................................................................................223 

6.5.3.2 Strength Determination by Back Analysis of Failures ..............................................223 

6.5.3.3 Hoek-Brown Strength Criteria for Fractured Rock Masses.......................................224 

6.5.4 Selection of Rock Shear Strength for Design ................................................................227 

CHAPTER 7 ................................................................................................... 229

7.1 INTRODUCTION............................................................................................................... 229 

7.2 LOESS.................................................................................................................................. 230 

7.2.1 Identification of Loess ...................................................................................................230 

7.2.2 Issues Related to Subsurface Exploration in Loess .......................................................231 

7.2.3 Laboratory Strength Testing of Loess............................................................................232 

7.2.4 Evaluation of Collapse Potential....................................................................................232 

7.3 EXPANSIVE SOILS........................................................................................................... 233 

7.3.1 Identification of Expansive Soils...................................................................................233 

7.3.2 Evaluation of Expansion (Swell) Potential ....................................................................235 

7.3.3 Shear Strength Evaluation of Expansive Soils...............................................................236 

7.4 ORGANIC SOILS AND PEAT ......................................................................................... 236 

7.4.1 Introduction....................................................................................................................236 

7.4.2 Identification of Organic Soils and Peat ........................................................................237 

7.4.3 Issues Related to Subsurface Exploration and Sampling of Organic Soils and Peat.....238 

7.4.4 Shear Strength of Organic Soils and Peats ....................................................................238 

7.4.5 Compressibility of Organic Soils and Peats...................................................................239 

7.5 COLLUVIUM AND TALUS ............................................................................................. 240 

7.5.1 Identification of Colluvium and Talus...........................................................................240 

7.5.2 Issues Related to Subsurface Exploration and Testing in Colluvium............................241 

7.5.3 Issues Related to Subsurface Exploration and Testing in Talus....................................241 

7.5.4 Compressibility of Colluvium and Talus.......................................................................242 

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x

7.5.5 Shear Strength of Colluvium and Talus.........................................................................242 

7.6 SHALES AND DEGRADABLE MATERIALS............................................................... 243 

7.6.1 Identification of Degradable Materials ..........................................................................243 

7.6.2 Slake Durability Test .....................................................................................................244 

7.6.3 Jar Slake Test .................................................................................................................244 

7.6.4 Point-Load Test..............................................................................................................245 

7.6.5 Use of Shale Material.....................................................................................................245 

7.7 CEMENTED SANDS.......................................................................................................... 246

7.7.1 Identification of Cemented Sands..................................................................................246 

7.7.2 Issues Related to Subsurface Exploration and Testing in Cemented Sands ..................247 

7.7.3 Interpretation of Laboratory and Field Testing Results in Cemented Sands .................247 

7.8 SENSITIVE CLAYS........................................................................................................... 248 

7.8.1 Identification of Sensitive Clays....................................................................................248 

7.8.2 Issues Related to Subsurface Exploration and Testing in Sensitive Clays ....................249 

7.8.3 Interpretation of Laboratory and Field Testing Results in Sensitive Soils ....................250 

7.9 PARTIALLLY SATURATED SOILS .............................................................................. 250 

7.9.1 Identification of Partially Saturated Soils ......................................................................250 

7.9.2 Sampling and Testing Partially Saturated Soils.............................................................251 

7.9.3 Interpretation of Laboratory and Field Testing Results of Partially Saturated Soils.....251 

CHAPTER 8 ................................................................................................... 253

8.1 INTRODUCTION............................................................................................................... 253

8.2 RESOLVING INCONSISTENCIES BETWEEN TEST RESULTS ............................. 253

8.3 ESTIMATING VARIABILITY OF SELECTED PARAMETERS............................... 254

8.4 FINAL SELECTION OF DESIGN PARAMETERS ...................................................... 255

REFERENCES............................................................................................... 258

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TABLE OF CONTENTS (continued) 

xi

APPENDIX A SOIL AND ROCK PROPERTY SELECTION EXAMPLES 

Example 1 Soft to Medium Clay and Overconsolidated Clay Crust 

Example 2 Piedmont Residual Soil, Weathered Rock, And Rock 

Example 3 Heavily Overconsolidated Clays 

APPENDIX B CALCULATION OF THE COEFFICIENT OF CONSOLIDATION 

FROM LABORATORY DATA 

APPENDIX C ALTERNATIVE APPROACH TO EVALUATE HORIZONTAL 

COEFFICIENT OF CONSOLIDATION VALUES FROM PIEZOCONE 

DISSIPATION TESTS 

Archived Document
LIST OF TABLES 

Table Page

xii 

1 Summary of information needs and testing considerations for a range of highway 

applications. ...................................................................................................................... 12 

2 Sources of historical site data. ...........................................................................................15

3 Guidelines for minimum number of investigation points and depth of investigation .......24 

4 Reference publications on in-situ testing...........................................................................27 

5(a) Boring methods (modified after Day, 1999)......................................................................30 

5(b) Rock core drilling methods (modified after Day, 1999)....................................................31 

5(c) Other exploratory techniques (modified after Day, 1999).................................................32 

6(a) Common samplers to collect disturbed soil samples (modified after NAVFAC, 

1982). ................................................................................................................................33 

6(b) Specialty samplers to collect disturbed soil samples (modified after NAVFAC, 

1982). ................................................................................................................................34 

7 Nominally undisturbed soil samplers (modified after NAVFAC, 1982)...........................41 

8 Factors affecting the SPT and SPT results (after Kulhawy and Mayne, 1990). ................46 

9 Corrections to the SPT (after Skempton, 1986).................................................................47 

10 In-situ testing methods used in soils. .................................................................................49 

11 In-situ testing methods used in rock. .................................................................................66 

12 Geophysical testing techniques..........................................................................................70 

13 Common soil laboratory tests. ...........................................................................................75 

14 Methods for index testing of soils......................................................................................76 

15 Methods for performance testing of soils. .........................................................................77 

16 Sample quality designation system (Lacasse et al., 1985).................................................81 

17 U.S. standard sieve sizes and corresponding opening dimension......................................86 

18 Time tf to reach failure (after Head, 1986). .......................................................................99 

19 Recommended maximum hydraulic gradient for permeability testing............................101 

20 Common rock laboratory tests.........................................................................................103 

21 Summary information on rock laboratory test methods. .................................................103 

22 SPT N value soil property correlations for granular soils (after AASHTO, 1988). ........110 

23 SPT N value soil property correlations for cohesive soils (after AASHTO, 1988).........110 

24 Casagrande method to evaluate σp′..................................................................................128 

25 Strain-energy method to evaluate σp′...............................................................................129 

26 Summary of correlations for Cc (modified after Holtz and Kovacs, 1986). ....................131 

27 Modified time factors, T*, for analysis of CPTu dissipation data (after Teh and 

Houlsby, 1991).................................................................................................................141 

28 Elastic constants of various soils based on soil type (modified after AASHTO, 

1996). ...............................................................................................................................148 

29 Elastic constants of various soils based on SPT N value (modified after AASHTO, 

1996). ...............................................................................................................................148 

30 Typical values of small-strain shear modulus..................................................................154 

31 Unconfined compressive strength of particles for rockfill grades in figure 73. ..............164 

32 Summary of issues relevant to shear strength evaluation in support of the design of 

typical geotechnical features............................................................................................168 

33 Conventional methods of interpretation for su from in-situ tests.....................................180

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LIST OF TABLES (continued) 

Table Page

xiii 

34 Relationship among relative density, SPT N value, and internal friction angle of 

cohesionless soils (after Meyerhof, 1956). ......................................................................184 

35 Comparison of hydraulic conductivity from empirical and field methods......................193 

36 Description of geological mapping terms........................................................................198 

37 Rock material strengths....................................................................................................199 

38 Weathering grades ...........................................................................................................199 

39 Rock quality description based on RQD..........................................................................202 

40 CSIR classification of jointed rock mass.........................................................................203 

41 Typical elastic constants for intact rock (after Wyllie, 1999)..........................................206 

42 Estimation of EM based on RQD (modified after Carter and Kulhawy, 1988)................207 

43 Typical ranges of friction angles for a variety of rock types (after Barton, 1973; 

Jaeger and Cook, 1976)....................................................................................................214 

44 Shear strength of filled discontinuities (modified after Hoek and Bray, 1977)...............220 

45 Approximate relationship between rock-mass quality and material constants used in 

defining nonlinear strength (Hoek and Brown, 1988). ....................................................225 

46 Parameters used to develop strength envelopes in figure 105. ........................................227 

47 Summary of sampling difficulties and engineering characteristics of special 

materials...........................................................................................................................229 

48 Qualitative assessment of collapse potential (after ASTM D 5333)................................233 

49 Organic soils and peat classification properties (after Landva et al., 1983)....................238 

50 Evaluation of jar slake index, IJ. ......................................................................................245 

51 Criteria for rockfill materials (after Strohm et al., 1978).................................................246 

52 Values of coefficient of variation, V, for geotechnical properties and in situ tests 

(after Duncan, 2000). .......................................................................................................256 

A-1 Summary of soil property selection examples................................................................ A-1 

A-2 Summary of CPT subsurface profile interpretation........................................................ A-7 

A-3 Summary of CPTu subsurface profile interpretation...................................................... A-9 

A-4 Summary of laboratory testing on varved clay samples. .............................................. A-15 

A-5 Parameters used for analysis of dissipation data........................................................... A-26 

A-6 Evaluation of ch from CPTu2 dissipation data at 6.68 m. ............................................. A-26 

A-7 Evaluation of ch from CPTu2 dissipation data at 17.65 m. ........................................... A-26 

A-8 Summary of triaxial test data. ....................................................................................... A-31 

A-9 Vane shear data............................................................................................................. A-37 

A-10 Index properties of Piedmont soils at the Alabama site................................................ A-44 

A-11 Summary of CPTu subsurface profile interpretation.................................................... A-46 

A-12 Summary of oedometer testing on Piedmont residual soils.......................................... A-54 

A-13 Test data and calculated elastic modulus from Menard PMT data at the Alabama 

site................................................................................................................................. A-55 

A-14 Test data and calculated unload-reload elastic modulus from full displacement PMT 

data (C-41) at the Alabama site. ................................................................................... A-56 

A-15 Summary of elastic modulus values from various tests................................................ A-61 

A-16 Stress-strain data for specimen B2-1-1 at 15 m............................................................ A-63 

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LIST OF TABLES (continued) 

Table Page

xiv 

A-17 Strength properties from CIUC triaxial tests on Piedmont soils from the Alabama 

site................................................................................................................................. A-64 

A-18 Soil stratigraphy from SPT and classification testing................................................... A-75 

A-19 Summary of oedometer testing on heavily overconsolidated clays.............................. A-81 

Archived Document
LIST OF FIGURES 

Figure Page

xv 

 

1 Soil and rock property selection flowchart. ......................................................................5 

2 Checklist items for site reconnaissance. .........................................................................19 

3 Large diameter auger boring...........................................................................................29 

4 Split barrel sampler.........................................................................................................33 

5 Thin walled (Shelby) tube for sampling (with end caps)................................................35 

6 Stationary piston sampler................................................................................................36 

7 Denison sampler..............................................................................................................37 

8 Pitcher sampler................................................................................................................38 

9 Single and double tube rock core barrels (after FHWA-HI-97-021, 1997)....................43 

10 SPT performed at the back of a drill rig. ........................................................................44 

11 Stress normalization parameter, CN, for sands................................................................48 

12 Cone and piezocone penetrometers. ...............................................................................52 

13 Cone penetration testing from cone truck.......................................................................53 

14 Measurement locations on cone penetrometers. .............................................................54 

15 Illustration of unequal end areas of CPT (after Kulhawy and Mayne, 1990).................55 

16 Flat plate dilatometer test equipment..............................................................................57 

17 Pre-bored pressuremeter equipment. ..............................................................................61 

18 Typical curves and characteristic pressures for pre-bored Menard pressuremeter.........61 

19 (a) Rectangular vane; and (b) Parameters used to define vane dimensions....................63 

20 Typical pressure-dilation graphs for a borehole dilatometer (after ISRM, 1987). .........67 

21 Typical setup for an in-situ direct shear test in an adit (after Saint Simon et al., 

1979). ..............................................................................................................................68 

22 Disturbance during sampling and trimming (after Ladd and Lambe, 1963). .................79 

23 Lab and field consolidation curves. ................................................................................80 

24 Equipment used for Atterberg limits testing of soil........................................................84 

25 Location of clay minerals on the Casagrande plasticity chart (Skempton, 1953). .........85 

26 Components of consolidation test...................................................................................89 

27 Incremental load oedometer............................................................................................90 

28 Failure of a loose sand specimen in a triaxial cell; and (b) Load frame, pressure 

panel, and computerized data acquisition system...........................................................93 

29 Direct shear testing box. .................................................................................................94

30 Soil sample mounted in direct shear testing apparatus. ..................................................95 

31 Rigid wall permeameter................................................................................................100 

32 Flexible wall permeameter............................................................................................101 

33 Point load strength test equipment................................................................................104 

34 Laboratory direct shear testing equipment for rock......................................................107 

35 Traditional drilling, sampling, and laboratory testing of collected samples.................109 

36 Variability of SPT N values..........................................................................................111 

37 Comparison of SPT (N1)60 and CPT qt values. ............................................................112 

38 Soil classification based on qt and FR (Robertson et al., 1986)....................................113 

39 Soil classification based on qt and Bq (Robertson et al., 1986)....................................113 

40 Typical CPTu log..........................................................................................................115

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LIST OF FIGURES (continued) 

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41 CPTu log with subsurface stratigraphy interpretation ..................................................116 

42 CPTu dissipation test in sand........................................................................................117 

43 Soil classification based on DMT.................................................................................118 

44 Summary plot of Atterberg limits data. ........................................................................120 

45 Boring location plan. ....................................................................................................121 

46 Interpreted subsurface profile. ......................................................................................121 

47 Definition of Cc, Cr, Cs, σp′. .........................................................................................126 

48 Profile of preconsolidation stress..................................................................................127 

49 Illustration of Casagrande method to evaluate preconsolidation stress........................129 

50 Illustration of strain-energy method to evaluate preconsolidation stress......................130 

51 Summary consolidation data showing cv. .....................................................................133 

52 Evaluation of Cα............................................................................................................135 

53 Correlation of σp′ with CPT qt data (after Kulhawy and Mayne, 1990).......................137 

54 Correlation of σp′ with CPTu u1 data (after Kulhawy and Mayne, 1990). ...................137 

55 Correlation of σp′ with CPTu u2 data (after Kulhawy and Mayne, 1990). ...................138 

56 Correlation of σp′ with DMT po data (after Kulhawy and Mayne, 1990).....................138 

57 Correlation of σp′ with self boring PMT pL data (after Kulhawy and Mayne, 

1990). ............................................................................................................................139 

58 Correlation of σp′ with VST su,VST data (after Kulhawy and Mayne, 1990)................139 

59 CPTu2 pore pressure dissipation curves........................................................................141 

60 Rigidity index (after Keaveny and Mitchell, 1986)......................................................142 

61 Preconsolidation stress from oedometer and DMT. .....................................................144 

62 Correlation of cv to LL..................................................................................................146 

63 Strength measured by in-situ tests at peak of stress-strain curve..................................149 

64 Stress-strain-strength curves for three geomaterials having the same strength yet 

different stiffness. .........................................................................................................149 

65 Variation of modulus with strain level..........................................................................150 

66 Field and laboratory methods to evaluate shear wave velocity. ...................................152 

67 Seismic dilatometer test. ...............................................................................................154 

68 Modulus degradation based on g=0.3. ..........................................................................155 

69 Shear wave velocity profile from seismic cone sounding. ...........................................157 

70 Various laboratory tests used to measure soil strength showing the imposed 

stresses and loading conditions.....................................................................................158 

71 Drained stress-strain behavior.......................................................................................160 

72 Mohr-Coulomb failure criteria......................................................................................161 

73 Typical ranges of friction angle for rockfills, gravels, and sands (Terzaghi, 

Peck, and Mesri, 1996). ................................................................................................164 

74 Relationship between φ′ and PI (Terzaghi, Peck, and Mesri, 1996).............................165 

75 Relationship between c′ and σp′ (Mesri and Abdel-Ghaffar, 1993). ............................166 

76 Residual friction angles for clayey soils (after Stark & Eid, 1994)..............................166 

77 Stress-strain and Mohr circle representation for four UU tests performed on the 

same soil (after Day, 1999)...........................................................................................170 

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78 Interpretation of UU test data. ......................................................................................171 

79 CIU triaxial compression test results. ...........................................................................173 

80 Effective stress path in undrained shear........................................................................174 

81 Direct shear test results. ................................................................................................176 

82 Area correction for direct shear test..............................................................................177 

83 Typical stress-strain curve and Mohr circle representation of the state of stress 

for an unconfined compression test. .............................................................................178 

84 Shear modes for an embankment slip surface...............................................................179 

85 Plasticity based VST correction factors........................................................................181 

86 Example su profile (after Finno and Chung, 1992).......................................................182 

87 Correlation of φ′ with SPT N60 data in clean sands. .....................................................185 

88 Correlation of φ′ with normalized CPT qt data in clean sands......................................186 

89 Correlation of φ′ with the DMT KD parameter for clean sands. ...................................186 

90 Range of hydraulic conductivity values based on soil type. ........................................192 

91 Range of hydraulic conductivity based on grain size (after GeoSyntec, 1991)............194 

92 Illustration of geological mapping terms (after Wyllie, 1999). ....................................196 

93 List of parameters and categories describing rock mass characteristics (after 

Wyllie, 1999). ...............................................................................................................197 

94 Calculation of core recovery and RQD.........................................................................201 

95 Axial and diametral stress-strain curves for intact rock tested in uniaxial 

compression. .................................................................................................................205 

96 Relationship between in situ modulus and rock mass rating (after Bieniawski, 

1978; Serafim and Pereira, 1983). ................................................................................207 

97 Pressure-displacement plot for borehole jack...............................................................210 

98 Curve of Etrue versus Ecalc (after ASTM D 4971)..........................................................210 

99 Relationships between shear stress and normal stress on rupture surface for five 

different geological conditions (TRB, 1996)................................................................213 

100 Effect of surface roughness and normal stress on the friction of a discontinuity 

surface...........................................................................................................................214 

101 Definition of joint roughness coefficient, JRC (Barton, 1973).....................................216 

102 Simplified division of filled discontinuities into displaced and undisplaced, and 

NC and OC categories. .................................................................................................219 

103 Results of direct shear test of filled discontinuity showing measurements of 

shear strength and roughness (after Wyllie, 1999). ......................................................222 

104 Typical curved shear strength envelope defined by Hoek-Brown theory for rock 

mass strength (Hoek, 1983). .........................................................................................224 

105 Illustration of use of nonlinear shear strength for three fractured rock mass 

types. .............................................................................................................................226 

106 Classification chart for swelling potential (after Seed et al., 1962)..............................234 

107 Guide to collapsibility, compressibility, and expansion based on in-situ dry 

density and liquid limit (after Mitchell and Gardner, 1975 and Gibbs, 1969)..............235 

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LIST OF FIGURES (continued) 

Figure Page

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108 Values of natural water content and compression index for peats, clays, and silts 

(after Mesri, Stark, Ajlouni, and Chen, 1997). .............................................................240 

A-1 Boring log B1............................................................................................................... A-4 

A-2 Summary data for samples from Borings B1, B2, and B3........................................... A-5 

A-3 Subsurface soil layering based on CPT data and friction ratio.................................... A-6 

A-4 Classification of soil type based on qt and FR. ............................................................ A-7 

A-5 Subsurface soil layering based on CPTu data, friction ratio, and Bq parameter.......... A-8 

A-6 Classification of soil type based on qt and Bq.............................................................. A-8 

A-7 Subsurface soil layering based on DMT index values............................................... A-10 

A-8 Classification and consistency of soil layers based on DMT data............................. A-10 

A-9 Index properties and “Po” diagram for Boring B-1.................................................... A-12 

A-10 Evaluation of σp′using the Casagrande method......................................................... A-14 

A-11 Evaluation of σp′ using the strain energy method...................................................... A-16

A-12 Profile of σp′ based on correlations with in-situ testing devices................................ A-17 

A-13 Sample graphical evaluation for Ccε and Crε.............................................................. A-19 

A-14 Empirical correlations of Cc and Ccε.......................................................................... A-20 

A-15 Secondary compression results for Connecticut Valley varved clays 

investigation............................................................................................................... A-21 

A-16 Laboratory-measured Ccε vs. depth............................................................................ A-22 

A-17 cv values from laboratory consolidation tests. ........................................................... A-23 

A-18 CPTu2 pore pressure dissipation curves..................................................................... A-25 

A-19 Normalized CPTu2 dissipation curves. ...................................................................... A-25 

A-20 Cαε values from laboratory consolidation tests.......................................................... A-28 

A-21 Comparison of Cαε and Ccε for laboratory consolidation tests................................... A-28 

A-22 Ko value calculated using in-situ testing devices....................................................... A-29 

A-23 Po diagram developed from 1-D consolidation data. ................................................. A-31 

A-24 Stress - strain curve and stress path for a UU test on a Connecticut Valley 

varved clay specimen................................................................................................. A-32 

A-25 Stress - strain curve and stress path for a CIUC test on a Connecticut Valley 

varved clay specimen................................................................................................. A-33 

A-26 Effective stress properties from CIUC test data. ....................................................... A-34 

A-27 Po diagram with OCR and su vs. depth. ..................................................................... A-35 

A-28 Normalized strength property relationships for various shear modes. ...................... A-35 

A-29 Estimation of shear strength from in-situ tests: (a) VST; (b) empirical CPT, 

CPTu, & DMT based on correlations in table 33; and (c) based on Equation 

A-14 with φ′=18.2o

..................................................................................................... A-39 

A-30 Residual soil classification and typical gneiss and schist weathering profile 

(after Sowers and Richardson, 1983)......................................................................... A-40 

A-31 Boring log for Alabama site....................................................................................... A-42 

A-32 Summary index test data for samples from borings B-1, B-3, B-4, and B-6 at 

Alabama site............................................................................................................... A-43

A-33 Subsurface layering based on CPTu data................................................................... A-45 

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A-34 Soil classification charts with CPTu data from Alabama site (signature 1). ............ A-47 

A-35 Soil stratigraphy from DMT index values. ................................................................ A-50 

A-36 Classification and consistency of soil at Alabama site based on DMT data ............ A-50 

A-37 Comparison of Alabama site data to published trend between CPT tip 

A-37 resistance and N-value as a function of mean grain size (Kulhawy & Mayne, 

1990). ......................................................................................................................... A-52 

A-38 Comparison of Opelika, Alabama data to published trend between dilatometer 

modulus and N-value in Piedmont sandy silts (Mayne & Frost, 1989)..................... A-52 

A-39 Comparison of Opelika, Alabama data to published trends between PMT 

A-39 modulus and N-Value in Piedmont soils (Martin, 1977)........................................... A-53 

A-40 Laboratory 1-D consolidation curves (a) Stress – Strain; (b) Stress – Strain 

Energy........................................................................................................................ A-54 

A-41 Menard pressuremeter data for the Alabama site. ..................................................... A-56 

A-42 Full displacement pressuremeter data for the Alabama site. ..................................... A-57 

A-43 Dilatometer modulus ED compared to pressuremeter modulus, EP. .......................... A-59 

A-44 Calculation of elastic modulus from shear wave velocity data.................................. A-60 

A-45 Stress-strain curve for CIUC tests on samples B2-1-1 at 15 m and B7-1 at 4 

m. A-62 

A-46 Effective stress paths for CIUC tests shown in figure A-45...................................... A-65 

A-47 Effective stress failure envelopes based on triaxial test data..................................... A-65 

A-48 Estimate of effective stress friction angle at Alabama site using in-situ test 

data. A-68 

A-49 Rock core log. ............................................................................................................ A-71 

A-50 Unconfined compression test results (sample depth 18 ft). ....................................... A-73 

A-51 Unconfined compression test results (sample depth 40 ft). ....................................... A-73 

A-52 Direct shear test results on rock joint (sample depth 74 ft). ...................................... A-73 

A-53 Boring Log UHSPT1. ................................................................................................ A-76 

A-54 Summary test data from samples across site.............................................................. A-77 

A-55 Atterberg limits results for soil samples. ................................................................... A-77 

A-56 Results of CPT measurements at site......................................................................... A-78 

A-57 Classification of soils using CPT derived parameters. .............................................. A-79 

A-58 Results of DMT measurements at site. ...................................................................... A-80 

A-59 Classification of soils using DMT derived parameters.............................................. A-81 

A-60 Evaluation of Ccε, Crε, and σp ′ using oedometer test results for sample at 

depth of 7 m. .............................................................................................................. A-82 

A-61 σp′ and OCR with depth from laboratory and in-situ tests. ....................................... A-83 

A-62 Calculated Ko with depth from in-situ tests. .............................................................. A-84 

A-63 Ko correlated with KD from DMT data (Kulhawy and Mayne, 1990)....................... A-85 

A-64 Ko correlated with qt from CPT data (Kulhawy and Mayne, 1990). ......................... A-85 

A-65 Undrained shear strength, su, from laboratory and in-situ testing.............................. A-86 

A-66 Guide to expansion and collapse potential (adapted from Holtz and Kovacs, 

1986). ......................................................................................................................... A-88 

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Figure Page

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B-1 Casagrande logarithm of time method for cv calculation..............................................B-1 

B-2 Taylor’s square root of time method for cv calculation. ...............................................B-3 

C-1 Comparison of measured and predicted piezocone dissipation data. ...........................C-2 

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LIST OF SYMBOLS 

 xxi 

A activity 

a cone radius 

AC corrected area for cone penetrometer 

Ac percentage of ash from organic content test 

an net area ratio for cone penetrometer 

As surface area of cone sleeve 

AR area ratio 

B Skempton’s pore pressure parameter 

Bq normalized pore pressure parameter for soil classification using CPTu 

c' effective stress cohesion intercept 

c T total stress cohesion intercept 

c′i instantaneous cohesion 

CB borehole diameter correction factor 

Cc compression index 

Ccε modified compression index or compression ratio 

CE energy correction factor 

CF clay fraction 

ch horizontal coefficient of consolidation 

CN SPT stress normalization parameter 

CP collapse potential 

CR rod length correction factor 

Cr recompression index 

Crε modified recompression index or recompression ratio 

CS sampling method correction factor or swelling index 

cv vertical coefficient of consolidation 

Cα coefficient of secondary compression 

Cαε modified coefficient of secondary compression 

D initial diameter of pressuremeter or vane diameter or distance between platens in a 

point-load strength test 

d10 effective grain size at 10% passing by weight 

d15 effective grain size at 15% passing by weight 

ds equivalent particle diameter 

De diameter at sampler cutting tip 

Di inside diameter of sampling tube 

Do outside diameter of sampling tube 

Dr relative density 

E Young’s modulus 

e void ratio 

ED dilatometer modulus 

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LIST OF SYMBOLS (continued) 

 xxii 

Ed deformation modulus of rock from borehole dilatometer 

EM rock mass modulus 

Emax, Eo small-strain Young’s modulus 

eo in-situ void ratio 

EP pressuremeter elastic modulus 

ER energy ratio 

ER intact rock deformation modulus 

Es equivalent elastic modulus 

Eu-r pressuremeter elastic modulus evaluated using unload-reload cycle 

FR friction ratio 

fs sleeve friction 

G shear modulus 

Gd interpreted shear modulus from borehole dilatometer test 

Gmax,Go small-strain shear modulus 

Gs specific gravity 

H vane height 

HDR drainage height 

Ho initial height 

I displacement influence factor 

i dilation angle 

iB angle of vane taper at bottom 

iT angle of vane taper at top 

I P, PI plasticity index 

ICR inside clearance ratio 

ID material index for dilatometer or slake durability 

IJ Jar Slake index 

IL incremental load 

Ir undrained rigidity index 

Is point load strength 

Is(50) size-corrected point load strength index 

JCS compressive strength of rock adjacent to the fracture surface 

JRC joint roughness coefficient 

k hydraulic conductivity 

KD horizontal stress index 

KE kinetic energy 

kh horizontal hydraulic conductivity 

km calibration pressure-dilation curve (for borehole dilatometer) 

Ko at-rest lateral earth pressure coefficient 

kPLT size correction factor for point load strength test 

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LIST OF SYMBOLS (continued) 

 xxiii 

kR stiffness of rock (for borehole dilatometer) 

ks stiffness of hydraulic system (for borehole dilatometer) 

kT stiffness of overall system plus rock (for borehole dilatometer) 

L length of cell membrane in borehole dilatometer 

LI liquidity index 

LL liquid limit 

Ms mass of soil solids 

mv volumetric compressibility coefficient 

N uncorrected Standard Penetration Test (SPT) blow count 

N60 standardized SPT blow count 

(N1)60 normalized and standardized corrected blow count 

Nmeas field recorded N-value 

Nk cone factor 

OCR overconsolidation ratio 

P point load breaking strength 

p' mean normal effective stress 

p1 expansion pressure 

Pa atmospheric pressure 

PE potential energy 

pf yield pressure 

Pi applied pressure 

pL limit pressure 

PL plastic limit 

po liftoff pressure or complete recompression pressure 

pr yield point 

pu minimum pressure during unloading in a pressuremeter unload/reload cycle 

q applied surface stress 

qc uncorrected cone tip resistance 

qt corrected cone tip resistance 

qu unconfined compressive strength 

RP uninflated radius of pressuremeter probe 

RT temperature correction factor 

S normally consolidated undrained strength ratio 

scollapse settlement due to collapse 

SL shrinkage limit 

sr remolded undrained shear strength 

St sensitivity 

su undrained shear strength 

suCIUC undrained shear strength from isotropically consolidated triaxial compression tests 

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LIST OF SYMBOLS (continued) 

 xxiv 

suDSS undrained shear strength from direct simple shear tests 

suVST undrained shear strength from vane shear test 

suUC undrained shear strength from unconfined compression tests 

suUU undrained shear strength from triaxial unconsolidated compression tests 

t time or vane edge thickness 

T torque and/or temperature 

T* modified time factor 

t100 time required to complete primary consolidation 

t50 time required to complete 50 percent primary consolidation 

t90 time required to complete 90 percent primary consolidation 

tf minimum time required to fail a sample 

Tmax maximum torque measured in vane shear test 

Tnet maximum torque minus rod friction torque 

tp time of primary consolidation 

Trod rod friction torque 

ts shear wave arrival times 

U degree of consolidation or normalized excess pore pressure 

u1 pore pressure at cone tip (CPTu1) 

u2 pore pressure at cone shoulder (CPTu2) 

ui initial pore pressure at t=0 

um measured pore pressure during CPT 

uo hydrostatic pore pressure or in-situ pore pressure condition 

ut pore pressure at time t 

Vm mean volume of pressuremeter probe 

Vo initial volume of pressuremeter probe 

Vs shear wave velocity or volume of soil solids 

wi initial moisture content 

wn natural moisture content 

zm gage offset zero reading in a flat plate dilatometer 

zw depth of water table 

%C percent collapse 

1-U normalized excess pore pressure 

α angle of taper 

γ unit weight 

γb buoyant unit weight 

γd, γdry dry unit weight 

γs shear strain 

γsat saturated unit weight 

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LIST OF SYMBOLS (continued) 

 xxv 

γt, γtot moist (total) unit weight 

∆A calibration factor for applied suction of membrane in air 

∆B calibration factor for applied expansion of membrane in air 

∆e change in void ratio 

∆Hc change in height of soil upon wetting 

∆Rm increase in radius of pressuremeter probe 

∆u change in pore pressure 

∆W change in work per unit volume 

δ deformation (of circular membrane) 

•

δ maximum rate of displacement 

δf displacement required to achieve peak strength of soil 

δn normal displacement 

δs shear displacement 

ε axial strain 

εf strain at end of increment 

εi strain at beginning of increment 

εp axial strain to reach peak strength 

εV vertical strain 

μ vane shear correction factor 

ν Poisson’s ratio 

νR Poisson’s ratio of rock 

ρ pump constant for borehole dilatometer 

ρd, ρdry dry mass density 

ρt, ρtot moist (total) mass density 

σ' effective stress 

σ1' major principal effective stress 

σ3' minor principal effective stress 

σc uniaxial compressive strength of rock or consolidation pressure 

σf stress at end of strain increment 

σho' in-situ horizontal effective stress 

σi stress at beginning of strain increment 

σn normal stress 

σn' effective normal stress 

σp' preconsolidation stress 

σv' vertical effective stress 

σvo total vertical stress 

σvo' in-situ vertical effective stress 

τ shear stress 

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LIST OF SYMBOLS (continued) 

 xxvi 

τmax peak shear strength 

τr residual strength 

φ friction angle 

φ ' effective stress friction angle 

φ i' instantaneous friction angle 

φ r' residual friction angle 

φ s' secant friction angle 

φ T total stress friction angle 

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1

CHAPTER 1 

INTRODUCTION TO 

GEOTECHNICAL ENGINEERING CIRCULAR ON 

SOIL AND ROCK PROPERTIES 

1.1 INTRODUCTION 

The purpose of this document is to provide the practicing highway design professional with specific 

recommendations regarding the appropriate methods to obtain engineering properties of soil and 

rock materials that may be encountered during construction of a wide range of transportation 

facilities. The target audience for Geotechnical Engineering Circular (GEC) No. 5 includes 

primarily geotechnical engineers and civil engineers who are responsible for establishing subsurface 

exploration programs and directing/overseeing the field and/or laboratory testing programs. The 

document is equally intended for structural engineers, engineering geologists, or geologists who may 

be responsible for these programs. The document is written to provide both general and specific 

information regarding the assessment of soil and rock properties to a potentially diverse audience. 

GEC No. 5 was developed in a very specific manner due in large part to the diverse background of 

potential readers. It is recognized that every civil engineer, and many geologists, have had classes 

related to this subject and there are several excellent textbooks and Federal Highway Administration 

(FHWA)-sponsored courses, demonstration projects, and guidance manuals dealing with the subject 

matter. Additionally, because site investigations and laboratory testing are critical steps in virtually 

all highway projects, every state Department of Transportation (DOT) has had to deal with the 

subject and resolve issues related to soil and rock properties on a day-to-day basis. Therefore, it is 

anticipated that almost every reader has an opinion, or at least an experience, regarding the subject of 

soil and rock properties. The purpose of this document is to provide each reader, regardless of their 

level of experience, with guidance related to appropriate techniques for evaluating soil and rock 

properties. 

1.2 BACKGROUND 

Textbooks and FHWA courses and documents are invaluable in terms of providing general 

recommendations concerning site investigation and laboratory testing. Basic information regarding 

how many holes to advance, how many samples to obtain, and how to conduct specific laboratory 

tests can be found in these sources. The one element missing from most of these documents/courses 

is a rationale for the “judgment” that must be applied to decide which specific tests are going to 

provide the specific information needed for a specific project and how to appropriately interpret 

these test results. Without judgment, the site exploration and field/laboratory testing phases can 

become prescriptive (e.g., advance ten boreholes to refusal, spaced at 60-m centers obtaining 

Standard Penetration Test (SPT) samples at 1.5-m vertical intervals). This approach may miss the 

fact that previous explorations in the area identified a weak 0.6-m thick clay layer at a depth of 5 m 

below ground surface that will control the stability of the proposed embankment constructed over the 

deposit. Therefore, there is a need to assist the design professional in making rationale decisions 

Archived Document
2

when encountering real (as opposed to textbook) soil deposits. The designer must make decisions 

regarding: (1) the layout and organization of the subsurface exploration program; (2) the type of in 

situ or laboratory test to run on the encountered material; and (3) the proper techniques to employ in 

interpreting multiple test results and developing design parameters. 

In reviewing the text and course materials that are currently available, it is recognized that two 

important areas are not addressed. First, over the past 15 to 20 years, several in situ testing 

techniques have moved from the arena of university research to routine engineering practice. Today, 

in situ testing plays a critical role in assessing soil properties and, to a lesser extent, rock properties, 

particularly by complementing laboratory-derived data. The designer needs to understand what in 

situ tests are applicable for specific soil deposits and specific soil properties. The second area that is 

not addressed in current literature is that the data resulting from the range of laboratory and in situ 

tests are often not completely consistent with other data obtained for the project and/or soil deposit. 

The designer needs to develop a rationale for accepting or rejecting data and for resolving potential 

inconsistencies between data provided by different laboratory and field tests. 

This document was prepared to provide the design professional with tools to assist in the rational 

development of subsurface investigation as well as in the execution and interpretation of laboratory 

and field testing programs. This document attempts to provide this rationale for the entire range of 

materials potentially encountered by a state DOT. This includes soft clay to intact rock and all 

variations of material between these extremes. 

1.3 DOCUMENT ORGANIZATION 

To assist the reader, the rationale for evaluating soil and rock properties has been organized into 

eight specific chapters. After this initial Chapter 1 that provides a general introduction to the entire 

document, the remainder of the GEC No. 5 is organized as follows: 

• Chapter 2 provides an overview of the remainder of the document by describing an FHWArecommended “process” for obtaining soil and rock properties. 

• Chapter 3 describes the rationale and procedures for developing a subsurface investigation 

program in soil and rock deposits. 

• Chapter 4 provides a general introduction and description of field and laboratory tests, 

focusing on specific tests, the resulting data from the tests, and the limitations of the tests. 

• Chapter 5 describes the procedures for selecting specific tests for soil and for interpreting 

the resulting data from these tests. 

• Chapter 6 describes rock mass classification and the procedures for selecting specific tests 

for rock and rock masses and for interpreting the resulting data from these tests. 

• Chapter 7 provides a discussion on special materials (e.g., expansive soils, loess, organic 

materials, colluvium, talus, and degradable materials) that could be encountered by a design 

professional. 

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3

• Chapter 8 provides a brief summary of techniques available to the engineer to resolve 

inconsistencies in test results from different tests and to appropriately select design values. 

The final portion of the document presents three Property Selection Examples in which the concepts 

identified in Chapters 2 through 8 are put into practice. For each of the examples, a “real” soil 

profile is used and the example provides a step-by-step approach in conducting the subsurface 

investigation program, performing laboratory and field tests, and interpreting the test results to 

obtain recommended soil and rock properties that may be required for typical highway related 

designs. 

A few final notes are warranted with regards to the document. It is expressly noted that the 

document presents a summary of the state of-the-practice as of the document date for a wide range 

of subsurface conditions. As such, many of the relatively “new” investigation and testing techniques 

that are currently available to DOTs, particularly in situ testing methods, have been included in the 

document. The document provides specific recommendations regarding the best tests to run for a 

given soil and the proper method to interpret these tests. By providing a rationale for obtaining 

realistic and appropriate soil and rock properties, it is anticipated that an “improved” or a “better” 

understanding and estimation of these properties will result relative to historical practices. The 

designer must assess how these properties affect the selected design, as many of the current design 

methods are semi-empirical and based largely on the historical (i.e., not the improved) techniques for 

evaluating soil and rock properties. Finally, it is intended that this document be read cover-to-cover, 

more as a “novel” than as a “cookbook.” In the latter case, the reader may desire specific 

information regarding a specific test and therefore can read a few pages from the text to secure an 

answer. With this document, recommendations are made as explicit paragraphs and sometimes the 

rationale is explained as part of the overall philosophy. Therefore, it is intended that the reader 

initially read the document cover-to-cover. Later, reference to specific sections and to 

recommendations can be made, but this later reference is provided once the spirit of the entire 

document has been understood. Although the authors have attempted to address the widest range of 

potential conditions encountered by the designer, there will certainly be cases that are not addressed 

in this document. By applying the philosophy and extrapolating the recommendations to these new 

and potentially more complex scenarios, the authors believe the reader will be able to use this 

document for even the most complex highway design applications. 

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4 

CHAPTER 2 

THE PROCESS FOR SOIL AND ROCK PROPERTY SELECTION 

2.1 INTRODUCTION 

Most geotechnical engineering textbooks provide information related to the mechanics of conducting 

field and laboratory tests to obtain soil and rock properties. In addition, the American Society for 

Testing and Materials (ASTM) and American Association of State Highway and Transportation 

Officials (AASHTO) standards provide excellent guidance related to the specific procedures for 

performing the actual field and laboratory tests. There are, however, few resources that provide 

guidance to the design professional related to a rational process for selecting appropriate critical 

locations in the geologic deposit and then developing a specific laboratory and field testing program 

to obtain soil and rock properties appropriate for design. The goal of this chapter is to describe a 

“process” that has been used on a variety of large and small projects to integrate the various decision 

steps necessary to arrive at the final design parameters. In addition to describing this step-by-step 

formal process, guidance is provided on the: (1) appropriate use of correlations to aid in engineering 

property selection; and (2) use of the Observational Method to refine and improve selected soil and 

rock properties used in design. 

2.2 PROCESS OF SOIL AND ROCK PROPERTY SELECTION 

A rational approach for selecting soil and rock properties for engineering design can be summarized 

as a logical twelve-step procedure that encompasses the general activities of site investigation and 

field testing, laboratory testing and interpretation, and engineering design. This step-by-step process 

is presented on the flow chart in figure 1. A brief description of each step of this process is 

presented below. More extensive discussion and the methods used to implement these steps are 

provided throughout the remainder of this document. 

Site Investigation and Field Testing

• Review Available Information: The best place to start the process of material property 

selection is to review any and all information that may be available. There are several 

sources for this information, many of the sources being in the public domain and readily 

available at modest expense. 

• Identify Required Material Properties: No investigation should be initiated without specific 

goals being established that are related to design and construction issues that must be 

considered (i.e., performance requirements), engineering properties that are needed, and the 

type of structure that is to be constructed. 

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Review available subsurface information and develop preliminary model of subsurface conditions

Conduct laboratory testing

Site Investigation and Field Testing

Laboratory Testing and

Test Interpretation

Engineering

Design

Identify material properties required for design and constructability and estimate scope of field program 

Plan site exploration and field test program

Conduct field investigations and field testing

Perform sample descriptions and laboratory index tests

Summarize basic soil/rock data and develop subsurface profile

Are results 

consistent with 

preliminary 

model?

Review design objectives and initial results

Are there additional 

data needs

Select representative soil/rock samples and details of laboratory testing 

Yes

No

Review quality of laboratory test data and summarize

Select material properties and finalize subsurface model

Are results 

consistent and valid

Is a Phase II 

Investigation

necessary?

Yes

No

Perform design and consider constructability issues

Phase II Investigation (if needed)

Yes

No

Yes

No

Figure 1. Soil and rock property selection flowchart. 

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• Plan Site Investigation: Historical information, which will provide anticipated subsurface 

conditions, coupled with knowledge of the specific design will allow an efficient sitespecific investigation strategy to be developed. Contingency plans should be considered 

based on anticipated variabilities in subsurface conditions. Sampling intervals should be 

identified and an in situ testing program should be developed. 

• Conduct Site Investigation and Field Testing: Once the investigation strategy is developed, 

it is ready to implement. Findings should be communicated to the geotechnical design 

engineer during the field work and modifications to the number and types of samples and 

testing should be determined, as required. 

• Describe Samples: Results from the field investigation program and subsequent laboratory 

identification of samples should be compared to the anticipated conditions based on 

historical information. Selected laboratory samples can be reviewed by the design engineer 

to obtain first-hand observations. These samples should be used for performing simple 

laboratory index tests. 

• Develop Subsurface Profile: Using results from the field investigation and the laboratory 

index tests, a detailed subsurface profile should be developed by the geotechnical design 

engineer. It is helpful at this step to review the initial site investigation objectives and 

expectations to be assured that the materials are consistent with expectations. 

• Review Design Objectives: An on-going evaluation of field and available laboratory data 

relative to the design objectives should be performed during the implementation of the site 

investigation. If adjustments are needed or if additional data needs are identified, 

procedures should be initiated to obtain the necessary information. 

Laboratory Testing and Test Interpretation

• Select Samples for Performance Testing: Prior to initiating the project-specific laboratorytesting program, the design engineer should review the recovered samples and confirm the 

testing that needs to be conducted (i.e., type, number, and required test parameters). If 

possible, selected samples should be extruded in the laboratory and reviewed by the design 

engineer. 

• Conduct Laboratory Testing: Once the samples have been reviewed and the testing program 

is confirmed, it is time to continue the index tests and initiate the performance-testing 

program (with index test correlation for quality assurance). Preliminary results should be 

provided to the design engineer for review. 

• Review Quality of Laboratory Data: If the data and interpreted laboratory test results are not 

consistent with expectations or if results indicate that the sample was disturbed, it is 

necessary to review progress and make adjustments. On some projects, results at this stage 

can be used to plan and initiate a more detailed and focused phase of investigation. A 

phased investigation approach is particularly helpful on large projects and in cases where 

there are many unknowns regarding the subsurface conditions or specific project 

requirements prior to conducting the proposed site investigation program. 

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• Select Material Properties: The laboratory and field test results should be interpreted and 

compared to project expectations and requirements. The role of the design engineer at this 

stage is critical as the full integration of field and laboratory test results must be coupled 

with the site-specific design. If test results are not completely consistent, the reason(s) 

should be evaluated, poor data should be eliminated, and similarities and trends in data 

should be identified. It may be necessary to return to the laboratory and conduct an 

additional review of sample extrusion, selection, and testing. 

Engineering Design

• Perform Design: At this final stage, the design engineer has the necessary information 

related to the soil and rock properties to complete the design. Additionally, the design 

engineer also has first-hand knowledge related to the variability of the deposit and of the 

material properties. Design activities can proceed with knowledge of these properties and 

variabilities. 

As referenced in chapter 1, this process is logical and is generally followed on many projects. In 

many cases, however, old “rules-of-thumb” and “status quo” approaches can result in an 

unconscious “by-passing” of critical steps. In particular, selection of the correct engineering 

property tests, their interpretation, and summarization are often poorly performed. Rigorous 

attention to this twelve-step procedure is required to assure efficient and thorough investigation and 

testing programs, especially since many projects are fragmented in which drilling, testing, and 

design are performed by different parties. 

2.3 USE OF CORRELATIONS TO ASSIST PROPERTY SELECTION 

If time and budget were not an issue, the design engineer could obtain as many samples as necessary 

and conduct as many laboratory or in situ tests as desired to obtain a complete assessment of 

subsurface soil and rock conditions. Engineering properties could be quantified and any inconsistent 

data could be set aside; additional testing could then be initiated. Unfortunately, time and budgets 

are major issues and the design engineer must make critical decisions at several steps throughout the 

design to obtain the most reliable and realistic soil and rock property information. As described 

previously, a critical step in obtaining these properties lies in the selection of a specific test and the 

interpretation of the test results. For any number of reasons (e.g., cost, sampling difficulties, etc.), it 

may be difficult to obtain the specific parameter(s) of interest. Fortunately, the design engineer can 

often use well-developed and/or site-specific correlations to obtain the desired parameter. Also, 

correlations serve as a quality assurance check on determined test results. 

Correlations to engineering properties come in many forms, but all have a common theme; 

specifically, the desired correlation utilizes a large database of results based on past experience. In 

the best case, the correlation and experience have been developed or “calibrated” using the specific 

local soil; in other cases the correlation may be based on reportedly similar soils. The reliance or use 

of correlations to obtain soil and rock properties is justified and recommended in the following 

cases: (1) specific data are simply not available and are only possible by indirectly comparing to 

other properties; (2) a limited amount of data for the specific property of interest are available and 

the correlation can provide a complement to these limited data; or (3) the validity of certain data is in 

question and a comparison to previous test results allows the accuracy of the selected test to be 

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assessed. Correlations in general should never be used as a substitute for an adequate 

subsurface investigation program, but rather to complement and verify specific project-related 

information. Examples of each of the three cases follows: 

• Specific Data are Unavailable: Several examples of this type exist. Most notable is the 

strength of uncemented clean sands. Undisturbed sampling is prohibitively expensive and 

correlations to Standard Penetration Testing (SPT), Cone Penetration Testing (CPT), and 

other in situ tests results have been shown to be quite reliable. As another example, suppose 

that the strength of a soil in triaxial extension is desired and only triaxial compression data 

are available. By reviewing previous published comparisons of compression and extension 

test results for similar soils, it is possible to approximate the triaxial test extension test 

results. 

• Limited Data are Available: For a given application, suppose that only a few, high-quality 

consolidation tests were performed. Compression properties were found to correlate well 

with Atterberg limits testing results. It is therefore concluded that additional consolidation 

test results are not required and that numerous Atterberg limits tests can be used to 

confidently assess compression properties. 

• Assessing Data Validity: Consider that results from tests on two similar soils are 

inconsistent. By comparing the results to those for similar soils it may be possible to 

identify whether the data are simply inconsistent of if some of the data are incorrect. 

There are several sources of correlation data for a range of geotechnical materials and properties. 

Many geotechnical textbooks and reference manuals include correlations as part of the text (e.g. 

Holtz and Kovacs, 1981; NAVFAC, 1982). The Electric Power Research Institute (EPRI) (Kulhawy 

and Mayne, 1990) commissioned the preparation of a very useful document that includes several 

correlations for laboratory and in situ tests. 

Regardless of the specific correlation, the following critical components need to be explicitly 

recognized: 

• The selected correlation is only as good as the data used to develop the correlation. Many 

correlations for sands were developed for clean, uncemented, uniform sands primarily for 

assessing liquefaction potential. Be careful in using this correlation to assess properties in a 

well-graded silty sand deposit. Select the appropriate correlation carefully. 

• A correlation provides an “approximate” answer and will undoubtedly exhibit scatter among 

the data points. Assess the data and the scatter by using upper and lower bound (i.e., best 

case/worst case) scenarios in the design calculations. 

• The selected correlation will be most accurate if “calibrated” to local soil conditions. Many 

state DOTs have developed useful correlations based on specific project experience in their 

state. 

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2.4 USE OF OBSERVATIONAL METHOD 

Once an appropriate value for the design has been selected, it is possible to complete the design and 

proceed to construction. There is one final step that can be performed to validate the data and 

possibly improve the accuracy of the selected value. This three step process involves: (1) using the 

design value and the actual estimated loading to predict a field response; (2) systematically monitor 

the field performance; and (3) “back calculate” the actual property of interest. This process of 

prediction, monitoring, and reassessment is known as the Observational Method (Terzaghi and Peck, 

1967). Two examples of this technique follow: 

• In soil, the use of piezometers to monitor the rate of pore pressure dissipation and measured 

settlements of a large area fill can result in a more accurate estimate of the compressibility 

and time rate of consolidation characteristics of soft soils as well as provide information to 

maximize the rate of fill placement. 

• In rock, the use of instrumented rock bolts and displacement monitoring instrumentation can 

provide valuable information regarding the kinematics of block stability and the strength of 

the jointed rock mass. 

The Observational Method is an invaluable aid and ideally should be a part of every geotechnical 

project. Sadly, this approach is often overlooked due to budget concerns, and in many cases is not 

even considered by the design engineer. Where appropriately used, the Observational Method can 

have significant benefits not only to the project at hand, but also for other projects in the area 

because a full-scale assessment of the engineering properties can be made. It is strongly 

recommended that the Observational Method be included as a basic tenet of all projects. 

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CHAPTER 3 

PLANNING A SUBSURFACE INVESTIGATION AND 

LABORATORY TESTING PROGRAM 

3.1 INTRODUCTION 

To evaluate soil and rock properties required for geotechnical design related to transportation 

projects, subsurface investigation and laboratory testing programs are developed and executed. The 

data collection efforts associated with these activities should occur early in the project; failure to 

conduct an appropriately scoped investigation and laboratory-testing program will result in potential 

data gaps and/or the need to re-mobilize to the site for supplementary testing. Data gaps can cause 

significant delays in the project and can potentially lead to either an overconservative and costly 

design or an unconservative and unsafe design. It is, therefore, imperative that the subsurface and 

laboratory testing programs be carefully planned to ensure that the information collected in the field 

and the laboratory will be sufficient to develop soil and rock properties for design and construction. 

This chapter will present general guidelines related to the development of subsurface investigation 

and laboratory testing programs for the evaluation of soil and rock properties. Since the selection of 

sampling and testing methods will be driven by the scope of the project and geologic conditions, 

critical project related issues must be understood prior to field and laboratory planning activities. 

For heterogeneous deposits and special materials (i.e., colluvium, organic soils, etc.) greater effort 

will be required during the planning stage of the investigation to assess the applicability of specific 

tools and sampling devices. 

3.2 PLANNING THE SUBSURFACE INVESTIGATION AND LABORATORY 

TESTING PROGRAM 

3.2.1 General 

Planning subsurface investigation and laboratory testing programs requires the engineer to be aware 

of parameters and properties needed for design and construction, as well as to understand the 

geologic conditions and site access restrictions. Specific steps include: (1) identify data needs; (2) 

gather and analyze existing information; (3) develop a preliminary site model; (4) develop and 

conduct a site investigation; and (5) develop and conduct a laboratory-testing program. Specific 

planning steps are addressed in the following sections. 

3.2.2 Identify Data Needs 

The first step of an investigation and testing program requires that the engineer understand the 

project requirements and the site conditions and/or restrictions. The ultimate goal of this phase is to 

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identify geotechnical data needs for the project and potential methods available to assess these 

needs. During this phase it is necessary to: 

• identify design and constructability requirements (e.g., provide a grade separation, transfer 

loads from bridge superstructure, provide for a dry excavation); 

• identify performance criteria (e.g., limiting settlements, right of way restrictions, proximity 

of adjacent structures) and schedule constraints; 

• identify areas of concern on site and potential variability of local geology; 

• develop likely sequence and phases of construction; 

• identify engineering analyses to be performed (e.g., bearing capacity, settlement); 

• identify engineering properties and parameters required for these analyses; 

• evaluate methods to obtain parameters and assess the validity of such methods for the 

material type and construction methods; and 

• evaluate number of tests/samples needed and appropriate locations for them. 

As an aid to assist in the planning of site investigation and laboratory testing, table 1 provides a 

summary of the information needs and testing considerations for various geotechnical applications. 

A discussion of specific field and laboratory test methods is provided in chapter 4. 

3.2.3 Gather and Analyze Existing Information 

Before any equipment is mobilized to the site, existing data for the site, both regionally and locally, 

should be evaluated as a logical first step in the investigation. This is an important and inexpensive 

step that is often overlooked. There are many readily available data sources that can be used to 

identify major geologic processes that have affected the site, site history, geologic constraints, manmade features, and access issues. The planning step can be extremely cost effective and productive. 

Existing data will provide information which can reduce the scope of the subsurface investigation, 

help guide the location of testing and sampling points, and reduce the amount of time in the field due 

to unexpected problems. For example, historical aerial photographs can be used to identify an area 

where fill had been placed, where a landslide scarp exists, or major geologic structures such as 

faults, bedding planes, and continuous joint sets. A list of potential information sources along with 

the type of information available is presented in table 2. 

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Table 1. Summary of information needs and testing considerations for a range of highway applic

Geotechnical 

Issues 

Engineering 

Evaluations 

Required Information 

for Analyses 

Field Testing(1) 

Shallow Foundations

• bearing capacity • settlement (magnitude & rate) • shrink/swell of foundation soils 

(natural soils or embankment fill) 

• chemical compatibility of soil and 

concrete 

• frost heave • scour (for water crossings) • extreme loading 

• subsurface profile (soil, groundwater, rock) • shear strength parameters • compressibility parameters (including consolidation, 

shrink/swell potential, and elastic modulus) 

• frost depth • stress history (present and past vertical effective 

stresses) 

• chemical composition of soil • depth of seasonal moisture change • unit weights • geologic mapping including orientation and 

characteristics of rock discontinuities 

• vane shear test • SPT (granular soils) • CPT • dilatometer • rock coring (RQD) • nuclear density • plate load testing • geophysical testing 

• • • • • • • • • • • • 

Driven Pile Foundations

• pile end-bearing • pile skin friction • settlement • down-drag on pile • lateral earth pressures • chemical compatibility of soil and 

pile 

• driveability • presence of boulders/ very hard 

layers 

• scour (for water crossings) • vibration/heave damage to nearby 

structures 

• extreme loading 

• subsurface profile (soil, ground water, rock) • shear strength parameters • horizontal earth pressure coefficients • interface friction parameters (soil and pile) • compressibility parameters • chemical composition of soil/rock • unit weights • presence of shrink/swell soils (limits skin friction) • geologic mapping including orientation and 

characteristics of rock discontinuities 

• SPT (granular soils) • pile load test • CPT • vane shear test • dilatometer • piezometers • rock coring (RQD) • geophysical testing 

• • • • • • • • • • • • • 

(1) Corresponding AASHTO and ASTM Standard references are provided in chapter 4 for field and laboratory tests. 

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Table 1. Summary of information needs and testing considerations for a range of highway applications 

Geotechnical 

Issues 

Engineering 

Evaluations

Required Information 

for Analyses

Field Testing(1) L

Drilled Shaft Foundations 

• shaft end bearing • shaft skin friction • constructability • down-drag on shaft • quality of rock socket • lateral earth pressures • settlement (magnitude & rate) • groundwater seepage/ dewatering • presence of boulders/ very hard 

layers 

• scour (for water crossings) • extreme loading 

• subsurface profile (soil, ground water, rock) • shear strength parameters • interface shear strength friction parameters (soil and 

shaft) 

• compressibility parameters • horizontal earth pressure coefficients • chemical composition of soil/rock • unit weights • permeability of water-bearing soils • presence of artesian conditions • presence of shrink/swell soils (limits skin friction) • geologic mapping including orientation and 

characteristics of rock discontinuities 

• degradation of soft rock in presence of water and/or air 

(e.g., rock sockets in shales) 

• technique shaft • shaft load test • vane shear test • CPT • SPT (granular soils) • dilatometer • piezometers • rock coring (RQD) • geophysical testing 

• 1-D • tria• gra• inte• pH, • perm• Att• mo• unit • org• coll• rock 

and 

• poin• slak

Embankments and Embankment Foundations 

• settlement (magnitude & rate) • bearing capacity • slope stability • lateral pressure • internal stability • borrow source evaluation (available 

quantity and quality of borrow soil) 

• required reinforcement 

• subsurface profile (soil, ground water, rock) • compressibility parameters • shear strength parameters • unit weights • time-rate consolidation parameters • horizontal earth pressure coefficients • interface friction parameters • pullout resistance • geologic mapping including orientation and 

characteristics of rock discontinuities 

• shrink/swell/degradation of soil and rock fill 

• nuclear density • plate load test • test fill • CPT • SPT (granular soils) • dilatometer • vane shear • rock coring (RQD) • geophysical testing 

• 1-D • tria• dire• gra• Att• org• mo• hyd• geo• shri• slak• unit 

Excavations and Cut Slopes 

• slope stability • bottom heave • liquefaction • dewatering • lateral pressure • soil softening/progressive failure • pore pressures 

• subsurface profile (soil, ground water, rock) • shrink/swell properties • unit weights • hydraulic conductivity • time-rate consolidation parameters • shear strength of soil and rock (including 

discontinuities) 

• geologic mapping including orientation and 

characteristics of rock discontinuities 

• test cut to evaluate stand-up 

time 

• piezometers • CPT • SPT (granular soils) • vane shear • dilatometer • rock coring (RQD) • in situ rock direct shear test • geophysical testing 

• hyd• gra• Att• tria• dire• mo• slak• rock 

and 

• poin

(1) Corresponding AASHTO and ASTM Standard references are provided in chapter 4 for field and laboratory tests. 

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Table 1. Summary of information needs and testing considerations for a range of highway applications 

Geotechnical 

Issues 

Engineering 

Evaluations

Required Information 

for Analyses

Field Testing(1) 

Fill Walls/ Reinforced Soil Slopes

• internal stability • external stability • settlement • horizontal deformation

• lateral earth pressures• bearing capacity

• chemical compatibility with soil 

and wall materials

• pore pressures behind wall • borrow source evaluation (available 

quantity and quality of borrow soil) 

• subsurface profile (soil, ground water, rock) • horizontal earth pressure coefficients • interface shear strengths • foundation soil/wall fill shear strengths

• compressibility parameters (including consolidation, 

shrink/swell potential, and elastic modulus) 

• chemical composition of fill/ foundation soils

• hydraulic conductivity of soils directly behind wall• time-rate consolidation parameters• geologic mapping including orientation and 

characteristics of rock discontinuities

• SPT (granular soils) • CPT • dilatometer • vane shear • piezometers • test fill• nuclear density

• pullout test (MSEW/RSS) • rock coring (RQD) • geophysical testing 

• • • • • • • • • • 

Cut Walls • internal stability 

• external stability • excavation stability • dewatering • chemical compatibility of wall/soil • lateral earth pressure • down-drag on wall • pore pressures behind wall • obstructions in retained soil 

• subsurface profile (soil, ground water, rock) • shear strength of soil • horizontal earth pressure coefficients • interface shear strength (soil and reinforcement) • hydraulic conductivity of soil • geologic mapping including orientation and 

characteristics of rock discontinuities 

• test cut to evaluate stand-up 

time 

• well pumping tests • piezometers • SPT (granular soils) • CPT • vane shear • dilatometer • pullout tests (anchors, nails) • geophysical testing 

• • • • • • • • • 

(1) Corresponding AASHTO and ASTM Standard references are provided in chapter 4 for field and laboratory tests. 

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Table 2. Sources of historical site data. 

Source Functional Use Location Exam

Utility Maps 

• Identifies buried utility locations • Identifies access restrictions • Prevents damage to utilities 

Local agencies/utility companies 

Power line identification priinvestigation prevents extenexpensive repairs, and bodil

Aerial Photographs 

• Identifies manmade structures • Identifies potential borrow source areas • Provides geologic and hydrological 

information which can be used as a basis for site reconnaissance 

• Track site changes over time 

Local Soil Conservation Office, United States Geological Survey (USGS), Local Library, Local & National aerial survey companies 

Evaluating a series of aerial area on site which was filled reviewed 

Topographic Maps 

• Provides good index map of site area • Allows for estimation of site topography • Identifies physical features in the site area • Can be used to assess access restrictions 

USGS, State Geological Survey 

Engineer identifies access arareas of potential slope instacut/fill capacity before visiti

Existing Subsurface Investigation Report 

• May provide information on nearby 

soil/rock type; strength parameters; hydrogeological issues; foundation types previously used; environmental concerns 

USGS, United States Environmental Protection Agency (USEPA), State DOTs 

A five year old report for a nproject provides geologic, hgeotechnical information for scope of the investigation 

Geologic Reports and Maps 

• Provides information on nearby soil/rock 

type and characteristics; hydrogeological issues, environmental concerns 

USGS and State Geological Survey 

A twenty year old report on identifies rock types, fracturgroundwater flow patterns 

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Table 2. Sources of historical site data (continued). 

Source Functional Use Location Exam

Water/Brine Well Logs 

• Provide stratigraphy of the site and/or 

regional area 

• Varied quality from state to state • Groundwater levels 

State Geological Survey/Natural Resources 

A boring log of a water suppthe site area shows site stratevaluations of required dept

Flood Insurance Maps 

• Identifies 100 and 500 yr. floodplains near 

water bodies 

• May prevent construction in a floodplain • Provide information for evaluation of 

scour potential 

Federal Emergency Management Agency (FEMA), USGS, State/Local Agencies 

Prior to investigation, the flosite is in a 100 yr floodplain structure is moved to a new 

Soil Survey 

• Identifies site soil types • Permeability of site soils • Climatic and geologic information 

Local Soil Conservation Service 

The local soil survey providsurface soils to facilitate preevaluation 

Sanborn Fire Insurance Maps 

• Useful in urban areas • Maps for many cities are continuous for 

over 100 yrs. 

• Identifies building locations and type • Identifies business type at a location (e.g., 

chemical plant) 

• May highlight potential environmental 

problems at an urban site 

State Library/Sanborn Company (www.sanborncompany.com) 

A 1929 Sanborn map of St. smelter was on site for 10 yeprevents an investigation in 

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Of the information sources listed in table 2, utility maps, existing subsurface investigation reports in 

the area, geologic reports/maps, and soil surveys often provide extremely valuable information. 

Although historic reports may provide useful data, the information they contain should be used with 

caution. The data should be scrutinized as to its reliability by evaluating the consistency of the 

information and the overall quality of the report. Review of construction or performance monitoring 

records will provide helpful information regarding subsurface conditions and variability. On some 

projects, it may be worthwhile to develop a geologic summary report that provides information on 

regional and local geology, hydrogeology, and unstable areas (e.g., karstic areas, fault zones, etc.). 

3.2.4 Conduct Site Visit 

After review of the available data and prior to mobilizing investigation equipment to the project site, 

a site visit should be performed. During the site visit, the engineer should carefully observe all 

relevant physical features of the area and record detailed notes. These observations should include: 

• utility locations (overhead and underground); 

• access issues (e.g., location, width, and condition of all potential access roads; trees; power 

lines; buildings; right-of-way); 

• conditions of nearby structures (record location, type, and depth of existing structures and 

foundations); 

• geologic constraints (e.g., rivers, streams, bluffs, outcrops); 

• topographic conditions (e.g., ditches, hills, valleys); 

• soil/rock type (e.g., clay, sand, rock outcrops, conditions when wet); 

• surface conditions (e.g., desiccated surface, lack of vegetation, debris, ponded water, 

deposits of colluvium or talus, evidence of rock/soil slope failures); 

• geomorphic controls (e.g., landslides, floodplains, karst, erosional/depositional conditions); 

• flood levels/drainage issues; 

• zonation of rock mass with information on location, orientation, and type of boundary 

between zones of relatively uniform geology; 

• adjacent property use; and 

• potential borrow source areas (if applicable). 

If possible, the engineer should meet with local landowners to obtain information on springs, historic 

landslides, old mine workings, etc. and to obtain permission for gaining access to private property. 

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Potential drilling locations should be noted for subsequent subsurface investigation. A topographic 

map of the project site should be prepared prior to the site visit and the map should be used by the 

engineer during the site visit. The previously referenced observations should be noted on the 

topographic map and a preliminary layout of the subsurface investigations borings/soundings should 

be marked. Temporary field stakes can often be established for these locations during the site visit. 

Additional information can be gained by driving along roads in the general site area and observing 

roadcuts and rock outcrops. In hilly areas, roadcuts provide some of the most accessible and useful 

means of obtaining subsurface information prior to the use of intrusive investigation methods. Seeps 

along the face of a roadcut may provide an idea of the relative proximity of the water table to the 

ground surface and the structure of the subsurface geologic units. 

Information gathered during a site reconnaissance visit will aid in the selection of a drill rig, sampler 

type, boring/sampling locations, personnel safety needs, and potential problems which may preclude 

construction of certain geotechnical design elements. The site visit also allows the engineer to make 

an initial estimate of subsurface conditions, as well as an estimation of the time needed to complete 

the field investigation and testing program. Conducting the site visit after collecting and interpreting 

existing data allows for an efficient data collection effort. Figure 2 provides a checklist of 

information items that should be collected during the site reconnaissance stage of a project. 

Checklists and/or field reconnaissance forms are useful while conducting site visits to ensure that 

pertinent data are not forgotten or overlooked. A copy of the field reconnaissance report should be 

provided to the drill crew prior to drilling. 

3.2.5 Develop Preliminary Site Model 

A preliminary site model should be developed using the information obtained from existing data and 

the site visit. The preliminary model should consist of the soil and rock stratigraphy, potential site 

restrictions, and anticipated groundwater levels. The model should be divided into zones of interest 

(i.e., geotechnical units) based on the necessary design parameters and objectives. This model will 

obviously change as results of the detailed investigation are collected. 

While developing the preliminary site model, particular attention should be given to the possibility 

of encountering heterogeneities. In many geologic settings, these include boulders and significant 

variations in bedrock surface elevation, which may affect the design and investigation. If boulders 

are suspected at a site, unexpected “top of rock” elevations should be questioned and it may be 

necessary to drill several feet into “rock” in order to resolve this issue. In addition, soils containing 

large amounts of gravel or boulders may cause damage to intrusive in situ testing equipment or limit 

the feasibility of these test methods. 

Based on the information gathered thus far in the planning phase, it may be possible to eliminate or 

favor specific designs. For example, the preliminary site model may indicate that extensive boulder 

and gravel deposits are present at the site, which could eliminate the possibility of a driven pile 

design. Alternatively, the preliminary site model might show subsurface conditions to be 

particularly favorable to a proposed design. 

Archived Document
Figure 2. Checklist items for site reconnaissance 

19 

 Date  Prepared by:  Organization 

 ACCESSIBILITY 

 Easy  By Vehicle only  Difficult by car - Walk only  Requires 4-wheeled drive  Dozer and Grading Required  Inaccessible  Details 

 VISIT TO SITE 

 Date/Time of Day  Visitors  Weather Conditions 

 Sunny  Cloudy  Rain  Snow  Icy  Freezing 

 GROUND COVER 

 Asphalt  Grass  Flowers  Bushes  Trees  Forest  Soil  Gravel  Concrete  Rock Outcroppings  Evidence of fill/debris  Prior Construction  Existing Buildings  Roadways  Other 

EXISTING TERRAIN 

 Level Ground  Sloping Conditions  Gentle Dip  Steep  Hummocky  Rolling Hills  Mountainous 

 Other remarks 

 SITE HYDROLOGY 

 Dry - Barren  Desert  Surface Water Conditions 

 None  Swampy  Pond  Lake  Ocean  Stream  River 

 Subsurface Water 

 None  Not Obvious  Major Aquifer  Water Wells  Pumping from deep wells  Other Details 

SITE DRAINAGE

 Runoff Features 

 Erosion  Ponding  Waterfalls  Piping  Swale  Other 

 Natural 

 Excellent  Good  Fair  Poor 

 Artificial Drains 

 Stormwater System  Retention Pond  Vertical wick drains  Pumping Stations  Other 

SOIL AND ROCK CONDITIONS 

 Surface Soils 

 Topsoil  Presence of Fills  Evidence of Debris  Pollutants/Contaminants  Agrarian types/farming  Evidence of slope instability 

 Landslides/slips 

 Creep 

 Cracking  Scour  Heave  Subsidence 

 Cut/Quarry Operations  Fill/Borrow  Other 

 Subsurface Soils 

 USCS soil types: 

 GM, GC, GP, GW  SM, SC, SP, SW  CL, CH, ML, MH  Pt, OL, OH  Other 

 Surface Rocks 

 Loose cobbles  Boulders  Rock outcroppings 

 Type of rocks 

 Igneous  Sedimentary  Metamorphic  Details 

 Rock Features 

 Jointing Patterns  Faults  Discontinuities  Weathering  Planes of weakness  Evidence of talus  Karst/sinkholes  Caves  Other 

 INVESTIGATIVE OPERATIONS 

 Existing test pits  Existing boreholes  Cased holes  Blasting operations 

 Dynamite  ANFO  Rippers  Percussive Drills 

 Erratics/ boulders 

 Coreholes  Diamond drilling  Wireline drilling 

 Exploratory Adits 

 Vertical shafts 

Other 

 PR

U

 N

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3.2.6 Developing a Site Investigation Program 

For many projects and for many site conditions, the most difficult and crucial part of the planning 

phase involves the decisions regarding sampling/investigation method, boring locations, number of 

samples, number and types of laboratory tests, and the number of confirmatory samples. At this 

stage, the types of potential sampling/investigation methods should have been identified and 

assessed. During the site visit, it is often possible to eliminate certain techniques. Specific tools for 

sampling and testing are discussed in chapter 4 and include: 

• Undisturbed sampling: Thin walled tubes, block samples, rock core 

• Disturbed sampling: SPT, augering, non-core rock drilling 

• Non-intrusive: Geophysical and remote sensing methods 

• In-situ soil and rock testing: CPT, flat-plate dilatometer test (DMT), vane shear test (VST), 

pressuremeter, borehole dilatometer, in situ direct shear test, etc. 

Many engineers underestimate the importance in selecting an appropriate investigation technique(s) 

and sampling method(s), since in-house SPT capabilities are a presumptive “one size fits all” 

investigation/sampling approach (e.g., SPT sampling at 1.5-m intervals in borings on 60-m spacing). 

Experience has shown the short-falls of this approach and the benefit of a focused site-specific 

strategy. For example, in cases where extensive laboratory tests are not needed but it is necessary to 

gather large amounts of data concerning stratigraphy and subsurface material variability, intrusive 

non-sampling methods such as the CPT and DMT could be considered. The CPT method is quick 

and allows for stratigraphic mapping over a site area much more quickly and economically than the 

use of other intrusive methods. Additionally, data from the CPT can be input into stress history and 

strength correlations to obtain specific design parameters. When the penetrometer is fit with a pore 

pressure transducer (CPTu), detailed stress history and strength correlations for clays can be 

evaluated, estimates of groundwater elevation can be predicted, and flow characteristics can be 

assessed. The DMT can provide stratigraphic information similar to that from the CPT, but the 

DMT data are better suited for assessment of modulus values for settlement calculations. 

The number of borings and their locations in a site area will depend on the proposed structure, 

design parameters, access issues, geologic constraints, and expected stratigraphy and heterogeneity. 

Some minimum guidelines for boring spacing are provided in table 3. This table was developed 

based on a number of FHWA documents including Drilled Shafts: Construction Procedures and 

Design Methods (FHWA-IF-99-025), Design and Construction of Driven Pile Foundations (FHWAHI-97-013), and Advanced Technologies for Soil Slope Stability (FHWA-SA-94-005). This table 

should be used only as a first step in estimating the minimum number of borings for a particular 

design, as actual boring spacings will be dependent upon the project type and geologic environment. 

In areas underlain by heterogeneous soil deposits and/or rock formations, it will probably be 

necessary to exceed the minimum guidelines from table 3 to capture variations in soil and/or rock 

type and to assess consistency across the site area. For situations where large-diameter rockArchived Document
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socketed shafts will be used or where drilled shafts are being installed in karstic formations, it may 

be necessary to advance a boring at the location of each shaft. In a laterally homogeneous area, 

drilling or advancing a large number of borings may be redundant, since each sample tested would 

exhibit similar strength and compressibility properties. In all cases, it is necessary to understand 

how the design and construction of the geotechnical feature will affect the soil and/or rock mass in 

order to optimize the investigation. 

The engineer or geologist should plot the proposed boring locations on a site topographic map prior 

to initiation of drilling. They should review the notes taken during the site visit considering access 

restrictions and not arbitrarily or randomly select boring locations. Alternate boring locations should 

be considered and a contingency plan (e.g., move a maximum of 15 m from a boring location if 

unable to drill at a particular location) should be discussed in case a boring needs to be relocated due 

to access restrictions or unexpected geologic conditions. Field personnel unfamiliar with the 

objectives and rational behind the planning of the site investigation should maintain contact with the 

office during field activities and discuss issues such as the relocation of a boring. Arbitrary or 

random boring selection will increase the chances of boring relocation, confusion, and wasted time 

in the field. Final boring locations should be surveyed and recorded as part of the permanent project 

record. 

The guidelines for depth interval selection should also be developed in recognition of specific 

site/project conditions and design property/parameter requirements. For preliminary screening, 

disturbed samples might be taken continuously in the upper 3 m, at 1.5 m intervals up to 30 m, and 

possibly every 3 m at depths greater than 30 m. For characterization and assessment of design 

properties in fine-grained soils, a minimum of one undisturbed sample should be taken for each 

stratum, with additional samples taken at 3 to 6 m intervals with depth. Undisturbed samples may 

not need to be taken in each boring if the deposit is relatively homogeneous with closely spaced 

borings. These are general minimum guidelines and intervals may need to be increased depending 

upon the project requirements and site geologic conditions. The sampling interval may need to be 

increased when soil/rock conditions change frequently with depth; however, these changes need to 

be considered in the context of the design. Therefore, ongoing communication with the 

office/design engineer is absolutely essential. Once the site stratigraphy has been established, it may 

not be necessary to sample every time there is a change in stratigraphy if the changes have no impact 

on design. For example, it may not be necessary to sample alternating layers of coarse grained 

deposits where settlement is of concern, and for designs concerned with bearing capacity, although 

samples below the anticipated extent of the area influenced by the load may be reduced, samples 

should be obtained in case the type of foundation changes between preliminary and final design. 

The sampling interval will vary between individual projects and regional geologies. If soils are 

anticipated to be difficult to sample or trim in the lab due to defects, etc., the frequency of sample 

collection should be higher than average to offset the number of samples that may be unusable in the 

lab for performance property evaluation (e.g., shear strength). When borings are widely spaced, it 

may be appropriate to collect undisturbed samples in each boring. For closely spaced borings or in 

deposits of lateral uniformity, undisturbed samples may only be needed in select borings. If a thin 

clay seam is encountered during drilling and not sampled, the boring may need to be offset and redrilled to obtain a sample. 

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It is often quite helpful to combine in situ soundings using CPT, CPTu, DMT, or pressuremeter with 

conventional disturbed/undisturbed sampling. For example, by performing CPT or CPTu soundings 

prior to conventional drilling and sampling, it may be possible to target representative and/or critical 

areas where samples can be later obtained. This combination may reduce some of the potential 

drilling redundancy in heterogeneous environments. The use of geophysical methods can be used to 

provide useful information on ground in-between boring locations. 

For investigations for rock slopes and foundations, it is important to consider structural geology, in 

addition to information obtained as part of a rock-coring program. For example, the orientation and 

characteristics of a clay-filled discontinuity is critical information that will be used to judge whether 

a rock slope will be stable or unstable or whether a structural foundation will undergo minor or 

significant settlement. However, a detailed structural geologic assessment may provide enough 

information to significantly limit the scope of a rock-coring program. For example, drilling and 

coring may not be required where applied loads are significantly less than the bearing capacity of the 

rock, where there is no possibility of sliding instability in a rock slope, or where there are extensive 

rock outcrops from which information can be obtained to confidently establish the subsurface 

conditions for design and constructability assessments (Wyllie, 1999). 

As the reader can see from this discussion, it is difficult and really unnecessary to establish a 

prescriptive drilling, sampling, and testing protocol that is applicable to all sites. The engineer is 

most effective when: (1) applying these conventional guidelines with site and/or project-specific 

requirements/constraints; and (2) recognizing the advantages and limitations of sampling equipment 

and in-situ testing methods. 

3.2.7 Developing a Laboratory Testing Program 

The final planning step includes the development of a laboratory-testing program. Prior to planning 

the laboratory-testing program, it is again necessary to review the design needs of the project. 

Specifying unnecessary laboratory tests will add time and cost to the project and consume samples. 

Table 1 lists laboratory tests that may be applicable to specific designs. One can see from this table 

that moisture content, Atterberg limits, grain size distribution, and unit weight tests are 

recommended for most design applications in soils. As will be discussed in chapter 5, index tests 

are not specifically used in the design but are invaluable in establishing general conditions and 

assessing inherent material variabilities. Although this table provides useful guidelines, additional 

tests may be required or tests may be eliminated depending on individual site conditions. 

Once a list of necessary tests has been developed and the field program has been executed, the 

engineer should review field notes, borings, and design plans to identify “critical areas”. Critical 

areas correspond to borings/locations where the results of the laboratory tests could result in a 

significant change in the proposed design. Samples from these critical areas should be identified for 

performance testing. As will be discussed in chapter 5, performance tests provide design-specific 

parameters. For example, if unexpected clay deposits were discovered in the area of a proposed 

footing, 1-D consolidation testing would be required to assess potential settlements. If the test 

results show that settlements may be excessive under the proposed design, the footing may have to 

be moved, redesigned, or deep foundation support considered. Similarly, if the same footing was 

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proposed for use on a rock slope, the presence of an unfavorably oriented clay-filled discontinuity 

would likely make it necessary to obtain an undisturbed rock core sample for direct shear testing of 

the clay-filled discontinuity. The engineer should also consider the site area in context of the 

required design parameters and the samples available for laboratory analysis. It is necessary to 

select samples for laboratory analysis that will accurately characterize the site. In heterogenous 

areas, many samples may be required to obtain comprehensive parameters; in homogeneous areas, 

few samples may be required. 

A laboratory-testing program should be performed on representative and critical specimens from 

geologic layers across the site. To assess the locations where tests should be performed, it is useful 

to evaluate sample location maps and geologic cross sections. By evaluating a sample location map, 

it will be easy to quickly identify the locations of disturbed and undisturbed samples that may be 

used in the laboratory testing program. The generation of detailed subsurface cross sections (see 

section 5.2), including stratigraphy, in-situ testing results, and laboratory index test results, if 

available, will be useful when identifying representative samples for laboratory performance testing. 

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Table 3. Guidelines for minimum number of investigation points and 

depth of investigation. 

Application 

Minimum Number of Investigation Points 

and Location of Investigation Points Minimum Depth of Invest

Retaining walls A minimum of one investigation point for each 

retaining wall. For retaining walls more than 30 m in length, investigation points spaced every 30 to 60 m with locations alternating from in front of the wall to behind the wall. For anchored walls, additional investigation points in the anchorage zone spaced at 30 to 60 m. For soil-nailed walls, additional investigation points at a distance of 1.0 to 1.5 times the height of the wall behind the wall spaced at 30 to 60 m. 

Investigate to a depth below bottom of wtimes the wall height or a minimum oInvestigation depth should be great enousoft highly compressible soils (e.g. pefine grained soils) into competent mbearing capacity (e.g., stiff to hard cohdense cohesionless soil, or bedrock). 

Embankment Foundations A minimum of one investigation point every 60 m 

(erratic conditions) to 120 m (uniform conditions) of embankment length along the centerline of the embankment. At critical locations, (e.g., maximum embankment heights, maximum depths of soft strata) a minimum of three investigation points in the transverse direction to define the existing subsurface conditions for stability analyses. For bridge approach embankments, at least one investigation point at abutment locations. 

Investigation depth should be, at a minithe embankment height unless a hard strabove this depth. If soft strata is encoundepth greater than twice the eminvestigation depth should be great enouthe soft strata into competent materiacohesive soil, compact to dense cobedrock). 

Cut Slopes A minimum of one investigation point every 60 m 

(erratic conditions) to 120 m (uniform conditions) of slope length. At critical locations (e.g., maximum cut depths, maximum depths of soft strata) a minimum of three investigation points in the transverse direction to 

define the existing subsurface conditions for stability analyses. For cut slopes in rock, perform geologic mapping along the length of the cut slope. 

Investigation depth should be, at a miniminimum elevation of the cut unless encountered below the minimum eleInvestigation depth should be great enouthrough soft strata into competent matercohesive soil, compact to dense cobedrock). In locations where the baground-water level, increase depth needed to determine the depth of underly

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Table 3. Guidelines for minimum number of investigation points and 

depth of investigation (continued). 

Application Minimum Number of Investigation Points and 

Location of Investigation Points 

Minimum Depth of Invest

Shallow Foundations For substructure (e.g., piers or abutments) widths less 

than or equal to 30 m, a minimum of one investigation point per substructure. For substructure widths greater than 30 m, a minimum of two investigation points per substructure. Additional investigation points should be provided if erratic subsurface conditions are encountered. 

Depth of investigation should be: (1) penetrate unsuitable foundation soils (e.g., fine grained soils) into competent matericapacity (e.g. stiff to hard cohesive socohesionless soil or bedrock) and; (2) at stress increase due to estimated footing lothe existing effective overburden stress aencountered before the depth required bachieved, investigation depth should be grea minimum of 3 m into the bedrock, but robe sufficient to characterize compressibilinear-horizontal to horizontal discontinuities

Deep Foundations For substructure (e.g., bridge piers or abutments) 

widths less than or equal to 30 m, a minimum of one investigation point per substructure. For substructure widths greater than 30 m, a minimum of two investigation points per substructure. Additional investigation points should be provided if erratic subsurface conditions are encountered. Due to large expense associated with construction of rock-socketed shafts, conditions should be confirmed at each shaft location. 

In soil, depth of investigation should extend pile or shaft tip elevation a minimum of 6 mtimes the maximum pile group dimensionAll borings should extend through unsuunconsolidated fill, peat, highly organic grained soils, and loose coarse-grained soils materials. For piles bearing on rock, a minimum of 3 obtained at each investigation point locatboring has not terminated on a boulder. For shafts supported on or extending into roof rock core, or a length of rock core equathe shaft diameter for isolated shafts or twshaft group dimension, whichever is greabelow the anticipated shaft tip elevation to characteristics of rock within the zone of fo

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CHAPTER 4 

TOOLS AND TECHNIQUES FOR 

MEASURING SOIL AND ROCK PARAMETERS

4.1 INTRODUCTION 

The purpose of this section is to provide information on various in-situ and laboratory testing 

methods that are currently used to establish site-specific soil and rock properties for design and 

construction. The execution of a conventional subsurface exploration and testing program usually 

includes rotary drilling, Standard Penetration Testing (SPT), disturbed and undisturbed sample 

recovery, and laboratory testing. Although procedures for these commonly performed activities are 

codified in AASHTO and ASTM standards and are well known to most geo-professionals, important 

testing details are sometimes overlooked that can result in marginal data quality. This section was 

prepared to help identify the importance of selecting and conducting the appropriate field and 

laboratory testing method. 

In-situ testing methods are becoming increasingly used on transportation projects, however testing 

procedures and testing limitations are not as well understood as more conventional methods of 

subsurface exploration and testing. In this section, procedures for various in-situ and laboratory 

testing methods are presented as they relate to obtaining high-quality data for evaluation of 

engineering properties. Information on equipment calibration, measured test parameters, quality 

control, and the appropriate range of ground conditions that are appropriate for each test is also 

presented. 

In-situ tests discussed in this section include: (1) SPT; (2) CPT; (3) piezocone penetration testing 

(CPTu); (4) seismic piezocone testing (SCPTu); (5) DMT; (6) pressuremeter testing (PMT); and (7) 

vane shear testing (VST). Many state DOTs perform these tests directly using agency-owned 

equipment. In many cases however, the agency may directly contract to an outside contractor for 

these services or they may be contracted for as part of an overall project development package. 

Several technical reports and manuals are available that describe these methods. A brief list of these 

references is provided in table 4. Those agencies that perform or contract for these testing services 

are encouraged to obtain the references identified in table 4. 

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Table 4. Reference publications on in-situ testing. 

Test Method 

AASHTO/ 

ASTM Designation Reference 

SPT AASHTO T206 

ASTM D 1586 

U.S. Department of Transportation, Federal Highway Administration, 

(1997) Subsurface Investigations, Training Course in Geotechnical and 

Foundation Engineering, FHWA HI-97-021 

CPT, CPTu, 

SCPTu 

ASTM D 3341, D5778 U.S. Department of Transportation, Federal Highway Administration, 

(1988) Guidelines for Using the CPT, CPTu, and Marchetti DMT for 

Geotechnical Design, FHWA-SA-87-023-024. 

