Design and Construction of Driven Pile

Foundations—Lessons Learned on the

Central Artery/Tunnel Project

PUBLICATION NO. FHWA-HRT-05-159 JUNE 2006

Research, Development, and Technology

Turner-Fairbank Highway Research Center

6300 Georgetown Pike

McLean, VA 22101-2296
FOREWORD 

The purpose of this report is to document the issues related to the design and construction of 

driven pile foundations at the Central Artery/Tunnel project. Construction issues that are 

presented include pile heave and the heave of an adjacent building during pile driving. 

Mitigation measures, including the installation of wick drains and the use of preaugering, proved 

to be ineffective. The results of 15 dynamic and static load tests are also presented and suggest 

that the piles have more capacity than what they were designed for. The information presented in 

this report will be of interest to geotechnical engineers working with driven pile foundation 

systems. 

Gary L. Henderson 

Director, Office of Infrastructure 

 Research and Development 

NOTICE 

This document is disseminated under the sponsorship of the U.S. Department of Transportation 

in the interest of information exchange. The U.S. Government assumes no liability for the use of 

the information contained in this document. 

The U.S. Government does not endorse products or manufacturers. Trademarks or 

manufacturers’ names appear in this report only because they are considered essential to the 

objective of the document. 

QUALITY ASSURANCE STATEMENT 

The Federal Highway Administration (FHWA) provides high-quality