Lunne, T., Robertson, P.K., and Powell, J.J.M. (1997) Cone Penetration 

Testing in Geotechnical Practice, E & F Spon, 312 pp. 

DMT Suggested ASTM 

Method 

Schmertmann, 1986 

U.S. Department of Transportation, Federal Highway Administration 

(1992) The Flat Dilatometer Test, FHWA-SA-91-044. 

Marchetti, S., and Crapps, D.K. (1981) Flat Dilatometer Manual, Internal 

Report of GPE, Inc. (Gainesville, FL), available at http://www.gpe.org. 

PMT ASTM D 4719 U.S. Department of Transportation, Federal Highway Administration, 

(1989) The Pressuremeter Test for Highway Applications, FHWA-IP-89-

008. 

Clarke, B.G. (1995) Pressuremeters in Geotechnical Design, Blackie 

Academic & Professional, 364 pp. 

VST ASTM D 2573 American Society of Testing and Materials, (1988) Vane Shear Strength 

Testing in Soils: Field and Laboratory Studies, ASTM STP 1014, 378 

pp. 

The remainder of this section is devoted to describing the specific tools and techniques used to drill, 

sample and test soil and rock materials. In addition, summary information on the use of geophysical 

methods as a supplement to drilling, sampling, in-situ testing, and laboratory testing of soil and rock 

is presented. 

4.2 BORING METHODS 

Geotechnical borings are a critical component of any subsurface exploration program. They are 

performed to satisfy several objectives including those listed below. 

• identification of the subsurface distribution of materials with distinctive properties, 

including the presence and geometry of distinct layers; 

• determination of data on the characteristics of each layer by retrieving samples for use in 

evaluating engineering properties; 

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• acquisition of groundwater data; and 

• provide access for introduction of in-situ testing tools. 

There are many types of equipment used in current practice for advancing a soil or rock boring. 

Typical types of soil borings are listed in table 5(a), rock coring methods in table 5(b), and other 

exploratory techniques in table 5(c). Detailed information on soil and rock boring procedures are 

beyond the scope of this document, but can be found in AASHTO (1988), FHWA HI-97-021 (1997), 

and ASTM D 4700. A brief description of typical soil boring methods is provided below (Day, 

1999). 

• Discontinuous Auger boring: An auger is an apparatus with a helical shaft that can be 

manually or mechanically advanced to bore a hole into soil. Auger drilling is shown in 

figure 3. The practice of advancing a borehole with a mechanical auger consists of rotating 

the auger while at the same time applying a downward pressure on the auger to penetrate 

soil and possibly weak or weathered rock. The augers may be continuous, where the length 

of the helix is along the entire length of the shaft, or discontinuous when the auger is at the 

bottom of the drill stem. There are basically two types of discontinuous augers: 

discontinuous flight augers and bucket augers. Commonly-available discontinuous flight 

augers have diameters ranging from 0.05 to 1.2 m and bucket augers have diameters ranging 

from 0.3 to 2.4 m. For discontinuous flight auger borings, the auger is periodically removed 

from the hole and the soil lodged in the grooves of the flight auger is removed. When using 

a bucket auger, the soil in the bucket is periodically removed. A casing is generally not used 

for discontinuous flight and bucket auger borings. Therefore, these methods are not 

recommended for boreholes deeper than 10 m where the hole may cave-in during the 

excavation of loose or soft soils, or when the boring is below the groundwater table. In firm 

stiff clays, discontinuous auger borings can be performed to depths in excess of 10 m. 

• Continuous flight auger. As the name implies, continuous flight augers have the auger 

flights continuous along the entire length of the auger. There are two types of continuous 

flight augers: solid stem and hollow stem. For both of these type augers the drill cuttings 

are returned to the ground surface via the auger flights. The solid stem auger must be 

removed from the borehole to allow access to the hole for sampling or testing. Because the 

auger must be periodically removed from the borehole, a solid stem auger is not appropriate 

in sands and soft soils, or in soil deposits exhibiting high groundwater. A hollow-stem flight 

auger has a circular hollow core that allows for sampling through the center of the auger. 

The hollow-stem auger acts like a casing and allows for sampling in loose or soft soils or 

when the excavation is below the groundwater table. A plug is necessary when advancing 

hollow stem augers to prevent cuttings from migrating through the hollow stem. The plug is 

removed to permit SPT sampling. In loose sands and soft clays extending below the water 

table, drilling fluids are often used (and maintained) in order to minimize and mitigate 

disturbance effects. 

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Figure 3. Large diameter auger boring. 

• Wash-type borings. Wash-type borings use circulating drilling fluid (e.g., water or mud), to 

remove cuttings from the borehole. Cuttings are created by the chopping, twisting, and 

jetting action of the drill bit that breaks the soil or rock into small fragments. If bentonite or 

polymeric drilling muds cannot be used to maintain an open borehole, casings are often used 

to prevent cave-in of the borehole. The use of casing will require a significant amount of 

additional time and effort but will result in a protected borehole. When drilling mud is used 

during subsurface boring, it will be difficult to classify the soil using auger cuttings. Also, 

the outside of samples may become contaminated with drilling mud. 

The previously described methods are typically used for soil exploration, while the following 

methods are primarily used in rock exploration. 

• Rotary coring. This type of boring equipment is most commonly used for rock exploration 

when an intact core of the rock is desired. This technique uses power rotation of the drilling 

bit as circulating fluid removes cuttings from the hole. The drilling bits are specifically 

designed to core rock, and inner/outer tubes or casings are used to capture the intact core. 

Table 5(b) lists various types of rotary coring techniques for rock, although many of these 

techniques are applicable in stiff soil. 

• Percussion drilling. This type of drilling equipment is often used to penetrate hard rock for 

subsurface exploration or for the purpose of drilling wells. The drill bit works much like a 

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jackhammer, rising and falling to break up and crush the rock material. Air is commonly 

used to clean the cuttings to the ground surface. Table 5(c) includes a description on use of 

the percussion drilling techniques. 

Table 5(a). Boring methods (modified after Day, 1999). 

Method Procedure Applications Limitations / Remarks 

Auger boring 

(ASTM D 1452) 

Dry hole drilled with hand or 

power auger; samples 

recovered from auger flights 

In soil and soft rock; to 

identify geologic units and 

water content above water 

table 

Soil and rock stratification 

destroyed; sample mixed with 

water below the water table 

Hollow-stem 

auger boring 

Hole advanced by hollow-stem 

auger; soil sampled below 

auger as in auger boring above 

Typically used in soils that 

would require casing to 

maintain an open hole for 

sampling 

Sample limited by larger gravel; 

maintaining hydrostatic balance 

in hole below water table is 

difficult 

Wash-type boring Light chopping and strong 

jetting of soil; cuttings removed 

by circulating fluid and 

discharged into settling tub 

Soft to stiff cohesive 

materials and fine to coarse 

granular soils 

Coarse material tends to settle 

to bottom of hole; Should not be 

used in boreholes above water 

table where undisturbed 

samples are desired. 

Becker Hammer 

Penetration Test 

(BPT) 

Hole advanced using double 

acting diesel hammer to drive a 

168 mm double-walled casing 

into the ground. 

Typically used in soils with 

gravel and cobbles; Casing 

is driven open-ended if 

sampling of materials is 

desired 

Skin friction of casing difficult 

to account for; unsure as to the 

repeatability of test 

Bucket Auger 

boring 

A 600 to 1200-mm diameter 

drilling bucket with cutting 

teeth is rotated and advanced. 

At the completion of each 

advancement, the bucket is 

retrieved from the boring and 

soil is emptied on the ground. 

Most soils above water 

table; can dig harder soils 

than above types and can 

penetrate soils with cobbles 

and boulders if equipped 

with a rock bucket 

Not applicable in running sands; 

used for obtaining large 

volumes of disturbed samples 

and where it is necessary to 

enter a boring to make 

observations 

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Table 5(b). Rock core drilling methods (modified after Day, 1999)(1). 

Method Procedure Type of sample Applications Limitations / Remarks 

Rotary coring of 

rock (ASTM D 

2113; AASHTO T 

225) 

Outer tube with diamond 

(or tungsten carbide) bit 

on lower end rotated to 

cut annular hole in rock; 

core protected by 

stationary inner tube; 

cuttings flushed upward 

by drill fluid 

Rock cylinder 22 

to 100 mm wide 

and as long as 3 

m, depending on 

rock soundness. 

Standard coring 

size is 54 mm 

diameter. 

To obtain continuous 

core in sound rock 

(percent of core 

recovered depends on 

fractures, rock 

variability, equipment, 

and driller skill) 

Core lost in fracture or 

variable rock; blockage 

prevents drilling in 

badly fractured rock; 

dip of bedding and joint 

evident but not strike 

Rotary coring of 

rock, wire line 

Same as ASTM D 2113, 

but core and stationary 

inner tube retrieved from 

outer core barrel by 

lifting device or 

“overshot” suspended on 

thin cable (wire line) 

through special largediameter drill rods and 

outer core barrel 

Rock cylinder 28 

to 85 mm wide 

and 1.5 to 3 m 

long 

To recover core better 

in fractured rock which 

has less tendency for 

caving during core 

removal; to obtain 

much faster cycle of 

core recovery and 

resumption of drilling 

in deep holes 

Core lost in fracture or 

variable rock; blockage 

prevents drilling in 

badly fractured rock; 

dip of bedding and joint 

evident but not strike 

Rotary coring of 

swelling clay, soft 

rock 

Similar to rotary coring 

of rock; swelling core 

retained by third inner 

plastic liner 

Soil cylinder 28.5 

to 53.2 mm wide 

and 600 to 1500 

mm long encased 

in plastic tube 

In soils and soft rocks 

that swell or 

disintegrate rapidly in 

air (protected by plastic 

tube) 

Sample smaller; 

equipment more 

complex than other soil 

sampling techniques 

Note: (1) See section 4.3.3 for additional discussion on types of core barrels (i.e., single-, double-, 

or triple-tube). 

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Table 5(c). Other exploratory techniques (modified after Day, 1999). 

Method Procedure Type of sample Applications Limitations / Remarks 

Borehole 

camera 

Inside of core hole viewed 

by circular photograph or 

scan 

No sample, but a 

visual 

representation of 

the material 

To examine 

stratification, fractures, 

and cavities in hole 

walls 

Best above water table 

or when hole can be 

stabilized by clear water 

Pits and 

Trenches 

Pit or trench excavated to 

expose soils and rocks 

Chunks cut from 

walls of trench; size 

not limited 

To determine structure 

of complex formations; 

to obtain samples of 

thin critical seams such 

as failure surface 

Moving excavation 

equipment to site, 

stabilizing excavation 

walls, and controlling 

groundwater may be 

difficult; useful in 

obtaining depth to 

shallow rock and for 

obtaining undisturbed 

samples on pit/trench 

sidewalls; pits need to 

be backfilled 

Rotary or 

cable tool 

well drill 

Toothed cutter rotated or 

chisel bit pounded and 

churned 

Pulverized To penetrate boulders, 

coarse gravel; to 

identify hardness from 

drilling rates 

Identification of soils or 

rocks difficult 

Percussive 

Method 

(jack 

hammer or 

air track) 

Impact drill used; cuttings 

removed by compressed air 

Rock dust To locate rock, soft 

seams, or cavities in 

sound rock 

Drill becomes plugged 

by wet soil 

4.3 SAMPLING METHODS 

4.3.1 Disturbed Sampling of Soil 

Disturbed sampling provides a means to evaluate stratigraphy by visual examination and to obtain 

soil specimens for laboratory index testing. Disturbed samples are usually collected using splitbarrel samplers (figure 4; AASHTO T206, ASTM D 1586), although several other techniques are 

available for disturbed sample collection in boreholes (see table 6(a) and 6(b)). Shallow disturbed 

samples can also be obtained using hand augers and test pits. Direct push methods, such as 

GeoProbe sampling, can be used to obtain continuous disturbed samples but have similar limitations 

in sampling depth as solid stem and bucket augers (i.e., less than 10 m unless in firm to stiff clays). 

Discrete direct push samples can be obtained at depth using free-floating or retractable piston 

samplers. Samples obtained via disturbed sampling methods can often be used for index property 

testing in the laboratory but explicitly should not be used to prepare specimens for consolidation and 

strength (i.e. performance) tests. 

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33 

Figure 4. Split barrel sampler. 

Table 6(a). Common samplers to collect disturbed soil samples (modified after NAVFAC, 1982). 

Sampler Typical 

Dimensions 

Soils that Give Best 

Results 

Method of 

Penetration 

Cause of 

Low 

Recovery 

Remarks 

Split Barrel Standard is 

50 mm 

outside 

diameter 

(OD) and 35 

mm inside 

diameter 

(ID); 

All soils finer than gravel 

size particles that allow 

sampler to be driven; 

gravels invalidate drive 

data; 

A soil retainer may be 

required in granular soils. 

64 kg (140 lb) 

hammer driven 

Gravel may 

block 

sampler 

A SPT is performed using 

a standard penetrometer 

and hammer (see text); 

samples are extremely 

disturbed 

Continuous 

helicalflight auger 

Diameters 

range 76 to 

406 mm; 

penetrations 

to depths 

exceeding 

15 m 

Most soils above water 

table; will not penetrate 

hard soils or those 

containing cobbles or 

boulders 

Rotation Hard soils, 

cobbles, 

boulders 

Method of determining 

soil profile, bag samples 

can be obtained; log and 

sample depths must 

account for lag time 

between penetration of bit 

and arrival of sample at 

surface, to minimize 

errors in estimated sample 

depths 

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Table 6(b). Specialty samplers to collect disturbed soil samples (modified after NAVFAC, 1982). 

Sampler Typical 

Dimensions 

Soils that Give Best 

Results 

Method of 

Penetration 

Cause of 

Low 

Recovery 

Remarks 

Disc auger Up to 1070 mm 

diameter; 

usually has 

maximum 

penetration 

depth of 8 m 

Most soils above 

water table; will not 

penetrate hard soils 

or those containing 

cobbles or boulders 

Rotation Hard soils, 

cobbles, 

boulders 

Method of determining soil 

profile, bag samples can be 

obtained; log and sample 

depths must account for lag 

time between penetration of 

bit and arrival of sample at 

surface, to minimize errors 

in estimated sample depths 

Bucket 

auger 

Up to 1220 mm 

diameter 

common; larger 

sizes available; 

with extensions, 

depth over 24 m 

are possible 

Most soils above 

water table; can 

penetrate harder soils 

than above types and 

can penetrate soils 

with cobbles and 

boulders if equipped 

with a rock bucket 

Rotation Soil too hard 

to penetrate 

Several bucket types 

available, including those 

with ripper teeth and 

chopping tools; progress is 

slow when extensions are 

used 

Test boring 

of large 

samples, 

Large 

Penetration 

Test (LPT) 

50- to 75-mm 

ID and 63- to 

89- mm OD 

samplers 

(examples, 

Converse 

sampler, 

California 

Sampler) 

In sandy to gravelly 

soils 

Up to 160 

kg (350 lb) 

hammer 

driven 

Large gravel, 

cobbles, and 

boulders may 

block sampler 

Sample is intact but very 

disturbed; A resistance can 

be recorded during 

penetration, but is not 

equivalent to the SPT Nvalue and is more variable 

due to no standard 

equipment and methods 

4.3.2 Undisturbed Sampling of Soil 

4.3.2.1 General 

Undisturbed soil samples are required for performing laboratory strength and consolidation testing 

on generally cohesive soils ranging from soft to stiff consistency. High-quality samples for such 

testing are particularly important for approach embankments and for structural foundations and wall 

systems that may stress compressible strata. In reality, it is impossible to collect truly undisturbed 

samples since changes in the state of stress in the sample will occur upon sampling. The goal of 

high-quality undisturbed sampling is to minimize the potential for: (1) alteration of the soil structure; 

(2) changes in moisture content or void ratio; and (3) changes in chemical composition of the soil. 

Due to cost and ease of use, the thin-walled Shelby tube (figure 5) is the most common equipment 

for obtaining relatively undisturbed samples of soils. Depending upon cohesive soil type (e.g., 

stiffness and whether significant granular material is in the soil matrix), alternative sampling 

equipment may be used to obtain nominally undisturbed soil samples including: (1) stationary piston 

(figure 6); (2) hydraulic piston; (3) Denison (figure 7); and (4) pitcher samplers (figure 8). Summary 

information on these samplers is provided in table 7 and detailed procedures for these sampling 

techniques are provided in FHWA HI-97-021. Although not common for typical transportationArchived Document
35 

related projects, a variety of special samplers are available to obtain samples of soil and soft rocks 

such as the retractable plug, Sherbrooke, and Laval samplers. 

When dealing with relatively shallow soils that are very stiff, brittle, partially cemented, or contain 

coarse gravel or stones, the best method to obtain large relatively undisturbed samples is by block 

sampling. Block sampling involves isolating a soil column, encasing it in paraffin wax, and 

covering it with an open-ended box or tube (usually about 30-cm square). The bottom is cut, sealed 

and covered, and the sample is transported to the laboratory. This technique is difficult to implement 

for deep deposits of materials. 

Figure 5. Thin walled (Shelby) tube for sampling (with end caps). 

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Figure 6. Stationary piston sampler.

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Figure 7. Denison Sampler.

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Figure 8. Pitcher sampler.

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4.3.2.2 Overview of Thin-Walled Tube Sampling 

The importance of appropriate sampling practice cannot be overemphasized. Poor sampling 

practices, exposure to extreme temperatures, and careless handling of samples will result in 

misleading test results that may result in uneconomical or unsafe designs. Issues related to good 

practice for undisturbed sampling are provided in this section. 

• Geometry of a Thin-Walled Tube: The inside clearance ratio (ICR) and the area ratio (AR) 

are parameters that are used to evaluate the disturbance potential for different types of soil 

samplers. These parameters are defined as follows: 

2

22

i

io

D

DD AR − = (Equation 1) 

e

ei

D

DD ICR − = (Equation 2) 

where De = diameter at the sampler cutting tip, Di = inside diameter of the sampling tube, and Do = 

outside diameter of the sampling tube. For a sample to be considered undisturbed, the ICR should 

be approximately 1 percent and the AR should be 10 percent or less. Using a tube with this ICR 

value minimizes the friction buildup between the soil sample and the sampler during the 

advancement of the sampler. Using a tube with an AR value less than 10 percent enables the 

sampler to cut into the soil with minimal displacement of the soil. Thin-walled tubes (e.g., Shelby 

tubes) are typically manufactured to meet these specifications, but a thicker walled tube with an ICR 

of zero is commonly used in the Gulf states (i.e., Texas, Louisiana) to sample very stiff 

overconsolidated clays. The use of the thicker walled tube minimizes buckling of the sampler in the 

stiff deposits, and the ICR of zero minimizes sample expansion within the tube. Additional 

information on suitable geometry for thin-walled tubes is provided in ASTM D 1587. 

• Sample Tube Inspection and Storage: Tubes received from the manufacturer should be 

inspected to assure that no damage has occurred to the ends of the tubes. Plastic end caps, 

which will later be used to facilitate securing of the sample, should be placed on the ends of 

the tube at this time. 

• Cleaning Borehole Prior to Sampling: Depending upon the methods used, drilling and 

sampling procedures will cause some disturbance in the vicinity of the advancing face of a 

borehole. This is especially the case if a sample is overdriven, if casing is advanced ahead 

of the borehole, or during continuous sampling operations. It is recommended that a 

borehole be advanced and cleaned to two to three diameters below the bottom of the 

previous sample to minimize disturbance. Additionally, after advancement of the borehole, 

caving may occur at the bottom of the hole. Thus, the bottom of the borehole should be 

cleaned out thoroughly before advancing the sampling device. Improper cleaning will lead 

to severe disturbance of the upper material (accumulated settled material), and possibly 

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disturbance of the entire sample. Cleaning is usually performed by washing materials out of 

the hole. It should be ensured that the jet holes are not directed downward, for this will 

erode soft or granular materials to an unknown depth. All settled material should be 

removed to the edge of the casing. In deep or wide borings, special cleaning augers may be 

used to decrease time for cleaning and produce a cleaner hole. 

• Tube Advancement and Retrieval: Tubes should be advanced without rotation in a smooth 

and relatively rapid manner. The length of the sampler advancement should be limited to 

610 mm for a 762-mm long tube to minimize friction along the wall of the sampler and 

allow for loose material in the hole. The amount of recovery should be compared to the 

advanced length of the sampler to assess whether material has been lost, the sample has 

swelled, or some caved material has been collected at the top of the tube. The possible 

presence of caved material should be noted at the top of the tube so that no laboratory 

moisture content or performance tests are performed on that material. After advancing to 

the target depth, the drill rod should be rotated one full turn to shear off the bottom of the 

sample. A waiting period of 5 to 15 minutes is recommended for tubes in soft soils to 

permit the sample to reach equilibrium inside the tube and prevent the sample from falling 

out the bottom of the tube during retrieval. This waiting period may be reduced for stiffer 

soils. 

• Preparation for Shipment: Upon removal of the sample from the borehole, the ends should 

be capped using the plastic end caps and the tube should be labeled. The label should be 

written directly on the tube with a permanent marking pen, and include: (1) tube 

identification number; (2) sample depth; (3) top and bottom of sample; (4) length of 

recovery; (5) sampling date; (6) job name and/or number; and (7) sample description. Tube 

samples that are intended for laboratory performance testing (i.e., strength, consolidation, 

hydraulic conductivity) should never be extruded from the tube in the field and stored in 

alternative containers. Samples should only be extruded in the laboratory under controlled 

conditions. After the sample is collected, seal the upper end of the tube with nonshrinking 

wax. After the upper wax has dried, remove at least 25 mm of material from the lower end 

of the tube and seal the bottom of the tube with nonshrinking wax. The use of relatively low 

temperature wax will minimize shrinkage and potential moisture migration within the 

sample. The space between the wax seal and the top of the tube should be filled with 

sawdust or moist sand. The tube should be kept vertical, with the top of the sample in the 

upright position. If the sample needs to be inverted for purposes such as sealing, care should 

be taken to ensure the sample does not slide within the tube. 

• Shipment: Sample tubes need to be packed in accordance with guidelines provided in ASTM 

D 4220, or in an equivalent sample box. Tubes should be isolated from other sample tubes, 

and fit snugly in the case to protect against vibration or shock. The cushioning material 

between the samples should be at least 25 mm thick, and the cushioning on the container 

floor should be at least 50 mm thick. The samples should not be exposed to extreme heat or 

cold. If possible, the engineer should deliver the samples to the lab or use an overland 

freight service to ship samples. Typical handling practices of air freight services will lead to 

additional disturbance of the sample. The use of a chain of custody form for sample 

traceability records is encouraged. 

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Table 7. Nominally undisturbed soil samplers (modified after NAVFAC, 1982). 

Sampler Typical 

Dimensions 

Soils that Give 

Best Results 

Method of 

Penetration 

Cause of 

Disturbance or Low 

Recovery 

Remarks 

Shelby tube 

(ASTM D 1587; 

AASHTO T 207) 

76 mm OD and 73 

mm ID most 

common; available 

from 50 to 127 mm 

OD; 760-mm 

sampler length 

standard 

Cohesive finegrained or soft 

soils; gravelly 

and very stiff 

soils will crimp 

tube 

Pressing with 

relatively 

rapid, smooth 

stroke; can be 

carefully 

hammer driven 

but this will 

induce 

additional 

disturbance 

Erratic pressure 

applied during 

sampling, 

hammering, gravel 

particles, crimping of 

tube edge, improper 

soil types for 

sampler, pressing 

tube greater than 80% 

of tube length 

Simplest device for 

undisturbed samples; 

boring should be clean 

before sampler is 

lowered; little waste area 

in sampler; not suitable 

for hard, dense or 

gravelly soils 

Stationary piston 76 mm OD most 

common; available 

from 50 to 127 mm 

OD; 760-mm 

sampler length 

standard 

Soft to medium 

clays and fine 

silts; not for 

sandy soils 

Pressing with 

continuous, 

steady stroke 

Erratic pressure 

during sampling, 

allowing piston rod to 

move during press, 

improper soil types 

for sampler 

Piston at end of sampler 

prevents entry of fluid 

and contaminating 

material requires heavy 

drill rig with hydraulic 

drill head; samples 

generally less disturbed 

compared with Shelby 

tube; not suitable for 

hard, dense, or gravelly 

soil 

Hydraulic piston 

(Osterberg) 

76 mm OD is most 

common; available 

from 50 to 101 mm 

OD; 910-mm 

sampler length 

standard 

Silts and clays, 

some sandy 

soils 

Hydraulic or 

compressed air 

pressure 

Inadequate clamping 

of drill rods, erratic 

pressure 

Needs only standard drill 

rods; requires adequate 

hydraulic or air capacity 

to activate sampler; 

samples generally less 

disturbed compared with 

Shelby tube; not suitable 

for hard, dense, or 

gravelly soil 

Denison 89 to 177 mm OD, 

producing samples 

60 to 160 mm; 

610-mm sampler 

length standard 

Stiff to hard 

clay, silt, and 

sands with 

some 

cementation, 

soft rock 

Rotation and 

hydraulic 

pressure 

Improper operation 

of sampler; poor 

drilling procedures 

Inner tube face projects 

beyond outer tube, which 

rotates; amount of 

projection can be 

adjusted; generally takes 

good samples; not 

suitable for loose sands 

and soft clays 

Pitcher sampler 105 mm OD; uses 

76-mm diameter 

Shelby tubes; 

sample length 610 

mm 

Same as 

Denison 

Same as 

Denison 

Same as Denison Differs from Denison in 

that inner tube projection 

is spring controlled; often 

ineffective in 

cohesionless soils 

Foil Sampler Continuous 

samples 50 mm 

wide and as long 

as 20 m 

Fine grained 

soils including 

soft sensitive 

clays, silts, and 

varved clays 

Pushed into the 

ground with 

steady stroke; 

Pauses occur to 

add segments 

to sample 

barrel 

Samplers should not 

be used in soils 

containing fragments 

or shells 

Samples surrounded by 

thin strips of stainless 

steel, stored above cutter, 

to prevent contact of soil 

with tube as it is forced 

into soil 

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4.3.3 Rock Coring 

When considering equipment for rock coring, the dimensions, type of core barrel, type of coring bit, 

and drilling fluid are important variables. The minimum depth of rock coring should be determined 

based on the local geology of the site and type of structure to be constructed. Coring should also be 

performed to a depth that assures that refusal was not encountered on a boulder. 

Four different types of core barrels are described in ASTM D 2113 including: (1) Single Tube 

(figure 9(a)); (2) Rigid Double Tube (figure 9(b)); (3) Swivel Double Tube (figure 9(c)); and (4) 

Triple Tube. A brief description of issues related to rock coring is provided subsequently. 

Additional information on drilling rigs, methods of circulating drill cuttings (i.e., fluid or air), hole 

diameters, and casings are provided in ASTM D 2113. 

Since the double core barrel isolates the rock from the drilling fluid stream to yield better recovery, it 

is the minimum standard of core barrel that should be used in practice when an intact core is 

required. The inner tube of a swivel-type core barrel does not rotate during drilling, resulting in less 

disturbance and better recovery in weak and fractured rock. Rigid type double tube core barrels 

should not be used where core recovery is a concern. Triple tube swivel-type core barrels will 

produce better recovery and less core breakage than a double tube barrel. 

Conventional drilling and wireline techniques are applicable for rock coring. Wireline techniques 

are generally preferred by drillers for deep cores since the production rates are higher and the 

method facilitates deeper sampling. Wireline drilling can be used for any project, but is typically 

used for coring at depths greater than 25 m. 

The standard size rock core is NX 54 mm (2 1/8 in.) diameter. Generally larger core sizes will lead 

to less mechanical breakage and yield greater recovery, but the associated cost for drilling will be 

much higher. Since the size of the core will affect the percent recovery, this should be clearly 

recorded on the log. Additionally, the core length can increase recovery in fractured and weathered 

rock zones. In these zones a core length of 1.5 meters is recommended, and core lengths should not 

be greater than 3 m under any conditions because of the potential to damage the long cores. 

The coring bit is manufactured in accordance with one of the following designs: (1) diamond; (2) 

carbide insert; or (3) sawtooth. Bit selection will be based on the anticipated rock formation as well 

as the expected drilling fluid. Diamond bits are applicable in all rock types, and permit higher rates 

of coring when compared to other bits. Carbide insert bits are cheaper than diamond bits and can be 

used in soft to medium-hard rock. While sawtooth bits are the cheapest of the three, they have no 

salvage value, lead to slower coring, and are typically only used in soft rock. 

Archived Document
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Figure 9. Single and double tube rock core barrels (after FHWA-HI-97-021, 1997). 

Observations of drilling conditions and measurements of core properties taken in the field should be 

recorded on the boring log. These include: (1) size, type, and design of the core barrel; (2) length of 

each core run, core recovery, and rock quality designation (RQD); (3) engineer or geologists 

 (c) 

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44 

description of the formation; (4) structural descriptions including dip of strata, dip of jointing, 

presence of fissures, presence of fractures, and any other pertinent information; (5) depth, thickness, 

and apparent nature of joint filling; and (6) notations concerning drilling time and character. Cores 

in a split core barrel should be photographed immediately upon removal from the borehole. Core 

recovery and RQD are parameters often used in shear strength evaluations for rock masses. These 

parameters are defined in section 6.2.2. 

4.4 STANDARD PENETRATION TEST (SPT) 

4.4.1 General 

The most commonly used in-situ test in the world is the Standard Penetration test (SPT) (AASHTO 

T206, ASTM D 1586). The SPT is a simple and rugged test suitable for most soil types except 

gravel and is usually performed using a conventional geotechnical drill rig (see figure 10). SPTs are 

recommended for essentially all subsurface investigations since a disturbed sample can be obtained 

for baseline soil property interpretation. Many engineers have experience using SPT resistance for 

design purposes, even though the standard accepted correlations are often based on limited 

laboratory reference tests. Additionally, variability associated with hammer types used (i.e., donut, 

safety, automatic) and specific testing errors result in relatively poor correlations for evaluating 

performance properties for design, especially for cohesive soils. The test does provide a rough index 

of the relative strength and compressibility of the soil in the vicinity of the test. 

Figure 10. SPT performed at the back of a drill rig. 

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4.4.2 Procedures 

Standard Penetration Test procedures consist of repeatedly dropping a 63.5-kg hammer from a 

height of 760 mm to drive a split-spoon (i.e., split-barrel) sampler three successive 150-mm long 

increments. The number of blows required to drive the sampler is recorded for each 150-mm 

increment. The initial 150-mm increment is considered a seating drive. The blows required for the 

second and third 150-mm increments are totaled to provide blows/300 mm. This total is referred to 

as the as the SPT resistance or “N-value”. Blow counts taken for each 150-mm interval should be 

recorded, even for the seating increment. Additionally, the total recovery of soil during the 450-mm 

drive should be recorded. Depending upon sampler geometry, resistance for a fourth 150-mm 

increment is sometimes recorded. Due to sampler side friction, the values for the fourth 150-mm 

increment should not be used in the calculation of N-values, but may provide additional insight into 

soil stratigraphy. 

Since the SPT is highly dependent upon the equipment and operator performing the test, it is often 

difficult to obtain repeatable results. The main factors affecting the SPT results are summarized in 

table 8. The SPT should not be relied on in soils containing coarse gravel, cobbles, or boulders, 

because the sampler can become obstructed, resulting in high and unconservative N values. The test 

should not be relied on for cohesionless silts because dynamic effects at the sampler tip can lead to 

erroneous strength and compressibility evaluations. The test also has little meaning in soft and 

sensitive clays (Kulhawy and Mayne, 1990). 

When performing an SPT and recording information on the field log, the following items are of note: 

(1) N is always recorded as an integer; (2) a test is ended and noted as “refusal” if 50 blows over a 

25-mm increment has been recorded. At this point, the blows per 25 mm (or in.) is recorded (i.e., 

100/50 mm or 50/ 25 mm); and (3) if the N-value is less than one, then the engineer or geologist 

should record that the penetration occurred due to the weight of the hammer (WOH) or the weight of 

rods (WOR). 

4.4.3 Parameters Measured 

The measured N-value is the number of blows required to drive the split spoon sampler a distance of 

300 mm. The efficiency of the system can be obtained by comparing the kinetic energy, KE, (i.e., 

KE=1⁄2mv2

), with the potential energy, PE, of the system, (i.e., PE=mgh). The energy ratio (ER) is 

defined as KE/PE. For routine engineering practice in the United States, correlations for engineering 

properties are based on SPT N values measured based on a system which is 60 percent efficient, i.e., 

ER=60 percent. The N values corresponding to 60 percent efficiency are termed N60. Numerous 

correction factors to the measured N-value are necessary because of energy inefficiencies and 

procedural variation in practice. When all factors are applied to the field recorded N-value (Nmeas), 

the corrected value is calculated as: 

N60= NmeasCECBCSCR (Equation 3) 

where correction factors are presented in table 9 and include the effects of energy (CE), borehole 

diameter (CB), sampling method (CS), and rod length (CR). As can be noted from table 9, values of 

Archived Document
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the correction term for energy, i.e., CE, vary over a relatively wide range. For this reason, accurate 

estimates of CE are more important than estimates of the other correction factors. More accurate 

estimates of CE should be evaluated by directly measuring the energy ratio (ER) of a particular SPT 

setup according to procedures in ASTM D 4633. Commercially available equipment can be used to 

perform this calibration. Hammer systems used for standard penetration testing should be 

periodically calibrated using the procedures outlined in ASTM D 4633. 

Table 8. Factors affecting the SPT and SPT results (after Kulhawy and Mayne, 1990). 

Cause Effects Influence on SPT 

N Value 

Inadequate cleaning of hole SPT is not made in original in-situ soil, 

and therefore soil may become trapped 

in sampler and may be compressed as 

sampler is driven, reducing recovery 

Increases 

Failure to maintain adequate head of 

water in borehole 

Bottom of borehole may become quick Decreases 

Careless measure of drop Hammer energy varies (generally 

variations cluster on low side) 

Increases 

Hammer weight inaccurate Hammer energy varies (driller supplies 

weight; variations of 5 – 7 percent 

common) 

Increases or 

decreases 

Hammer strikes drill rod collar 

eccentrically 

Hammer energy reduced Increases 

Lack of hammer free fall because of 

ungreased sheaves, new stiff rope on 

weight, more than two turns on cathead, 

incomplete release of rope each drop 

Hammer energy reduced Increases 

Sampler driven above bottom of casing Sampler driven in disturbed, artificially 

densified soil 

Increases greatly 

Careless blow count Inaccurate results Increases or 

decreases 

Use of non-standard sampler Correlations with standard sampler 

invalid 

Increases or 

decreases 

Coarse gravel or cobbles in soil Sampler becomes clogged or impeded Increases 

Use of bent drill rods Inhibited transfer of energy of sampler Increases 

Archived Document
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Table 9. Corrections to the SPT (after Skempton, 1986). 

Factor Equipment Variable Term Correction 

Energy 

Ratio 

Donut Hammer 

Safety Hammer 

Automatic Hammer 

CE = ER/60 0.5 to 1.0(1)

0.7 to 1.2(1)

0.8 to 1.5(1)

Borehole 

Diameter 

65 to 115 mm 

150 mm 

200 mm 

CB 1.0 

1.05 

1.15 

Sampling 

method 

Standard sampler 

Non-standard sampler 

CS 1.0 

1.1 to 1.3 

Rod Length 3 to 4 m 

4 to 6 m 

6 to 10 m 

10 to >30 m 

CR 0.75 

0.85 

0.95 

1.0 

1

 Values presented are for guidance only. Actual ER values should be measured per ASTM D 4633 

Since N-values of similar materials increase with increasing effective overburden stress, the 

corrected blowcount (N60) is often normalized to 1-atmosphere (or about 100 kPa) effective 

overburden stress using overburden normalization schemes. The normalized corrected blowcount is 

referred to as (N1)60, and is equal to: 

(N1)60=CN N60 (Equation 4) 

where CN is the stress normalization parameter calculated as: 

CN = (Pa/σvo′)

n

 (Equation 5) 

where Pa is atmospheric pressure in the same units as σvo′, and n is a stress exponent typically equal 

to 1 in clays (e.g., Olsen, 1997; Mayne & Kemper, 1988) and 0.5 to 0.6 in sands (e.g., Seed et al., 

1983; Liao & Whitman, 1986; Olsen, 1997). Figure 11 illustrates a correlation for CN that is used 

for sands. There exist several soil specific correlations for CN, as reported in the literature. CN may 

vary slightly from the general values identified previously, depending upon the specific correlation. 

Archived Document
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0

0.5

1

1.5

2

2.5

0 50 100 150 200 250 300

VERTICAL EFFECTIVE STRESS (kPa)

CN

Sand (n=0.5)

Figure 11. Stress normalization parameter, CN, for sands. 

4.5 CONE PENETRATION TESTS (CPT / CPTU / SCPTU) 

4.5.1 General 

The cone penetration test (CPT) involves the hydraulic push of an instrumented steel probe at 

constant rate to obtain continuous vertical profiles of stress, pressures, and/or other measurements. 

No borehole, cuttings, or spoil are produced by this test. Testing is conducted in accordance with 

ASTM D 5778. The cone penetration test can be conducted without the use of a pore pressure 

measurement (i.e., CPT) or can be conducted using a device to measure penetration pore pressure 

using a piezocone (i.e., CPTu). Some equipment includes the ability to measure the propagation of 

shear waves using a seismic piezocone; this test designated as SCPTu. Details concerning cone 

penetration tests are summarized in table 10. Additionally, other in-situ tests that will be discussed 

subsequently are presented in this table. 

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Table 10. In-situ testing methods used in soil. 

Method Procedure Applicable 

Soil Types 

Applicable Soil 

Properties 

Limitations / Remarks 

Electric Cone 

Penetrometer 

(CPT) 

A cylindrical probe is 

hydraulically pushed 

vertically through the soil 

measuring the resistance at the 

conical tip of the probe and 

along the steel shaft; 

measurements typically 

recorded at 2 to 5 cm intervals 

Silts, sands, 

clays, and 

peat 

Estimation of soil 

type and detailed 

stratigraphy 

Sand: φ′, Dr, σho′

Clay: su, σp′

No soil sample is obtained; 

The probe may become 

damaged if testing in gravelly 

soils is attempted; Test results 

not particularly good for 

estimating deformation 

characteristics 

Piezocone 

Penetrometer 

(CPTu) 

Same as CPT; additionally, 

penetration porewater 

pressures are measured using 

a transducer and porous filter 

element 

Silts, sands, 

clays, and 

peat 

Same as CPT, 

with additionally: 

Sand: uo / water 

table elevation 

Clay: σp′, ch, kh 

OCR 

If the filter element and ports 

are not completely saturated, 

the pore pressure response may 

be misleading; Compression 

and wear of a mid-face (u1) 

element will effect readings; 

Test results not particularly 

good for estimating 

deformation characteristics 

Seismic 

CPTu 

(SCPTu) 

Same as CPTu; additionally, 

shear waves generated at the 

surface are recorded by a 

geophone at 1-m intervals 

throughout the profile for 

calculation of shear wave 

velocity 

Silts, sands, 

clays, and 

peat 

Same as CPTu, 

with additionally: 

Vs, Gmax, Emax, 

ρtot, eo

First arrival times should be 

used for calculation of shear 

wave velocity; If first 

crossover times are used, the 

error in shear wave velocity 

will increase with depth 

Flat Plate 

Dilatometer 

(DMT) 

A flat plate is hydraulically 

pushed or driven through the 

soil to a desired depth; at 

approximately 20 to 30 cm 

intervals, the pressure required 

to expand a thin membrane is 

recorded; Two to three 

measurements are typically 

recorded at each depth. 

Silts, sands, 

clays, and 

peat 

Estimation of soil 

type and 

stratigraphy 

Total unit weight 

Sand: φ′, E, Dr, mv

Clays: σp′, Ko, su, 

mv, E, ch, kh

Membranes may become 

deformed if overinflated; 

Deformed membranes will not 

provide accurate readings; 

Leaks in tubing or connections 

will lead to high readings; 

Good test for estimating 

deformation characteristics at 

small strains 

Pre-bored 

Pressuremeter 

(PMT) 

A borehole is drilled and the 

bottom is carefully prepared 

for insertion of the equipment; 

The pressure required to 

expand the cylindrical 

membrane to a certain volume 

or radial strain is recorded 

Clays, silts, 

and peat; 

marginal 

response in 

some sands 

and gravels 

E, G, mv, su Preparation of the borehole 

most important step to obtain 

good results; Good test for 

calculation of lateral 

deformation characteristics 

Full 

Displacement 

Pressuremeter 

(PMT) 

A cylindrical probe with a 

pressuremeter attached behind 

a conical tip is hydraulically 

pushed through the soil and 

paused at select intervals for 

testing; The pressure required 

to expand the cylindrical 

membrane to a certain volume 

or radial strain is recorded 

Clays, silts, 

and peat in 

sands 

E, G, mv, su Disturbance during 

advancement of the probe will 

lead to stiffer initial modulus 

and mask liftoff pressure (po); 

Good test for calculation of 

lateral deformation 

characteristics 

Archived Document
Table 10. In-situ testing methods (continued). 

50 

Method Procedure Applicable 

Soil Types 

Applicable Soil 

Properties 

Limitations / Remarks 

Vane Shear 

Test (VST) 

A 4 blade vane is slowly 

rotated while the torque 

required to rotate the vane is 

recorded for calculation of 

peak undrained shear strength; 

The vane is rapidly rotated for 

10 turns, and the torque 

required to fail the soil is 

recorded for calculation of 

remolded undrained shear 

strength 

Clays, 

Some silts 

and peats if 

undrained 

conditions 

can be 

assumed; 

not for use 

in granular 

soils 

su, St, σp′ Disturbance may occur in soft 

sensitive clays, reducing 

measured shear strength; 

Partial drainage may occur in 

fissured clays and silty 

materials, leading to errors in 

calculated strength; Rod 

friction needs to be accounted 

for in calculation of strength; 

Vane diameter and torque 

wrench capacity need to be 

properly sized for adequate 

measurements in various clay 

deposits 

Symbols used in table 10. 

φ′: effective stress friction angle 

Dr: relative density 

σho′: in-situ horizontal effective stress 

su: undrained shear strength 

σp′: preconsolidation stress 

ch: horizontal coefficient of consolidation 

kh: horizontal hydraulic conductivity 

OCR: overconsolidation ratio 

Vs: shear wave velocity 

Gmax: small-strain shear modulus 

G: shear modulus 

Emax: small-strain Young’s modulus 

E: Young’s modulus 

ρtot: total density 

eo: in-situ void ratio 

mv: volumetric compressibility coefficient 

Ko: coefficient of at-rest earth pressure 

St: sensitivity

4.5.2 Equipment 

Equipment necessary for performing a cone penetration test includes a penetrometer, cone rod or 

drill rod, electrical cable, a data acquisition system, and hydraulic actuator attached to equipment 

exhibiting sufficient reaction mass to advance the penetrometer. This can be a conventional drilling 

rig or, more commonly, a dedicated CPT truck commonly weighing 20 to 25 tons. 

A standard cone penetrometer is a 35.7-mm diameter cylindrical probe with a 60o

 apex at the tip, 10-

cm

2

 projected tip area, and a 150-cm2 sleeve surface area. More robust penetrometers are available 

with a 44-mm diameter body, a 15-cm2 projected tip area, and 200- to 225-cm2

 sleeve surface area. 

A 15-cm2 penetrometer will generally provide the same response as a 10-cm2 probe. The size of a 

cone is identified by the projected tip area, e.g. a 10-cm2 cone or a 15-cm2 cone. Figure 12 shows a 

number of different cone penetrometers and piezocones. Standard cone rod is typically 1 m in 

length with a 35.7 mm outer diameter and a 22 mm inner diameter. Alternatively, the penetrometer 

can be pushed with standard AW or EW drill rod. 

A cone cable will run through the hollow cone/drill rods and attach to a data acquisition system at 

the ground surface. The data acquisition system will generally consist of an analog signal 

conditioner, an analog to digital (A-D) converter, and computer processor. Current data acquisition 

Archived Document
51 

systems are attached to one or two computer monitors so the operator and engineer can observe data 

recorded during the sounding in real time. Real time monitoring allows for decisions to be made in 

the field with respect to the sounding. This is helpful if auxiliary tests, such as a pore pressure 

dissipation tests, are to be performed in certain soil layers, or if the test is to be terminated once a 

certain layer is encountered. Printers can be attached to the computer processor to obtain a real-time 

printout of the data during the test. Printed data are a good backup in case an unforeseen incident 

causes the computer to crash and lose the data. Data are typically recorded every 2 to 5 cm of 

vertical penetration. 

For a piezocone penetration test (CPTu), the penetration porewater pressures are monitored using a 

transducer and porous filter element. The filter element position can be located at the mid-face on 

the cone (u1) or behind the cone tip at the shoulder (u2), with the latter required for the correction of 

tip resistance. Filter elements consist of high-density polypropylene, ceramic, or sintered metal. 

The high-density polypropylene filters wear easily in sands and may clog in clays. Since 

polypropylene filters are inexpensive and disposable, they should generally be replaced after each 

test. Ceramic and sintered metal can be re-used many times, but should be switched frequently and 

cleaned to prevent clogging. Filter compression and wear may influence u1 results, and the position 

of the u1 filter may vary between cone manufacturers. The u2 position is more standardized and is 

required for tip correction. It is not as affected by filter compression as the u1 location. In most 

cases it is recommended to specify the u2, but on critical projects where detailed stratigraphy and or 

stress history are desired it may be advantageous to have the subcontractor bring additional tips with 

u1 filter locations. This allows for flexibility in testing, if it is necessary to switch filter location in 

the middle of the program due to encountered soil conditions. A u1 element is recommended for 

profiling in stiff fissured geomaterials where u2 ≈ 0. 

For the seismic piezocone test, a geophone is located approximately 500 mm uphole from the cone 

tip. The geophone detects shear waves generated at the ground surface at intervals of approximately 

1-meter, corresponding to successive rod additions. 

Archived Document
52 

Figure 12. Cone and piezocone penetrometers (note the quarter for scale). 

4.5.3 Procedures 

Test procedures for the CPT consist of hydraulically pushing the cone at a rate of 2 cm/s in 

accordance with ASTM D 5778 using either a standard drill rig or specialized cone truck (see figure 

13). The advance of the probe requires the successive addition of rods at approximately 1 m or 1.5 

m intervals. Readings of tip resistance (qc), sleeve friction (fs), inclination (i), and pore pressure (um) 

are taken at least at every 5-cm (i.e., 2.5-sec interval). For the seismic cone test, shear wave arrival 

times (ts) are typically recorded at rod breaks corresponding to 1-m or 1.5-m intervals in the strata. 

Careful saturation of the porous filter and transducer ports is paramount for piezocone testing. Poor 

saturation will lead to a compressible measurement system, and thus the full magnitude of the 

penetration pore pressure response will not be recorded. If water or water mixtures are used as the 

saturation fluid, a fluid filled membrane should be wrapped around the element to maintain 

saturation until the probe enters the ground. Glycerin and silicon oil are typically viscous enough to 

prevent desaturation of the element before penetration into the ground. 

Archived Document
53 

Typically a pause in penetration will occur to add new rods. This is referenced as the rod break. 

The depth at each rod break should be recorded and compared to the expected depth. Inconsistent 

depths typically result from: (1) multiple readings at a pause in penetration due to decompression of 

the rod string; or (2) double recording of an interval due to the depth sensor being activated during 

retraction of the pushing head. The extra readings typically will show significantly reduced tip 

resistance and sleeve friction values due to decoupling of the cone from the soil matrix. These 

readings should be deleted from the CPT record and the depth of the sounding should be corrected. 

For the seismic piezocone (SCPTu), downhole shear wave velocity tests are performed at each 1-m 

rod break. A special instrumented hammer is used to trigger a surface source rich in shear waves 

(e.g., typically a horizontal steel beam). The steel beam is coupled to the ground under a hydraulic 

outrigger of a cone truck or drill rig, or under the tire of a support vehicle. The horizontal distance 

between the source beam and cone rod should be minimized (typically < 1.5 m) to ensure a relatively 

vertically-propagating shear wave. A horizontal geophone located within the penetrometer serves as 

a receiver for the signal, which is displayed on the screen of an oscilloscope. First arrival times for 

shear waves are recorded with respect to depth, to provide interpretations of shear wave velocity of 

the overlying soil material. 

Figure 13. Cone penetration testing from cone truck. 

Archived Document
54 

4.5.4 Parameters Measured 

Electric and electronic penetrometers have standard readings of tip resistance (qc) and sleeve friction 

(fs), as shown in figure 14(a). Piezocone penetrometers measure penetration porewater pressures 

using filters located at the shoulder (u2; figure 14(b)) or the midface (u1; figure 14(c)). A horizontal 

geophone in the seismic piezocone (figure 14(d)) can be used to record mechanically induced shear 

waves from the surface, leading to determination of shear wave arrival time (ts) and shear wave 

velocity (Vs). 

Figure 14. Measurement locations on cone penetrometers: 

a. Electric Cone Penetrometer, CPT; b. Piezocone Penetrometer (filter behind tip), CPTu2; 

c. Piezocone Penetrometer (mid-face filter) CPTu1; d. Seismic Piezocone, SCPTu2. 

The cone tip resistance (qc) is the measured axial force over the projected tip area. It is a point stress 

related to the bearing capacity of the soil. In sands, the tip resistance is primarily controlled by the 

effective stress friction angle (φ′), relative density (Dr), and effective horizontal stress-state (σho′). 

For intact clays, the tip resistance is primarily controlled by the undrained shear strength (su) and 

preconsolidation stress (σp′). Particularly in clays and silts, the measured qc must be corrected for 

porewater pressures acting on the cone tip geometry, thus obtaining the corrected tip stress, qt

(Lunne, et al., 1997): 

qt = qc + (1-an)u2 (Equation 6) 

where an is the net area ratio determined from laboratory calibration and u2 is the shoulder 

penetration porewater pressure. The net area ratio is approximated as the ratio of the unequal end 

areas of the cone (see figure 15). The net area ratio (an) is penetrometer-specific and is obtained by 

isotropic pressurization of the cone in a triaxial cell. It is best to use penetrometers with an an > 0.80 

 a. b. c. d. 

Archived Document
55 

to minimize the necessary correction, although many 15 cm2 cones have an ≈ 0.75. Contract 

specifications should always request the an calibration curves and clear indication that qc readings 

have been adjusted to provide the proper qt values. A general rule of thumb is that qt > 5 MPa in 

sands, while qt < 2 MPa in soft to medium clays and silts. It should be noted that the tip resistance in 

overconsolidated clays can occasionally be larger than 5 MPa. 

Figure 15. Illustration of unequal end areas of CPT (after Kulhawy and Mayne, 1990). 

The sleeve friction (fs) is a shear stress determined as load acting over the cylindrical surface area of 

a smooth sleeve. This value is often expressed as the Friction Ratio (FR) given as fs / qt x 100, 

which is indicative of soil type (Lunne et al., 1997). Often, FR is less than one percent in clean 

sands and greater than four percent in clays and silts of low to medium sensitivity. In highly 

sensitive clays, FR may be approximately one percent or less. 

The penetration porewater pressures are monitored using a transducer and porous filter element. 

These readings represent the fluid pressures between the soil particles during penetration. At the 

shoulder position, the pressures are near hydrostatic in sands (u2 ≈ uo) whilst considerably higher 

than hydrostatic (u2 > uo) in soft to firm to stiff intact clays. Using values for total stress, σvo, and 

Archived Document
56 

hydrostatic pore pressure, uo, the pore pressure parameter, Bq = (u2 - uo) / (qt - σvo), is used as a 

means to normalize CPTu data for the purpose of soil classification and undrained shear strength 

estimation (see chapter 5 for discussion on the use of Bq for soil classification). At the mid-face 

location (u1), penetration porewater pressures are always positive, while at the u2 location 

measurements range from positive in intact (i.e., non fissured) materials to as low as negative one 

atmosphere (-100 kPa) in fissured clays and dense silts. The parameter resulting from a CPT, CPTu, 

and SCPTu are combined to provide several useful index and performance parameters. Details of 

these parameters are provided in chapter 5. 

4.6 FLAT DILATOMETER TEST (DMT) 

4.6.1 General 

The flat plate dilatometer (DMT) involves pushing an instrumented steel blade into the subsurface 

soils and periodically stopping the penetration to obtain specific pressure measurements at the 

selected depth. No borehole cuttings or spoil are generally produced by this test, although it is 

possible to advance a conventional soil boring and then perform the DMT downhole within the 

borehole. 

4.6.2 Equipment 

The flat plate dilatometer system consists of a high strength steel blade, tubing, pressure gauge 

readout unit, and a nitrogen gas tank. The tapered steel blade is approximately 240-mm long, 95-

mm wide, 15-mm thick, with an 18o

 wedge tip. A 60-mm diameter inflatable steel membrane exists 

on the face of the blade. The DMT tubing contains a wire, which is connected to an audible alarm 

within the control panel. The alarm notifies the operator to take specific pressure readings (i.e., the 

A, B, and C pressures, discussed subsequently). Figure 16 shows a flat plate dilatometer blade and 

the associated equipment. 

4.6.3 Procedures 

While there is no ASTM standard for the flat plate dilatometer (DMT), a suggested method has been 

outlined in Schmertmann (1986). Since the DMT operates using a pressure-based system, all 

connections should be sufficiently tightened with a wrench to eliminate leaks. 

Archived Document
57 

Figure 16. Flat plate dilatometer test equipment. 

The blade is pushed into the ground at 20 mm/sec utilizing a similar truck and hydraulic system as 

described for the cone penetration test. Discrete tests are performed typically at 200-mm intervals. 

For each test, a 60-mm diameter flexible circular membrane on the face of the blade is inflated with 

nitrogen pressure to provide the A- and B-readings corresponding to the deformation, δ, of the 

circular membrane. Specifically the A-reading: (A) is the lift-off pressure where δ=0; and the Breading (B) is the expansion pressure where δ=1.1 mm. A third reading may be recorded during 

deflation of the membrane and is noted as the closing pressure, or the C-reading. 

The test should start within 15 seconds of reaching the desired depth, the A-reading should be taken 

within 15-30 seconds after beginning the test, and the B-reading should be taken within the next 15-

30 seconds. After the B-reading is attained, the system is vented to prevent overexpansion of the 

membrane. The C-reading is recorded when the membrane is flush with the face of the blade during 

slow venting of the gas pressure. After deflation of the membrane, the blade is pushed to the next 

test depth and the inflation cycle is repeated. 

Low 

Reading 

Gauge Pressure 

Increase 

Vent 

Pressure 

Relief 

(C-Reading) 

High 

Reading 

Gauge 

Regulator 

Dilatometer Blade 

Nitrogen 

Tank 

Archived Document
58 

4.6.4 Parameters Measured 

The A- and B- readings need to be corrected for membrane stiffness effects to obtain the liftoff 

pressure, po, and expansion pressure, p1. Correction of the readings has been presented by 

Schmertmann (1986): 

po = 1.05(A + ∆A -zm) -0.05(B - ∆B - zm) (Equation 7) 

p1 = B - ∆B - zm (Equation 8) 

where ∆A and ∆B (reported as positive absolute values) are the calibration factors for applied 

suction and expansion of the membrane in air, respectively, and zm is the gage offset zero reading. 

For a new gage, zm is often zero. 

The two dilatometer pressures, po and p1, are combined with the hydrostatic water pressure, uo, to 

provide three index parameters developed by Marchetti (1980). Hydrostatic water pressure should 

be evaluated based on available pore pressure information. The material index, ID, is related to the 

soil classification and is presented as: 

ID = (p1 - po) / (po - uo) (Equation 9) 

The dilatometer modulus, ED, is related to the compressibility of the soil. The equation for this 

parameter is based on elastic theory and is presented as: 

ED = 34.7 (p1 - po) (Equation 10) 

The horizontal stress index, KD, is related to the in-situ horizontal stress-state of the soil. The index 

KD will always be greater than Ko due to disturbance caused during insertion of the blade. This 

parameter is presented as: 

KD = (po - uo) / σvo' (Equation 11) 

A soil classification and consistency scheme has been developed utilizing the material index, ID, and 

Dilatometer Modulus, ED, and is presented in Schmertmann (1986). Other useful correlations to 

engineering properties are presented in chapter 5. 

4.7 PRESSUREMETER TEST (PMT) 

4.7.1 General 

The pressuremeter test (PMT) involves inflating a cylindrical probe against the sidewalls of a boring. 

In general, the instrument is placed in a pre-bored hole prior to expansion, although it is possible to 

self-bore the instrument to the test location. The pressuremeter can be used to obtain specific 

strength and deformation properties of the subsurface soils. 

Archived Document
59 

4.7.2 Equipment 

Pressuremeter equipment consists of an expandable cylindrical probe, inner rubber membrane, 

pressure lines, and an outer slotted tube or Chinese lantern. The pressuremeter probe may consist of 

either one cell or three cells that are hydraulically or pneumatically expanded to obtain a pressure 

versus volume curve or pressure versus radial strain curve. The inner rubber membrane is expanded 

at a specific test depth, and the outer slotted tube or Chinese screen expands with the membrane. 

The slotted tube or screen protects the membrane from punctures induced by gravel or other sharp 

objects. 

Developments in pressuremeter testing include the self-boring pressuremeter, full displacement 

pressuremeter, and the cone pressuremeter. These devices do not require the independent 

preparation of a borehole. The latter two are thus a potentially higher production test. The selfboring pressuremeter is primarily used for research. Unfortunately, the probe penetration during a 

full displacement test leads to a highly disturbed zone of soil, which complicates interpretation of the 

field data. These instruments have not yet been extensively used on DOT projects. 

Pressuremeter dimensions have not been standardized, which may lead to error when attempting to 

compare test data from different probes. Commonly, a 76-mm diameter probe is used and a thin 

walled Shelby tube can be conveniently pushed and extracted prior to the PMT. Typical probe 

diameters for Menard type (i.e., pre-bored) pressuremeters are presented in ASTM D 4719, and it is 

recommended that the length to diameter (L:D) ratio be at least 6.5L:1D to minimize end effects (see 

figure 17). For the full displacement and cone pressuremeters, the diameter is typically that of a 15 

cm2

 cone (43.7 cm) and the suggested length to diameter ratio is 10L:1D. 

4.7.3 Procedures 

The most important aspect of the pressuremeter test (ASTM D 4719) is the preparation of the 

borehole. The sidewalls of the borehole should be smooth, consistent, and of the appropriate 

uniform diameter. The borehole should not be less than 1.03D, or greater than 1.2D, where D is the 

initial diameter of the pressuremeter. 

The PMT proceeds by incrementally increasing the inflation pressure while monitoring the radial 

deformation or volume. A loading increment should be selected to yield accurate results without 

producing an excessively long test. Seven to ten load increments of 25 to 200 kPa are typically used, 

depending upon the anticipated soil conditions. Loading increments are generally applied at 

typically one to three minute intervals. Operator judgment and experience are typically used in 

deciding when it is appropriate to increase or decrease the load. In general, the load versus 

deformation response is monitored and when the response stabilizes under a given load, the next 

load increment is applied. 

It is good practice to perform the PMT using a phase of monotonic loading, a phase including a 

drained creep test, and at least one unload-reload cycle. The purpose of the creep test is to assess the 

time-dependent deformation behavior of the material. In a creep test, a constant pressure is 

maintained in the pressuremeter and the corresponding time-dependent deformation is recorded. 

Archived Document
60 

Depending on the soil type and the magnitude of the staged load, the creep testing stage may take 

several minutes to as long as an hour. In general, this test should be conducted for approximately 5 

to 10 minutes for a clayey soil, for 3 to 5 minutes for silty soils, and for 2 to 3 minutes for sandy 

soils. Within these time periods, the displacement of the membrane is anticipated to have generally 

stabilized. Soft soils at a high load ratio will require the longest time to complete. 

The purpose of unload-reload cycle is primarily to assess the elastic behavior of the material. Since 

the initial loading includes disturbance effects as well as both recoverable and irrecoverable 

deformations, an evaluation of the “elastic” modulus based on the initial pressuremeter response will 

likely underestimate the actual small-strain deformations of the material during loading under the 

field-applied stress. In general, it is recommended to include approximately two unload steps and 

three to four reload steps for each unload-reload cycle. To avoid inducing plastic deformations 

during the unloading cycle, it is recommended that no more than 80 percent of the applied pressure 

be removed before the reload cycle is initiated. In addition, care must be taken to release the 

pressure slowly during the unload cycle to allow the pressure in the pressuremeter membrane to 

equilibrate with the gage pressure when the pressure and deformation readings are recorded. If not 

conducted carefully, the membrane will appear to move outward when the pressure is decreased, 

because the pressure in the membrane will actually be higher than the gage pressure. If time permits 

(i.e., an unload–reload cycle may take nearly as long to complete as a monotonic-loading only 

pressuremeter test), it is recommended to run more than one unload-reload test. Finally, it is 

desirable to conduct one of the unload-reload cycles prior to achieving the yield pressure, pf. 

A typical pressuremeter curve for pre-bored Menard type pressuremeter is shown on figure 18. This 

shows a monotonic loading stage, followed by an unload-reload cycle, a creep stage, a second 

unload-reload cycle and a final monotonic load increase. 

4.7.4 Parameters Measured 

The following characteristic pressures on the pressuremeter curve are shown on figure 18: 

• po – the pressure at which recompression of the disturbed soil is complete and expansion 

into undisturbed soil is started. 

• pf – an inflection point known as the creep or yield pressure where the soil changes from 

pseudo elastic to plastic where shear is initiated. 

• Creep test – prior to performing an unload reload test, a creep test should be performed. 

This allows deformation to continue at a constant pressure until strain rates of 0.1%/min are 

recorded. 

• pu – the minimum pressure during unloading, in the unload – reload cycle. 

• pr – the yield point during the reloading portion of an unload – reload cycle where 

recompression ends and the soil reinitiates plastic shearing. 

Archived Document
61 

Figure 17. Pre-bored pressuremeter equipment. 

• pL – the limit pressure where the curve becomes asymptotic on a pressure versus volume 

curve. The value of pL is taken as the extrapolated pressure when the volume is equal to 

2xVo, where Vo is the initial volume of the pressuremeter. 

Figure 18. Typical curves and characteristic pressures for pre-bored Menard pressuremeter. 

pL is the 

pressure 

corresponding 

to 

∆V = Vo

or when 

V = 2Vo

Pressure

Volume

p o

p f

p L

Creep

Test

pu

p r

Pseudo

Elastic Plastic

Unload-reload cycle

Archived Document
62 

The pseudo-elastic modulus from pressuremeter tests is calculated by one of two methods, 

depending upon whether the data are plotted on an applied stress versus probe volume curve or an 

applied stress versus probe radius curve. The equation to calculate the pressuremeter elastic 

modulus from an applied stress versus probe volume curve is (ASTM D 4719): 

( )( ) V

P EP VV m ∆∆ 12 ν 0 ++= (Equation 12) 

where ν is the Poisson ratio (usually taken as 0.33), Vo is the initial volume of the probe, Vm is the 

average volume of the probe over the stress range of consideration (i.e., (Vfinal − Vinitial)/2 for a given 

stress increment), and ∆P / ∆V is the slope of the linear portion of the stress versus probe volume 

curve (between po and pf or between pu and pr). If the slope is taken as the unload-reload portion of 

the curve, this should be noted as Eu-r. 

The equation to calculate the pressuremeter elastic modulus from an applied stress versus probe 

radius curve is (ASTM D 4719): 

( )( ) Rd

P EP P RR m ∆∆ 1 ν ∆++= (Equation 13) 

where ν is the Poisson ratio (usually taken as 0.33), RP is the uninflated radius of the probe, ∆Rm is 

the increase in radius of the probe to where Ep is measured, and ∆P / d∆R is the slope of the linear 

portion of the applied stress versus radius curve between po and pf or pu and pr. 

4.8 VANE SHEAR TEST (VST) 

4.8.1 General 

The vane shear test (VST) involves the use of a simple rotated blade to evaluate the undrained shear 

strength in soft to stiff clays and silts. The use of the VST should be limited to soils in which slow 

(6o

 / min) rotation of the blade will lead to undrained shearing. 

4.8.2 Equipment 

Equipment necessary for a VST includes a four-sided vane with a height to diameter (H/D) ratio of 

2, rods, and a torque-measuring device. Additional equipment may include a rod sheath, protection 

shoe, or slip coupling to account for rod friction, and a gear drive to provide a constant rate of 

rotation. 

The standard vane has a diameter (D) of 65 mm, a height (H) of 130 mm, and a vane edge thickness 

(t) of 2 mm. However, vane sizes range from a diameter of 38 to 92 mm, a height of 76 to 184 mm, 

a blade thickness between 1.6 and 3.2 mm, and are attached to a 12.7-mm diameter rod. The rod 

diameter may need to be increased in stiff materials to prevent yielding of the rod during rotation. 

The vane may be rectangular, double-tapered, or single-tapered (i.e., tapered at the bottom of the 

Archived Document
63 

vane). Using a tapered vane will facilitate placement of the vane in stiffer materials. A typical vane 

and a summary of the geometric parameters used to identify a vane are provided in figure 19. 

Vane size selection is a function of the anticipated strength of the soil and accuracy of the torque 

wrench. Larger vanes are typically used in soft soils and smaller vanes used in stiffer soils. While a 

large vane will provide better resolution than a smaller vane, it may cause more disturbance during 

insertion, be more difficult to rotate and thus lead to additional disturbance, or result in loads that 

overstress the capacity of the torque wrench. A number of different sized vanes, as well as torque 

wrenches with varying capacity, should be brought to the field to accommodate potentially variable 

conditions. 

Figure 19. (a) Rectangular vane; and (b) Parameters used to define vane dimensions. 

4.8.3 Procedures 

Procedures for the vane shear test are outlined in ASTM D 2573. Depending upon what type of 

vane shear device is used, there is the potential for friction to develop along the rod. The torqueArchived Document
64 

measuring equipment can record this friction. This rod friction needs to be minimized and 

accounted for in the calculation of shear strength. Typical methods to account for rod friction 

include: (1) rods protected within a sheath; or (2) measurement of rod friction with a slip coupling. 

If the vane test is performed below the bottom of a borehole, the depth to the top of the vane from 

the bottom of the borehole should be equal to about 4 borehole diameters to minimize disturbance 

effects. 

The VST should be started within 5 minutes of insertion, and the vane should be rotated at 6o

 per 

minute. This usually results in a time to failure between 2 and 5 minutes and is usually fast enough 

to assure undrained conditions, but time to failure may be up to 10 to 15 minutes in very soft clays. 

Readings of torque at 30 second to 1-minute intervals are recommended to provide an assessment of 

the soil response. Additionally, systems are currently available which record the applied torque with 

time or angle of rotation. Systems that use a gear drive to rotate the vane can eliminate some of the 

potential operator error, but limit the flexibility necessary to test extremely soft soils. 

4.8.4 Parameters Measured 

Three parameters can be obtained from the vane shear test: (1) undrained shear strength (su,VST); (2) 

remolded undrained shear strength (sr,VST); and (3) sensitivity (St,VST). It should be noted that the 

subscript VST is added to each parameter to note that the parameter was obtained using vane shear 

test data. The undrained shear strength, remolded shear strength, and sensitivity will differ 

depending upon the mode of shear, as well as the strain level and degree of remolding. 

During rotation, the torque (T) is measured and the maximum torque (Tmax) is used to calculate the 

undrained shear strength based on the vane geometry. Prior to calculation of undrained shear 

strength (su,VST), the torque associated with rod friction (Trod) must be subtracted from the measured 

torque (Tnet = Tmax – Trod). Best practice involves using a sheath (or a slip coupling) to eliminate rod 

friction, and thus Tnet would equal Tmax. The undrained shear strength for a standard rectangular 

vane (H/D = 2) is expressed as: 

, 3 7

6

D

T

s net

VSTu

π = (Equation 14) 

where D is the diameter of the vane (see figure 19). For the general case of rectangular or tapered 

vanes, the following equation is used to calculate undrained shear strength: 





 



 ++

=

H i

D

i D D

T

s

T B

net VSTu

6

cos cos

12

2 ,

π

 (Equation 15) 

where D is the diameter of the vane, H is the height of the vane, iB is the angle of taper at the bottom, 

and iT is the angle of taper at the top. These parameters are graphically displayed in figure 19(b). 

The remolded strength is achieved in the same manner as the peak strength, except the torque 

reading is taken during rotation of the vane following 10 rapid turns. The torque associated with rod 

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friction should be recorded prior to the remolded test, and then subtracted from the maximum torque 

recorded for calculations of sr,VST. The sensitivity of the soil from vane shear tests is expressed as: 

VSTr

VSTu

VSTt

s

s

S

,

,

, = (Equation 16) 

4.9 USE OF DRILL RIGS TO PERFORM IN-SITU TESTS 

When using a drill rig to perform in-situ tests such as the CPT, DMT, and PMT, it will be necessary 

to carefully pass wires, cables, and tubing through the drill rod and couplings. The inside diameter 

of the coupling will typically be the limiting diameter. The ends of cables will typically be the 

largest diameter of the cable or tubing system and should therefore be smaller than the diameter of 

the drill rod couplings. Typically, AW drill rod is used when CPTs are performed from drill rigs. 

Wires, cables, and tubing will need to exit through a coupling at the ground surface. The top 

coupling should have a slot in the side to enable cables and tubing to pass through without being 

pinched or cut. 

It is important to ensure that the adapter for the in-situ testing device be appropriately attached to the 

drill rod. Threads for the CPT and DMT are usually proprietary and manufacturers of CPT and 

DMT equipment do not typically have drill rod adapters readily available. For this reason, it is best 

practice to have a drilling equipment supplier weld the CPT or DMT adapter to the appropriate drill 

rod coupling. 

Casing may be needed to support the borehole above the zone of testing for the VST and PMT tests. 

Typical casing sizes are provided in FHWA HI-97-021 (1997). Appropriate casing sizes and 

diameters will be a function of the probe diameter and are outlined in ASTM D 2573 for the VST 

and ASTM D 4719 for the PMT. 

If non-standard equipment is used to advance the probe, such as a drill rig or GeoProbe truck, the 

time for advancement of each rod should be recorded for CPT and DMT tests. A CPT contractor 

may be able to use the data acquisition system to record time of penetration along with depth, tip 

resistance, sleeve friction, and penetration pore pressure. It should be specified that the 

subcontractor provide time information along with depth information, if their system is capable of 

recording time. The penetration should be maintained relatively constant, at a rate of 20 + 5 mm/sec 

for the CPT and DMT. Any pauses in penetration should be noted on a log. 

4.10. IN-SITU TESTING IN ROCK 

4.10.1 General 

In-situ testing to evaluate rock mass deformation modulus and shear strength is sometimes required 

for the design of foundations for major structures such as dams and bridges, however, such testing is 

not performed for structural foundations or slopes associated with typical highway applications. 

Circumstances where in-situ rock testing may be carried out for highway projects include 

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foundations comprising closely fractured and weak rock that could compress, resulting in settlement 

of the structure, or continuous, low strength discontinuities on which sliding could take place. The 

need for in-situ testing would arise when it is not possible to obtain undisturbed samples, or 

sufficiently large samples, for laboratory testing. This section provides an overview of specific insitu testing methods that may be used for rock (see table 11) and is based largely on information 

presented in Wyllie (1999). 

Three methods of in-situ deformation modulus and shear strength testing are described in this 

section. These include: (1) borehole dilatometer; (2) borehole jack; and (3) in-situ direct shear test. 

Table 11. In-situ testing methods used in rock. 

Method Procedure Rock 

Properties 

Limitations / Remarks 

Borehole 

Dilatometer 

The dilatometer is lowered to 

the test elevation and the 

flexible rubber membrane is 

expanded exerting a uniform 

pressure on the sidewalls of 

the borehole. 

Modulus of 

fractured rock 

mass 

Poisson’s ratio of rock must be 

assumed; test only affects a 

small area of the rock mass but 

several tests over the depth of 

influence can be performed 

Borehole Jack Jacks exert a unidirectional 

pressure to the walls of a 

borehole by means of two 

opposed curved steel platens 

Modulus of 

fractured rock 

mass 

Measured modulus value must 

be corrected to account for 

stiffness of steel platens; test 

method can be used to provide 

an estimate of anisotropy 

Plate Load Test Load is applied to a steel plate 

or concrete foundation using a 

system of hydraulic jacks and 

a reaction frame anchored to 

the foundation rock 

Modulus of 

fractured rock 

mass 

Loaded area is limited so may 

not be effectively testing rock 

mass if joints are widely 

spaced; modulus values 

corrected for plate geometry, 

effect of rock breakage, rock 

anisotropy, and steel plate 

modulus 

In-situ Direct Shear 

Test 

Testing is typically performed 

in an adit where reaction for 

the shear load is provided by 

an adit wall. Normal load 

applied via a jack system that 

uses the adit roof for reaction 

Peak and residual 

shear strength of 

discontinuity or 

discontinuity 

infilling 

Need to isolate a block of rock 

above the discontinuity surface 

without disturbing the infilling 

4.10.2 Borehole Dilatometer 

The borehole dilatometer is similar to the pressuremeter used in soil. At each testing depth, a 

uniform radial pressure is exerted on the walls of the drill hole by means of a flexible rubber sleeve. 

As with the pressuremeter, the volumetric expansion of the borehole can be measured by the 

inflation medium (generally oil or water) as the pressure is raised, or by electronic transducers that 

measure radial displacement of the inside of the sleeve. For the latter type, the measurement devices 

are generally arranged at right angles that further enables the anisotropy of the rock to be evaluated. 

The expansion volume of the borehole can be measured with a calibrated hand-operated screw 

pump. Alternatively, the volumetric expansion can be measured directly in the probe. 

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Because of the generally stiff character of most rock (even fractured rock), the hydraulic system 

should be relatively stiff and the system should be calibrated prior to and after testing. Figure 20 

shows typical pressure-dilation graphs for a calibration test carried out in a material of known 

modulus; this figure also shows the result of a test carried out in rock. A complete test usually 

consists of three loading and unloading cycles, with dilation and pressure readings being taken on 

both the loading and unloading cycles. 

Figure 20. Typical pressure-dilation graphs for a borehole dilatometer (after ISRM, 1987). 

4.10.3 Borehole Jack 

As an alternative to the flexible dilatometer and in rock that may be too stiff for the dilatometer, the 

borehole jack (ASTM D4971) can be used to measure rock mass deformability in a drill hole. The 

jack exerts a directional pressure by means of semi-cylindrical steel loading platens, with the 

deformation being measured with linear variable differential transformers (LVDTs) built into the 

jack. Calculation of the modulus is carried out in a similar manner to that of the dilatometer, except 

that allowance must be made for the more complicated boundary conditions (i.e., pressure is not 

exerted uniformly in the borehole) and the relative stiffness of the steel platens compared to the insitu rock mass. 

4.10.4 In-situ Direct Shear Testing 

In-situ direct shear testing is typically not performed due to the expense associated with the test 

setup. This test may be appropriate, however, for critical cases in which the shear strength of an 

undisturbed, potentially sensitive, infilling is required for design analyses. In cases where the 

infilling is displaced, it is likely that laboratory direct shear tests carried to residual conditions on 

recompacted infilling material would be sufficient. Figure 21 shows a schematic illustration of an 

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in-situ direct shear test performed in an adit. Testing can be conducted on a rock surface using 

cables anchored into the rock adjacent to the test site to supply the reaction for loading in the 

direction normal to the shear load. 

Legend: 

1. Rock anchor 

2. Hand-placed concrete 

3. Wide flange steel beam 

4. Hardwood 

5. Steel plates 

6. 30 ton jack 

7. Dial gauge 

8. Steel rollers 

9. Reinforced concrete pad 

10. Bearing plate 

11. Styrofoam 

12. 50 ton jack 

13. Steel ball 

Figure 21. Typical setup for an in-situ direct shear test in an adit (after Saint Simon et al., 1979). 

4.11 GEOPHYSICAL TESTING 

Geophysical testing is often used as part of the initial site exploration phase of a project and/or to 

provide supplementary information collected by widely-spaced observations (i.e., borings, test pits, 

outcrops etc.). Geophysical testing can be used for establishing stratification of subsurface 

materials, the profile of the top of bedrock, depth to groundwater, limits of types of soil deposits, 

rippability of hard soil and rock, and the presence of voids, buried pipes, and depths of existing 

foundations. Data from geophysical testing should always be correlated with information from 

direct methods of exploration. 

Geophysical testing offers some notable advantages and some disadvantages that should be 

considered before the technique is recommended for a specific application. The advantages are 

summarized as follows: 

• Many geophysical tests are non-invasive and thus offer significant benefits in cases where 

conventional drilling, testing, and sampling are difficult (e.g., deposits of gravel, talus 

deposits) or where potentially contaminated soils may occur in the subsurface. 

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• In general, geophysical testing covers a relatively large area, thus providing the opportunity 

to characterize large areas with few tests. It is particularly well-suited to projects that have 

large longitudinal extent compared to lateral extent (such as for new highway construction). 

• Geophysical measurement assesses the characteristics of soil and rock at very small strains, 

typically on the order of 0.001 percent thus providing information on truly elastic properties. 

• For the purpose of obtaining information on the subsurface, geophysical methods are 

relatively inexpensive when considering cost relative to the relatively large areas over which 

information can be obtained. 

Some of the general disadvantages of geophysical methods include: 

• Most methods work best for situations in which there is a large difference in stiffness 

between adjacent subsurface units. 

• It is difficult to develop good stratigraphic profiling if the general stratigraphy consists of 

hard material over soft material 

• Results are generally interpreted qualitatively and therefore useful results can only be 

obtained by an experienced engineer or geologist familiar with the particular testing method. 

• Specialized equipment is required (compared to more conventional subsurface exploration 

tools). 

There are a number of different geophysical in-situ tests that can be used for stratigraphic 

information and in the determination of engineering properties. Table 12 provides a summary of the 

various geophysical methods that are currently available in U.S. practice. Additional information on 

the procedures used for these methods is provided in AASHTO (1988), FHWA-HI-97-021 (1997), 

and Campanella (1994). Additional general discussion regarding the major headings in table 12 is 

presented, with particular emphasis on the potential application to highway engineering. 

• Seismic Methods: These methods are becoming increasingly popular for highway and 

general geotechnical engineering practice, as they have the potential to provide quantitative 

data regarding the shear wave velocity of the subsurface materials. The shear wave velocity 

is directly related to small-strain material stiffness, which in turn, is often correlated to 

strength and soil/rock type. As such, these techniques are often used for assessing the 

vertical stiffness profile in a soil deposit and for assessing the interface between soil and 

rock. 

• Electrical Methods: These methods are usually used when attempting to locate voids or 

locally distinct materials. With regards to highway applications, these procedures may be 

applicable for assessing the potential for karst activity along a potential transmission 

corridor, or for locating specific underground drums and/or voids. The techniques provide 

qualitative information only and are usually part of a two- or three-phased investigation 

program. 

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Table 12. Geophysical testing techniques. 

Method Basic Field Procedures Applications Li

SEISMIC METHODS Seismic Refraction Impact load is applied to the ground surface. 

Seismic energy refracts off soil/rock layer interfaces and is recorded on the ground surface using several dozen geophones positioned along a line or performing repeated events using a single geophone. 

• depth to bedrock 

• depth to water table • thickness and relative stiffness soil/rock 

layers 

• does not work 

depth or if soft 

• works best 

discontinuity is 

Spectral-Analysisof-Surface-Waves (SASW) (additional information provided in section 5.5.4.2) 

Impact load is applied to the ground surface. Surface waves propagate along ground surface and are recorded on the ground surface with two geophones positioned along a line. 

• depth to bedrock 

• measurement of shear wave velocity • thickness and stiffness of surface pavement 

layer 

• qualitative indicator of cracking in 

pavement 

• resolution dec

increasing dept

• accurate interp

significant amo

• interpretation i

overlies a soproperties are d

ELECTRICAL METHODSDC Resistivity DC current is applied to the ground using 

electrodes. Voltages are measured at different points on the ground surface with other electrodes positioned along a line. 

• depth to water table 

• inorganic groundwater contamination • groundwater salinity • soil layer thickness • delineation of certain vertical features 

(e.g., sinkholes, contamination plumes, waste trenches) 

• slow; must ins

the ground 

• resolution dec

increasing dept

• resolution is 

heterogeneous 

Electromagnetics Electrical coils are held over the ground. 

Current passing through the coils induces a magnetic field in the ground, which is measured with receiver coils. 

• groundwater salinity 

• inorganic groundwater contamination • detection of buried metal objects • delineation of certain vertical features 

(e.g., sinkholes, contamination plumes, waste trenches) 

• extra effort is 

depth of target 

• resolution dec

increasing dept

Ground Penetrating Radar (GPR) 

Electromagnetic energy is pulsed into the ground. This energy reflects off boundaries between different soil layers and is measured at the ground surface. 

• depth to water table 

• identification of buried objects • thickness of pavement layers • void detection 

• not effective b

clay 

• depth of penetr

meters 

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Table 12. Geophysical testing techniques (continued). 

Method Basic Field Procedures Applications Li

GRAVITY AND MAGNETIC METHODSGravity The Earth’s gravitational field is measured at the 

ground surface. 

• identification of subsurface voids • identification of large objects possessing 

unusually high or low densities 

• results are non-

subsurface conresults) 

• primarily, large

applications in 

Magnetics The Earth’s magnetic field is measured at the 

ground surface. 

• identification of ferrous materials 

• identification of soil/rock containing large 

amounts of magnetic minerals 

• results are non-

subsurface conresults) 

• primarily a l

tool; applicatilimited 

NEAR-SURFACE NUCLEAR METHODS Neutron Moisture Content 

Instrument is placed on the ground surface and neutrons are emitted into the ground. Energy of returning neutrons is related to the moisture content in the ground (hydrogen atoms decrease the energy of the neutrons detected at the sensor). 

• estimate of water content in compacted soil 

• estimate of asphalt content in asphalt 

concrete 

• can be quantitative if properly calibrated to 

site conditions 

• limited investig• possible healt

operators not p

• will detect hyd

non-water bear

Gamma Density Instrument is placed on the ground surface and 

gamma radiation is emitted into the ground. Returning gamma energy is a function of material density (denser materials absorb more gamma energy so less is detected at the sensor) 

• estimate of density of soil or asphalt 

concrete 

• limited investig

foot); 

• investigation d

few inches if g

• possible healt

operators not p

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Table 12. Geophysical testing techniques (continued). 

Method Basic Field Procedures Applications Li

BOREHOLE METHODSCrosshole/ Downhole (additional information provided in section 5.5.4.2) 

Energy sources and geophones are placed in boreholes and/or on ground surface; interval travel times are converted into seismic wave velocity as a function of depth in the borehole 

• measurement of wave velocities for 

seismic site response analysis 

• depth to water table • correlation of lithologic units with surface 

seismic 

• identification of thin layers at depth 

• requires one 

significant supp

Suspension Logger Field instrument is placed in a fluid-filled 

borehole and used to measure P- and S-wave velocities in surrounding soil or rock. 

• measurement of wave velocities for 

seismic site response analysis 

• correlation of lithologic units with surface 

seismic 

• identification of thin layers at depth 

• requires boreho

field equipmen

• borehole must b

Electrical Logging Field instrument is placed in a borehole. 

Electrical fields are directly applied or electromagnetically induced into surrounding soil or rock and electrical resistivity is measured. 

• estimate of soil/rock permeability or 

porosity 

• identification of inorganic contaminant 

plumes or saltwater intrusion 

• identification of thin layers at depth 

• requires boreho

field equipmen

• generally cann

borehole 

• may require flu• results may be 

mud salinity 

Nuclear Logging Field instrument is placed in a borehole. 

Surrounding soil or rock is irradiated with neutrons particles and/or gamma energy. Energy and neutrons returning to the instrument are measured and related to rock density, porosity and pore fluid type. 

• estimate of soil/rock type, density, 

porosity, and pore fluid density 

• identification of thin layers at depth 

• requires boreho

field equipmen

• possible healt

operators are n

Lithology Logging Field instrument is placed in a borehole; 

naturally occurring electrical fields and radiation levels are related to soil or rock type 

• classification of soil or rock type 

• identification of thin layers at depth 

• requires boreho

field equipmen

• may require flu• results are dep

conditions and/

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• Gravity and Magnetic Methods: These methods are similar to the previously described 

electrical methods, except that they rely on the correlations between the influence of voids 

and subsurface anomalies and differences in the earth’s micro-gravitational field and/or the 

magnetic fields, rather than the changes in the electrical fields. 

Near-surface Nuclear Methods: These techniques have been used for several years in the field 

of soil construction. Through careful calibration, it is possible to reliably assess the 

moisture content and density of compacted soils. These techniques have gained widespread 

adoption as reliable quantitative techniques 

Borehole Methods: Downhole geophysical techniques have been recognized as providing 

reliable indications of a wide range of soil properties. The downhole/crosshole techniques 

have proven to provide reliable measure of shear wave velocity. As reported previously, 

this parameter is directly related to small-strain stiffness and is correlated to strength and 

soil type. The downhole logging techniques have seen little use in highway construction, 

but they have been the mainstay for deep geologic characterization in oil exploration and 

deep geologic characterization. The principal advantage is the ability to obtain several 

different geophysical tests/ indicators by “stringing” these tools together in a deep boring. 

With specific regards to highway construction, the authors have identified a few typical examples 

where geophysical testing could be used to compliment conventional exploration. 

Highly Variable Subsurface Conditions: In several geologic settings, the subsurface conditions 

along a transportation corridor may be expected to be variable. This variability could be 

from underlying karst development above limestone, alluvial deposits, including buried 

terrace gravels, across a wide floodplain, buried boulders in a talus slope. For these cases, 

conventional exploration techniques may be very difficult and if “refusal” is encountered at 

one depth, there is a strong likelihood that different materials underlie the region. In these 

cases, a preliminary subsurface characterization profile using geophysical testing could 

prove advantageous in designing future focused investigations. 

Regional Studies: Along a transmission corridor it may be necessary to assess the depth to (and 

through) rippable rock. Alternative alignments may or may not be possible, but the cost 

implications may be significant. Therefore, it is important to obtain a profile related to 

rock/soil stiffness. Geophysical testing is a logical consideration for this application, as a 

precursor to invasive investigation. 

Settlement Sensitive Structures: The prior two examples related to cases where the geophysical 

testing served as the front–end of a multi-phase project. In the case where a settlementsensitive structure is to be founded on deposits of sands, the in situ modulus of the sand 

deposit is critical. After assessing the characteristics of the site, it may be helpful to 

quantify the deformation modulus via geophysical testing at the specific foundation site. 

These examples demonstrate that geophysical testing has a potentially important role in the 

subsurface characterization of soils and rocks. Like the other “tools” described in this document, the 

particular selection of the appropriate technology is very much a function of the site conditions and 

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the goals of the characterization program. In this document, focus is placed on geophysical testing 

techniques that can be used to measure soil shear wave velocity, Vs, such as seismic refraction, 

SASW, and seismic cone penetrometer. The relevance and evaluation of this parameter for static 

deformation analysis of various geotechnical structures is provided in chapter 5. 

4.12 LABORATORY SOIL TESTING 

4.12.1 Introduction 

Laboratory testing of soil samples recovered during subsurface explorations is the most common 

technique to obtain engineering properties necessary for design. A laboratory-testing program 

consists of index tests to obtain general information on material consistency and performance tests to 

measure specific properties (e.g., shear strength, compressibility, hydraulic conductivity) for design 

and constructability assessments. This section provides information on common laboratory test 

methods for soil including testing equipment, general procedures related to each test, and parameters 

measured by the tests. A discussion of the interpretation of these tests to obtain properties is 

provided in chapter 5. 

Laboratory soil testing will be required for most projects and it is therefore necessary to 

appropriately select the types and quantities of laboratory tests to be performed. A careful review of 

all data obtained during the field investigation is essential to developing an appropriately scoped 

laboratory-testing program. In some cases, owners may wish to hire external testing laboratories to 

perform select tests. It is necessary that testing requests be clear and sufficiently detailed. Unless 

specialized testing is required, the owner should require that all testing be performed in accordance 

with appropriate specification for laboratory testing such as those codified in AASHTO and ASTM. 

Table 13 provides a listing of commonly-performed laboratory tests. Tables 14 and 15 provide a 

summary of typical soil index and performance tests, respectively. Additional information on these 

tests is provided in subsequent sections. 

4.12.2 Quality Assurance for Laboratory Testing 

4.12.2.1 Sample Tracking 

Whether the laboratory testing is performed in-house or is subcontracted, samples will likely be 

assigned a laboratory identification number that differs from the identification number assigned in 

the field. A list should be prepared which matches the laboratory identification number with the 

field identification number. This list can also be used to provide tracking information to ensure that 

each sample arrived at the lab. When requesting laboratory testing, both the field identification 

number and the laboratory identification number should be used on the request form. A spreadsheet 

or database program is useful to manage sample identification data. 

Archived Document
Table 13. Common soil laboratory tests. 

75

Test 

Category Name of Test AA

Visual Identification 

Practice for Description and Identification of Soils (Visual-Manual Procedure) 

Practice for Description of Frozen Soils (Visual-Manual Procedure) 

Index Properties Test Method for Determination of Water (Moisture) Content of Soil by Direct Heating Method 

Test Method for Specific Gravity of Soils Method for Particle-Size Analysis of Soils Test Method for Classification of Soils for Engineering Purposes 

Test Method for Amount of Material in Soils Finer than the No. 200 (75-μm) Sieve 

Test Method for Liquid Limit, Plastic Limit, and Plasticity Index of Soils T 

Compaction Test Method for Laboratory Compaction Characteristics of Soil Using Standard Effort (600 kN. m/m3) 

Test Method for Laboratory Compaction Characteristics of Soil Using Modified Effort (2,700 kN.m/m3) 

Strength Properties Test Method for Unconfined Compressive Strength of Cohesive Soil 

Test Method for Unconsolidated, Undrained Compressive Strength of Cohesive Soils in Triaxial Compression 

Test Method for Consolidated, Undrained Compressive Strength of Cohesive Soils in Triaxial Compression 

Method for Direct Shear Test of Soils under Consolidated Drained Conditions Test Methods for Modulus and Damping of Soils by the Resonant-Column Method Test Method for Laboratory Miniature Vane Shear Test for Saturated Fine-Grained Clayey Soil 

Test Method for CBR (California Bearing Ratio) of Laboratory-Compacted Soils 

Test Method for Resilient Modulus of Soils Test Method for Resistance R-Value and Expansion Pressure of Compacted Soils 

Test Method for One-Dimensional Consolidation Properties of Soils Test Method for One-Dimensional Consolidation Properties of Soils Using Controlled-Strain Loading Test Methods for One-Dimensional Swell or Settlement Potential of Cohesive Soils 

Consolidation and Swelling Properties Test Method for Measurement of Collapse Potential of Soils 

Permeability Test Method for Permeability of Granular Soils (Constant Head) 

Test Method for Measurement of Hydraulic Conductivity of Saturated Porous Materials Using a Flexible Wall Permeameter 

Corrosivity Test Method for pH for Peat Materials 

Test Method for pH of Soils Test Method for pH of Soil for Use in Corrosion Testing Test Method for Sulfate Content Test Method for Resistivity Test Method for Chloride Content 

Organic Content Test Methods for Moisture, Ash, and Organic Matter of Peat and Other Organic Soils 

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Table 14. Methods for index testing of soils. 

Test Procedure Applicable 

Soil Types 

Applicable Soil 

Properties 

Limitations / Remarks 

Moisture 

Content, wn

Dry soil in oven at 100 + 5 o

C Gravel, sand, 

silt, clay, peat 

eo, γ Simple index test for all 

materials 

Unit 

Weight and 

Density 

Extract a tube sample; measure 

dimensions and weight; 

Soils where 

undisturbed 

samples can 

be taken, i.e., 

silt, clay, peat 

γtot, γdry, ρtot, ρdry, σvo Not appropriate for clean 

granular materials where 

undisturbed sampling is not 

possible. Very useful index 

test. 

Atterberg 

Limits, LL, 

PL, PI, SL, 

LI 

LL – Moisture content associated with 

failure at 25 blows of specimen in 

Casagrande cup 

PL – Moisture content associated with 

crumbling of rolled soil at 3.2 mm 

Clays, silts, 

peat; silty and 

clayey sands 

to determine 

whether SM 

or SC 

Soil classification Not appropriate in nonplastic granular soil. 

Recommended for all 

plastic materials. 

Mechanical 

Sieve 

Place air dry material on a series of 

successively smaller screens of known 

opening size and vibrate to separate 

particles of a specific equivalent 

diameter 

Gravel, sand, 

silt 

Soil classification Not appropriate for clay 

soils. Useful, particularly 

in clean and dirty granular 

materials 

Wash Sieve Flush fine particles through a U.S. No. 

200 sieve with water; 

Sand, silt, 

clay 

Soil classification Needed to assess fines 

content in dirty granular 

materials 

Hydrometer Allow particles to settle, and measure 

specific gravity of the solution with 

time. 

Fine sand, 

silt, clay 

Soil classification Helpful to assess relative 

quantity of silt and clay 

Specific 

Gravity 

The volume of a known mass of soil is 

compared to the known volume of 

water in a calibrated pyncnometer 

Sand, silt, 

clay, peat 

Used in calculation of eo Particularly helpful in cases 

where unusual solid 

minerals are encountered 

Organic 

Content 

After performing a moisture content 

test at 110 o

C, the sample is ignited in 

a muffle furnace at 440 o

C to measure 

the ash content. 

All soil types 

where organic 

matter is 

suspected to 

be a concern 

Not related to any specific 

performance parameters, 

but samples high in 

organic content will likely 

have high compressibility 

Recommended on all soils 

suspected to contain 

organic materials 

Symbols used in table 14.

eo: in-situ void ratio 

γ: unit weight 

γtot: total unit weight 

γdry: dry unit weight 

ρtot: total density 

ρdry: dry density 

σvo: total vertical stress 

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Table 15. Methods for performance testing of soils. 

Test Procedure Applicable 

Soil Types 

Soil 

Properties 

Limita

1-D Oedometer Incremental loads are applied to a soil specimen 

confined by a rigid ring; deformation values are recorded with time; loads are typically doubled for each increment and applied for 24 hours each. 

Primarily clays and silts; Granular soils can be tested, but typically are not. 

σp′, OCR, Cc, 

Ccε, Cr, Crε, Cα,, Cαε, cv, k 

Recommended for fine useful index to other cri

Constant rate of Strain Oedometer 

Loads are applied such that ∆u is between 3 and 30 

percent of the applied vertical stress during testing 

Clays and silts; Not applicable to free draining granular soils. 

σp′, Cc, 

Ccε, Cr, Crε, cv, k 

Requires special testing testing time significantl

Unconfined Compression (UC) 

A specimen is placed in a loading apparatus and sheared under axial compression with no confinement. 

Clays and silts; cannot be performed on granular soils or fissured and varved materials 

su,UC Provides rapid means to 

strength, but disturbancmigration will effect res

Unconsolidated Undrained (UU) Triaxial Shear 

The specimen is not allowed to consolidate under the confining stress, and the specimen is loaded at a quick enough rate to prevent drainage 

Clays and silts su,UU Sample must be nearly 

and rate effects will affe

Isotropic consolidated drained compression (CIDC) 

The specimen is allowed to consolidate under the confining stress, and then is sheared at a rate slow enough to prevent build-up of porewater pressures 

Sands, silts, clays φ′, c′, E Can be run on clay spec

triaxial test to obtain de

Isotropic consolidated undrained compression (CIUC) 

The specimen is allowed to consolidate under the confining stress with drainage allowed, and then is 

sheared with no drainage allowed, but porewater pressures measured 

Sands, silts, clays, peats φ′, c′, su,CIUC, 

E 

Recommended to measuUseful test to assess eff

Not for measuring defor

Direct Shear The specimen is sheared on a forced failure plane at 

a constant rate, which is a function of the hydraulic 

conductivity of the specimen 

Compacted fill materials; sands, silts, and clays 

φ′, φ′r Requires assumption of 

easy strength test. 

Flexible Wall Permeameter 

The specimen is encased in a membrane, consolidated, backpressure saturated, and measurements of flow with time are recorded for a specific gradient 

Relatively low permeability materials (k < 1x10-5 cm/s); clays & 

silts 

k Recommended for fine 

saturation required. Coprovided. System permof magnitude greater thneeded to allow inflow 

Rigid Wall Permeameter The specimen is placed in a rigid wall cell, vertical 

confinement is applied, and flow measurements are recorded with time under constant head or falling 

head conditions 

Relatively high permeability materials; sands, gravels, and silts 

k Need to control gradien

soils. Monitor for sidew

Symbols used in table 15. φ′: peak effective stress friction angle OCR: overconsolidation ratio Ccε: modified compression index φ′r residual effective stress friction angle cv: vertical coefficient of consolidation Cr: recompression index c′: effective stress cohesion intercept E: Young’s modulus Crε: modified recompression index su: undrained shear strength k: hydraulic conductivity Cα: secondary compression index σp′: preconsolidation stress Cc: compression index Cαε: modified secondary compression index 

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4.12.2.2 Sample Storage 

Undisturbed soil samples should be transported and stored so that that the moisture content is 

maintained as close as possible to the natural conditions (AASHTO T 207, ASTM D 4220 and 5079). 

Samples should not be placed, even temporarily, in direct sunlight. Undisturbed soil samples should be 

stored in an upright position with the top side of the sample up. 

As storage time increases, moisture will migrate within a tube. Potential for disturbance and moisture 

migration within the sample will increase with time, and samples tested after 30 days should be noted on 

the laboratory data sheet. Excessive storage time can lead to additional sample disturbance that will 

affect strength and compressibility properties. Additionally, stress relaxation, temperature changes, and 

storage in a room with humidity below 90 percent will have detrimental effects to the samples. Longterm storage of soil samples should be in temperature and humidity controlled environments. The 

temperature control requirements may vary from sub freezing to ambient and above, depending on the 

environment of the parent formation. The relative humidity for soil storage normally should be 

maintained at 90 percent or higher. 

Long-term storage of soil samples in sampling tubes is not recommended. During long term storage the 

sample tubes may corrode. This accompanied by the adhesion of the soil to the tube may develop such 

resistance to extrusion that some soils may experience internal failures during extrusion. Often these 

failures cannot be seen by the naked eye; x-ray radiography (ASTM D 4452) will likely be necessary to 

confirm the presence of such conditions. If these samples are tested as “undisturbed” specimens, the 

results may be misleading. 

4.12.2.3 Sample Handling 

Careless handling of undisturbed soil samples may cause major disturbances that could lead to serious 

design and construction consequences. Samples should always be handled by experienced personnel in 

a manner that ensures that the sample maintains structural integrity and natural moisture condition. 

Saws and knives used to prepare soil specimens should be clean and sharp. Preparation time should be 

kept to a minimum, especially where the maintenance of the moisture content is critical. Specimens 

should not be exposed to direct sun, freezing, or precipitation. 

4.12.2.4 Specimen Selection 

The selection of representative specimens for testing is one of the most important aspects of sampling 

and testing procedures. Selected specimens must be representative of the formation or deposit being 

investigated. The senior laboratory technician, the geologist and/or the geotechnical engineer should 

study the drilling logs, understand the geology of the site, and visually examine the samples before 

selecting the test specimens. Samples should be selected on the basis of their color, physical 

appearance, and structural features. Specimens should be selected to represent all types of materials 

present at the site, not just the worst or the best. Samples with discontinuities and intrusions may 

prematurely fail in the laboratory. If these features are small and randomly located, however, they 

would not necessarily cause such failures in the field. Such local failures should be noted but not 

selected as representative of the deposit. 

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4.12.3 Effects of Sample Disturbance 

As a result of drilling, sampling, sample extrusion, and trimming to form a specimen for testing, 

nominally undisturbed specimens from samples obtained using methods presented in table 7 will 

become disturbed. These processes change the effective stress condition in the soil sample; that is, the 

effective stress in the soil at the time after a sample is trimmed and prepared for testing is different than 

that of the same soil in the ground. Figure 22 provides an illustration of the stress changes that a soil 

undergoes as it is transferred from the ground (given by point A at an at-rest (Ko) condition) to a 

laboratory specimen. At point F in figure 22, significant stress relief has occurred, and the specimen is 

at a higher void ratio than the in-situ condition. Since the laboratory sample is at a higher void ratio 

(i.e., less dense) than it was in-situ, a laboratory strength test performed on the specimen would likely 

exhibit a lower strength and a higher compressibility relative to the same soil at the in-situ void ratio. 

Figure 22. Disturbance during sampling and trimming (after Ladd and Lambe, 1963). 

The effects of disturbance listed above provide a basis for evaluating laboratory shear strengths from 

unconsolidated-undrained (UU) tests and consolidated-undrained (CIU) strength tests. It should be 

noted that with CIU tests there is some recompression and restoration of void ratio, while results of a 

UU test will be much more affected by disturbance induced changes in void ratio and structure of the 

specimen. Figure 23 illustrates the “field” or undisturbed consolidation curve for a normally 

consolidated sample as well as the disturbed curve developed from laboratory data. The void ratio in the 

ground is given by point A. A UU test on a disturbed specimen will be performed at a void ratio shown 

by point B since the sample will be unloaded prior to testing, whereas for a CIU test on a disturbed 

sample, in which the sample is reconsolidated to the in-situ effective stress, the void ratio will be at point 

C. Since soil strength is a function of void ratio, among other properties, it can be seen that the UU 

strength may be less than that for the in-situ sample and the CIU strength may be greater. In summary, 

for both a UU and a CIU test, if the sample is disturbed it will not be possible to reestablish in the 

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laboratory the unique void ratio and effective stress combination that existed in the ground, although 

clearly the CIU test can provide a much better representation of the in situ soil conditions as compared 

to a UU test. Efforts must therefore concentrate on obtaining high quality, undisturbed samples for 

laboratory strength testing. 

Additional test-related issues such as direction of shearing, sample disturbance, and rate effects, will 

play a role in differences between laboratory and field values of strength. For example, shear rates used 

in a UU test are much faster than that which would be expected for most field situations. In general, the 

faster a clay sample is sheared under undrained conditions, the greater will be the measured undrained 

strength. Based on this discussion, it would seem as though disturbance and shear rate produce 

“compensating errors” in the UU test. Also, many engineers assume that the disturbance effect is 

greater than the shear rate effect leading to the conclusion that UU strengths must be less than the actual 

undrained strength of the soil so that characterization of undrained strengths from UU tests is therefore 

conservative. Such a generalization should not be relied upon when selecting undrained strengths for 

design. Undrained strengths based on UU testing must be supplemented with additional CIU testing 

and/or in situ testing. Additional discussion on the selection of shear strength values based on UU and 

CIU testing is provided in chapter 5. 

log σvo’

Void Ratio, e C

σvo’

A

B

Figure 23. Lab and field consolidation curves. 

Based on the previous discussion, it is clear that minimization of disturbance during sampling and 

testing is required. Also, since some disturbance is inevitable, procedures and “rules-of-thumb” are 

available to assess sample disturbance and the potential effects of this disturbance on soil properties. 

Quantification of sample disturbance in cohesive soils can be assessed using: (1) laboratory odometer 

curves; and (2) radiographic testing. 

Sample disturbance will generally affect laboratory oedometer curves in three ways (see figure 23): 

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• The strain to reach the in-situ effective stress will be greater for the disturbed sample when 

compared to the strain (or change in sample height) of an undisturbed specimen (compare point 

C to point A in figure 23). 

• The preconsolidation stress is difficult to determine due to rounding of the consolidation curve. 

This generally leads to an under estimation of the preconsolidation stress. 

• The recompression curve for the disturbed sample will have a steeper slope than that of an 

undisturbed specimen, while the apparent virgin compression curve will likely be flatter than 

that of the undisturbed specimen. 

Depending upon the stress history of the deposit and the stress ranges of the design, sample disturbance 

can effect the selection of soil properties to different degrees. As shown in table 16, the quality of a 

specimen can be assessed based on results of oedometer tests. Additionally, the quality of a specimen 

prepared for strength testing can be assessed if the strain to reach σvo′ during the consolidation phase is 

monitored and an estimate of OCR for the specimen is known. Table 16 attempts to correlate 

disturbance to the initial elastic stiffness of a sample. That is, for a given OCR, the loading of the 

sample up to the preconsolidation stress is “elastic” and the response should be relatively stiff. As the 

amount of strain required to reach the initial effective stress in the ground increases, so does the 

likelihood that the sample is disturbed. 

Table 16. Sample quality designation system (Lacasse et al., 1985). 

Range 

of 

OCR 

Depth 

Interval 

(m) 

Very good 

test if 

εv < 

(%) 

Acceptable 

test if 

< εv < 

(%) 

Likely to be very 

disturbed if 

εv > 

(%) 

1 – 1.2 0 – 10 

10 – 50 

3 

2 

3 – 5 

2 – 4 

5 

4 

1.2 – 1.5 0 – 10 

10 – 50 

2 

1 

2 – 4 

1 – 3 

4 

3 

1.5 – 2 0 – 10 

10 – 50 

1.5 

1.0 

1.5 – 3.5 

1 – 2.5 

3.5 

2.5 

2 – 3 0 – 10 

10 – 50 

1.0 

0.75 

1 – 3 

0.75 – 2 

3 

2 

3 – 8 0 – 10 

10 – 50 

0.5 

0.5 

0.5 – 1 

0.5 – 2 

1 

2 

εv = strain required to reach σvo′

X-ray photographs of soil specimens can be used to assess sample quality. Radiography (ASTM D 

4452) utilizes X-ray photographs to assess density variation or consistency of a sample, and thus identify 

potential areas of defects and disturbance. X-ray photographs may be taken on samples within tubes or 

liners, or on extruded samples. Radiography can be used to identify: 

• variation in soil types; 

• macrofabric features such as bedding planes, varves, fissures, and shear planes; 

• presence of intrusions such as gravel, shells, calcareous soils, peat, and drilling mud; 

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• presence of voids and cracks; and 

• variation in the degree of disturbance that may range from curvature of soil layers near the tube 

edges to extreme disturbance noted by large voids and cracks (typically at the end of the tubes). 

Since these features are often within the sample and not apparent from visual identification, radiography 

provides a non-destructive means for selecting representative samples for laboratory performance 

testing. Radiography is particularly useful where a limited number of samples are available for testing 

or complexities in sampling are likely to induce disturbance. Radiographic testing requires special 

testing equipment (usually from an outside laboratory) but the testing is not expensive. The 

radiographic images provide information to ensure that high quality samples are used for laboratory 

performance tests. 

4.12.4 Laboratory Index Tests for Soils 

4.12.4.1 General 

Data generated from laboratory index tests provide an inexpensive way to assess soil consistency and 

variability among samples collected from a site. Information obtained from index tests is used to select 

samples for engineering property testing as well as to provide an indicator of general engineering 

behavior (e.g., high plasticity clay based on plasticity index (PI) may indicate high compressibility, low 

hydraulic conductivity, and high swell potential). Common index tests discussed in this section include 

moisture content, unit weight (wet density), Atterberg limits, particle size distribution, visual 

classification, specific gravity, and organic content. Index testing should be conducted on soil materials 

from every project. Information from these tests should be assessed prior to a final decision regarding 

the specimens selected for subsequent performance testing. 

4.12.4.2 Moisture Content 

The moisture (or water) content test is one of the simplest and least expensive laboratory tests to 

perform. Moisture content is defined as the ratio of the mass of the water in a soil specimen to the dry 

mass of the specimen. Natural moisture contents (wn) of sands are typically 0 ≤ wn ≤ 20 %, whereas for 

inorganic and insensitive silts and clays, general ranges are: 10 ≤ wn ≤ 40 %. However, depending upon 

the mineralogy, formation environment, and structure of clay, it is possible to have more water than 

solids (i.e., w > 100%). Therefore soft and highly compressible clays, as well as sensitive, quick, or 

organically rich clays, can exhibit water contents 40 ≤ wn ≤ 300 % or more. 

Moisture content can be tested in a number of different ways including: (1) a drying oven (ASTM D 

2216); (2) a microwave oven (ASTM D 4643); or (3) a field stove or blowtorch (ASTM D 4959). While 

the microwave or field stove (or blowtorch) methods provide a rapid evaluation of moisture content, 

potential errors inherent with these methods require confirmation of results using ASTM D 2216. The 

radiation heating induced by the microwave oven and the excessive temperature induced by the field 

stove may release water entrapped in the soil structure that would normally not be released at 110o

 C, 

yielding higher moisture content values than would occur from ASTM D 2216. 

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Field measurements of moisture content often rely on a field stove or microwave, due to the speed of 

testing. When dealing with compacted material, it is common to use a nuclear gauge (ASTM D 3017) in 

the field to rapidly assess moisture contents. It is noted that nuclear gage readings may indicate widely 

varying moisture contents for micaceous soils. Results from these techniques should be “calibrated” or 

confirmed using the drying oven (ASTM D 2216). 

Sampling, handling, and storage may alter the in-situ moisture content tests. Because the top end of the 

sample tube may contain water or collapse material from the borehole, moisture content tests should not 

be performed on material near the top of the tube. Also, as storage time increases, moisture will migrate 

within a specimen and lead to altered moisture content values. If the sample is not properly sealed, 

drying of the sample and moisture loss will likely occur. 

4.12.4.3 Unit Weight 

In the laboratory, soil unit weight and mass density are easily measured on tube samples of natural soils. 

The moist (total) mass density is ρt = Mt/Vt, whereas the dry mass density is given by ρd = Ms/Vt. The 

moist (total) unit weight is γt = Wt/Vt, whereas the dry unit weight is defined as γd = Ws/Vt. The 

interrelationship between the total and dry mass density and unit weight is given by: 

ρd = ρt/(1+wn) (Equation 17) 

and the relationship between total and dry unit weight is given by: 

γd = γt/(1+wn) (Equation 18) 

The terms density and unit weight are often incorrectly used interchangeably. The correct usage is that 

density implies mass measurements while unit weight implies weight measurements. For this document 

they will be referenced as “density (unit weight)”, if the usage is independent of the specific definition. 

Field measurements of soil mass density (unit weight) are generally restricted to shallow surface 

samples, usually when placing compacted fills, and can be accomplished using drive tubes (ASTM D 

2937), sand cone method (ASTM D 1556), or nuclear gauge (ASTM D 2922). To obtain unit weights or 

mass densities with depth, either high-quality thin-walled tube samples must be obtained (ASTM D 

1587), or relatively expensive geophysical logging by gamma ray techniques (ASTM D 5195) can be 

employed. 

4.12.4.4 Atterberg Limits 

The Atterberg limits of a fine grained (i.e., clayey or silty) soil represent the moisture content at which 

the behavior of the soil changes. The tests for the Atterberg limits are referred to as index tests because 

they serve as an indication of several physical properties of the soil, including strength, permeability, 

compressibility, and shrink/swell potential. These limits also provide a relative indication of the 

plasticity of the soil, where plasticity refers to the ability of a silt or clay to retain water without 

changing state from a semi-solid to a viscous liquid. In geotechnical engineering practice, the Atterberg 

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limits generally refers to the liquid limit (LL), plastic limit (PL), and shrinkage limit (SL). These limits 

are defined below. 

• Liquid Limit (LL) - This upper limit represents the moisture content at which any increase in 

moisture content will cause a plastic soil to behave as a liquid. The LL is defined as the 

moisture content at which a standard groove cut in a remolded sample will close over a distance 

of 1⁄2 inch at 25 blows of the liquid limit device (figure 24). 

• Plastic Limit (PL) - This limit represents the moisture content at which the transition between 

the plastic and semisolid state of a soil. The PL is defined as the moisture content at which a 

thread of soil just crumbles when it is carefully rolled out to a diameter of 3.2 mm. 

• Shrinkage Limit (SL) – The moisture content corresponding to the behavior change between the 

semisolid to solid state of the soil. The SL is also defined as the moisture content at which any 

further reduction in moisture content will not result in a decrease in the volume of the soil. 

A measure of a soils plasticity is the plasticity index (PI) which as calculated as PI = LL – PL. 

The PI is a useful index since numerous engineering correlations have been developed relating 

PI to clay soil properties, including undrained and drained strength and compression index. 

Results are typically presented on Casagrande’s Plasticity chart (see figure 25). On this chart, 

the equation for the A-line and U-line are, respectively: 

− PIlineA = LL − )20(73.0: (Equation 19) 

LLPIlineU −=− )8(9.0: (Equation 20) 

Figure 24. Equipment used for Atterberg limits testing of soil. 

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Figure 25. Location of clay minerals on the Casagrande plasticity chart (Skempton, 1953). 

Other indices based on the Atterberg Limits include the liquidity index (LI) and the activity (A) of a soil. 

These are defined as: 

LI = (wn-PL) / PI (Equation 21) 

A = PI / CF (Equation 22) 

where wn is the moisture content of the soil and CF is the clay fraction that corresponds to the 

percentage of particles exhibiting an equivalent diameter (ds) < 0.002 mm. The use of the liquidity 

index and activity can provide very useful information concerning the likely behavior of a soil, even 

though Atterberg limits are performed on completely remolded materials. For example, a LI less than or 

equal to zero is generally indicative of a heavily overconsolidated soil that may be desiccated or highly 

expansive. A soil with a LI equal to unity implies that the soil is at its liquid limit and is likely to be 

relatively weak and compressible. A LI greater than unity indicates that the soil is sensitive. Soils with 

a LI greater than approximately 0.7 will likely undergo significant consolidation settlements when 

loaded. 

Values for activity can be correlated to the type of clay mineral that, in turn, provides important 

information relative to the expected behavior of a clay soil. A clay soil that predominantly comprises 

montmorillonite behaves very differently from a clay soil comprising kaolinite. Figure 25 also shows 

the activities of various clay minerals and location on the Casagrande’s plasticity chart. 

A=1.5 (Calcium) 

A=4 to 7 (Sodium)

A = 0.5 to 1.3 

A=0.3 to 0.5 

A=0.1 (hydrated) 

A=0.5 (dehydrated) 

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4.12.4.5 Particle Size Distribution 

Particle size distribution by mechanical sieve and hydrometer are useful for soil classification purposes. 

Procedures for grain size analyses are contained in ASTM D 422 and AASHTO T88. Testing is 

accomplished by placing air-dried material on a series of screens of known opening size. Each 

successive screen has a smaller opening to capture progressively smaller particles. Testing of the finer 

grained particles is accomplished by suspending the chemically dispersed particles in water column and 

measuring the specific gravity of the liquid as the particles fall from suspension. 

Particle size testing is relatively straightforward, but results can be misleading if procedures are not 

performed correctly and equipment is not maintained in good condition. If the sieve screen is distorted, 

large particles may be able to pass through sieve openings that typically would retain the particles. 

Material lodged within the sieve from previous tests could become dislodged during shaking, increasing 

the weight of material retained on the following sieve, thus sieves should be cleaned after each test. A 

wire brush may distort finer sieve meshes during cleaning, so a plastic brush should be used to clean the 

U.S. No. 40 sieve and finer. Table 17 shows U.S. standard sieve sizes and associated opening sizes. 

Table 17. U.S. standard sieve sizes and corresponding opening dimension. 

U.S. Standard 

Sieve No. 

Sieve Opening 

(mm) 

Sieve Opening 

(in.) 

3 6.35 0.25 

4 4.75 0.187 

6 3.35 0.132 

8 2.38 0.0937 

10 2.00 0.0787 

12 1.68 0.0661 

16 1.20 0.0469 

20 0.85 0.0331 

30 0.60 0.0232 

40 0.425 0.0165 

50 0.30 0.0117 

60 0.25 0.0098 

70 0.21 0.0083 

100 0.15 0.0059 

140 0.106 0.0041 

200 0.075 0.0029 

270 0.053 0.0021 

400 0.0375 0.0015 

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Representative samples with fines (particles with diameter less than 0.075 mm or the U.S. No. 200 

sieve) should not be oven dried prior to testing because some particles may cement together leading to a 

calculated lower fines content from mechanical sieve analyses than is actually present. When finegrained particles are a concern, a wash sieve (ASTM D 1140) should be performed to assess the fines 

content. Additionally, if the clay content is an important parameter, hydrometer analyses need to be 

performed. It should be noted that the hydrometer test provides approximate analysis results due to 

oversimplified assumptions, but the obtained results can be used as a general index of silt and clay 

content. Depending upon the chemical makeup of the fine grained particles, the traditional sodium 

hexametaphosphate solution used to disperse the clay particles may not provide adequate dispersion. If 

the clay particles are not dispersed, the hydrometer data would lead to the interpretation of a low clay 

content. In some cases the concentration of dispersing agent may need to be increased or a different 

dispersing agent may need to be used. 

4.12.4.6 Laboratory Classification 

In addition to field identification (ASTM D 2488), soils should be classified in the laboratory using the 

Unified Soil Classification System (USCS) in accordance with ASTM D 2487 or the AASHTO soil 

classification system (AASHTO T 145). The differences in these two systems are beyond the scope of 

this document, and the reader is referred to Holtz and Kovacs (1981) and AASHTO (1988) for a 

comparison of these two classification systems. The USCS will be used throughout the remainder of 

this document. Classification in the laboratory occurs in a controlled environment and more time can be 

spent on this classification than the identification exercise performed in the field. Laboratory and/or 

field identification is important so that defects and features of the soil can be recorded that would not 

typically be noticed from index testing or standard classification. Some of the features include mica 

content, joints, and fractures. 

4.12.4.7 Specific Gravity 

The specific gravity of solids (Gs) is a measure of solid particle density and is referenced to an 

equivalent volume of water. Specific gravity of solids is defined as Gs = Ms/(Vs × γw) where Ms is the 

mass of the soil solids and Vs is the volume of the soil solids. Since many sands are comprised of quartz 

and/or feldspar minerals and many clays consist of the kaolinite and/or illite clay minerals in 

composition, and since the specific gravity of these minerals are confined to a relatively narrow range, 

the typical values of specific gravity of most soils also lie within the narrow range of Gs = 2.7 ± 0.1. 

Exceptions include soils with appreciable organics (i.e., peat), ores (mine tailings), or calcareous (high 

calcium carbonate content) constituents. It is common to assume a reasonable Gs value, although 

laboratory testing by AASHTO T100 or ASTM D 854 or D 5550 can be used to verify and confirm its 

magnitude, particularly on projects where little previous experience exists and unusually low or high 

unit weights are measured. 

4.12.4.8 Organic Content 

A visual assessment of organic materials may be very misleading in terms of engineering analysis. 

Laboratory test method AASHTO T194 or ASTM D 2974 should be used to evaluate the percentage of 

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organic material in a specimen where the presence of organic material is suspected based on field 

information or from previous experience at a site. The test involves heating a sample to temperatures of 

440°C and holding this temperature until no further change in mass occurs. At this temperature, the 

sample turns to ash. Therefore, the percentage of organic matter is (100% - % ash) where the % ash is 

the ratio of the weight of the ash to the weight of the original dried sample. The sample used for the test 

is a previously dried sample from a moisture content evaluation. Usually organic soils can be 

distinguished from inorganic soils by their characteristic odor and their dark gray to black color. In 

doubtful cases, the liquid limit should be determined for an oven-dried sample (i.e., dry preparation 

method) and for a sample that is not pre-dried before testing (i.e., wet preparation method). If drying 

decreases the value of the liquid limit by about 30 percent or more, the soil may usually be classified as 

organic (Terzaghi, Peck, and Mesri, 1996). 

Soils with relatively high organic content have the ability to retain water, resulting in high moisture 

content, high primary and secondary compressibility, and potentially high corrosion potential. Organic 

soils may or may not be relatively weak depending on the nature of the organic material. Highly organic 

fibrous peats can exhibit high strengths despite having a very high compressibility. 

4.12.4.9 Electro Chemical Classification Tests 

Electro chemical classification tests provide the engineer or geologist with quantitative information 

related to the aggressiveness of the soil conditions and the potential for deterioration of a foundation 

material. Electro chemical tests include (1) pH; (2) resistivity; (3) sulfate ion content; and (4) chloride 

ion content. If the pH of the soil is below 4.5 or the resistivity is less than 1000 ohms/cm, the soil 

should be treated as an aggressive environment. If the soil resistivity is between 3000 ohms/cm and 

5000 ohms/cm, chloride ion content and sulfate ion content tests should be performed. If results from 

these tests indicate chloride ion content greater than 100 ppm or sulfate ion content greater than 200 

ppm, then the soil should be considered as aggressive. Tests to characterize the aggressiveness of a soil 

environment are important for design applications that include metallic elements, especially for ground 

anchors comprised of high strength steel and for metallic reinforcements in mechanically stabilized earth 

walls. 

4.12.5 Laboratory Performance Tests for Soils 

4.12.5.1 General 

Design soil properties for deformation, shear strength, and permeability characteristics are evaluated 

using laboratory-testing methods. To contrast most index tests, performance tests are usually more 

costly and time consuming. The results, however, provide specific data regarding engineering 

performance. This section provides information on equipment and testing procedures for consolidation, 

shear strength, and permeability testing. 

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4.12.5.2 Consolidation 

Equipment 

The oedometer (or one-dimensional consolidometer) is the primary laboratory equipment used to 

evaluate consolidation and settlement potential of cohesive soils. Results from oedometer tests can be 

used to assess the rate of consolidation (t100), creep characteristics (Cα), stress history (σp′), and swell 

potential. A consolidation test is typically performed on undisturbed samples obtained from the deposit 

to evaluate settlement potential of in-situ foundation soils, however recompacted materials can also be 

tested to assess the settlement performance of compacted fills. The equipment for a consolidation test 

includes: (1) a loading device that applies a vertical load to the soil specimen; (2) a metal ring (fixed or 

free) that laterally confines the soil specimen and restricts deformation to the vertical direction only; (3) 

porous plates placed on the top and bottom of the sample; (4) a dial indicator or linear variable 

differential transducer (LVDT); (5) a timer; and (6) a surrounding container to permit the specimen to 

remain submerged during the test. Figure 26 shows the components of a consolidation test. 

Figure 26. Components of consolidation test (photograph courtesy of GeoTesting Express). 

The consolidation-loading device may be a weighted lever arm (figure 27), a pneumatic device, or an 

automated loading frame (figure 28(b)). Automated loading frames are recommended for use in 

production testing because they provide the most flexibility in testing options. The pneumatic device 

provides flexibility in loads and load increment ratios that can be applied during testing. A weighted 

lever arm provides a robust, relatively simple system for consolidation testing, however, because data 

are generally recorded manually, it is difficult to expedite testing or vary the loading schedule since data 

reduction cannot typically be performed in real time. 

Consolidation cells may be either fixed ring or floating ring. Friction and drag are created in the ring as 

the specimen compresses in relation to the ring. In a fixed ring test the sample compresses from the top 

only, potentially resulting in high incremental side shear forces. In a floating ring test the sample 

compresses from the top and bottom thus providing the advantage of minimizing drag forces. However, 

the floating ring method has the disadvantages of being more difficult to set up, allowing for potential 

sidewall leakage that would result in an inaccurate assessment of the coefficient of consolidation, and 

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soil may squeeze out near the junction of the sidewall and the bottom porous plate. For these reasons, 

the fixed ring method is most commonly used. 

Measurement of deflection can be done with either dial indicators or electronic LVDT. Properly 

calibrated, each device should provide the same accuracy, but the electronic output of a LVDT can be 

incorporated into automated recording systems for quicker, more efficient, and higher resolution 

readings. A timer will be required to assess duration of loading increments. Monitoring of time for 

manual systems can be accommodated using a wall clock with a second hand. The internal clock of a 

computer is used for automated systems. 

Figure 27. Incremental load oedometer. 

Procedures 

Consolidation properties of clay soils are evaluated in the laboratory using the one-dimensional 

consolidation test. The most common laboratory method is the incremental load (IL) oedometer (ASTM 

D 2435) shown in figure 27. High-quality undisturbed samples using thin-walled tubes (ASTM D 

1587), piston samplers, or other special samplers are required for laboratory consolidation tests. 

The consolidation test is relatively expensive and time consuming as compared to simpler index type 

tests, but results are necessary to assess the compressibility properties of the soil. As will be shown in 

subsequent sections of this document, this test is one of the most valuable tests for fine grained soil as it 

provides valuable data regarding stress history, as well as compressibility. It is important to carefully 

consider all laboratory testing variables and their potential effects on computed properties. Information 

that will need to be provided to a laboratory for a consolidation test will include the loading schedule 

(i.e., loads and duration of loads). It is important to carefully evaluate the loading schedule to be used. 

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The loading sequence selected for a consolidation test will depend on the type of soil being tested and 

the particular application (e.g., embankment, shallow foundation) being considered for the project. The 

selection of a loading schedule should never be left to the discretion of the laboratory. As an example, if 

the clay soil is heavily overconsolidated, it is possible that a laboratory-determined maximum load for 

the consolidation test will not be sufficient to exceed σp′. 

The range of applied loads for the test should well exceed the effective stresses that are required for 

settlement analyses. This range should cover the smallest and largest effective stresses anticipated in the 

field and will depend on depth, foundation loads, and excavations. The anticipated preconsolidation 

stress should be exceeded by at least a factor of four during the laboratory test. If the preconsolidation 

stress is not significantly exceeded during the loading schedule, σp′, and Cc (or Ccε) may be 

underestimated due to specimen disturbance effects. A load-increment-ratio (LIR) of 1 defined as 

∆σV/σV′ = 1, is commonly used for most tests corresponding to a doubling of the vertical stress applied 

to the specimen during each successive increment. As the stress approaches the value of σp′, smaller 

LIR increments are recommended to facilitate an accurate estimate of σp′. Typically, laboratories 

provide a unit cost for a consolidation test that may be based on 6 to 8 load increments with a separate 

cost for each additional increment. 

It is recommended that an unload-reload cycle be performed, especially for cases where accurate 

settlement predictions are required, specifically to obtain a value for Cr. Since most samples will 

inevitably be somewhat disturbed, a Cr value based on the initial loading of a consolidation test sample 

will be higher than that for an undisturbed sample, resulting in an overestimation of settlements in the 

overconsolidated region. A Cr value based on an unload-reload cycle is likely to be more representative 

of the actual modulus in the overconsolidated region. It is recommended that the unload-reload cycle be 

performed at a stress slightly less than σp′. 

The duration of each load increment should be selected to ensure that the sample is approximately 100 

percent consolidated prior to application of the next load. For relatively low to moderate plasticity silts 

and clays, durations of 3 to 12 hours will be appropriate for loads in the normally consolidated range. 

For fibrous organic materials, primary consolidation may be completed in 15 minutes. For high 

plasticity materials, the duration for each load may need to be 24 hours or more to ensure complete 

primary consolidation and to evaluate secondary compression behavior. Conversely, primary 

consolidation may occur in less than 3 hours for loads less than σp′. If the time period is too short for a 

given load increment (i.e., the sample is not allowed to achieve approximately 100 percent consolidation 

before the next load increment is applied), then values of Cc (or Ccε) may be underestimated and values 

of cv may be overestimated. The duration of time required, however, can be optimized using pneumatic, 

hydraulic, or electro-mechanical loading systems that include automated loading and data acquisition 

systems. Continuous deformation versus time measurements and the square root of time method 

(described in chapter 5) can be used to estimate the beginning and end of primary consolidation during 

the test. Once the end of primary consolidation is detected, the system can automatically apply the next 

load increment. Alternatively, some laboratories can provide real-time deformation versus time plots to 

enable the engineer to evaluate whether 100 percent primary consolidation has been achieved. 

The constant rate of strain (CRS) version of the consolidation test (ASTM D 4186) applies the loading 

continuously and measures stress and pore pressures by transducers in real time, thereby reducing 

testing times from approximately 1 week by IL oedometer to about 1 day by a CRS consolidometer. 

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While expediting the testing time duration, the CRS consolidation test requires special instrumentation 

and equipment that is not normally available in state DOT laboratories. A discussion of the CRS and 

other consolidation methods are described in Head (1986) and Lowe et al. (1969). 

Secondary compression should be assessed on the basis of the deformation versus a log-time response 

(see section 5.4.2.7). The consolidation test for each load increment should be run long enough to 

establish a log-linear trend between time and deformation. 

4.12.5.3 Soil Strength 

General 

Soil shear strength is influenced by many factors including the effective stress state, mineralogy, 

packing arrangement of the soil particles, soil hydraulic conductivity, rate of loading, stress history, 

sensitivity, and other variables. As a result, shear strength of soil is not a unique property. 

Laboratory measured shear strength values will also vary because of boundary conditions, loading rates, 

and direction of loads. In this section, typical laboratory strength tests including unconfined 

compression (AASHTO T208; ASTM D 2166), triaxial (AASHTO T234; ASTM D 4767), and direct 

shear (AASHTO T236; ASTM D 3080) are introduced and in chapter 5, the evaluation of soil strength 

from these laboratory tests is provided. A detailed discussion on testing equipment and procedures is 

beyond the scope of this document. The interested reader should review the AASHTO and ASTM 

standards for detailed information on these tests. This section also describes information that must be 

conveyed to a laboratory-testing firm to ensure that the strength testing performed is consistent with the 

requirements imposed by the design (e.g., selection of confining pressures consistent with the imposed 

loads). 

Unconfined Compression (UC) Tests 

The unconfined compression test is a quick, relatively inexpensive means to obtain an approximate 

estimation of undrained shear strength of cohesive specimens. This test unfortunately is commonly used 

in practice because of its simplicity and low cost. In most cases, undrained strength results from an 

unconfined compression test are conservative. The maximum stress, qu, measured at failure is equal to 

two times the undrained strength (su). In this test a cylindrical specimen of the soil is loaded axially, 

without any lateral confinement to the specimen, at a sufficiently high rate to prevent drainage. Since 

there is no confinement, residual negative pore pressures that may exist in the sample following sample 

preparation control the state of effective stress. This test cannot be performed on granular soils, dry or 

crumbly soils, silts, peat, or fissured or varved materials. Because there is no control on the effective 

stress state of the specimen, this test is not recommended for evaluating strength properties for 

compressible clay soils subjected to embankment or structural foundation loads. The reliability of this 

test decreases with respect to increasing sampling depth because the sample tends to swell after 

sampling resulting in greater particle separation and reduced shear strength. Testing the full diameter 

extruded specimen as soon as possible after removal from the tube can minimize swelling. This reduces 

disturbance and preserves natural moisture content. 

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Triaxial Tests 

Equipment – Triaxial systems today use electronic instrumentation to provide continuous monitoring of 

test data (see figure 28). Force is measured using a force transducer or load cell that is typically 

mounted outside the triaxial cell, but more advanced systems have incorporated the transducer within 

the testing cell to reduce rod-friction effects. LVDTs are used to monitor deformations. Additionally, 

volume measurements can be taken with a device that makes use of an LVDT to measure the rise or fall 

of a bellofram cylinder. This change in movement is calibrated to the volume of water the sample will 

take in or push out. Pressure transducers are mounted on the base of the test cell to monitor the 

confining pressure and the pore pressure within the sample. 

Figure 28. (a) Failure of a loose sand specimen in a triaxial cell; and 

(b) Load frame, pressure panel, and computerized data acquisition system 

Unconsolidated-Undrained (UU) Test – In this test, no drainage or consolidation is allowed during either 

the application of the confining stress or the shear stress. This test models the response of a soil that has 

been subject to a rapid application of confining pressure and shearing load. It is difficult to obtain 

repeatable results for UU testing due to sample disturbance effects. Like the UC tests, the accuracy of 

the UU test is also dependent on the soil sample retaining its original structure until testing occurs. The 

undrained strength of the soil, su, is measured in this test. 

Consolidated-Drained (CD) Test – In this test, the specimen is allowed to completely consolidate under 

the confining pressure prior to performing the shearing portion of the test. During shearing, load is 

applied at a rate slow enough to allow drainage of pore water and no buildup of pore water pressures. 

The time required to conduct this test in low permeability soil may be as long as several months; 

therefore it is not common to conduct this test on low permeability soils. This test models the long-term 

(drained) condition in soil. Effective stress strength parameters (i.e., φ′ and c′) are evaluated in this test. 

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Consolidated-Undrained (CU) Test – The initial part of this test is similar to the CD test in that the 

specimen is allowed to consolidate under the confining pressure. Shearing occurs, however, with the 

drainage lines closed, thus during shearing there is continual pore water pressure development. The rate 

of shearing for this test is more rapid than that for a CD test. Pore pressures should be measured during 

shearing so that both total stress and effective stress strength parameters can be obtained. It is noted that 

the effective stress parameters evaluated for most soils based on CU testing with pore pressure 

measurements is similar to that obtained for CD testing, thus making CD tests not necessary for typical 

applications. 

Direct Shear Tests 

Equipment – The apparatus and procedures for direct shear testing are discussed in ASTM D 3080. A 

specimen is prepared in a split square or circular box (see figure 29), and is sheared as one box is 

displaced horizontally with respect to the other using upper and lower loading frames (see figure 30). 

Load cells are used to monitor the shear force and LVDTs are used to monitor both horizontal and 

vertical deformation. Using this instrumentation, as well as a loading frame that provides a constant rate 

of horizontal deformation, it is possible to automate the direct shear test. 

Figure 29. Direct shear testing box (photograph courtesy of GeoTesting Express). 

Test Procedures – In the direct shear test, the soil is first consolidated under an applied normal stress. 

The soil is then sheared at a constant rate after consolidation is completed (which will be instantaneous 

in cohesionless soils), which should be selected as a function of the hydraulic conductivity of the 

specimen. Direct shear testing is commonly performed on compacted materials used for embankment 

fills and retaining structures. This testing can also be performed on natural materials; however, the lack 

of control on soil specimen drainage makes the evaluation of undrained strength unreliable. This test 

can be used to evaluate the drained strength of natural materials by shearing the sample at a slow enough 

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rate to reasonably ensure that no porewater pressures develop. In addition to peak effective stress 

friction angle (φ′), the direct shear test can be used for the evaluation of effective stress residual 

strengths (c′r ≈ 0; φ′r). These parameters are necessary for stability and landslide analyses. Residual 

strengths are associated with very large deformations and shear strains along a predefined or preferential 

slip surface. A reversing direct shear test can be used to evaluate residual shear strengths. In this test, 

the direction of shearing in the test is reversed several times thereby causing the accumulation of 

displacements at the slip surface. 

For designs involving geosynthetics, the strength of the interface between the soil and geosynthetic or 

geosynthetic and geosynthetic are often necessary parameters. Direct shear machines have been 

modified to test the shear strength of various interfaces, as described in ASTM D 5321. 

Figure 30. Soil sample mounted in direct shear testing apparatus (photograph courtesy of GeoTesting 

Express). 

Factors Affecting Strength Testing Results 

In chapter 2, a flowchart describing the process of evaluating design soil and rock properties was 

presented. That flowchart included a step wherein information collected from subsurface investigations 

and baseline laboratory test data (i.e., index test results) are summarized to facilitate development of a 

subsurface profile and as a tool to assist in developing a program for required laboratory performance 

tests such as strength and consolidation testing on specimens from tube samples. This section provides a 

discussion of specific factors relevant to laboratory unconfined compression, triaxial, and direct shear 

testing and how these factors affect measured shear strength values. In chapter 5, specific 

recommendations are developed with respect to selecting a particular test for a particular application 

where multiple tests may be appropriate. 

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Sample Disturbance

The degree of disturbance affecting samples will vary according to the type of soil, sampling method, 

and skill of the driller. All samples will experience some degree of disturbance due to the removal of insitu stresses during sampling and laboratory preparation for testing. Due to disturbance-induced 

alteration of the in-situ soil structure, internal migration of pore water, and reduction in the effective 

stress state of the sample, shear strength values obtained from UC and UU tests will be unrepeatable and 

may be higher or lower than corresponding field strengths. Recompression of a sample during the 

consolidation phase of a CIU test will reduce the void ratio of the specimen that may lead to higher 

laboratory strengths relative to the in-situ condition, but destruction of natural bonding during sampling 

will typically more than offset this strength increase. Table 16 should be used to judge the degree of 

disturbance of the samples and strength results from samples designated as very good or acceptable can 

be considered to be appropriate. Shear strengths from samples designated as likely to be very disturbed 

should be used with caution for design calculations. 

Mode of Shearing

Experience has shown the undrained soil shear strength also depends on the direction of shearing. That 

is, a soil loaded in compression will likely have a shear strength that is different than if the soil is loaded 

in extension. The effects are not as recognized for drained (effective stress) strength in compression and 

extension, or partially drained conditions. Most triaxial tests will be performed with isotropic 

consolidation and vertical compression, as most commercial laboratories are not equipped to perform 

various modes of shearing. The engineer must consider how the actual strength mobilized under field 

conditions differs from that measured using laboratory (or in-situ) methods. For example, when 

considering design of a temporary anchored wall in soft to medium clay, the undrained strength used to 

evaluate the earth pressures acting on the wall may be determined from a triaxial compression test. The 

lateral capacity of the wall toe, however, is more appropriately evaluated using the undrained strength 

from a triaxial extension test. The extension loading path more accurately approximates the unloading 

caused by a soil excavation as compared to a compression loading path and, more importantly, 

experience has shown that the shear strength of the soil located inside the excavation can be less than 

that of the soil located in the retained ground for certain clay soils. 

It is recognized that for most typical projects the use of alternative loading paths is not practical. 

However, information and existing correlations relating undrained strength from isotropically 

consolidated (CU) triaxial tests to other loading paths can be used to adjust the CU strength to a value 

more appropriate for the loading condition imposed by the structure to be built. This is discussed further 

in chapter 5. 

Confining Pressures

Soil shear strength is governed by the effective stresses in the soil. Therefore, it is important to carefully 

consider the range of effective stresses that a soil will be subjected to during the design life of a 

structure. These stresses will be affected by changes in the level of the ground water table, effects of 

capillary rise, design loads of potential structures, as well as many other possibilities. For laboratory 

testing considerations, this means that for each sample tested, the in-situ (or current) effective stress 

condition and that which will exist after the design feature (e.g., shallow foundation, embankment, 

retaining wall, cut slope) has been constructed needs to be calculated. 

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For laboratory strength testing, three different confining stresses are generally used for each sample at a 

unique depth, thus requiring three specimens from the same undisturbed sample. For each specimen, the 

shear strength is measured and a shear strength envelope is developed. Details regarding the evaluation 

of shear strength are provided in chapter 5. In some cases, it may be difficult to prepare three specimens 

from one tube due to defects, fissures, poor recovery, etc. Testing can be done using only one or two 

specimens if the number of specimens is limited, but the engineer should investigate the reasons for the 

inability to prepare three specimens and should exercise judgment in interpreting the test results. 

Loading ranges typically include the effective overburden stress at the sample depth, one half the 

effective overburden stress at the sample depth, and a third stress condition superceding the anticipated 

design load or two to four times the effective overburden stress at the depth sampled, whichever is 

greater. To calculate the range of effective stresses, the final effective stress at the elevation of the 

sample should be plotted as a function of depth. For surface loadings such as that due to embankments 

and shallow foundations, stress distributions with depth should be calculated using appropriate methods. 

For UU tests, the soil specimen is not re-consolidated to the effective stress in the ground. In selecting 

confining pressures, the total stress at the elevation of the soil sample in the ground is reapplied to the 

specimen with the test apparatus drain lines closed. If it is assumed that the water content of the 

specimen just prior to testing is the same as that in the ground and if the sample is saturated, the 

reapplication of total stresses equal in magnitude to those which were in the ground, should theoretically 

restore the sample to its in-situ effective stress condition. Because of inevitable sample disturbance, 

more pressure is transferred to the porewater resulting in a lower effective stress as compared to that in 

the ground. The lower preshear effective stress results in lower than actual shear strength. UU test 

results are considered unreliable at depths greater than 6 m for normally consolidated samples and over 

12 m for overconsolidated soils because of this reduction in effective stress. 

Specimen Size

The specimen size for testing must be provided to the laboratory. Triaxial testing specimens are 

cylindrical with a minimum diameter of 33 mm, and a length to diameter (L:D) ratio between 2 and 2.5. 

Undisturbed samples from tubes, which are typically 76 mm in diameter, need to be trimmed to fit the 

caps and bases of the triaxial device. A specimen trimmed with care to 35.6 mm diameter is generally 

the best practice for triaxial test specimen preparation for CU or CD testing to minimize the disturbance 

related to the side walls of the samples. If UU tests are performed, specimens should be extruded 

directly from the sampling tube and tested untrimmed at full diameter to minimize disturbance effects. 

Laboratory testing for undrained strength of heavily overconsolidated, fissured soils is difficult since 

typical sample diameters may not be sufficient to capture the affects of fissures and cracks on strength. 

In many cases, the actual strength of the soil can be up to 50 percent less than that measured in the 

laboratory. Additional discussion on the strength testing of stiff, fissured soil materials are provided in 

chapter 5. 

Saturation

Backpressure saturation procedures are typically used to saturate soil samples for triaxial testing. A 

backpressure of at least 1 atm (100 kPa) should be applied, but 2 to 3 atm (200 to 300 kPa) of 

backpressure are recommended. Samples need to be saturated for drained tests to permit volume change 

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measurements to be made and, for undrained tests, to permit pore pressures during shearing to be 

measured. Saturation by backpressure methods involves raising the pressure inside the specimen to 

dissolve gas into the pore fluid. Since the cell pressure is raised an equal value along with the internal 

specimen pressure, the effective stress of the sample remains constant. The pore pressure parameter, 

B=∆u/∆σ3, should be equal to at least 0.95 for the specimen to be considered saturated. If the B-value 

remains constant as the back pressure is increased, the specimen can be considered essentially saturated. 

Displacement at Failure

The engineer should estimate the amount of deformation or strain necessary to achieve the ultimate 

strength of the material in a laboratory strength test. The purpose of this is to ensure that the full stressstrain curve of the sample is recorded during the test. For example, large-displacement (or residual) 

shear strength values may be required to perform stability analyses for a preexisting slip plane, such as 

the case for a landslide. The engineer should provide the laboratory with a minimum strain (or 

displacement) value to ensure that the laboratory does not prematurely stop a test. In most cases, UC 

and triaxial tests run to 15 percent axial strain will be sufficient. For truly normally consolidated 

samples tested in compression, strains on the order of 20 to 25 percent may be required to reach the peak 

soil strength. 

Rate of Shearing

The rate of shearing needs to be carefully considered before beginning either a triaxial or direct shear 

test, especially for fine grained soils. The selection of shearing rates for undrained shearing tests on 

clays needs to be slow enough to ensure equalization of pore pressures within the sample. For drained 

tests on clays, the shearing rate must be slow enough to allow for excess pore pressures to dissipate 

through the pervious boundaries. For both CU and CD tests, the time to failure, tf, is estimated using 

table 18. This table also shows the affect of using side drains. Typical triaxial tests incorporate porous 

stones on the top and bottom of the specimen. The use of filter strips along the side of the specimen 

(i.e., side drains) serves to reduce the time required to dissipate excess pore pressures in the specimen by 

allowing drainage in the radial direction. The t100 value in table 18 is the time to complete primary 

consolidation, which can be evaluated using time rate versus deformation data from the consolidation 

portion of the strength test. Since most laboratories do not record time rate consolidation data during the 

consolidation phase of a CIU test, this data will need to be requested from the testing laboratory prior to 

testing. Next, the axial strain to reach peak strength, εp, is estimated. Strains required to reach peak 

conditions depend on the type of clay, OCR of the clay, and the imposed loading during shear (i.e., 

compression, extension, etc.). Typical values for compression loading of an isotropically consolidated 

specimen are 20 to 25 percent at OCR = 1 and decreasing to a few percent at high OCR (>20). A 

maximum rate of displacement, •

δ , can then be calculated so that εp is reached after tf for a specimen 

with an initial height Ho as: 

•

δ =εpHo/(12.7 t100) (Equation 23) 

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Table 18. Time tf to reach failure (after Head, 1986). 

Type of test No side drains With side drains 

CU 0.51 t100 1.8 t100

CD 8.5 t100 14 t100

Although not commonly performed on fine-grained soils, in the case of consolidated drained (CD) direct 

shear tests, the shearing rate can be selected based on ASTM 3080 wherein the minimum time required 

to fail a sample, tf, is calculated as: 

 tf = 50xt50 = 11.7xt90 (Equation 24) 

where t50 and t90 are the times required to complete 50 percent and 90 percent primary compression, 

respectively. The times t50 and t90 may be evaluated using the square root of time or logarithm of time 

method (see chapter 5) to assess the vertical displacements measured with time under the constant 

normal load prior to shearing. Once tf is calculated, the displacement required to achieve the peak 

strength of the soil (using a conventional size direct shear box), δf, can be estimated as 1 to 2 mm for 

hard clays, 2 to 5 mm for stiff clays, and 8 to 10 mm for plastic clay (Bardet, 1997). The maximum 

shear rate for the CD test is then selected so that the test takes at least as long as tf to reach the 

displacement δf. 

4.12.5.4 Permeability 

Equipment 

Laboratory permeability testing is performed to determine the hydraulic conductivity of a soil specimen. 

For natural soils, tests are conducted on specimens from tube samples and for fill and borrow soils, tests 

are made on recompacted materials. Two types of tests are commonly performed, the rigid wall test 

(AASHTO T215; ASTM D 2434) and the flexible wall test (ASTM D 5084). Rigid wall permeameters 

are not recommended for low permeability (i.e., k ≤ 10-6 cm/s) soils due to the potential for sidewall 

leakage. The equipment for the rigid wall test includes a rigid wall permeameter, water tank, vacuum 

pump, and manometer tubes (see figure 31). The permeameter must be large enough to minimize 

sidewall leakage, and thus the diameter of the rigid wall device should be at least 8 to 12 times that of 

the largest soil particle. Porous stones and filter paper must not restrict flow through the soil, but must 

also retain the fine particles. A vacuum pump is used to remove air from the samples and to saturate the 

specimens prior to testing. Manometer outlets should be available on the sides of the cell to measure 

head loss over the specimen. 

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Figure 31. Rigid wall permeameter (photograph courtesy of GeoTesting Express). 

Constant head and falling head tests can be performed using a flexible wall permeameter cell, cell 

reservoir, headwater reservoir, tailwater reservoir, top and base caps, flexible membrane, porous stones, 

and filter paper. With this equipment, the specimen can be tested at a range of confining stresses under 

backpressure saturation. The separate headwater and tailwater reservoirs can be monitored, and falling 

head or constant head tests can be performed. Since the flexible membrane encases the specimen, side 

leakage is prevented. Flexible wall permeameter cells (see figure 32) consist of influent and effluent 

lines as well as pore stones and filter paper. The hydraulic conductivity of the system may be found to 

be lower than that of the material due to small diameter flow lines and low permeability porous stones. 

This is not acceptable. The hydraulic conductivity of the system should be tested to ensure it is at least 

one order of magnitude greater than that anticipated for the material. 

Procedures 

Laboratory determinations of hydraulic conductivity (k) for materials with a hydraulic conductivity less 

than or equal to 1x10-3 cm/s are usually performed in flexible wall permeameters (ASTM D 5084). 

Rigid wall permeameters are typically used for sandy and gradually soils (ASTM D 2434) with a 

hydraulic conductivity greater than 1x10-3 cm/s. The flexible membrane used to encase the specimen 

prevents sidewall leakage for fine-grained soils that would occur in a rigid wall system. Rigid wall 

systems are used for granular materials because the permeability of the flexible wall system (valves, 

pore stones, tubing, etc.) may be less than that of the specimen, and thus the test would reflect a 

permeability lower than that of the specimen. 

The confining stress of the hydraulic conductivity test will need to be specified to the laboratory. As 

confining stress increases, the hydraulic conductivity of fine grained soils will typically decrease due to 

consolidation of the specimen and reduction of void ratio. The confining stress should be equal to the 

anticipated effective stress-state in the soil. 

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Figure 32. Flexible wall permeameter (photograph courtesy of GeoTesting Express). 

The hydraulic gradient, defined as the difference in hydraulic head across the specimen divided by the 

length of the specimen, should also be specified to the laboratory. Typical hydraulic gradients in field 

situations are less than 5, however the use of such a small gradient will result in extremely long testing 

times for materials having hydraulic conductivities less than 1x10-6 cm/s. If the hydraulic gradient 

across the specimen is too high, turbulent flow will occur and result in hydraulic conductivity values that 

are too low. Suggested hydraulic gradient values are a function of the anticipated hydraulic 

conductivity, as presented in table 19. 

Table 19. Recommended maximum hydraulic gradient for permeability testing. 

Hydraulic Conductivity, 

cm/sec 

Recommended Maximum 

Hydraulic Gradient 

1 x 10-3 to 1 x 10-4 2 

1 x 10-4 to 1 x 10-5 5 

1 x 10-5 to 1 x 10-6 10 

1 x 10-6 to 1 x 10-7 20 

less than 1 x 10-7 30 

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Saturation of the specimen is necessary to achieve accurate results. Backpressure saturation procedures 

should be used to achieve B-values greater than 0.95, or constant B-values over a range of increasing 

back pressure values. In a rigid wall test, air is evacuated from the sample using a pump and then 

saturation from the bottom of the specimen upward is performed (see ASTM D 2434). 

A hydraulic conductivity test should be ended when steady flow is occurring. The hydraulic 

conductivity is considered to be steady when four or more consecutive hydraulic conductivity 

measurements fall within ±25 percent of the average k value if k is greater than 1 x 10-10 m/s, or if four 

or more measurements fall within ±50 percent of the average if k is less than 1 x 10-10 m/s. 

Since the hydraulic conductivity is defined at a temperature, T (in degrees Celsius), of 20 o

C, a 

temperature correction factor, RT, may need to be incorporated into the calculations. To apply this 

temperature correction factor, the temperature of the permeant will need to be provided by the 

laboratory. The corrected hydraulic conductivity value, i.e., k20 is given as kmeasured × RT. Temperature 

correction factors are presented in ASTM D 5084 and can also be calculated according to equation 25. 

 RT = (-0.02452 T +1.495) (Equation 25) 

4.13 LABORATORY ROCK TESTS 

4.13.1 Introduction 

This section provides information on common laboratory test methods for rock including testing 

equipment, general procedures related to each test, and parameters measured by the tests. A discussion 

of the interpretation of these tests to obtain properties is provided in chapter 6. Table 20 provides a list 

of commonly performed laboratory tests for rock associated with typical projects for highway 

applications. Although other laboratory test methods for rock are available including triaxial strength 

testing, rock tensile strength testing, and durability testing related to rock soundness, most design 

procedures for structural foundations and slopes on or in rock are developed based on empirical rules 

related to RQD, degree of fracturing, and to the unconfined compressive strength of the rock. The use 

of more sophisticated rock laboratory testing is usually limited to the most critical projects. Details on 

other laboratory rock testing procedures are provided in FHWA- HI-97-021 (1997). Table 21 provides 

summary information on typical rock index and performance tests. 

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Table 20. Common rock laboratory tests. 

Test 

Category Name of Test ASTM Test Designation 

Point Load Strength Suggested method for evaluating point load 

strength D 5731 

Compressive Strength Compressive strength of intact rock core 

specimen (in unconfined compression) D 2938 

Direct Shear Strength Laboratory direct shear strength tests for rock 

specimens under constant normal stress D 5607 

Durability Slake durability of shales and similar weak 

rocks D 4644 

Strength-Deformation Elastic moduli of intact rock core specimens in 

uniaxial compression D 3148 

Table 21. Summary information on rock laboratory test methods. 

Test Procedure Applicable 

Rock Types 

Applicable 

Rock 

Properties 

Limitations / Remarks 

Point-Load 

Strength Test 

Rock specimens in the form of core, 

cut blocks, or irregular lumps are 

broken by application of concentrated 

load through a pair of spherically 

truncated, conical platens. 

Generally not 

appropriate for 

rock with uniaxial 

compressive 

strength less than 

25 MPa 

Provides an 

index of 

uniaxial 

compressive 

strength 

Can be performed in the field 

with portable equipment or in the 

laboratory; in soft or weak rock, 

test results need to be adjusted to 

account for platen indentation 

Unconfined 

Compressive 

Strength of 

Intact Rock 

Core 

A cylindrical rock specimen is placed 

in a loading apparatus and sheared 

under axial compression with no 

confinement until peak load and failure 

are obtained. 

Intact rock core Uniaxial 

compressive 

strength 

Simplest and fastest test to 

evaluate rock strength; fissures or 

other anomalies will often cause 

premature failure 

Laboratory 

Direct Shear 

Test 

A rock specimen is placed in the lower 

half of the shear box and encapsulated 

in either synthetic resin or mortar. The 

specimen must be positioned so that 

the line of shear force lies in the plane 

of the discontinuity to be investigated. 

The specimen is then mounted in the 

upper shear box and the normal load 

and shear force are applied. 

Used to assess 

peak and residual 

shear strength of 

discontinuity 

Peak and 

residual 

shear 

strength 

May need to perform in-situ 

direct shear test if design is 

controlled by potential slip along 

a discontinuity filled with very 

weak material 

Elastic Moduli 

of Intact Rock 

Core 

Procedure is similar to that for 

unconfined compressive strength of 

intact rock. Lateral strains are also 

measured 

Intact rock core Modulus and 

Poisson’s 

ratio 

Modulus values (and Poisson’s 

ratio) vary due to nonlinearity of 

stress-strain curve. 

Slake 

Durability(1)

Dried fragments of rock are placed in a 

drum made of wire mesh that is 

partially submerged in distilled water. 

The drum is rotated, the sample dried, 

and the sample is weighed. After two 

cycles of rotating and drying, the 

weight loss and the shape of size of the 

remaining rock fragments are recorded. 

Shale or other 

soft or weak 

rocks 

Index of 

degradation 

potential of 

rock 

Note: (1) This test is described further in chapter 7. 

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4.13.2 Laboratory Testing of Rock 

4.13.2.1 Point-Load Strength Test 

The point load strength test is an appropriate method used to estimate the unconfined compressive 

strength of rock in which both core samples and fractured rock samples can be tested. The test is 

conducted by compressing a piece of the rock between two points on cone-shaped platens (see figure 33) 

until the rock specimen breaks in tension between these two points. Each of the cone points has a 5-mm 

radius of curvature and the cone bodies themselves include a 60° apex angle. The equipment is portable, 

and tests can be carried out quickly and inexpensively in the field. Because the point load test provides 

an index value for the strength, usual practice is to calibrate the results with a limited number of uniaxial 

compressive tests on prepared core samples. The point load test is also used with other index values to 

assess degradation potential of shales (see chapter 6). 

Figure 33. Point load strength test equipment. 

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If the distance between the platens is D and the breaking load is P, then the point load strength, Is is 

calculated as: 

2

e

s

D

P I = (Equation 26) 

where De is the equivalent core diameter which is given by: (1) De = D2

 for diametral tests; or (2) De = 

4×A for axial, block, or lump tests where A = W×D. The area A is the minimum cross-sectional area of 

a lump sample for a plane through the platen contact points where W is the specimen width. 

The size-corrected point load strength index, Is(50) of a rock specimen is defined as the value of Is that 

would have been measured by a diametral test with D = 50 mm. For tests performed on specimens other 

than 50 mm in diameter, the results can be standardized to the size-corrected point load strength index 

according to: 

s sPLT = IkI )50( (Equation 27) 

The value of the size correction factor, kPLT, is given by: 

)()50/( 45.0

PLT = Dk mminD (Equation 28) 

It has been found, on average, that the uniaxial compressive strength, σc, is about 20 to 25 times the 

point load strength index, with a value of 24 commonly used, i.e., 

c 24 s )50( σ = I (Equation 29) 

However, tests on many different types of rock show that the ratio can vary between 15 and 50, 

especially for anisotropic rocks. Consequently, the most reliable results are obtained if a series of 

uniaxial calibration tests are carried out. Point load test results are not acceptable if the failure plane lies 

partially along a pre-existing fracture in the rock, or is not coincident with the line between the platens. 

For tests in weak rock where the platens indent the rock, the test results should be adjusted by measuring 

the amount of indentation and correcting the distance D (Wyllie, 1999). 

4.13.2.2 Unconfined Compressive Strength of Intact Rock Core 

The unconfined compressive strength of intact rock core can be evaluated reasonably accurately using 

ASTM D 2938. In this test, rock specimens or regular geometry, generally rock cores, are used. The 

rock core specimen is cut to length so that the length to diameter ratio is 2.5 to 3.0 and the ends of the 

specimen are machined flat. The ASTM test standard provides tolerance requirements related to the 

flatness of the ends of the specimen, the perpendicularity of the ends of the specimens, and the 

smoothness of the length of the specimen. The specimen is placed in a loading frame. Axial load is 

then continuously applied to the specimen until peak load and failure are obtained. The unconfined (or 

uniaxial) compressive strength of the specimen is calculated by dividing the maximum load carried by 

the specimen during the test by the initial cross-sectional area of the specimen. 

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This test is more expensive than the point load strength test, but is also more accurate. Careful 

consideration of the design requirements should be made in deciding whether to perform this test or the 

simpler point load strength test. 

4.13.2.3 Elastic Moduli of Intact Rock Core 

This test is performed similarly to the unconfined compressive test discussed above, except that 

deformation is monitored as a function of load. This test is performed when it is necessary to estimate 

both elastic modulus and Poisson’s ratio of intact rock core. Because of this, it is common to measure 

both axial (or vertical) and lateral (or diametral) strain during compression. It is preferable to use strain 

gauges glued directly to the rock surface as compared to LVDT mounted on the platens since slight 

imperfections at the contact between the platens and the rock may lead to movements that are not related 

to strain in the rock (Wyllie, 1999). 

4.13.2.4 Laboratory Direct Shear Test 

The apparatus and procedures for direct shear testing are discussed in ASTM D 5607. This test is 

typically used to evaluate the shear strength of a rock discontinuity. Overall, the equipment for the 

direct shear test on rock is similar to that for soil including a direct shear testing machine, a device for 

applying normal pressure, and displacement monitoring devices. A schematic for the test set up is 

shown in figure 34. For testing of rock specimens, an encapsulating material such as a high strength 

gypsum cement is poured around the specimen in the upper and lower holding ring. The specimen is 

sheared as one holding ring is displaced horizontally with respect to the other such that the discontinuity 

surface is exactly parallel to the direction of the shear load. Load cells are used to monitor the shear 

force and LVDTs or dial gauges are used to monitor both horizontal and vertical deformation. Multiple 

LVDTs should be used to monitor vertical deformation and potential overturning of the specimen. 

In this test, plots of shear stress versus shear displacement and normal displacement and shear 

displacement are prepared. Normal stresses should be adjusted to account for potential decreases in the 

shear contact area. After the sample is sheared, the sample is then reset to its original position, the 

normal load is increased, and another test is performed. Each test will produce a pair of shear stress and 

normal stress values for both peak and residual conditions. From this, the friction angle of the 

discontinuity surface can be evaluated. Evaluation of the friction angle is described in chapter 6. 

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Figure 34. Laboratory direct shear testing equipment for rock (after Wyllie, 1999, Foundations on 

Rock, Figure 3.19, p. 74, E&FN Spon). 

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CHAPTER 5 

INTERPRETATION OF SOIL PROPERTIES 

5.1 INTRODUCTION 

The interpretation of soil properties based on measured data and/or derived parameters from 

laboratory and in-situ tests is presented in this chapter. Numerous textbooks on soil mechanics and 

foundation engineering describe the basic methods used to interpret soil properties from laboratory 

tests such as one-dimensional consolidation and triaxial shear tests (e.g., Lambe and Whitman, 1969; 

Bardet, 1997). In this chapter, guidance on soil property interpretation is presented with a specific 

focus on practical issues, including the assessment of data quality and the selection of properties for 

design where multiple methods are available to obtain a property. The soil properties presented in 

this section include: (1) subsurface stratigraphy; (2) in-situ stress state; (3) deformation (i.e., 

consolidation, elastic deformation properties, and secondary compression); (4) shear strength; and 

(5) hydraulic conductivity. 

In this section, the selection of properties for design calculations concerning, for example, 

settlements, time rate of settlement, and shear strength is presented. In selecting design soil 

properties, it is typical to use terminology such as “average”, “conservative”, “lower-bound”, or 

“worst-case” in the context of describing a single design property or set of design properties. The 

selection of properties must consider the design application, potential construction issues related to 

the implementation of the design, the design life of the structure, and the criticality of the structure. 

5.2 INTERPRETATION OF SUBSURFACE STRATIGRAPHY 

5.2.1 General 

Methods used to provide information on subsurface stratigraphy and soil type include the SPT with 

recovery of disturbed samples, electric cone penetration test (CPT), piezocone penetration test 

(CPTu), and flat plate dilatometer test (DMT). In this section, the evaluation of soil stratigraphy 

based on data collected from each of these tests is described. This section also provides information 

on developing a subsurface profile for use in geotechnical design analyses. 

5.2.2 Soil Classification by Soil Sampling and Drilling

Routine sampling practice involves the recovery of auger cuttings, drive samples, and thin-walled 

tubes pushed into the ground. First, a vertical boring is advanced using rotary drilling methods that 

may involve solid flight augers (limited to depths < 10 m), hollow-stem augers (limited to depths < 

30 m), and wash-boring techniques. At select depths, a hollow thick-walled tube is lowered to the 

bottom of the borehole and driven into the ground to recover a disturbed soil sample. The driving 

record (i.e., number of blows to drive the spoon 300 mm) is also the basis for the SPT resistance, or 

N value. The 450-mm long drive samples are usually collected at discrete 1.5-m intervals, although 

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smaller and larger testing intervals have been used. Figure 35 depicts the general process of 

traditional drilling, sampling, and laboratory testing of collected samples. 

CaCO3

Layer 3

Layer 2

Layer 4

Layer 1

Bedrock

Collected Samples are:

• marked and sealed;

• stored upright; and

• transported to the lab.

Undisturbed

Samples

Consists of

hydraulically pushing

a thin walled tube to

recovery relatively

undisturbed soil

samples for

laboratory testing of

mechanical

properties.

Disturbed

Samples

Drive samples are

taken using a

drop hammer

system and a

split-barrel

sampler (steel

hollow tube).

1.5 m

Disturbed

Samples

Sieve

Analysis

Atterberg

Limits

 Moisture Content

 Organic Content

Undisturbed

Samples

Classification and Index

Properties

2 - 3 days

Mechanical Properties

2 - 3 Weeks

Hydraulically

extrude and

trim soil

specimen

Consolidation

Compression

Direct Shear

Strength

Triaxial

Shear Strength

and Modulus

Casagrande Fall Cone

Cup Carbonate

Content

HCl Acid Plastic Limit

Diamond Bit

Rock Coring

Figure 35. Traditional drilling, sampling, and laboratory testing of collected samples. 

More recently, sampling by a combination of direct-push and percussive forces has become available 

(e.g., Geoprobe sampling), whereby 25-mm diameter continuously-lined plastic tubes of soil are 

recovered. Although disturbed samples of soil are recovered, the full stratigraphic profile can be 

examined for soil type, layers, seams, lenses, and color changes. 

Standard penetration testing coupled with disturbed sample description, identification, and 

classification is the most commonly used method to obtain stratigraphic profiling information for 

most traditional projects. Although SPT samples are disturbed, the samples are usually of sufficient 

quality to permit visual identification and laboratory classification. These classifications are used 

with recorded SPT N values to develop a subsurface profile and to facilitate “first-order” estimates 

of engineering properties of the soil. Tables 22 and 23 provide baseline correlations between N 

values and relative density (granular soils) and consistency (cohesive soils). When establishing a 

stratigraphic profile using a SPT boring, significant changes in N values should be used to identify 

changes in material types. 

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Table 22. SPT N value soil property correlations for granular soils (after AASHTO, 1988). 

N value Relative Density 

0-4 Very loose 

5-10 Loose 

11-24 Medium Dense 

25-50 Dense 

> 50 Very Dense 

Table 23. SPT N value soil property correlations for cohesive soils (after AASHTO, 1988). 

N value Consistency 

0-1 Very soft 

2-4 Soft 

5-8 Medium Stiff 

9-15 Stiff 

16-30 Very Stiff 

31-60 Hard 

>60 Very Hard 

As discussed in chapter 4, N values are highly dependent on the energy efficiency of the SPT 

method. Inconsistency in the N values across a site may be due to variations in energy efficiency 

between different drill rigs and crews; therefore, N values should be corrected to (N)60, as described 

in section 4.4.3. Failure to correct these N values could result in confusion when attempting to 

correlate subsurface units between borings. Figure 36 shows an example of actual variability of SPT 

N value measurements due to varying energy efficiencies for three different drill rigs and crews. For 

the data included in this figure, each crew used hollow stem auger techniques and a safety hammer 

with a cathead and rope system. The borings were performed in Piedmont residual soil consisting of 

silty fine sands. From this figure it is noted that variability in N-values by 50 percent or more in 

similar soils is not uncommon. Similar variations have been reported with different hammers. 

During drilling operations, the field engineer or geologist should work closely with the driller. An 

experienced driller can distinguish changes in material type and can often identify the material based 

on resistance. The driller should be instructed at the beginning of the job and on a daily basis to alert 

the engineer or geologist of important observations such as, for example, loss of drilling fluid, odors, 

or auger refusal. Such observations should be noted on the field log since this information will be 

useful in developing the site stratigraphy for design analyses. For example, the loss of drilling fluid 

may occur in fractured rock, highly permeable units, or large void spaces; organic odors from the 

borehole may indicate an organic unit such as peat; chemical odors may indicate the presence of 

contaminated soil; and refusal may indicate the depth to bedrock or the presence of an obstruction 

such as a buried object or boulder. 

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Figure 36. Variability of SPT N values. 

The engineer or geologist should record for each boring the initial depth to water, the groundwater 

elevation at the end of drilling, and the groundwater elevation 24 hours after drilling at each location. 

If possible, at the end of the investigation, the groundwater levels should also be recorded in each 

boring. Groundwater level information is most valuable if the levels in all of the borings can be 

measured in one day, in the shortest amount of time possible since groundwater levels may fluctuate 

daily. Cave-in depths should also be recorded as these may indicate groundwater activity, even 

though no water may have been observed. For more reliable groundwater levels, all borings should 

be protected from surface water infiltration at the end of drilling. 

Sampling intervals for SPTs are typically performed at 1.5 m intervals, although continuous 

sampling may be used for the upper reaches (up to 6 m) of a soil deposit. Where very detailed 

stratigraphy is required, it is more cost-effective to supplement SPTs with other in-situ devices such 

as the CPT or CPTu. These devices can be used to evaluate very thin soil layers. The ability to 

accurately characterize the thickness of potentially problematic soil layers cannot be overemphasized 

as costs associated with improving such layers are directly related to the thickness of the layer. 

Additionally, the materials in these layers may be shown to effectively control the entire design. As 

an example, figure 37 shows a summary of SPT (N1)60 values for two borings advanced at site near 

Charleston, South Carolina. Based on SPT (N1)60 values alone, it may be concluded that the 

thickness of a potentially liquefiable layer (i.e., as evidenced by SPT N values less than say 10 for 

this site) may be on the order of 5 to 13 ft (1.5 to 4 m). Two cone penetration tests were performed 

near the two borings and the presence of a potentially liquefiable layer was confirmed (based on low 

qt values and classification as silty sand to sand; see discussion of classification using CPT in section 

5.2.3). Additional cone testing was performed across the site to delineate the extent of the 

liquefiable zone. 

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40

50

60

70

80

90

100

110

0 10 20 30 40 50

(N1)60 (BLOWS/FT), qt (TSF)/10

ELEVATION (FT MSL)

Boring G1

Boring G5

CPTu-03

CPTu-07

Figure 37. Comparison of SPT (N1)60 and CPT qt values. 

5.2.3 Soil Classification by Cone Penetration Testing 

Cone penetration testing (CPT) may be used to obtain detailed site stratigraphy information. The 

CPT is rapid, relatively inexpensive, and facilitates the collection of continuous subsurface 

stratigraphy data. Soil classification using the CPT is based on measured values for cone tip 

resistance (qt), and sleeve friction (fs) presented as friction ratio (FR) defined as the ratio of sleeve 

friction to cone tip resistance (i.e., FR = Rf = fs/qtx100). The piezocone (CPTu) can also be used for 

soil classification based on measured pore pressure (u2) during advancement of the penetrometer. 

In general, the tip resistance (qt) in sands is greater than 5 MPa whereas the tip resistance for clayey 

and silty materials is typically less than 2 MPa. Penetration porewater pressures in sands are often 

near hydrostatic values since the permeability of sands is relatively high such that negligible excess 

penetration porewater pressures are induced by penetration of the piezocone. In soft to firm, 

normally to lightly overconsolidated clays, measured u2 values are often 3 to 10 times the value of 

hydrostatic porewater pressure. When using a penetrometer with the pore pressure transducer 

located at the shoulder position (i.e., the u2 position), measured porewater pressures in heavily 

overconsolidated and fissured clays and dense silts are often negative due to the demonstrated 

behavior for these materials to dilate during undrained shearing. It is noted that porewater pressures 

measured using a penetrometer with the pore pressure transducer located at the tip position (i.e., the 

u1 position) are always positive, regardless of the soil. Penetration porewater pressures that are 

lower than expected may indicate poor saturation of the filter element. 

In addition to the previously referred general characteristic, friction ratio, tip resistance, and pore 

pressure data have been used with empirical soil behavioral classification charts for detailed soil 

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classification. Examples of these charts are shown in figures 38 and 39 and the use of these charts is 

illustrated in Soil Property Selection Example No. 1 and No. 2. 

When using pore pressure data from CPTu, the normalized pore pressure parameter Bq should be 

used: 

Normalized Pore Pressure Parameter,

vot

o

q q

uu B −σ− = 2 (Equation 30) 

It is noted that the term “classification” used here does not refer to a standard classification system 

such as USCS. Since more traditional laboratory classification is based on a mechanical analysis of 

“disassembled” particles and remolded fines material, it is impossible for a CPT to provide the same 

“classification” information since the CPT readings provide information on the in-situ soil. 

Therefore, the CPT and CPTu should be considered to provide an assessment of soil behavior type 

and an indirect assessment of soil “classification”. 

The CPT and CPTu typically record data (qc, fs, u2) at 1-sec duration time intervals, therefore 

permitting nearly continuous subsurface profiling. Typically, these data are not plotted manually on 

the classification charts (i.e., figures 38 and 39) since most CPT contractors have software available 

that automatically classifies each set of data using these charts. 

Figure 38. Soil classification based on 

qt and FR (Robertson et al., 1986).

Figure 39. Soil classification based on 

qt and Bq (Robertson et al., 1986).

Soil Behavior Type (Robertson et al.,1986; Robertson & Campanella, 1988) 

1 – Sensitive fine grained 5 – Clayey silt to silty clay 9 – sand 

2 – Organic material 6 – Sandy silt to silty sand 10 – Gravelly sand to sand 

3 – Clay 7 – Silty sand to sandy silt 11 – Very stiff fine grained* 

4 – Silty clay to clay 8 – Sand to silty sand 12 – Sand to clayey sand* 

 * Overconsolidated or cemented 

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To develop a subsurface profile for design analyses based on CPT or CPTu data, the logs should first 

be reviewed to establish soil units based on distinct breaks in the qt, fs, FR, and u2 signatures. Figure 

40 shows a CPT log for an embankment constructed over dredged material, natural clays, and sands. 

The soil classification column shows numerous classifications based on the nearly continuous data. 

To develop a profile for design, distinct breaks in the actual cone data (e.g., qt signatures) should be 

reviewed. A review of figure 41 indicates that a significant change in material type occurs at 

approximately a depth of 18 ft (5.5 m) as defined by a distinct drop in qt and fs along with a sharp 

increase in u2. These trends indicate that at this depth the CPT has penetrated into a clay unit 

characterized by low qt, high FR, and increased pore pressures due to the lower hydraulic 

conductivity of the clay. Another distinct break occurs at 32 ft (9.8 m) as evidenced by increased qt, 

decreased FR, and decreased u2 to essentially hydrostatic pressure. These trends in the data indicate 

that a sandy material has been encountered. Also shown on this profile are SPT blowcount data 

from an adjacent boring. Thus, the CPT characterization presented here is consistent with the SPT 

boring. It is worth noting that although the SPT N value is similar for the lower portion of the dike 

sands and the dredged material (with SPT N = 11), the CPT signature does indicate that there is a 

distinction between these two materials as evidenced by the rapid decrease in tip resistance. The 

distinction between these two layers was also confirmed by reviewing disturbed SPT samples for the 

project. 

Once the main units have been identified, the soil classification columns of the logs should be 

reviewed. For the example shown in figure 41, the upper sand unit has been classified into three soil 

types (i.e., sand, silty sand, and sandy silt). Figure 38 and 39 shows the close proximity of each of 

these categories to each other and how scatter in the data from one unit could be plotted in several 

different zones. Such variations within a unit should generally be considered negligible, meaning 

that such distinctions will not typically need to be considered in developing a subsurface profile for 

design analyses. When dividing the subsurface into units, more attention should be given to overall 

groupings rather than to individual classifications. Experience has shown that minor variations in 

grain size throughout a unit across a site explain this subtle classification difference, although these 

changes generally have little to no significant change in engineering properties. 

Groundwater elevations can also be evaluated using the CPTu provided that relatively clean sands 

exist within the subsurface. Under hydrostatic conditions, the groundwater elevation can be 

obtained by extrapolating the water pressure readings in a relatively clean sand unit back to zero (see 

figure 40). Pore pressure readings in clay or silt should not be used for this purpose. Alternatively, a 

pore pressure dissipation test can be performed in a sand layer to ensure that all penetration pore 

pressures have dissipated. Figure 42 shows the results of a dissipation test in a sand layer in which 

dissipation occurs instantaneously. For the example shown in figure 42, the dissipation test was 

performed at a depth of 35.6 ft. The pore pressure head measured was approximately 24 ft 

indicating that the ground water table was located at approximately a depth of 12 ft. The in-situ pore 

pressure condition, uo, can also be evaluated by performing full dissipation tests in more fine-grained 

materials such as clays and silts, however, the tests will take much longer due to the lower hydraulic 

conductivity of these materials. Estimation of uo from CPTu pore pressure dissipation tests in low 

hydraulic conductivity materials is usually not practical. On the other hand, in a Piedmont silty soil, 

full dissipation occurs within 1 to 2 minutes and thus u2 data at rod breaks can be used to estimate uo. 

The use of dissipation tests to obtain time rate of settlement information is discussed in section 5.4.4. 

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Figure 40. Typical CPTu log. 

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Figure 41. CPTu log with subsurface stratigraphy interpretation.

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Figure 42. CPTu dissipation test in sand. 

The CPT is most useful when results are correlated to known subsurface stratigraphy conditions. If 

the CPT is used in conjunction with other intrusive investigation methods (e.g., SPT), at least one 

CPT boring should be placed adjacent to a boring. If possible, the CPT results should be correlated 

with the physical samples recovered from an adjacent boring. With this correlation, similar CPT 

results across the site, possibly far from a sampling location, can be related to a particular soil type. 

5.2.4 Soil Classification using the Flat-Plate Dilatometer 

The DMT uses empirical charts for indirect soil classification. The chart shown in figure 43 utilizes 

the relationship between dilatometer modulus (ED) and material index (ID) to evaluate: (1) soil type 

(i.e., clay, silt, sand); and (2) consistency (soft vs. hard, loose vs. dense). Soil type is categorized by 

the following ranges: 

• Clay: ID < 0.6 

• Silt: 0.6 < ID < 1.8 

• Sand: ID > 1.8 

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In the absence of other data, this DMT classification can be used in preliminary design, but should 

be confirmed by comparing classification results with other site-specific testing of either soil type 

and/or behavior such as that from laboratory classification, SPT, and CPT data (if available). 

When using a DMT for soil classification and stratigraphy, the previously referenced comments 

regarding the CPT are applicable. Specifically, the DMT should be correlated on a site-specific 

basis with SPT and/or tube samples wherever possible. Additionally, it is recognized that since the 

DMT (like the CPT) provides quantified data on regular intervals, subtle differences in the data can 

result in a different “classification” while not representing a difference in engineering parameters. 

Low Density

Low Rigidity

Loose

Soft

Med. Rigidity

Medium D...

Compressible

Medium

Rigid

Dense

Very Rigid

Very Dense

Hard

CLAY

SILTY CLAYCLAYEY SILTSILTSANDY SILTSILTY SANDSAND

10

100

1000

0.1110

Material Index, ID

Dilatometer Modulus, E

D (atm)

V. SOFT CLAY/PEAT

Figure 43. Soil classification based on DMT. 

5.2.5 Generating a Subsurface Profile 

A primary objective of any subsurface investigation is to obtain sufficient data to enable a twodimensional drawing depicting the site geostratigraphy. This drawing is called a subsurface profile. 

A sufficient quantity of data must be collected to characterize soil variability with respect to aerial 

and vertical distribution across the site. The purpose of stratigraphic profiling is to develop a simple 

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working model that depicts the major subsurface layers exhibiting distinct engineering 

characteristics. 

Once the stratigraphic information is compiled from the previously discussed methods (i.e., borings, 

CPT soundings, DMT soundings), final logs should be generated. The final logs are created using 

field notes and boring logs. Results of each boring, CPT, and DMT should be used to develop a 

preliminary model of soil units at each boring/sounding location, including elevation and depth 

intervals. This model should be added to and modified as each log is analyzed. Once an initial 

screening has been conducted on all of the logs, each of the completed logs should be placed on a 

large flat surface that is scaled to present the data in close proximity to each other, yet spaced 

horizontally and vertically on the surface to reflect their relative location in the field. The distinct 

units in each log should be compared with those in the adjacent logs. Attention to depth, soil type, 

distinct features (e.g., shell fragments, root matting, etc.), and engineering properties (e.g., Atterberg 

limits, strength, settlement parameters, etc.) will help in the correlation. The engineer or geologist 

should try to match these units across the site area, keeping in mind that conditions may change 

laterally depending on the geologic conditions. For example, alluvial units in floodplain deposits 

may “pinch out” laterally (i.e., localized lenses of material may exist over relatively short lateral 

distances). Thus, the importance of local and regional geology cannot be over emphasized. 

Index properties such as Atterberg Limits, moisture content, and density provide information 

required to establish the stratigraphy in relation to the design parameters. Plotting these values with 

depth (see figure 44) can provide baseline information on the engineering properties of fine-grained 

soils. For example, natural moisture contents near or greater than the liquid limit indicate that a soil 

is normally consolidated and likely to be sensitive (i.e., loss of soil structure and strength upon 

disturbance). Overconsolidated clays or desiccated crusts may be identified by natural moisture 

contents near the plastic limit. Significant changes in index properties with depth should be used in 

conjunction with other data (e.g., CPT) to identify subsurface units. 

After completing this exercise and grouping subsurface units based on their engineering properties, 

the engineer or geologist should create cross-sections of the site subsurface conditions. This entails 

plotting the borings at their respective elevations and relative positions horizontal to one another. 

Often, the vertical scale is exaggerated compared to the horizontal scale. The different strata are 

grouped based on color, texture, consistency, and soil type of the recovered samples or profiles from 

the soundings. If possible, at least two cross-sections, roughly perpendicular to each other should be 

constructed, as these will provide for an understanding of lateral trends in stratigraphy. Groundwater 

elevations should be plotted on the cross-section based on site-wide measurements taken on one day. 

It is likely that the conceptual model based on these cross sections will be modified as additional 

data are collected and evaluated. An example boring location plan and subsurface profile are shown 

in figures 45 and 46. If sufficient information is available, an additional graphical figure with threedimensional perspective (i.e., a fence diagram) can be developed. 

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Figure 44. Summary plot of Atterberg limits data. 

Historical boring logs from the area can be used to supplement current boring logs in developing a 

subsurface profile, however, such historical logs need to be carefully reviewed well in advance of 

drilling activities to ensure that the data are accurate. In some cases, boring log locations are 

referenced to the center alignment of a roadway or highway without actually surveying the location 

of the borehole. It is imperative to ensure that a consistent coordinate system has been used to 

establish the correct relative location of all borings. Since borings would have likely been performed 

over an extended period of time or for different contracts along a roadway alignment (i.e., project 

centerlines are commonly changed during project development), it is possible that coordinate 

systems will not be consistent. Simply stated, if a historical boring cannot be confidently located on 

a site plan due to these issues, then the boring has limited usefulness for establishing stratigraphy. 

Also, it is likely that different drill rigs with different operators and different energy efficiencies 

were used in the collection of SPT data on historical boring logs. This must be recognized when 

attempting to correlate engineering properties to SPT blow count values. The engineer, however, 

should consider that while these issues represent potential limitation in the usage of historical 

borings, it is necessary to review these borings relative to the design under consideration. As an 

example, a historical boring may indicate a thick layer of very soft clay as evidenced by weight of 

rod/weight of hammer. While shear strength and consolidation properties cannot be reliably 

estimated based on SPT blow count values, the historical boring may provide useful information 

concerning the depth to a firm stratum. 

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Figure 45. Boring location plan 

Figure 46. Interpreted subsurface profile. 

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Most DOTs have collected large amounts of subsurface data from previous investigations within 

their states. Unfortunately, much of these data are archived with related project data once the project 

has been completed, and thus may not be readily available or accessible for use during future 

projects. Additionally, the subsurface data may not be fully utilized if the locations of the borings 

are not identified or if the plan drawing of the project site is not maintained with the boring logs. To 

overcome this problem, many DOTs currently use longitude and latitude to identify the boring 

locations, in lieu of (or in companion with) the conventional positioning format that uses station and 

offset. To fully appreciate and use these historical data, the profession now realizes the benefits of 

using electronic boring logs and/or establishing an electronic database of the subsurface information. 

Unfortunately, the vast majority of the historical subsurface boring information is provided via paper 

copies, as it was not common (and still is relatively uncommon) to record electronic boring records. 

Several DOTs have recently commenced using electronic boring records for their projects. Not only 

does this provide a redundant record to compliment the paper copy, but it also preserves data in a 

way that has the potential for automated electronic data management. One of the most useful 

techniques applied and recommended by the authors, and increasingly used by DOTs, involves the 

use of a centralized database and Geographic Information System (GIS) techniques that are 

maintained by the DOT. In its most simplistic form, the electronically stored data are managed and 

assessed visually using GIS software, where each boring location is identified on a plan map. The 

authors have found that an appropriately developed database and GIS can be used to great advantage 

by the DOT. Specifically, in addition to the previously mentioned advantages related to electronic 

data records, to compliment the paper logs, it is possible to: (1) catalog the borings that previously 

have been conducted; (2) inventory data regarding specific problematic formations across the state; 

and (3) develop cross sections that depict subsurface conditions across a site or within a region. This 

type of application tool can facilitate the development of subsequent subsurface investigations that 

are appropriately focused and that optimize the utility of existing data. 

Recently, the authors have been involved in projects that augment and extend the concepts of 

database management and GIS technologies described above. Specifically, the authors note that a 

forms-based or map-based system can be used to develop electronic data records related to 

subsurface characterization. Using this technique, a handheld computational device (e.g., handheld 

computer, Palm, Jornada, etc.) can be used in such a manner that the inspector is directed to 

complete an electronic checklist on the handheld computer. The checklist includes information 

regarding the top and bottom of the encountered stratigraphic units, characteristics of the strata (e.g., 

primary and secondary descriptors, color, consistency) in accordance with recommendations 

presented in ASTM D 2487. In addition, the inspector is prompted to provide information regarding 

any obtained samples, including the sampling interval, type of sample, etc. In short, using the 

handheld computer and a properly organized electronic form, it is possible to systematically collect 

all subsurface information that would normally appear on a conventional boring log. The primary 

difference is that now the information is available as electronic data, and not just a paper record. 

After the subsurface data are recorded, the handheld computer is synchronized with a host computer. 

At this point, the collected field data are transferred to a geotechnical database that is maintained on 

the host computer. Information in the database can be interpreted/viewed using GIS technologies. 

Importantly, this database can be queried to provide specific information for input into a dedicated 

boring log program, where a paper copy record of the boring log can be printed. Using this system 

of data management, three advantages are realized: (1) the field data do not have to manually entered 

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into a database or directly typed on a boring log; (2) the prompts that are programmed on the 

handheld computer assure a degree of consistency and uniformity in the recorded data that has 

heretofore not been possible; and (3) the recorded data become part of a database where they can be 

treated as data, not merely electronic information. Edits to the information in the database that are 

noted after synchronization and printing can simply be made in the database itself, and do not 

involve the handheld computer. 

5.3 IN-SITU STRESS STATE 

5.3.1 General 

The current state of stress in the ground is required for the design of most geotechnical features. 

This state of stress is defined by the total and effective vertical and horizontal stresses, as well as the 

preconsolidation stress. The overburden stresses depend on gravity and buoyancy forces whereas 

horizontal stresses and the preconsolidation stress are more complex and depend upon the stress 

history of the soil. 

5.3.2 Overburden Stresses

This section provides a summary of the methods used to calculate effective overburden stresses. 

The total vertical (overburden) stress is calculated as the cumulative sum of the total unit weights 

(γT) with depth, σvo = Σ(γT)dz, summed from the ground surface (z = 0) to the depth of interest. If 

the depth to the water table is designated zw and the depth of interest is z, then uo =γw(z-zw). The 

effective vertical stress is then calculated as: 

vo σσ −= uovo

' (Equation 31) 

The unit weight or density of the soil can be assessed using techniques described previously in 

chapter 4 for recovered samples. Alternatively, for soils located at depths below the groundwater 

table, soils can be assumed to be completely saturated and the saturated unit weight of soil is taken 

as: 

)1(

)(

o

osw

sat e

eG

+

+ = γ

γ (Equation 32) 

where Gs = specific gravity of solids (≈ 2.7 ± 0.1 for many soils) and eo = in-situ void ratio. 

Above the water table, the soil may be dry, partially dry, partially saturated, or completely saturated 

due to capillarity effects. In the case of clean sands, it is often assumed that the soil is dry, and 

therefore, the unit weight of soil is given by: 

)1()1( n

sat

o

sw

d

we

G

+ = + = γ γ γ (Equation 33) 

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Moreover, for a completely dry soil, the pore pressure is zero. Therefore, the total and effective 

overburden stresses are equal. In clays, however, capillarity effects can result in complete saturation 

for heights of 10 meters or more above the water table and therefore γsat may be appropriate. A 

corresponding negative hydrostatic porewater pressure occurs above the phreatic surface (free water 

table) which is calculated by the relation: uo =γw (z-zw). In this case of negative pore pressure, the 

effective vertical stress is larger than is the case of dry soil at the same depth. 

5.3.3 Horizontal Stresses 

The horizontal state of stress in natural soils is a difficult parameter to assess accurately, however, 

for most design applications involving transportation projects, the horizontal stress can be estimated 

with sufficient accuracy using the correlations described in this section. 

The effective horizontal stress in the ground (σho′) is related to the vertical effective stress by the atrest lateral earth pressure coefficient, Ko: 

'' K = σσ vohoo (Equation 34) 

In the absence of direct measurements, laboratory testing on a variety of soils using oedometer and 

triaxial specimens have shown that the magnitude of Ko depends strongly on the stress history and 

frictional characteristics of the deposit. For normally consolidated soils, the value of Ko can be 

estimated using the following relationship: 

o φ−= 'sin1K (Equation 35) 

The magnitude of Ko increases during unloading (such as that caused by removal of soil overburden) 

and becomes greater than unity for OCR values exceeding about 4. For overconsolidated soils 

(clays, silts, sands, and gravels), Ko can be calculated according to the following equation: 

'sin

o OCR)'sin1(K φ φ−= (Equation 36) 

Studies involving field tests in these materials have shown generally good agreement with these 

correlations, however additional factors such as cementation, aging, structuring, and desiccation may 

alter the magnitude of in-situ Ko. Where a more accurate assessment of Ko is required for design, it 

is recommended that either laboratory triaxial tests on undisturbed samples subjected to Ko

consolidation or in-situ pressuremeter, total stress cells, Iowa stepped blade, or hydraulic fracture 

testing be performed. Information on the latter three in situ tests can be found in Tavenas et al. 

(1975), Handy et al. (1982), and Jamiolkowski et al. (1985). For Ko consolidation in a triaxial test, 

cell pressures are automatically controlled to maintain a condition of zero lateral strain (i.e., the Ko

condition). 

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5.4 CONSOLIDATION PROPERTIES OF SOIL 

5.4.1 General 

In this section, the interpretation of laboratory-measured consolidation properties for cohesive soils 

is presented. These properties relate to primary settlement, (Cc and Cr), secondary settlement (Cα), 

time rate of settlement (cv), and stress history (σp′). As will be demonstrated later, the consolidation 

characteristics of a soil provide invaluable information regarding the strength characteristics of the 

same soils. Interpretation of σp′ from various in-situ test methods and evaluation of ch from CPTu 

dissipation data is also provided. 

Consolidation properties are required to perform time-dependent settlement analyses of soils for 

embankments, structural foundations, and retaining structures. Consolidation refers to the time 

dependent decrease in volume of a soil due to the dissipation of pore pressures within the soil mass. 

Shortly after a soil is first loaded (i.e., after immediate or undrained distortional settlements have 

occurred) the stresses are transmitted to the pore fluid in the soil mass resulting in excess pore 

pressures. As these pore pressures dissipate with time, the load is gradually transferred to effective 

stress within the soil skeleton. The resulting increase in effective stress results in a decrease in 

volume that causes settlements. The general consolidation characteristics of various major soil types 

can be used to evaluate whether a settlement problem should be anticipated for a particular 

geotechnical feature according to the following: 

• Gravels, sands, and non-plastic silts: These soils consolidate rapidly under load and do not 

typically present settlement problems unless close tolerances are required for the project. 

• Plastic silt-clay mixtures: Soft silts and clays are more compressible than stiff silts and clays. 

Settlement may continue long after construction is complete. 

• Organic soils: These soils are very compressible as well as biodegradable and can result in 

large settlements that occur for many years. 

5.4.2 Laboratory Consolidation Tests 

5.4.2.1 General 

Consolidation properties of cohesive soils are typically evaluated in the laboratory using the onedimensional consolidation test. The most common laboratory method is the incremental load (IL) 

oedometer (ASTM D-2435). The constant rate of strain (CRS) version (ASTM D-4186) applies the 

loading continuously while measuring stress and pore pressures by transducers, thereby reducing 

testing times from 1 to 2 weeks by IL oedometer to say 1 day by the CRS consolidometer. Highquality undisturbed samples using thin-walled tubes, piston samplers, or other special samplers are 

required for laboratory consolidation tests. Tests performed on disturbed samples will result in 

computed consolidation properties that may result in conservative or unconservative design 

evaluations. Important issues relative to consolidation testing such as sample disturbance effects 

have been described in chapter 4. 

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5.4.2.2 Soil Parameters from Laboratory Consolidation Tests 

Consolidation test results are traditionally shown in a graph of void ratio (e) versus logarithm of 

applied vertical effective stress (σv') in the oedometer, as illustrated in figure 47. Each of the 

compression indices are defined by the change in void ratio per log cycle stress (= ∆e/∆logσv') for 

the respective ranges of recompression (Cr), virgin compression (Cc), and swelling or rebound (Cs). 

An alternate version of presenting consolidation test results is using a plot of vertical strain as the 

ordinate axis, whereby εv = ∆e/(1+eo). In this case, the compression indices are reported as the 

recompression ratio, Crε = Cr /(1+eo), and compression ratio, Ccε = Cc /(1+eo). 

Figure 47. Definition of Cc, Cr, Cs, and σp′. 

The maximum preconsolidation stress delineates the region of semi-elastic behavior (corresponding 

to overconsolidated states) from the region of primarily plastic behavior (associated with normal 

consolidation). The degree of preconsolidation is expressed using the overconsolidation ratio, where 

OCR = σp'/σvo'. Other important parameters obtained from the consolidation test include the 

constrained modulus (D = ∆σv /∆εv = 1/mv), and time-dependent parameters: coefficient of vertical 

consolidation (cv) and coefficient of secondary compression (Cα). 

5.4.2.3 Selection of Samples for Laboratory Consolidation Testing 

Information on subsurface stratigraphy and basic index properties should be used to select the 

number and depths of undisturbed samples for laboratory consolidation testing. The number of 

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samples should be selected to facilitate the development of a profile of preconsolidation stress with 

depth. The selection of samples for consolidation testing (and shear strength testing) should be 

performed in a rational manner since such laboratory tests are time consuming, expensive, and 

provide important performance properties. 

The procedure used to select samples includes developing a plot of the vertical effective stress, σvo′, 

with depth (see Soil Property Selection Example No. 1 in appendix A). The location of each 

undisturbed sample should be plotted for its specific boring and the vertical effective stress 

evaluated. For consolidation tests, it is imperative that the test be carried out to stress levels much 

greater than the estimated preconsolidation stress to facilitate an accurate evaluation of the actual 

preconsolidation stress. It is useful to develop an estimated profile of preconsolidation stress with 

depth so that critical points can be noted. The engineer should recognize that the preconsolidation 

stress (σp′) (or alternatively the OCR) of fine-grained materials is an important property because it 

significantly affects settlement calculations and the evaluation of undrained shear strength. An 

example plot showing preconsolidation stress with depth is shown in figure 48. 

Figure 48. Profile of preconsolidation stress. 

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128 

5.4.2.4 Evaluation of σp′ from Laboratory Consolidation Tests 

Standard oedometer tests are conducted on small specimens taken from the field and the 

characteristic e-logσv' graphs show a change in slope at the preconsolidation stress. Sampling 

disturbance effects typically “lower” the overall e-logσv' curve relative to field conditions (see figure 

23). Consequently, the value of σp' is often underestimated in routine testing and interpretation. 

Sample disturbance can be quantified using the measured strain level that corresponds to σvo' (see 

table 16). 

The preconsolidation stress from oedometer tests is normally interpreted from the e-logσv' 

relationship using the Casagrande graphical technique. In many clays and silts, this approach may 

be adequate in evaluating a reasonable value of σp' for engineering purposes. The steps involved in 

the Casagrande method are provided in table 24. 

In very soft, sensitive, or structured materials, particularly those affected by sample disturbance, 

swelling, and the release of stress associated with removal from the ground, alternative graphical 

techniques can be utilized to better delineate the magnitude of σp′. Herein, the strain-energy method 

(Becker et al., 1987) is recommended for interpreting σp′. The strain-energy method, however, is 

valid for evaluating σp’ for any laboratory consolidation test, not just for samples that are considered 

to be disturbed. The strain-energy method involves plotting the cumulative strain energy (i.e., the 

product of stress times strain) for each load increment in a laboratory consolidation test. The point 

where the strain energy plot exhibits a large incremental increase represents the preconsolidation 

stress for the soil. The strain-energy method is outlined in table 25. 

Example calculations of σp′ using both the Casagrande and strain-energy methods are provided in 

Soil Property Selection Example No. 1 and No. 2 in appendix A. 

Table 24. Casagrande method to evaluate σp′. 

1. Construct a line tangent to the steepest portion of the consolidation curve (within the normally 

consolidated range). 

2. Locate the point of maximum curvature in the area of the laboratory curve where the slope 

transitions from shallow to steep. Construct a horizontal line from this point of maximum 

curvature. 

3. Construct a tangent line to the curve from the point of maximum curvature. 

4. Construct a line that bisects the angle between the horizontal line constructed in Step 2 and 

the tangent line constructed in Step 3. 

The point of intersection between the bisector line (Step 4) and the first tangent line (Step 1) is the 

location of the preconsolidation pressure. A diagram of the Casagrande method is provided in 

figure 49. 

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Table 25. Strain-energy method to evaluate σp′. 

1.Compute the change in work per unit volume for each strain increment according to: 

( )if

fi W εε

σσ − +

=∆ *

2

)''( (Equation 37) 

where: ∆W = change in work (kPa/m3

) per unit volume 

σi = stress at beginning of strain increment (kPa) 

σf = stress at end of strain increment (kPa) 

εi = strain at beginning of increment (dimensionless) 

εf = strain at end of increment (dimensionless) 

2. Plot stress versus summation of work for each respective stress increment. For plotting purposes, it is 

assumed that the stress value corresponding to the summation of work is the stress at the end of the 

strain increment. 

3. A noticeable change in slope should be evident when data are plotted. A curve connecting the data 

should have a sharp transition from a flatter slope in the recompression range (slope 1) to a steeper slope 

(slope 2) in the virgin compression range. Construct a trend line through the data that represent a line 

with slope 1. Construct a second line through the data that represent a line with slope 2. 

4. The stress where these two trend lines intersect is the preconsolidation stress. The strain-energy method 

is illustrated in figure 50. 

0.25

0.35

0.45

0.55

0.65

0.75

0.85

10 100 1000 10000

Effective Consolidation Stress (kPa)

Void Ratio, e

σp' = 230 kPa

4

3

2

1

Figure 49. Illustration of Casagrande method to evaluate preconsolidation stress. 

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0

100

200

300

400

500

600

700

800

0 2000 4000 6000 8000 10000

Unit Strain Energy (kPa/m3

)

Stress (kPa)

σp' = 250 kPa

1

2

0

10

20

30

40

0 150 300 450 600

Unit Strain Energy (kPa/m3

)

Stress (kPa)

σp '= 250 kPa

1

2

Figure 50. Illustration of strain energy method to evaluate preconsolidation stress. 

5.4.2.5 Evaluation of Cc and Cr 

Compression parameters can be estimated by correlation or by direct laboratory-measurement. The 

compression index, Cc, and recompression index, Cr, are index values required for primary 

consolidation settlement predictions. These parameters have been defined in figure 47. The value of 

Cc is evaluated by drawing a best-fit tangent line to data on an e-log σvc′ representation of 

consolidation data along the virgin (i.e., part of curve where stresses are greater than σp′) portion of 

the curve. The modified compression index, Cce, is evaluated similarly on a plot of εv-log σvc′. The 

line drawn to evaluate Cc (or Ccε) should include the stress range defined by the calculated σp′ and 

the final effective stresses in the ground. Typically, values of Cr (or Crε) are 10 to 20 percent of the 

value of Cc (or Ccε). 

Numerous correlations relating simple soil classification properties (e.g., LL, wn) to Cc and Ccε are 

available in the literature for silts and clays. These correlations can be used to make first-order 

predictions of settlements, but should not be relied upon for final design, unless the correlation has 

been developed using site-specific laboratory consolidation test data. These correlations may be of 

limited value, however, for highly structured soils that are sensitive. Several correlations that are 

based on relatively large databases are provided in table 26. Other correlations are presented as part 

of Soil Property Selection Example No. 1 in appendix A. 

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Table 26. Summary of correlations for Cc (modified from Holtz and Kovacs, 1986). 

Equation Applicable Soils 

Cc = 0.009 (LL –10) Undisturbed clays of low to medium sensitivity 

Cc= 0.007 (LL - 7) Remolded clays 

Cc = 0.01 wn Chicago clays 

Interpreting a compression and recompression index value for design should be based on a rational 

assessment of the data. The objective is to assign a value to each behaviorally different subsurface 

layer or to assign some representative value for the entire subsurface. Assessments to be made in 

evaluating compression data include: (1) depth ranges where the material is more silty or sandy as 

compared to other depth ranges; (2) depth of transition from a crust layer to an underlying softer clay 

layer; and (3) assessment of sampling disturbance. 

The sampling disturbance indices previously discussed offer a useful means to evaluate calculated Cc

or Ccε values. As previously discussed measured Cc (or Ccε) values from disturbed samples will be 

lower than those measured from undisturbed samples. Available data in the literature indicates that 

modified compression index values for disturbed samples may be, on average, about 15 to 25 

percent lower than values from “field-measured” (i.e., undisturbed) curves. It is noted, however, that 

the effects of sample disturbance on the evaluation of Cc (or Ccε) can be minimized by carrying the 

consolidation test out to a stress level of approximately 8 times the estimated preconsolidation stress. 

A reasonably accurate assessment of Cc (or Ccε) can then be made by drawing a tangent line to the 

straight line portion of the consolidation curve at these high stress levels. It is recognized that such 

high stress levels may not be able to be achieved for standard laboratory equipment, especially if the 

estimated preconsolidation stress is itself quite large. However, since soils with high 

preconsolidation stresses pose little concern for typical geotechnical features (since loading will not 

likely cause the soil to experience stresses greater than its preconsolidation stress), then a sufficiently 

accurate estimate of Cc (or Ccε) can be made for these soils at more moderate loading levels. 

5.4.2.6 Laboratory Evaluation of cv

The time rate of settlement is typically represented in geotechnical practice via the coefficient of 

vertical consolidation, cv and the coefficient of horizontal consolidation, ch. The parameter cv is used 

for evaluating time rate of settlement for shallow foundations and large aerial fills whereas the 

parameter ch is used for estimating pore pressure dissipation around driven piles and for designing 

wick drains. Graphical procedures used to evaluate cv from laboratory consolidation data include 

Casagrande’s logarithm of time method and Taylor’s square root of time fitting method. 

Casagrande’s method uses the time to complete 50 percent primary consolidation and evaluates cv

according to: 

50

2 197.0

t

H

c DR

v = (Equation 38) 

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where HDR is the drainage height (equal to one-half the average thickness of the oedometer test 

specimen for each load increment for a double drained specimen) and t50 is the time required to 

achieve 50 percent primary consolidation. For the square root of time method, the time for 90 

percent primary consolidation, t90, is used and cv is calculated according to: 

90

2 848.0

t

H

c DR

v = (Equation 39) 

Procedures for these methods are provided in appendix B. Each of these methods is approximate 

and will result in different calculated values for cv even though the same deformation-time data are 

used for both methods. The use of the logarithm of time method requires that a substantial part of 

the consolidation curve be defined after 90 percent consolidation is complete to facilitate a good 

approximation of t90. For the square root of time method, the test does not need to be carried out 

beyond t90. It is noted, however, that if secondary compression is to be evaluated, the test will need 

to be carried out beyond the time to reach 100 percent of primary consolidation. For some soils, 

such as soils which undergo significant secondary compression, these graphical methods may be 

difficult to implement to obtain the time to the end of primary consolidation since their deformation 

versus time response may not “look” like those used to develop the graphical methods. Recalling 

that the definitions of the time of primary consolidation, tp, is the time at which all excess pore 

pressure has dissipated for a given load (i.e., ∆u=0), it may be beneficial to perform CRS oedometer 

tests for these soils since pore pressures are measured during the test. 

Significant scatter of calculated cv values in the overconsolidated range may result from the 

following: (1) consolidation occurs quite rapidly at these load levels making the determination of the 

time for the end of primary consolidation difficult; and (2) in very stiff clays, fissures may exist at 

low stress levels which will affect drainage rates. Some of the inherent variability associated with 

evaluation of this parameter can be minimized by concentrating the interpretation on values 

corresponding to a reload cycle and to values associated with virgin compression. Values from 

initial loading should not be used due to inevitable sample disturbance effects. 

A value for cv should be calculated for each load increment for each consolidation test. For each 

test, values of cv should be plotted as a function of log σvc′ (i.e., cv vs. log σvc′ plot). High quality 

data will typically exhibit a sharp reduction in calculated cv values near the preconsolidation stress. 

Figure 51 shows a summary plot for a consolidation test. Index properties should also be reported 

with the consolidation test plots. 

Engineers typically use a constant value of cv to perform time rate of settlement analyses, however, it 

should be noted that cv is not a constant value for a test on a particular soil (see figure 51). Values 

for cv depend on many factors including whether the preconsolidation stress has been exceeded. For 

load increments less than the preconsolidation stress, consolidation occurs relatively rapidly and cv

values can be relatively high. The typical trend for most clayey soils is that cv values are higher in 

the overconsolidated range and exhibit a relatively rapid decrease as the preconsolidation stress is 

approached. Values for cv will be lowest for effective stresses that exceed σp′. 

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Figure 51. Summary consolidation data showing cv. 

A rational approach to selecting a cv value for design would include first assessing whether the 

design loads are sufficiently high to consolidate the compressible layer to stress levels beyond the 

preconsolidation stress. If stress levels after consolidation are below approximately 0.8σp′, an 

average cv value associated with stress levels below 0.8σp′ should be used for design. If the 

compressible layer will consolidate to stress levels in excess of σ′p (or greater than 0.8σp′ as 

described herein), then the average cv value for the range of stresses from σp′ to the final vertical 

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effective stress in the ground should be used. This value, however, should not exceed the average 

value obtained for the overconsolidated stress range (i.e., less than 0.8σp′). 

A value for cv based on laboratory consolidation results may vary significantly from actual values in 

the field. As previously discussed, time rate of settlement analyses based on consolidation test 

results on disturbed samples will likely result in an overprediction of the time actually required for 

primary consolidation to be completed in the field. In addition, laboratory tests only simulate 

vertical drainage whereas most natural soil deposits have interbedded seams or layers of more 

permeable material within the low permeability layer. These smaller layers will permit lateral 

drainage as well, which will tend to decrease the time required to complete primary consolidation. 

All of these factors tend to make predictions of the time rate of settlement to be conservatively 

overestimated. The coefficient of consolidation can be obtained most accurately in the field using 

measured pore pressures from piezometers installed at several depths within a clay layer. 

Alternatively, monitoring the actual time rate of settlement in the early stages of loading can be used 

to assess the appropriate value of cv and this value can be used to refine the predicted time to 

complete primary consolidation. 

The parameter cv is also important for developing a fill placement plan for an embankment that is to 

be constructed over clays. Failures of embankments constructed over clays are often caused by 

placing the fill too rapidly, i.e., not providing sufficient time to enable construction-induced pore 

pressures in the clay foundation to stabilize. If time permits, staged construction represents the most 

cost-effective solution and it is therefore worth the effort to perform high quality laboratory tests to 

evaluate the cv parameter. Oftentimes, staged construction over an extended period of time is 

deemed inappropriate because of the need to put the structure into service as soon as possible. With 

this, ground improvement techniques may be necessary. 

5.4.2.7 Evaluation of Cαε

In geotechnical design analyses, it is assumed that secondary settlement occurs after primary 

consolidation is completed. As noted earlier, secondary compression settlements may be relatively 

large for organic soils. For normally consolidated soils, the ratio of the coefficient of secondary 

compression to the compression index (Cα/Cc = Cαε/Ccε) is relatively constant for a given soil. On 

average, the value of Cα/Cc is 0.04±0.01 for inorganic clays and silts. For organic clays and silts, the 

value averages 0.05±0.01. For peats, the value averages 0.06±0.01. These values may be used to 

assess actual values from laboratory tests or for preliminary analyses. If the final effective stress in 

the ground is less than the preconsolidation stress, then Cr should be used instead of Cc to estimate 

the secondary compression index. 

The coefficient of secondary compression is calculated using the portion of the deformation versus 

time plot for a consolidation test that corresponds to a time after primary consolidation is completed. 

For soils that are expected to demonstrate significant secondary settlements, it is important that load 

durations extend to times after primary consolidation is completed. Figure 52 shows a plot of 

deformation (for a particular load increment) versus logarithm of time for a soil sample. The 

coefficient of secondary compression, Cα, is evaluated according to the following equation: 

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1

2 log t

t

e C ∆ α = (Equation 40) 

where ∆e is the change in void ratio over an elapsed time equal to t2-t1. The times t1 and t2 occur 

after the time to the end of primary consolidation, tp. Alternatively, the modified coefficient of 

secondary compression, Cαε, can be evaluated similarly using a plot of volumetric strain versus 

logarithm of time. 

2.25

2.3

2.35

2.4

2.45

2.5

2.55

2.6

2.65

0.1 1 10 100 1000 10000

Elapsed Time (min)

Void Ratio (e)

tp

One Log 

Cycle

∆e=Cα=0.06

Stress Range

40 - 80 kPa

Figure 52. Evaluation of Cα. 

Values of Cαε (or Cα) are affected by stress history in that maximum values for Cαε are evaluated for 

stress levels greater than the preconsolidation stress (i.e., at stresses corresponding to virgin 

compression). Therefore, to assess a value of Cαε (or Cα) to be used for design analyses, the final 

effective stress in the ground after primary consolidation is completed should be evaluated. If the 

final effective stress is less than approximately 0.8σp′, then an average value of Cαε (or Cα) evaluated 

in the overconsolidated range may be used for design. If final effective stresses in the ground exceed 

σp′, then it is conservative to select Cαε (or Cα) value corresponding to stresses in the range of 1 to 2 

times σp′. It is noted that for samples that are disturbed, laboratory-measured values of Cα (or Cαε) 

may be underestimated in the normally consolidated range as compared to results from high-quality 

undisturbed samples. Therefore, the effects of disturbance on measured secondary compression 

values need to be carefully assessed, especially for soils that may undergo significant secondary 

compression such as organic clays and peats. 

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5.4.3 Evaluation of σp′ from In-situ Test Methods

The use of in-situ tests for profiling σp' in clayey soils is attractive since it may be possible to discern 

a rather complicated and varied stress history that includes multiple effects (e.g., erosion, reloading, 

aging, plus cementation). Also, in-situ tests are conducted quickly and inexpensively, thus allowing 

an immediate assessment of the state of overconsolidation and its variation across a site. However, 

in-situ data provide only indirect measures of preconsolidation and therefore should be used to 

supplement values obtained from laboratory consolidation tests. Calibration of in-situ test 

correlations is prudent and accomplished by benchmarking the data against values provided from 

consolidation, triaxial, and index tests run on undisturbed samples, as well as the stress history based 

on an understanding of the local geologic setting. Site-specific correlations between in-situ tests and 

laboratory tests, however, may be cost-prohibitive on small projects. 

Analytical derivations are available to evaluate the OCR of cohesive soils using the cone 

penetrometer, piezocone, flat dilatometer, and pressuremeter. These methods, via a correlation, 

relate OCR to a specific measured parameter in the in-situ test. These correlations are presented 

elsewhere (see Kulhawy & Mayne, 1990). Herein, simple empirical and statistical expressions for 

estimating σp' from in-situ test results are presented. These correlations can be used to provide firstorder estimates of σp' and to complement preconsolidation stress values evaluated from laboratory 

oedometer tests. Preliminary estimates of σp' for intact (i.e., not fissured) natural clays may be made 

using the approximate generalized trends as follows for each of the in-situ tests: 

Cone Penetration Test (figure 53): σp' = 0.33 (qT - σvo) (Equation 41) 

Type 1 Piezocone (face element) (figure 54): σp' = 0.47 (u1 - u0) (Equation 42) 

Type 2 Piezocone (shoulder element) (figure 55): σp' = 0.54 (u2 - u0) (Equation 43) 

Flat (Plate) Dilatometer Test (figure 56): σp' = 0.51 (p0 - uo) (Equation 44) 

Self Boring Pressuremeter Test (figure 57): σp' = 0.45 pL (Equation 45) 

Field Vane Test (figure 58): σp' = 3.54 (su, VST) (Equation 46) 

Note that stress measures in these figures are normalized to atmospheric pressure, Pa. 

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Figure 53. Correlation of σp′ with CPT qt data (after Kulhawy and Mayne, 1990). 

Figure 54. Correlation of σp′ with CPTu u1 data (after Kulhawy and Mayne, 1990). 

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Figure 55. Correlation of σp′ with CPTu u2 data (after Kulhawy and Mayne, 1990). 

Figure 56. Correlation of σp′ with DMT po data (after Kulhawy and Mayne, 1990). 

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Figure 57. Correlation of σp′ with self-boring PMT pL data (after Kulhawy and Mayne, 1990). 

Figure 58. Correlation of σp′ with VST su, VST data (after Kulhawy and Mayne, 1990). 

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Based on data shown in figures 53 through 58, it can be concluded that values of σp′ from these 

empirical correlations compare favorably to values obtained from laboratory testing, except as noted 

below. 

• The piezocone with a shoulder element (i.e., CPTu2) will measure negative penetration pore 

water pressures in fissured clays and dense silts indicating that the above correlation for σp′

is not valid for such soils. 

• Each correlation tends to underpredict values of σp′ for stiff fissured clays. The use of these 

correlations for stiff fissured clays is therefore conservative. 

Although not shown, correlations have been developed for correlating SPT N values to σp'. This 

correlation is poor and therefore SPT results should not be used, even as a first-order estimator, to 

evaluate σp' for soft to medium clays. 

5.4.4 Evaluation of ch from CPTu Dissipation Data 

In this section, procedures are presented to evaluate the coefficient of lateral consolidation, ch, from 

piezocone dissipation test results. To calculate ch using a piezocone, the modified time factor, T* 

must be evaluated. The parameter T* is related to the degree of consolidation. Values of T* are 

shown in table 27 for various stages of dissipation. The degree of consolidation, U, is defined with 

relation to the dissipation of excess penetration porewater pressure as: 

oi

ot

uu

uu U −− 1−= (Equation 47) 

where: ut = pore pressure at time t; 

 ui = initial pore pressure at t=0; and 

uo = hydrostatic pore pressure at equilibrium (i.e., static conditions based on elevation of 

ground water table). 

Based on this equation, when ut = ui, the degree of consolidation is zero whereas if ut = uo, the 

degree of consolidation is 100 percent. 

Normalized excess pore pressures (defined as 1-U) are based on the dissipation from the initial 

excess pore pressure not the maximum excess pore pressure measured during a dissipation test. This 

is specifically noted since, in some cases, the maximum excess pore pressure does not correspond to 

the initial excess pore pressure. Figure 59 shows pore pressure dissipation test results at two depths 

within a clay deposit. 

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Table 27. Modified time factors, T*, for analysis of CPTu dissipation data 

(after Teh and Houlsby, 1991). 

Degree of 

Consolidation, U 

(%) 

Normalized Excess 

Pore Pressure, 

(1 – U) 

T* for Piezocone with 

Element Located MidFace, u1

T* for Piezocone with 

Element Located 

Behind the Tip, u2

20 0.8 0.014 0.038 

30 0.7 0.032 0.078 

40 0.6 0.063 0.142 

50 0.5 0.118 0.245 

60 0.4 0.226 0.439 

70 0.3 0.463 0.804 

80 0.2 1.04 1.60 

0

100

200

300

400

500

600

0.01 0.1 1 10 100

Time (min)

Porewater Pressure Behind Tip (u2)

Depth = 6.68 m

Depth = 17.65 m

Figure 59. CPTu2 pore pressure dissipation curves. 

Values for ch can be estimated for various degrees of consolidation; however evaluations for design 

analyses should be based on a degree of consolidation of 50 percent or greater. The parameter ch is 

calculated according to the following: 

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t

IaT

c r

h

2 * = (Equation 48) 

where a is the radius of the cone (equal to 1⁄2 D, see figure 15), Ir is the undrained rigidity index 

(=G/su where G is the soil shear modulus and su is the undrained strength), and t is the time at which 

the pore pressure measurement is taken. The parameter Ir can be evaluated directly for use in 

equation 48 from figure 60. Calculations are provided in Soil Property Selection Example No. 1 in 

appendix A that demonstrates the application of equation 48. An alternate method to evaluate ch

based on piezocone dissipation test results is provided in appendix C. 

0

50

100

150

200

250

300

1 10

Overconsolidation Ratio, OCR

Undrained Rigidity Index

PI=10

PI=20

PI=30

PI=40

PI=50

Figure 60. Rigidity index (after Keaveny and Mitchell, 1986). 

5.4.5 Selection of Design Values for Consolidation Analyses 

General 

With respect to consolidation parameters and properties, the design engineer must consider whether 

the magnitude of settlement must be controlled to strict tolerances to ensure good performance of the 

structure or whether settlements must only be controlled to a more nominal tolerance. For example, 

settlements of embankments are typically not very critical due to the flexibility of the embankment; 

however, the settlement of an approach embankment needs to be very accurately assessed in an 

attempt to “match” the settlement of the bridge abutment. For mechanically stabilized earth (MSE) 

structures, allowable differential settlement along the wall alignment can vary between allowing 

significant differential settlement for geotextile-reinforced slopes to more stringent requirements for 

walls with full-height precast concrete facings. Such relative tolerances will ultimately guide the 

selection of settlement parameters. Where tolerances are very strict, conservative parameters will be 

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selected and site-specific variations will be assessed; where tolerances are more relaxed, average 

parameters will likely be selected for design. Ultimately the design engineer must ask the question, 

“what is the implication on the performance of the structure (for both limit states and serviceability 

states) if the selected design properties are significantly different than actual properties?” 

Compression Parameters 

Compressibility properties including Cc (or Ccε), Cr (or Crε), and Cα (or Cαε) should be evaluated 

from one-dimensional consolidation tests. As discussed, computed values will depend on degree of 

sample disturbance, stress level at which the parameter is computed relative to the preconsolidation 

stress, and inherent variability. For settlement calculations, the range of calculated values and 

average values for these parameters for each compressible soil unit should be evaluated and plotted 

versus depth. Settlement analyses can be carried out extremely rapidly using spreadsheets or 

settlement analysis programs (e.g., EMBANK (FHWA, 1993)) so that the effect of parameter values 

at the upper and lower end of the range can be evaluated. 

Preconsolidation Stress 

Both laboratory and in-situ tests can be used to evaluate preconsolidation stress, recognizing that 

laboratory test data are required to correlate values derived from in-situ tests. As an example, figure 

61 shows a comparison of calculated values of σp′ from laboratory oedometer curves and from the 

correlation described previously for the DMT (i.e., equation 44). Results compare favorably and 

demonstrate the utility of in-situ testing for profiling preconsolidation stress. For the particular 

project example illustrated in figure 61, oedometer tests were performed at approximately 1.5 to 3-m 

intervals. For more typical transportation-related projects, oedometer tests will likely be performed 

at greater intervals based on project schedule and cost. For this project example, if the number of 

oedometer tests were reduced to be more consistent with that for a typical project, then the 

interpreted DMT data would accurately fill in data gaps between oedometer test results. Once again, 

however, high-quality oedometer tests must be performed to establish values of preconsolidation 

stress for which in-situ test results can be compared. 

A profile of σp′ (or OCR = σp′/σvo′) with depth should be developed for the site. This parameter 

represents the most important value relative to settlement and shear strength evaluations for designs 

involving cohesive deposits. As with compressibility properties, an upper and lower bound profile 

should be developed based on laboratory tests and plotted with a profile based on particular in-situ 

test(s) (if used). There are no specific rules to judge the appropriateness of any particular value, 

however sample disturbance will typically be the most important factor affecting results. The 

sample disturbance indices will be useful in assessing why, for example, a particular test provides a 

value of preconsolidation stress that seems too low. 

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Figure 61. Preconsolidation stress from oedometer and DMT. 

Specific data quality issues that should be considered in evaluating preconsolidation stress are 

described below: 

• Assess whether a consolidation curve that does not exhibit a sharp break is the result of 

sample disturbance or that the soil sample may contain relatively significant amounts of 

non-cohesive material (i.e., sand, non-plastic silt), which will result in a consolidation curve 

without a distinct break. 

• Make sure that the consolidation curve used to compute the preconsolidation stress 

corresponds to end of primary consolidation conditions. Most laboratories do not perform 

analyses to evaluate the end of primary condition, but simply provide consolidation curves 

that represent the end of the test (see figure 48 for comparison of σp′ for laboratory tests 

after 24 hours and for tests carried out to the end of primary consolidation). 

• The preconsolidation stress can only be reasonably estimated from laboratory oedometer 

tests if the test has been carried out to sufficiently high pressures. As a general rule of 

thumb, tests should be carried out to loads equal to approximately 4 to 8 times the estimated 

σp′. Previous testing and/or geologic information should be used to estimate σp′. 

It is particularly important to accurately compute preconsolidation stress values for relatively 

shallow depths where in-situ effective stresses are low. An underestimate of the preconsolidation 

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stress at shallow depths will result in overly conservative estimates of settlement for shallow soil 

layers. 

Coefficient of Consolidation 

Laboratory consolidation test data are used to estimate the vertical coefficient of consolidation, cv

whereas a piezocone can be used with pore pressure dissipation test data to estimate the horizontal 

coefficient of consolidation, ch. The general trend associated with coefficient of consolidation is that 

the cv values are stress dependent and that higher values are to be expected for effective stresses less 

than the preconsolidation stress. Due to the numerous simplifying assumptions associated with 

conventional consolidation theory, for which the coefficient of consolidation is based, it is unlikely 

that even the best estimates of cv or ch from high-quality laboratory tests will result in predictions of 

time rate of settlement in the field that are significantly better than a prediction within one order of 

magnitude. In general, the in-situ value of cv is larger than the value measured in the laboratory test. 

Therefore, a rational approach is to select average, upper, and lower bound values for the appropriate 

stress range of concern for the design application. These values should be compared to values 

obtained from previous work performed in the same soil deposit. Under the best-case conditions, 

these values should be compared to values computed from measurements of excess pore pressures 

during construction of other structures. Figure 62 provides a general correlation for cv that can be 

used to judge the reasonableness of computed cv values obtained from oedometer test data. 

Calculations performed concerning time rate of settlement for the actual field conditions should be 

performed using the average, upper, and lower bound values discussed above. The results of these 

calculations should be compared relative to specific project constraints. For example, a range of cv

values may result in calculated times to the end of primary consolidation for field conditions of 6 to 

60 days. This would indicate that even under worst-case conditions, it is likely that consolidation 

will be completed during construction. Where calculated cv or ch values result in a range of time 

rates of settlement which encompass “go, no-go” decisions, then the installation of piezometers 

should be considered. 

Values for cv obtained from laboratory oedometer tests should be compared to values for ch obtained 

from piezocone tests. Although the time rate of consolidation in the lateral direction should be 

greater than that in the vertical direction, for most relatively isotropic soils the ratio of lateral to 

vertical consolidation may only be as high as two. Soils that are highly stratified such as varved 

clays may have a ratio on the order of ten or more. 

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Figure 62. Correlation of cv to LL. 

Coefficient of Secondary Compression 

The analysis of secondary compression settlements includes evaluating time-dependent (creep) 

settlements for a given period of time. In engineering analyses, it is commonly assumed that the 

coefficient of secondary compression can be represented as a constant value over the period of 

interest implying that the rate of secondary compression is independent of the thickness of the 

compressible layer. The accurate interpretation of Cα and Cαε is directly linked to the ability to 

accurately evaluate the time to the end of primary consolidation in a laboratory consolidation test. 

As discussed, the coefficient of secondary compression is expected to vary with final effective stress 

level and an average value can be developed based on the recommendations provided in section 

5.4.2.7. Computed values from consolidation tests should be compared to computed values for Cc

and Cr and the ratios Cα/Cc and Cα/Cr should be compared to the expected values listed in section 

5.4.2.7 (e.g., Cα/Cc = 0.05 ±0.01 for organic silts and clays). Additional discussion on the evaluation 

of secondary compression properties for organic clays, organic silts, and peats is provided in chapter 

7. 

5.5 GENERAL STRESS-STRAIN AND STIFFNESS PROPERTIES 

5.5.1 Background 

There are numerous semi-empirical methods available for calculating settlements in soils. Some of 

these methods apply elasticity theory to calculate settlements based on a selected elastic modulus 

value. The difficulties associated with selecting a single modulus value to characterize the soil have 

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resulted in very conservative assessments of settlements in soils resulting from foundation loadings. 

This section describes a method that uses the shear wave velocity of the soil as a means to evaluate 

an appropriate modulus value for use in evaluating settlements in soils subject to vertical (or axial) 

loadings. Although not presented here, the method can be extended to evaluate primary 

consolidation settlements in cohesive soils. 

For completeness, several historical correlations for elastic modulus based on soil type and SPT N 

values are presented first. These values can be used in preliminary calculations for foundation 

settlements. However, it is strongly recommended that where “go, no-go” decisions regarding the 

use of shallow foundations as compared to deep foundations need to be made, that the method 

presented herein for evaluating elastic modulus values based on soil shear wave velocity be 

considered. 

5.5.2 Settlement Analysis for Soils 

Settlements are often calculated based on results from in-situ tests and used either in empirical 

relationships or using equations from elasticity theory. The empirical expressions are particularly 

dependent on site-specific geologies, local practices, and specific test method (e.g., SPT, CPT, PMT, 

DMT). A review of the traditional methods used to evaluate settlements in sands is provided in 

FHWA/RD-86/185 (1987). Several methods use an elastic modulus value and a general elasticity 

equation to calculate settlements in granular soils and immediate settlements in clayey soils. 

The general equation for displacement at the center of an applied surface loading is of the form: 

)1(

2 −⋅= ν

Es

qBI s (Equation 49) 

where q = applied surface stress (∆σv at z = 0), B = width of the loaded area, I = displacement 

influence factor, Es = equivalent elastic modulus, and ν = Poisson's ratio of the soil. The influence 

factors (I) depend on surface area geometry, layer thickness, degree of homogeneity, and relative 

rigidity of the loaded area. Values of I are given in table and chart forms for foundations (e.g., Harr, 

1966; Poulos & Davis, 1974) and embankments (e.g., Giroud, 1968), or else can be generated on a 

simple spreadsheet using stress distributions (e.g., Mayne & Poulos, 1999). Equations are also 

available for evaluating settlement of a deep foundation element (i.e., a driven pile or drilled shaft) 

subject to axial compression loading. 

Table 28 and 29 provide correlations for Es based on soil type and from uncorrected SPT N values. 

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Table 28. Elastic constants of various soils based on soil type (modified after AASHTO, 1996). 

Soil Type Range of Equivalent Elastic 

Modulus (kPa) 

Clay 

• Soft sensitive 

• Medium stiff 

• Very stiff 

• 2,500 to 15,000 

• 15,000 to 50,000 

• 50,000 to 100,000 

Loess • 15,000 to 60,000 

Silt • 2,000 to 20,000 

Fine sand 

• Loose 

• Medium dense 

• Dense 

• 8,000 to 12,000 

• 12,000 to 20,000 

• 20,000 to 30,000 

Sand 

• Loose 

• Medium dense 

• Dense 

• 10,000 to 30,000 

• 30,000 to 50,000 

• 50,000 to 80,000 

Gravel 

• Loose 

• Medium dense 

• Dense 

• 30,000 to 80,000 

• 80,000 to 100,000 

• 100,000 to 200,000 

Table 29. Elastic constants of various soils based on SPT N value (modified after AASHTO, 1996). 

Soil Type Equivalent Elastic Modulus (kPa)

Silts, sandy silts, slightly cohesive 

mixtures 

400 (N1)60

Clean fine to medium sands and 

slightly silty sands 

700 (N1)60

Coarse sands and sands with little 

gravel 

1,000 (N1)60

Sandy gravels 1,200 (N1)60

5.5.3 Method to Evaluate Equivalent Elastic Modulus 

For deformation analyses, correlations such as that presented above for the SPT have been 

developed to obtain a modulus value from penetration test data (e.g, Schmertmann, 1970; Mitchell & 

Gardner, 1975). As shown in figure 63, however, a relevant soil modulus for a deformation analysis 

is situated close to the initial stress state of the material, while the penetration readings correspond to 

the peak (i.e., fully mobilized strength) of the stress-strain-strength curve. 

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Figure 63. Strength measured by in-situ tests at peak of stress-strain curve. 

Although the strength of a soil is in many cases proportional to its stiffness characteristics, different 

materials can have similar shear strength but have significant differences in stiffnesses (see figure 

64). 

Figure 64. Stress-strain-strength curves for three geomaterials having the 

 same strength yet different stiffnesses. 

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In certain geologic materials, it has been possible to develop calibrated correlations between specific 

tests (e.g., PMT, DMT) with performance-monitored data obtained from full-scale structures, 

including foundations and embankments, or with reference values from laboratory tests. Generally, 

these tests provide a modulus somewhere along the stress-strain curve, yet the specific position 

along the stress-strain curve is not well-defined, as indicated by figure 65. 

Figure 65. Variation of modulus with strain level. 

Of particular note, the small-strain modulus from geophysical tests provides an excellent reference 

value, as this is the maximum stiffness that the soil can exhibit at a given void ratio and confining 

stress. Herein, a generalized approach based on the small strain stiffness from shear wave 

measurements will be discussed, whereby the initial modulus (E0) is reduced to a value consistent 

with an appropriate working stress level for the desired factor of safety (FOS) for the structure. This 

reduction factor is also referred to as the modulus degradation value, E/Eo. 

The equivalent elastic modulus value, Es, can be calculated as: 

o

o

s E

E

E E 















= (Equation 50) 

where Eo is the small-strain elastic Young’s modulus. This modulus is related to the small-strain 

shear modulus, Go, according to: 

GE oo += ν )1(2 (Equation 51) 

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Recent research has confirmed that Poisson's ratios for many soils are lower than once thought. For 

drained loading, appropriate values are 0.1 < ν < 0.2 for stages of loading up to typical working load 

levels (e.g. Jamiolkowski, et al. 1994). 

The small strain (initial) shear modulus of soil, Go, is related to the shear wave velocity, Vs, and the 

mass density of the soil, ρ, by the equation: 

2

o = ρ VG s (Equation 52) 

Mass density of the soil is related to total unit weight of the soil, γt, by the acceleration of gravity, g: 

g

t γ

ρ = (Equation 53) 

The mass density of soils can be reasonably estimated from soil classification and location relative to 

the water table. Combining equations 50 to 52 results in the following expression for the equivalent 

elastic modulus: 

)1(2 2 +νρ 















= s

o

s V

E

E E (Equation 54) 

The required parameters necessary to evaluate the equivalent elastic modulus (Es), therefore include: 

(1) Poisson’s ratio; (2) shear wave velocity, Vs; and (3) modulus degradation value, E/Eo. A value of 

0.1 for Poisson’s ratio may be assumed for drained loadings in granular materials at working load 

levels. In subsequent sections, field procedures are described for evaluating shear wave velocity. 

Correlations are also provided which relate small strain shear modulus, Go, to measured parameters 

from in-situ tests. These correlations may be used if shear wave velocity information is not available 

for a particular project. Finally, a chart solution is provided for evaluating the modulus degradation 

value. 

5.5.4 Evaluation of Shear Wave Velocity 

5.5.4.1 General 

Measurements of shear wave velocity, Vs, can provide a reliable means for evaluating the small 

strain shear modulus (Go). The parameter Go is applicable to static monotonic loading and dynamic 

loading and for drained and undrained behavior. Thus, the value of Go can be established as a 

benchmark of stiffness for deformation analysis of all geotechnical problems. In addition, the small 

strain characteristics may be most appropriate for seismic design considerations. 

5.5.4.2 Field Measurements of Shear Wave Velocity 

The measurement of Vs (and G0, see Equation 52) in soils can be accomplished by a variety of 

laboratory and field methods, as depicted in figure 66. Provided that high-quality undisturbed 

samples can be obtained during field drilling explorations, laboratory tests can be conducted on 

specimens of natural soils. Current practice, however, relies more often on field measurement of Vs

and Go. 

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Impulse

(Vs)vh

waves

Downhole

Vertically

Propagating /

Horizontally

Polarized

Oscilloscope

Source Plank

Coupled to

Ground with

Static Load

Seismic

Cone or

Dilatometer

Geophone

(Vs)hv

Vs

Crosshole

Horizontally

Propagating /

Vertically Polarized

Cased Boreholes

P- and S-Wave

Logging

Drilling and

Sampling

Undisturbed

Tube

Sample

LABORATORY

FIELD

σvo'

σ ho'

θ

L

r

F ixe d Ba s e

L

r

s

θ γ =

τ

Torsional

Shear

Resonant

Column

Bender

Elements σvo'

σho'

τ

τ

σvo'

σ ho V ' s

σvo'

σ ho'

Triaxial, internal

local strain

Impulse

Hammer

Seismic

Refraction

Geophone

Seismic

Refraction

Striking

Plates &

Impulse

Figure 66. Field and laboratory methods to evaluate shear wave velocity. 

Geophysical techniques for subsurface exploration are described in detail by Woods (1994). 

Geophysical techniques commonly used in geotechnical practice are briefly summarized in the 

following paragraphs. Two general types of techniques are available to measure shear wave velocity 

in the field: 

• intrusive techniques whereby measurements are made using probes and sensors that are 

lowered in boreholes or pushed into the ground; and 

• non-intrusive techniques whereby the measurements are made using sources and receivers at 

the ground surface. 

Borehole Seismic Surveys 

In a borehole seismic survey, one or more boreholes are drilled into the soil to the desired depth of 

exploration. Wave sources and/or receivers are then lowered into the boreholes to perform the 

desired tests. There are three approaches to borehole seismic surveys: 

• Cross Hole Survey: In a cross hole survey, the energy source is located in one boring and 

the detector (or detectors) is placed at the same depth as the energy source in one or more 

surrounding boreholes at a known spacing. Travel time between source and receiver is 

measured to determine the wave velocity. 

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• Down Hole Surveys: In a down hole survey, the energy source is located on the surface and 

the detector, or geophone, is placed in the borehole. The travel time is measured with the 

geophone placed at progressively increasing depth to evaluate the wave velocity profile. 

• Up Hole Surveys: Geophones are laid out on the surface in an array around the borehole. 

The energy source is set off within the borehole at successively decreasing depths starting at 

the bottom of the hole. The travel times from the source to the surface are analyzed to 

evaluate wave velocity versus depth. The energy source is usually either explosives or a 

mechanical pulse instrument composed of a stationary part and a hammer held against the 

side of the borehole by a pneumatic or hydraulic bladder. 

The cross hole technique is generally the preferred technique for a borehole survey as it offers the 

highest resolution and greatest accuracy. However, cross hole measurements require a very precise 

evaluation of the distance between the energy source and the detector, as well as multiple vertical 

holes. An inclinometer reading is generally performed in the boreholes used in a cross hole survey 

to correct the results for deviation of the boreholes from verticality. Cross hole geophysical testing 

is codified in ASTM D 4428. 

It has become convenient and economical to incorporate geophones within penetrometers and probes 

used for in-situ soil tests, and thus permit downhole testing in conjunction with routine in-situ testing 

soundings. The commercially-available seismic cone (Robertson, et al. 1986; Campanella, 1994) 

and seismic dilatometer (Martin and Mayne, 1998) are hybrid tests that combine penetration testing 

with downhole geophysics, thus allowing the determination of both small-strain and high-strain 

properties within a single sounding. Figure 67 shows a schematic illustration of a seismic 

dilatometer test. 

Seismic Refraction and Seismic Reflection Methods 

Seismic refraction and reflection exploration surveys are conducted from the surface and do not 

require boreholes. The resolution of the methods is relatively poor and decreases with depth. These 

methods are most suitable as a means of identifying the depth to competent rock and the location of 

prominent soil horizons that have a large contrast in density and stiffness compared to the overlying 

soil. Additional discussion is not warranted because these technologies generally cannot provide 

sufficiently accurate deformation parameters. 

Spectral Analysis of Surface Waves (SASW) 

Spectral Analysis of Surface Waves (SASW) is a non-intrusive geophysical technique used primarily 

for evaluating subsurface shear wave velocity profiles. SASW testing evaluates shear wave velocity 

indirectly by direct measurement of Rayleigh, or surface wave, velocity. Rayleigh wave velocity is 

closely related in magnitude to the shear wave velocity. SASW results are representative of the 

average properties of an assumed relatively large mass of material. SASW can be a very costeffective method of investigation. Ease and speed of field measurements and automated algorithms 

for data processing and inversion allow for evaluation of subsurface conditions at a relatively large 

number of points at a fraction of the cost of conventional intrusive exploration techniques. 

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Oscilloscope

DMT

Control

Panel

Source

Geophone

DMT

Tubing

Drill Rods

DMT Blade

N2

Gas

Receiver

Geophone

Oriented

Parallel to

Plank

Impulse by

Sledge Hammer

Plank with

Static Load

SH waves

Receiver

Geophone Cable

Receiver Geophone

Figure 67. Seismic dilatometer test. 

5.5.5 Correlations for Small-Strain Shear Modulus 

In addition to using shear wave velocity information, the small-strain shear modulus, Go, can be 

estimated using empirical correlations. Table 30 presents the typical range for Go for several generic 

soil types. 

Table 30. Typical values of small-strain shear modulus. 

Soil Type Small-strain shear modulus, Go (kPa) 

Soft clays 2,750 to 13,750 

Firm clays 6,900 to 34,500 

Silty sands 27,600 138,000 

Dense sands and gravels 69,000 to 345,000 

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The small-strain shear modulus can be correlated to the SPT N60 value and to the CPT qc value 

according to the following: 

68.0

60 Go = N )(560,15 (Equation 55) 

375.0'25.0 Go = qc σ vo )()(634,1 (Equation 56) 

Units of kPa are used in equations 55 and 56. Additional correlations for small-strain shear modulus 

are provided in FHWA-SA-97-076 (1997). 

5.5.6 Evaluation of Modulus Degradation Value 

The small-strain shear modulus (and hence the small-strain Young’s modulus) should be reduced for 

use in foundation deformation calculations since strains associated with foundation loadings are a 

few orders of magnitude greater than those corresponding to the small-strain values for Go and Eo. A 

graphical procedure can be used as a simple means to reduce the small-strain stiffness (E0) to those 

at working stress levels. To use this graphical procedure, the stress ratio, as defined as q/qult, at the 

desired working stress (q) is required. By definition q/qult = 1/FOS. Figure 68 illustrates the 

suggested trends for intact (i.e., non-fissured) clays and uncemented sands. For monotonic loading 

of these soils, a value of the dimensionless parameter g = 0.3 can be used. The parameter g is simply 

a curve-fitting parameter for the hyperbolic relationship shown in figure 68. The value of 0.3 has 

been confirmed as reasonable for first-order approximations based on results from laboratory shear 

tests as well as backcalculations from full-scale foundation load tests (Burns and Mayne, 1996; 

Kates, 1996). It is noted that the use of g=0.3 for highly structured soils and cemented materials is 

conservative (i.e., the calculated modulus value using g=0.3 for these soils is likely to result in a 

conservative estimate of settlements). 

0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

0.8

0.9

1

0 0.2 0.4 0.6 0.8 1

q/qult (=1/FOS)

E/Eo

E/Eo = 1 - (q/qult)

0.3

Figure 68. Modulus degradation based on g= 0.3. 

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5.5.7 Summary 

A method for evaluating an equivalent elastic modulus has been presented. The modulus value can 

be used in settlement calculations of structures in soils for which the settlement analysis method 

used requires an elastic modulus. The method uses the soil shear wave velocity and small-strain 

modulus value (i.e., Eo or Go) in the evaluation of the modulus. The evaluation of the equivalent 

elastic modulus is summarized below and is illustrated in Soil Property Selection Example No. 2 in 

appendix A. 

• Develop a profile of shear wave velocity versus depth at the location of the structure. Shear 

wave velocity should be evaluated using geophysical or in-situ testing techniques. Figure 69 

shows a shear velocity profile obtained using a commercially available seismic piezocone. 

• Develop a profile of small strain shear modulus, Go, using the shear wave velocity profile 

and unit weight measurements or from correlations. 

• Divide the profile into distinct zones based on the shear modulus profile or develop a single 

average value over each distinct zone within the depth of interest. 

• Based on the design FOS for the structure, evaluate E/Eo using figure 68 and the modified 

hyperbola with a “g” exponent equal to 0.3. 

• Calculate the equivalent elastic modulus, Es, using equation 54 and use this value in 

settlement analyses such as that of the form given in equation 49. 

5.6 SHEAR STRENGTH PROPERTIES OF SOIL 

5.6.1 Introduction 

As part of the design analyses for most geotechnical projects, the load-carrying capacity of the 

supporting soil is evaluated. These capacity evaluations include, for example, bearing capacity of a 

shallow foundation, side and tip resistance of drilled shafts and driven piles, stability analyses of 

slopes and embankments and passive soil capacity for the toe of a retaining structure. These 

analyses are concerned with comparing actual loads imposed on the system to the limit (or failure) 

state of the supporting soil. This limit state of stress corresponds to the shear strength of the soil. 

For a given soil, shear strength is anything but a unique property. This concept is often overlooked, 

even by experienced geotechnical engineers. The soil strength to be used in design analyses must be 

qualified in relation to whether the appropriate strength is: drained or undrained, peak, fully softened

or residual, intact or remolded, static or cyclic, compression or extension; and other facets, such as 

direction of loading, rate of loading, and boundary conditions. As a consequence, soil strength is not 

a fundamental property, but instead, a specific behavioral response to a certain set of loading 

conditions. 

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Figure 69. Shear wave velocity profile from seismic cone sounding. 

Figure 70 provides an overview of the various laboratory tests used to measure soil strength showing 

the imposed stresses and loading conditions. Laboratory testing for most highway projects may 

include unconfined compression, triaxial shear, and direct shear. Other tests including direct simple 

shear, plane strain, torsional shear, hollow cylinder, cubical triaxial, and ring shear are also available, 

but are most often used on large projects (e.g., use of direct simple shear testing on the Central 

Artery Project in Boston) or in universities. 

A variety of in-situ testing devices are also available for evaluating drained soil strength in granular 

soils and undrained strength in cohesive soils through the use of engineering correlations and 

theoretical equations. Currently, however, no in-situ test provides a reliable evaluation of effective 

stress strength (c′ and φ′) of clays, however ongoing research with piezocones appears promising. 

Field penetration tests (SPT, CPT, DMT) can be readily used to evaluate the strength of sands (φ′). 

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σvo' - ∆σv'

σho'

Simple

Shear

Direct

Simple

Shear

Direct

Shear Box

Triaxial

Compression

(Loading)

Triaxial

Extension

(Unloading)

Triaxial

Compression

(Unloading)

Triaxial

Extension

(Loading)

Plane Strain

Compression

Plane Strain

Extension Cubical

Triaxial

Hollow

Cylinder

Torsional

Shear

Ring

Shear

Torsional

Shear

σvo'

σho'

θ

L

r

Fixed Base

θ

γ =

τ

σvo'

τ Specimen

σvo' + ∆σv

σho'

εz = 0

 σvo'

σho' + ∆σh'

εz = 0

σh1' + ∆σh1'

σvo' + ∆σv

σh2' + ∆σh2'

σvo' + ∆σv'

σho'

σvo'

σho'

- ∆σh

σvo'

σho'

+ ∆σh

γs

Shear

Stress

 τ

Normal

Stress

σvo'

γs

Shear

Stress

 τ

Normal

Stress

σvo'

γs h

δ

≈

Shear

Stress

 τ

Normal

Stress

σvo'

σvo' + ∆σvo'

σho' + ∆σho'

Fixed Base

τ ∆τ +

Figure 70. Various laboratory tests used to measure soil strength showing the imposed stresses and 

loading conditions. 

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This section provides state-of-the practice information on the evaluation of soil shear strength (i.e., 

φ′, c′, and su) for design analyses. Information on laboratory, in-situ, and correlative techniques to 

evaluate shear strength are presented. Also, fundamental information relating to shear strength 

evaluation including concepts of drained and undrained behavior and total and effective stress 

analyses are presented as they relate to shear strength evaluation. The numerous factors that affect 

the selection of shear strength values for design analyses are presented in terms of their relevance to 

specific design requirements for typical highway and transportation design and construction 

applications. A description of the process used to evaluate shear strength data from several test 

methods is provided at the end of this section for effective stress and total stress strength parameters 

for soils. 

5.6.2 Fundamental Concepts of Soil Shear Strength 

5.6.2.1 Drained versus Undrained Loading 

In geotechnical practice, it is most important to distinguish between "drained" and "undrained" 

strengths. These terms refer to the ability of the porewater in the soil to move between soil particles 

resulting in volume change, and the accompanying generation (or lack) of excess porewater 

pressures, ∆u. Soils can also exhibit any number of partially drained strengths, however, design 

analyses are typically performed using drained and undrained strengths, as these represent limits to 

the expected range of behavior. 

For a saturated soil subjected to undrained loading, no drainage of porewater from the void spaces 

can occur, and thus the soil undergoes no change in volume. During undrained loading, changes in 

total stress (∆σ) cause the development of either positive porewater pressures (∆u > 0) that will tend 

to decrease the effective stress in the soil or negative porewater pressures (∆u < 0) that will tend to 

increase the effective stress in the soil. 

The drained loading of a saturated soil means that the water in the void spaces is free to move so 

that no excess pore water pressures develop (∆u = 0). There is usually a change (i.e., increase or 

decrease) in void ratio and a corresponding change in volume. Again, water may be present, but is 

free to move either out of the soil mass (termed contractive soil behavior) or into the soil mass 

(termed dilatant soil behavior). Contractive behavior results in a decrease in volume (e.g., 

settlement) and dilative behavior causes an increase in volume (e.g., swelling). Clean sands have 

such a high permeability (e.g., k > 10-3 cm/s) that, under static loading, they are almost always 

drained. Sands, however, will behave in an undrained mode when subjected to rapid loading, such 

as that imposed by an earthquake whereby the entire deposit is engaged and water cannot drain. 

All clays exhibit drained behavior when the rate of loading is very slow, so slow that it does not 

interfere with the rate of water migration that is controlled by its low permeability (e.g., k < 10-6

cm/s). Drained behavior in clays should be considered in evaluating the long-term stability (and 

ground movements) of cut slopes. The same clay that behaves in a drained manner in these 

applications, however, may initially behave in an undrained mode in the short-term if the rate of 

loading is too fast to permit water inflow. The short-term stability of excavations and slopes 

constructed in soft to medium clays is represented by undrained loading conditions. 

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5.6.2.2 Drained Stress-Strain-Strength Behavior 

In this section, the stress-strain-strength behavior of soils is introduced for the simple case of drained 

loading. To illustrate this, a graph of measured shear stress (τ) versus shear strain (γs) from a direct 

simple shear test is used (figure 71). It is recognized that the direct simple shear test is not 

commonly performed, but it is most useful in introducing drained stress-strain strength behavior of 

soils. The more common direct shear (box) test is very similar to the direct simple shear and thus the 

same basic principles apply. The slope of the τ-γs curve is the shear modulus (G), as designated by 

point 1 in figure 71. The term modulus can have several definitions, including initial

(corresponding to small-strains), secant (always through the origin of the measured response or G = 

τ/γs), and tangent (local incremental slope or G = ∆τ/∆γs). An unload-reload cycle can also be 

introduced and this is defined by the tangent value. 

The maximum stress on the τ-γs curve is commonly interpreted as the peak shear strength (τmax), 

corresponding to point 2 in figure 71. Figure 71 shows two stress-strain curves, each one 

corresponding to different effective consolidation stresses (i.e., σv1′ and σv2′). For each specimen, 

the measured peak strength (τmax) is plotted versus effective consolidation stresses as shown in figure 

71. A linear fit is generally forced (minimum of two data sets) to provide the simplified straight line 

Mohr-Coulomb failure criterion: 

τmax = c′ + σ′ tan φ′ (Equation 57) 

where φ′ = effective stress friction angle and c′ = effective cohesion intercept. 

tan φ'NC (fully-softened)

+ δv,

dilative

soil

Shear Stress, τ

σv' = effective overburden

stress

Go = small-strain shear modulus

Gsecant = 

s γ

τ

Gtangent = 

s γ

τ

∆

∆

τmax

τmax

Shear Strain, γs

Post-peak

Strain-softening

peak

peak

σv1'

σv2'

Fully-softened

Strength

residual

residual

effective

consolidation

stress

1 4

3

2

Shear Stress, τ

Normal Stress, σv'

τmax

τmax

1

3

4

2

tan φ'r (residual strength)

tan φ'peak

c'

τ = c' + σ' tan φp'

τ = σ' tan φ'

τr = σ' tan φr'

τ =

Shear

Stress

− δv,

contractive

soil

τ

σv'

 σv1′ σv2′

Figure 71. Drained stress-strain behavior. 

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During drained shearing, the soil specimen will likely undergo a change in total volume. If the soil 

decreases in volume during shear, the response is termed contractive behavior. This response is 

indicative of loose sands and soft clays. If the soil increases in volume during shearing, a dilative (or 

dilatant) behavior is observed. This response is common in hard clays and dense sands. If no 

change in volume occurs during drained shear loading (∆V/V0 = 0), the corresponding stress state is 

called the critical state. The complete description of soil behavior by this arrangement is termed 

critical-state soil mechanics, which encompasses normally- and overconsolidated soils for 

undrained, semi-drained, and drained loading, for both contractive and dilatant behavior (Schofield 

and Wroth, 1968; Wood, 1990). However, in current U.S. practice, the concept of critical state soil 

mechanics is not consistently recognized, although this concept is an excellent representation of soil 

behavior. In current U.S design practice., drained and undrained are not specifically viewed as they 

relate to volume change conditions. Rather, the adoption of the simplified Mohr-Coulomb strength 

criterion is widespread and it is normally taken that “drained conditions” are analyzed using 

effective stress parameters (i.e., φ′, c′) and “undrained conditions” are analyzed by total stress 

methods (i.e., c=cu=su). For cohesive soils, the drained analysis corresponds to long-term conditions 

and undrained analyses to short-term conditions. For cohesionless soils (i.e., those with relatively 

high hydraulic conductivity values), drained analyses alone are performed. The Mohr-Coulomb 

strength parameters, φ′ and c′, are defined in figure 72. 

Figure 72. Mohr-Coulomb failure criteria. 

Referring to drained strength characteristics in figure 71, after the peak shear strength (τmax) is 

reached during drained loading, the shear stress reduces to a stable value termed the fully-softened 

strength, depicted as point 3 in figure 71. The fully softened strength is intermediate between the 

peak strength and the residual strength and there are no specific procedures to identify the fully 

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softened strength. Conceptually, the fully softened strength is close in value to the peak strength of 

the same soil in a normally-consolidated condition. This can be expressed as: 

τ = σ′ tan φNC′ (Equation 58) 

where φNC′ is the peak strength (or critical-state strength) of the normally-consolidated soil. Note 

that for long-term analyses, the effective cohesion intercept is a small value for the normally 

consolidated case and is therefore assumed to be zero (c′ = 0). 

For clays, if drained loading continues for very large shearing strains, the shear stress value drops 

even further to the residual strength, denoted τr and indicated by point 4. The residual strength is 

related to the mineralogical frictional characteristics of the soil in which the plate-like clay particles 

align themselves in a direction parallel to the shear plane that is developed at these very large strains. 

The residual strength can be represented by: 

τr = σ′ tan φr′ (Equation 59) 

In commercial practice, φr′ is obtained using 8 to 10 repeated cycles of shearing on the same 

specimen in a direct shear box using the same direction of shear and the same normal load. The 

more elaborate ring shear device is purposely suited for obtaining true residual values of φr′. The 

ring shear testing device is illustrated in figure 70 and additional information on the ring sheartesting device can be found in Terzaghi, Peck, and Mesri (1996). 

For design of geotechnical features constructed on or in clay soils that exhibit peak, fully softened, 

and residual shear strength, the design engineer must consider the level of deformations that may be 

expected within the soil mass to appropriately select the strength to be used for design calculations. 

As an example, consider the evaluation of soil pressures acting on an anchored wall constructed in 

overconsolidated clay. Since the portion of shear strength at peak resulting from the cohesion 

intercept (c′ in figure 71) tends to reduce relatively rapidly with increasing deformations beyond 

peak, soil deformations associated with flexible anchored walls may be sufficient to appreciably 

reduce this cohesion. Therefore, unless local experience indicates that a particular value of cohesion 

intercept can be reliably accounted for, zero cohesion should be used in the analyses of flexible 

anchored walls for long-term (drained) conditions. Since the anticipated deformations are relatively 

small, conservative drained shear strength for analysis of anchored walls is therefore the fully 

softened strength. 

Residual strengths should be used for geotechnical projects that are designed for a location in which 

there is evidence of an existing failure surface within the clay (e.g., a structural system used to 

stabilize an active landslide) or in cases where large deformations are anticipated. For these 

conditions, it can be assumed that sufficiently large deformations have occurred to reduce the 

strength to a residual value and therefore, the residual strength can be used for design analyses. 

5.6.2.3 Undrained Stress-Strain-Strength Behavior 

The undrained shear stress-shear strain curve is similar to that observed for drained loading, except 

that excess porewater pressures are also generated (∆u ≠ 0). During undrained shearing, a 

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contractive soil will exhibit positive pore pressures, while a dilative soil will show negative pore 

pressures. From the undrained τ-γs curve, the peak value of τmax is designated as the undrained shear 

strength (su or cu). 

Instability under undrained conditions develops mainly for a contractive soil where the soil attempts 

to mobilize frictional shearing resistance and which also causes the soil to contract under the 

prevailing confining stresses. This tendency to contract during shear is typical for normally to 

lightly overconsolidated soft to medium clay soils. Since this tendency cannot be realized, due to the 

clay soil permeability in relation to the rate of shearing, positive porewater pressures are generated in 

the soil which reduce the effective stress and hence the mobilized frictional shearing resistance. In 

such cases the short term undrained shearing resistance of the soil is less than would have been the 

case if drainage (contraction of the soil volume) could have occurred. The short-term condition is 

critical for temporary walls constructed in, and for embankments constructed on, normally to lightly 

overconsolidated clay soils. 

In clay soils subjected to unloading conditions that result from excavation to form a highway cut 

slope, the soil attempts to expand as it mobilizes frictional shearing resistance. This is resisted 

causing negative porewater pressure to be developed that increases the effective stress in the soil and 

hence increases the mobilized frictional shearing resistance. Thus, in overconsolidated clay subject 

to excavation, the short-term (undrained) strength and stability potentially exceeds that which would 

apply once drainage has occurred. For the examples cited, the engineer needs to be careful to assess 

both short-term and long-term strengths in their analyses. For dilative soils, the tendency to 

dissipate pore pressure will reduce the effective stress and thus the strength. 

5.6.2.4 Effective Stress Parameters 

Granular soils such as gravels, sand, and non-plastic silts have effective stress failure envelopes that 

pass through the origin indicating that c′ = 0 for these materials. In fact, only cemented sands and 

some overconsolidated clay appear to have a true c' value (i.e., c′ ≠ 0). The value of φ′ for sands 

depends on mineralogy and packing arrangement that is related to relative density and effective 

confining stress level (Bolton, 1986). Ranges of φ′ for clean quartzitic (silica) sands are typically 

30° ≤ φ′ ≤ 50°, whereas calcareous (corraline) sands may exhibit somewhat higher values. The value 

of φ′ generally is recorded at low confining stresses. Figure 73 shows typical ranges of friction angle 

for rockfills, gravels, and sands over a wide range of confining stresses and with initial porosities 

ranging from 0.17 to 0.48. The rockfill grades shown on figure 73 are summarized in table 31. To 

use this figure, the engineer must select the range of confining stresses that the granular soil will be 

subject to in the field. If this range is relatively large, the friction angle of the soil will vary over the 

range of confining stresses. A conservative single value can be selected based on the largest 

confining stresses, or a strength envelope can be developed (in particular for slope stability analyses) 

in which values of φ′ are selected to be representative of smaller confining stress increments. 

Alternatively, values of φ′ used for design analyses for cohesionless soils are typically based on 

correlation to a measured parameter value from a penetration test (i.e., N60 from SPT, qt from CPTu, 

or KD from DMT) as discussed in section 5.6.6. 

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Table 31. Unconfined compressive strength of particles for rockfill grades in figure 73. 

Rockfill Grade Particle unconfined 

compressive strength (MPa) 

A ≥220 

B 165 to 220 

C 125 to 165 

D 85 to 125 

E ≤85 

Figure 73. Typical ranges of friction angle for rockfills, gravels, and sands 

(Terzaghi, Peck, and Mesri, 1996). 

For clays, empirical correlations have been developed to relate φ′ to the plasticity characteristics of 

the soil. Figure 74 shows a slight trend of φ′ decreasing with increasing PI (Mesri and AbdelGhaffar, 1993), yet values can be ± 8° in variance. Considering the overall importance of φ′ in 

stability calculations, foundations, and landslide analyses, it is essential to directly assess φ′ by 

means of consolidated drained direct shear tests, consolidated drained triaxial tests, or consolidated 

undrained triaxial tests with porewater pressure measurements. The consequences of estimating φ′

can be economically unwise. As an example, for relatively long, shallow slip surfaces that may be 

associated with a landslide, the required forces that would need to be resisted by some form of 

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stabilization system (e.g., retaining wall, micropiles) would vary significantly depending on the 

drained friction angle of the soil. It is highly recommended that state DOTs develop historical data 

summaries of φ′ versus PI to check validity of future test results. 

20

20

40

40

50

0 60 80 100 200 400 600 1000 0

10

Soft Clays

Mexico City

Clay

Attapulgite

Soft and Stiff Clays

Shales

Clay Minerials

30

Fricttion Angle, (degrees)

Plasticity Index, I (%) p

’

Figure 74. Relationship between φ′ and PI (Terzaghi, Peck, and Mesri, 1996). 

The short-term value of effective cohesion intercept is related to the preconsolidation stress and 

current effective stress state, as shown by figure 75. However, for long-term analyses involving 

most insensitive clays, silts, and uncemented sands, it is best to adopt c′ = 0, unless extensive 

laboratory testing is conducted or sufficient experience exists to prove bonding or cementation. 

Conservative recommended values of effective cohesion intercept are as follows: 

Short Term: c′ = 0.024 σ p′

Long Term: c′ = 0 

The correlation shown in figure 76 (Stark and Eid, 1994) can be used to estimate residual friction 

angles for preliminary analyses that involve clayey soils that have been subject to large 

displacements. 

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Figure 75. Relationship between c′ and σp′ (Mesri and Abdel-Ghaffar, 1993). 

Figure 76. Residual friction angles for clayey soils (after Stark and Eid, 1994). 

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5.6.2.5 Total Stress Parameters 

Total stress analyses for soils that do not drain during the loading period involve the principle that if 

an element of soil in the laboratory is subjected to the same changes in stress under undrained 

conditions as an element of the same soil in the field, the same excess pore pressures will develop. 

Thus, if the total stresses in the laboratory and the field are the same, the effective stresses will also 

be the same. Because soil strength is governed by effective stresses, the strength measured in 

laboratory tests should be the same as the strength in the field when the pore pressures and total 

stresses are the same. Thus, under undrained conditions, strengths can be related to total stresses, 

making it unnecessary to specify undrained excess pore pressures for design analyses. 

Although the total stress principle is simple, experience has shown that many factors influence the 

pore pressures that develop under undrained loading. As a result, determining undrained strengths 

by means of laboratory and in-situ testing requires considerable attention to detail if reliable results 

are to be achieved. Shear strengths for use in undrained total stress analyses must be measured using 

test specimens and loading conditions that closely duplicate the conditions in the field. 

Alternatively, they can be reliably measured using the appropriate in-situ test. 

Total stress type analyses using the “φ = 0” approach are perhaps the most widely used form of 

analyses performed by highway engineers for design capacity analyses involving clays. For this 

condition, the Mohr-Coulomb relationship reduces to the form σ1′ = σ3′ + 2c or c=cu=su= 1⁄2 (σ1′-σ3′) 

= undrained shear strength. Since su is stress dependent, its value is commonly normalized by the 

vertical effective overburden stress (σvo′) at the depth where su is measured. The interpretation of su

from laboratory and in-situ tests is discussed subsequently in this chapter. 

5.6.3 Relevance of Design Applications to Soil Shear Strength Evaluation 

The selection of an appropriate value for soil shear strength to be used in the analysis of a particular 

geotechnical structure should consider, at a minimum, the following: (1) how fast will construction 

occur relative to the hydraulic conductivity of the soil (i.e., drained or undrained strengths); (2) how 

does the direction of applied load affect measured shear strengths and the appropriate strength to be 

used for an analysis; (3) how do the expected levels of deformation for the geotechnical structure 

affect the selection of shear strength; and (4) how does the manner in which the feature is 

constructed affect the shear strength to be used in analysis. Table 32 provides a summary of specific 

issues related to the design and construction of typical highway design elements that should be 

considered in developing and implementing a laboratory and in-situ testing program for evaluating 

soil shear strength. Issues related to the evaluation of shear strength for rock are described in chapter 

6. 

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Table 32. Summary of issues relevant to shear strength evaluation in support of the design of 

typical geotechnical features. 

Design 

Element Issues Relevant To Shear Strength Evaluation 

Shallow 

Foundation 

• Soil shear strength information required for depths up to 2 times the width of the footing, unless 

weak zones are found below this depth. The depth of bottom of footing will be based, in part, on 

requirements with respect to frost penetration depths and scour depths. 

Drilled Shaft • The excavation of a hole to construct a drilled shaft results in stress relief and disturbance in the soil 

that ultimately results in a reduction in shear strength from that corresponding to in-situ (i.e., before 

construction) conditions. The magnitude of the stress relief and disturbance will depend upon the 

method of construction, soil type, saturation condition, and type of strength (e.g., side shear or end 

bearing). 

Driven Pile • The shear strength of the soil may vary significantly between the time when the pile (or pile group) 

is first driven and tested to the time when the superstructure loads are applied to the pile (or pile 

group). The time-dependent phenomena of strength increase is referred to as “pile set-up” and is 

often observed for driven piles in saturated NC to moderately OC clay and fine-grained material. A 

decrease in strength with time is referred to as “relaxation” and is often observed for heavily OC 

clays, dense silts, dense fine sands, and weak laminated rock. Shear strengths should therefore be 

evaluated for both long and short term conditions. 

• Changes in site conditions that affect the in-situ the effective stress state may increase or decrease 

shear strength and pile capacity. These may include site dewatering or additional surface loading 

from an embankment. 

• An increase in granular soil strength may occur due to densification during driving. This increased 

strength will need to be considered such that an appropriate pile driving system can be selected for 

construction. 

Retaining 

Walls 

• The analysis of non-gravity cantilevered and anchored walls require an evaluation of earth pressures 

on the active side and passive side of the excavation. For undrained loadings in some clayey soils, 

particularly low to medium plasticity materials, there can be a large difference in undrained strength 

between the strength used for the active side and the passive side of the excavation. 

• For soils that may exhibit peak, fully softened, and residual conditions, an estimate of the tolerable 

deflection of the wall system needs to be made and this deflection used to select the appropriate 

strength condition for analysis. 

Slopes • The shear strength of discontinuities (e.g., fissures) in soil (and rock) needs to be evaluated since it 

may represent the critical (i.e., lowest) shear strength for design. 

• Weathering and other physiochemical reactions may occur at a quick enough rate to weaken soil 

bonds and reduce shear strength. 

• Strength loss may occur in cut slopes due to soil softening (in presence of water) and continuing 

deformations. Large deformation residual strengths should be used for long-term analyses. 

5.6.4 Laboratory Testing Methods for Evaluating Soil Shear Strength 

5.6.4.1 Selection of Laboratory Testing Method 

The laboratory shear strength testing method and testing parameters used should model the actual 

field problem being considered and provide a reliable assessment of shear strength. As previously 

noted, laboratory strength testing of natural granular soils is not typically performed due to sampling 

difficulties, however, the shear strength of compacted granular materials such as that used for 

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backfill in retaining structures and embankment fill is typically evaluated using the direct shear 

device. 

The well known unconfined compression (UC) and unconsolidated undrained (UU) tests are often 

used to evaluate su in cohesive materials. Although simple to perform, the UC and UU tests are 

particularly misleading because of difficulties caused by sampling disturbance, high strain rate, and 

uncertain drainage effects, and therefore they are not recommended as the sole method to evaluate 

undrained shear strength for design analyses, but can be used to complement other strength data. 

Hand torvane and pocket penetrometer tests should not be considered at all for the evaluation of 

undrained strength for design analyses. These tests provide an index of soil consistency as an aid in 

selecting the type of testing device for measuring undrained strength and should not be relied upon 

for engineering analyses. 

Ideally, laboratory specimens for shear strength testing should be reconsolidated in the laboratory 

such that at least the original effective stress states (σvo′ and σho′) in the ground are applied to the 

specimen to simulate in-situ conditions. In research laboratories, it is possible to consolidate the 

specimens under K0 conditions using automated triaxial devices. In most commercial laboratory 

triaxial systems, an isotropic stress state (σc′ = σvo′) is imposed because of simplicity in procedure. 

These isotropic triaxial tests are suitable in providing an appropriate effective stress friction angle 

(φ′) for most soil types, however, the undrained shear strengths from isotropically consolidated 

triaxial tests will likely be higher than those that should be used for design analyses of 

embankments, foundations, and retaining walls (e.g., Mayne, 1988). Procedures are available, 

however, to adjust the measured undrained strength in a triaxial compression test to values 

specifically applicable to particular field loading conditions. These procedures are discussed in this 

chapter. 

In this section, the interpretation of soil shear strength using triaxial, direct shear, and unconfined 

compression direct simple shear testing methods are presented. 

5.6.4.2 Triaxial Testing 

On transportation-related projects, triaxial shear strength testing is performed on undisturbed 

samples of cohesive soils. Triaxial testing can be generally classified as: (1) unconsolidatedundrained (UU); (2) consolidated-undrained (CU); and (3) consolidated-drained (CD). 

Unconsolidated Undrained Triaxial Test 

In the UU test, the total stress undrained shear strength of the soil is calculated based on the 

measured compressive strength of the soil. Shear strengths calculated from UU tests correspond to 

the depth at which the sample was taken from in the ground. Figure 77 shows the stress-strain and 

Mohr circle representation for four UU tests performed on the same soil. Theoretically, if each 

sample is 100 percent saturated and at the same moisture content, the shear strength (which is the 

radius of the Mohr circle) will be the same. This is because shear strength is directly related to the 

void ratio of the sample. If the void ratio of each sample is the same, then since volume change is 

not allowed during shearing, the measured shear strength from each UU test will be the same. In 

most cases, however, due to sampling difficulties and inevitable differences in moisture content, 

samples taken from approximately the same depth will not exhibit this response. 

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Figure 77. Stress-strain and Mohr circle representation for four UU tests performed on the same 

soil (after Day, 1999). 

UU test results should be interpreted using the φ=0 concept. As an example, consider figure 78 in 

which the measured shear strength was not the same for each specimen. This may be due to sample 

disturbance or other factors. However, some commercial laboratories will report a best-fit envelope 

through triaxial data, including UU data, such as that shown in figure 78. Such an interpretation is 

incorrect; for each test, the undrained strength (i.e., radius of each individual Mohr circle) should be 

evaluated. 

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Figure 78. Interpretation of UU test data. 

For soils that are partially saturated, such as compacted clays or naturally occurring clays above the 

water table, undrained strengths should be measured from UU tests performed on specimens with the 

same void ratio and degree of saturation as the soil in the field. Such soils will not exhibit a constant 

value for su. It is therefore important that tests be performed over a range of confining pressures that 

represent the range of stresses to be expected in the field. For each test, plot the confining pressure 

and corresponding undrained shear strength. This exercise will likely result in a nonlinear envelope 

of shear strength. Many slope stability programs enable the user to input an envelope of shear 

strength as a series of points (i.e., normal stress and shear stress). 

There are significant problems associated with interpreting undrained shear strengths via UU tests 

for soils that are saturated. First, inevitable sample disturbance will result in measured shear 

strengths less than the actual strengths. Second, since the shearing portion of the test occurs so 

rapidly, it is likely that the measured strengths will be higher than a strength corresponding to typical 

rates of shear. Third, when the sample is removed from the tube, the total stress on the sample is 

zero, implying that residual negative pore pressures are in the sample. These pore pressures will 

obviously affect the effective stress condition in the sample and will therefore influence the 

measured strength. Although these factors may result in “compensating errors”, it is not advisable to 

rely on undrained shear strengths obtained from UU data alone as the sole source of undrained 

strength information for design analyses. Data from UU tests should be used to complement data 

from consolidated undrained triaxial tests and from interpreted undrained strengths from in-situ 

testing, as discussed subsequently. 

Consolidated Undrained Triaxial Test 

The CU triaxial test provides data that can be used to interpret total stress strength parameters. 

Consolidated undrained triaxial testing with pore pressure measurements provides data for both total 

and effective stress strength interpretation. Unlike the UU test, the sample in a CU test is 

reconsolidated in the laboratory to a predetermined consolidation pressure. 

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Results of triaxial tests can be plotted either as Mohr circles or as stress path plots. Stress path plots, 

also referred to as p′-q or p-q plots, depict a series of points that represent the maximum shear stress 

on the Mohr Circles (i.e., τ corresponding to the top of the Mohr circle at a normal stress of σn = 1⁄2 

( σ1 + σ3). The coordinates for each point are p= 1⁄2 ( σ1 + σ3) and q =1⁄2 ( σ1 - σ3). A line drawn 

through these points represents the total stress path for the triaxial test. If effective stresses are used, 

then p′= 1⁄2 ( σ1′ + σ3′) = p - ∆u. The parameter q is the same since (σ1′ - σ3′ ) = ( σ1 - ∆u – (σ3 – 

∆u)) = σ1 – σ3. A graphical example of the effective stress path for three specimens consolidated to 

different consolidation stresses and sheared in undrained triaxial compression (see figure 79 for 

stress-strain and pore pressure-strain curves for the three specimens) is shown in figure 80. As with 

a Mohr circle evaluation of the strength parameters, a best-fit line may be drawn through the data 

and the test result parameters α′ and a′ can be calculated. To evaluate the Mohr-Coulomb effective 

stress strength parameters from an effective stress path plot the following equations are used: 

'cos/''

'tan'sin

φ

αφ

=a

=

c (Equation 60) 

As noted above, both total stress and effective stress parameters can be obtained from CU tests with 

pore pressure measurements. Effective stress parameters may be used for the evaluation of longterm conditions for cohesive soils since φ′ and c′ can be directly calculated. It is noted, however, 

that an effective stress analysis also requires knowledge of the in-situ pore pressures. For many 

design applications, it is very difficult to estimate these pore pressures with reasonable certainty. 

Analyses for short-term conditions in cohesive soils are performed using total stress parameters. 

The recommended approach for evaluating undrained shear strengths from CU tests for saturated 

soils for use in short-term total stress analyses includes developing a relationship between undrained 

shear strength su (which assumes φ = 0) and vertical effective stress. 

As discussed in chapter 4 and as illustrated on figure 23, for soils that are normally to lightly 

overconsolidated (i.e., OCR < 1.5), a CU test performed on a disturbed laboratory specimen that is 

reconsolidated back to the in-situ effective stress will likely overestimate the undrained strength 

since the specimen will be more dense (i.e., be at a lower void ratio) at a particular effective stress in 

the laboratory as compared to that at the same effective stress in the ground at the depth of the 

sample. Because of this, CU tests should be performed at consolidation pressures greater than the 

effective stress in the ground to compensate for the effects of disturbance. Because consolidation to 

higher pressures will result in higher undrained strengths, the undrained strength measured using a 

CU test at consolidation pressures greater than those corresponding to the depth at which the sample 

was taken is not a correct measure of the undrained strength for the depth in the ground where the 

sample for the CU test was taken. Using this information, however, a relationship between 

undrained strength and consolidation pressure can be developed. Typically, the undrained strength 

ratio, defined as su/σvo′, is calculated and used for this relationship. For a slope stability analysis, for 

example, the program will calculate the effective vertical stress for each slice and then based on the 

relationship between undrained strength and effective vertical stress, the appropriate undrained 

strength will be assigned. Alternatively, a constant value of su can be assigned to each soil layer in 

the stability analysis. 

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Figure 79. CIU triaxial compression test results. 

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Figure 80. Effective stress path in undrained shear. 

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An important issue relative to the interpretation of CU tests for total stress undrained analyses, is that 

a total stress friction angle and cohesion intercept, i.e., φΤ and cT that may be calculated from a Mohr 

circle representation of the total stresses at failure from CU tests, should not be used for analyses that 

assume undrained conditions. Undrained analyses are based on no consolidation of cohesive 

deposits under the application of, for example, an embankment loading. As an example, the 

incorrect effect of using φΤ and cT evaluated based on CU test results as input data for a stagedloading stability analysis would be that as soon as fill is placed, the foundation soils would 

“experience” an increase in shear strength corresponding to the equation: 

TTn tan (Equation φστ += c 61) 

Since the normal stress, σn, will increase due to the placement of fill, the shear strength of the soil 

will increase instantaneously according to equation 61. This is incorrect. To correctly model the 

effect of placing fill, but not realizing an immediate increase in shear strength, requires that the shear 

strength be written according to the φ=0 concept, such that: 

n uu =στ )0(tan =+ ss (Equation 62) 

In this case, the undrained strength of the foundation soils will not increase as a result of the 

placement of the fill but will instead be correctly based on the undrained strength corresponding to 

the state of stress in the soil before the load was placed. For a staged construction analysis, this 

would include the increased effective stresses in the ground resulting from the previously placed fill 

which has been allowed time to consolidate and gain strength. 

5.6.4.3 Direct Shear Testing 

Direct shear testing is commonly performed on compacted materials used for embankment fills and 

retaining structures. This testing can also be performed on natural materials; however, the lack of 

control on soil specimen drainage makes the evaluation of undrained strength unreliable. This test 

can be used to evaluate the drained strength of natural materials by shearing the sample at a slow 

enough rate to reasonably ensure that no porewater pressures develop. 

During the direct shear test, the normal stress and shear stress are measured on a horizontal failure 

surface. In interpreting the shear strength of a soil from a direct shear test, it is assumed that the 

normal stress on the failure plane is equal to the confining pressure placed on the sample at the 

beginning of the test (corrected for changes in specimen area as the specimen is sheared) and that the 

shear stress on the failure plane is calculated as the applied shear force divided by the corrected area. 

For a given soil, usually three tests are performed, each at a different confining pressure, σn. The 

results are plotted as shown in figure 81. 

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Figure 81. Direct shear test results. 

As shown in figure 82, the contact area between the two specimen halves varies with the relative 

displacement, δ, between the upper and lower shear box. The corrected area Ac of the sheared 

specimen for a square box length, a is: 

aaA −= δ )( c (Equation 63) 

and for the cylindrical box of internal diameter D: 

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



 



 −= θ δ θ sin

2

2

D

D Ac (Equation 64) 

where θ = cos-1(δ/D) (in radians). The area correction needs to be applied to both the normal stress 

and the shear stress. For a typical sample diameter of 6.3 cm, the error on shear and normal stresses 

may be 20% when δ =1 cm (Bardet, 1997), so it is important to make the appropriate correction, 

particularly if specific values of τ are used in analyses (e.g., when using the test results to obtain 

undrained shear strengths). 

Figure 82. Area correction for direct shear test. 

Direct shear testing performed on recompacted soils should be performed for a normal stress range 

consistent with the anticipated range of stresses for the field application. Tests should be performed 

at varying compaction and density conditions to be representative of the anticipated compaction 

conditions that can be achieved in the field. Consideration should be given to the potential for the 

soils to become saturated after construction and, if saturation is possible, tests should be performed 

on saturated (or submerged) samples. Tests should always be carried out until a stable largedisplacement shear stress is measured. 

The direct shear test is applicable to evaluating the shear strength of stiff, fissured materials or other 

materials that may contain pre-existing shear surfaces. In those cases, the specimen may be trimmed 

in such a manner so that the existing shear surface is oriented horizontally. 

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5.6.4.4 Unconfined Compression Testing 

This test is typically performed on cohesive soils and provides a rapid means to obtain an 

approximate value of the undrained shear strength of cohesive soils. In this test, a cylindrical soil 

specimen is loaded axially in compression. This test cannot be performed on granular soils, or 

fissured or varved materials. This test represents a special case of the UU test wherein the total 

confining pressure is equal to zero. Since residual negative pore pressures reside within the 

specimen, however, the actual state of stress prior to shear is unknown. 

Figure 83 shows a representative stress-strain curve and Mohr circle representation of the state of 

stress for a soil tested in unconfined compression. In a UC test, a peak in the stress-strain curve 

usually evidences failure, although a limiting strain value (e.g., say 15%) may be used to define 

failure for soils that do not exhibit a discernable peak value. When the failure envelope is assumed 

to be purely cohesive (i.e., φ=0) as shown on figure 83, the Mohr circle is tangent to the horizontal 

line τ=su. 

Figure 83. Typical stress-strain curve and Mohr circle representation of the state of stress for an 

unconfined compression test. 

In most cases, the undrained strength measured in an unconfined compression test is less than that 

which would be measured in the field or for a triaxial compression test, even a laboratory UU triaxial 

test result. The unconfined compression test does not include a reconsolidation phase wherein the 

stresses in the ground are applied to the specimen. Also, sample disturbance can result in significant 

underestimation of actual strength. Because the undrained strength of a soil is affected by moisture 

content where, for example, a small change in moisture content may result in a significant change in 

undrained strength, it is necessary to take great care in preserving the in-situ (or field) moisture 

content of the soil specimen, particularly during sample preparation. If possible, all sample 

preparation work should be performed in a humidity controlled room. 

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5.6.4.5 Relevance of Laboratory Strength Tests to Field Conditions 

As noted, the undrained strength to be used for design analyses depends on the direction of loading. 

For example, figure 84 shows an embankment with an assumed slip surface. This figure indicates 

that three different modes of soil shearing would be involved in evaluating the average shear 

strength along the failure surface. Therefore, accurate evaluation of undrained strength to be used in 

the stability analysis of an embankment constructed over soft to medium clays must somehow 

recognize the differences in measured undrained strengths for each of these shearing modes. Also, 

many field situations involve plane strain conditions such as that for a continuous footing, retaining 

wall, or long embankment; therefore the use of triaxial testing conditions may not be accurate. It is 

recognized, however, that to perform all laboratory tests that may be pertinent to a particular field 

loading condition may be impractical, especially for relatively small projects. Therefore, CU triaxial 

tests with pore pressure measurements are recommended as a standard reference test. Using the 

results of the CU test, the results of all other tests can be compared conveniently, even those tests 

following different stress paths to failure than the CU test. This is demonstrated in subsequent 

sections. 

Figure 84. Shear modes for an embankment slip surface. 

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5.6.5 Undrained Shear Strength from In-situ Tests 

The interpretation of undrained strength from in-situ tests involves a direct relationship to convert 

the in-situ test measurement (e.g., KD from DMT, qt from CPT) to su, based on empirical, analytical, 

or numerical methods. The common approaches for the vane, cone, piezocone, pressuremeter, 

dilatometer, plate, and SPT are summarized in table 33. The correlations presented in this table can 

be used to provide preliminary estimates of su. The use of the correlations in table 33 is illustrated in 

Soil Property Selection Example No. 1 in appendix A. 

Table 33. Conventional methods of interpretation for su from in-situ tests. 

IN-SITU TEST COMMENTS REFERENCES 

VST: suv = 6T/(7π D3) 

for H/D = 2 

Static equilibrium analysis 

Empirical: μ ≈ 2.5(PI)-0.3 ≤ 1.1 

Chandler (1988, ASTM 1014) 

PMT: supmt = dp/d(ln εv) 

supmt = (pL-po)/Nc

Cavity expansion theory 

Empirical bearing factor Nc = 5.5 

Windle & Wroth (1977, ICSMFE). 

Baguelin et al. (1972, JSFMD). 

SPT: su(N60) = f1N60pa/100 Empirical: f1 = 4.5 for PI = 50 

Empirical: f1 = 5.5 for PI = 15 

Stroud (1974, ESOPT-1) 

Stroud (1989, PTUK) 

CPT: sucpt = (qc-σvo)/Nc 

sucpt = (qt-σvo)/Nkt

Limit plasticity theory 

Cavity expansion theory 

Corrected cone tip resistance, qt

NkT = 10 (TC) 

= 15 (DSS) 

= 20 (TE) 

Meyerhof (1951, Geotechnique) 

Vesic (1977, NCHRP) 

Aas, et al. (1986, ASCE GSP 6) 

CPTu2: sucptu2 = ∆u/N u N u2 = 7.9 (uncorrected vane) 

Charts: N u = f(Ir, Af, u1 or u2) 

Cavity expansion + critical state 

Tavenas, et al. (1982, ESOPT). 

Robertson and Campanella (1983) 

Mayne & Bachus (1989, ISOPT) 

DMT: 

suDMT = 0.22 σvo'(1⁄2KD)

1.25 

suDMT = (po-uo)/10 

suDMT = dsσvo' (0.5KD)

1.25

Based on mix of UU, UC, VST 

Cavity expansion theory 

Empirical and test-dependent: 

TC: ds = 0.20 

VST: ds = 0.19 

DSS: ds = 0.14 

Marchetti (1980, JGE). 

Schmertmann (1991) 

Lacasse & Lunne (1988, ISOPT) 

PLT: suplt = qult/6.18 Limit plasticity theory Meyerhof (1951, Geotechnique) 

Symbols used in table 33

 VST = vane shear test CPTu2 = piezocone test 

 PMT = pressuremeter test DMT = flat dilatometer test 

 SPT = standard penetration test PLT = plate load test 

 CPT = cone penetration test 

The correlations shown in table 33 that convert a measured in-situ test parameter to an undrained 

strength are widely used in practice. The “conversion factors” used in these correlations (e.g., NKT

for CPT) must be calibrated to the results from high-quality laboratory tests for the specific soil 

deposit and to a particular mode of shearing. For example, some practicing engineers have adopted 

a value of 15 for NKT for CPT, although the appropriate value for a particular deposit may be 

significantly higher or lower than 15. Also, many engineers use the vane shear test to evaluate su for 

soft clays that cannot be easily sampled for laboratory tests. Engineers use the vane shear correction 

factor μ to account for shear rate and strength anisotropy effects. This μ factor is a function of PI 

and can be written as: 

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μ ≈ 2.5 (PI)-0.3 < 1.1 (Equation 65) 

Although the vane shear test is widely accepted and used, the scatter of the data originally used to 

develop the μ factors (see figure 85) is significant. Additional limitations associated with the use of 

these correlations are provided below. 

Figure 85. Plasticity based VST correction factors. 

Several difficulties exist in using the approaches summarized in table 33 to convert in-situ tests 

directly to undrained strength. First, each in-situ testing device loads the soil in a different direction 

at a different rate, and therefore, the effects of boundary conditions, strength anisotropy, and strain 

rate influence the results. Second, each field test utilizes different models as the basis for 

interpretation (e.g., limit plasticity, limit equilibrium, cavity expansion, numerical method, or 

empirical correlation) such that large inconsistencies exist when the results are compared. Thirdly, 

each field test has been calibrated to its own particular laboratory reference test (e.g., vane shear test 

and consolidated anisotropically undrained compression (CAUC) triaxial test as is reported, 

Chandler, 1988), and consequently, a wide number of interrelationships would be required to convert 

results from one test to another test. Given this, it is clear that measured undrained strengths from 

different in-situ tests and laboratory tests are expected to be quite different, resulting in difficulties in 

selecting appropriate undrained strengths for design applications. See the example profile of su with 

depth for a soft clay/silt site in figure 86 as an illustration. 

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Figure 86. Example su profile (after Finno and Chung, 1992) (Note: for clarity, the comparison of 

testing results has been shown on two separate figures). 

To use in-situ tests to evaluate su, it is recommended that the in-situ results be referenced to the 

preconsolidation stress (σp′). This is advantageous since each of the laboratory test modes (e.g., 

isotropically consolidated undrained compression (CIUC), plane strain compression (PSC), and 

other modes corresponding to isotropic and anisotropic consolidation, compression, extension, or 

simple shear loading, and triaxial or plane strain conditions) all provide a different measured su value 

for a given soil, whereas the preconsolidation stress (σp′) is uniquely defined from one-dimensional 

consolidation. In using in-situ tests, the equations developed in section 5.4.3 that relate σp′ to 

measured in-situ test parameters can be used with the results of one-dimensional laboratory 

consolidation tests to develop a profile of σp′ with depth. The recommended relationship between su

and σp′ is described below. 

Undrained shear strength can be expressed in a form that relates to the stress history (i.e., OCR) of 

the deposit: 

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su/σvo' = S OCRm (Equation 66) 

where S = normally-consolidated undrained strength ratio (also given in symbol form as 

[(su/σvo′)NC)] and m = exponent. Experimental studies have shown the parameters S and m to vary 

with test mode (Jamiolkowski, et al. 1985; Koutsoftas and Ladd, 1985; Kulhawy and Mayne, 1990). 

Values for S typically range from 0.15 to 0.30 and m is typically 0.8. Theoretical relationships 

based on critical-state soil mechanics (Wroth, 1984) have shown that S depends on test mode and 

increases with φ′ of the clay, while the exponent parameter relates to the compression indices by: 

m ≈ 1 - Cs/Cc.

 (Equation 67) 

As previously discussed, the laboratory-shearing mode that best approximates the range of 

conditions encountered for typical design analyses is the direct simple shear mode. The undrained 

shear strength from this test, i.e., suDSS, represents an overall "average" strength that is intermediate 

between triaxial compression and triaxial extension, as well as between plane strain compression and 

extension. The suDSS has been shown suitable for direct use in analyses of slopes, foundation bearing 

capacity, and embankment stability (e.g. Larsson, 1980; Ladd, 1991). In this regard, for 

overconsolidated intact clays, the recommended relationship is: 

(su/σvo')DSS = 0.23 OCR0.80 (Equation 68) 

The value of 0.23 for (su/σvo′)NC was developed from a large database of DSS testing results. For 

soils with preferential shear planes that are near horizontal (e.g., varved clays), the value for 

(su/σvo′)NC may be as low as 0.16. Work performed by Mayne (1988), however suggests that the 

value of (su/σvo′)NC is influenced by the drained friction angle of the soil and can be written in 

general as: 

( ) 2

'sin '/ φ σ = u vo NC s (Equation 69) 

Therefore, equation 68 can be written as: 

( ) 80.0

2

'sin vo '/ DSS OCR φ su σ = (Equation 70) 

For very soft clays where the degree of overconsolidation is small (OCR < 2), the expression can be 

approximated by: 

suDSS ≈ 0.21 σp' (Equation 71) 

The use of these equations coupled with the interpreted profile of preconsolidation stress, σp′, can be 

easily implemented into a spreadsheet and used to generate profiles of undrained shear strength with 

depth for cohesive soil deposits based on in-situ and laboratory oedometer results. Profiles of 

undrained strength developed from these equations can also be used to judge potential outlier data 

from laboratory and in-situ tests. The information presented in this section can be used to develop 

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undrained strength profiles for use in design analyses. Section 5.6.7 provides a detailed discussion 

on the development of undrained strength profiles. 

5.6.6 Drained Friction Angle of Granular Soils from In-situ Tests 

It is common practice to evaluate the effective stress or drained friction angle of granular materials 

from in-situ penetration tests via a correlation to a measured test parameter. In this section, 

correlations are presented for the SPT, CPT, and DMT. These correlations can be easily 

implemented into a spreadsheet to evaluate friction angle as a function of depth within granular 

deposits. 

Friction Angle Based on SPT 

Table 34 presents baseline relationships for evaluating the drained friction angle of cohesionless 

soils. This table is based on data for relatively clean sands. Given this, selected values of φ′ based 

on SPT N values should be reduced by 5° for clayey sands and the value from the table should be 

increased by 5° for gravelly sands. 

Table 34. Relationship among relative density, SPT N value, and internal friction angle of 

cohesionless soils (after Meyerhof, 1956). 

State of 

Packing 

Relative Density 

(%) 

Standard Penetration Resistance, N 

(blows/300 mm) 

Friction angle, φ′

(°) 

Very loose 

Loose 

Compact 

Dense 

Very dense 

<20 

20-40 

40-60 

60-80 

>80 

<4 

4-10 

10-30 

30-50 

>50 

< 30 

30-35 

35-40 

40-45 

>45 

Note: N = 15 + (N′ - 15) / 2 for N′ > 15 in saturated very fine or silty sand, where N′ = measured 

blow count and N = blow count corrected for dynamic pore pressure effects during the SPT. 

Equation 72 is a derived correlation between φ' and normalized SPT resistance ((N1)60), where highquality undisturbed frozen samples of natural sands were obtained that permitted direct 

measurements of φ' in triaxial cells (Hatanaka and Uchida, 1996). The data were obtained using an 

automatic trip hammer system where energy efficiency was reported as 78 percent. For an average 

state-of-practice with 60% efficiency in the U.S., the expression for peak φ' is given as: 

φ = N 601 20)(4.15' °+ (Equation 72) 

The well-known correlation between N60 and φ′ developed by Schmertmann (1975) is shown in 

figure 87. Results from this correlation tend to be somewhat conservative, especially for shallow 

depths (i.e., less than 2 m). The correlation shown in figure 87 can be approximated as: 

[ ( )] 34.0 '

60

1 /3.202.12/tan' φ ≈ N + σ Pavo

− (Equation 73) 

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0

20

40

60

0 100 200 300

VERTICAL EFFECTIVE STRESS σ'vo (kPa)

SPT BLOWCOUNT (N60)

25°

30°

35°

40°

Ф'=50° 45°

Figure 87. Correlation of φ′ with SPT N60 in clean sands. 

It is important to recognize that these correlations have been developed for relatively clean sands. 

Use of these correlations in, for example, micaceous sands is not recommended. The presence of 

mica in sand will tend to reduce the SPT blowcount significantly below that which would be 

measured for the same sand without mica. However, the actual friction angle of the clean sand and 

the micaceous sand may not be significantly different when measured in laboratory triaxial tests. 

Laboratory triaxial tests should be performed on silty sand soils where more exact values of φ′ are 

required. 

The SPT (and hence the correlations noted above) should not be used to estimate the drained friction 

angle of mostly gravel soil, unless the correlations are used conservatively or can be modified based 

on local experience. The size of gravel particles is often larger than the inner diameter of the split 

sampler used in the SPT test, thus it is likely that the SPT will overestimate the penetration 

resistance of a gravelly soil. In most cases, however, shear strengths of gravelly soils are sufficiently 

high for most engineering applications. Issues related to gravelly soils relate mostly to constraints 

associated with installation of driven piles and/or drilled shafts. However, in some highly seismic 

regions, it may be necessary to assess the liquefaction resistance of gravelly soils. In those cases, 

large penetration tests (LPT) and Becker penetration tests (BPT) may need to be performed. 

Friction Angle Based on CPT

Bearing capacity theory is used to correlate φ' in sands to the measured point resistance in cone 

penetration testing. Robertson and Campanella (1983) developed the relationship between peak φ' 

and normalized cone tip resistance shown in figure 88. This may be approximated by: 

[ ( )'/log38.01.0arctan' ] φ tc = + q σ vot (Equation 74) 

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0

50

100

150

200

250

300

350

400

0 5 10 15 20 25 30 35 40 45 50

CONE RESISTANCE qt (MPa)

VERTICAL EFFECTIVE STRESS 

σ'vo (kPa) 

34° 40°

46°

30° 32° 36° 38°

42°

44°

Ф'=48°

Figure 88. Correlation of φ′ with normalized CPTqt data in clean sands. 

Friction Angle Based on DMT

Campanella and Robertson (1991) developed an equation relating the dilatometer horizontal stress 

index, KD, to soil friction angle. This correlation is theoretically based as it considers the geometry 

of the dilatometer and an assumed soil failure surface around the dilatometer. Marchetti (1997) 

developed a lower bound envelope for φ′ given by: 

φ' ≈ 28° + 14.6 log KD - 2.1 log2

 KD (Equation 75) 

This correlation is plotted in figure 89. 

25

30

35

40

45

0 5 10 15 20

KD

Ф'

Figure 89. Correlation of φ′ with the DMT KD parameter for clean sands. 

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5.6.7 Selection of Total Stress Parameters (su) for Undrained Strength Design Analyses 

For design analyses of short-term conditions in normally to lightly overconsolidated cohesive soils, 

the undrained shear strength, su, is commonly evaluated. A profile of su with depth should be 

developed at several locations across the site. Since undrained strength is not a unique property, 

profiles of undrained strength developed using different testing methods will be different. Typical 

practice on transportation projects is to develop profiles of su based on laboratory CU and UU testing 

and, for cases where undisturbed sampling is very difficult, field vane testing. Other in-situ methods 

have been presented which can also be used to develop a profile of su with depth. In this section, 

guidance on the development of undrained strength profiles is provided. A specific example of 

undrained strength profile development is provided in Soil Property Selection Example No. 1. 

Specific issues that should be considered when developing a profile of undrained shear strength with 

depth are described below. 

• Strength measurements from hand torvanes, pocket penetrometers, or unconfined 

compression tests should not be used solely to evaluate undrained shear strength for design 

analyses. Consolidated undrained triaxial tests with pore pressure measurements and in-situ 

tests should be used. 

• All available undrained strength data should be plotted with depth. The type of test used to 

evaluate each undrained strength should be clearly identified. Known soil layering should 

be used so that trends in undrained strength data can be developed for each soil layer. 

• Review data summaries for each laboratory strength test method. Moisture contents of 

specimens for strength testing should be compared to moisture contents of other samples at 

similar depths. Significant changes in moisture content will affect measured undrained 

strengths. Review Atterberg limits, grain size, and unit weight measurements to confirm 

soil layering. 

• CU tests on normally to lightly overconsolidated samples that exhibit disturbance should 

contain at least one specimen consolidated to approximately at least 4 times σp′ to permit 

extrapolation of the undrained shear strength at σp′. 

• Undrained strengths from CU tests correspond to the effective consolidation pressure used 

in the test. This effective stress needs to be converted to the equivalent depth in the ground. 

• A profile of σp′ (or OCR) should be developed and used in evaluating undrained shear 

strength. 

• Correlations for su based on in-situ test measurements (i.e., those based on table 33) should 

not be used for final design unless they have been calibrated to the specific soil profile under 

consideration. 

A plot of an undrained strength profile should be developed with undrained strength on the x-axis 

and depth on the y-axis. Laboratory test data including CU and UU testing should be plotted at the 

correct depths. Typically, CU strengths will be greater than UU strengths, and this should be used to 

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judge the quality of the data. In-situ test data, which has been correlated to undrained strength, 

should be plotted at the depth where the measurement was taken. For strengths developed based on 

CPT, CPTu, and DMT, for which numerous measurements may have been taken, the data should be 

plotted as points without connecting the data with lines. This will facilitate visual identification of 

upper and lower bounds and anomalous data. 

For intact clays (i.e., not fissured), the relationships developed for undrained strength (i.e., equation 

70 should be plotted on the profile as well. The average effective stress friction angle from CU tests 

and the profile of preconsolidation stress is used to calculate the undrained strength using these 

relationships. This relationship has been developed from a database that includes over 100 clays and 

it can therefore be used as a means to evaluate potential data outliers. For example, strengths 

measured in CU tests should be higher than that for DSS. Therefore, the quality of the CU data can 

be judged by reviewing the overall trend in the CU data relative to the relationships based on the 

DSS, i.e., on average, the CU strengths should be higher. Also, a lower bound line can be drawn to 

represent the undrained shear strength assuming that the entire soil layer is normally consolidated 

(i.e., OCR = 1). This would be a line defined by su=(0.5 sin φ′) σvo′. Data that fall below this line 

are likely to be anomalous and should not be included in the interpretation. 

With all the data plotted, envelopes of undrained strength for design analyses should be developed. 

Also, the best-fit profile line should be developed using data corresponding to the likely range of 

stresses in the field. Undrained strength data corresponding to vertical effective stresses outside the 

range for the field application can be used for confirmation of the profile selected, but should not 

have equal weighting as for data points within the range of anticipated field stresses. It is noted that 

reported variations in undrained strength have been up to 40 percent (Duncan, 2000) meaning that 

typical standard variation in undrained strength may be equal to 40 percent of the average value. 

These variations are based on measurements of undrained strength using various testing methods for 

the same soil. Therefore, specific rules regarding the selection of a single profile line to represent 

undrained shear strength with depth cannot be developed. However, recognizing that undrained 

shear strength is a parameter used for limit state analyses (as compared to analyses of serviceability 

states), it is appropriate to develop an envelope for assumed worst credible case or critical case 

conditions. That is, analyses should be performed with a lower bound envelope to assure that 

calculated factors of safety are greater than 1.0. This lower bound envelope, however, needs to be 

developed for relevant data only; that is anomalous low values must be removed from the data set 

prior to developing this envelope. Unlike effective stress analysis wherein a lower bound friction 

angle can be selected with reasonable confidence (e.g., select a residual friction angle based on soil 

PI), the selection of lower bound undrained strength is much more difficult. 

For heavily overconsolidated, fissured clays, the undrained strength in the field can be reduced by as 

much as 50 percent of the values obtained from laboratory testing because of the added macrofabric 

of cracks and fractures. When testing overconsolidated, fissured clays in the laboratory, this 

difference in measured strength can be minimized by testing samples with a maximum practical 

diameter (e.g., use full diameter samples from thin-wall tubes for UU testing). At relatively shallow 

depths, the effect of fissures on measured laboratory undrained strengths needs to be carefully 

considered since it is possible that overburden pressures in the field will not be large enough to close 

the fissures. At higher pressures, fissures may close and the undrained strength in the field will be 

closer to those measured in the laboratory. Due to the fissured nature of overconsolidated clays, 

which can permit relatively rapid local drainage at the level of the discontinuities in the clay (which, 

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in turn, may cause significant softening of the soil), it is generally difficult to define with any 

certainty the period of time during which the undrained strength of the clay may reliably be assumed 

to apply. Therefore, in overconsolidated, fissured clays, design analyses should be performed in 

terms of drained effective stress parameters, especially for cases where the clay soil is subject to 

unloading during the short-term (e.g., unloading due to soil removal at the base of an excavation or 

due to cut slope construction). 

It is recognized, however, that certain state-of-the-practice design analyses for short-term conditions 

such as capacity evaluations of driven piles and drilled shafts are performed in terms of the 

undrained strength. Many of these design analyses are semi-empirical and based on successful 

design in these materials. Engineers involved in the design of geotechnical features in heavily 

overconsolidated, fissured materials should be cognizant of the fact that the short-term strength of 

these materials is typically much greater than the long-term drained strength. Design analyses 

should thus consider both cases. 

5.6.8 Selection of Effective Stress Parameters (φ′, c′) for Design Analyses 

Long-term effective stress strength parameters (c′ and φ′) of clays are best evaluated by slow 

consolidated drained direct shear box tests, consolidated drained triaxial compression tests, or 

consolidated undrained triaxial tests with pore pressure measurements. In laboratory tests, the rate 

of shearing should be sufficiently slow to ensure substantially complete dissipation of excess pore 

pressure in the drained tests or, in undrained tests, complete equalization of pore pressure throughout 

the specimen. Information on appropriate shearing rates has been provided in section 4.12.5.3. 

Laboratory tests should always be carried out to displacements sufficiently large to reach a stable, 

post-peak shear stress. As previously discussed, the selection of peak, fully-softened, or residual 

strength for design analyses must be made based on a careful review of the expected or tolerable 

displacements of the soil mass. Where pre-existing weak interfaces are present, direct shear methods 

should be used, if practical, to set up the sample so that the interface strength can be directly 

evaluated. Alternatively, soil material from within the shear zone should be sampled and 

reconstituted to the in-situ moisture content for shear testing. 

The use of a cohesion intercept (c′) for long-term analyses in natural materials must be carefully 

assessed. With continuing displacements, it is likely that the cohesion intercept value will decrease 

to zero for long-term conditions, especially for highly plastic clays. 

The strength of granular strata can be assessed from penetration test data particularly the SPT, CPT, 

and DMT. Since laboratory testing on undisturbed samples of granular materials is impractical, it is 

necessary to rely on correlations to obtain the effective stress friction angle. 

5.7 HYDRAULIC CONDUCTIVITY PROPERTIES OF SOIL 

5.7.1 Introduction 

The hydraulic conductivity of soil is important in the design of many transportation structures 

because it provides information related to the role that water is expected to have on the design, 

constructability, and performance of the structure. It is recognized that knowing what role water will 

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play (e.g., drained versus undrained, static versus flowing, etc.) is critical to understanding how the 

soil will behave. 

Hydraulic conductivity data, whether obtained from the field or the laboratory, should be scrutinized 

due to test complexities (discussed in Section 4.12.5.4). These data should be considered in the 

context of the site geology and the method used to obtain the values. They are only as reliable as the 

method used to obtain them, keeping in mind that even excellent laboratory and field methodology 

may only provide values within an order of magnitude of actual conditions. 

5.7.2 Laboratory Output/Data Reduction 

The following parameters should be required in the laboratory output: 

• Hydraulic conductivity recommendation; 

• k versus head, k versus q, qout/qin information/plots; 

• Sample dimensions (initial and final); 

• Water content (initial and final); 

• Sample weight (initial and final); 

• Degree of saturation (initial and final); 

• Permeameter type; 

• Hydraulic gradient; 

• Effective stress conditions; 

• Temperature readings during the test period. 

The resultant k value is meaningless without detailed knowledge of how the test was performed. 

These tests are difficult to perform accurately and results can be highly variable and highly 

dependent on laboratory conditions. The information listed above is essential and should be 

carefully analyzed to ensure that the test conditions were as close to field conditions as possible and 

that they match those conditions specified by the engineer. Special attention should be given to the 

possibility of sample disturbance, sidewall leakage, and other test errors that could affect the results. 

Trends in k and temperature during the test, gradients, stress conditions, and the degrees of 

saturation are the most important parameters to consider. Hydraulic conductivity values should have 

stabilized long before test termination, the gradients should be well below those listed in the ASTM 

D 5084 standard, stress conditions should match those from the zone of sample collection, and the 

degree of saturation should approach 100% (keeping in mind that it is impossible to obtain 100% 

saturation). Temperature changes during laboratory testing will affect the viscosity of water and, 

therefore, the resulting k values, so controlled laboratory tests conditions are essential. 

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The hydraulic conductivity data should be accompanied by Atterberg limits and grain size 

distribution information for correlation with field conditions and for textbook 

correlations/approximations of k. After receiving the laboratory data, the k values should be 

checked against the estimated values as discussed in the following section. If there is a major 

discrepancy (more than two orders of magnitude) this could be a sign of laboratory error, or possible 

anomalous soils. 

5.7.3 Correlation Methods 

Prior to reducing hydraulic conductivity data, hydraulic conductivities should be estimated using 

published charts, local data (if available), and correlations. Almost every soils, geology, and 

hydrogeology textbook contains charts of soil/rock type versus hydraulic conductivity. Figure 90 is 

a reprint of a chart created by Casagrande and later modified by Holtz and Kovacs (1981). This 

chart is useful because it provides information on appropriate laboratory conditions as well as 

hydraulic conductivity ranges for various soil types. Local data, available through the USDA Soil 

Conservation Service, USGS, state geologic survey, or local water supply district (if groundwater 

sources are used in the area), may provide the best initial estimate of k for the site area. Soil surveys 

will typically contain k values for all of the surficial soil types in the district; however, if subsurface 

information is required, the survey will provide little to no information. 

Hazen’s equation is the most common correlation equation used to estimate hydraulic conductivity 

for sands (k > 10-3 cm/s). This equation is written as: 

k = C(d10)

2 (Equation 76) 

where: k is the hydraulic conductivity in cm/s; 

 C = coefficient ranging from 0.4 to 1.2 depending on sand size/sorting; and 

 d10 = effective grain size in mm at 10% passing by weight. 

This equation is based solely on grain size and it requires input (d10) from particle size distribution 

curves and the use of a coefficient estimated based on sand type (e.g., fine sand, poorly sorted, etc.). 

Hazen’s equation should be used with caution since it only provides very approximate k estimates 

applicable only to clean sands (with less than 5% passing the No. 200 sieve) with d10 sizes between 

0.1 and 3.0 mm (Holtz and Kovacs, 1981). Table 35 contains a comparison of Hazen’s equation 

estimates with slug test information from an alluvial sand aquifer. One can see that in this deposit, 

the error ranged from slight to over one order of magnitude. The error would be expected to 

decrease as the sand becomes more uniform. Local knowledge of the soil conditions can allow sitespecific values of C to be established. 

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Good drainage

10-8 10-9 10-7 10-6 10-5 10-4 10-3 10-2 10-1 10 1.0 1 102 10-8 10-9 10-7 10-6 10-5 10-4 10-3 10-2 10-1 10 1.0 1 102

Poor drainage Practically impervious

Clean gravel Clean sands, 

Clean sand and gravel mixtures

Impervious sections of earth dams and dikes 

“Impervious” soils which are modified by the 

effect of vegetation and weathering; fissured, 

weathered clays; fractured OC clays

Pervious sections of dams and dikes

Very fine sands, organic and 

inorganic silts, mixtures of sand, silt, 

and clay glacial till, stratified clay 

deposits, etc. 

“Impervious” soils 

e.g., homogeneous 

clays below zone of 

weathering

Drainage property

Application in earth 

dams and dikes

Type of soil

Direct determination 

of coefficient of 

permeability

Indirect determination 

of coefficient of 

permeability

*Due to migration of fines, channels, and air in voids.

Direct testing of soil in its original position (e.g., well points). 

If properly conducted, reliable; considerable experience required. (Note: Considerable experience 

also required in this range.)

Constant Head Permeameter; 

little experience required.

Constant head test in triaxial cell; 

reliable w ith experience and no leaks.

Reliable;

Little experience

required

Falling Head Permeameter;

Range of unstable permeability;* 

much experience necessary to 

correct interpretation

Fairly reliable; 

considerable experience necessary 

(do in triaxial cell)

Computation:

From the grain size distribution

(e.g., Hazen’s formula). Only 

applicable to clean, 

cohesionless sands and gravels

Horizontal Capillarity Test:

Very little experience necessary; especially 

useful for rapid testing of a large number of 

samples in the field w ithout laboratory facilities.

Computations:

from consolidation 

tests; expensive 

laboratory 

equipment and 

considerable 

experience 

required.

10-8 10-9 10-7 10-6 10-5 10-4 10-3 10-2 10-1 10 1.0 1 102 10-8 10-9 10-7 10-6 10-5 10-4 10-3 10-2 10-1 10 1.0 1 102

COEFFICIENT OF PERMEABILITY

CM/S (LOG SCALE)

Figure 90. Range of hydraulic conductivity values based on soil type. 

Another method of estimating hydraulic conductivity empirically through grain size was developed 

by GeoSyntec (1991) and is presented as Figure 91. As with Hazen’s equation, grain size 

distribution information is necessary to develop the input parameter (i.e., d15, grain size at 15% 

passing by weight) of the material. Using this value and the band of values corresponding to 

gradient and confining stress, a k value can be estimated. This method, as with other empirical 

methods, can only be used to provide an estimated value within 1 to 2 orders of magnitude of the insitu condition. The lower the permeability and the higher the variability of grain size in the soil, the 

higher the error in using empirical relationships based on uniform particle distribution. 

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Table 35. Comparison of hydraulic conductivity from empirical and field methods. 

Effective 

Particle Size, 

d10

(mm) 

Coefficient, 

C (Hazen) 

Hydraulic 

Conductivity 

(Hazen) 

(cm/s) 

Log of k 

(Hazen) 

Slug Test k 

Results 

(cm/s) 

Log of k from 

Slug Test 

0.8 1 6.4E-1 -0.194 4.3E-3 -2.367 

0.3 1 9.0E-2 -1.046 1.3E-2 -1.886 

0.9 1 8.1E-1 -0.092 2.2E-3 -2.658 

1.3 1 1.7E0 0.230 1.2E-2 -1.921 

2.0 1 4.0E0 0.602 1.6E-2 -1.796 

0.3 1 9.0E-2 -1.046 6.9E-3 -2.161 

0.8 1 6.4E-1 -0.194 1.8E-3 -2.745 

0.5 1 2.5E-1 -0.602 2.0E-3 -2.699 

Average 2.8E-1 5.0E-1 1.0E-2 9.4E-3 

5.7.4 Interpretation Methods 

As previously mentioned, hydraulic conductivity will vary across the site from point to point; 

however, it is important to separate potential error sources from actual field variations. Calculation 

of hydraulic conductivity from ch interpretations using piezocone dissipation test data has been 

shown to be useful in silts and clays. These in-situ test results should be compared to values from 

the laboratory, but potential for flow anisotropy should be considered in this evaluation. The 

engineer should analyze the data collected during the field investigation (e.g., CPT and boring logs) 

paying close attention to: high permeability units/lenses within an overall low permeability unit; 

potential confining units (i.e., zones of low k overlying zones of high k); fractured zones; and other 

features which will: (1) result in varying k data across the site; (2) control groundwater flow; or (3) 

result in unexpected or anomalous data. 

Incorrect Correct 

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1.E-09

1.E-08

1.E-07

1.E-06

1.E-05

1.E-04

1.E-03

1.E-02

1.E-01

1.E+00

1.E+01

1000 100 10 1 0.1 0.01 0.001 0.0001

CHARACTERISTIC PARTICLE SIZE, d15, OF THE SOIL (mm)

HYDRAULIC CONDUCTIVITY, ks (cm/s) 

Low Gradient ~ 1 to 2

High Gradient ~ 10

Confining Stress ~ 35 kPa

Confining Stress ~ 210 kPa

Figure 91. Range of hydraulic conductivity based on grain size (after GeoSyntec, 1991). 

Considering the site geology, the laboratory and field data should be tabulated with other known data 

for the sample/test location and with depth, soil/rock type, grain size distribution, Atterberg limits, 

and water content. This table should also include important test information such as: stress 

conditions, gradients, and test method. Once this table is constructed it will be much easier to group 

like soil types and k values, to delineate distinct areas within the site, and to eliminate potentially 

erroneous data. Once these values have been grouped together and potentially erroneous values 

eliminated, it may be useful to compute an average value for each grouping. When averaging, the 

log of the hydraulic conductivity value must be taken before performing an arithmetic mean or 

incorrect results will be produced. First, the logarithm of each value should be taken. Second, an 

average value should be calculated from these logarithmic values. Finally, the antilog of this 

average value should be taken to calculate the average hydraulic conductivity value. Table 35 

illustrates how to calculate the mean of the log of k data and compares this value with an incorrect 

direct arithmetic mean. 

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CHAPTER 6 

INTERPRETATION OF ROCK PROPERTIES 

6.1 INTRODUCTION 

In chapter 4, laboratory and in-situ soil and rock tests were introduced. Chapter 5 focused on the 

interpretation of soil properties, starting with identification and classification of the material and 

proceeding through the interpretation of both routine and somewhat specialized laboratory and insitu tests. The emphasis of this chapter is on rock properties. The type of tests described in this 

section range from relatively simple visual assessments to in-situ tests. There is, however, a 

conscious decision to place more emphasis on the visual assessment of rock, relative to the emphasis 

placed on the interpretation of laboratory and in-situ tests. Additionally, there is more emphasis 

placed herein on the visual assessment of rock than was devoted to the visual assessment of soil. 

The rationale for these two points of emphasis bears some introductory comments. 

• Why devote effort on visual observations, when laboratory and in-situ tests to assess specific 

properties are available? In general, most soil is considered to behave as a continuous 

medium, although it is recognized that the soil is composed of discrete particles. From 

chapter 5, we know that the exception to this characterization is the behavior of highly overconsolidated clays that are often fractured. For these stiff and hard materials, the engineer 

must be careful to distinguish between the behavior of the stiff intact material and that of a 

“blocky” discontinuous matrix since the overall behavior of stiff soil is often governed by 

the discontinuities rather than by the intact “blocks”. The behavior of rock can be 

considered to be an extreme example of the behavior of stiff soil; if the rock has 

discontinuities (e.g., fractures, bedding planes, joints, etc.), then the rock mass behavior will 

almost always be governed by the behavior of the discontinuities. Therefore, the orientation 

and characteristics (e.g., length, roughness, etc.) of the discontinuity, as well as the behavior 

of the material within the discontinuity (e.g., gouge, etc.) is critical to assessing the response 

of the rock mass to loading. Fortunately, there are several simple and inexpensive tests, as 

well as focused visual observations, that can be used to provide qualitative and quantitative 

information regarding the nature and extent of the discontinuities and the rock mass. In 

many cases, these visual observations are sufficient to provide accurate and adequate 

information for use in evaluating the engineering properties of the in-place rock mass. 

• Why devote more emphasis on visually assessing rock, compared to the effort to visually 

assess soil? Again, it is noted that most soil is generally characterized as a relatively 

continuous medium. Visually assessing discrete soil particles may prove to be helpful, 

particularly with regards to angularity in sands or silt/clay distinctions in fine-grained soils. 

Laboratory and/or in-situ tests, however, are necessary to assess the response of the soil as a 

continuum. The rock counterpart to a soil continuum includes intact, non-fractured rock, 

relatively massive sedimentary rock that does not have weak bedding planes, and relatively 

massive, non-jointed igneous and metamorphic rocks. While these type rocks exist, it is 

with a much lower frequency than discontinuous or fractured rock masses of the same 

geologic origin. This is due in large part to the fact that even a few strategically located 

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discontinuities can control behavior of the rock mass. Because of the importance of the 

discontinuities in rock, and the fact that most rock is much more discontinuous than soil, 

emphasis is placed on visual assessment of the rock and the rock mass. Laboratory and in 

situ testing can be helpful in characterizing both discontinuous and intact rock. 

6.2 ROCK MASS CLASSIFICATION 

6.2.1 Description of Rock Masses 

Standardized geologic mapping and logging procedures should be used for describing rock masses. 

The types of information collected will depend on site access, extent of rock outcrops, and the 

criticality of the proposed structure to be constructed on or in the rock mass. A method proposed by 

the International Society of Rock Mechanics (ISRM) (1981) provides standardized quantitative and 

qualitative information on rock masses. This and other rock mass classification systems are 

described in ASTM D5878. To introduce and explain the ISRM method, several figures and tables 

were prepared, primarily to provide a standardized definition of terms. Figure 92 provides an 

illustration of a rock mass and the 13 parameters that are included in a detailed rock description. 

Figure 93 shows how these parameters are divided into five categories. Table 36 provides a brief 

description of each of the terms. In addition to ISRM (1981), details on the use of this method to 

characterize rock masses are provided in Wyllie (1999) and FHWA-HI-99-007 (1998). 

Figure 92. Illustration of geological mapping terms (after Wyllie, 1999, Foundations on Rock, 

Figure 4.4b, p. 101, E&FN Spon). 

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Figure 93. List of parameters and categories describing rock mass characteristics 

(after Wyllie, 1999). 

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Table 36. Description of geological mapping terms. 

Term Description 

Rock Type The rock type is defined by the origin of the rock (i.e., sedimentary, metamorphic, igneous), color (including 

whether light or dark minerals predominate), texture or fabric ranging from crystalline, granular, or glassy, and 

grain size ranging from boulders to silt/clay size particles. 

Wall Strength The compressive strength of the rock forming the walls of discontinuities will influence shear strength and 

deformability. Rock compressive strength categories and grade vary from extremely strong (> 250 MPa grade R6) 

to extremely weak (0.25 to 1 MPa grade R0) (see table 37). 

Weathering Reduction of rock strength due to weathering will reduce the shear strength of discontinuities as well as reduce the

shear strength of the rock mass due to the reduced strength of the intact rock. Weathering categories and grades are 

summarized in table 38. 

Discontinuity 

Type 

The discontinuity type range from smooth tension joints of limited length to faults containing several centimeters of 

clay gouge and lengths of many kilometers. Discontinuity types include faults, bedding, foliation, joints, cleavage, 

and schistosity. 

Discontinuity 

Orientation 

The orientation of discontinuities is expressed as the dip and dip direction of the surface. Alternatively, the 

discontinuity can be represented by strike and dip. The dip of the discontinuity is the maximum angle of the plane 

to the horizontal (angle ψ in figure 92) and the dip direction is the direction of the horizontal trace of the line of dip, 

measured clockwise from north (angle α in figure 92). 

Roughness Roughness should be measured in the field on exposed surfaces with lengths of at least 2 m. The degree of 

roughness can be quantified in terms of the Joint Roughness Coefficient (JRC) as described in section 6.5.2.3. Wall 

roughness is an important component of shear strength, especially in the case of undisplaced and interlocked 

features (e.g., unfilled joints). 

Aperture Aperture is the perpendicular distance separating the adjacent rock walls of an open discontinuity (thereby 

distinguishing it from the width of a filled discontinuity), in which the space is air or is water filled. Categories of 

aperture range from cavernous (> 1 m) to very tight (< 0.1 mm). 

Infilling Type 

and Width 

Infilling is the term for material separating the adjacent walls of discontinuities such as fault gouge; the 

perpendicular distance between adjacent rock walls is termed the width of the filled discontinuity. Filled 

discontinuities can demonstrate a wide range of behavior and thus their affect on shear strength and deformability 

can vary widely. 

Spacing Discontinuity spacing can be mapped in rock faces and in drill core; spacing categories range from extremely wide 

(> 6000 mm) to very narrow (< 6 mm). The spacing of individual discontinuities has a strong influence on the mass 

permeability and seepage characteristics of the rock mass. 

Persistence Persistence is the measure of the continuous length or area of the discontinuity; persistence categories range from

very high (> 20 m) to very low (< 1 m). This parameter is used to define the size of blocks and the length of 

potential sliding surfaces. Persistence is important in the evaluation of tension crack development behind the crest 

of a slope. 

Number of 

Sets 

The number of sets of discontinuities that intersect one another will influence the extent to which the rock mass can 

deform without failure of the intact rock. As the number of sets increases and the block sizes reduce, the greater the 

likelihood for blocks to rotate, translate, and crush under applied loads. 

Block Size 

and Shape 

The block size and shape are determined from the discontinuity spacing, persistence, and number of sets. Block 

shapes include blocky, tabular, shattered and columnar, while block size ranges from very large (> 8 m3

) to very 

small (< 0.0002 m3

). 

Seepage Observations of the seepage from discontinuities should be provided. Seepage quantities in unfilled discontinuities 

range from very tight and dry to continuous flow. Seepage quantities in filled discontinuities range from dry in 

heavily consolidated infillings to filling materials that are washed out completely and very high water pressures are 

experienced. 

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Table 37. Rock material strengths. 

Grade Description Field Identification 

Range of Uniaxial 

Compressive 

Strength (MPa) 

R0 Extremely weak 

rock 

Indented by thumbnail 0.25 – 1.0 

R1 Very weak rock Crumbles under firm blows with point of geological 

hammer; can be peeled by a pocket knife 

1.0 – 5.0 

R2 Weak rock Can be peeled by a pocket knife with difficulty; shallow 

indentations made by firm blow with point of geological 

hammer 

5.0 – 25 

R3 Medium strong 

rock 

Cannot be scraped or peeled with a pocket knife; specimen 

can be fractured with single firm blow of geological 

hammer 

25 – 50 

R4 Strong rock Specimen requires more than one blow of geological 

hammer to cause fracture 

50 – 100 

R5 Very strong rock Specimen requires many blows of geological hammer to 

cause fracture 

100 – 250 

R6 Extremely strong 

rock 

Specimen can only be chipped with geological hammer > 250 

Table 38. Weathering grades. 

Term Description Grade 

Fresh No visible sign of rock material weathering; slight discoloration on major discontinuity 

surfaces is possible 

I 

Slightly 

weathered 

Discoloration indicates weathering of rock material and discontinuity surfaces. All the 

rock material may be discolored by weathering and the external surface may be 

somewhat weaker than in its fresh condition. 

II 

Moderately 

weathered 

Less than half of the rock material is decomposed and/or disintegrated to a soil. Fresh or 

discolored rock is present either as a discontinuous framework or as corestones. 

III 

Highly 

weathered 

More than half of the rock material is decomposed and/or disintegrated to a soil. Fresh 

or discolored rock is present either as a discontinuous framework or as corestones. 

IV 

Completely 

weathered 

All rock material is decomposed and/or disintegrated to soil. The original mass structure 

is still largely intact. 

V 

Residual soil All rock material is converted to soil. The mass structure and material fabric are 

destroyed but the apparent structure remains intact. There may be a large change in 

volume, but the soil has not been significantly transported. 

VI 

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Using the terms described in table 36 and the first category in figure 93, a typical rock material 

description would be as follows (Wyllie, 1999):

slightly weathered, crystalline, gray, fine grained, medium strong basalt 

An example of a rock mass description using table 36 and the four remaining categories in figure 93 

would be as follows: 

Columnar jointed basalt with vertical columns and one set of horizontal joints, spacing of 

vertical joints is very wide, spacing of horizontal joints wide, joint lengths are 3 to 5 m 

vertically, and 0.5 to 1 m horizontally; the discontinuity infilling is very soft clay with widths of 

2 to 5 mm. The vertical columnar joints are smooth, while the horizontal joints are rough. No 

seepage observed. 

6.2.2 Core Recovery and Rock Quality Designation 

The easiest way to characterize the amount of material recovered during rock coring is to calculate 

“core recovery” as the amount (i.e. length) of recovered material divided by the total length of the 

core run (presented as a percentage). Rock Quality Designation (RQD) (ASTM D6032) is a 

modified core recovery percentage in which the lengths of all sound rock core pieces over 100 mm 

in length are summed and divided by the length of the core run. Pieces of core that are not hard and 

sound should not be included in the RQD evaluation even if they are at least 100 mm in length. The 

purpose of the soundness requirement is to downgrade rock quality where the rock has been altered 

and/or weakened by weathering. For the RQD evaluation, lengths must be measured along the 

centerline of the core. Figure 94 illustrates the correct procedure for calculating core recovery and 

RQD. Table 39 presents the correlation of RQD to rock quality. 

The RQD is appropriate for use with all core sizes except for BQ and BX core with NX and NQ core 

size being optimal. Core breaks caused by the drilling process should be fitted together and counted 

as a single piece of sound core. Drilling breaks are usually evidenced by rough fresh surfaces, 

however for laminated rocks (i.e., rocks containing horizontally oriented fracture surfaces), it may 

be difficult to identify core breaks caused by drilling. In this case, the RQD should be estimated 

conservatively; for shear strength characterization it is conservative to not count the length near 

horizontal breaks whereas for estimates of rock blasting requirements, it is conservative to count the 

length near horizontal breaks. 

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lengthruncoreTotal

eredrerockoflengthTotal Core CRery cov ,covRe =

mm

CR

200,1

++++++ 

1208060190250200250(

=

CR= %96

lengthruncoreTotal

soundofLength pieces mm

RQD ∑ > = 100

%100*

200,1

)200190250(

mm

mm RQD ++ =

RQD= %53

Figure 94. Calculation of core recovery and RQD. 

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6.2.3 CSIR Classification 

The ISRM (1981) procedures, coupled with core recovery and RQD designations, provide vehicles 

for describing or characterizing rock and rock mass. A consistent method for rock mass 

classification is essential for evaluating rock properties used in design. The two most widely used 

rock classification systems were developed by the Council for Scientific and Industrial Research 

(CSIR) in South Africa (referred to as the “Q-rating system”) and the Norwegian Geotechnical 

Institute (NGI) (referred to as the Geomechanics system). Both of these methods are described in 

detail in FHWA-HI-99-007 (1998). The CSIR classification system is the most widely used 

procedure in the US. Therefore, in this section an overview of the CSIR system is presented. 

Table 39. Rock quality description based on RQD. 

RQD Value Description of Rock Quality 

0-25 % Very poor 

25-50 % Poor 

50-75 % Fair 

75-90 % Good 

90-100 % Excellent 

The CSIR classification system considers the specific properties or conditions of the rock/rock mass, 

as well as an adjustment for the orientations of the joints. The following properties and conditions 

of the rock or rock mass are explicitly considered: (1) compressive strength of the intact rock 

(discussed in section 6.3); (2) RQD value; (3) joint spacing; (4) condition of the joints; and (5) 

groundwater conditions. As shown in table 40, each of these parameters is given a numerical rating 

based on the relative importance of the specific parameter on the behavior of the rock mass. This 

rating is adjusted to account for joint orientation depending on the favorability of the joint 

orientation for the specific project. The overall rating of the rock mass, termed the rock mass rating 

(RMR), is calculated as the sum of the individual ratings for each of the five parameters minus the 

adjustment for joint orientation (if applicable). Based on the final RMR, the rock mass is classified 

as: (1) Class I-very good rock; (2) Class II-good rock; (3) Class III-fair rock; (4) Class IV-poor rock; 

and (5) Class V-very poor rock. The RMR can be used to estimate rock mass deformation modulus 

and shear strength of the rock mass and can be an invaluable design aid when used appropriately. 

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Table 40. CSIR classification of jointed rock mass. 

A. CLASSIFICATION PARAMETERS AND THEIR RATINGS 

PARAMETER RANGES OF VALUES 

Point load strength 

index 

>8 MPa 4 to 8 Mpa 2 to 4 MPa 1 to 2 MPa 

unia

Strength of intact 

rock material Uniaxial 

compressive 

strength 

>200 MPa 100 to 200 MPa 50 to 100 MPa 25 to 50 MPa 10 t

M

1 

Relative Rating 15 12 7 4 2 

Drill core quality RQD 90% to 100% 75% to 90% 50% to 75% 25% to 50% 

2 Relative Rating 20 17 13 8 

Spacing of joints >3 m 1 to 3m 0.3 to 1 m 50 to 300 mm 

3 Relative Rating 30 25 20 10 

Condition of joints 

Very rough 

surfaces 

Not continuous 

No separation 

Hard joint wall 

rock 

Slightly rough 

surfaces 

Separation <1mm 

Hard joint wall rock 

Slightly rough surfaces 

Separation <1mm 

Soft joint wall rock 

Slickensided surfaces or 

Gouge <5 mm thick 

or 

Joints open 1 to 5 mm 

Continuous joints 

So

4 

Relative Rating 25 20 12 6 

Inflow per 10 m tunnel length None <25 liters/min 25 to 125 liters/min 

OR OR OR OR Ratio= joint water 

pressure/major principal stress 

0 0.0 to 0.2 0.2 to 0.5 

Ground water 

General Conditions OR Completely Dry OR Moist only (interstitial water) OR Water under moderate pressure OR 

5 

Relative Rating 10 7 4 

B. RATING ADJUSTMENT FOR JOINT ORIENTATIONS 

Strike and dip orientations of joints Very favorable Favorable Fair Unfavorable 

Tunnels 0 -2 -5 -10 

Ratings Foundations 0 -2 -7 -15 

Slopes 0 -5 -25 -50 

C. ROCK MASS CLASSES DETERMINED FROM TOTAL RATINGS 

RMR Rating 100 to 81 80 to 61 60 to 41 40 to 21 

Class No. I II III IV 

Description Very good rock Good rock Fair rock Poor rock 

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6.3 ROCK UNIAXIAL COMPRESSIVE STRENGTH 

For the CSIR classification, as well as for several design applications, it is necessary to measure (or 

at least estimate) the uniaxial compressive strength of a rock core. The most reliable results for the 

uniaxial compressive strength of rock core are obtained when a series of uniaxial compressive 

strength calibration tests are carried out. 

An estimate of the compressive strength of rock can be made using available correlation information 

based on the point load strength test. It has been found, on average, that the uniaxial compressive 

strength, σc, is about 20 to 25 (average is approximately 24) times the point load strength index. 

Tests on many different types of rock, however, show that the ratio can vary between 15 and 50, 

especially for anisotropic rocks. The point load strength test is not recommended for very weak 

rocks where the uniaxial compressive strength is less than approximately 25 MPa or where 

reasonably accurate values for compressive strength of intact rock core. 

In the point load strength test, the fractured pieces of core should be examined after the point load 

strength test is completed. If a clean fracture runs from one loading point indentation to the other, 

the test results are considered acceptable. If the fracture runs across some other plane or if the 

loading points sink into the rock surface causing excessive crushing or deformation, the test result 

should be rejected (Hoek and Bray, 1977). When performing the point load strength test, 

information on the degree of weathering should be noted since the compressive strength of 

weathered material along the joint may be less than the uniaxial compressive strength. 

If a uniaxial test is conducted in accordance with ASTM D2938, it is important to follow the 

preparation guidelines established in ASTM D4543. Most importantly, the specimen should be 

selected to provide a length/diameter ratio of at least 2 but less than 2.5 and the ends must be 

smooth, parallel and perpendicular to the long axis of the core. It is also noted that for a given rock 

mass, as the size of the tested rock core increases, the uniaxial compression strength decreases, due 

primarily to the fact that as the specimen size increases there is a higher likelihood of encountering 

discontinuities (albeit potentially small) in the rock core. An advantage of the point load strength 

test is that raw core can be used. 

6.4 ROCK DEFORMATION MODULUS VALUES 

6.4.1 Intact Rock Modulus 

For certain design applications, it may be necessary (or desirable) to assess the deformation modulus 

of the rock core or the rock mass. The deformation modulus of intact rock, ER, is evaluated by 

performing uniaxial compression tests (ASTM D3148) on pieces of rock core obtained from drilling 

using a diamond core barrel. The stress-strain behavior of the intact rock can be measured in the 

laboratory and plotted as shown in figure 95. The axial and diameter strain are measured using very 

sensitive deformation instruments or, more commonly, using direct bond strain gages. Microstrain is 

defined where: 

6

dimension original

∆ original dimension

με 10 − = × (Equation 77) 

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The plots in figure 95 show two cycles of a compression test on a sample of strong and stiff gneissic 

rock that exhibits approximately perfectly elastic behavior with no hysteresis and no permanent 

deformation at the level of test deformations. 

The elastic constants calculated from the plots in figure 95 over the linear portion of the stress-strain 

curve include the Youngs modulus and Poisson ratio, and are calculated as follows (Wyllie, 1999): 

Young's modulus = Vertical stress/Vertical strain 

= 50.0 MPa / 610E – 6 

= 82 GPa (11.9 × 106

 psi) 

 Poisson's ratio = Diametral strain/Vertical strain 

= 150E – 6 / 610E - 6 

= 0.25 

Figure 95. Axial and diametral stress-strain curves for intact rock tested in uniaxial compression 

(after Wyllie, 1999, Foundations on Rock, Figure 3.3, p. 52, E&FN Spon). 

Table 41 shows the results of uniaxial compression tests carried out to determine the elastic 

constants of a variety of rock types (Lama and Vutukuri, 1978a, b). The Youngs modulus for intact 

rock is significantly larger than that for soil and typically varies between 0.1 to 100 GPa, but as can 

be seen from table 41, the values are rock formation specific. Calculated values of Poisson’s ratio 

typically range from 0.25 to 0.4, but these too are often rock specific. Typically, the settlement of a 

rock foundation will be controlled by the deformation modulus corresponding to the overall rock 

mass (see section 6.4.2) and will not be controlled by the deformation modulus of intact rock. 

However, the appropriate rock mass deformation modulus for use in design can be evaluated using 

procedures which reduce the intact rock modulus based on simple parameters such as joint opening 

size, RQD, and/or RMR, as discussed subsequently. 

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Table 41. Typical elastic constants for intact rock (after Wyllie, 1999) (see Wyllie, 1999 for original 

references for tests). 

Rock Type Young’s modulus 

GPa (psi × 106

) 

Poisson’s 

Ratio Reference 

Andesite, Nevada 37.0 (5.5) 0.23 Brandon (1974) 

Argillite, Alaska 68.0 (9.9) 0.22 Brandon (1974) 

Basalt, Brazil 61.0 (8.8) 0.19 Ruiz (1966) 

Chalk, USA 2.8 (0.4) - Underwood (1961) 

Chert, Canada 95.2 (13.8) 0.22 Herget (1973) 

Claystone, Canada 0.26 (0.04) - Brandon (1974) 

Coal, USA 3.45 (0.5) 0.42 Ko and Gerstle (1976) 

Diabase, Michigan 68.9 (10) 0.25 Wuerker (1956) 

Dolomite, USA 51.7 (7.5) 0.29 Haimson and Fairhurst (1970) 

Gneiss, Brazil 79.9 (11.6) 0.24 Ruiz (1966) 

Granite, California 58.6 (8.5) 0.26 Michalopoulos and Triandafilidis (1976) 

Limestone, USSR 53.9 (8.5) 0.32 Belikov (1967) 

Salt, Ohio 28.5 (4.1) 0.22 Sellers (1970) 

Sandstone, Germany 29.9 (4.3) 0.31 Van der Vlis (1970) 

Shale, Japan 21.9 (3.2) 0.38 Kitahara et al. (1974) 

Siltstone, Michigan 53.0 (7.7) 0.09 Parker and Scott (1964) 

Tuff, Nevada 3.45 (0.5) 0.24 Cording (1967) 

6.4.2 Rock Mass Modulus 

6.4.2.1 Method Based On Rock Mass Rating 

It is not possible to assess rock mass deformation characteristics in the laboratory since modulus 

values are highly dependent on the size of the sample. Bieniawski (1978) proposed a method of 

estimating the in situ rock mass modulus using the rock mass rating (RMR) previously described. 

The advantage of this approach is that RMR is evaluated from easily measured parameters (see table 

40). It is explicitly noted that discontinuity orientation will have an affect on the rock mass 

modulus. For example, an unfavorable gouge-filled joint orientation with respect to settlement 

would be in a direction at right angles to the load direction resulting in closure of discontinuities and 

settlement whereas for rock slope stability analyses, an unfavorable orientation with respect to 

sliding would be in a direction parallel to the load direction or inclined and dipping towards an open 

face. 

The empirical relationship between the RMR rating value and the in situ rock mass modulus is 

shown in figure 96. Based on a database review, the following equation for rock mass modulus, EM, 

exhibiting a RMR > 50 was developed (Bieniawski, 1978): 

GPaE RMR −= 1002)( M (Equation 78) 

Additional studies carried out on rock masses with RMR values ranging from approximately 20-85 

resulted in the following modification (Serafim and Periera, 1983): 

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40/)10( 10)( − = RMR

M GPaE (Equation 79) 

For rock masses with a RMR greater than approximately 80 (i.e., “very good rock” according to the 

CSIR classification system), the value for EM evaluated based on equation 79 should be compared to 

the deformation modulus of intact rock core as determined using ASTM D3148, and the lower value 

should be used for design calculations. 

Figure 96. Relationship between in situ modulus and rock mass rating (after Bieniawski, 1978; 

Serafim and Pereira, 1983). 

6.4.2.2 Method Based on RQD 

An alternate empirical method for evaluating the in situ rock mass modulus has been developed by 

Carter and Kulhawy (1988) and is widely used in practice to establish reliable estimates. This 

method uses the RQD value, an estimate of the intact rock modulus, ER, and a determination of 

whether the rock joints are open or closed to evaluate the rock mass modulus, EM (see table 42). In 

using table 42, values intermediate between tabulated entry values may be obtained by linear 

interpolation. It is necessary to estimate ER (see section 6.4.1) to establish a specific value of EM. 

Table 42. Estimation of EM based on RQD (modified after Carter and Kulhawy, 1988). 

EM/ER

RQD (in percent) Closed Joints Open Joints 

100 1.00 0.60 

70 0.70 0.10 

50 0.15 0.10 

20 0.05 0.05 

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6.4.2.3 Use of In situ Tests to Evaluate Rock Mass Modulus 

The methods previously described for estimating the in situ rock mass modulus based on RMR and 

RQD can be used for typical foundation and rock slope applications on highway projects. For very 

critical structures such as dams and tunnels, it may be necessary to perform in situ tests to “directly” 

measure the in-situ rock mass modulus. In chapter 4, specific in situ tests were introduced for 

assessing rock mass modulus. In this section, the evaluation of the in situ rock mass modulus using 

the borehole dilatometer and borehole jack is described. 

With all in situ tests there will be some disturbance of the rock, particularly where blasting must be 

used to prepare the site. The interpretation of the test should therefore recognize the potential for, 

and thus be able to assess the extent of this disturbance. Excavation of the foundation may also 

involve some disturbance to the rock and it is important to make an assessment of the degree of 

disturbance at the test site during testing compared with the likely condition at the foundation site at 

the time it is constructed. 

Borehole Dilatometer 

A typical borehole dilatometer curve for a test conducted in rock has been shown in figure 20. 

Using the borehole dilatometer, the shear modulus, Gd, and the modulus of elasticity, Ed, of the rock 

in the tested borehole are given by (ISRM, 1987): 

ρ

π 2 Ld = kG Rd (Equation 80) 

d GvE dR += )1(2 (Equation 81) 

where L is the length of the cell membrane; d is the diameter of the drill hole; νR is Poisson's ratio of 

the rock; and ρ is the pump constant defined as the fluid volume displaced per turn of pump wheel. 

The stiffness of rock over the length of the cell membrane, kR is: 

)/( )(turnMPa

kk

kk k

Ts

Ts

R − =

 (Equation 82) 

where ks its is the stiffness of the hydraulic system and kT is the stiffness of overall system plus rock 

(ratio D/C in figure 20). There are other versions of the borehole dilatometer that can be used and 

the details of these (as of the ISRM system) are beyond the scope of this document. It is, however, 

critical to note that when testing rock, several calibrations must be performed (see Wyllie, 1999). 

• The kR term is calculated from calibration of the hydraulic system and the results of a 

pressure-dilation test carried out in a calibration cylinder of known modulus. The stiffness of 

the hydraulic system, ks, is calculated from the stiffness of the calibration cylinder and the 

slope of the calibration pressure-dilation curve, km (ratio B/A in figure 20). 

• It is also necessary to make a correction for pressure losses due to the rigidity of the 

membrane. This is determined by inflating the dilatometer in the air without confinement to 

show the pressure required to inflate the membrane and the hydraulic system. It is usually 

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necessary to “select” membranes to match the stiffness of the tested rock. For example, if a 

very stiff membrane is used to perform a test in soft rock, the calibration will dominate the 

field response. 

• It is necessary to account for loss of volume in the hydraulic system that takes place in 

inflating and seating the membrane. 

Should it be necessary to conduct borehole dilatometer tests, it will likely require the services of an 

organization with specific borehole dilatometer experience and the specialized equipment. It is 

important to request the specifics on the equipment and assure that the calibration tests are provided. 

Borehole Jack 

Figure 97 shows a typical pressure-displacement plot for a borehole jack test carried out in strong 

limestone. Details of the borehole jack test and methods are provided in ASTM D4971. For the 

condition of full contact between the borehole jack and the rock sidewall and for a rock Poisson’s 

ratio of 0.25, the calculated in situ mass modulus, Ecalc is given by 

)/(

15.1

DD

Q E h

calc ∆

∆ = (Equation 83) 

where ∆Qh is the increment in hydraulic pressure, ∆D is the change in hole diameter resulting from 

∆Qh, and D is the hole diameter. Equation 83 will vary depending on the assumed Poisson’s ratio 

(see ASTM D4971 for modification based on Poisson’s ratio values other than 0.25), however the 

effect of a different Poisson’s ratio is relatively minor on the calculated value for Ecalc. Using the 

data in figure 97 for the third loading cycle (for a pressure increment of 30 to 40 MPa), the measured 

diametral displacement is approximately 0.09 mm. With a hole diameter of 76.2 mm, equation 83 is 

used to calculate an Ecalc of 9.7 GPa. For the borehole jack, the calculated value of rock modulus 

may need to be corrected to account for the variation in the ratio between the modulus of steel and 

the modulus of the intact rock. When the modulus of the steel is much greater than that of the intact 

rock, the correction factor is negligible because there is little deformation of the steel platens as the 

pressure is applied. However, when the intact rock modulus is high compared with the steel 

modulus, the modulus value calculated from the jack test is less than the “true” rock modulus and 

the “true” in situ mass modulus, Etrue should be evaluated using figure 98. For a calculated modulus 

of 9.7 GPa, figure 98 indicates the Etrue is approximately 12 GPa. 

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Figure 97. Pressure-displacement plot for borehole jack. 

Figure 98. Curve of Etrue versus Ecalc (after ASTM D 4971). 

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6.4.3 Selection of Rock Deformation Modulus for Design 

Various methods have been described for evaluating the deformation modulus in rock masses. For 

most highway design applications in which deformation evaluations are required, however, the 

empirical methods described based on either RMR or RQD can be used. As indicated, the use of insitu test methods to evaluate a deformation modulus is typically only performed for critical or large 

structures such as major bridges or dams. Alternatively, if design analyses based on the use of either 

RMR or RQD result in predicted settlements that exceed target values for a particular foundation 

type, then it may be appropriate to perform in situ tests to develop a more accurate assessment of the 

in situ modulus. The most important consideration is to use the rock mass characteristics, not the 

characteristics of the intact rock core. 

For shallow foundations, the average rock mass modulus over a depth equal to approximately 2 

times the width of the foundation should be assessed. This can be done by measuring the intact rock 

modulus via uniaxial compression tests on samples of intact core from successive core samples and 

then evaluating the rock mass modulus based on table 42. 

The selection of modulus values for a drilled shaft will depend on whether the shaft design is for end 

bearing only, side resistance only from a rock socket, or a combination of both. Many methods exist 

for evaluating deformations in drilled shafts and some are based on specific semi-empirical methods 

of evaluating rock mass modulus values based on the results of settlement data for previously 

constructed drilled shafts in particular rock types (see FHWA-IF-99-025, 1999). In general, 

however, it is recommended to use the procedures outlined in section 6.4.2 to obtain an estimate of 

the rock mass modulus. For drilled shafts with rock sockets in which load is carried in end-bearing 

and side resistance, an evaluation of the average rock mass deformation modulus around the rock 

socket and at the base of the shaft (at a depth of up to 2 times the diameter of the shaft below the 

base of the shaft) are required. Intact rock cores can be tested in uniaxial compression and then 

corrected to a rock mass modulus using table 42. The deformation modulus value can also be 

evaluated based on RMR ratings for rock near the shaft base and around the socket using figure 96. 

6.5 ROCK SHEAR STRENGTH 

6.5.1 Mohr-Coulomb Materials 

For geotechnical design analyses, rock (i.e. intact rock and jointed rock masses) is generally 

assumed to be a Mohr-Coulomb material in which the shear strength of the rupture surface is 

expressed in terms of the cohesion intercept (c) and the friction angle (φ). When an effective normal 

stress σ' acts on the rupture surface, the shear stress (τ) developed is: 

tanφστ += c ' (Equation 84) 

Figure 99 illustrates the relationship between the typical strength parameters for five geological 

conditions. A description of these conditions is as follows: 

• Curve 1. Infilled discontinuity: If the infilling is a weak clay or fault gouge, the 

friction angle of the infilling is likely to be low, and there may be some cohesion if the 

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infilling is undisturbed. Alternatively, if the infilling is strong (e.g. calcite which produces a 

healed surface) then the cohesive strength and the friction angle may be significant. 

• Curve 2. Smooth discontinuity: A smooth, clean discontinuity in the parent rock will 

have little or no cohesion, and the friction angle will be that of the parent rock. The friction 

angle of rock is related to the rock grain mineralogy and size; the friction angle is generally 

lower in fine-grained rocks than in coarse-grained rocks. 

• Curve 3. Rough discontinuity: Clean, rough discontinuity surfaces will have little or 

no cohesion, and the friction angle will be made up of the rock material friction angle (φ), 

and a component related to the roughness, or asperities, of the surface (see section 6.5.2.3). 

• Curve 4. Fractured rock mass: The shear strength of a fractured rock mass, in which 

the rupture surface lies partially on discontinuity surfaces and partially passes through intact 

rock, can be expressed as a curved envelope. At low normal stresses where individual 

fragments may move and rotate, the cohesion is low but the friction angle is high because 

the sliding surface is effectively rough and may involve non-frictional material. In addition, 

at the low normal stresses, the response may be dilative. At higher normal stresses, dilation 

is reduced and crushing of the rock fragments begins to take place and the friction angle 

reduces. The shape of the strength envelope is related to the degree of fracturing and the 

strength of the intact rock. 

• Curve 5. Weak intact rock: The shear strength of weak, but intact, rock is governed by 

the combination of cohesion and parent rock friction. Some rock (e.g. tuff) comprises finegrained material that has a low friction angle. However, due to its intact nature, the 

cohesion can be higher than that of a closely fractured well indurated rock. 

In this section, a discussion of the shear strength behavior of infillings, smooth and rough 

discontinuities, and fractured rock masses are presented. Following this discussion, the methods 

commonly used to evaluate shear strength for these rock classes are presented. The direct shear 

testing method (laboratory or in situ) is commonly used to evaluate the shear strength of infillings 

and discontinuity surfaces. Shear strength for fractured rock masses is evaluated using back analysis 

methods or semi-empirically based on characterization information such as RMR and rock type. 

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Figure 99. Relationships between shear stress and normal stress on rupture 

surface for five different geological conditions (TRB, 1996). 

6.5.2 Shear Strength Of Discontinuities 

6.5.2.1 General 

If discontinuities are identified within a rock mass, it will be necessary to evaluate the friction angle 

and cohesion intercept along the potential rupture surface to perform design analyses. The 

investigation program should also obtain information on important fracture characteristics such as 

continuous length, surface roughness, thickness and characteristics of any infilling material, water 

occurrence, as well as the effect of water on the properties of the infilling. Shear strength properties 

of various discontinuity types are described in this section. 

6.5.2.2 Friction Angle of Rock Surfaces 

For a planar, clean (no infilling) discontinuity, the shear strength will be defined solely by the 

friction angle. The friction angle of the rock material is related to the size and shape of the grains 

exposed on the fracture surface and the rock mineralogy. Thus, a fine-grained rock and/or rock with 

a high mica content aligned parallel to the surface will tend to have a low friction angle, while a 

coarse-grained rock will have a high friction angle. Table 43 shows typical ranges of friction angles 

for a variety of rock types (Barton, 1973; Jaeger and Cook, 1976). The friction angles listed in table 

43 should be used as a guideline only because actual values will likely vary from site to site, for a 

given rock type, and may vary slightly across a single site. 

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Table 43. Typical ranges of friction angles for a variety of rock types (after Barton, 1973; Jaeger 

and Cook, 1976). 

Rock Class Friction Angle Range(1) Typical Rock Types 

Low Friction 20 to 27° Schists (high mica content), shale, marl 

Medium Friction 27 to 34° Sandstone, siltstone, chalk, gneiss, slate 

High Friction 34 to 40° Basalt, granite, limestone, conglomerate 

Note: (1) Values assume no infilling and little relative movement between joint faces. 

6.5.2.3 Surface Roughness 

Surface irregularities or asperities produce interlock between discontinuity surfaces that increase the 

resistance to sliding (or shearing). Asperities can be considered as a series of saw teeth. When 

normal and shear forces are applied to a rock surface containing a clean, saw tooth fracture, the 

shear strength of the fracture surface is defined as follows: 

)(tan ++= ci ' φστ (Equation 85) 

where i is the inclination of the saw teeth as shown in figure 100. This relationship shows that the 

effective friction angle of a rough surface is equal to the sum of the friction angle of the rock and the 

inclination of the asperities. Another way of viewing this increased strength due to asperities is that 

it requires work for the rock mass to dilate and move over the asperities. 

Figure 100. Effect of surface roughness and normal stress on the friction of a discontinuity surface 

(after Wyllie, 1999, Foundations on Rock, Figure 3.15, p. 70, E&FN Spon). 

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During shearing, the asperities can be sheared off, resulting in a reduction in the friction angle. With 

increasing stress levels, there is a transition from dilation to shearing. At this point, the degree to 

which the asperities are sheared off will depend on both the magnitude of the normal force in 

relation to the compressive strength of the rock on the fracture surface, and the displacement 

distance. A rough surface that is initially undisturbed and interlocked will have a friction angle of (φ

+ i), which represents the peak shear strength. With increasing normal stress and displacement, the 

asperities will be sheared off, and the friction angle will progressively reduce to a minimum value, 

or residual, friction angle of the rock, resulting in a curved failure envelope as shown in figure 100. 

To quantify the relationship between the total friction angle (φ + i), the rock strength, and the normal 

stress acting on the rupture surface, Barton (1976) studied the shear strength behavior of artificially 

produced rough, clean "joints". The study showed that the shear strength of a rough rock surface can 

be defined by the following empirical equation 

tan[ )'/(log ] 10

' φστ += JRC JCS σ (Equation 86) 

where: JRC = joint roughness coefficient; JCS = compressive strength of the rock adjacent to the 

fracture surface; and σ′ = effective normal stress. 

The roughness of the fracture surface is defined by the joint roughness coefficient, JRC. Barton 

carried out direct shear tests on a large number of natural discontinuities and calculated JRC values 

corresponding to the surface roughness of the different shear test specimens. From these tests a set 

of typical roughness profiles with specified JRC values were prepared (figure 101). By comparing a 

fracture surface with these standard profiles, the JRC value can be evaluated (Barton, 1976). 

The term JRC log10(JCS/σ′) is equivalent to the inclination of the sawteeth or the roughness angle i 

in equation 85. At high stress levels when JCS/σ′ = 1 and the asperities are sheared off, the term 

JRC log10(JCS/σ′) equals zero. At low stress levels, the ratio JCS/σ′ becomes very large and the 

roughness component of the strength also becomes very large. In order that realistic values of the 

roughness component are used in design, the fracture surface (i.e. joint, etc.) should be visually 

inspected and the surface compared to the surfaces shown in figure 101. Regardless of the 

calculated values, for design the term (φ+i) should not exceed 50 degrees and the useful range for 

the ratio JCS/σ′ is between 3 and 100. As more progressive displacement or relative movement 

occurs across the fracture surface, it should be recognized that the value (φ+i) reduces. 

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Figure 101. Definition of joint roughness coefficient, JRC (Barton, 1973). 

6.5.2.4 Measurement of Surface Roughness 

The method described in section 6.5.2.3 to evaluate the roughness angle (i) is typically used during 

preliminary design or where exposures of discontinuity surfaces are not available at the project site. 

If exposures of discontinuity surfaces on which sliding may occur are available at a project site, 

measurements can be made of the surface roughness. If the exposure is extensive, then the actual 

roughness angle (i) can be measured and used in equation 85 in evaluating the shear strength of the 

surface. Methods of making these measurements are beyond the scope of this document, but 

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information can be found in FHWA-HI-99-007 (1998) and Wyllie (1999). In critical cases, the joint 

surface can be exhumed and a laboratory direct shear test can be performed at the range of projectrepresentative normal stresses to physically assess the strength of the rock along the fracture surface. 

6.5.2.5 Discontinuity Infilling 

If a discontinuity contains an infilling, the shear strength properties of the fracture will be influenced 

by the thickness and properties of the infilling. For example, for a relatively thick clay-filled fault 

zone in granite, the clay and not the granite would control the shear strength of the discontinuity; if 

the gouge in fairly thin and the granite surfaces are in contact along much of the fracture, the 

strength value will be intermediate between that of the parent rock plus asperities and the gouge 

material. In the case of a healed, calcite-filled fracture, a high cohesion would be used in design if 

the discontinuity remained healed after any construction-induced disturbance. With respect to the 

thickness of an infilling, if it is more than about 25 to 50 percent of the amplitude of the asperities, 

there will be little or no rock-to-rock contact, and the shear strength properties of the fracture will be 

controlled by the shear strength properties of the infilling (Goodman, 1970). 

Infillings are generally divided into the following two groups: 

• Clays – For montmorillonite and bentonitic clays, and clays encountered during coal 

mining, tests have been conducted that demonstrate friction angles ranging from about 8 to 

20 degrees and cohesion intercept values ranging from 0 to about 200 kPa. 

• Faults, shears and breccias – For these rocks, the material formed in fault zones and shears 

in rocks such as granite, diorite, basalt, and limestone will typically contain clay as well as 

granular fragments. The gouge materials associated with the rock types typically have 

friction angles ranging from about 25 to 45 degrees, and cohesion intercept values ranging 

from 0 to about 100 kPa. Fault gouge derived from coarse-grained rocks such as granites 

tend to have higher friction angles than those from fine-grained rocks such as limestones. 

As part of a Barton (1976) study, it was found that the residual friction angle of filled discontinuities 

was about 2 to 4 degrees less than the peak friction angle, while the residual cohesion intercept was 

essentially zero. 

Filled discontinuities can be divided into two general categories, depending on whether there has 

been previous displacement of the discontinuity (Barton, 1974). These categories are further 

subdivided into either normally-consolidated (N-C) or over-consolidated (O-C) materials (figure 

102): 

• Recently displaced discontinuities - These types of discontinuities include faults, shear 

zones, clay mylonites, and bedding-surface slips. In faults and shear zones, the infilling is 

formed by the shearing process which may have occurred many times and produced 

considerable displacement. The gouge formed in this process may include both clay-size 

particles, and breccia with the particle orientation and striations of the breccia aligned 

parallel to the direction of shearing. In contrast, the mylonites and bedding-surface slips are 

discontinuities that were originally clay bearing, and along which slip occurred during 

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218 

folding or sliding. For these types of discontinuities their shear strength will likely be at or 

close to the residual strength (figure 102, graph I). 

• Undisplaced discontinuities - Infilled discontinuities that have undergone no previous 

displacement include igneous and metamorphic rocks that have weathered along the 

discontinuity to form a clay layer. For example, diabase usually weathers to amphibolite 

and eventually to clay. Other undisplaced discontinuities include thin beds of clay and weak 

shales that are found with sandstone in interbedded sedimentary formations. Hydrothermal 

alteration may form infillings that can include low strength materials such as 

montmorillonite, and high strength materials such as quartz and calcite. 

The infillings of undisplaced discontinuities can be divided into normally consolidated (N-C) and 

overconsolidated (O-C) materials that have significant differences in peak strength values. This 

strength difference is illustrated on figure 102 in graphs II and III. While the peak shear strength of 

over-consolidated clay infillings may be high, there can be a significant loss of strength due to 

softening, swelling, and pore pressure changes on unloading. Unloading occurs when rock is 

excavated for a slope or foundation, for example. Strength loss also occurs on displacement in 

brittle materials such as calcite. 

For these conditions, it is important that the design assess the type of infilling and the degree of 

potential displacement along the discontinuity. As shown in table 44, the effects of the type of 

infilling and parent rock can have a significant effect on strength parameters. 

6.5.2.6. Effect of Water on Shear Strength 

The most significant effect of ground water on slope stability is the reduction in the effective normal 

stress produced by water pressures acting within the slope. A reduction in the effective normal 

stress results in a corresponding reduction in the shear strength of the rupture surface. Generally, the 

strength properties of rock are the same whether they are wet or dry; however some rocks, such as 

shales containing swelling clay, exhibit a loss of strength in the presence of water. Therefore, it is 

necessary to assess whether the infilling material is adversely affected by the occurrence of water. 

To this end, it is important to note that in some cases the in situ condition has tight and relatively dry 

fractures. However, due to construction, there could be sufficient movement to allow a pathway for 

water to become introduced to the rock and the rock infill material. 

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219 

Figure 102. Simplified division of filled discontinuities into displaced and undisplaced, and NC and 

OC categories (after Wyllie, 1999, Foundations on Rock, Figure 3.18, p. 73, E&FN Spon). 

Archived Document
Table 44. Shear strength of filled discontinuities (modified after Hoek and Bray, 1977) (Note: 

missing data implies shear strength not reported in original table; see Hoek and Bray, 1977 for 

references for testing data). 

220 

Peak Strength Residual Strength 

Rock Type Description c′ (kPa) φ′ c′ (kPa) φ′

Igneous Rocks

Basalt 

Diorite 

Granite 

Clayey basaltic breccia, wide variation 

from clay to basalt content 

Clay gouge (2 percent clay, PI=17) 

Clay-filled faults 

Weakened with sandy-loam fault 

filling 

Tectonic shear zones, schistose and 

broken granites, disintegrated 

rock and gouge 

240 

0 

0 to 100 

50 

242 

42 

26.5 

24 to 45 

40 

42 

 

Metamorphic Rocks

Schists, quartzites, 

and siliceous schists 

Slates 

10- to 15-cm thick clay filling 

Stratification with thin clay 

Stratification with thick clay 

Finely laminated and altered 

30 to 80 

610 to 740 

380 

50 

32 

41 

31 

33 

 

Sedimentary Rocks

Dolomite 

Limestone 

Limestone, marl, and 

lignites 

Limestone 

Sandstone 

(Graywacke) 

Strata containing coal 

Lignite 

Altered shale bed, approximately 15-

cm thick 

6 mm clay layer 

Interbedded lignite layers 

lignite/marl contact 

Marlaceous joints, 2-mm thick 

1 to 2-mm thick clay in bedding layer 

Clay mylonite seams, 1 to 2.5 cm 

thick 

Layer between lignite and underlying 

clay 

41 

80 

100 

0 

11 to 13 

14 to 30 

14.5 

38 

10 

25 

16 

15 to 17.5 

22 

0 

0 

0 

0 

17 

13 

15 to 24 

21 

11 to 11.5 

Archived Document
Table 44. Shear strength of filled discontinuities (modified after Hoek and Bray, 1977) (Note: 

missing data implies shear strength not reported in original table; see Hoek and Bray, 1977 for 

references for testing data). 

221 

Peak Strength Residual Strength 

Rock Type Description c′ (kPa) φ′ c′ (kPa) φ′

Sedimentary Rocks 

(clays and shales)

Bentonite 

Bentonitic shale 

Clays 

Clay Shale 

Clay Shale 

Montmorillonite Clay 

Bentonite seam in chalk 

Thin layers 

Triaxial tests 

Triaxial tests 

Direct shear tests 

Overconsolidated slips, joints, and 

minor shears 

Triaxial tests 

Stratification surfaces 

Montmorillonite 

8 cm seams of montmorillonite clay in 

chalk 

15 

90 to 120 

60 to 100 

0 to 270 

0 to 180 

60 

360 

16 to 20 

7.5 

12 to 17 

9 to 13 

8.5 to 29 

12 to 18.5 

32 

14 

7.5 to 11.5 

30 

0 to 3 

0 

80 

8.5 

10.5 to 16 

19.5 

11 

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222 

6.5.2.7 Laboratory Direct Shear Testing 

In a laboratory direct shear test on rock, the measured friction angle is the sum of the friction angle 

of the rock (φ), and the roughness of the surface (i) at each normal stress level. The roughness of the 

surface is calculated from the plots of shear and normal displacement (δs and δn, respectively, on 

figure 103(a) as follows: 





 



 = −

s

n i

δ

1 δ tan (Equation 87) 

Figure 103. Results of direct shear test of filled discontinuity showing measurements of shear 

strength and roughness (after Wyllie, 1999, Foundations on Rock, Figure 3.20, p. 75, E&FN Spon). 

This value of i is the shear strength component due to dilation of the rock upon shear. This value is 

Archived Document
223 

subtracted from the friction angle calculated from the plot of shear and normal stresses at failure 

(considering peak conditions) to obtain the friction angle of the rock. For example, figure 103(a) 

indicates that the peak strength is 1.02 MPa. For a normal stress of 1.32 MPa, this results in a 

friction angle of 37.7°. Therefore, for a dilation angle of 1.6°, the friction angle of the parent rock is 

36.1°. This friction angle is actually an “instantaneous” friction angle and represents the tangent to 

the strength envelope at a normal stress of 1.32 MPa. 

In laboratory direct shear testing on rock, it is typical to test each sample at a minimum of three 

normal stress levels, with the sample being reset to its original position between tests. When the 

tests are run at progressively higher normal stress levels, the total friction angle of the surface will 

reduce with each test if the asperities are progressively sheared. The maximum design load should 

be considered when establishing the normal stresses as used in the test. 

It can be difficult to measure the cohesion of a surface with the direct shear test. For cases where the 

cohesion is very low, it may not be possible to obtain an undisturbed sample so the cohesion may be 

reduced or lost during sampling. If the cohesion is high and the sample is intact, the material 

holding the sample will have to be stronger than the infilling material if the sample is to shear. 

Where it is important that the cohesion of a weak infilling be measured, an in situ direct shear test of 

the undisturbed material may be required. 

6.5.3 Shear Strength Of Fractured Rock Masses 

6.5.3.1 General 

A rupture surface in fractured rock for which there is no distinct discontinuity surface comprises 

both natural discontinuities aligned on the rupture surface, together with some shear failure through 

intact rock. It is difficult and expensive to sample and test large samples of fractured rock, therefore, 

two empirical methods of evaluating the friction angle and cohesion intercept of fractured rock 

masses have been developed and are described in this section. In both methods, it is necessary to 

categorize the rock mass in terms of both the intact rock strength and the characteristics of the 

discontinuities. It is advisable to compare the strength values obtained by both methods to improve 

the reliability of values used in design. 

6.5.3.2 Strength Determination by Back Analysis of Failures 

The most reliable method of evaluating the strength of a rock mass is to back analyze a failed, or 

failing, slope. This involves carrying out a stability analysis with the factor of safety set at 1.0 and 

using available information on the position of the rupture surface, the groundwater conditions at the 

time of failure and any external forces such as foundation loads and earthquake motion, if 

applicable. Additional information on performing back analyses as a means to evaluate fractured 

rock mass shear strength is provided in FHWA-HI-99-007 (1998). 

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224 

6.5.3.3 Hoek-Brown Strength Criteria for Fractured Rock Masses 

The shear strength of fractured rock masses can also be evaluated using the method developed by 

Hoek (1983) and Hoek and Brown (1988, 1997) in which the shear strength is represented as a 

curved envelope (figure 104). The three parameters defining the curved strength envelope of the 

fractured rock mass are the uniaxial compressive strength of the intact rock, σc, and two 

dimensionless constants m and s. The values of m and s are defined in table 45. The constants m

and s depend on the rock type and the degrees of fracturing of the rock mass. In table 45, rock types 

are grouped into five classes depending on the grain size and crystal structure. The six qualities of 

the rock mass are defined in table 45 either by the brief descriptions of the geology or by rock mass 

rating (RMR) that assign point scores to a number of characteristics of the rock mass (see table 45). 

Figure 104. Typical curved shear strength envelope defined by Hoek-Brown theory for rock mass 

strength (Hoek, 1983). 

The equation for the curved shear strength envelope is: 

8 )cos(cot ' ' c

i i

mσ −= φφτ (Equation 88) 

where τ is the shear stress at failure, σc is the uniaxial compressive strength of the intact rock and φi′

is the instantaneous friction angle at given values of τ and σ′. The value of φi′ is the inclination of 

the tangent to the Mohr failure envelope at the point (σ′, τ) as shown in figure 104 and is given by: 

( [ ] ) 21 ' 1 2 231 1)(sin33.030cos4tan − − −− φ i = h + h − (Equation 89) 

where 

c

c

m

sm h

σ

σσ

2

'

3

)(16 1 +

+= (Equation 90) 

Archived Document
225 

The instantaneous cohesion (ci′) is the intercept of the line defining the friction angle on the shear 

stress axis and is given by: 

' ''

i tan i c −= φστ Equation (91) 

Table 45. Approximate relationship between rock-mass quality and material 

constants used in defining nonlinear strength (Hoek and Brown, 1988). 

Empirical failure criterion 

σ′1 = σ′3 + (mσc σ′3 + sσc

2

)

1/2

σ′1 = major principal effective stress

σ′3 = minor principal effective stress 

σc = uniaxial compressive strength of intact 

rock, and m and s are empirical 

constants 

Rock Quality 

CARBONATE ROCKS WITH WELL DEVELOPED CRYSTAL CLEAVAGE – dolomite, limestone and marble 

LITHIFIED ARGRILLACEOUS ROCKS – mudstone, siltstone, shale and slate (normal to cleavage) ARENACEOUS ROCKS WITH STRONG CRYSTALS AND POORLY DEVELOPED CRYSTAL CLEAVAGE – sandstone and quartzite FINE GRAINED POLYMINERALLIC IGNEOUS CRYSTALLINE ROCKS – andesite, dolerite, diabase and rhyoliteCOARSE GRAINED POLYMINERALLIC IGNEOUS & METAMORPHIC CRYSTALLINE ROCKS – amphibolite, gabbro gneiss, granite, norite, quartz-diorite 

INTACT ROCK SAMPLES 

Laboratory size specimens free from 

discontinuities 

CSIR rating(1): RMR = 100

m 

s 

7.00 

1.00 

10.00 

1.00 

15.00 

1.00 

17.00 

1.00 

25.00 

1.00 

VERY GOOD QUALITY ROCK MASS 

Tightly interlocking undisturbed rock with 

unweathered joints at 1 to 3 m 

CSIR rating: RMR = 85

m 

s 

2.40 

0.082 

3.43 

0.082 

5.14 

0.082 

5.82 

0.082 

8.567 

0.082 

GOOD QUALITY ROCK MASS 

Fresh to slightly weathered rock, slightly 

disturbed with joints at 1 to 3 m 

CSIR rating: RMR = 65 

m 

s 

0.575 

0.00293 

0.821 

0.00293 

1.231 

0.00293 

1.395 

0.00293 

2.052 

0.00293 

FAIR QUALITY ROCK MASS 

Several sets of moderately weathered joints 

spaced at 0.3 to 1 m 

CSIR rating: RMR = 44 

m 

s 

0.128 

0.00009 

0.183 

0.00009 

0.275 

0.00009 

0.311 

0.00009 

0.458 

0.00009 

POOR QUALITY ROCK MASS 

Numerous weathered joints at 30-500 mm; 

some gouge. Clean compacted waste rock. 

CSIR rating: RMR = 23 

m 

s 

0.029 

0.000003 

0.041 

0.000003 

0.061 

0.000003 

0.069 

0.000003 

0.102 

0.000003 

VERY POOR QUALITY ROCK MASS 

Numerous heavily weathered joints spaced < 

50 mm with gouge. Waste rock with fines. 

CSIR rating: RMR = 3 

m 

s 

0.007 

0.0000001 

0.010 

0.0000001 

0.015 

0.0000001 

0.017 

0.0000001 

0.025 

0.0000001 

Notes: (1) CSIR = Council of Scientific and Industrial Research (Bieniawski, 1974 and table 40). 

Archived Document
226 

 

The features of the curved shear strength envelope are that, at low normal stress levels, the blocks of 

rock are interlocked and the instantaneous friction angle is high, whereas at higher normal stress 

levels, shearing of the rock is initiated with the result that the friction angle reduces. The 

instantaneous cohesion progressively increases in direct proportion to the normal stress as a result of 

the greater confinement and interlock of the rock mass. 

The procedure for using the curved strength envelopes in a stability analysis is first to evaluate the 

range of effective normal stresses acting along a potential rupture surface in the slope, and second to 

calculate the instantaneous cohesion values and friction angles (ci′, φi′) in this stress range. The 

stability analysis is carried out in a traditional stability calculation, except that values of ci′ and φi′

are used corresponding to the variation in normal stress along the sliding surface. Many commercial 

limit equilibrium stability analysis programs enable the user to input the actual shear stress-normal 

stress curve when performing the analysis. 

Figure 105 illustrates the application of the information in table 45 for three different conditions. 

These curves cover the range from very poor quality rock mass to very good quality rock mass. 

Table 46 shows values for the parameters used to develop figure 105. The reader of this document 

can verify the correct implementation of equations 88 to 91 by comparing to values on table 46. 

0

500

1000

1500

2000

2500

3000

3500

4000

4500

5000

-200 0 200 400 600 800 1000 1200

NORMAL STRESS (kPa)

SHEAR STRESS (kPa)

m=0.007; s=0.0000001; 

comp. strength =20 MPa

m=1.231; s=0.00293;

comp. strength = 60 MPa

m=5.14; s=0.082; comp.

strength = 80 MPa

Figure 105. Illustration of use of nonlinear shear strength for three fractured rock mass types. 

Archived Document
227 

Table 46. Parameters used to develop strength envelopes in figure 105. 

Curve 1 

m=0.007; s=0.0000001; 

σc=20 MPa 

Curve 2 

m=1.231; s=0.00293; 

σc=60 MPa 

Curve 3 

m=5.14; s=0.082; 

σc=80 MPa 

σ′

(kPa) 

h φ′i c′i 

(kPa

) 

τ

(kPa

) 

h φ′i c′i 

(kPa

) 

τ

(kPa) 

h φ′i c′i 

(kPa

) 

τ

(kPa

) 

-1,000 - - - - - - - - 1.004 70.5 3,868 1,049

-500 - - - - - - - - 1.010 65.0 3,326 2,254

0 1.011 64.5 0.8 0.8 1.010 64.8 412.0 412.0 1.017 61.9 3,250 3,250

100 4.820 15.5 18.2 46.1 1.017 61.5 424.8 608.9 1.018 61.4 3,252 3,435

170 7.487 12.3 26.1 63.2 1.023 59.8 441.5 733.1 1.019 61.0 3,255 3,562

250 10.535 10.3 33.6 79.0 1.028 58.1 463.8 865.8 1.020 60.7 3,261 3,705

300 12.439 9.5 37.7 87.7 1.032 57.2 478.6 944.8 1.020 60.4 3,265 3,794

1,000 39.106 5.3 78.4 171.2 1.083 49.6 699.9 1874.6 1.030 57.8 3,373 4,963

6.5.4 Selection of Rock Shear Strength for Design 

This section provides a summary of the information presented on the evaluation of shear strength for 

discontinuities, infillings, and fractured rock masses and some basic recommendations on selection 

of shear strength for design. 

Smooth Discontinuity 

As discussed, the shear strength for a smooth discontinuity is represented by a friction angle of the 

parent rock material. To evaluate the friction angle of this type of discontinuity surface for design, 

direct shear tests on samples can be performed. Samples can be formed in the laboratory by cutting 

samples of intact core. The range of typical friction angles provided in table 43 should be used in 

evaluating measured values. 

Rough Discontinuity 

The shear strength of a discontinuity surface is related to the roughness of the discontinuity. The 

magnitude of additional strength resulting from roughness for a given normal load will depend on 

the roughness profile and the compressive strength of the joint wall. The term JRC log10(JCS/σ′) 

can be used to obtain an estimate of the joint roughness angle, i, for cases where the normal stresses 

for design are within the range of 1 to 30 percent of the joint compressive strength. These stress 

ranges are common for many rock slopes. For values less than 1 percent or greater than 30 percent, 

this procedure is not valid. As previously noted, the maximum value for (φ+i) should be 50° and the 

minimum value is simply the friction angle of the intact rock corresponding to stress and 

deformation levels sufficiently large to have sheared through all asperities (i.e., zero dilation). A 

typical value for the basic rock friction (i.e., for a smooth discontinuity) is 30°. 

Archived Document
228 

It is noted that for very critical structures, it may be necessary to perform in situ direct shear tests to 

estimate the shear strength of a discontinuity surface (regardless of whether the surface is smooth, 

rough, or infilled). Moreover, designers should recognize that accurate measurement of the 

roughness angle is difficult. Therefore, if design analyses indicate that a relatively large, although 

seemingly achievable, value for i is required to demonstrate a stable slope (or other structure), than 

more detailed field measurements of surface roughness (as compared to that obtained using figure 

100) are necessary to confirm the value for i. For these critical cases, additional effort should be 

expended to assess the orientation of the discontinuity in relation to design conditions. 

Infillings 

When a major discontinuity with a significant thickness of filling is encountered in a rock mass in 

which a slope is to be excavated or where the discontinuity is adversely dipping, the designer should 

assume that shear failure would occur through this material. Figure 102 presents a general method 

of characterizing joint fillings according to whether the filling material is undisplaced or displaced 

and whether the filling material is normally consolidated or overconsolidated. For fillings that are 

displaced, it should be assumed that the material is at or near residual strength conditions in the 

field. The residual strength of the material can be evaluated in the laboratory using a direct sheartesting device in which samples of the filling material are reconstituted into the direct shear-testing 

device at the in situ moisture content. If the filling is judged to be undisplaced, then the designer 

needs to evaluate whether the proposed structure can be expected to undergo minimal deformations 

such that the peak strength of the infilling can be used in design. If it becomes critical to evaluate 

the peak strength of the infilling (including both friction angle and cohesion intercept), it will likely 

be necessary to perform laboratory or in situ direct shear tests of the infilling material. If laboratory 

testing is selected, great care must be exercised in obtaining samples for testing. The sample halves 

should not be displaced relative to each other and the samples should be sealed after collection to 

minimize moisture losses. 

Fractured Rock Masses 

In conditions where a fractured rock mass can be described using the RMR, the shear strength can be 

evaluated using equations 88 to 91. In all cases, however, and for shear strength considerations for 

smooth discontinuities, rough discontinuities, infillings, and weak rock, the designer should perform 

a detailed review of previous reports on shear strength evaluations performed in similar geologic 

environments. Also, the design engineer should perform a sensitivity study to evaluate whether 

small changes in shear strength (i.e., friction angle, cohesion intercept, roughness angle) result in 

relatively large changes in a calculated factor of safety. If this is the case, it may be prudent to 

reevaluate the investigation and testing program and perform high quality laboratory and/or in situ 

tests. 

Archived Document
229 

CHAPTER 7 

INTERPRETATION OF PROPERTIES FOR SPECIAL MATERIALS 

7.1 INTRODUCTION 

This section of the document provides information on the evaluation of properties for several 

“special” soil and rock deposits that may be encountered on typical transportation projects. The 

special deposits discussed in this section include: (1) loess; (2) expansive soils; (3) organic soils and 

peat; (4) colluvium and talus; (5) shales and degradable materials; (6) cemented sands; (7) sensitive 

clays; and (8) partially saturated soils. A summary of sampling/testing difficulties and engineering 

characteristics associated with these materials is provided in table 47. The remainder of this chapter 

provides specific guidance on the evaluation of index and engineering properties for these special 

materials. 

The laboratory and in-situ test methods described previously can be used for the special deposits 

described herein, subject to various practical limitations. These limitations may include difficulty in 

obtaining “undisturbed” samples for laboratory testing and problems associated with laboratory 

testing of partially saturated soils. Because the methods described in chapter 5 generally can be 

modified to evaluate properties for the special materials discussed herein, the most important aspect 

of property characterization in these special materials is the ability to identify them. 

Table 47. Summary of sampling difficulties and engineering characteristics of special materials. 

Soil Description Soil Identification Sampling and Testing 

Difficulties Engineering Characteristics 

Loess and 

Collapsible Soils 

wind-blown sand, silt, 

and clay with weak binder 

(loess); silty clays 

exhibiting loose structure 

and weak interparticle 

bonds (collapsible soils) 

low moisture content and unit 

weight, low SPT N values, 

sensitive structure that may 

collapse during undisturbed 

sampling, in-situ testing using CPT 

and DMT good for in-situ 

properties 

erodible, irrecoverable 

collapse upon inundation, 

need to assess material 

characteristics at design 

moisture contents 

Expansive Soils 

clay-rich soils in arid and 

semi-arid regions that are 

subject to wet/dry and 

freeze/thaw cycles 

resulting in deep 

desiccation cracking 

low moisture content when dry and 

heavily fissured; difficult to obtain 

undisturbed sample 

undergoes significant 

recoverable volume changes 

(both shrinkage and swelling) 

when moisture content 

changes; need to assess range 

of moisture content variation 

and depth of seasonal 

moisture content change for 

design 

Organic Soils 

and Peat 

high moisture content 

relative to plasticity and 

fibrous texture; high loss 

of mass upon heating 

b

samples may be very soft and 

fibrous and difficult to recover; 

in-situ test may over estimate 

strength due to fibrous nature of 

i l

highly compressible and 

subject to high secondary 

compression; potentially 

corrosive 

Archived Document
Table 47. Summary of sampling difficulties and engineering characteristics of special materials 

(continued). 

230 

Soil Description Soil Identification Sampling and Testing 

Difficulties Engineering Characteristics 

above 440oC material 

Colluvium and 

Talus 

weathered materials that 

migrate and accumulate 

on the sides and at the toe 

of slopes; fine grained 

material with rock 

fragments (colluvium); 

coarse grained material 

with boulders (talus) 

Colluvium: thin lenses of material 

likely; rock fragments make 

sampling difficult; test pits and 

trenches helpful 

Talus: boulders preclude invasive 

investigation techniques; 

geophysical techniques including 

Spectral Analysis of Surface 

Waves (SASW) can be used to 

evaluate talus thickness 

Colluvium: can form thin 

weak layer that influences 

slope stability; water 

accumulates at colluvium/rock 

interface 

Talus: in-situ material may 

exist near angle of repose (i.e., 

FOS=1) 

Shale and other 

Degradable 

Materials 

poorly indurated shale, 

claystone and mudstone 

that degrades to the parent 

soil material upon contact 

with water and air 

in-situ materials may be rock-like 

and difficult to sample, requiring 

rock drilling equipment; water 

introduced during drilling may 

initiate degradation 

material used as rockfill may 

unexpectedly degrade upon 

contact with water; slope 

stability will progressively 

decrease 

Cemented Sands 

sandy soil with salt or 

calcareous bonding at 

points of grain-to-grain 

contact; cementing agent 

may be soluble or 

insoluble 

high blow counts make material 

appear as dense sand; penetration 

tests meet refusal; rock coring 

equipment may be necessary, but 

water may cause dissolution of 

binder 

soil is brittle and appears to be 

very strong; potential for 

collapse or failure if binder is 

soluble; long-term strength 

may degrade to that of 

uncemented sand 

Sensitive Clays 

marine deposit generally 

consisting of silty, lowplasticity clays that have 

been leached of salt 

resulting in metastable 

structure 

metastable structure makes 

undisturbed sampling difficult; 

may require special foil sampler 

in-situ soils can collapse with 

little or no warning due to 

inherent metastable structure 

Partially 

Saturated Soils 

almost all in-situ soils that 

exist above the 

groundwater table 

partially saturated soils are 

generally stronger than saturated 

soils and present no special 

sampling difficulties as a result of 

partial saturation; some soils may 

be brittle at natural moisture 

contact; in-situ testing may be 

helpful in establishing existing 

properties but uncertain drainage 

conditions during in-situ tests 

make property evaluation difficult 

these soils exist naturally but 

in-situ strengths may be 

significantly different than 

laboratory strengths because 

lab tests rely on saturated 

specimens; in-situ tests results 

may not be applicable for 

design conditions 

Archived Document
231 

7.2 LOESS 

7.2.1 Identification of Loess 

Collapsible soils generally classify as soils that undergo a relatively significant and sudden decrease 

in volume when water is introduced. Types of collapsible soil include: (1) loose man-made fills; (2) 

colluvium (discussed subsequently in section 7.5); and (3) loess. Collapsible soils usually exist in 

the ground at relatively low values of dry unit weight and moisture content. At these conditions the 

materials are moderately strong and exhibit a slight but characteristic apparent cohesion. In their 

natural state, such soils can support moderate loads and undergo relatively small settlements. Upon 

wetting, however, the cohesion in the soil is lost and large settlements can occur even if the loading 

remains constant. If subsurface exploration results indicate the presence of collapsible soils, then 

laboratory testing of undisturbed samples (discussed subsequently) should be performed to quantify 

the magnitude of volume reduction upon wetting. In this section, the engineering properties of loess 

are specifically described, although other collapsible soils can be expected to behave similarly. 

Loess, which is common in the central U.S. as well as portions of Washington, Oregon, Idaho, and 

Alaska, is a uniform cohesive wind-blown (i.e., aeolian) soil consisting primarily of silt-sized 

particles. The apparent cohesion of the material is the result of calcareous clay binder that holds the 

silt particles together. The clay coating and wind blown formation create a very loose soil structure 

with little true particle-to-particle contact especially at low confining pressures. For loess (as well as 

for man-made fills), relatively low SPT N values can be used as an indicator of the potential for 

collapse. The engineering behavior of loess is affected by whether the loess is sandy, silty, or 

clayey. Silty loess is characterized as being extremely erodible; sandy and clayey loess are 

collapsible but are much less erodible than silty loess. Hydrometer testing should be performed on 

the material passing a No. 200 sieve to evaluate the quantity of sand, silt and clay. 

7.2.2 Issues Related to Subsurface Exploration in Loess 

Typical drilling and disturbed sampling procedures can be used to obtain loess samples for 

laboratory sieve analysis, hydrometer, soil classification, and Atterberg limits testing. In regions of 

the U.S. where loess is present, engineers have developed design charts for cut slopes in loess. 

These charts are used for relatively low-height cut slopes; however, for designs involving high cut 

slopes, undisturbed samples should be collected for laboratory strength and consolidation testing. 

Because of the collapsible nature of loess, it is absolutely critical to use sampling and sample 

handling procedures that minimize disturbance. Such procedures have been discussed in chapter 4. 

Oftentimes, critical potential failure surfaces for slope stability analyses are relatively shallow for 

soils exhibiting relatively small cohesion intercepts such as loess. It is therefore important to 

accurately characterize the shear strength of the loess at relatively shallow depths. For samples to be 

collected at shallow depths, it may be prudent to obtain block samples from trenches or test pits. 

The shear strength of loess is greatly affected by the degree of saturation of the soil. Therefore, it is 

essential to develop an accurate estimate of the position of the depth to groundwater and to assess 

whether the degree of saturation for the deposit will likely change during the design life (i.e., will 

there be a change in the position of the groundwater table). If the loess may eventually become 

saturated, it is explicitly noted that in-situ tests results alone should not be used to develop design 

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parameters. Typical in-situ testing procedures including the SPT, CPT, and DMT can be used to 

assess the current shear strength properties of the soil deposit only if the current state will be 

maintained during the project design life. Additionally, in-situ testing procedures should only be 

used to assess shear strength properties if the results are correlated to laboratory strength tests 

performed on samples at similar moisture contents and saturation conditions. 

7.2.3 Laboratory Strength Testing of Loess 

A common issue related to loess (and all soils that are partially saturated), is that the shear strength 

at low confining pressures, derived primarily by a cohesion intercept, is greatly affected by the 

moisture content (and saturation condition) of the sample. Saturation of the loess material “softens” 

the calcareous clay binder and greatly decreases strength. This, in turn, may lead to slope instability 

and accelerated erosion. Prior to performing laboratory shear strength tests, a critical assessment of 

the expected saturation conditions of the loess soil for the projected design life must be considered. 

The saturation condition will be affected by the location of the groundwater table and surface 

drainage measures. It is conservative to perform CIU triaxial tests on saturated loess samples and 

use the calculated undrained strength parameters for stability analyses. 

7.2.4 Evaluation of Collapse Potential 

In addition to potential strength loss due to saturation, loess (and other collapsible soils) can undergo 

volume change due to collapse of the soil structure when the sample becomes inundated. For 

situations in which it is necessary to construct a facility in and around collapsible soils such as loess, 

it is of primary importance to estimate the magnitude of potential collapse that may occur if the soil 

becomes wetted. To do this, a one-dimensional collapse potential test can be performed in an 

oedometer on undisturbed or recompacted samples according to ASTM D 5333, “Standard Test 

Method for Measurement of Collapse Potential of Soils”. For this test, a sample is placed in an 

oedometer and the vertical pressure on the sample is increased to the anticipated final loading in the 

field. At this load level, water is introduced to the sample and the resulting deformation due to 

collapse is recorded. The percent collapse (%C) is defined as: 

o

c

H

H C ∆ =100 % (Equation 92) 

where ∆Hc is change in height upon wetting and Ho is the initial height of the specimen. For a soil 

layer with a given thickness, H, the settlement due to collapse, scollapse, may be calculated as: 

( )

( ) H100

C%

scollapse = (Equation 93) 

The collapse potential (CP) is calculated as the percent collapse (%C) of a soil specimen subjected 

to a total load of 200 kPa as measured using ASTM D 5333. The CP is an index value used to 

compare the susceptibility of collapse for various soils. Table 48 provides a relative indication of 

the degree of severity for various values of CP. 

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Table 48. Qualitative assessment of collapse potential (after ASTM D 5333). 

Collapse Potential (CP) Severity of Problem 

0 None 

0.1 to 2% Slight 

2.1 to 6% Moderate 

6.1 to 10% Moderately Severe 

>10% Severe 

7.3 EXPANSIVE SOILS 

7.3.1 Identification of Expansive Soils 

Expansive soils are typically clayey soils that undergo large volume changes in direct response to 

moisture changes in the soil. Unlike collapsible soils, expansive soils tend to increase in volume 

(i.e., swell) as the moisture content of the soil is increased and decrease in volume (i.e., shrink) as 

the moisture content of the soil is decreased. Although the expansion potential of a soil can be 

related to many factors (e.g., soil structure and fabric, environmental conditions, etc.), it is primarily 

controlled by the clay mineralogy. Soils that contain low-plasticity kaolinite will tend to exhibit a 

lower shrink/swell potential than soils containing high-plasticity montmorillonite. Expansive soils 

are found throughout the U.S., however, damage caused by expansive clays is most prevalent in 

California, Wyoming, Colorado, and Texas where the climate is considered to be semi-arid and 

periods of intense rainfall are followed by long periods of drought. This pattern of wet and dry 

cycles results in periods of extensive near-surface drying and desiccation crack formation. During 

intense precipitation, water is added to the deep cracks permitting the soil to swell; upon drying, the 

soil will shrink. This weather pattern results in cycles of swelling and shrinking that can be 

detrimental to the performance of pavements, slabs on-grade, and retaining walls.

Deep-seated volume changes in expansive soils are rare. More common are volume changes within 

the upper few meters of a soil deposit. These upper few meters are more likely to be affected by 

seasonal moisture content changes due to climatic changes. The zone over which volume changes 

are most likely to occur is defined as the active zone. The active zone can be evaluated by plotting 

the moisture content with depth for samples taken during the wet season and for samples taken 

during the dry season. The depth at which the moisture content becomes nearly constant is the limit 

of the active zone depth (which is also referred to as the depth of seasonal moisture change). The 

active zone is an important consideration in foundation design. In the design of piles or drilled 

shafts, it is important to recognize that full side friction resistance may not be realized in this zone. 

As the soil undergoes cycles of shrinking and swelling, it may lose contact with the pile or shaft. 

Alternatively, as the soil swells, it may impose significant uplift pressures on the foundation 

element. 

In the field, the presence of surface desiccation cracks and/or fissures in a clay deposit is an 

indication of expansion potential. Experience has indicated that the most problematic expansive 

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near-surface soils are typically highly plastic, stiff, fissured overconsolidated clays. To identify 

expansive soils in the laboratory, several classification methods have been developed. Currently, 

there is not a standard classification procedure; different methods are used in various locations 

across the U.S. Typically, methods include the use of Atterberg Limits and/or clay soil percentage 

to qualitatively describe a soil as having low, medium, high, or very high expansion potential. 

Generally, soils with a plasticity index less than 15 percent will not exhibit expansive behavior. For 

soils with a plasticity index greater than 15 percent, the clay content of the soil should be evaluated 

in addition to the Atterberg Limits. Figure 106 shows the swelling potential of a remolded soil as 

related to the soil activity and clay fraction. For the purposes of evaluating expansion potential of a 

soil, activity can be defined as: 

( ) Clay Fraction ( ) CF

Plasticity )PI(Index AActivity = (Equation 94) 

where CF is the clay fraction that corresponds to the percentage of particles exhibiting an equivalent 

diameter (ds)<0.002 mm as calculated from a hydrometer test performed in accordance with ASTM 

D422, Standard Test Method for Particle Size Analysis of Soils. 

Figure 106. Classification chart for swelling potential (after Seed et al., 1962)

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Figure 107 relates expansion potential to liquid limit and in-situ dry density based on the experience 

of the U.S. Bureau of Reclamation. Additional tests for the qualitative assessment of expansion 

potential include percent swell calculated from the California Bearing Ratio (CBR) test (ASTM 

D4429), the free swell test, and the expansion index test. These tests and others are discussed in 

Chen (1988) and Nelson and Miller (1992). Such correlations are semi-empirical and should only be 

used for an initial assessment of the expansion potential of a soil. If any of the above tests indicate a 

potentially expansive soil, laboratory testing, as described in the next section, should be conducted 

on undisturbed samples to assess the potential swelling pressures of the material. 

Figure 107. Guide to collapsibility, compressibility, and expansion based on in-situ dry density and 

liquid limit (after Mitchell and Gardner, 1975 and Gibbs, 1969) 

7.3.2 Evaluation of Expansion (Swell) Potential 

For situations in which it is necessary to construct a facility in and around expansive soils, it will be 

necessary to estimate the magnitude of swell (i.e., surface heave) and the corresponding swelling 

pressures that may occur if the soil becomes wetted. A one-dimensional swell potential test can be 

performed in an oedometer on undisturbed or recompacted samples according to AASHTO T256 or 

ASTM D4546, “Standard Test Method for One-Dimensional Swell or Settlement Potential of 

Cohesive Soils”. In this test, the swell potential is evaluated by observing and measuring the swell 

of a laterally confined specimen when it is surcharged and flooded with water. Alternatively, after 

the specimen is inundated, the height of the specimen is kept constant by adding load. The swelling 

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pressure is then defined as the vertical stress necessary to maintain zero volume change. Swelling 

pressures in some expansive soils may be relatively large such that the loads imposed from 

lightweight structures or pavements do little to counteract swelling (heave). 

The use of the one-dimensional swell potential test to evaluate in-situ swell potential of natural and 

compacted clay soils has limitations including: 

• Lateral swell and lateral confining pressure are not simulated in the laboratory. The 

calculated magnitude of swell in the vertical direction may not be a reliable estimate of soil 

expansion for structures which are not confined laterally (e.g., bridge abutments); 

• The rate of swell calculated in the laboratory will not likely be indicative of rates of swell 

experienced in the field. Laboratory tests cannot simulate the actual availability of water in 

the field. 

• Long-term swell may be significant for some soils and should be added to primary swell, 

especially if the design is anticipated to have long-term access to water. 

7.3.3 Shear Strength Evaluation of Expansive Soils 

As noted previously, potentially problematic expansive soils are located near the ground surface and 

are heavily overconsolidated. These soils usually contain montmorillonite or other highly plastic 

mineral and have natural moisture contents equal or less than the plastic limit of the soil. Practically 

speaking, the shear strength of these materials is relatively high and therefore presents no major 

concerns relative to, for example, shallow foundation bearing capacity or embankment slope 

stability. However, if active controls are not incorporated into the design to minimize the potential 

for large moisture changes and subsequent expansion, then laboratory strength testing needs to be 

performed for the most critical saturation condition the soil is expected to incur in the field. If the 

near-surface soils will undergo relatively large seasonal changes in moisture content, then shear 

strength testing should be performed on specimens at the anticipated highest moisture content. Less 

conservative shear strength assessments (i.e., shear strengths corresponding to moisture contents 

other than the highest anticipated) can be used for cases where active measures are incorporated into 

the design to assure that a reasonably constant moisture content is maintained in the field. 

7.4 ORGANIC SOILS AND PEAT 

7.4.1 Introduction 

Organic soils (i.e., organic clays and organic silts) present similar engineering challenges as soft silts 

and clays, including low undrained shear strengths and high compressibility. In addition, organic 

silts and clays undergo significant secondary (or creep) deformations. Such long-term, continuous 

deformation can present significant maintenance issues for embankments and other structures that 

may be founded over such materials. Like other organic soils, peats also undergo significant 

secondary deformations. In this section, the identification and classification of organic soils and 

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peats are discussed along with information on evaluating shear strength and compression properties 

for these materials. 

7.4.2 Identification of Organic Soils and Peat 

Organic soils and peats are evidenced during subsurface exploration based on the presence of 

decaying vegetative matter and a strong odor. Typically, the materials are greenish, dark gray, or 

black in color and can have very fibrous structures with wood fragments and plant remains. 

Disturbed sampling techniques (e.g., auger cuttings that are brought to the surface) are used to 

provide a visual confirmation of the organic material. In-situ testing can also be used to identify 

subsurface layers of organic material. For example, these materials can be identified using the CPT 

by low cone tip resistance and relatively high friction ratios (see figure 38). Classification charts for 

the CPTu (figure 39) and the DMT (figure 43) have also been developed to identify organic 

materials and peat. However, it is recommended that disturbed samples be obtained to correlate 

CPT, CPTu, and DMT readings as it may be difficult to rely solely on in-situ testing. Oftentimes, 

clayey soils and organic soils do not have significantly different sounding information (i.e., stiffer 

organic soils will be classified as clays using the DMT classification charts). 

Samples of soil layers noted as organic on a boring log should be tested in the laboratory to evaluate 

the percentage of organic matter. The non-organic portion of the sample will control the engineering 

behavior of soils when the organic content is less than approximately 20 percent (Arman, 1970). 

Organic content is evaluated in the laboratory using one of several methods available in AASHTO 

T194 and ASTM D 2974, “Standard Test Method for Moisture, Ash, and Organic Matter of Peat 

Material and Other Organic Soils.” In this test, a sample of the material is dried, weighed, and then 

heated at high temperature. After cooling, the sample is reweighed and the loss of mass is 

calculated. After heating, ash is left over; therefore, organic content is equal to 100-Ac where Ac is 

the percentage of ash. The liquid limit test can also be used as a qualitative index of organic matter 

by comparing the liquid limit value for a sample that was dried before testing to the liquid limit 

value for a sample that was not dried. If the liquid limit of the sample that was dried is less than 75 

percent of the value of the liquid limit for the sample that was not dried then the soil may be 

classified as organic. 

Landva et al. (1983) developed a system for classifying organic soils and peats. In this system, peats 

and organic soils are divided into four groups: (1) peats (Pt); (2) peaty organic soils (PtO); (3) 

organic soils (O); and (4) silts and clays with organic content (MO and CO, respectively). Table 49 

summarizes the typical ranges of the properties used in this classification system. 

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Table 49. Organic soils and peat classification properties (after Landva et al., 1983). 

Material Ac

Moisture 

Content (wn) 

Specific Gravity 

(Gs) 

Fiber 

Content 

Pt <20% >500% < 1.7 > 50% 

PtO 20 - 

40% 150 - 800% 1.6 – 1.9 <50% 

O 40 - 

95% 100 – 500% > 1.7 Insignificant 

MO, CO 95 - 

99% <100% > 2.4 None 

7.4.3 Issues Related to Subsurface Exploration and Sampling of Organic Soils and Peat 

Organic soils and non-fibrous peats can be very weak and compressible due to a high moisture 

content and large void ratio. Like most soft clays, it can be difficult to obtain undisturbed samples 

of organic soils for laboratory performance testing. Thin walled fixed piston samplers are most 

suitable for undisturbed sampling of organic soils. Undisturbed sampling in fibrous peats, however, 

is very difficult due to the likely compression of the peat fibers during sampler advancement. Where 

undisturbed sampling with piston samplers is attempted, it is critical that the sampler have a sharp 

edge. If possible, fibrous peats should be sampled using block-sampling techniques. 

Engineering properties of peats and organic soils can vary significantly both spatially and with 

depth. Samples obtained within a few feet of each other may exhibit vastly different behaviors 

during loading. Therefore, subsurface investigations that encounter organic soils and peats should 

involve more sampling and testing as compared to inorganic soils to adequately characterize the 

materials. 

7.4.4 Shear Strength of Organic Soils and Peats 

Like all soils, the shear strength of organic soils is directly related to the effective stress in the 

ground and stress history of the deposit. Since organic soils are relatively lightweight (i.e., low dry 

density), saturated, and have no significant stress history, their strengths are usually very low. 

Where good quality undisturbed samples can be obtained, laboratory triaxial strength testing should 

be performed to obtain undrained shear strength information for design. Like most clayey soils, 

organic soils typically have very low hydraulic conductivities, therefore the assumption of undrained 

failure imposed by the CPT, CPTu, and VST allow for the use of these in-situ testing devices as a 

means to correlate undrained shear strength. The relationships given in table 33 for these in-situ 

tests can be used for testing in organic soils. The various factors (i.e., Nk, μ) should be developed 

based on site-specific data and laboratory triaxial data for a few samples. Correlations based on 

cone tip resistance may be difficult to use since it is likely that the tip resistance in organic soils will 

be extremely low, thus requiring high resolution load cells. 

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In using the vane shear test in organic soils, it is noted that the time to failure in the test will have an 

impact on the measured vane shear strength. Also, organic soils may possess enough fibers to act as 

localized reinforcement and lead to vane shear strengths that are too high. For these reasons, use of 

the equation su=μsu(VST) may overestimate the undrained strength of the organic clay or silt. Because 

of fiber content, the vane shear test is inappropriate for relatively high fiber organic soils and for 

fibrous peats. Moreover, due to the fibrous nature of such soils, hydraulic conductivities are 

relatively large (compared to silts and clays), and the appropriate shear strength for design may be 

more appropriately defined using drained strength parameters, as discussed below. 

In fibrous peats, the fibers tend to act as tensile reinforcement in the soil. Therefore, the strength 

behavior is almost entirely frictional with large friction angles (i.e., greater than 35o

). However, the 

shearing resistance varies significantly depending on the orientation of the failure plane relative to 

the general alignment of the fibers. Generally, the fibers are oriented horizontally. Therefore, 

failure planes in triaxial test samples may intersect the fibers resulting in high shear strengths. It is 

also noted that peat samples may require significant deformations (i.e., greater than 20 percent axial 

strain in drained laboratory triaxial tests) to mobilize the full strength of the material. This needs to 

be considered in stability analysis so that the mobilized strength for the analysis is consistent with 

the expected level of deformation. 

7.4.5 Compressibility of Organic Soils and Peats 

If a shallow foundation is to be constructed in and around an area with a near-surface deposit of 

organic soil or peat, the layer will likely be excavated and replaced prior to foundation construction. 

For embankments and foundations that are constructed over organic soil or peat layers, primary 

settlement will occur over a relatively short time (i.e., within a few days or months), and the 

majority of the total settlements will result from the long-term secondary compression of these soils. 

Therefore, secondary settlement will be the dominant component of settlement during the design life 

of the structure and should be evaluated as presented in section 5.4.2.7. 

Because of high in-situ void ratios, organic soils and peats will have high Cc values (i.e., >1). Figure 

108 can be used to estimate Cc values for clays, silts, and peats based on in-situ moisture content. 

As discussed in chapter 5, Cα/Cc for organic silts and clays is approximately 0.05 + 0.01 and for 

peats, the value is approximately 0.06 + 0.01, the highest value for geotechnical materials (Mesri et 

al., 1997). It is noted that standard one-dimensional oedometer tests and interpretation methods (as 

described in chapter 5) are used to evaluate Cc and Cα in organic soils and peats. In peaty soils, it 

may be difficult to identify the time to the end of primary consolidation, tp, since the transition from 

primary to secondary compression is not easily identified using standard interpretation methods. 

However, reasonably accurate estimates of Cα can still be made. The maximum vertical effective 

stress in the ground needs to be assessed so that a reasonably conservative estimate of secondary 

compression can be assessed. Like other soils, organic soils and peats will show an increase in Cα

for stress levels just in excess of the preconsolidation stress, σp′. It is therefore important to perform 

laboratory oedometer tests on organic soils and peats to stress levels greater than σp′. Calculated 

values of Cα will likely decrease for stress levels significantly greater than σp′, so tests should be 

performed to sufficiently high loads to ensure that the maximum value for Cα is calculated. 

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Figure 108. Values of natural water content and compression index for peats, clays, and silts (after 

Mesri, Stark, Ajlouni, and Chen, 1997). 

7.5 COLLUVIUM AND TALUS 

7.5.1 Identification of Colluvium and Talus 

Colluvium (colluvial soil) and talus are generally relatively loose deposits found near the base of 

slopes and may accumulate in valleys, swales, or other low-lying topographic features. Colluvial 

soils generally result from a two-stage process of: (1) in place weathering of the parent rock; and (2) 

subsequent migration downslope primarily by gravity through creep. Colluvium commonly consists 

of rock fragments in a heterogeneous clayey to sandy matrix. Talus, like colluvium, also 

accumulates at the base of slopes, but generally consists of the mechanically weathered granular 

component of the parent rock. Talus slopes are commonly characterized as well-graded boulders to 

sand- or silt-sized particles. In general, colluvium occurs in temperate and humid environments 

while talus predominantly occurs in arid and semi-arid regions. The characteristics of the parent 

bedrock and the climate in which the weathering and migration/transportation take place determine 

the characteristics of the colluvium and talus deposits. 

Accumulations of colluvium and talus on the sideslopes and at the base of sideslopes are often 

associated with slope stability problems. To route transportation passageways, cut slopes often have 

to be made in colluvium and talus deposits located near the base of a slope. In many cases, the cut 

slope often exposes the colluvium/talus material. Because these materials typically form by 

migration and sliding along the slope, they are often only marginally stable in their natural state. 

Therefore, the cut slopes made in these deposits tend to disrupt the natural equilibrium, thus 

requiring aggressive monitoring and maintenance. 

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7.5.2 Issues Related to Subsurface Exploration and Testing in Colluvium 

Typical drilling and sampling techniques can be utilized in colluvium to obtain samples for 

evaluating the physical characteristics of the material, provided it consists primarily of fine-grained 

particles. In some cases, colluvium exists as a massive deposit upward of 5 meters in thickness, 

while in other cases, the colluvium may occur as only a relatively thin (i.e., <1-m thick) weathered 

zone. In cases where the colluvial material occurs as only relatively thin lenses within a soil stratum 

and is highly weathered, it is easy to drill through the material and not realize its presence. Drilling 

and sampling in colluvial soils with large rock fragments can prove to be difficult. In addition to the 

above concerns, potential access problems due to relatively steep slopes and marginal stability and 

safety concerns typically make drilling and sampling in colluvium relatively difficult and expensive. 

Test pits and trenches represent economical alternative exploration methods for colluvium. This 

approach allows visual observation of the subsurface conditions in these materials. Test pits and 

trenches are typically backhoe-excavated; however, shallow hand-excavated test pits can provide 

subsurface data in inaccessible areas. Undisturbed block samples of colluvium can be cut from the 

sidewalls of the test pits and trenches. In cases where rock fragments do not preclude sampling, the 

colluvium may be sampled by manually advancing (i.e., pushing or driving) Shelby tubes or other 

types of sampling tubes from the sidewalls of the test pit. Test pits and trenches can also be used to 

evaluate the depth to bedrock and the groundwater conditions within the colluvium. 

Inclinometers and piezometers can be used as another means to characterize the engineering 

performance of in-situ colluvium deposits. Inclinometers socketed into underlying bedrock can be 

used to assess whether the colluvium is actively creeping down the slope. The measured 

displacements provide an indication of the relative stability of the colluvial soil slope. Piezometers 

installed at various depths within the colluvium, especially at the colluvium/bedrock interface, 

provide vital information regarding the groundwater conditions within the colluvium. It is common 

that the colluvium/rock interface provides a shallow aquitard that allows water to accumulate in the 

colluvium at the interface. This water can foster the further degradation of the parent rock 

(particularly if the underlying rock is poorly indurated shale and/or mudstone) and may cause 

seepage to occur along the interface. Both of these factors tend to reduce stability of the colluvium 

veneer along the slope. 

7.5.3 Issues Related to Subsurface Exploration and Testing in Talus 

Because the materials in talus slopes includes particles that range from silt and sand to boulders, site 

exploration is extremely difficult. Although individual samples of talus components may be 

recovered, they are often of little use because these small samples are indicative of only the matrix 

component of the talus, and are not at all representative of the mass itself. Thus, characterizing the 

global behavior or global properties is extremely difficult when using conventional drilling and 

sampling equipment. Similarly, any in-situ test requiring penetration is not useful because the 

penetration would likely meet refusal on a large rock/boulder. 

The occurrence of talus can probably best be identified by the use of aerial photographs and site 

reconnaissance. Site subsurface characterization is probably best performed by geophysical methods 

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that are non-invasive (e.g., SASW). In most cases, foundation support will need to be achieved in 

competent rock below the talus material. For this reason, geophysical techniques that can identify 

the depth to competent rock are particularly useful, especially where characterization over great 

lengths is required. 

7.5.4 Compressibility of Colluvium and Talus 

The compressibility of the colluvial soils can generally be assessed in the laboratory using the 

standard oedeometer test, provided that undisturbed samples can be obtained in the field. 

Additionally, settlements for these materials can be calculated based on the results of geophysical 

tests and the use of elastic theory as discussed in section 5.5. The shear wave velocity can be 

obtained from geophysical testing and provides a reliable means of evaluating the small strain shear 

modulus of a soil. 

Talus often exists in-situ as a loose mixture of granular material and boulders that may experience 

settlements during construction. Compression properties are difficult to assess due to the “scale” of 

the materials (i.e., large boulders to fine-grained soils) and the heterogeneity of the material (i.e., 

large rocks and large void spaces). The void spaces create the potential for large concrete over 

pours and/or lost grout in the construction of drilled shafts and anchors. Therefore, the assumed side 

friction capacity of the design may not be realized. Because of this and the compression potential of 

the material, foundation systems should not be designed to derive resistance from talus materials and 

should, instead, be founded in competent material below the talus. 

7.5.5 Shear Strength of Colluvium and Talus 

Shear strength of colluvium can be assessed from laboratory tests on undisturbed samples. In some 

cases, however, the colluvium is relatively strong and it is instead the colluvial material and 

potentially weathered rock at the colluvial soil/rock interface that represents the weakest material 

and where the greatest potential for sliding instability exists. As mentioned previously, water may 

accumulate at the colluvium/rock interface, weakening the soil and rock materials at this interface. 

The shear strength of the material at the colluvium/rock interface can be evaluated by performing 

laboratory direct shear tests on remolded samples of the material at the expected in-situ moisture 

content. In the test, residual conditions should be evaluated, especially if there is a potential that the 

material had been previously displaced as a result of landsliding or creep. Figure 76 can be used to 

make preliminary estimates of the drained residual friction angle. 

A reliable method to evaluate the strength of both colluvium and talus is to back analyze a failed or 

failing slope in close proximity to the project location. For talus materials, because of sampling 

difficulties, back analysis is often the only method for evaluating shear strength. Back analyses 

involve performing stability analyses where the values of the cohesion intercept and friction angle 

are modified to achieve a factor of safety of 1.0. This method for determining shear strengths is only 

accurate if a thorough assessment of the slope geometry and groundwater table are made. 

Additionally, the depth to competent rock and the location of a known or anticipated slip surface 

should be assessed via borings or by geophysical methods. 

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7.6 SHALES AND DEGRADABLE MATERIALS 

7.6.1 Identification of Degradable Materials 

On many transportation projects, construction activities involve the use of potentially degradable 

materials. Additionally, in many parts of the U.S., foundations are commonly built on these 

materials. Although the material may, at first, exhibit rock-like characteristics, it has the potential to 

degrade to soil-size particles. The gradual but ultimate degradation of the rock to the original parent 

soil material can occur within minutes or after several years of exposure to air and/or water. Shale, 

the most common member of this family of materials, can generically be considered to include 

claystone, siltstone, and mudstone. 

In many parts of the U.S., high-quality granular material is not locally available for use as borrow 

material in the construction of earth embankments and/or rockfill. As a result, degradable materials 

that, at first, appear to be competent granular materials are used. However, once in contact with 

water, these materials may degrade causing problems and/or failures during the service life of the 

structure. Foundations built on these materials are also at risk for failure. Drilling and other 

construction activities often introduce water, initiating the degradation process. Deep foundations, 

especially rock-socketed drilled shafts, which may be designed considering intimate contact with 

and support from the rock interface may fail as the materials degrade and contact is lost. 

Additionally, highway cuts are often excavated in degradable rock. The exposure of the cut to 

atmospheric conditions may result in significant degradation during the service life of the cut slope. 

Many rock types are prone to degradation when exposed to the cyclic wet/dry and freeze/thaw 

weathering processes. Rock types that are particularly susceptible to degradation due to these 

processes are poorly indurated shale and claystone exhibiting high clay content. The degradation 

can take the form of swelling, weakening, and ultimately disintegration. The effect of degradation 

on slope stability can range from surficial sloughing and gradual retreat of the face, to catastrophic 

slope failures resulting from the significant loss of strength. In sedimentary rock formations 

comprising alternating beds of resistant sandstone and relatively degradable shale, the weathering 

process can develop overhangs in the sandstone and produce a rockfall hazard. In shallow 

foundations, the assumed bearing capacity of the material may decrease as the foundation material 

degrades resulting in settlements and/or foundation failures. In deep foundations, as discussed 

above, the assumed end bearing and/or side friction may decrease over time. 

When potentially degradable materials are encountered, it is essential to establish the anticipated 

competency of the materials over the service life of the project. An assessment of the time required 

for significant degradation relative to the service life of the structure should be evaluated. 

Commonly, the point load test, the slake durability test, and the jar slake test are used to make this 

assessment. Classification systems and index values have also been developed for identifying the 

behavior of shales (see chapter 21 of TRB, 1996). 

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7.6.2 Slake Durability Test 

A simple index test to assess the tendency of rock to weather and degrade is the slake durability test, 

ASTM D4644, “Standard Test Method for Slake Durability of Shales and Similar Weak Rocks”. 

The test procedure includes placing the sample in a 2 mm square wire mesh drum and drying it in an 

oven to constant weight at 110oC. The test sample should consist of ten representative, intact, 

approximately equidimensional fragments weighing 40 to 60 grams each. The wire mesh drum is 

then partially submerged in water and rotated at 20 revolutions per minute for a period of 10 

minutes. The drum and its contents are then dried a second time and the loss of weight is recorded. 

The test cycle is then repeated a second time and the slake durability index, ID, is calculated as the 

ratio (reported as a percent) of final to initial dry weights of the sample. Earlier versions of this test 

(and some classification index systems) included up to five wet/dry cycles. The slake durability 

index is usually combined with other indices to formally assess rock durability. In general, a high 

slake durability index indicates that the rock is not particularly susceptible to degradation when 

exposed. For highly degradable rocks (i.e., a low Id), it is useful to carry out soil classification tests 

such as Atterberg limits and X-ray diffraction tests to identify clay mineral types and evaluate if 

swelling clays (e.g., montmorillonites) are present and responsible for the degradation. 

7.6.3 Jar Slake Test 

Like the slake durability test, the jar slake test requires that pieces of the rock be immersed in water 

and that evidence of degradation be noted. This test is somewhat easier and less quantitative than 

the slake durability test. The procedure for the jar slake test for shale materials is summarized as 

follows (TRB, 1996): 

• A piece of oven-dried shale is immersed in enough water to cover it by 15 mm. 

• After immersion, the piece is observed continuously for the first 10 minutes, followed by an 

additional 20 minutes of discontinuous, but careful, monitoring. If a reaction between the 

water and the rock is to occur, it will usually occur during this time period. A final 

observation of the condition of the rock is made after 24 hours. 

• Based on the visual observations, the Jar Slake Index, IJ, is established using the criteria 

described in table 50. 

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Table 50. Evaluation of jar slake index, IJ. 

Jar Slake Index, IJ General behavior during test 

1 Degrades rapidly to a pile of flakes or mud 

2 Breaks rapidly and/or forms many chips 

3 Breaks slowly and/or forms few chips 

4 Breaks rapidly and/or develops several fractures 

5 Breaks slowly and/or develops few fractures 

6 No change 

Like the slake durability index, the jar slake index is commonly used in combination with other 

indices to systematically classify the durability of the rock. This is described in section 7.6.5. 

7.6.4 Point-Load Test 

The point load test was initially developed to estimate the compressive strength of rock in which 

both rock core samples and fractured rock can be tested. However, because the testing equipment is 

portable and the test itself is inexpensive, the point load test is also used to assess the degradation 

potential of materials such as shales. The test is conducted by compressing a piece of rock to failure 

between two points on cone-shaped platens. Each of the cone points has a 5-mm radius of curvature 

and the cone bodies themselves include a 60° apex angle. The point load strength, Is is calculated as: 

2 D

P Is = (Equation 95) 

where P is the point load breaking strength and D is the distance between the platens (i.e., the 

approximate diameter of the rock). Details of this test are provided in FHWA HI-99-007 (1998) and 

in Section 4.13.2. Using the point load strength, the material can be classified as durable and strong 

for IS > 6 MPa, conditionally durable for 2 MPa < IS < 6 MPa, or non-durable and weak for IS < 2 

MPa (after Welsh et al., 1991). Because the point load test was not specifically developed for 

estimating the degradation potential of shales, it should only be used to make preliminary 

assessments of degradation potential. If the test indicates a potential for degradation, further testing 

should be conducted in the laboratory using the methods discussed previously. 

7.6.5 Use of Shale Material 

The laboratory and field index tests described above can be used to assess whether a shale (or other 

potentially degradable material) may be considered to be rockfill, soil, or as intermediate material 

between soil and rockfill. Intermediate materials are considered to be non-durable materials and 

need to be conditioned to be soil-like prior to use as a construction material. Strohm et al. (1978) 

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developed the criteria presented in table 51 based on values of ID and IJ. Other shale rating systems 

and rock durability classification criteria are available to assist the designer. 

Table 51. Criteria for rockfill materials (after Strohm et al., 1978). 

Slake 

Durability 

Index, ID

Jar 

Slake 

Index, IJ

Category 

>90 =6 

Durable rockfill materials, if minus 

gravel-sized fraction is less than 20 to 

30 percent 

60 to 90 3 to 5 Hard, Non-durable intermediate 

material 

<60 <2 Soft, non-durable materials treated as 

soil 

7.7 CEMENTED SANDS 

7.7.1 Identification of Cemented Sands 

Cemented sands are naturally occurring granular materials that have a cementing material either in 

the void space between individual grains or at the points of grain-to-grain contact. The result of the 

cementing action is that the sand exhibits a true cohesion (i.e., a component of shear strength that is 

independent of confining pressure). As a result, cemented sands are usually “stronger” than 

uncemented sands. This increased strength, however, presents several problematic characteristics: 

(1) the cementing agent may be so weak that it is destroyed during sampling, thus resulting in a poor 

characterization of the in-situ deposit; (2) the cementing agent may be soluble or extremely weak, 

such that the cohesion cannot be relied upon for long-term design conditions; and (3) at failure, the 

cemented sand exhibits a brittle load-deformation response that explicitly must be recognized. 

The primary consequence of cementation is to increase strength and reduce compressibility relative 

to uncemented materials. Thus, from this perspective, it is recognized that cementation improves the 

engineering characteristics of the deposit. Weakly cemented sands can form due to grain-to-grain 

point-contact welding as a result of aging or from a clay-silt binder that accompanies a wind-blown 

dune deposit. At the other extreme, the cemented sand can be characterized as weak sandstone, 

where a carbonate bond may occur at the grain contacts. Regardless of the method of origin, 

cementation results in stiff, but brittle, load-deformation behavior. Such behavior must be carefully 

assessed when evaluating shear strength for design, as discussed subsequently. 

7.7.2 Issues Related to Subsurface Exploration and Testing in Cemented Sands 

Cemented sands often present sampling difficulties. If the materials are weakly cemented, the 

cemented structure might not be recognized when conventional soil sampling techniques are 

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attempted. In these cases, the SPT blow counts may appear to be uncharacteristically high, but the 

recovered materials may appear to be uncemented. This is because the relatively weak cementing 

materials can be easily disrupted during drilling and cemented sand, while relatively weak, may be 

sufficient enough to preclude sampling using Shelby tubes. In both of these cases, the 

driving/pushing resistance is mistakenly interpreted as being characteristic of dense uncemented 

sands. At low working stress levels, this mischaracterization may be conservative; but at high 

working stresses, a dense sand may, in fact be stronger than a loose cemented sand, resulting in 

unconservative estimates of strength. At the other extreme, if the subsurface materials are 

characterized as soft rocks, conventional rock sampling/coring techniques may result in very poor (if 

any) sample recovery, due to the disturbance induced by the drilling equipment and the effects of 

water on the cemented structure. 

It is often possible to correctly identify and characterize cemented sands through careful observation 

during field reconnaissance, drilling, and sampling. The easiest technique to correctly identify 

cemented materials is to visually observe an exposed excavation or natural cut (e.g., stream bank 

erosion) in the material. For shallow deposits, test pits permit observation of the material and allow 

access for recovery of a block sample. For deep deposits, the following two techniques can be 

employed by the field engineer or geologist: (1) carefully observe the recovered cuttings and attempt 

to recover small pieces of the cemented materials that are returned within the matrix of uncemented 

sands; and (2) be observant of the potential reasons for “anomalous” behavior during drilling and 

sampling (e.g., uncharacteristically high blow counts during soil sampling, poor core recovery 

despite relatively uniform moderate drilling resistance, etc.). In some cases, double coring 

techniques employing Pitcher barrel samplers (soils) and Dennison samplers (rock) can be used to 

provide representative samples of deep deposits for visual characterization of the cemented nature of 

the material. 

With regards to laboratory testing, it is necessary to be extremely careful when extruding and 

handling the cemented materials. The relatively weak and brittle cementing agent can yield to even 

gentle fingertip pressure during handling and trimming. If possible, the recovered cemented 

materials should not be trimmed, but rather tested at the as-recovered diameter. For recovered block 

samples, the material should be carefully “shaved” in the laboratory to obtain a specimen suitable for 

testing. 

7.7.3 Interpretation of Laboratory and Field Testing Results in Cemented Sands 

As mentioned previously, an indicator of cemented sands is uncharacteristic or unanticipated 

behavior. Specifically, unusually high blow counts and/or poor recovery are characteristics of 

cemented sands. There are, however, several other reasons for high blow counts and poor recovery, 

so these by themselves cannot be taken as reliable indicators of the presence of cemented sands. If 

sufficient materials are recovered to allow laboratory testing, index tests and performance tests can 

reliably identify the existence and the character of the cemented material. In the case of index tests, 

a simple unit weight determination on an undisturbed specimen can quickly and reliably identify 

whether the material is as dense as the high blowcount response would indicate. It may be possible 

to either immerse a sample in water or simply add water to a piece of the intact sample to assess 

whether the cementing agent is soluble or if the material softens when inundated with water. If 

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either of these responses is identified, a careful assessment must be made of whether the service 

conditions will result in the introduction of (and the effect of) water. 

If specific strength/deformation characteristics are needed, laboratory triaxial shear or direct shear 

testing is recommended. Because of the sensitive and brittle nature of the cementing materials these 

tests must be carefully conducted and interpreted. It is important that representative samples be 

selected for laboratory testing, because one of the effects of sample disturbance is that only the 

strongest materials may survive the drilling/sampling process. Block sampling has been shown to be 

an effective technique for obtaining samples of cemented sands suitable for laboratory testing. Insitu testing using a pressuremeter has also been used effectively to provide quantitative strength and 

stiffness information. Other in-situ testing techniques, specifically the SPT and dilatometer, provide 

useful qualitative results, but must be calibrated to specific site/material conditions to provide 

quantitative information. 

The load versus deformation response of cemented sands must also be reviewed from the 

perspective of the brittle character of the material. At low confining pressures, the response due to 

cementation will dominate the frictional response. This results in an initial stiff response due to the 

cementation, followed by a strain softening response associated with the frictional characteristics of 

the sand after rupture of the cementing bonds. As the confining pressure increases for a specific 

degree of cementation, the difference between the peak and post-peak strengths decrease. At the 

limit at high confining pressures, it is possible that application of the confining pressure may result 

is disruption of cementing bonds, resulting in a load-deformation response that is consistent with 

that of an uncemented sand. The lesson from this general response characteristic is that the range of 

test confining pressures must be carefully selected to match the anticipated service conditions. 

Additionally, because the cementing bonds can be disrupted at low strains, the anticipated strains 

under the anticipated working stress should be assessed to allow the engineer to decide whether the 

peak (i.e., cemented) or the large-displacement (i.e., uncemented) strengths should be used in design. 

As an example, large-displacement pile driven into calcium carbonate sands will likely exhibit 

frictional capacities significantly less than that which would be predicted based on peak strengths 

derived from undisturbed samples. Driving piles can destroy bonds and disrupt the delicate structure 

of the cemented material (see Focht, 1994). 

7.8 SENSITIVE CLAYS 

7.8.1 Identification of Sensitive Clays 

Most geotechnical engineers are aware of the strain softening response (i.e., a peak strength 

followed by a reduction in strength with ongoing deformation) of soils under drained loading. In 

general, over-consolidated clays, dense sands, and most soils tested under low confining pressures 

tend to exhibit a strain softening response under drained loading conditions. This behavior is due to 

the dilation (i.e., positive volume change) characteristic of these materials. For these soils under 

undrained loading conditions the strain softening behavior generally is not observed, due primarily 

to the generation of negative pore water pressure that prohibits the increase in volume. In two 

notable soils, however, there is a post-peak reduction in the undrained strength: (1) extremely loose 

sands; and (2) sensitive clays. Extremely loose sands that tend to strain soften can be subject to 

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catastrophic static liquefaction. Fortunately, the natural occurrence of these materials is very rare, as 

they are usually associated with hydraulically filled structures. Sensitive clays, while naturally 

occurring, are generally confined to specific geographic regions of the world where there was 

extensive glaciation followed by isostatic uplift. The strain softening response in undrained shear 

loading for both loose hydraulically placed sands and sensitive clays are attributed to their unique 

metastable structure. Because strain-softening loose sands rarely are encountered, and are almost 

never encountered as part of highway construction, additional discussion is not provided for these 

materials. Sensitive clays, while confined to specific geographically regions, warrant a brief 

discussion in this document. 

The metastable structure of sensitive clays is established when relatively low plasticity clays are 

deposited in brackish (i.e., salty) waters in a flocculated particle orientation. The resulting high void 

ratio soil structure, due to the edge-to-face alignment of the clay plates, is stable although the soils in 

the deposit have high natural moisture content and a moderately high liquidity index. In this state, 

the clay is weak, but not likely sensitive. The sensitive characteristics are introduced when fresh 

water is leached through the uplifted deposit, replacing the brackish water with the fresh water. As a 

result of the replacement with fresh water, the electro-chemical interaction between clay plates is 

altered to the point where the Atterberg limits reduce and the liquidity index increases. In this state, 

the clay “favors” a dispersed soil structure; however, the existing flocculated structure is maintained 

until some mechanical action causes a disruption of the structure (i.e., the peak strength is 

mobilized). Because the disturbed clay now prefers a dispersed soil structure, there is a significant 

reduction in the post-peak shear strength due to the large amount of unbound pore water associated 

with the dispersed soil structure. 

7.8.2 Issues Related to Subsurface Exploration and Testing in Sensitive Clays 

Sampling and laboratory testing of sensitive clays is extremely difficult, primarily because the 

disturbance typical of conventional sampling and testing techniques may be sufficient to disrupt the 

in-situ soil structure. Essentially, this disturbance has the effect of mobilizing the peak strength of 

the material. In the extreme case of the glacial quick clays, the in-situ material may currently exist 

at a liquidity index that is greater than unity (i.e., the in-situ moisture content is greater than the 

liquid limit). Therefore, if the sample is disturbed to a degree sufficient to disrupt the in-situ soil 

structure, the once-solid soil will behave as a viscous liquid. 

When sampling and testing sensitive soils, extreme care must be exercised to minimize the potential 

for disrupting the in-situ soil structure. Conventional thin-walled tube samplers that are deployed 

rapidly may be appropriate for moderately sensitive materials. In some cases, using samplers larger 

than the conventional 3-inch diameter Shelby tubes will provide better recovery of less disturbed 

material. Regardless of the sample size, rapid sampler deployment has been shown to be effective at 

minimizing disturbance induced by friction along the walls of the sampler. In very sensitive soils, 

specially developed samplers that rely on foil liners to encapsulate the samples are used. This 

equipment is specialized, but may be common in areas of the country where sensitive clays are 

encountered (e.g., the northeastern seaboard, St. Lawrence Seaway valley, etc.). For shallow 

deposits, excavation of block samples is an excellent alternative. The key to laboratory testing 

follows the guidelines from field sampling, that is, handle the material with extreme care. The 

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samples should not be trimmed to a smaller diameter. It is best to extrude the sample directly onto 

the testing pedestal and provide confinement using a thin membrane. 

Field testing in sensitive soils using a field vane is an excellent alternative (or compliment) to 

laboratory testing. Local experience has been developed in parts of the country where sensitive soils 

are common. Specifically, the size of the vane and the methods for deployment should be matched 

to the strength of the material. One of the advantages of this type of field testing is that the 

equipment can be extremely portable and easy to hand-deploy over sensitive (or very soft) materials. 

Similarly, there are encouraging results from other types of in-situ tests, specifically the piezocone 

penetrometer (CPTu) and flat-plate dilatometer (DMT), conducted in soft and/or sensitive soils. The 

key to testing using these devices is to match the resolution of the testing equipment (i.e., capacity of 

load cells, membrane stiffness calibration, etc.) to the anticipated strength of the tested materials. 

7.8.3 Interpretation of Laboratory and Field Testing Results in Sensitive Soils 

Regarding the interpretation of the laboratory and field tests conducted on or in soft and/or sensitive 

soils, it is important to recognize that the stress-strain behavior of the soil is a function of the soil 

structure, and that the structure can potentially change during the test. In the cases of both 

consolidation and shear testing, the change in structure can result in the apparent “collapse” of the 

initial structure, meaning that there are two distinct and different regions of behavior associated with 

the pre-collapse and the post-collapse structure. When interpreting the test results, it is important to 

recognize whether a structure change occurred during the test. For example, in moderately sensitive 

materials, the specimen may have been sufficiently disturbed to mobilize the peak strength prior to 

conducting the laboratory test. In these cases, the laboratory test results will need to be associated 

with a post-peak response. It is explicitly noted that because of the structure-specific test results, 

most results of laboratory tests on sensitive soils cannot be normalized. Any test that requires 

confinement or testing above the in-situ state of stress is prone to initiate irreversible structural 

changes in the soil. For these reasons, it is recommended that field vane shear tests always be 

conducted in soft and/or sensitive soils. This test has been widely used throughout the world and 

can be modified for almost any soft soil condition. In addition, the test can be readily used to assess 

sensitivity because the peak- and post-peak strengths are explicitly provided. 

7.9 PARTIALLLY SATURATED SOILS 

7.9.1 Identification of Partially Saturated Soils 

There has been a recent interest in understanding the behavior of partially saturated (aka 

“unsaturated”) soils. This interest generally focuses on the relative roles of pore water and pore air 

pressures in determining the behavior of soil. The profession has learned from this focused interest 

that the roles of the pore fluids can change and are dependent on the relative moisture content of the 

tested material. The implications from these studies is that it may be difficult, if not impossible, to 

assess the stress-strain behavior of partially saturated soils using only the results of tests conducted 

on saturated specimens. This is important because: (1) almost all laboratory tests for strength and 

compressibility are conducted on saturated soil specimens; and (2) the majority of soils that occur in 

nature are not completely saturated. It is recognized that sands, silts and clays obtained from below 

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(and in many cases above) the water table are nearly saturated. In addition, for most of these soils, 

the effect of partial saturation is to stiffen and strengthen the tested material. The degree of 

stiffness/strength change is inversely related to the degree of saturation, with the low saturation 

specimens showing the largest differences. Therefore, relying on the results of tests conducted on 

saturated soil specimens to assess the properties of their in-situ unsaturated condition generally 

provides a conservative assessment of the materials strength and compressibility relative to the 

properties of the in-situ material. 

Relative to this section of the document, it can be argued that unsaturated soils are not really 

“unique” or “special”. What is important is that when attempting to assess performance in 

comparison to design, it is important to recognize that the in-situ strength and compressibility of 

unsaturated soils will likely differ from the properties that were used in the design. Because the 

saturated soil response will likely represent a worst-case representation of field conditions, it is 

entirely reasonable that the design parameters be developed based on saturated soil conditions. 

7.9.2 Sampling and Testing Partially Saturated Soils 

Since most in-situ soils exist in an unsaturated condition, conventional sampling and specimen 

preparation procedures specifically recognize the characteristics of unsaturated soils. In general, the 

air component of the pore volume interacts with the pore water to form menisci between the solid 

soil particles. The surface tension resulting from this interaction essentially increases the relative 

confinement on the sample. This, in turn, explains why soil specimens that are usually thought to be 

“cohesionless” can be collected in the field and will remain intact. Sampling partially saturated soils 

can be extremely difficult when the natural materials exist at a low degree of saturation (i.e., 

typically less than 50 percent). In this state, the surface tension is very large but the available water 

is so widely distributed that the natural material can be extremely brittle. In a sandy material the 

material is subject to “collapse” from even the slightest disturbance during sampling or material 

handling. In cohesive soils, the material may become too hard to effectively sample using a thinwalled sampler. This is often the case of colluvial or residual soil materials. In the cases where 

sampling is difficult and the deposit is shallow, block sampling and test pit exploration can prove to 

be very helpful. 

7.9.3 Interpretation of Laboratory and Field Testing Results of Partially Saturated Soils 

In general, soils should be tested in their unsaturated state. Most laboratory tests are conducted on 

saturated specimens to accommodate the measurement of either the excess pore pressure (undrained) 

or volume change (drained) characteristics of the soil. For tests that are conducted on partially 

saturated specimens, it is explicitly noted that the measurement of excess pore pressure and volume 

change is difficult because there will be a differential pressure in the two pore fluids, depending on 

the rate of testing, the type of soil, and the degree of saturation. As a result, total stress tests are 

generally conducted wherein the pore pressure/volume change characteristics are explicitly not 

measured. In interpreting the results from these tests, it should be noted that the effect of the 

air/water menisci is to provide a negative pore pressure and thus a potentially significant effective 

confining pressure. The result of this will be that the material is stronger and stiffer than would be 

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anticipated for a saturated material, particularly in a condition where no (or very little) confining 

pressure is externally applied. From the interpretation of strength properties perspective, this means 

that the material will exhibit an apparent cohesion intercept and an unconfined compressive strength. 

In some cases, it may be appropriate to use these apparent strengths in design, particularly for shortterm loading conditions or in cases where the material will likely never become saturated. 

In the case of the interpretation of field tests, it must be explicitly recognized that the natural 

materials are unsaturated and that the measured response will represent the actual strength/stiffness 

of the material in the in-situ condition. Because many interpretation procedures explicitly assume 

saturated conditions, the interpreted in-situ properties may be significantly stronger/stiffer than 

would be anticipated based on simple index tests and visual classification of the materials. In these 

cases, the test result is reporting the actual response and behavior of the material. The potential 

shortcoming is the inability of the conventional interpretation techniques to distinguish the influence 

of partial saturation. 

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CHAPTER 8 

APPLYING JUDGEMENT IN SELECTING 

SOIL AND ROCK PROPERTIES FOR DESIGN 

8.1 INTRODUCTION 

The reader is referred back to the flow chart presented in chapter 2, with a specific charge to review 

the flow chart in the context of what information and guidance was provided in chapters 3 through 7 

of this document. In these sections, several "how to" recommendations were made relative to the 

measurement and assessment of soil and rock properties. These aspects included recommendations 

for planning and executing a subsurface investigation, selecting drilling and sampling techniques, 

selecting sampling intervals and number of samples, developing the field and laboratory testing 

programs, and conducting and interpreting the field and laboratory tests. Recommendations have 

been made regarding which laboratory and in-situ tests are appropriate for obtaining specific soil and 

rock properties that may be needed in design. Finally, in chapters 5 and 6, recommendations were 

made regarding the interpretation of the laboratory and field tests. Notice that at this point, the 

reader should be near the end of the flow chart, and specifically at the step identified as Select 

Material Properties and Finalize Subsurface Model. At this point, two potentially significant 

questions are raised. Specifically, what should the engineer do if during the material property 

selection process it is found that: (1) there are inconsistencies between the results of selected tests; or 

(2) there is a significant variability in a selected parameter within the assumed relatively “uniform” 

subsurface model? To resolve these two issues, the engineer should carefully scrutinize the 

available data, assess the application where these data are needed, and ultimately apply judgment to 

select the appropriate value(s) considered for the design. From the early days as students in the 

introductory geotechnical engineering classroom, the term “judgment” has been used to describe 

how the geotechnical engineer makes decisions. While it is difficult to describe how to apply 

judgment or how to define what is “good” or “bad” judgment, this section will address the steps that 

should be followed to in making a judgment in response to the two previously posed questions. 

8.2 RESOLVING INCONSISTENCIES BETWEEN TEST RESULTS 

One of the goals of this document has been to make recommendations regarding the type of tests to 

conduct for a given application or for a specific soil/rock property. In many cases, multiple 

tests/resources were identified to assist the engineer in obtaining the needed information. Inevitably, 

because of the inherent variability of the soil deposit itself, the variations due to laboratory and field 

testing conditions, and the variation inherent in engineering property correlations, the 

calculated/estimated results will not be entirely consistent. Typical common examples of scenarios 

where inconsistent data are generated include: (1) SPT blow counts that are conducted across the 

entire site to correlate specific properties for the type of encountered soil and a limited number of 

index and/or performance tests conducted on a few “representative” samples; and (2) field vane 

shear test (VST) to provide a nearly continuous undrained strength profile versus depth at a project 

site and a few “high quality” unconsolidated undrained (UU) triaxial shear tests, again on recovered 

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representative samples. The following provides a step-by-step summary of how the engineer may 

resolve these differences. 

1. Data Validation: Assess the field and the laboratory test results to determine whether the 

reported test results are accurate and are recorded correctly for the appropriate material. 

Disregard (or at least downplay) potentially questionable results. 

2. Historical Comparison: Assess results with respect to anticipated results based on site 

and/or regional history. If the new results are inconsistent with a wealth of site/regional 

data, it will be necessary to assess whether the new data are anomalous or whether the new 

site conditions are anomalous. 

3. Performance Comparison: Assess results with respect to historic performance of structures 

at the site or within the region, if possible. For example, if settlements at adjacent structures 

or wall movements have been recorded historically, it may be possible to back-calculate 

actual in-situ soil properties. These can then be compared to the measured/correlated 

properties that are in question. 

4. Correlation Calibration: Develop site-specific correlations using the new field and 

laboratory data. Assess whether this correlation is within the range of variability typically 

associated with the correlation based on previous historic data used to develop the generic 

correlation. 

5. Assess Influence of Test Complexity: Assess results from the perspective of the tests 

themselves. Some tests may be easy to conduct and calibrate, but provide data of a 

“general” character (e.g., Atterberg limits) while other tests are complex and subject to 

operator influences, yet provide “specific” test results (e.g., stress path triaxial tests). It may 

be found that certain tests consistently provide high (or low) values compared to the 

anticipated results. Alternatively, certain field tests may be assumed to be undrained, when 

in fact the test duration allows drainage. 

The ultimate goal of this activity is to confirm that the data are valid for the test considered. In many 

cases, after considering all of these steps, it may be found that the actual anticipated range of 

properties is much less than originally reported. However, it is also common to find that the results 

of this step will indicate that all of the data are apparently valid and, thus, the material at the site is 

inherently variable; this by itself is a significant finding. The net outcome of this step is to provide a 

summary of valid data that is representative of the soils at the site. The fact that these data may be 

variable is discussed subsequently. 

8.3 ESTIMATING VARIABILITY OF SELECTED PARAMETERS 

In the previous section, it was noted that variability (or uncertainty) may be characteristic of the 

material or of the soil deposit, and the variability includes the combination of both inherent material 

variability and testing variability. Just as judgment was used to assess data quality and accuracy, it 

is necessary to apply similar judgment to estimate the variability for each of the selected parameters 

that will be considered in the design. There are several techniques to estimate this variability. 

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1. Experience: In some cases the engineer may have accumulated extensive experience in the 

region that it is possible to accurately select an average, typical or design value for the 

selected parameter, as well as the appropriate variability for the parameter. 

2. Statistics: If the engineer has extensive experience in a region, there may be sufficient data 

to formally establish the average value and the variability (i.e., mean and standard deviation) 

for the specific design parameter. 

3. Published Values for Variability: It has been found that engineers can estimate average 

values for specific parameters with a relatively high degree of accuracy, if sufficient data are 

presented. Unfortunately, it has also been found that engineers typically believe that there is 

less variability in these soil parameters than the data actually support. In recognition of 

these phenomena, Duncan (2000) prepared a table that presents an accumulation of 

experience regarding the coefficient of variation for several soil parameters. The values for 

the coefficient of variation (defined as the standard deviation divided by the average value) 

suggested by Duncan are presented in table 52. 

4. Establish Best-Case and Worst-Case Scenarios: Once again, relying on the experience of 

the engineer, it may be possible to establish not only the average value for a selected 

parameter, but also an “absolute” upper-bound and a lower-bound (i.e., best-case and worstcase) estimate for the parameter. Once these extreme values are defined, it is possible to 

utilize the “three-sigma rule” as described in Duncan (2000) to define the highest 

conceivable value (HCV) and the lowest conceivable value (LCV) for the parameter. It can 

be shown that 99.73 percent of the data for the specific parameter value will be within the 

range of the HCV and the LCV. Therefore, these values can be used to assess the standard 

deviation and thus the coefficient of variation. Soil Property Selection Example No. 2 

illustrates the use of the “three-sigma rule” and the LCV and HCV. 

The outcome of this step is to provide a best estimate of the specific parameter and a quantifiable 

measure of anticipated variability of the parameter. Regardless of the procedures used to formally 

estimate the variability of the selected parameter, it is important to review information from the 

previous section to confirm that the coefficient of variability resulting from this step encompasses 

the actual variability for the parameter as measured (or estimated) from the data. In the unlikely 

event that the variability resulting from section 8.2 is higher than the values resulting from 

calculations in section 8.3, it is likely that distinct material differences at the site exist, rather than 

simply inherent material variations. 

8.4 FINAL SELECTION OF DESIGN PARAMETERS 

A combination of judgment, experience, and actual data were used to assess the validity of the data 

and the anticipated variability of the data that are needed for design. This final step again involves 

judgment and the use of information from the previous two sections to select the appropriate design 

parameters. Using information provided from section 8.2 (and hopefully section 8.3), there are three 

approaches to this final selection. 

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Table 52. Values of coefficient of variation, V, for geotechncial properties and in situ tests (after 

Duncan, 2000) (see Duncan, 2000 for original references on reported values of V). 

Measured or interpreted parameter value Coefficient of Variation, V 

(%) 

Unit weight, γ 3 to 7 % 

Buoyant unit weight, γb 0 to 10 % 

Effective stress friction angle, φ′ 2 to 13 % 

Undrained shear strength, su 13 to 40 % 

Undrained strength ratio (su/σv′) 5 to 15 % 

Compression index, Cc 10 to 37 % 

Preconsolidation stress, σp′ 10 to 35 % 

Hydraulic conductivity of saturated clay, k 68 to 90 % 

Hydraulic conductivity of partly-saturated clay, k 130 to 240 % 

Coefficient of consolidation, cv 33 to 68 % 

Standard penetration blowcount, N 15 to 45 % 

Electric cone penetration test, qc 5 to 15 % 

Mechanical cone penetration test, qc 15 to 37 % 

Vane shear test undrained strength, suVST 10 to 20 % 

1. Semi-Deterministic: The engineer may simply estimate the design value by some form of 

judgment based on experience. For example, the undrained strength profile could be 

established by using the average of the field VST and the lab UU test results. The engineer 

may not use the UU test results for samples at significant depths. Similarly, the compression 

index for a clayey sand could be established based on historical regional correlations to SPT 

results. This selected value could be used in the requisite calculations to assess the design 

factor of safety (or the anticipated settlements). If acceptable performance is calculated 

(e.g., global calculated factor of safety (FOS) greater than 1.5 or total calculated settlements 

of less than 1 inch), the engineer may conclude that the design is complete. This approach 

has been adopted by many but is not recommended because it does not account for the 

heretofore described inherent variability of the selected parameters. Additionally, and 

probably most importantly, this procedure potentially allows the design engineer to use the 

rule-of-thumb or the experience of others without first-hand knowledge of the limitations 

and assumptions inherent to the assumptions. 

2. Statistics and Sensitivity Analyses: The engineer can use experience to assess the average 

value for the selected parameter, but explicitly also considers a reduced estimate of this 

value for design. For example, using the coefficient of variability resulting from section 8.3, 

the engineer may also select a reduced estimate (e.g., the mean value minus 1 standard 

deviation ...aka “mean minus 1 sigma") for consideration in design. By conducting 

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analyses at these two potential values, an assessment is made of the sensitivity of the 

analysis results to a range of potential design values. If these analyses indicate that 

acceptable results are provided and that the analyses are not particularly sensitive to the 

selected parameters, the engineer may be comfortable with concluding the analyses. If, on 

the other hand, the engineer determines that the calculation results are marginal or that the 

results are sensitive to the selected parameter, additional data review and parameter selection 

are warranted. If the engineer is experienced, this technique can be quite effective, 

particularly when it can be shown that additional analyses are not warranted. One potential 

disadvantage of this method is that “experience” can be used to selectively use parameters 

that provide the desired calculation result. 

3. Probabilistic Analyses: The previously referenced paper by Duncan (2000) presents an 

approach that explicitly combines variability in the design calculations and concepts of 

probability to estimate the potential for failure (or poor performance). Although not 

described in detail in this document, the technique involves the following three steps: (1) 

perform calculations (e.g., slope stability or settlement calculations) explicitly using the 

HCV and LCV values previously identified for each significant parameter or property; (2) 

use the results of these calculations to establish the variability of the calculation results to 

the selected parameter values; and (3) use the calculated variability to assess the probability 

of failure. The authors believe that, in time, this procedure (or something similar) will be 

incorporated into conventional practice, particularly in cases where highly variable 

properties are encountered and extensive investigation and testing are the only other 

techniques available to assess the values of the parameters. 

Regardless of the technique used to select the specific design value (or range of design values), the 

engineer will be provided with a measure of the calculated performance. The advantages of the first 

technique are that it is relatively simple and likely follows traditional practice. The latter two 

approaches incorporate a formal assessment of the variability in the calculation results to variations 

in material parameter selection. The authors believe that each of these approaches has merit and is 

appropriate under certain design conditions. Semi-deterministic techniques are recommended to be 

used only when the individual engineer has extensive experience in the project area and with the 

materials involved in the design. In this case, experience is extensively relied upon and should only 

be considered for non-critical applications. The use of statistics and sensitivity analyses should 

generally be considered the minimum practice. For this case, the engineer should be experienced 

with the materials and the application so that an interpretation of the sensitivity studies can be 

effectively understood. Ideally, the calculations can be considered complete when the calculation 

results are shown to be relatively insensitive to the anticipated variation in the selected parameter. 

Probabilistic techniques should be considered whenever the variability in the parameters is large or 

where the consequence of failure is high, and when the costs of acquiring additional confirming data 

are large. 

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In Situ Determination of Ko in Soft Sensitive Clays.” Proceedings, In-Situ Measurement of Soil 

Properties, Vol. I, ASCE, Raleigh, North Carolina, pp. 450-476. 

Tavenas, F., Leroueil, S., and Roy, M. (1982). “The Piezocone Test in Clays: Use and Limitations.” 

Proceedings, 2nd European Symposium on Penetration Testing, Vol. 2, Amsterdam, 889-894. A.A. 

Balkema, Rotterdam. 

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Teh, C.I. and Houlsby, G.T. (1991). “An Analytical Study of the Cone Penetration Test in Clay.” 

Geotechnique, Vol. 41, No.1, pp. 17-34. 

Terzaghi, K., and Peck, R.G. (1967). Soil Mechanics in Engineering Practice. John Wiley & Sons, 

Inc., New York, N.Y. 

Terzaghi, K., Peck, R.G., and Mesri, G. (1996). Soil Mechanics in Engineering Practice, John Wiley 

& Sons, Inc., New York, N.Y., pp. 549. 

Transportation Research Board (TRB) (1996). Landslides: Investigation and Mitigation, Special 

Report 247. National Academy Press, Washington D.C., 674 pp. 

Vesic, A.S. (1977). “Design of Pile Foundations.” Synthesis of Highway Practice 42, 

Transportation Research Board, Washington, D.C., 68 pp. 

Vinson, J., and Brown, D.A. (1997). “Site Characterization of the Spring Villa Geotechnical Test 

Site and a Comparison of Strength and Stiffness Parameters for a Piedmont Residual Soil.” Report 

No. IR-97-04, Highway Research Center, Harbert Engineering Center, Auburn University, AL, 385 

pp. 

Welsh, R.A. Jr., Vallejo, L.E., Lovell, L.W., and Robinson, M.K. (1991). “The U.S. Office of 

Surface Mining (OSM) Proposed Strength-Durability Classification System.” Proceedings, 

Symposium on Detection of and Construction at the Soil/Rock Interface, W.F. Kane and B. Amadei, 

Eds., ASCE Geotechnical Special Publication No. 28, American Society of Civil Engineers, New 

York, N.Y. 

Windle, D., and Wroth, C.P. (1977). “In Situ Measurement of the Properties of Stiff Clays.” 

Proceedings, 9th International Conference on Soil Mechanics and Foundation Engineering, Tokyo, 

Japan, pp. 347-352. 

Wood, D.M. (1990). Soil Behavior and Critical State Soil Mechanics. Cambridge University Press, 

U.K., 462 pp. 

Woods, R.G. (1994). “Laboratory Measurement of Dynamic Soil Properties.” Dynamic 

Geotechnical Testing II, ASTM, Special Technical Publication No. 1213, Philadelphia, PA, 165-

190. 

Wroth, C.P. (1984). “The Interpretation of In-Situ Soil Tests.” Rankine Lecture, Geotechnique, 

Vol. 34, No. 4, pp. 449-489. 

Wyllie, D.C. (1999). Foundations on Rock. Routledge, New York, N.Y., 401 pp. 

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APPENDIX A 

SOIL AND ROCK PROPERTY SELECTION EXAMPLES 

OVERVIEW 

Soil and rock property selection examples are provided in this section. The purpose of these 

examples is to illustrate the application of the methods and procedures described in chapters 1 

through 8 of this GEC. For each example, a different soil profile is used. Table A-1 provides a 

summary of the examples included in this section along with a list of the particular properties and 

parameters that are evaluated. It is noted that the methods used to evaluate certain properties and/or 

parameters do not necessarily vary significantly depending on the specific soil profile. For example, 

the methods and techniques used to evaluate subsurface stratigraphy from SPT and disturbed sample 

information and in-situ testing data does not vary significantly in soft cohesive soils and heavily 

overconsolidated soils. For this reason, the first two examples have been developed to be 

comprehensive with respect to evaluating all potentially relevant properties and parameters. The 

remaining example focuses on the evaluation of properties and parameters that are considered to be 

unique to the particular soil profile. 

Table A-1. Summary of soil property selection examples. 

Example No. Subsurface Profile Properties and Parameters Evaluated 

1 Soft to Medium Clay and 

Overconsolidated Clay Crust 

• Development of a Subsurface Profile (soil 

type, wn, LL, PL, PI, LI, γtot, γdry) 

• Stress History (σp′) 

• Deformation Parameters (Cc, Ccε, Cr, Cα) 

• Time Rate of Consolidation (cv, ch) 

• In-Situ Horizontal Stresses (Ko) 

• Undrained Shear Strength (su,UU, su,CIUC, 

su,VST, su,CPT, su,CPTu, su,DMT) 

2 Piedmont Residual Soil, 

Weathered Rock, and Rock 

• Development of a Subsurface Profile (soil 

type, GSD, wn, LL, PL, PI, LI, γdry) 

• Deformation Parameters (EPMT, EDMT) 

• Drained Strength (c′, φ′) 

3 Heavily overconsolidated 

clays 

• Development of a Subsurface Profile (soil 

type, wn, LL, PL, PI, LI, γtot, γdry) 

• Stress History (σp′) 

• Deformation Parameters (Cc, Ccε) 

• In-Situ Horizontal Stresses (Ko) 

• Undrained Strength Properties 

• Swell Potential 

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SOIL AND ROCK PROPERTY SELECTION EXAMPLE NO. 1 

Soft to medium clay and overconsolidated clay crust

INTRODUCTION 

The purpose of this example is to describe procedures and interpretation methods available in 

current practice for evaluating soil properties and parameters for soft to medium clays that may 

include overconsolidated clay crusts. Additionally, guidelines for sampling, testing, and property 

selection will be discussed. This example covers properties necessary for design on soft clay 

including: (1) subsurface stratigraphy (i.e., soil layering and groundwater levels); (2) in-situ stress 

state; (3) shear strength; and (4) consolidation. Various laboratory and in-situ testing devices may 

be used to evaluate clay soil properties. Several methods are presented for each property to 

demonstrate the utility and limitations of each and the rationale is provided to evaluate a particular 

property value for design when multiple values from several tests are recorded. This information 

may also be used by design engineers to plan an appropriately scoped subsurface investigation 

program. 

In developing this example, it is assumed that the engineer has reviewed engineering reports, boring 

logs, and other sources of information on the local geology for the area to be investigated. The data 

used for this example are from projects, located in the vicinity of the Connecticut River Valley of 

western Massachusetts. Subsurface soils in this area include an overconsolidated clay crust 

overlying a thick layer of soft to medium clay with silt and sand varves. The general subsurface 

stratigraphy consists of 1 m of fill soil overlying a thick deposit of Connecticut Valley varved clay. 

Below the varved clay are dense silt deposits overlying bedrock. The thickness of individual silt or 

clay varves is typically approximately 2 to 8 mm and the varves are generally horizontally oriented. 

The upper 5 m of the varved clay deposit is overconsolidated because of surface erosion, 

desiccation, and seasonal groundwater level fluctuations. Below this weathered crust the soils 

become soft and near normally consolidated with increasing depth. For this example, the ground 

water table is located approximately 1 m below the ground surface. 

DATA REQUIREMENTS 

A detailed review of existing geotechnical data for the site will likely reveal information on 

subsurface stratigraphy and engineering properties of the soil similar to that information provided in 

the Introduction. For this site, it is likely that structural foundations required to support relatively 

large loads would consist of deep foundation elements (e.g., driven piles, drilled shafts) founded on 

or within the dense soil/rock layers below the varved clays. Relatively small loads may be able to be 

supported by a structural element founded within the upper overconsolidated clay crust. Table 1 

provides information on analysis requirements, soil properties, and laboratory and field-testing 

required to perform a design for specific geotechnical applications that may be constructed on the 

soil profile used for this example as well as other subsurface profiles. This table provides a link 

between design applications and engineering properties required for design and constructability 

assessments. 

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SUBSURFACE EXPLORATION AND TESTING PROGRAM 

A subsurface exploration and testing program for a soft to medium clay site including the number, 

location, and depth of borings, test pits, and hand augers, and number, location, and depth of 

disturbed and undisturbed samples will depend on factors which range from the specific design 

application, local geology, and the performance requirements for the design application. 

For this site, basic soil data (e.g., visual identification, classification, and index tests) obtained using 

disturbed split spoon samples and SPT blowcount values may be sufficient to develop a subsurface 

profile to be used for design analyses. Using these methods would likely require continuous 

sampling in the upper reaches of the subsurface to discern the transition from the crust to the softer 

clay soils. This transition could also be accurately evaluated using cone penetration testing and /or 

dilatometer testing. However, since split barrel samples obtained during a SPT cannot provide 

specimens suitable for performance-evaluation testing, and the SPT blowcount values cannot be 

used to assess shear strength or compressibility properties of clayey soils for final design, other insitu and laboratory tests are required. Therefore, in-situ testing methods combined with laboratory 

tests on specimens from high quality undisturbed samples are necessary to develop design 

parameters 

For most applications involving relatively soft cohesive soils, it is necessary to develop shear 

strength and compressibility parameters to be used for design. These parameters should be 

evaluated using laboratory testing methods and high quality undisturbed cohesive soil samples. 

Procedures for obtaining undisturbed soil samples have been described in chapter 4 of this 

document. It may be necessary or preferable to use in-situ testing devices to obtain data to 

complement the laboratory testing, especially if the project site is relatively large, if the project 

schedule is critical, if the soil deposit is highly stratified or has a highly developed macrofabric (e.g., 

an overconsolidated clay crust), or if it is difficult to obtain undisturbed samples. Local correlations 

between fundamental properties and in-situ test parameters are desirable, but global correlations 

confirmed with site specific laboratory test data may be equally appropriate. 

DEVELOPMENT OF SUBSURFACE PROFILE FOR DESIGN 

General

In this example, data collected from the SPT with recovery of disturbed samples, electric cone 

penetration test (CPT), piezocone penetration test (CPTu), and flat plate dilatometer test (DMT) are 

used to develop a subsurface profile for the site. When using CPT, CPTu, and DMT data, derived 

parameters are calculated and incorporated into empirical classification charts. The soil profiles 

developed from each test method are compared and a single subsurface profile is developed. 

Use of SPT and Disturbed Sampling

A boring log for the site is shown on figure A-1. A review of all available boring logs indicates that 

the subsurface profile consists of up to 5 m of relatively stiff varved clay and varved silt. Below this 

is a layer of soft varved clay that varies in thickness from 16 to 29 m. Below the varved clay is a 

layer of very stiff to hard varved silt. 

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Figure A-1. Boring log B1.

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 A-5 

The interpreted soil profile based on SPT data, visual identification and laboratory soil classification, 

and index testing of disturbed samples indicates two primary soil layers, Layer A and Layer B. 

Since this example focuses on overconsolidated crusts and soft to medium clays, the dense silt and 

bedrock below the soft varved clay are not considered further in this example. For clayey and silty 

soils, consistency can be evaluated using SPT blowcount (N) values (see table 23) and comparison 

of natural moisture content (wn) to Atterberg Limits using the Liquidity Index (LI). 

A summary of SPT N values, index test results, and calculated LI values are provided in figure A-2 

for three borings (B1, B2, and B3). Interpretation of soil layers A and B is described below. 

0

10

20

30

40

50

60

70

0 10 20 30 40 50 60

N - Value

Depth (m)

0

10

20

30

40

50

60

70

0 50 100

Moisture Content

0

10

20

30

40

50

60

70

-1 0 1 2 3

Liquidity Index

B1

B2

B3

0

10

20

30

40

50

60

70

7 10 13 16

γ dry (kN/m3

)

Figure A-2. Summary data for samples from Borings B1, B2, and B3. 

Layer A 

Layer A is located between the ground surface and a depth of 5 m. SPT N values for the three 

borings range from 6 to 21 blows/300 mm over this 5-m depth range indicating a medium stiff to 

stiff soil. Liquidity indices increase from approximately –0.5 for the upper 1 m to approximately 0.7 

at a depth of 4 m. These data indicate that the upper varved clay and varved silt crust is 

overconsolidated. Values of LI approach 1 at a depth of 5 m. Liquidity indices less than 

approximately 0.7 generally indicate that the soil is overconsolidated. 

Layer B 

Layer B is located below a depth of approximately 5 m. This layer is a very soft varved clay based 

on SPT N values (0 to 1 blows/300 mm) and on visual identification of disturbed samples. This 

layer extends to various depths in each boring. Atterberg limits data indicate that the material is a 

low plasticity silt. Grain size information for samples collected in Boring B1 indicate that the soil is 

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a silty clay to clayey silt. Note that the Atterberg limits testing was performed on a bulk specimen 

and that the Atterberg limits data for individual clay and silt varves would likely vary. The 

calculated LI is greater than one indicating that the soil is normally consolidated to lightly 

overconsolidated. 

CPT Stratigraphy

Measurements of CPT tip resistance (qt), sleeve friction (fs), and calculated friction ratio can be used 

to identify major stratigraphy changes in soil profiles. A CPT record for the example soil is shown 

on figure A-3. The profile shown for the two-layer system discussed previously (i.e., Layer A and 

Layer B) reveals additional subtle signatures of the layer, specifically the transition into and between 

these two major layers. Four layers (A1, A2, B1, and B2) are shown in figure A-3 and the 

interpretation of these layers is discussed subsequently. Individual measurements of qt and friction 

ratio are plotted on the classification chart (figure A-4). With reference to figure A-4, it is noted that 

tip resistance values less than 5 MPa are indicative of clayey and silty materials, while tip resistance 

values greater than 10 MPa are associated with sandy materials. Soil layers with tip resistance 

values between 5 MPa and 10 MPa range from loose sands or silt mixtures to overconsolidated 

clays. Tip resistance values for the example soil deposit are less than 5 MPa, indicating that the 

interpreted soil layering is a result of soil stiffness variation in clayey or silty material rather than a 

change in soil type. Table A-2 summarizes results of the CPT subsurface profile interpretation. 

0

5

10

15

0246

qt (MPa)

Depth (m)

0

5

10

15

012345

Friction Ratio (%)

0

5

10

15

0 50 100 150

fs (kPa)

Figure A-3. Subsurface soil layering based on CPT data and friction ratio. 

B1 

A1

A2 

B2 

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Figure A-4. Classification of soil type based on qt and friction ratio. 

Table A-2. Summary of CPT subsurface profile interpretation. 

Layer Depth Range 

(m) 

qt-FR zones Classification 

A1 0-0.7 3, 4 Silty clay 

A2 0.7-3.95 5, 6, 7 Clayey silt 

B1 3.95-5.25 4, 5 Silty clay 

B2 5.25 to top of 

dense silt layer 

1,3 ,4, 5 Clay 

CPTu Stratigraphy 

The piezocone provides the same information as the electric cone but also provides readings of 

penetration pore pressure with depth. The calculated pore pressure parameter (i.e., Bq=(u2-uo)/(qtσvo)) provides information that is used for stratigraphy profiling. Determination of stratigraphy in 

this section will concentrate on information obtained from the u2 penetration porewater pressure and 

Bq parameter. Five layers are shown in figure A-5, as interpreted using qt, friction ratio, u2, and the 

calculated Bq parameter. Individual measurements of Bq and qt are plotted on the Robertson et al. 

(1986) classification chart (figure A-6). Table A-3 summarizes the results of the CPTu subsurface 

profile interpretation. 

Layer A1

Layer A2

Layer B1

Layer B2

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 A-8 

0

5

10

15

0246

qt (MPa)

Depth (m)

0

5

10

15

-100 100 300 500 700

u2 (kPa)

uo

0

5

10

15

-0.2 0 0.2 0.4 0.6 0.8 1 1.2 1.4

Bq

0

5

10

15

0 50 100 150

fs (kPa)

Figure A-5. Subsurface soil layering based on CPTu data, friction ratio, and Bq parameter. 

Layer A1

Layer A21

Layer A22

Layer B1

Layer B2

Figure A-6. Classification of soil type based on qt and Bq. 

A1 

A21

A22

B1 

B2 

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Table A-3. Summary of CPTu subsurface profile interpretation. 

Layer Depth Range 

(m) 

Qt-FR zones Bq-FR zones Classification 

A1 0-0.7 3, 4 Silty clay 

A21 0.7-2.65 5, 6, 7 6, 7, 8 Clayey silt 

A22 2.65-3.95 5, 6 5, 6 Clayey silt 

B1 3.95-5.25 4, 5 3, 4 Silty clay 

B2 5.25 to top of 

dense silt layer 

1,3 ,4, 5 1,3 , 4 Clay 

For this example, the data obtained from penetration pore pressure using the piezocone were used to 

further delineate soil layer A2 between a depth of 0.7 and 3.95 m. This layer is further delineated 

and referred to as Layers A21 and A22. Additional discussion on the characterization of these layers 

is provided below. 

Layer A21 

Layer A21 exhibits tip resistance typical of a stiff clay to a very loose silty sand (3 to 4 MPa). Data 

points primarily plot in zones 5 through 7 of the qt-FR and qt-Bq classification charts. While this 

layer classifies as a clayey silt to a sandy silt, further review of the raw u2 pore pressure data 

indicates that the layer is an overconsolidated clay. The u2 pore pressure readings are about -50 kPa. 

Negative pore pressures may be measured in dense sands, however, the measured tip resistance in 

this layer is less than about 5 MPa, which, as previously noted, is indicative of silty and clayey 

materials. 

Layer A22 

During a cone test, there is a slight delay in penetration while a cone rod is added at 1-m intervals. 

For sandy materials, rapid dissipation of penetration pore pressures is noted during these brief 

delays. Although Layer A22 may exhibit behavior of a sandy silt (Zone 6), the rapid dissipation of 

pore pressure during rod breaks was not apparent, indicating that the layer may be more 

representative of a silt. 

Stratigraphy based on DMT Data

Like the CPT, empirical charts that incorporate index parameters based on testing device 

measurements are used for soil classification from DMT data. Figure A-7 shows raw DMT data and 

derived indices (i.e., ID, ED, KD) along with five interpreted soil layers for the example soil profile. 

Figure A-8 shows the material index values, ID, and dilatometer modulus values, ED, within each 

layer plotted on a soil classification chart developed by Marchetti & Crapps (1981). The layering 

that was delineated using the DMT matches closely that evaluated based on the CPTu data. The 

same layer designators are used for the DMT data as was used for the CPTu data. 

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 A-10 

0

2

4

6

8

10

12

14

16

18

20

0 500 1000 1500

Pressure (kPa)

Depth (m)

po

p1

p2

uo

0

2

4

6

8

10

12

14

16

18

20

0.1 1 10

ID

0

2

4

6

8

10

12

14

16

18

20

0 10 20 30 40

ED (MPa)

0

2

4

6

8

10

12

14

16

18

20

0 5 10 15 20

KD

Figure A-7. Subsurface soil layering based on DMT index values. 

10

100

1000

0.1110

Material Index, ID

E

D (atm)

Layer A1

Layer A21

Layer A22

Layer B1

Layer B2

 Clay Silt Sand

Figure A-8. Classification and consistency of soil layers based on DMT data. 

Archived Document
 A-11 

Layer A1 

Layer A1 extends to 1.2 m below the ground surface and is classified as compressible silt to sandy 

silt based on figure A-8. The boundary between Layer A1 and Layer A21 is noted by the increase in 

ED (see figure A-7). 

Layer A21 

Layer A21 is located at depths between depths of 1.2 m and 2.4 m, and is classified as a medium 

rigidity silty sand based on figure A-8. High values of overconsolidation ratio (OCR) are associated 

with high KD values. Since the crust is variable and has a high silt content with the potential for 

some sand (see figure A-2 for Boring B1), this layer is classified as an overconsolidated sandy silt. 

The boundary between Layer A21 and Layer A22 is evidenced by a decrease in ID, ED, and KD. 

Layer A22 

Layer A22 is located at depths between 2.4 m and 3.6 m. This layer classifies as a medium dense 

silt. Since the KD value is still high in this layer, it is likely that the layer is overconsolidated. The 

layer is classified as an overconsolidated silt. The boundary between layer A22 and Layer B1 is 

evidenced by a decrease in ID, ED, and KD. 

Layer B1 

Layer B1 is located at depths between 3.6 m and 4.0 m. Based on figure A-8, this layer classifies as 

a medium consistency silty clay to clayey silt. The KD value is greater than 5, which is indicative of 

a slightly overconsolidated material. This layer is classified as a slightly overconsolidated varved 

silty clay to clayey silt. The boundary between layer B1 and Layer B2 is evidenced by a decrease in 

ID, ED, and KD. 

Layer B2 

Layer B2 is located at depths greater than 4 m to the end of the sounding (20 m). This layer 

classifies as a soft clay based on figure A-8. 

Summary 

Results indicate that the primary difference between the methods presented is in the resolution of the 

thickness and soil behavioral characteristics of the upper overconsolidated clay crust. The most 

detail is provided by the CPTu and DMT, followed by the CPT and SPT with disturbed sampling. 

Any one of these methods, however, provides sufficient detail to develop a subsurface profile for use 

in most geotechnical design analyses. It is likely that for settlement and stability analyses, it would 

only be necessary to perform analyses assuming a two-layer system, that being an overconsolidated 

clay crust overlying soft clay. Where detailed stratigraphic profiling is required (e.g., identification 

of liquefiable materials or slip surface), the CPTu, CPT, or DMT are relatively fast and inexpensive 

as compared to continuous sampling. 

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 A-12 

BASELINE DATA COLLECTION AND INTERPRETATION 

Index Data Interpretation 

In addition to data collected as part of the field investigation, the following laboratory data should be 

collected for disturbed and undisturbed soil samples: (1) natural moisture content; (2) visual 

identification; (3) dry unit weight; and (4) Atterberg limits. A summary of these data has been 

provided on figure A-2 for Boring B1, B2, and B3. 

Calculation of Overburden Stresses 

The vertical effective stress in the ground, σvo′, is illustrated graphically by plotting effective stress 

with depth. This diagram is sometimes referred to as a “Po diagram”. Pressure diagrams are shown 

for boring location B1 in figure A-9. This diagram also shows calculated total stresses and pore 

pressures in the ground. The ground water table is located approximately 1 m below the ground 

surface. 

0

5

10

15

20

25

30

35

40

Depth (m)

Brow n-Gray Varved Clay

Gray Varved Clay (CL)

Gray Varved Silt (ML)

Brow n-Gray Varved Silt

0

5

10

15

20

25

30

35

40

0 50 100

Moisture Content

0

5

10

15

20

25

30

35

40

0 500 1000

Pressure Diagram

tot

stress

u

eff

stress

0

5

10

15

20

25

30

35

40

7 10 13 16

γ dry (kN/m3

)

Figure A-9. Index properties and “Po diagram” for Boring B1. 

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 A-13 

To develop the diagram for the example soil profile, the average total unit weight of each soil layer 

in the profile was evaluated using laboratory results. For each layer, the total unit weight (γt) was 

calculated using laboratory-measured dry unit weight (γd) and natural moisture content (wn) values 

according to the equation: 

)w1( dt +γ=γ n (Equation A-1) 

For the example profile, it is sufficient to evaluate an average total unit weight value for the upper 

crust layer and the lower soft varved clay layer. Typically, scatter in unit weight data is small and 

average values can be used for design analyses. Using the information in figure A-2, the average 

total unit weight for the upper crust layer is 17 kN/m3

, and the average unit weight for the lower soft 

varved clay layer is 16.3 kN/m3. These values were used to develop the Po diagram. 

CONSOLIDATION PROPERTIES OF SOIL 

Overview

In this section, laboratory and in-situ testing methods to evaluate consolidation properties are 

presented. These properties relate to primary and secondary consolidation of clayey soils. Firstorder predictors (e.g., correlations to simple index properties) for consolidation properties are 

presented as well as more detailed methods for evaluating site-specific design properties. 

Selection of Undisturbed Samples for Laboratory Consolidation Tests

Information on subsurface stratigraphy and basic index properties should be used to select the 

number and depths of undisturbed samples for laboratory consolidation testing. The number of 

samples should be selected to facilitate the development of a profile of preconsolidation stress, σp′, 

with depth. For the example profile, at least two consolidation tests should be performed for the 

soils in the crust layer (between approximately 1 and 5 m), one test just below the transition depth 

from the crust to the underlying varved clay (at approximate depth of 5 to 6 m) and then at 3 to 5-m 

intervals within the varved clay deposit. 

Evaluation of Preconsolidation Stress from Laboratory Testing

Standard one-dimensional consolidation tests (ASTM D2435) were performed on specimens from 

undisturbed tube samples at the approximate depths noted above. During sampling, transportation, 

extrusion, and trimming, samples must be carefully handled to prevent disturbance. For the 

consolidation test results presented here, it is assumed that the test has been performed over the 

required range of vertical stresses. These vertical stresses should be selected based on the specific 

loads imparted to the soil for the project conditions. 

Typically, testing will incorporate a loading schedule with a load increment ratio (LIR) of one (1). 

This implies that each successive load is twice as large as the previous load. To obtain quality data, 

the engineer should estimate the σp′. With σp′ estimated, smaller load increments can be specified 

for stress levels near σp′. This will facilitate a more accurate assessment of the σp′. It is 

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recommended that an unload-reload cycle be performed just before σp′ and recompression behavior 

be evaluated from this as opposed to using the initial portion of the consolidation curve up to σp′. 

Figure A-10 shows the evaluation of σ′p for a specimen using the Casagrande method and figure A11 shows the evaluation of σ′p for the same specimen using the strain energy method. Results 

summarized in table A-4 indicate that σ′p values calculated from the strain energy method are at least 

equal to those calculated from the Casagrande method. Also shown on table A-4 are measures of 

sample disturbance using the sampling disturbance index described in chapter 4. Results show that 

where disturbance measures indicate that the sample is relatively undisturbed (i.e., rankings of VG 

or AT), that the Casagrande σ′p is no less than 90 percent of the σ′p value from the strain-energy 

method. Where the disturbance measures indicate that the sample is likely to be disturbed (i.e., 

ranking of DT), then the Casagrande σ′p value was as low as 75 percent of the strain-energy value. 

Where disturbance is relatively large, graphical techniques such as Schmertmann’s method (see 

Holtz and Kovacs, 1981) may be used to develop a “field-corrected” consolidation curve. 

Alternatively, the observations from this example support the recommendation to use the strainenergy method for all evaluations, especially when sample disturbance is notable. 

Figure A-10. Evaluation of σp′ using the Casagrande method. 

Archived Document
A-15 

Table A-4. Summary of laboratory testing on varved clay samples. 

Test # Block # Depth wn LL PI ρdry eo σvo' σp' σp' OCR OCR ε at SQR Ccε Cc 

(m) (kPa) Casa SE Casa SE σvo'

- - 1.52 23 37.4 11.1 - 0.667 14.7 160 170 10.9 11.6 0.1 VG 0.064 0.11 

137 F1-T1-E3 2.72 39 44 13.6 1.31 1.21 39.1 250 - 6.4 - 1.5 DT 0.16 0.35 138 F1-T1-E3 2.72 38.3 44 13.6 1.31 1.21 39.1 280 - 7.2 - 1.1 DT 0.13 0.29 

- - 2.72 39 44 13.6 - 1.131 39.1 300 320 7.7 8.2 1.4 DT 0.192 0.41 - - 3.05 33 37.4 9.9 - 0.957 50.1 220 265 4.4 5.3 2.6 DT 0.215 0.42 

120 F1-T3-E1 3.3 41.2 46.4 16.5 1.32 1.20 43.8 330 - 7.5 - 1.0 AT 0.17 0.37 121 F1-T3-E1 3.3 38.4 46.4 16.5 1.32 1.20 43.8 420 - 9.6 - 0.8 AT 0.15 0.33 140 F1-T5-E1 4.27 42.4 42.5 13 1.19 1.44 50.7 250 - 4.9 - 1.0 AT 0.21 0.51 139 F1-T5-E1 4.27 46.7 42.5 13 1.19 1.44 50.7 260 - 5.1 - 1.4 DT 0.19 0.46 141 F1-T7-E1 5.26 49.4 40 11.9 1.13 1.57 57.8 170 - 2.9 - 1.3 AT 0.21 0.54 142 F1-T7-E1 5.26 48.9 40 11.9 1.13 1.57 57.8 150 - 2.6 - 2.2 AT 0.16 0.41 

- - 6.1 60 46 17.5 - 1.74 59.1 200 175 3.4 3.0 1.1 DT 0.221 0.61 

144 F1-T9-E1 6.3 56 45.9 17.5 1.08 1.69 65.1 135 - 2.1 - 1.1 AT 0.21 0.56 143 F1-T9-E1 6.3 59.8 45.9 17.5 1.08 1.69 65.1 135 - 2.1 - 0.9 VG 0.17 0.46 

- - 6.3 56 45.9 17.5 - 1.624 65.1 125 135 1.9 2.1 1.0 VG 0.22 0.58 

145 F1-T11-E1 7.34 52.9 44.7 12.8 1.16 1.50 71.9 155 - 2.2 - 1.9 AT 0.22 0.55 146 F1-T11-E1 7.34 46.7 44.7 12.8 1.16 1.50 71.9 140 - 1.9 - 2.2 AT 0.15 0.38 117 F1-T13-E1 8.61 61 49.8 18.2 1.07 1.71 80.2 160 - 2.0 - 2.7 AT 0.19 0.51 147 F1-T13-E1 8.61 57.2 49.8 18.2 1.07 1.71 80.2 135 - 1.7 - 2.4 AT 0.21 0.57 

- - 9.14 63.4 47 16.9 - 1.839 78.9 120 160 1.5 2.0 3.7 AT 0.185 0.53 - - 10.67 59.2 50 20.3 - 1.717 91.1 105 - 1.2 - 6.0 DT 0.196 0.53 - - 12.19 65 49.2 18.7 - 1.885 103.4 111 - 1.1 - 6.2 DT 0.169 0.49 - - 18.29 43 44.4 14.5 - 1.247 128.9 120 120 0.9 0.9 6 DT 0.258 0.58 - - 21.3 39.9 - NP - 1.157 161.3 160.2 - 1.0 - 5.05 DT 0.049 0.11 - - 24.38 56.1 - NP - 1.627 194.4 250 - 1.3 - 10.5 DT 0.185 0.48 - - 27.43 32 - NP - 0.928 223.9 205 260 0.9 1.2 4.2 DT 0.084 0.16 

Archived Document
 A-16 

Figure A-11. Evaluation of σp′using the strain energy method. 

In-Situ Test Methods to Evaluate σ′p

In this section, estimates of σ′p using in-situ testing methods are presented. Recommended specific 

equations are as follows: 

Cone Penetration Test: σp' = 0.33 (qT - σvo) (Equation A-2) 

Type 2 Piezocone (shoulder element): σp' = 0.53 (u2 – uo) (Equation A-3) 

Flat (Plate) Dilatometer Test: σp' = 0.51 (p0 - uo) (Equation A-4) 

Field Vane Test: σp' = 3.54 (su,VST) (Equation A-5) 

Figure A-12 shows a profile of σp′ based on the aforementioned correlations for in-situ tests and the 

profile presented previously. Also shown on this figure are the results of the laboratory oedometer 

tests. 

Calculated values of σp′ compare favorably to the values obtained from the laboratory testing, except 

as noted subsequently. 

Archived Document
 A-17 

• Since the piezocone with a shoulder element (i.e., CPTu2) measures negative penetration 

pore water pressures in the upper 4 m, the correlation for σp′ is not valid in this region. 

• Each correlation tends to underpredict values of σp′ for stiff fissured clays. The use of these 

correlations for stiff fissured clays is therefore conservative. 

0

5

10

15

20

25

30

0 100 200 300 400

PRECONSOLIDATION STRESS (kPa)

DEPTH (m)

CPTU1 - (U2 corr.)

CPTU1 - (Qt corr.)

DMT4 (Po corr.)

FVT-1 (Su corr.)

Oedometer

(Casagrande)

Oedometer (StrainEnergy)

Figure A-12. Profile of σp′ based on correlations with in-situ testing devices. 

In figure A-12, the average trend line using equations A-2 through A-5 is plotted. For each of these 

correlations, the standard deviation (SD) is approximately 100 kPa (see figures 53, 55, 56, and 58 for 

SD values). The results in figure A-12 indicate that the proposed first-order correlation for σp′

which is based on a large database of clay soil sites, is appropriate for the varved clay soils of this 

site. For each test device, the resulting correlated σp′ value was within ±1 SD of a profile drawn 

through the laboratory testing results. 

Archived Document
 A-18 

The utility of both in-situ testing devices and laboratory oedometer testing could be realized for a 

reasonably large project where several boring locations would be required. If, for example, a project 

required twenty geotechnical test borings to achieve appropriate coverage for comprehensive 

characterization, it likely would be faster and more cost-effective to replace borings with additional 

locations for in-situ testing such as CPT or DMT. In addition, as shown previously, additional data 

such as stratigraphic profiling can be provided. By comparing laboratory-measured σp′ values with 

an in-situ testing device result, the correlations previously described could be easily adjusted to more 

closely match the laboratory-measured results for σp′. The value of the in-situ testing device is that 

comparisons of in-situ testing results from additional locations could be compared to that of the 

baseline location to verify similar conditions. If trends in the data were similar, but correlated σp′

values were higher than those from the baseline, then this information could be used to justify a 

different σp′ profile with depth at the other location. If trends in the data were very different, then 

additional geotechnical borings and undisturbed sampling could be performed at that location and 

additional laboratory consolidation tests performed. 

Evaluation of Cc and Cr

Overview 

Compression parameters Cc and Cr can be interpreted by correlation or by direct laboratorymeasurement. The value of Cc is evaluated by drawing a best-fit tangent line to data on an e-log σ′vc

representation of consolidation data along the virgin (i.e., part of curve where stresses are greater 

than σ′p) portion of the curve. The modified compression index, Ccε, is evaluated similarly on a plot 

of εv-log σ′vc. An example evaluation of Ccε for the example soil profile is shown on figure A-13. 

Data Interpretation 

Numerous correlations relating simple soil index properties (e.g., LL, wn) to Cc and Ccε are available 

in the literature for silts and clays. These correlations are often used for first-order predictions of 

settlements, but should not be relied upon for final design, unless the correlation has been developed 

using site-specific laboratory consolidation test data. For most clays, values of Ccε tend to increase 

with increasing moisture contents, however, significant scatter exists for the data. Several empirical 

correlations are shown in figure A-14 along with actual data from the example soil profile. A review 

of figure A-14 indicates that correlations proposed by Terzaghi and Peck (1967) and Lambe and 

Whitman (1969) generally underpredict compression parameter values (i.e., Cc or Ccε) as compared 

to laboratory-measured values. On average, the correlation developed by Azzouz et al. (1976) for 

Chicago clays provides a reasonable approximation of Cc. 

It is interesting to note that the correlation developed for Hackensack Valley varved clays, which 

should be more representative for the example varved clay deposit, on average, overpredicts 

laboratory-measured Ccε values. Figure A-15 shows the results of an extensive investigation of the 

Connecticut Valley varved clays. This figure depicts the large scatter that exists in Ccε data for this 

deposit. Figures A-14 and A-15 demonstrate that significant uncertainty may exist when using 

published correlations for clay compression indices to perform design calculations. 

Archived Document
 A-19 

Figure A-13. Sample graphical evaluation for Ccε and Crε. 

Archived Document
 A-20 

0.00

0.10

0.20

0.30

0.40

0.50

0.60

0.70

0.80

0 20 40 60 80

wn

Cc

Azzouz et al., 1976

Cc = 0.01wn

Chicago Clays

0.00

0.10

0.20

0.30

0.40

0.50

0.60

0.70

0.80

0 20 40 60 80 100

LL

Cc

Terzaghi & Peck, 1948

Cc = 0.009 (LL-10)

Undisturbed Clays

0

0.1

0.2

0.3

0.4

10 30 50 70

wn

Ccε

Lambe & Whitman, 1969

Ccε = f (wn)

Variety of Clays

0

0.05

0.1

0.15

0.2

0.25

0.3

0.35

0.4

0 0.5 1 1.5 2

eo

Ccε

Azzouz et al., 1976

Ccε = 0.156eo +0.0107

Saxena et al., 1978

Ccε = 0.37(eo-0.65) 

Hackensack

Valley Varved

Clay

Figure A-14. Empirical correlations of Cc and Ccε. 

Archived Document
 A-21 

Figure A-15. Secondary compression results for Connecticut Valley varved clays investigation. 

Cαε

Ccε

Tentative Recommendation 

Cαε (%) = 4 Ccε

Range for Inorganic Clays 

from Ladd (1973) 

Cαε (%) = (4±1) Ccε

Archived Document
 A-22 

Figure A-16 shows the laboratory-measured Ccε versus depth. For the varved clay, Ccε ranges from 

0.07 to 0.22 with an average value of 0.19. Interpreting a value for design should be based on a 

rational assessment of the data. The objective is to assign a Ccε value to each behaviorally different 

subsurface layer or to assign some representative value for the entire subsurface. Assessments to be 

made in evaluating the Ccε data include: (1) depth ranges where the material is more silty or sandy as 

compared to other depth ranges; (2) depth of transition from crust layer to underlying varved clay 

deposit; and (3) assessment of sampling disturbance. 

The variability in Ccε as shown in figure A-16 is likely the result of: (i) some samples being more 

disturbed than others; and (ii) the possibility that, due to the varved nature of the deposit, there is 

relatively significant variation between samples as to the amount of silt, sand, and clay. Lower 

values of Ccε would be expected for samples that contain higher amounts of silt and sand as 

compared to those with a higher percentage of clay. 

0

5

10

15

20

25

30

0.05 0.1 0.15 0.2 0.25

Modified Compression Index (Ccε )

Depth (m)

Disturbed Test (DT)

Acceptable Test (AT)

Very Good Test (VG)

Figure A-16. Laboratory-measured Ccε vs. depth. 

It is inferred from the calculated sample disturbance indices for samples from the overconsolidated 

clay crust (i.e., depths less than approximately 4 m), that 5 of the 9 tests are designated as disturbed. 

This is likely due to the fissured nature of this material and that relatively large volume changes may 

have occurred during the initial portion of the test to close the fissures. It appears as though the 

single value of Ccε = 0.07 at 1.5 m may be an outlying data point. For the normally consolidated to 

lightly overconsolidated portion of the profile (i.e., for depths greater than approximately 4 m), the 

vast majority (9 out of 10) of the samples between a depth of 4 and 8.6 m are designated as 

acceptable to very good with respect to the sample disturbance index. For this depth range, Ccε

Archived Document
 A-23 

ranges from 0.15 to 0.22 with an average value of 0.20. The three samples below a depth of 8.5 m 

were designated as disturbed and have an average Ccε of 0.18. The sample at a depth of 27.5 m was 

mostly silt and not considered to be representative of the varved clay. 

Since the varved clay deposit is “consistently heterogeneous” in that the profile contains thin layers 

of silt and clay, it is appropriate to use an average value for Ccε for design settlement analyses. 

Based on this, a value of 0.20 is appropriate for Ccε for the entire profile. The average 

recompression ratio is 0.03 ± 0.01. For design, it would be reasonable to select a value for design of 

0.04 for the entire varved clay and varved silt profile. 

Laboratory Evaluation of cv

Figure A-17 shows a plot of calculated cv values for reload and virgin compression for five varved 

clay samples tested in one-dimensional compression. The cv values are plotted as a function of 

normalized stress that is defined as σvc′ /σp′, where σvo′ is the vertical effective stress in the ground 

at the depth where the sample was taken. Each sample exhibits the same trend of cv significantly 

reducing for load levels in the normally consolidated range (i.e., for σvc′/σp′>1). Scatter in the 

overconsolidated range is typical and may result from the following: (1) consolidation occurs quite 

rapidly at these load levels making the determination of the time for the end of primary 

consolidation difficult; and (2) in very stiff clays, fissures may exist at low stress levels which will 

affect drainage rates. Some of the inherent variability associated with evaluation of this parameter 

can be minimized by concentrating the interpretation on values corresponding to a reload cycle and 

to values associated with virgin compression. Values from initial loading should not be used due to 

inevitable sample disturbance effects. Data from a reload cycle is likely more representative of the 

“field curve”. 

0

2

4

6

8

10

12

14

16

0.01 0.1 1 10

σvc/σ'p

cv (cm2/s *10-3)

B142 - depth = 15.1 m

B174 - depth = 10.5 m

B182 - depth = 12 m

B174 - depth = 13.6 m

B182 - depth = 7.4 m

Figure A-17. cv values from laboratory consolidation tests. 

Archived Document
 A-24 

A rational approach to selecting a cv value would include first assessing whether the applied loads 

are sufficiently high to consolidate the compressible layer to stress levels beyond σp′. If stress levels 

after consolidation are below approximately 0.8σp′, an average cv value associated with stress levels 

below 0.8σp′ should be used for design. If the compressible layer will consolidate to stress levels in 

excess of σp′ or greater than 0.8σp′ as described here, then the average cv value for the range of 

stresses from σp′ to the final vertical effective stress in the ground should be used. This value, 

however, should not exceed the average value obtained for the overconsolidated stress range (i.e., 

less than 0.8σp′). 

The data shown in figure A-17 indicate that measured cv values for the sample from B174 at a depth 

of 10.5 m are greater than those for the other four samples, although the difference is much less in 

the normally consolidated range. This may be due to more silt in the sample as compared to other 

samples, however, such scatter can be expected even for similar deposits due to the evaluation 

method used to obtain cv. These data indicate that a reasonable average value for the normally 

consolidated range is 0.0008 cm2

/s and, for the overconsolidated range, a value of 0.003 cm2/s is 

reasonable. 

Evaluation of ch from CPTu Dissipation Data

In this section, dissipation test results from piezocone testing are used to estimate the horizontal 

coefficient of consolidation, ch, in the varved clay profile at two different depths in the deposit. To 

perform a dissipation test, cone penetration is paused at a test depth and pore pressures are recorded 

with time. Pore pressure dissipation curves for the two dissipation tests are shown on figure A-18. 

The procedures used to evaluate ch have been provided in section 5.4.4. 

To calculate ch using a piezocone, the modified time factor, T* must be evaluated. The parameter 

T* is related to the degree of consolidation, U. Values of T* are presented in section 5.4.4 (see table 

27) for various stages of dissipation. The parameter U has been defined in equation 47. Figure A-19 

shows normalized excess pore pressure curves for the two dissipation tests. 

Archived Document
 A-25 

0

100

200

300

400

500

600

0.01 0.1 1 10 100

Time (min)

Porewater Pressure Behind Tip (u2)

Depth = 6.68 m

Depth = 17.65 m

Figure A-18. CPTu2 pore pressure dissipation curves. 

0.0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

0.8

0.9

1.0

1.1

0.01 0.1 1 10 100

Time (min)

Normalized Excess Pore Pressure, U

Depth = 6.68 m

Depth = 17.65 m

Figure A-19. Normalized CPTu2 dissipation curves 

The parameters used for this example to evaluate ch are shown in table A-5. A standard cone with a 

projected tip area of 10 cm2 and a diameter of 35.6 mm was used in the evaluation. Tables A-6 and 

A-7 present calculations of ch at various degrees of consolidation using normalized curves for the 

two depths presented above. Values for ch can be estimated for various degrees of consolidation, 

Archived Document
 A-26 

however, evaluations for design should be made based on a degree of consolidation of at least 50 

percent. The parameter ch is calculated according to the following: 

( )

t

IaT

c r

h

2 * = (Equation A-6) 

Table A-5. Parameters used for analysis of dissipation data. 

Depth 

(m) 

Filter 

Location 

Cone 

Radius, a 

(cm) 

Plasticity 

Index 

(PI) 

OCR 

Rigidity 

Index, Ir

(figure 60) 

6.68 u2 1.78 10 2 230 

17.65 u2 1.78 15 1 210 

Table A-6. Evaluation of ch from CPTu2 dissipation data at 6.68 m 

Normalized 

Excess Pore 

Pressure, 

U 

T* for u2 

element 

location 

Cone 

Radius, a 

(cm) 

Square Root 

of Rigidity 

Index, 

r I

Time to 

Normalized Excess 

Pore Pressure, U 

(min) 

ch 

(cm2

/s)*10-3

0.8 0.038 1.78 15.2 1.9 16 

0.7 0.078 1.78 15.2 3.1 20 

0.6 0.142 1.78 15.2 4.6 25 

0.5 0.245 1.78 15.2 6.9 29 

0.4 0.439 1.78 15.2 10.4 34 

0.3 0.804 1.78 15.2 15.9 41 

0.2 1.60 1.78 15.2 26 49 

Table A-7. Evaluation of ch from CPTu2 dissipation data at 17.65 m 

Normalized 

Excess Pore 

Pressure, 

U 

T* for u2 

element 

location 

Cone 

Radius, a 

(cm) 

Square Root 

of Rigidity 

Index, 

r I

Time to 

Normalized Excess 

Pore Pressure, U 

(min) 

ch 

(cm2/s)*10-3

0.8 0.038 1.78 14.5 0.75 39 

0.7 0.078 1.78 14.5 1.4 43 

0.6 0.142 1.78 14.5 2.4 45 

0.5 0.245 1.78 14.5 3.8 49 

0.4 0.439 1.78 14.5 5.6 60 

0.3 0.804 1.78 14.5 8.5 72 

0.2 1.60 1.78 14.5 13.9 88 

Archived Document
 A-27 

A comparison of the ch values reported in tables A-6 and A-7 to the cv values from figure A-17 for 

laboratory testing on varved clay samples indicates that the horizontal coefficient of consolidation 

may be on the order of 50 times greater than the vertical coefficient of consolidation. Typically, ch

is greater than cv in clays due to the presence of localized pockets of higher permeability material. 

In the case of varved clays, this affect is magnified due to the continuous nature of the silt and sand 

varves. 

The selection of an appropriate value for a coefficient of consolidation (i.e., cv or ch) will depend on 

the design application. For large area fills (i.e., where the width of the loaded area is greater than 

say 2 times the thickness of the cohesive soil deposit), a cv value should be used. Where wick drains 

are used or for the evaluation of pile driving induced pore pressures, a ch value should be used. For 

each specific design, the engineer should evaluate the drainage distances and likely direction of 

consolidation water flow and select a cv, ch, or intermediate value accordingly. Alternatively, the 

best assessment of time rate of settlement (or time for consolidation) can be achieved using 

piezometers. 

Evaluation of Cα

In geotechnical design analyses, it is assumed that secondary settlement occurs after primary 

consolidation is completed. As noted earlier, secondary compression settlements may be relatively 

large for organic soils. For the soil profile used in this example, secondary compression settlements 

would be expected to be insignificant for typical highway projects since the soils are not organic. 

However, the procedures to calculate Cαε are the same regardless of the soil type. This example 

demonstrates the evaluation and interpretation of Cαε. 

Figure A-20 shows calculated values of Cαε versus σvc′ /σp′ for reload and virgin compression for the 

same five consolidation tests discussed in the section of this example on laboratory evaluation of cv. 

For each load increment, the time to complete primary consolidation was evaluated and Cαε was 

evaluated over a period after primary consolidation was completed. This plot indicates that values 

of Cαε are affected by stress history in that maximum values for Cαε are evaluated for stress levels 

greater than the preconsolidation stress (i.e., at stresses corresponding to virgin compression). 

Therefore, to assess a value of Cαε to be used for design analyses, the final effective stress in the 

ground after primary consolidation is completed should be evaluated. If the final effective stress is 

less than approximately 0.8σp′, then an average value of Cαε evaluated in the overconsolidated range 

may be used for design. If final effective stresses in the ground exceed σp′, then it is conservative to 

select Cαε value corresponding to stresses in the range of 1 to 2 times σp′. 

For normally consolidated soils, the ratio of the coefficient of secondary compression to the 

compression index (Cα/Cc = Cαε/Ccε) is relatively constant for a given soil. On average, the value of 

Cαε/Ccε is 0.04±0.01 for inorganic clays and silts. This value may be used to assess the values from 

the laboratory tests. Figure A-21 shows a plot of Cαε versus Ccε for the five consolidation tests. The 

values of Ccε were obtained from the laboratory consolidation curves (not provided herein) and are 

representative of the slope of the strain vs. log σvc′ curve at each stress level. The average Ccε value 

from the five consolidation tests is 0.35. The results indicate a range for Cαε/Ccε from 0.02 to 0.04. 

This range of values is, on average, slightly less than that reported in the literature for inorganic 

clays and silts. It is noted that for samples that are disturbed, laboratory-measured values of Cα (or 

Archived Document
 A-28 

Cαε) may be underestimated in the normally consolidated range as compared to results from highquality undisturbed samples. For design, an average value corresponding to Cαε/Ccε = 0.03 for an 

average Ccε = 0.35 could be used, i.e., Cαε=1.05 (say 1 percent). This value is consistent with the 

range shown in figure A-20 for Cαε values in the range for σvc′ /σp′ > 1. 

0

0.2

0.4

0.6

0.8

1

1.2

1.4

0.01 0.1 1 10

σ'vc/σ'p

Cαε (%)

B142 - depth = 15.1 m

B174 - depth = 10.5 m

B182 - depth = 12 m

B174 - depth = 13.6 m

B182 - depth = 7.4 m

Figure A-20. Cαε values from laboratory consolidation tests. 

Cαε/Ccε = 0.04

Cαε/Ccε=0.02

0

0.05

0.1

0.15

0.2

0.25

0.3

0.35

0.4

0.45

0 0.2 0.4 0.6 0.8 1 1.2 1.4

Cαε (%)

Ccε

B142 - depth=15.1 m

B174 - depth =10.5 m

B182 - depth = 12 m

B174 - depth =13.6 m

B182 - depth = 7.4 m

Figure A-21. Comparison of Cαε and Ccε for laboratory consolidation tests. 

Archived Document
 A-29 

EVALUATION OF IN-SITU HORIZONTAL STRESSES 

For overconsolidated soils (clays, silts, sands, and gravels), Ko can be calculated according to the 

following equation: 

'sin

o OCR)'sin1(K φ φ−= (Equation A-7) 

Using results from undrained triaxial compression tests and one-dimensional consolidation tests, 

values of φ′ and OCR for the example soil profile can be substituted into equation A-7 to develop a 

profile of Ko with depth. This profile is shown on figure A-22. Also shown on this figure are 

profiles of Ko with depth based on in-situ testing devices. To develop these profiles using in-situ 

test data, the preconsolidation stress was calculated at each measurement point using equations A-2 

through A-5 for estimating preconsolidation stress from measured in-situ testing device parameters. 

From this, the OCR value was calculated and Ko was evaluated using equation A-7. 

0

5

10

15

20

25

30

0.0 0.5 1.0 1.5 2.0

Ko

DEPTH (m)

CPT (U2)

CPT (Qt)

DMT4

FVT-1

Oedometer (Casagrande)

Oedometer (Strain Energy)

Figure A-22. Ko value calculated using in-situ testing devices. 

Archived Document
 A-30 

For the soft normally to lightly overconsolidated clay layer, which exists below approximately a 

depth of 5 m, all methods provide reasonable values of Ko. The maximum difference between any 

two calculated Ko values at any depth is approximately 0.2. It is noted that the variation is caused by 

differences in calculated values of preconsolidation stress (and thus OCR). This variation should not 

result in a significant effect on design calculations that require estimates of Ko in soft to medium 

clays. 

The methods described in this section for evaluating Ko are well-suited for evaluations in normally 

to lightly overconsolidated clay soils and for intact (i.e., non-fissured) overconsolidated soils. In 

addition, the use of the in-situ testing method correlations likely results in lower values of Ko than 

actual values. Whether or not values that are less than actual values for Ko are conservative or 

unconservative needs to be evaluated based on the specific design application. Where a more 

accurate assessment of Ko is required for design, it is recommended that either laboratory triaxial 

tests on undisturbed samples subjected to Ko consolidation or in-situ self-boring pressuremeter 

testing be performed. 

UNDRAINED SHEAR STRENGTH 

Overview

The evaluation of undrained shear strength, su, for the soft to medium varved clay deposit is 

provided in this section. Specifically, laboratory strength data from CIUC and UU testing as well as 

derived su values from CPT, CPTu, DMT, and VST are presented. Based on these data, a 

recommended undrained strength profile is developed. 

Baseline Data for Undrained Strength Evaluation

The methods presented in chapter 5 indicate that the preconsolidation stress profile should be 

developed for a cohesive deposit and that this information is useful in evaluating undrained strength. 

For this section of the example, the profile of preconsolidation stress shown in figure A-23 is used. 

Also shown on this figure is a plot of the vertical effective stress with depth (i.e., the “Po diagram”). 

Undrained Shear Strength Evaluation from Laboratory Shear Strength Tests

Table A-8 shows a summary of the triaxial tests (UU and CU) performed and the results of this 

testing. For the CU tests, the isotropic consolidation stress used was selected to be equal to the 

vertical effective stress in the ground. Therefore, the undrained strength measured in the CU tests 

corresponds to the depth at which the sample was taken. Example stress-strain and pore pressurestrain curves are shown below for a UU and CU test. 

Archived Document
 A-31 

0

5

10

15

20

25

30

0 100 200 300 400

σ'p (kPa)

Depth (m)

Effective Stress

0

5

10

15

20

25

30

12345

OCR

Figure A-23. Po diagram developed from 1-D consolidation data. 

Table A-8. Summary of triaxial test data. 

Boring Sample Test 

No. 

Test 

Type 

Depth 

(m) 

w 

(%) 

LL PI γdry Cell 

Stress 

(kPa) 

Peak 

su 

(kPa) 

∆uf 

(kPa) 

σp′

(kPa) 

OCR φ′

S-43 UP-1C R-1 CIUC 8.5 41.4 46 21 12.9 119.6 62.8 56.4 275 2.30 31.3 

S-52 UP-3B R-4 CIUC 9.4 52.5 56 30 11.4 88.3 44.1 47.1 212 2.40 30.8 

S-43 UP-4C R-2 CIUC 17.1 51.6 52 25 11.6 191.2 70.6 109.8 200 1.05 28.6 

S-43 UP-6C R-3 CIUC 23.2 49.2 52 26 11.9 239.3 77.5 147.1 250 1.04 27.5 

BR-35 UP-1 Q14 UU 6.4 53.3 56 29 11.1 71.6 27.5 NA 300 4.2 NA 

S-52 UP-2 Q-9 UU 6.4 52.3 60 32 11.1 59.8 44.1 NA 300 5.0 NA 

S-56 UP-2 Q-8 UU 6.4 48.7 49 25 11.7 71.6 33.3 NA 300 4.2 NA 

BR-16 UP-1 Q-1 UU 6.7 44.9 51 25 12.1 100.0 34.3 NA 300 3.0 NA 

BR-24 UP-1 Q-6 UU 7.9 59.4 60 33 10.3 71.6 27.5 NA 250 3.5 NA 

BR-8 UP-1 Q12 UU 7.9 51.6 50 25 11.1 83.4 34.3 NA 250 3.0 NA 

BR-11 UP-1 Q-7 UU 9.4 52.3 50 25 11.1 100.0 37.3 NA 275 2.7 NA 

BR-16 UP-2 Q-2 UU 11.0 54.6 53 27 10.9 139.3 34.3 NA 250 1.8 NA 

S-43 UP-2 Q-4 UU 11.0 48.1 52 25 11.9 143.2 42.2 NA 250 1.7 NA 

S-52 UP-4 Q10 UU 12.5 47.6 45 21 11.7 107.9 31.4 NA 250 2.3 NA 

BR-35 UP-4 Q15 UU 15.5 51.5 48 23 11.1 143.2 19.6 NA 225 1.6 NA 

S-43 UP-4 Q-5 UU 17.1 51.6 52 25 11.3 191.2 32.4 NA 200 1.0 NA 

BR-11 UP-3 Q11 UU 21.6 41.7 39 17 12.6 196.1 39.2 NA 250 1.3 NA 

Archived Document
 A-32 

Laboratory Shear Strength Testing

Unconsolidated Undrained (UU) Triaxial Test 

Unconsolidated Undrained (UU) tests were performed on samples of a Connecticut valley varved 

clay associated with a highway project in Connecticut. A stress–strain strain curve from an 

unconsolidated undrained test is presented in Figure A-24 (a). The undrained shear strength from 

this test would be half the maximum principal stress difference, ∆σmax. For this test ∆σmax = 160.8 

kPa, so the undrained shear strength from this UU test is 80.4 kPa. 

Since shear strength will vary for different test types, it is best to designate the shear strength using 

the acronym for the associated laboratory or field test. For the unconsolidated undrained (UU) test, 

the undrained shear strength is designated su, UU. 

Total stress paths can be presented for an unconsolidated undrained test with the total stress state 

variables q and p defined as: 

 ( )

2

−σσ 31 q = and ( )

2

+σσ 31 p = (Equation A-8) 

Figure A-24. Stress – strain curve and stress path for a UU test on Connecticut Valley varved clay 

specimen. 

A total stress path for a UU test is presented in figure A-24 (b). Since q is equal to the principal 

stress difference divided by 2, it can be seen in figure A-24 (b) that the failure point from the stress 

path plot is equal to the undrained shear strength, su, UU = 80.4 kPa.

Isotropically Consolidated Undrained Compression (CIUC) Triaxial Test 

Isotropically consolidated undrained compression (CIUC) triaxial tests with pore pressure 

measurements were also performed on samples of the Connecticut valley varved clay. For a triaxial 

compression test, the undrained shear strength will be determined as half the maximum principal 

stress difference, ∆σmax. Figure A-25 (a) shows a stress strain curve from a CIUC test of 

Connecticut valley varved clay. For this test ∆σmax = 160.7 kPa, so the undrained shear strength 

0

20

40

60

80

100

120

140

160

180

0.00 5.00 10.00 15.00 20.00 25.00

Axial Strain, ε v (%)

Principal Stress Difference, (kPa)

0

10

20

30

40

50

60

70

80

90

0 50 100 150 200 250 300

p (kPa)

q (kPa)

UU TSP

Archived Document
 A-33 

from this CIUC test is 80.4 kPa. For the isotropically consolidated undrained compression (CIUC) 

test, the undrained shear strength is designated su, CIUC. 

Figure A-25. (a) Stress – strain curve and stress path for a CIUC test on a Connecticut Valley 

varved clay specimen. 

The total and effective stress paths for this CIUC test are shown in Figure A-25 (b). The effective 

stress path variable, p′, is equal to: 

p′ = p - ∆u (Equation A-9) 

The horizontal difference between the total and effective stress paths is the change in pore pressure 

(∆u) induced by the shearing of the soil and the triaxial test. 

Drained Strength Evaluation 

Data from the CIUC tests was used to estimate the drained friction angle, φ′. To define the failure 

envelope, values for p′ and q at failure were plotted. Figure A-26 shows the effective stress failure 

points for the CIUC tests, and the interpreted failure envelope. 

The failure envelope exhibits a break at about p′=120 kPa. Therefore, a friction angle of 31o

 is 

calculated for the low stress range and a friction angle of 18.2o

 with an effective cohesion intercept 

of 26 kPa is calculated for stress ranges higher than p′ = 120 kPa. 

Summary of Laboratory Shear Strength Testing Results and Design Shear Strength Profile 

Figure A-27 shows a plot of undrained strength vs. vertical effective stress for the UU and CU 

results. Included on this figure is the relationship given by equation 68 and 70. Recall that equation 

68 derived for the direct simple shear mode was developed based on a large database of shear 

strength testing in clays. This relationship uses the preconsolidation stress profile shown in figure 

A-23 to evaluate OCR and a drained friction angle of 18.2 degrees. The drained friction angle that 

should be used for the database approach (i.e., equation 70) should correspond to the direct simple 

0

10

20

30

40

50

60

70

80

90

0 50 100 150 200 250 300 350

p, p' (kPa)

q (kPa)

CIUC Effective

CIUC Total

∆ u

0

20

40

60

80

100

120

140

160

180

0 5 10 15 20 25

Axial Strain, ε v (%)

Principal Stress Difference, 

(kPa)

0

20

40

60

80

100

120

140

160

180

Pore Pressure (kPa)

Shear Stress

Pore Pressure

Archived Document
 A-34 

shear mode. For most soils, there will only be a minor difference between the drained friction angle 

from a TXC test and a DSS test. However, because of the horizontal stratification of the varved clay 

deposit, the friction angle from DSS loading will be significantly lower than that from TXC. 

Available data in the literature indicates that φ′ for the DSS mode for varved clays is on the order of 

16 to 20 degrees. Therefore, the interpreted value of 18.2 degrees was used herein. Figure A-28 

shows a plot of su/σvo′ vs. OCR. This plot indicates that the laboratory strength test measurements 

(both UU and CU) indicate that the soil can be characterized using a relationship similar to equation 

70. 

q = 0.517p

sinφ' = tan α'

φ' = 31o

q = 0.3122p + 24.902

sinφ' = tan α'

c' = a / cos φ'

c' = 26 kPa φ' = 18.2o

0

25

50

75

100

0 50 100 150 200

p' (kPa)

q (kPa)

Figure A-26. Effective stress properties from CIUC test data. 

The results from UU tests are consistently below those from CU testing and from equation 68. 

Based on laboratory testing alone, it appears as though an average profile line between the UU and 

CU strengths would be appropriate, but perhaps still on high side. The results of CU and UU testing 

for varved materials needs to be carefully considered relative to the design application. Because of 

the horizontal stratification of varved materials, it is likely that the lowest shear strength would be 

associated with shearing in a near-horizontal direction. Therefore, failure surfaces that develop in 

UU and CU tests may not be consistent with potential critical conditions for field applications. For 

this reason, it may be necessary to perform direct simple shear testing for projects involving varved 

clays and potentially marginal stability conditions. 

Archived Document
 A-35 

0

5

10

15

20

25

30

0 100 200 300 400

σp' (kPa)

Depth (m)

CIUC tubes

UU tubes

Effective Stress

0

5

10

15

20

25

30

12345

OCR

0

5

10

15

20

25

30

0 25 50 75 100

su (kPa)

CIUC

UU

0.23OCR^0.8

sinf'/2OCR^0.8

NC Strength

Figure A-27. Po diagram with OCR and su vs. depth. 

y = 0.148x0.83

R2 = 0.8189

93083 - UU

y = 0.3137x0.7007

R2 = 0.966

93083 - CIUC

0

0.2

0.4

0.6

0.8

1

1.2

1.4

1.6

1.8

1 10

OCR

su/ vo'

93083 - UU

93083 - CIUC

0.23 OCR^0.8

sinf'/2*OCR^0.8

(f'=18.2)

 2 3 4 5 6 7 8 9 

Figure A-28. Normalized strength property relationships for various shear modes. 

Archived Document
 A-36 

As previously discussed in chapter 5, the DSS is considered to represent an average condition 

between compression and extension loading and one which is applicable to most stability and 

bearing capacity evaluations. For most soils, the trend in strength is su(compression) > su (DSS) > su

(extension). However, for varved clay soils, the DSS mode represents the most critical (i.e., lowest) 

undrained strength because shearing occurs parallel to the varves. Therefore, a design shear strength 

envelope (which is intended to reflect the average undrained shear strength) for a varved clay 

deposit should be representative of an average between direct simple shear and compression. 

Therefore, based on figure A-27, it would be reasonable to assign a constant value of about 45 kPa 

for the undrained strength over a depth of 5 to 23 m. 

STRENGTH EVALUATION FROM IN-SITU TESTS 

Overview

The use of in-situ tests for evaluating undrained shear strength is presented in this section. In-situ 

test data used herein were obtained for a soft to medium varved clay deposit located near Amherst, 

Massachusetts. Relationships between in-situ test parameters and undrained strength have been 

summarized in table 33. The relationships presented in table 33 can be used for first-order 

evaluations of undrained strength. As discussed in section 5.6.5, for some in-situ testing devices, 

available correlations relate an in-situ testing parameter to a particular undrained strength that 

corresponds to a mode of shear. Herein, the direct simple shear mode is used. 

As discussed in section 5.6.5, the recommended approach to using in-situ test data to evaluate 

undrained shear strength for design is to use the data to develop a profile of σp′ (and thus OCR) 

using the relationships provided in section 5.4.3 (i.e., equations 41 through 46). From the OCR 

profile, an undrained strength profile can be developed using equation 68 or 70. This method is also 

illustrated in this section. 

Vane Shear Test

A summary of vane shear testing results is shown in table A-9. At each testing depth, the maximum 

torque, Tmax, required to turn the vane device was recorded. The undrained strength corresponding 

to peak conditions from the vane su(VST) is calculated as: 

( ) 3

max

7

6

D

T

s VSTu

π = (Equation A-10) 

where D is the diameter of the vane which equals 5.1 cm for these tests. The average plasticity 

index (PI) of the soil is less than 20 indicating that the reduction factor, μ, can be assumed to equal 

1.0. 

Archived Document
 A-37 

Table A-9. Vane shear data. 

Depth Vane 

Diameter, D 

Vane 

Height 

Torque su,VST 

(m) (cm) (cm) (Nm) (kPa) 

3.66 5.08 10.16 40.3 83.8 

4.27 5.08 10.16 25.7 53.5 

4.88 5.08 10.16 20.9 43.6 

5.49 5.08 10.16 16.4 34.1 

6.1 5.08 10.16 17.4 36.3 

6.71 5.08 10.16 16.8 34.9 

7.32 5.08 10.16 17.6 36.7 

7.92 5.08 10.16 17.2 35.9 

8.53 5.08 10.16 16.7 34.7 

9.14 5.08 10.16 16.0 33.3 

9.75 5.08 10.16 17.4 36.3 

10.36 5.08 10.16 16.1 33.6 

10.97 5.08 10.16 16.7 34.8 

11.58 5.08 10.16 16.9 35.2 

12.19 5.08 10.16 15.4 32 

12.8 5.08 10.16 16.5 34.4 

13.41 5.08 10.16 16.0 33.3 

14.02 5.08 10.16 14.5 30.1 

14.63 5.08 10.16 17.0 35.4 

15.24 5.08 10.16 11.9 24.8 

Note: μ = 1 for all test depths 

Piezocone Penetration Test (CPTu)

Results of a piezocone sounding were use to evaluate undrained strength according to the following 

correlations: 

( )

k

vot

CPTu N

q

s −σ = (Equation A-11) 

( )

um

om

CPTu N

uu

s u

∆

− = (Equation A-12) 

The correlation for su(CPT) can be compared to the DSS mode of shear with an assumed Nk value of 

15 (see table 33). The piezocone used was a type 2 (indicating that the filter element is located at 

the shoulder of the penetrometer). The correlation for su(CPTu) can be compared to the vane shear 

testing device using a N∆um factor equal to 7.9 (see table 33). 

Archived Document
 A-38 

Flat Plate Dilatometer (DMT)

A DMT was performed in the varved clay deposit as well. According to table 33, the KD parameter 

from a DMT can be correlated to undrained strength (for the DSS mode of shear) according to: 

( ) ( ) 25.1DMTu 14.0 σvo 5.0 K D s ′ = (Equation A-13) 

Comparison of Undrained Strength Results using In-situ Tests

Calculated undrained shear strengths from the correlations for VST, CPT, CPTu, and DMT are 

compared to laboratory DSS tests. The DSS test results were obtained for varved clay soils from the 

Amherst site. A review of figure A-29 (a) and (b) indicates the following: 

• The undrained shear strength estimated using the VST is approximately 50 percent greater 

than the laboratory DSS undrained strengths. Undrained strengths measured using the VST 

are typically greater than those from DSS tests. 

• The use of NK equal to 15 results in good agreements to the laboratory DSS data for depths 

greater than 4 m. At depths less than 4 m, the clay crust is fissured and it is likely that 

partial drainage occurs. Since the cone factor is based on analyses that consider undrained 

conditions, significant error, as is seen here, can be expected when using such a correlation 

in a soil that behaves in a partially drained manner during penetration of the penetrometer. 

• Use of a piezocone pore pressure factor N∆um = 7.9 results in calculated undrained strengths 

that are up to 50 percent greater than the laboratory DSS. This is likely due to the fact that 

the piezocone pore pressure factor of 7.9 is based on a database in which piezocone results 

are calibrated to vane shear test results. Drops in su at 1-m intervals correspond to 

dissipation of penetration prorewater pressure at pauses in penetration. These pauses are 

necessary for addition of rods to advance the penetrometer. 

• The DMT data matches laboratory DSS test data reasonably well. 

As previously noted, in-situ data can be correlated to σp′ for use in equation 70. Figure A-29 shows 

calculated profiles of undrained strength for the CPT, CPTu, DMT, and VST. Equations A-2, A-3, 

A-4, and A-5 were used to evaluate σp′ and the undrained strength was calculated according to 

( )' 2

'sin 8.0

u OCR vo s σ φ = ⋅ (Equation A-14) 

When compared to figures A-29(a) and A-29(b), the use of equation A-14 results in better agreement 

to the laboratory DSS for VST and CPTu and slightly better agreement for the CPT. 

Archived Document
 A-39 

0

3

6

9

12

15

0 25 50 75 100

su (kPa)

Depth (m)

VST

DSS

0

3

6

9

12

15

0 25 50 75 100

su (kPa)

CPT; Nk = 15

CPTu; Nu = 7.9

DMT

Laboratory DSS

0

3

6

9

12

15

0 25 50 75 100

su (kPa)

CPT

CPTu

DMT

VST

Laboratory

DSS

Figure A-29. Estimation of shear strength from in-situ tests: (a) VST; (b) empirical CPT, CPTu, & 

DMT based on correlations in table 33; and (c) based on equation A-14 with φ′=18.2o

.

a. c. b. 

Archived Document
A-40 

SOIL AND ROCK PROPERTY SELECTION EXAMPLE NO. 2 

Piedmont Residual Soil, Weathered Rock, and Rock 

INTRODUCTION 

This example focuses on evaluating residual soil properties including: (1) subsurface stratigraphy; 

(2) elastic and time dependent deformation characteristics; and (3) drained strength properties. 

Index properties, deformation modulus, and strength properties of the rock are also discussed. Field 

and laboratory data are presented, analyzed, and recommendations are provided on the selection of 

appropriate design values. 

For this example, the project site is located in the southwest portion of the Piedmont Province, in 

west-central Alabama (the Alabama site). The site is underlain by residual soils formed from the 

weathering of gneiss and schist. Rock data used for this example are from sites in the Atlanta, 

Georgia area. In residual soils, understanding the weathering profile is necessary in the 

interpretation of subsurface stratigraphy and engineering properties. Residual soil classification 

schemes and a generic weathering profile of a formation derived from gneiss and schist are 

presented in figure A-30. 

Sowers, 

1963 

Deere & Patton, 

1971 

Soil IA- Horizon 

5 < N < 50 IB- Horizon 

I. Residual 

Soil IC- Horizon, 

Saprolite Saprolite 

5 < N < 50 

IIA- Transition 

Saprolite to 

Weathered 

Rock 

Partially 

Weathered Rock, 

Alternate Hard & 

Soft Seams 

N > 50 

II. 

Weathered 

Rock IIB- Partially 

Weathered 

Rock 

Rock 

RQD > 75% 

III. Unweathered Rock 

RQD > 75% 

Figure A-30. Residual soil classification and typical gneiss and schist weathering profile 

(after Sowers and Richardson, 1983) 

The primary residual soil horizons based on Sowers (1963), Deere & Patton (1971), and Sowers & 

Richardson (1983) are the soil zone, the saprolite zone, partially weathered rock, and rock. The soil 

zone consists of a sandy upper leached zone and a more clayey oxidized accumulation zone 

(horizons IA- and IB- respectively). This soil horizon (I) is typically completely weathered with 

Archived Document
A-41 

little to no residual soil structure remaining. These layers are typically on the order of 1 meter thick 

in the Piedmont province, but may be thicker in tropical residual soils. The saprolite horizon (II) 

consists of highly variable, highly anisotropic materials, reflecting the structure of the original parent 

rock. This layer is composed of silty sands and sandy silts which may contain some mica. The 

presence of mica will lead to an increased void ratio and significantly affect the engineering 

properties of the material. Relic fractures and joints of the parent rock are maintained in the soil 

structure, as well as slickensided planes of weakness. The depth to and thickness of the partially 

weatherd rock zone is highly variable. More weathering typically occurs along fractures, and new 

fractures open as weathering continues. This will lead to alternating hard weathered rock and soft 

weathered soil zones until competent rock is reached. For classification purposes of residual soil 

profiles, competent rock is defined as material having a RQD greater than 75 percent. 

SUBSURFACE EXPLORATION AND TESTING PROGRAM 

A subsurface exploration in residual soils will be primarily concerned with estimating the extent of 

zones within the weathering profile. The thickness and variability of the saprolite layer will be of 

concern for most design problems, and the depth to weathered rock and depth to competent rock 

may be of concern depending upon the project specifications and local geology. Generally, SPTs on 

1.5-m intervals can provide information on the thickness of the layers, and depth to weathered 

bedrock. Rock coring should be used to confirm competent bedrock, when depth to and consistency 

of rock is a design parameter. Experience using in-situ tests, such as the PMT, CPT, CPTu, and 

DMT, have shown their applicability to design problems in Piedmont residual soils. The CPT and 

CPTu can provide additional information that can be used to develop a more detailed subsurface 

profile. Based on previous local experience, the PMT and DMT can be used to provide estimates of 

in-situ elastic deformation properties. 

DEVELOPMENT OF SUBSURFACE SOIL PROFILE FOR DESIGN 

General

At the west-central Alabama site, in-situ test methods were used to develop a subsurface profile. 

This section provides information on soil profiles developed from drilling and sampling information, 

piezocone data, and flat plate DMT data. SPT’s are performed at 1.5-m intervals, CPTu data is 

available every 5 cm, and the flat dilatometer tests were performed every 30 cm at this site. Results 

from series of each test type are considered individually within separate subsections. 

Use of SPT and Disturbed Sampling

A geotechnical boring with Standard Penetration Testing at the site is shown on figure A-31. Eight 

borings were performed at the site, SPTs were performed in four of the borings and the other borings 

were primarily used to collect Shelby Tube samples. The investigation was performed to assess soil 

conditions for installation and testing of 12-m long drilled shafts, thus borings significantly deeper 

than 12 m are not available at this site. The water table varies from 2 to 5 meters below ground 

surface. 

Archived Document
A-42 

Figure A-31. Boring log for Alabama site. 

Archived Document
A-43 

Index testing and SPT N-values are presented in figure A-32. Index tests results are summarized in 

table A-10. A CME automatic hammer was used to perform the SPTs, with an assumed energy 

efficiency of 90 percent. The energy efficiency should be calibrated periodically in accordance with 

ASTM D4633, and efficiency values should be obtained from the drilling subcontractor. The dry 

density of the soil ranges from 13 to 17 kN/m3

, with a majority of the data around 15 kN/m3. Low 

dry unit weight values, and thus high void ratios, may indicate the presence of mica. This will 

significantly affect compressibility properties of the soils, and thus the presence of mica based on 

visual identification should be noted in the boring logs as well as in laboratory data sheets. 

0

2

4

6

8

10

12

14

16

0 5 10 15 20 25

SPT N-value 

Depth (m)

B-1

B-3

B-4

B-6

A.

B.

0

2

4

6

8

10

12

14

16

10 12 14 16 18

γ dry (kN/m3

)

GWT

0

2

4

6

8

10

12

14

16

0 20 40 60 80 100

w(%) & Atterberg Limits

`

PL w LL

0

2

4

6

8

10

12

14

16

0 20 40 60 80 100

Grain Size Distribution

%P200

% Clay

Figure A-32. Summary index test data for samples from borings B-1, B-3, B-4, and B-6 at Alabama 

site. 

From hydrometer analyses, the clay content (finer than 0.002 mm) ranges from up to 30 percent clay 

in the upper 3 meters and below 15 percent clay for the remaining depth. The tested samples show 

between 27 and 85 percent fines (passing No. 200 sieve) from mechanical grain size analyses. Since 

the mean grain size for Piedmont soils is near the opening size for the U.S. No. 200 sieve (0.075 

mm), it appears as if there is significant variability in the fines content. The soil acts as a dual 

symbol material, (SM-ML) exhibiting characteristics of fine-grained soils (undrained behavior) and 

coarse-grained soils (drained) when subject to different types of loading. For most transportation 

projects, loading conditions in the Piedmont province will occur under drained conditions. 

While Atterberg limits data were obtained for several samples, about 50 percent of the samples 

tested at the site classified as non-plastic. 

Archived Document
A-44 

Table A-10. Index properties of Piedmont soils at the Alabama site. 

Boring Depth % 

Sand 

% 

Silt 

% Clay % 

Fines 

LL PL PI MC LI 

 (m) (%) 

B1 1.4 53.7 21 25.3 46.3 48 38 10 22.9 -1.51 

B1 1.4 47.5 30.2 22.3 52.5 48 38 10 22.9 -1.51 

B3 1.4 42.2 31 26.8 57.8 54 40 14 24.5 -1.11 

B3 1.4 44 28.6 27.4 56 56 40 16 24.5 -0.97 

B4 4.3 57.1 33.6 9.3 42.9 40 30 10 31.1 0.11 

B4 4.3 55.3 35.8 8.9 44.7 40 30 10 31.1 0.11 

B3 4.4 71.3 24.2 4.5 28.7 30 26 4 24.3 -0.43 

B1 4.5 53.8 38.3 7.9 46.2 45 40 5 35.8 -0.84 

B1 4.5 53.2 39 7.8 46.8 45 40 5 35.8 -0.84 

B6 5.9 61.2 32.3 6.5 38.8 39 31 8 28.8 -0.28 

B4 7.4 70.6 23.8 5.6 29.4 40 37 3 21.6 -5.13 

B4 7.4 68.8 25.9 5.3 31.2 40 37 3 21.6 -5.13 

B1 7.5 15.2 - - 84.8 44 39 5 26.4 -2.52 

B1 10.6 30.9 57.2 11.9 69.1 54 46 8 44.6 -0.18 

B3 12 50 38 12 50 40 36 4 38.1 0.53 

B4 12 25.4 - - 74.6 37 31 6 25.9 -0.85 

B6 12 58.4 37.9 3.7 41.6 37 34 3 27.1 -2.30 

B6 12 57.9 38.4 3.7 42.1 37 34 3 27.1 -2.30 

B6 15 66.9 28.4 4.7 33.1 34 29 5 25.6 -0.68 

B6 15 64.2 31.5 4.3 35.8 34 29 5 25.6 -0.68 

B1 15.2 10.5 - - 89.5 49 41 8 46.9 0.74 

 Average - 49.3 28.8 16.4 50.7 40.56 33.96 6.6 28.4 -1.07 

LL = Liquid Limit; PL =Plastic Limit; PI = Plasticity Index; MC = Moisture Content; LI = Liquidity Index 

A review of all boring logs and laboratory index test data at the site show about 3 to 4 meters of 

sandy silt with clay over silty sand. These two zones are labeled A and B in figure A-32. Since this 

site is located at the southwest end of the Piedmont, additional heat and humidity of the subtropical 

environment coupled with the rise and fall of the near surface water table have led to increased 

weathering and desiccation of the near surface soils. The increased weathering results in a higher 

clay content, while the desiccation resulted in higher N-values. Dry unit weight values were 

typically between 14 and 16 kN/m3

, but values were as low as 13 kN/m3 and as high as 16.8 kN/m3. 

This variation in unit weight may have been induced by the weathering profile, may be indicative of 

a localized softer zone, may have resulted from samples taken on a relic fissure, or may have 

resulted from disturbance during sampling and testing. The assessment of potential thin zones of 

soft soils by use of conventional drilling and sampling methods is difficult in the Piedmont, since 

layers may be missed during the investigation. Sample disturbance is likely, which may affect the 

outcome and interpretations of laboratory test data. 

Archived Document
A-45 

Stratigraphy from CPTu Data

Figure A-33 presents the tip resistance (qt), sleeve friction (fs), friction ratio (FR = fs/qtx100), and 

penetration pore pressure measured behind the tip (u2) with depth for two CPTu soundings at the 

Alabama site. The interpreted layering profile and differences between the measured signatures for 

the two different soundings (signature 1 and signature 2) are discussed with respect to an interpreted 

weathering profile. Figure A-34 shows CPTu data for one sounding (signature 1) from the Alabama 

site presented on the Robertson et al. (1986) soil behavior type charts. 

Figure A-33. Subsurface layering based on CPTu data. 

Two primary layers (A and B) are selected in the upper 15 meters of the profile. Based on 

differences in cone signature, Layer B is divided into four subsections (B1, B2, B3, B4). Below 21 

to 23 meters, the Layer C is distinguished based on the piezocone signature. The pore pressure (u2) 

and tip resistance (qt) response tend to control the selected stratigraphy, while the friction ratio (FR) 

displays a range of soil response from sand to clay. Table A-11 summarizes the results of the CPTu 

profile interpretation for signature 1. 

Layer A 

Layer A is designated as a desiccated zone of partially-saturated soils. The u2 pore pressure 

response is around zero, but may be positive or negative. These varying conditions may be an 

indication of current degree of saturation, depending upon the humidity, infiltration, and prior 

rainfall around the time of testing. Since this layer is a partially saturated zone, the pore pressure 

response will not be used for classification purposes. Tip resistance and sleeve friction values have 

parallel responses, increasing and decreasing at similar depths within the layer. Signature 2 initially 

has a lower tip resistance (about 1.2 MPa) and higher friction ratio (about 7) than signature 1, but the 

two signatures converge at 2.5 meters with a tip resistance around 2.7 MPa and a friction ratio 

A

B1

B2

B3

B4

C

0

5

10

15

20

25

30

35

0 5 10 15 20

qt ( MPa)

Dept h BGS ( m)

Si gnat ur e 1

Si gnat ur e 2

0

5

10

15

20

25

30

35

0 200 400 600

f s ( kPa)

0

5

10

15

20

25

30

35

- 100 100 300 500

u2 ( kPa)

uo

0

5

10

15

20

25

30

35

02468

FR ( %)

Archived Document
A-46 

around 5. The difference in these two signatures is likely indicative of differential near-surface 

weathering. While this layer covers 4 zones of the classification chart presented in figure A-34, the 

average response indicates a silty clay to clay with the variation in the response indicating the 

presence of some sand. 

Table A-11. Summary of CPTu subsurface profile interpretation. 

Layer Depth 

Range 

qt – FR zones qt – Bq

zones 

Soil Behavior Type 

Classification 

Residual Soil 

Weathering 

 (m) (Robertson et al., 1986) Horizon 

A 0 to 4.5 3,4,5,6 - Clayey silt to silty 

clay with sand 

Horizon IA/IB 

B1 4.5 to 

7.75 

3,4,5,6 6,7 Clayey silt to silty 

clay with sand 

Saprolite 

B2(1) 7.75 to 

11.5 

3,4,5 5,6,7 Clayey silt to silty 

clay with sand 

Variable Saprolite, 

some softer material 

B3(1) 11.5 to 

15.6 

3,4 4,5,6 Silty clay to clay with 

sand 

Softer Saprolite 

B4 15.6 to 

23 

3,4,5,6,7,8,11 7,8 Silty sand with sand Denser Saprolite 

C 23 + 3,4,5,6,7,8,9, 

11,12 

8,9,10 Alternating seams of 

silty sand with sand 

and gravel (rock 

fragments) and 

clayey silt to sandy 

silt with clay 

Partially Weathered 

Rock 

(1) While zone B2 and B3 have similar soil behavior types and residual soils horizons, they are primarily separated to 

discuss differences between cone signature 1 and cone signature 2.

Layer B 

Variation in the upper weathered soil portion of the profile is inferred from the two cone signatures 

between 4.5 and approximately 22 meters depth. 

Layer B1 

This sublayer generally has a similar response for both cone signatures with a tip resistance around 

2.6 MPa, a friction ratio around 5, and negative u2 penetration pore water pressures up to about 

negative 100 kPa. Penetration pore pressures greater than negative 100 kPa (negative 1 atmosphere) 

have not been reported for any soil type, and it its believed that a limiting negative pore pressure is a 

function of cavitation of the porewater at about negative one atmosphere. 

Archived Document
A-47 

Layer A

Layer B1

Layer B2

Layer B3

Layer C

Layer D

Robertson et al. 

(1986) Soil 

Behavior Type

1 - Sensitive

 fine grained

2 - Organic 

 material

3 - Clay

4 - Silty clay to 

 clay

5 - Clayey silt to

 silty clay

6 - Sandy silt to

 clayey silt

7 - Silty sand to 

 sandy silt

8 - Sand to silty

 sand

9 - Sand

10 - Gravelly 

 sand to sand

11 - Very stiff

 fine grained*

12 - Sand to

 clayey sand*

*Overconsolidated or 

Cemented

Figure A-34. Soil classification charts with CPTu data from Alabama site (signature 1). 

Layer B4

Layer C

Archived Document
A-48 

Layer B2 

Layer B2 shows quite a different response for all 3 piezocone parameters when signature 1 is 

compared to signature 2. Signature 1 has a qt of about 2 MPa, fs of about 100 kPa, FR of about 5, 

and positive u2 penetration porewater pressures up to about 300 kPa. Signature 2 has a qt of about 4 

MPa, fs of about 200 kPa, FR of about 5, and u2 penetration porewater pressures of about negative 80 

kPa. The contrasts in this layer are interesting in that the tip resistance and sleeve friction values are 

proportionally higher in signature 2 when compared to signature 1, and the u2 penetration porewater 

pressures are negative in the stiffer soils (signature 2) and mostly positive in the softer soils 

(signature 1). The transition from negative to positive u2 penetration porewater pressures is likely 

indicative of transition from a dense dilative sandy silt material to a looser more clayey silt. 

 

Layer B3 

Layer B3 shows the convergence of signatures 1 and 2 to qt values of about 2 MPa, fs of about 100 

kPa, FR of about 5, and positive u2 penetration porewater pressures up to about 500 kPa. At both 

test locations a softer clayey silt has been encountered. 

Layer B4 

At a depth of about 15.6 meters the tip resistance and sleeve friction values increase to about 4.7 

MPa and 200 kPa, respectively, yielding a friction ratio between 4 and 4.5. The u2 penetration 

porewater pressures become negative again in this layer. Due to the higher tip resistance and 

negative u2 penetration porewater pressures, it is inferred that this layer consists of a dense silty 

sand. 

Layer C 

The transition into Layer C is not abrupt, and likely occurs between 21 and 23 meters depth in this 

profile. The cone test that produced signature 2 met refusal at about 21 meters, while the cone test 

that produced signature 1 was continued until about 32 meters. According to table A-11, Layer C 

classifies as every soil behavior type except sensitive fine grained, organic material, or gravelly 

sand. While this may seem prohibitive to classification of the layer, the soil layer becomes obvious 

when you consider the residual soil weathering profile. Alternating layers of very stiff material (qt ≈

20+ MPa) and softer zones (qt ≈ 5 to 15 MPa) are indicative of the partially weathered rock horizon. 

The u2 penetration pore pressures are still negative, showing no change from the saprolite horizon. 

The friction ratio values are still variable, but on average are lower in the very stiff soils and higher 

in the soft zones. 

The depth to competent rock is quite variable in the Piedmont. Cone penetration testing through 

weathered rock is difficult and may require termination of a sounding before competent rock is 

confirmed, due to inclination, and overstressing of load cells. For design problems where depth to 

rock is important, it should be ensured that rock is reached rather than an intermediate layer within 

the partially weathered rock zone. Rock coring of at least 1.5 m is recommended to accompany CPT 

and CPTu investigations in residual soils where depth to competent rock is needed. 

Archived Document
A-49 

Stratigraphy from DMT Data

Figure A-35 shows the results of four DMTs performed at the Alabama site with the open symbols 

representing the po lift-off pressure and closed symbols representing the p1 expansion pressure. In 

figure A-36, the index parameters ID and ED are used to estimate classification and consistency. 

The material index, ID, indicates that the site is primarily clayey silts and silts. One of the soundings 

also indicates the possibility of thin clayey seams at 5.5 and 10.5 meters. While the material index is 

uniform throughout the site, the increase in the horizontal stress index parameter, KD, is used to 

separate a stiff overconsolidated crust at depths above about 3.5 meters. This crust is labeled Layer 

A, while the underlying soils are labeled Layer B. It is inferred from figure A-36 that the clayey 

silts, silts, and sandy silts are of medium dense to dense consistency. 

Summary

Similar soil profiles were developed in the upper 16 meters using the SPT, CPTu, or DMT. A stiff 

desiccated crust exists up to around 2.75 to 4.5 meters, which is underlain by silty sands to sandy silt 

with some clay to 16 meters. Using data from the piezocone penetrometer, the saprolite profile was 

interpreted to continue to about 22 meters where it transitioned to partially weathered rock to the end 

of the sounding at 32 meters. The distinction between partially weathered rock and competent 

bedrock cannot be confirmed using the cone penetrometer. The CPTu provided the most detail for 

evaluating site stratigraphy. The u2 penetration porewater pressure measurements provided 

additional insight into the weathering profile where conventional laboratory index tests did not show 

distinct differences in soil type. The potential for significant variability in the weathering profile can 

be addressed with piezocone testing, since the CPTu can provide rapid and inexpensive profiling of a 

site when compared to traditional drilling. While the piezocone provided good resolution of layering 

and changes in the weathering profile, companion drilling, sampling, and visual classification are 

recommended so the engineer can get a better idea of important defects and features. These may 

include mica content, the presence of original rock structure, and slickensided planes of weakness in 

the soil matrix. 

Archived Document
A-50 

0

2

4

6

8

10

12

0 500 1000 1500 2000

Pressure (kPa)

Depth (m) 9901-o 9901-1 9902-o 9902-1

SDMT01-o SDMT01-1 SDMT02-o SDMT02-1

 po p1

0

2

4

6

8

10

12

0.1 1 10

ID

C lay Silt Sand

0

2

4

6

8

10

12

0 25 50 75

ED (MPa)

0

2

4

6

8

10

12

0 5 10 15 20

KD

A.

B.

Figure A-35. Soil stratigraphy from DMT index values. 

Low Density

Low Rigidity

Loose

Soft

Med. Rigidity

Medium Dense

Compressib

M

le

edium

Rigid

Dense

Very Rigid

Very Dense

Hard

10

100

1000

0.1 1 10

Material Index, ID

E D (atm)

9901-A

9901-B

9902-A

9902-B

SDMT01-A

SDMT01-B

SDMT02-A

SDMT02-B

 Clay Silt Sand

Figure A-36. Classification and consistency of soil at Alabama site based on DMT data.

Archived Document
A-51 

RELATIONSHIP BETWEEN SPT N VALUES AND IN-SITU TEST PARAMETERS 

Substantial design and experience in residual soils of the Piedmont province have been based on 

SPT blow counts. This example presents strength and compression properties derived from the SPT 

as well as other in-situ tests. This section provides correlation between the SPT N-value and other 

in-situ test reference parameters, such as CPT tip resistance, qt. Due to the large amount of scatter in 

correlations between N-value and other in-situ test resistance parameters, it is not recommended to 

use these correlations directly in design. Rather, these correlations are presented as a reference so 

that the engineer with many years of design experience with the SPT can get a feel for other in-situ 

test parameters. Assessment of engineering properties using individual in-situ and laboratory tests 

will be presented in the following sections. 

Figure A-37 provides a correlation which relates the ratio of cone tip resistance to N-value as a 

function of mean particle size, D50 (Kulhawy & Mayne, 1990). The soils at the Alabama site with a 

higher fines content plot with qc/N values between 1 and 2, while the more granular material (D50 >

0.075 mm) plot with qc/N ratios of between 2 and 4. These data are in agreement with previously 

proposed correlations, and thus first order data quality seems acceptable. 

Figure A-38 compares dilatometer modulus, ED, corresponding to a factor of safety of 3 (i.e., backcalculations corresponding to working load levels) to N-value using a correlation presented in 

Mayne & Frost (1988). Data from the Alabama site as well as data from additional sites in the 

Piedmont province are included for comparison to the proposed trend. The Alabama data follow the 

previously proposed trend and matches well with the additional data provided. The first order 

quality of the DMT data appears to be reasonable. 

Figure A-39 compares the initial elastic modulus from pressuremeter tests to N-value using a 

correlation for Piedmont soils presented in Martin (1977). Menard prebored pressuremeter initial 

elastic modulus data from the Alabama site is evaluated along with cone pressuremeter (CPM) 

unload-reload (U-R) modulus data from the Alabama site. The initial modulus values from the 

Menard PMTs fit within the range of uncertainty of the Martin (1977) correlation, but the U-R 

modulus from the CPM is one to two orders of magnitude higher than expected. This is likely due to 

disturbance inherent in the CPM test as well as possible additional errors in test procedure that will 

be discussed in a latter section on deformation data from the PMT. 

Archived Document
A-52 

Figure A-37. Comparison of Alabama site data to published trend between CPT tip resistance and 

N-value as a function of mean grain size (Kulhawy & Mayne, 1990). 

Figure A-38. Comparison of Opelika, Alabama data to published trend between dilatometer 

modulus and N-value in Piedmont sandy silts (Mayne & Frost, 1988). 

Opelika, AL

0

200

400

600

800

0.00 10.00 20.00 30.00 40.00 50.00 60.00

GT

NCSU

Olympic Stadium

Opelika, AL

Archived Document
A-53 

1

10

100

1000

10000

1 10 100

SPT N-value

EPMT (atm)

Menard PMT (Opelika)

Full Displacement PMT U-R

(Opelika)

Martin, 1977

Figure A-39. Comparison of Opelika, Alabama data to published trends between PMT 

modulus and N-Value in Piedmont soils (Martin, 1977). 

DEFORMATION CHARACTERISTICS OF RESIDUAL SOILS 

General

While compression properties from laboratory oedometer tests are commonly used to assess 

settlements for clayey soils, the elastic modulus is typically used to calculate settlements in granular 

soils. Residual soils are intermediate and thus oedometer test results and methods relying on elastic 

modulus values are used in practice to evaluate settlements in residual soils. Oedometer test results 

for the soils at the Alabama site are presented in this section and information on potential errors and 

limitations in estimating settlements of residual soils based on oedometer tests are presented. 

Information on evaluating an elastic modulus in residual soils using various in-situ tests is presented 

in this section. Elastic moduli calculated from pressuremeter test (PMT) data and flat plate 

dilatometer test (DMT) data, as well as interpreted from shear wave velocity data are presented. 

Evaluation of Compression Properties from Laboratory Oedometer Tests

Laboratory consolidation test results for residual soils are affected by the high sand and silt content, 

as well as disturbance that typically occurs during sampling, extraction, and trimming of the 

specimens. Most conventional consolidation curves for residual soil are generally rounded due to 

disturbance. Predicted settlements using conventional consolidation curves for Piedmont residual 

soils are usually 1.25 to 1.5 times greater than actual measured values. This overprediction of 

settlement may result in the use of deep foundation systems where shallow foundations could have 

Archived Document
A-54 

been used. One-dimensional consolidation curves for specimens from the Alabama site are shown in 

figure A-40(a). The strain energy method is used to evaluate the preconsolidation stress since it can 

lead to a less ambiguous estimation of preconsolidation stress from oedometer tests on residual soils, 

as shown in figure A-40(b). Data from laboratory oedometer tests on Piedmont residual soils at the 

Alabama site are shown in table A-12. 

Sample disturbance was evaluated using the procedures outlined in chapter 4, and results are 

presented in table A-12. The strain levels to reach in-situ vertical effective stress indicate that all 

specimens were likely to be disturbed. The curves never reached a truly linear virgin compression 

line, increasing the difficulty in selecting a preconsolidation stress from the test data. Unload-reload 

cycles performed immediately before the anticipated preconsolidation stress will result in a more 

accurate evaluation of the recompression index. 

0

5

10

15

20

25

30

35

1 10 100 1000 1000

Effective Consolidation Stress, σv ' (kPa)

Vertical Strain, εv (%)

B9 - 3 m eters; eo = 0.95

B9 - 4 m eters; eo = 1.04

B7 - 5 m eters; eo = 1.22

B8 - 7 m eters; eo = 0.95

B7 - 9 m eters; eo = 1.72

 

0

100

200

300

400

0 500 1000 1500 2000

Stress (kPa)

Strain Energy (kPa)

3 meters

4 meters

5 meters

7 meters

9 meters

0

10

20

30

40

50

60

0 100 200 300 400 500 600

Stress (kPa)

Strain E nergy (kPa)

Figure A-40. Laboratory 1-D consolidation curves 

(a) Stress – Strain; (b) Stress – Strain Energy

Table A-12. Summary of oedometer testing on Piedmont residual soils. 

Boring Depth LL PI wn eo σvo' Casagrande Strain Energy Strain Sample 

 

(m) 

 

(%) 

 

(kPa) 

σp′

(kPa) 

OCR σp′

(kPa) 

OCR to σvo′

(%) 

Quality1

B9 3 37 10 30.2 0.95 62.2 200 3.2 375 6.0 2.0 DT 

B9 4 39 12 27.2 1.04 79.8 190 2.4 400 5.0 5.4 DT 

B7 5 28 4 32.8 1.22 88.5 220 2.5 375 4.2 4.4 DT 

B8 7 - NP 25.4 0.95 104.4 215 2.1 250 2.4 4.0 DT 

B7 9 - NP 45.3 1.72 122.2 100 < 1 155 1.3 8.3 DT 

1

 Evaluated using estimated OCR and strain at in-situ vertical effective stress; DT – Disturbed test 

Archived Document
A-55 

In the next sections, calculations of elastic modulus are presented for various in-situ testing methods. 

In order to compare those values to values obtained from consolidation data, typical recompression 

parameters can be used to calculate an elastic modulus. For example, using the consolidation data 

for Boring B9 at 4 meters results in a value of Crε of about 0.05. The elastic modulus can be 

calculated as: 





 





−

−+ = )1(

3.2 )21()1( '

ν

σ νν

rε

v

C

E (Equation A-15) 

Using a vertical effective stress of 100 kPa and a Poisson’s ratio of 0.1, results in a calculated elastic 

modulus, E, of 4,500 kPa. 

Elastic Modulus from Pressuremeter Data

The pressuremeter elastic modulus estimated from the applied stress – volume curve can be used for 

elastic settlement analysis as outlined in FHWA-IP-89-008 (1989). Settlement analyses for 

Piedmont residual soils based on elastic modulus values from pressuremeter testing have been 

shown to be more accurate than those based on conventional settlement analyses using compression 

properties from laboratory oedometer testing. Menard pressuremeter data from the Alabama site are 

shown in figure A-41 and full displacement pressuremeter data from the same site is shown in figure 

A-42. Data summaries are provided in tables A-13 and A-14. 

The calculation of the pressuremeter elastic modulus has been presented in section 4.7. Using the 

applied stress – probe volume curve at 8 m in boring B-5, the elastic modulus is calculated as: 

( )( ) ( )( ) ( )

( ) kPa

cm

kPa

cm cm

V

P EP VV m 995,4

110295

150500 12 79033.012 5.202 3

3 3

0 = −

− += +

∆

∆

ν ++= (Equation A-16) 

where ν is the Poison ratio taken as 0.33, Vo is the initial volume of the probe, Vm is the mean 

volume of the probe over the stress range of consideration (i.e., (110 cm3 + 295 cm3)/2 = 202.5 cm3), 

and ∆P / ∆V is the slope of the linear portion of the stress – volume curve (between po and pf or pu

and pr). 

Table A-13. Test data and calculated elastic modulus from 

Menard PMT data at the Alabama site.

Boring Depth 

(m) 

po

(kPa) 

pf

(kPa) 

pL 

(kPa) 

∆p 

(kPa) 

∆V 

(cm3

) 

Vo 

(cm3) 

Vm 

(cm3) 

EP

(kPa) 

B-5 1 64 436 NA 372 91 790 234.5 11,140 

B-5 2 95 341 NA 246 71 790 267.5 9,746 

B-5 4 151 410 NA 259 142 790 414 5,841 

B-5 6 169 600 NA 431 190 790 314 6,662 

B-5 8 150 500 NA 350 185 790 202.5 4,995 

B-5 10 255 537 NA 282 100 790 160 7,126 

B-5 12 246 573 NA 327 277 790 179 3,043 

B-5 15 475 874 NA 399 142 790 396.5 8,868 

Archived Document
A-56 

Figure A-41. Menard pressuremeter data for the Alabama site. 

Table A-14. Test data and calculated unload-reload elastic modulus from 

full displacement PMT data (C-41) at the Alabama site. 

Depth 

(m) 

pu

(kPa) 

pr

(kPa) 

pL 

(kPa) 

∆p 

(kPa) 

∆V 

(cm3

) 

Vo 

(cm3) 

Vc

(cm3

) 

Vm 

(cm3) 

Eu-r

(kPa) 

Eu-r/EP

1 97.8 617.4 940 519.6 2.7 192 34.67 14.15 123,276 11 

2 117 584 800 467 1.8 192 34.67 15 166,782 17 

3 137 610 900 473 2.3 192 34.67 15.25 132,339 NA 

4 136 526 740 390 1.4 192 34.67 14.9 179,003 31 

6 127 486 782 359 1.6 192 34.67 16.1 144,894 22 

8 72 645 1000 573 0.5 192 34.67 15.05 736,850 148 

10 113 900 1600 787 2.1 192 34.67 12.65 238,570 33 

Archived Document
A-57 

Figure A-42. Full displacement pressuremeter data for the Alabama site. 

The data presented in tables A-13 and A-14 indicate that the initial modulus data from the PMT (i.e., 

Ep) is over an order of magnitude lower than the unload-reload modulus (i.e., Eu-r) from the full 

displacement pressuremeter. This result is likely caused by the following: (1) disturbance during 

preparation of the borehole for the Menard pressuremeter tests; and (2) the lack of a creep stage 

during the full displacement pressuremeter tests. 

The correlation presented in figure A-39 can be used to evaluate the elastic modulus values obtained 

from the two different methods of pressuremeter testing. This figure indicates that the Ep values 

obtained are in good agreement with the Martin (1977) correlation and the Eu-r values are 

excessively high. It is noted that the Martin (1977) correlation used the initial elastic modulus from 

prebored pressuremeter data, and thus disturbance during borehole preparation likely led to a lower 

Archived Document
A-58 

slope of the expansion pressure curve, and thus a lower value of the initial pressuremeter elastic 

modulus. This disturbance and low elastic modulus from Menard PMT data was also likely at the 

Alabama site. To account for this disturbance, at least one unload-reload curve should be performed. 

Below the yield stress (pf) the soil will generally act elastically, and after pF the unload-reload elastic 

modulus will gradually decrease with increasing strain. It is good practice to perform the unloadreload curve near the anticipated yield stress (pF) to estimate the maximum elastic modulus. 

Very high unload-reload elastic modulus values were calculated from full displacement 

pressuremeter data. Typically the unload-reload modulus in Piedmont soils will be 2 to 4 times 

greater than the initial pressuremeter modulus, but the ratio of Eu-r to Ep presented in table A-14 

varied from 11 to 148. High modulus values from full displacement pressuremeter data are likely 

due to disturbance from insertion of the probe as well as the lack of a creep test prior to performing 

the unload-reload loop. 

From the test data presented, the pressuremeter elastic moduli produced by the Menard test would be 

recommended for design. Values from the full displacement pressuremeter tests appear 

unconservatively high. The pressuremeter modulus – SPT N-value correlation for Piedmont data 

(figure A-39) was used to identify potentially erroneous data. It is recommended that creep tests and 

unload–reload curves be performed for future investigations. 

Elastic Modulus from DMT Data

As described in section 5.5, the selection of an elastic modulus for use in deformation analyses 

depends on the soil stress level. That is for designs with very high factors of safety, it is likely that 

the initial small-strain modulus could be used whereas for more typical factors of safety some 

reduced value of the initial modulus is appropriate for settlement analyses. For most designs, the 

elastic modulus corresponding to 25 percent of failure stress, E25, may be used. In Piedmont residual 

soils, the use of the dilatometer modulus, ED, as equal to E25 has been shown to provide reasonably 

accurate predictions of settlement (Mayne & Frost, 1988). Dilatometer modulus values from this 

investigation resemble previously proposed correlations to SPT N-value as shown in figure A-38. 

Figure A-43 compares the dilatometer elastic modulus, ED, to the pressuremeter elastic modulus, Ep.

Figure A-43 indicates that the dilatometer modulus is 2 to 5 times greater than the initial elastic 

modulus from the PMT with an overall average value of Ed = 21,200 kPa (210 atm). Since unloadreload tests were not performed during the Menard pressuremeter tests, it is anticipated that the 

reported pressuremeter modulus values are low and that modulus values based on unload-reload tests 

would be more representative of the elastic modulus of the soil. It is anticipated that a properly 

performed unload-reload cycle after a creep test would yield a pressuremeter modulus that is 2 to 4 

times greater than that reported. Therefore, for elastic settlement calculations using the available 

data from the Alabama site, the DMT modulus, ED, could be used directly in calculations provided 

that working stress levels are consistent with a mobilized soil strength of about 25 percent. 

Archived Document
A-59 

0

2

4

6

8

10

12

14

16

0 100 200 300 400 500 600

 Modulus, E D or EP (atm)

Depth (m)

Average Dilatometer Modulus, ED

Pressuremeter Modulus, EP

Figure A-43. Dilatometer modulus ED compared to pressuremeter modulus, EP 

Elastic Modulus from Shear Wave Velocity Data

As discussed in section 5.5, the initial elastic modulus can also be calculated from shear wave 

velocity data and reduced to an appropriate value consistent with anticipated working stress levels 

for the design. Shear wave velocity, interpreted mass density, and elastic modulus profiles for the 

Alabama site are shown in figure A-44. The shear wave velocity was recorded using a seismic 

piezocone and the mass density was estimated using a correlation to shear wave velocity, Vs, (in 

m/s) and depth, z, (in m) as presented in Mayne et al. (1999): 

( ) s

t /095.1log7.58614.0 Vz

1 1

+ +

ρ +≈ (Equation A-17) 

Laboratory values of mass density are also shown on figure A-44, and agree well with the 

correlation in equation A-17. 

The initial elastic modulus is calculated using equations 51 and 52. As discussed for the DMT 

elastic modulus, the use of E as equal to E25 for design purposes has been shown to be reasonable in 

Piedmont soils. The associated strain level corresponds to a factor of safety value, where 

FS=1/(q/qu), equal to 4 and a modulus ratio of: 

Archived Document
A-60 

1 ( ) 34.025.01 3.0

3.0

=−= 













 −=

o qu

q

E

E (Equation A-18) 

0

2

4

6

8

10

12

0 100 200 300 400

Vs (m/sec)

Depth (m)

0

2

4

6

8

10

12

1.5 2 2.5

ρtot

Vs

correlation

Lab Data

0

2

4

6

8

10

12

0 1000 2000 3000

E (atm)

Emax

E25

Avg

0

2

4

6

8

10

12

0 500 1000 1500

Gmax (atm)

Figure A-44. Calculation of elastic modulus from shear wave velocity data. 

For the profile shown in figure A-44, the average shear wave velocity to a depth of about 12 m is 

204 m/s. Using equation A-17, an average shear wave velocity of 204 m/s, and an average depth of 

6 m, the estimated total mass density, ρtot, is 1.87. From equation 52, the average maximum shear 

modulus would equal, Go = ρtotxVs

2

 = 1.87 (204)2 = 77,822 kPa = 77.8 MPa. To convert between the 

small strain shear modulus and small strain elastic modulus, equation 51 is used with a Poisson ratio 

(ν) equal to 0.1, Eo=2xGox(1+ν)=2x77,822 kPa x (1+0.1) = 171,208 kPa = 171 MPa. As shown in 

equation A-18, the E/Eo reduction value associated with a FS equal to 4 (E25) is 0.34, yielding E25 = 

0.34xEo = 0.34x171,208 kPa = 58,210 kPa = 58 MPa. 

The appropriate E value can also be calculated directly from shear wave velocity using equation 54: 

Es ( ) 210,581.0120487.1234.0 kPa 58MPa 573atm 2 =+⋅⋅⋅⋅= == (Equation A-19) 

This value is about 2.7 times greater than the E25 value calculated from the DMT. Since Piedmont 

soils are considered to be structured due to the relict bonding and features of the parent rock, it can 

be assumed that the small strain shear wave velocity measurements would capture the effects of 

these structural features whereas the DMT and PMT would cause some disturbance. It is noted that 

if a FS value of 2.5 (i.e., mobilized strength corresponding to 40 percent of the ultimate strength) 

was used for design, the associated E/Eu reduction factor would be 0.24, and the E40 value would be 

about 41,089 kPa. 

Archived Document
A-61 

Summary

Calculations of compression properties for Piedmont residual soils have been presented based on 

laboratory oedometer tests, Menard pressuremeter, full-displacement pressuremeter, and the DMT. 

Shear wave velocity data has also been used. Table A-15 shows a summary table of the average 

elastic modulus values measured (or evaluated) based on these tests. 

Table A-15. Summary of elastic modulus values from various tests. 

Test Type Elastic Modulus (kPa) 

Laboratory oedometer 4,500 

Menard pressuremeter 7,585 

Full-displacement pressuremeter >100,000 

Dilatometer 21,200 

Shear Wave Velocity correlation 58,210 (E25) 

41,089 (E40) 

SPT N correlation (see table 29) 

Es=700 (N1)60

10,500 

As previously discussed, modulus values from the full-displacement pressuremeter are clearly too 

large for the reasons discussed. It is important to note that modulus values from laboratory 

oedometer tests and from a correlation using SPT N values underestimates the modulus (relative to 

other tests) by a factor of 2 to greater than 10. Moreover, where settlements are critical, it can be 

seen that more sophisticated methods such as the DMT, PMT, or shear-wave velocity evaluations 

may be warranted. Perhaps more importantly, this assessment of compression properties 

demonstrates the importance of using high-quality undisturbed samples for laboratory testing. 

Where it is clear that obtaining undisturbed samples will be difficult and impractical, consideration 

should be given to performing appropriate in-situ testing. 

STRENGTH CHARACTERISTICS OF PIEDMONT RESIDUAL SOILS 

General

Selection of design strength properties in residual soils is particularly difficult due to the 

intermediate drainage conditions, as well as potential variability in the weathering profile and 

presence of slickensided planes of weakness. In-situ penetration tests in silty and clayey materials 

are typically undrained. In some cases in Piedmont residual soils, dissipation of excess penetration 

pore pressures occur within 1 to 3 minutes of paused penetration indicating partially drained 

conditions. Therefore, correlations between in-situ test parameters and either undrained strength or 

drained strength parameters are not appropriate. Additionally, the presence of mica may 

significantly reduce SPT blowcounts and could affect other in-situ test parameters to varying 

degrees, but will not reduce the effective stress friction angle to the same degree. In summary 

though, due to the relatively high hydraulic conductivity values of residual soils, most loading 

Archived Document
A-62 

conditions associated with typical transportation construction activities will occur over such a time 

period that the prevailing drainage condition in the residual soil is drained. Analysis of data for this 

example is therefore referenced to the drained condition. 

In this example, laboratory strength characteristics of Piedmont soils are evaluated using 

isotropically consolidated undrained triaxial compression tests with pore pressure measurements 

(CIUC). Correlations between strength parameters and in-situ test parameters are compared to the 

Alabama site laboratory data and recommendations for design strength values are presented. 

Summary of CIUC Test Data

Fifteen CIUC triaxial tests were performed on nominally undisturbed Shelby tube samples from the 

Alabama site. A typical stress-strain curve and pore pressure-strain curve for two specimens are 

shown in figure A-45 and detailed testing data for specimen B2-1-1 during the shearing portion of 

the test are provided in table A-16. Triaxial test summary information is presented in table A-17. 

0

25

50

75

100

125

150

175

200

0 1 2 3 4 5 6 7 8 9 10

Axial Strain (%)

Applied Stress, (kPa)

0

25

50

75

100

125

150

175

200

Pore Pressure, u (kPa)

B2-1-1 @ 15m (Stress)

B7-1 @ 4m (Stress)

B2-1-1 @15m (Pore Pressure)

B7-1 @ 4m (Pore Pressure)

Point of Maxim um 

Obliquity B2 Point of Maximum

Obliquity B7

Figure A-45. Stress-strain curve for CIUC tests on samples B2-1-1 at 15 m and B7-1 at 4 m. 

Failure was defined at the point of maximum obliquity. The point of maximum obliquity occurs 

when the ratio of σ1′ to σ3′ is a maximum. For test B2-1-1, the point of maximum obliquity occurs 

at an axial strain of approximately 8.2%, and the obliquity value corresponding to that strain level is: 

( )

( )

( )

( ) 744.3

46.11402.171

46.1142.326

'

'

3

1

3

1 = −

− = ∆−

∆− = uu

σ

σ

σ

σ (Equation A-20) 

Archived Document
A-63 

where σ3 is the minor principal stress (which is equal to the consolidation stress in a CIUC test), σ1

is the major principal stress (which is equal to the applied vertical stress in a CIUC test), the prime 

(′) sign indicates effective stress which requires the subtraction of excess pore pressure, ∆u, from 

each principal stress. While the point of maximum obliquity may occur near the end of the test, as 

shown for specimen B2-1-1 in figure A-45, it may also occur earlier in the test, as shown for 

specimen B7-1 in figure A-45. This point is easy to find when using a spreadsheet application for 

data analysis, as shown in table A-16. 

Table A-16. Stress-strain data for specimen B2-1-1 at 15 m. 

Axial 

Deflect. 

(cm) 

Axial 

Strain, 

εv (%) 

Axial 

Load 

(N) 

Corrected 

Area 

(cm2

) 

∆σ 

(kPa) 

σ3 

(kPa)

σ1 

(kPa)

∆u 

(kPa) 

p 

(kPa) 

p' 

(kPa) 

q 

(kPa) 

(σ1′/σ3′) 

0.000 0.00 0.0 39.2 0.0 171.0 171.0 0.0 171.0 171.0 0.0 1.000 

0.036 0.26 289.1 39.3 73.5 171.0 244.5 44.1 207.8 163.6 36.7 1.579 

0.071 0.51 359.9 39.4 91.2 171.0 262.3 62.7 216.6 153.9 45.6 1.843 

0.107 0.77 409.7 39.5 103.6 171.0 274.6 75.8 222.8 147.0 51.8 2.089 

0.142 1.02 452.8 39.6 114.2 171.0 285.2 85.5 228.1 142.6 57.1 2.336 

0.178 1.28 472.8 39.7 119.0 171.0 290.0 91.0 230.5 139.5 59.5 2.487 

0.213 1.54 485.7 39.9 121.9 171.0 292.9 95.8 232.0 136.1 60.9 2.621 

0.249 1.79 500.4 40.0 125.2 171.0 296.3 100.0 233.6 133.7 62.6 2.763 

0.284 2.05 512.4 40.1 127.9 171.0 298.9 102.7 235.0 132.2 64.0 2.873 

0.320 2.30 527.1 40.2 131.2 171.0 302.3 105.5 236.6 131.1 65.6 3.003 

0.356 2.56 536.5 40.3 133.2 171.0 304.2 106.9 237.6 130.8 66.6 3.077 

0.391 2.82 544.9 40.4 135.0 171.0 306.0 108.3 238.5 130.2 67.5 3.150 

0.427 3.07 552.5 40.5 136.5 171.0 307.5 110.3 239.3 128.9 68.2 3.248 

0.462 3.33 558.3 40.6 137.5 171.0 308.6 111.0 239.8 128.8 68.8 3.292 

0.498 3.59 567.6 40.7 139.5 171.0 310.5 111.7 240.8 129.1 69.7 3.351 

0.533 3.84 575.6 40.8 141.1 171.0 312.1 112.4 241.5 129.2 70.5 3.406 

0.569 4.10 581.8 40.9 142.2 171.0 313.2 113.1 242.1 129.0 71.1 3.454 

0.605 4.35 586.7 41.0 143.0 171.0 314.0 113.8 242.5 128.8 71.5 3.498 

0.640 4.61 594.7 41.1 144.6 171.0 315.6 113.8 243.3 129.5 72.3 3.525 

0.676 4.87 600.5 41.2 145.6 171.0 316.6 113.8 243.8 130.0 72.8 3.543 

0.711 5.12 606.7 41.4 146.7 171.0 317.7 113.8 244.4 130.6 73.4 3.562 

0.747 5.38 613.0 41.5 147.8 171.0 318.8 114.5 244.9 130.5 73.9 3.613 

0.782 5.63 618.3 41.6 148.7 171.0 319.7 114.5 245.4 130.9 74.3 3.629 

0.818 5.89 622.8 41.7 149.4 171.0 320.4 114.5 245.7 131.2 74.7 3.641 

0.853 6.15 627.6 41.8 150.1 171.0 321.1 114.5 246.1 131.6 75.1 3.654 

0.889 6.40 632.5 41.9 150.9 171.0 321.9 114.5 246.5 132.0 75.4 3.668 

0.925 6.66 636.1 42.0 151.3 171.0 322.3 114.5 246.7 132.2 75.7 3.675 

0.965 6.95 641.9 42.2 152.2 171.0 323.2 114.5 247.1 132.7 76.1 3.691 

1.001 7.21 644.5 42.3 152.4 171.0 323.4 114.5 247.2 132.8 76.2 3.695 

1.036 7.46 649.9 42.4 153.3 171.0 324.3 114.5 247.7 133.2 76.6 3.710 

1.072 7.72 653.9 42.5 153.8 171.02 324.8 114.46 247.9 133.5 76.9 3.719 

1.107 7.98 659.7 42.6 154.7 171.02 325.7 114.46 248.4 133.9 77.4 3.735 

1.143 8.23 663.7 42.8 155.2 171.02 326.2 114.46 248.6 134.2 77.6 3.744 

1.179 8.49 667.7 42.9 155.7 171.02 326.7 113.77 248.9 135.1 77.9 3.720 

Notes: The highlighted point refers to the maximum principal stress ratio, and thus failure at maximum obliquity. 

Failure Envelopes and Design Strength

Table A-17 provides a summary of the results of the CIUC triaxial tests conducted. The table 

contains parameters at failure, including strain (εv,ff), excess pore pressure (∆uff), Skempton’s pore 

pressure parameter (Aff), the stress path parameters (p, p′, and q), and the secant effective stress 

friction angle (φ’). For example, Skempton’s Aff parameter for specimen B5-1 at 4 m shown in table 

A-17 was calculated as: 

Archived Document
A-64 

21.0

02.158

04.33

31

3 = −

− = ∆−∆

∆−∆ = kPakPau kPakPa Aff σσ

σ (Equation A-21) 

The stress path parameters for specimen B5-2-2 at 15 m shown in table A-17 were calculated as: 

( ) ( ) kPa kPa kPa

p 65.665

2

7.4136.917

2

31 = + = + = σσ (Equation A-22) 

' upp =−= kPa − kPa = 1.4676.19865.665 kPa (Equation A-23) 

( ) ( ) kPa kPa kPa

q 95.251

2

7.4136.917

2

31 = − = − = σσ (Equation A-24) 

The effective stress path is shown on figure A-46.

Table A-17. Strength properties from CIUC triaxial tests on Piedmont soils 

from the Alabama site. 

Boring Depth 

(m) 

σ3C’ 

(kPa) 

∆σff

(kPa) 

σ1,ff

(kPa) 

εv,ff

(%) 

∆uff

(kPa) 

Aff p 

(kPa) 

p' 

(kPa) 

q 

(kPa) 

Secant φ' 

(deg) 

B5-1 4 52.3 158.2 210.5 13.8 33.4 0.21 131.4 98.0 79.1 53.8 

B7-1 4 84.8 155.8 240.6 3.6 23.4 0.15 162.7 139.3 77.9 34.0 

B7-2 4 162.0 188.0 350.0 6.8 62.1 0.33 256.0 193.9 94.0 29.0 

B7-1 6 179.3 248.5 427.7 4.4 108.4 0.44 303.5 195.1 124.2 39.6 

B8-2 6 179.3 248.5 427.7 4.7 108.4 0.44 303.5 195.1 124.2 39.6 

B7-2-1 8 206.8 233.9 440.7 4.0 136.9 0.59 323.7 186.8 116.9 38.8 

B7-2-2 8 310.3 351.9 662.2 10.7 201.5 0.57 486.3 284.7 176.0 38.2 

B7-1 10 120.7 183.3 303.9 4.1 72.6 0.40 212.3 139.7 91.6 41.0 

B7-2 10 241.3 313.8 555.1 9.3 159.2 0.51 398.2 239.0 156.9 41.0 

B2-1 12 115.6 116.5 232.1 15.0 46.9 0.40 173.9 127.0 58.3 27.3 

B2-6 12 115.6 105.0 220.6 8.2 54.5 0.52 168.1 113.6 52.5 27.5 

B5-1 12 127.2 174.2 301.4 7.3 46.3 0.27 214.3 168.0 87.1 31.2 

B5-2 12 102.0 116.4 218.4 5.0 47.2 0.41 160.2 113.0 58.2 31.0 

B2-1-1 15 171.0 155.2 326.2 8.2 114.5 0.74 248.6 134.2 77.6 35.3 

B5-2-2 15 413.7 503.9 917.6 2.9 198.6 0.39 665.6 467.1 251.9 32.6 

Notes: The subscript ff refers to measurement made at failure. Failure is defined as the point of maximum obliquity.

The secant effective stress friction angle for specimen B7-2-2 at 8 m is calculated as: 

( ) o

p

q 2.38

7.284

176 arcsin ' arcsin'tanarcsin'  =



 



 =













 φ = ψ = (Equation A-25) 

Figure A-47 shows a range of effective stress failure envelopes for the data. The trend line shown in 

this figure was developed assuming the effective stress cohesion intercept is zero. A regression 

analysis was also performed (although not shown) to evaluate the best-fit cohesion intercept. This 

analysis resulted in φ′=33.8° and c′=4.9 kPa. Based on this, it does not appear that there is a 

significant cohesion intercept in the laboratory specimens tested from this site. Since the cohesion 

intercept is fairly insignificant, the effective stress shear strength properties of the soil at this site will 

be modeled solely using a friction angle of 35.2° based on regression analyses. 

Archived Document
A-65 

0

50

100

150

200

250

300

0 50 100 150 200 250 300 350 400 450 500

p' (kPa)

q (kPa)

B2-1-1 @ 15m Effective Stress

B7-1 @ 4m Effective Stress

Effective Stress Failure Envelope

Figure A-46. Effective stress paths for CIUC tests shown in figure A-45. 

0

50

100

150

200

250

300

0 50 100 150 200 250 300 350 400 450 500

p' (kPa)

q (kPa)

+ 1 Standard Deviation

s = 5.3o

29.9o < φ' < 40.5o

Conservative Lower Estimate

φ' = 27.3o

Linear Regression

φ' = 35.2o

R2 = 0.9334

Figure A-47. Effective stress failure envelopes based on triaxial test data. 

a. 

Archived Document
A-66 

Statistical Analysis of Strength Data

In this section, a more detailed assessment of the laboratory strength data is presented. Basic 

statistical measures and statistical analyses are illustrated using the laboratory strength data. 

Review of the individual data points in figure A-47 indicates that the effective stress failure envelope 

from linear regression of the data points (i.e., φ′=35.2°) may be unconservative at effective stresses 

greater than 300 kPa. In addition, the scatter of the data suggests that it may not be prudent to adopt 

the average friction angle for design analyses without also performing design analyses with a more 

conservative representation of the effective stress friction angle. 

To assess the laboratory strength data further, the standard deviation in the secant friction angle data 

was calculated. For this calculation, the mean effective stress friction angle ( i φ ' ) was taken as 35.2o

from the linear regression. Since the measured data is actually the slope of the failure envelope 

given by tan ψ′, the mean of sin φ′=tan ψ′ ( i x ) was used. This value was calculated as 0.5764. The 

number of samples (n) is 15. The standard deviation of sin φ′ was calculated as: 

( ) ( )

0934.0

14

122.0

115

5764.0

1

1

2

1

2

== −

−

= −

−

=

∑∑= =

n

i

i

n

i

ii x

n

xx

s (Equation A-26) 

To determine the standard deviation of φ′, the arcsine (s) was calculated to be 5.3°. The effective 

stress friction angle of this material is likely between the average (35.2o

) and + one standard 

deviation, resulting in an effective stress friction angle between 29.9 and 40.5 degrees. The average 

and the average + one standard deviation are plotted on figure A-47. The average minus 1 standard 

deviation contains the data at high effective stresses, but does not adequately encompass the scatter 

at effective stresses around 100 kPa to 200 kPa. To account for scatter at lower effective stresses, a 

conservative lower estimate of 27.3 degrees is also shown in figure A-47. This statistical analysis is 

useful since it provides a rational approach to develop conservative design properties. That is, for 

this laboratory data, it would be appropriate to confidently select a friction angle of 29.9° as lower 

bound value for design. 

The statistical analysis performed above was possible due to the relatively large size of an available 

database of friction angle data (i.e., 15 triaxial tests). It is possible, however, to assess the statistical 

variability of the friction angle of a soil (or other properties) without a relatively significant database. 

Published values of the coefficient of variation, V, defined as the standard deviation divided by the 

average value, are available for various soil properties (see Duncan, 2000 and table 52) and these can 

be used to facilitate a simple statistical analysis without requiring significant amounts of data. For 

example, table 52 indicates that the coefficient of variation for soil effective stress friction angle is 

estimated at 2 to 13 percent, and thus the standard deviation estimated from this laboratory data 

would be between: 

oo Vs φ =⋅=⋅= 7.02.3502.0' (Equation A-27) 

Archived Document
A-67 

oo Vs φ =⋅=⋅= 6.421.3513.0' (Equation A-28) 

From this calculation, it is evident that the calculated standard deviation from laboratory tests on 

Piedmont soils from the Alabama site is actually slightly greater than a maximum standard deviation 

calculated using V=13 percent. 

A third method for assessing standard deviation from minimal laboratory or field test results uses the 

three-sigma rule (Duncan, 2000). The three-sigma rule says that 99.73 percent of all values of a 

normally distributed parameter fall within three standard deviations of the mean. Therefore, if HCV 

= highest conceivable value of the parameter and LCV = lowest conceivable value of the parameter, 

these are approximately three standard deviations above and below the average value. The standard 

deviation can then be estimated from: 

6

HCV LCV

s − = (Equation A-29) 

Using secant effective stress friction angle laboratory data from this example and assuming that the 

maximum and minimum values for the friction angle represent the HCV and the LCV, respectively, 

results in: 

4.4

6

3.278.53

6 = − = − =

o o LCVHCV

s ° (Equation A-30) 

This leads to a coefficient of variation of: 

%13125.0

2.35

4.4

' ==== φ

s V (Equation A-31) 

This value for the coefficient of variation is in good agreement with published values and with the 

value estimated from all the laboratory data. Difficulties in assessment of the coefficient of variation 

from the three-sigma rule lie in the estimation of the highest and lowest conceivable value. The 

HCV and LCV are typically underestimated, and would likely have been underestimated in this case 

if the large amount of laboratory data were not available. 

Shear Strength based on Correlation to In-Situ Test Parameters

Correlations for the SPT, CPT, and DMT are compared to a best-fit linear regression effective stress 

friction angle failure envelope in figure A-48. The following correlations between in-situ test 

parameters and effective stress friction angle are recommended for clean sands, but are used herein 

for Piedmont residual silty sands to demonstrate potential errors associated with applying 

correlations incorrectly. 

Standard Penetration Test: φ′ = [15.4x(N1)60]

0.5 + 20o (Equation A-32) 

Cone Penetration Test: φ′ = 17.6o

 + 11.0xlog [(qc/pa)/(σvo′/pa)

0.5] (Equation A-33) 

Flat Dilatometer Test: φ′ = 28o

 + 14.6xlog KD –2.1xlog2KD (Equation A-34) 

Archived Document
A-68 

The applicability of these correlations are difficult to assess since: (1) residual soils are not clean 

sands, and in-situ test parameters are likely effected to some degree by fines content as well as mica 

content; (2) there is significant variability in residual soil deposits, so in-situ test data may not be 

directly comparable to the laboratory test data at equal depths; and (3) the penetration may be 

undrained, partially drained, or drained. Figure A-48 shows friction angle from laboratory data 

compared to friction angle estimated from in-situ test data using the equations presented above. 

Adjacent to the friction angle plot is the profile of u2 penetration porewater pressure with depth. 

From CIUC triaxial test data shown in figure A-48, the friction angle is estimated to be 35.2 degrees 

using best-fit linear regression. Tests were performed at depths of 4, 6, 8, 10, 12, and 15 m. Above 

3 to 4 m, the soil consists of an overconsolidated partially saturated crust with higher fines content as 

compared to the soil at greater depth, thus the correlations will not be discussed for material from the 

ground surface to a depth of 4 m. It is noted, however, that the correlated effective stress friction 

angles approach high values at low depths. 

0

2

4

6

8

10

12

14

16

25 30 35 40 45

Effective Stress Friction Angle, φ ' 

Depth (m)

Laboratory

SPT

DMT

CPT2

0

2

4

6

8

10

12

14

16

-200 0 200 400

CPT u2 (kPa)

Depth (m)

uo

Figure A-48. Estimate of effective stress friction angle at Alabama site 

using in-situ test data.

Effective stress friction angle estimated from SPT (N1)60 values using the Hatanaka & Uchida 

correlation slightly underestimate the laboratory friction angle value by about 1 degree until about 12 

to 16 m where the φ′ values are underpredicted by about 2 to 3 degrees. While the correlation seems 

to work well, the effects of fines content and mica content on this correlation are unknown and it is 

likely that the good agreement is due to compensating errors. The energy efficiency of the automatic 

hammer was assumed to be 90 percent, but may have varied from 50 to 90 percent. The variation in 

Archived Document
A-69 

energy efficiency alone would result in a reduction of up to about 4 degrees of the estimated friction 

angle. This shows that while correlations of effective stress friction angle of clean sands using SPT 

(N1)60 data from Piedmont soils at the Alabama site match well, the correlation is sensitive to the 

energy efficiency of the SPT hammer which adds considerable uncertainty. There is little 

confidence in the use of this clean sand correlation in Piedmont soils without site-specific laboratory 

confirmation. 

Results from the correlation between CPT tip resistance and effective stress friction angle of clean 

sands slightly underpredicts laboratory CIUC data by about 1 degree for a majority of the sounding, 

but may underpredict φ′ by about 6 degrees or more. It is noted that sharp differences in estimated 

effective stress friction angle from CPT data and laboratory values occur where the u2 penetration 

porewater pressure switches from negative to positive. The positive u2 penetration porewater 

pressures are coupled with a reduction in tip resistance that leads to the reduction in estimated 

friction angle. Since the zone of positive excess (greater than hydrostatic) penetration porewater 

pressures is indicative of partially undrained penetration, the tip resistance in those zones is likely 

not a function of effective stress friction angle and the correlation between tip resistance and 

effective stress friction angle is not valid for these zones. Additionally, fines and mica content will 

affect CPT tip resistance. Therefore, the good agreement between the tip resistance-φ′ correlation in 

the upper 10 meters may be a result of compensating errors.

Effective stress friction angle estimated from correlations to DMT KD parameter match well with φ′

from laboratory CIUC tests. Since the DMTs are performed at pauses in penetration, testing under 

essentially drained conditions is possible. The DMT-φ′ correlation is anticipated to be a lower 

bound relationship, and φ′ estimates from DMT data match well from 4 to 8 meters and slightly 

underpredicts the friction angle from 8 to 12 m. Once again, the effect of drainage conditions and 

fines/mica content on the estimated friction angle has not been verified for DMT testing in residual 

soils. 

Recommended Drained Strength for Design

For strength characterization, it is recommended to perform laboratory tests at a range of confining 

stresses that encompass the design loads. Due to the inherent variability of Piedmont soils, upper 

and lower bound values of effective stress friction angle should be evaluated along with a best 

estimation of friction angle. If sufficient data is not available to assess the standard deviation of the 

friction angle, then published coefficient of variation values can be used to estimate the standard 

deviation from the average friction angle. 

Selection of an appropriate failure envelope for design from laboratory data has been discussed in 

previous sections, along with statistical variability in friction angle. While correlations between φ′

and in-situ test parameters show good agreement at some locations, the variability in the soil profile 

and uncertainty in the effects on test readings due to fines and mica content may lead to acceptable 

agreement due to compensating errors. 

Archived Document
A-70 

PROPERTIES OF WEATHERED ROCK AND ROCK 

In this section, information and property evaluation for weathered rock and rock commonly found in 

the Piedmont province is presented. Rock core information and laboratory testing data are from a 

project site located in Atlanta, Georgia. 

Stratigraphy Information

Figure A-49 shows a rock core log from a depth of 17 to 57 ft (5.2 to 17.4 m). The overburden soils 

at this site (not depicted on the log shown on figure A-49) consist primarily of silty sands (SM) 

underlain by partially weathered rock. For this project, partially weathered rock was defined as 

material which exhibits SPT N values greater than 100-blows/300 mm. The partially weathered rock 

is a transition zone from the overburden (residual) soils to the underlying sound rock. A 3-ft (1-m) 

thick layer of partially weathered rock was encountered at a depth of 14 ft below the ground surface. 

Below the partially weathered rock, is gneiss. A review of the log indicates very good core recovery 

and high RQD values (greater than 90 percent) indicating excellent rock. 

Rock Strength and Deformation Properties

Figure A-50 and A-51 show the results of unconfined compression strength tests on rock core from 

this boring. These samples are taken from depths of 18 ft (just below the partially weathered rock) 

and 40 ft. The stress-strain curve for the sample at 18 ft shows a common feature of laboratory 

compression testing on rocks that are fractured, that being an initial response that is “softer” than the 

subsequent response. This is due to the initial compression of the joints. The unconfined 

compressive strength of these samples is given on the data sheets. A deformation modulus was 

calculated (see figures A-50 and A-51) for each sample. The sample at 18 ft has a modulus of 16.1 

GPa and the sample at 40 ft has a modulus of 22.1 GPa. 

Figure A-52 shows the results of a laboratory direct shear test on a rock joint. This joint is relatively 

planar (i.e., smooth) and is filled with soil material. For this testing, two normal stress levels were 

selected, 4.3 ksf (206 kPa) and 8.6 ksf (412 kPa). For this test series, the sample was sheared at one 

confining stress until both peak and residual strength had been reached. For subsequent confining 

stresses, shearing was carried out to a residual shear strength value. Measured friction angles are 

shown on figure A-52. The irregular (as compared to soils) shear stress-deformation curve results 

from the opening and closing of the joint as the sample is sheared. 

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A-71 

Figure A-49. Rock core log (page 1 of 2). 

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A-72 

Figure A-49. Rock core log. (page 2 of 2). 

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A-73 

0

1

2

3

4

5

6

0 0.1 0.2 0.3 0.4 0.5

Axial Stress, (ksi)

Axial Strain, (%)

ED = ( 4.75 - 3 ) ksi

( 0.325 - 0.25 ) %

2.3 x 10 psi

16.1 GPa

= ED

6

=

BORING NO.: NC-122 DEPTH: 18'

DESCRIPTION: Hard to very hard light

gray and gray biotite

muscovite gneiss

ULTIMATE STRENGTH:

UNIT WEIGHT, DRY:

UNIT WEIGHT, SOAKED: 177.7 pcf

7.20 ksi

177.2 pcf

0

2

4

6

8

10

12

0 0.1 0.2 0.3 0.4 0.5

Axial Stress, (ksi)

Axial Strain, (%)

ED = ( 7.5 - 3.5 ) ksi

( 0.275 - 0.15 ) %

3.2 x 10 ksi

22.1 GPa

= ED

6

=

BORING NO.: NC-122 DEPTH: 40'

DESCRIPTION: Hard to very hard dark

gray-green biotite

amphibole (hornblende)

ULTIMATE STRENGTH:

UNIT WEIGHT, DRY:

UNIT WEIGHT, SOAKED: 176.0 pcf

9.92 ksi

175.5 pcf

gneiss

Figure A-50. Unconfined compression test 

results (sample depth 18 ft). 

 Figure A-51. Unconfined compression 

test results (sample depth 40 ft). 

BORING NO.: NC-230 PEAK ANGLE OF SHEAR RESISTANCE, :

UNIT WEIGHT, DRY:

UNIT WEIGHT, SOAKED: 173.3 pcf

65°

170.7 pcf

0

0

Sheer Stress, (ksf)

2

4

6

8

10

12

14

16

0.2 0.4 0.6 0.8 1.0 1.2 1.4 1.6 0 2 4 6 8 10 12 14 16

Normal Stress, (ksf) N Cumulative Horizontal Displacement, (inches)

SAMPLE DEPTH: 74.0' RESIDUAL ANGLE OF SHEAR RESISTANCE, : 55° R

CORE DESCRIPTION: Hard dark gray

muscovite biotite gneiss

JOINT SURFACE DESCTIPTION: Open,

planar, secondary deposits, 34° dip

Figure A-52. Direct shear test results on rock joint (sample depth 74 ft). 

Archived Document
A-74 

SOIL AND ROCK PROPERTY SELECTION EXAMPLE NO. 3 

Heavily overconsolidated clays

INTRODUCTION 

This example problem describes procedures and interpretation methods for evaluating properties for 

heavily overconsolidated clays. Specifically, this problem focuses on: (1) subsurface profile 

development from laboratory and in-situ test results; (2) consolidation parameters; (3) in-situ stress 

state (i.e., Ko); (4) undrained shear strength; and (5) swell potential. The evaluation procedures for 

properties of overconsolidated clays such as subsurface profile, consolidation, in-situ stress state, 

and shear strength are similar to those previously described in Property Selection Example Nos. 1 

and 2. In this Property Selection Example, issues specifically relevant to the evaluation of these 

properties as they relate to heavily overconsolidated materials are highlighted. The reader is referred 

to the previous example problems for more detail on the baseline procedures for property evaluation. 

The project site is located near Houston, Texas. The subsurface soils at the Site consist primarily of 

clays that have been preconsolidated throughout their thickness as a result of dessication. Thin 

seams of fine sand and silt are also present throughout the soil profile. The depth to groundwater is 

approximately 2 m. The clayey soils of this region have a complex structure including joints, 

fissures, and silt and sand seams. 

The subsurface stratigraphy at the Site consists of the upper “Beaumont” formation to a depth of 

approximately 8-12 m and the lower “Montgomery” formation with a reported thickness of 

approximately 150 m. The Beaumont formation comprises primarily stiff plastic clays with thin 

seams of sand and silt, while the Montgomery formation comprises stiff clays of lower plasticity 

with more silt and sand layers. Available geologic information indicates that the surface of the 

Montgomery formation became highly weathered before deposition of the Beaumont, and 

consequently much of the clay was leached from the surface. As a result, a significant silt and sand 

region exists at the Beaumont/Montgomery interface and the transition is typically readily apparent 

in both laboratory and in-situ tests. The soils in this region are highly variable due to depositional 

history (location relative to a distributary channel) resulting in highly variable engineering properties 

across the Site. 

SUBSURFACE EXPLORATION AND TESTING PROGRAM

A basic description of the subsurface stratigraphy at the Site was evaluated from SPT blow count 

values and soil index data collected from disturbed split spoon samples across the Site. In-situ 

testing including CPTu, DMT, and self-boring pressuremeter (SBPMT) results are also provided. 

Undisturbed samples for laboratory consolidation and strength testing were obtained from thinwalled Shelby tubes. 

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A-75 

DEVELOPMENT OF A SUBSURFACE PROFILE FOR DESIGN 

As discussed in the previous example problems, methods commonly employed to determine the 

subsurface profile include the SPT, CPTu, and the DMT. In this example, data collected from each 

of these tests are presented and compiled to develop a subsurface profile. 

Use of SPT and Disturbed Sampling

Figure A-53 shows a boring log for the Site. The log indicates an 8 m thick layer of stiff to very stiff 

clay overlying about 1 m of silty clay. Below this, there is an 11-m thick layer of stiff clay with sand 

and clayey silts. The log indicates a definitive transition at approximately a depth of 8 m where SPT 

blowcounts decrease from 18 to 7 within 1.5 meters. Correlation with SPT blowcount values 

indicates that the clayey soils are medium stiff to very stiff clays and silts. 

Index test results are included in figure A-54. These results indicate that the total unit weight is 

approximately constant within the Beaumont and Montgomery layer with average values of 19.9 

kN/m3 (Beaumont) and 20.7 kN/m3

 (Montgomery). The moisture content is typically around 20 

percent with a decrease at the Beaumont/Montgomery interface. In the Beaumont formation, 

moisture contents are, overall, very close to the plastic limit of the soil indicating that the soils in this 

formation are heavily overconsolidated. Soils within the Beaumont are much more plastic than the 

Montgomery as shown in the plasticity chart (figure A-55). 

Table A-18 provides a detailed assessment of the stratigraphy based on the boring information and 

index testing results. Six individual layers are identified based on this information. This layering is 

used to evaluate results from CPT, CPTu, and DMT in the following sections. 

Table A-18. Soil stratigraphy from SPT and classification testing. 

Layer 

ID(1)

Depth 

(m) Classification and Description 

1 0 – 3 Very stiff clay of varying plasticity (CL-CH). 

2 3 - 7 Stiff to very stiff clay of high plasticity (CH). Highly plastic soils 

with liquid limits greater than 50. 

3 7 - 8.5 Medium Stiff Silty Clay (CL). Transition zone from Beaumont to 

Montgomery. Moisture content and plasticity decreases. 

4 8.5 - 

14 

Stiff to very stiff sandy clay with sand pockets (CL). N values 

increase. Soils less plastic with lower moisture content than the 

Beaumont soils. 

5 14 - 18 Dense silt with clayey silt and sand layers (ML). Soil layer is 

sandier as indicated by higher blowcounts and less plastic soils. 

6 18 - ? Stiff clay (CL). 

(1)The Layer ID numbers in table will correspond to the Layer ID numbers for CPTu and DMT results 

to be discussed subsequently. 

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A-76 

Figure A-53. Boring Log UHSPT1 

Archived Document
A-77 

15 20 25

Unit Weight, γt (kN/m3

)

0 20 40 60 80

Moisture Content, wn (%)

wn

LL

PL

0

2

4

6

8

10

12

14

16

18

20

0 20 40 60

N-Value

Depth (m)

Beaumont

Montgomery

Figure A-54. Summary test data from samples across site. 

0

10

20

30

40

50

60

70

80

90

0 20 40 60 80 100 120

Liquid Limit (LL)

Plasticity Index (PI)

Beaumont

Montgomery

A-Line

U-line

Figure A-55. Atterberg limits results for soil samples. 

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A-78 

Stratigraphy from CPTu Data

Three CPTu soundings designated UHDU1, UHDU2, and UHDU3 are shown in figure A-56. For 

UHDU1 and UHDU3, pore pressures were measured on the sleeve just behind the cone tip (u2), and 

for UHDU2, pore pressures were measured at the cone tip (u1). The six layer designations 

previously developed based on boring log information (see table A-18) are used in the subsurface 

characterization based on CPTu measurements. 

-100 50 200 350 500

u2 (kPa)

UHDU3

02468

FR (%)

UHDU1

UHDU2

0 500 1000

fs (kPa)

0

2

4

6

8

10

12

14

16

18

20

0 10 20 30

qt (MPa)

Depth (m)

-500 2000 4500

u1 (kPa)

2

3

4

5

6

1

Layer

ID

Figure A-56. Results of CPT measurements at site. 

Based on a review of the CPTu data, the following interpretations can be made: 

• A transition from Layer 1 to Layer 2 can be inferred from the significant reduction in pore 

pressure in the u2 measurement. These negative pore pressure values indicate that this layer 

(Layer 2) is overconsolidated. 

• The transition between Layer 2 and Layer 3 can be inferred by an overall reduction in 

friction ratio and a significant decrease in penetration pore pressure (u1). This indicates that 

Layer 3 is more silty and/or sandy than the more clayey Layer 1. 

• Layer 4 is clearly evidenced for the depth range of 8.5 to 14 m by a slight increase in the qt

with depth. The increase in qt coupled with a slight decrease in FR indicates that Layer 4 is 

siltier than Layer 3 as shown in the classification charts presented in figure A-57. Also, the 

qt response is more variable than for previous depth increments indicating the presence of 

harder silt or sand seams within the silty clay to clayey silt layer. 

• Below 14 m the increase in qt is consistent with the increase in blowcounts shown in the 

SPT and indicates sandy silt to sand. The spikes in the qt response showing increases 

correlate to similar spikes showing decreased penetration pore pressures, thus indicating 

sandier seams. 

Archived Document
A-79 

Figure A-57. Classification of soils using CPT derived parameters. 

Robertson et al. 

(1986) Soil 

Behavior Type

1 - Sensitive

 fine grained

2 - Organic 

 material

3 - Clay

4 - Silty clay to 

 clay

5 - Clayey silt to

 silty clay

6 - Sandy silt to

 clayey silt

7 - Silty sand to 

 sandy silt

8 - Sand to silty

 sand

9 - Sand

10 - Gravelly 

 sand to sand

11 - Very stiff

 fine grained*

12 - Sand to

 clayey sand*

*Overconsolidated 

or Cemented

Figure EX3-5. Classification of soils using CPT derived parameters.

Beaumont Montgomery

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A-80 

Stratigraphy from DMT Data

Figure A-58 shows the results of four DMTs performed at the Houston site. In figure A-59, the 

index parameters ID and ED are used to estimate classification and consistency. 

The material index, ID, indicates that the site soils range from silty clays to silty sands. Soils closer 

to and within the “sand” classification are primarily within the thin sand and silt seams of the 

Montgomery formation. Borderline points (i.e., at the sand/silt transition around ID = 2) are near the 

Beaumont/Montgomery interface at a depth of approximately 8 m. 

Overall, the ED parameter increases with depth indicating that the Montgomery formation is stiffer 

than the Beaumont formation. Above 8 m, different soundings show relatively similar results. 

Below 8 m, sounding results are much more variable. This is consistent in that the Montgomery 

formation is known to be more variable (i.e., more silt and sand content/seams within clayey matrix). 

0

2

4

6

8

10

12

14

16

18

20

0 2000 4000

Pressure (kPa)

Depth (m)

p0

p1

uo

ED (MPa)

1 10 100

UHDMT1

UHDMT2

UHDMT3

UHDMT4

KD

0.1 1 10 0 10 20

ID Layer

ID

2

3

4

5

6

1

Figure A-58. Results of DMT measurements at site. 

Archived Document
A-81 

10

100

1000

0.1 1 10

Material Index, ID

ED (atm)

Beaumont

Montgomery

 Clay Silt Sand

Figure A-59. Classification of soils using DMT derived parameters. 

CONSOLIDATION TEST RESULTS 

Laboratory one-dimensional consolidation tests were performed on four samples at various depths. 

The results of the testing are summarized in table A-19. Figure A-60 shows the determination of 

Ccε, Crε, and σp′ for a sample from a depth of 7 m.

Table A-19. Summary of oedometer testing on heavily overconsolidated clays. 

Boring Depth γd σvo' σp' OCR Ccε Crε Strain to σ'vo Sample Quality

(m) (kN/m3

) (kPa) (kPa) (%)

UHS2 2.9 16.5 56 400 7.1 0.19 0.02 0.4 Very Good

UHS2 7.0 14.6 97 1,200 12.4 0.18 0.03 1.5 Acceptable

UHS3 10.1 18.5 128 250 2.0 0.06 0.01 1.0 Acceptable

UHS2 11.3 17.6 140 650 4.6 0.08 0.01 2.0 Disturbed

UHS2 14.3 16.9 171 1000 5.8 0.10 0.01 1.2 Acceptable

Note: Preconsolidation stress was calculated using the Stain Energy Method. 

Compression Parameters

Based on these data, a Ccε for design of 0.20 is appropriate for the Beaumont layer and 0.10 is 

appropriate for the Montgomery layer. This difference in compression index is consistent with the 

fact that the Montgomery formation is sandier than the Beaumont formation. As can be seen from 

figure A-60, these clayey soils undergo significant rebound (or swelling) upon unloading. For this 

reason, the practice of loading the soil to just before the preconsolidation stress and then unloading 

and using information from the unload-reload cycle to obtain a recompression index is likely to be 

conservative because of the significant swelling that occurs during unloading. 

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A-82 

Also, if significant swelling is allowed to occur, particularly at low stress levels, then this swelling 

may cause the soil to become significantly destructured as compared to the in-situ material. In this 

case, if the soil is then reloaded into the virgin compression region, the interpreted value of Ccε may 

not be representative for design calculations. In other words, if the design will only include 

compressive loadings and consolidation and no unload cycles, then the value for Ccε obtained in the 

laboratory for a soil that has been subjected to swelling may not be representative. This issue should 

be considered when developing a laboratory consolidation-testing program. 

0

50

100

150

200

250

300

0 1000 2000 3000 4000

Stress (kPa)

Work (kJ/m3)

σ'

p

0

5

10

15

20

25

1 10 100 1000 10000

Effective Consolidation Stress, σ'

p (kPa)

Vertical Strain, 

εv (%)

Ccε

Crε σ'

p

Figure A-60. Evaluation of Ccε, Crε, and σp′ using oedometer test results for sample at depth of 7 m. 

Preconsolidation Stress

Figure A-61 summarizes the results of the evaluation of preconsolidation stresses from laboratory 

oedometer testing (using the Strain Energy approach) and from correlations to in-situ testing 

parameters. These data indicate that the DMT data closely matches the preconsolidation stresses 

Archived Document
A-83 

evaluated based on laboratory oedometer tests. These data also indicate a somewhat typical result 

for overconsolidated deposits, that being higher OCR values at shallower depths. Also, the data 

indicate that the deposit is still overconsolidated at depths of up to 14 m. 

Significant scatter is apparent in the CPTu correlations due to the presence of sandy and silty layers 

and seams. It is noted, however, that if the CPTu data from depths less than 2 m (above the ground 

water table and where desiccation effects are most prevalent) and from depths greater than 8 m are 

omitted, the CPTu correlation for σp′ tends to envelope the values obtained from the DMT and 

oedometer with an average OCR over this depth range of about 6 to 8. At depths greater than 8 m 

(i.e., the Montgomery formation), the CPTu correlations consistently overpredict OCR values 

relative to the DMT and oedometer test results. Based on CPTu classification, the soils in the 

Montgomery formation have tip resistances on the order of 2 to 5 MPa and a friction ratio of about 3 

indicating a more silty and sandier material. The correlations are not appropriate for estimating 

preconsolidation stresses in these materials. Since the correlations are not appropriate for a depth 

range of 8 to 14 m, it would be appropriate to adopt an OCR profile based mostly on oedometer 

results. 

Based on these evaluations, it would be appropriate to use a range of OCR profiles for design 

analyses. One such profile would assume a constant value of OCR equal to 7 for a depth range of 0 

to 8 m and a constant value of OCR of about 4 for a depth range of 8 to 14 m.

0

2

4

6

8

10

12

14

16

18

20

0 1000 2000 3000 4000 5000

Preconsolidation Stress, σ'p (kPa)

Depth (m)

Oedometer

CPTu (Eq. 41)

CPTu (Eq. 42)

DMT

0 5 10 15

OCR

Oedometer

CPT (Eq. 41)

CPT (Eq. 42)

DMT

Figure A-61. σp′ and OCR with depth from laboratory and in-situ tests. 

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A-84 

IN-SITU HORIZONTAL STRESS STATE 

In heavily overconsolidated soils, Ko can be an important design parameter since horizontal stresses 

can be significantly greater than vertical stresses. Figure A-62 shows values of Ko with depth for the 

site using various correlations. The SBPMT is a device that is capable of providing a direct 

measurement of the in-situ horizontal stress. For the SBPMT, Ko is calculated as the ratio of 

effective horizontal stress divided by vertical effective stress. The effective horizontal stress is given 

by (po –uo) where po is the pressuremeter lift-off pressure and uo is the hydrostatic pore pressure in 

the ground. 

0

2

4

6

8

10

12

14

16

18

20

012345

Ko

Depth (m)

SBPMT

DMT (Figure A-63)

CPT (Figure A-64)

Equation 35

Figure A-62. Calculated Ko with depth from in-situ tests. 

Parameters from other in-situ testing devices such as the SPT, CPT, CPTu, and DMT have been 

correlated to Ko. Figures A-63 and A-64 are used to evaluate Ko from CPTu and DMT data. Ko for 

the DMT is calculated based on the equation shown in Figure A-63 assuming βk equal to 0.9 for 

fissured clays. Other correlations (not shown here) are available based on SPT N values and 

Archived Document
A-85 

penetration pore pressures from piezocones. For the methods shown in figures A-63 and A-64 (and 

others), the standard deviation is approximately 0.5. Values for Ko were also calculated according to 

equation 36 using OCR values from the oedometer test results with an assumed value for φ′of 25o

. 

For all methods, the overall trend of the data is similar; values of Ko closely follow the trend in OCR 

with higher values for Ko evident at the shallower depths (based on DMT and SBPMT) and 

progressively decreasing with depth. In selecting a profile for design, an upper and lower bound 

profile of Ko should be evaluated and then the design engineer should evaluate which Ko profile is 

critical. For example, a high Ko value may indicate that excellent resistance could be developed in a 

tiedown anchor while at the same time, a high value would indicate that a diaphragm wall would be 

subjected to potentially very large lateral loads during service.

Figure A-63. Ko correlated with KD from DMT data (Kulhawy and Mayne, 1990). 

Figure A-64. Ko correlated with qt from CPT data (Kulhawy and Mayne, 1990). 

Archived Document
A-86 

UNDRAINED SHEAR STRENGTH 

Figure A-65 provides a summary of undrained shear strength, su, with depth for the Site. Laboratory 

UU and CIU triaxial tests are shown. The UU test results show significant scatter due in part to the 

local joint structure and the presence of sand and silt seams. Above 2 m, the UU test results show 

undrained strengths of approximately 150 kPa, on average. Below 2 m, the CIU test results are more 

consistent than the UU test results in that there is a general trend of increasing su with depth. 

0

2

4

6

8

10

12

14

16

18

20

0 100 200 300 400

su (kPa)

Depth (m)

UU

CU

CPT (Nk=19 (< 8m); Nk=23 (> 8m))

DMT (ds=0.20 for TXC)

DMT (ds=0.14 for DSS)

CPT (Nk=30)

Figure A-65. Undrained shear strength, su, from laboratory and in-situ testing. 

Several correlations using in-situ tests were also used to evaluate su. Undrained strength using CPTu 

results was calculated according to the following: 

k

vot

u N

q sσ− = (Equation A-35) 

where Nk is equal to 19 in the Beaumont formation and 23 in the Montgomery formation. These 

values for Nk have been developed based on local calibration of CPT tip resistance results for soil 

Archived Document
A-87 

formations in the Houston area. In figure A-65, the data used for the CPT correlation are the average 

of soundings UHDU1 and UHDU3 (in which qt can be evaluated since u2 measured). 

An additional correlation using the CPTu data is shown in figure A-65 for an assumed Nk factor 

equal to 30. As shown by Powell and Quarterman (1988), the effect of fissuring is to increase the 

correlated Nk factor, resulting in a lower correlated undrained shear strength for a given cone tip 

resistance. It is not possible to establish specific guidance relating the degree of fissuring to 

undrained shear strength, however, for highly fissured materials, the undrained shear strength 

appropriate for design analyses can be only 50 percent of the measured intact shear strength of the 

soil. This highlights the importance of understanding the degree of fissuring (e.g., size of soil 

“blocks” compared to cone tip size) when attempting to use undrained strength properties for design. 

The correlation for the DMT uses the relationship suDMT = dsσvo' (0.5KD)

1.25 where ds depends on the 

shear mode used. In figure A-65, values for ds of 0.14 and 0.20 were used corresponding to triaxial 

compression and direct simple shear, respectively. The DMT data is the average of three soundings 

completed at the Site. The profiles of su from the DMT indicate that the correlation based on a 

triaxial compression mode of shear, on average, provides higher values of su as compared to other 

methods whereas the DMT correlation based on the direct simple shear mode appears to provide a 

good average fit to all the data. As with the interpretation of OCR, the correlations are much better 

for depths less than 8 m in the Beaumont formation. 

Based on these data and the uncertainties associated with the evaluation of undrained shear strengths 

in fissured materials, it would be reasonable to select a constant value of su for design of 100 kPa for 

depths from 0 to 8 m. It is not clear whether an undrained strength characterization would be 

appropriate for depths greater than 8 m due to the increased sand content of this layer. However, 

based on figure A-65, depths up to 14 m could be conservatively assigned a constant undrained 

strength of 100 kPa or an undrained strength increasing from 100 kPa at 8 m to 150 kPa at 14 m. 

SWELL POTENTIAL 

Figure A-66 was used to evaluate swell potential for the soils at the Site. In using this figure, 

average index properties (i.e., in-situ dry density and liquid limit) were calculated for the soils of the 

Beaumont formation and for the Montgomery formation. The average data plotted on figure A-66 

indicate that the Beaumont soils have high expansion potential compared to low expansion potential 

for Montgomery soils. As discussed in chapter 7, there are other methods available to evaluate swell 

potential, however, the information shown in figure A-66 would be sufficient to alert the engineer to 

the potential for significant swelling in the Beaumont formation (i.e., upper 8 or 9 m). Conservative 

local practice in these soils would likely preclude the use of shallow foundations unless specific 

measures were designed to minimize moisture changes in the upper reaches of the soil profile. To 

develop such designs would require more detailed information on swell potential such as that 

available from one-dimensional swell tests. 

Archived Document
A-88 

Figure A-66. Guide to expansion and collapse potential (adapted from Holtz and Kovacs, 1986). 

Archived Document
 B-1 

APPENDIX B 

CALCULATION OF THE COEFFICIENT OF CONSOLIDATION 

FROM LABORATORY DATA 

B.1 OVERVIEW

The calculation of the coefficient of consolidation using laboratory data is typically performed using 

the following methods; (1) Casagrande’s logarithm of time method; and (2) Taylor’s square root of 

time method. These graphical methods provide an estimate of the time to reach a certain percentage 

of consolidation (e.g., 50 percent consolidation for the logarithm of time method and 90 percent 

consolidation for the square root of time method). Step by step procedures for the logarithm of time 

and square root of time methods are presented in the following sections. 

B.2 LOGARITHM OF TIME METHOD

During a specific loading increment of an oedometer test, the deformation data will be recorded with 

time. To assess the vertical coefficient of consolidation, cv, the data may be plotted versus time on a 

log scale. A graphical construction can be performed to assess the time to reach 50 percent 

consolidation, t50. This procedure is described below and presented in figure B-1. 

Figure B-1. Casagrande logarithm of time method for cv calculation. 

1. Extend the straight-line portions of primary and secondary consolidation so that they intersect. 

This point is marked A on figure B-1, and represents 100 percent primary consolidation. 

2. Since the curve is on a log scale, the deformation value at t=0 will need to be estimated. The 

initial portion of the deformation – log time curve is approximated as a parabola. Select two 

Archived Document
 B-2 

times, t1 and t2, such that t2 = 4t1. The difference in deformation values associated with these 

two time increments is identified as x. 

3. Draw a horizontal line at the deformation value corresponding to x less than the deformation 

value at t1 (line DE in Figure B-1). It may be desirable to perform this construction for a number 

of time pairs to increase the accuracy of the t=0 estimate. This deformation value (i.e., that 

corresponding to line DE) approximates zero percent consolidation. 

4. Draw a horizontal line equal to the deformation value half way between the deformation at zero 

percent consolidation and the deformation value corresponding to 100 percent consolidation. 

This line intersects the time rate consolidation curve (point F on figure B-1) at a time equal to t50. 

5. The cv value for this increment is calculated as: 

50

2 197.0

t

H

c DR

v

⋅ = (Equation B-1) 

where HDR is the drainage height for the specimen. In a double drained specimen, HDR is equal to 

the specimen height, and in a specimen with single drainage, HDR is equal to half the specimen 

height. 

B.3 SQUARE ROOT OF TIME METHOD

For this procedure, the deformation value for each time increment is plotted versus the square root of 

time. A graphical construction can be performed to assess the time to reach 90 percent 

consolidation, t90. This procedure is described below and presented in figure B-2. 

1. Extend the linear portion of the early part of the time rate consolidation curve, as indicated by 

line AB on figure B-2. 

2. Draw a second line with the same origin on the y-axis (point A), and intersecting the x-axis at a 

value 15 percent greater than the line drawn for step 1. This implies that line OC=1.15 OB. 

3. The square root of time value corresponding to the intersection of the line drawn for step 2 and 

the time rate consolidation curve (point D) is assumed equal to the square root of the time to 

reach 90 percent consolidation. The value for t90 is obtained by squaring this value. 

4. The cv value for this increment is calculated as: 

90

2 848.0

t

H

c DR

v

⋅ = (Equation B-2)

Archived Document
 B-3 

Figure B-2. Taylor’s square root of time method for cv calculation. 

Archived Document
C-1 

APPENDIX C 

ALTERNATIVE APPROACH TO EVALUATE HORIZONTAL 

COEFFICIENT OF CONSOLIDATION VALUES 

FROM PIEZOCONE DISSIPATION TESTS 

OVERVIEW 

A method for evaluating the horizontal coefficient of consolidation, ch, from piezocone dissipation 

test data (Burns & Mayne, 1998) is presented in this appendix. The solution presented below has 

been shown to work well for monotonic decay (∆u always decreasing) as well as a dilatory response 

(∆u increases for some or all of the dissipation). 

CALCULATION OF RIGIDITY INDEX 

To calculate ch, the rigidity index, Ir, of the soil is estimated. This property is the ratio of the soil 

shear stiffness, G, to the undrained shear strength, su, and can be calculated from piezocone 

penetration test data and complimentary strength test data as: 





 



 −















−

− ⋅



 



 = + 925.2 925.2 5.1

exp

uq 2

q

M

I

t

vot

R

σ (Equation C-1) 

where qt is the cone tip resistance corrected for porewater pressures, σvo is the total vertical stress, u2

is the penetration pore pressure measured behind the tip, and M is the slope of the critical state line 

equal to: 

'sin3

'sin6

φ

φ

− M = (Equation C-2) 

where φ′ is the effective stress friction angle of the soil. 

CALCULATION OF ch

In lieu of merely matching one point on the dissipation curve (i.e., t50), the entire curve is matched 

using the method presented herein to provide the best estimate of the value of ch. The graph of pore 

pressure decay is plotted on a normalized scale as the ratio of excess pore pressure at time t, ∆ut, to 

the initial value of excess pore pressure during penetration, ∆ui. 

The measured initial excess porewater pressure (∆ui = u2 - uo) is expressed as: 

( ) ( ) i oct i shear i ∆+∆=∆ uuu (Equation C-3) 

The initial octahedral component of the excess porewater pressure (∆uoct)i is equal to: 

( ) ( )( ) ( ) oct i Rvo OCRMu ln2/3/2 I ' σ

Λ =∆ (Equation C-4) 

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C-2 

where OCR is the overconsolidation ratio, Λ is the critical state pore pressure parameter equal to 

about 0.8, and other parameters are as defined above. 

The initial shear induced component of the excess porewater pressure (∆ushear)i is equal to: 

 ( ) ( ) [ ] ' 2/1 shear i OCR vo u σ

Λ −=∆ (Equation C-5) 

The porewater pressures at any time are obtained in terms of the modified time factor, T*, from: 

 ( ) [ ] ( ) [ ] 1 1 *501 *50001 − − uTuu +∆++∆=∆ T t oct i shear i (Equation C-6) 

 T* = (cht)/(a

2

IR

0.75) (Equation C-7) 

where a is the radius of the penetrometer. 

To estimate ch, an iterative solution will need to be performed.

1. Plot normalized dissipation data from the test on a semi-log scale of time vs. normalized 

porewater pressure, ∆u/∆ui (see figure C-1 as an example). 

2. Recalculate ∆ut using equation C-6, and iterate so that a good match of the data from step (1.) is 

computed. Equations C-1, C-2, C-4, C-5, and C-7 will need to be used for generation of the 

calculated dissipation curve. By varying the value of ch in equation C-7, the value of T* will 

change and thus alter ∆ut. The value of ch that yields the best match of field data is the best 

estimate of the field horizontal coefficient of consolidation. 

Figure C-1. Comparison of measured and predicted piezocone dissipation data. 

Archived Document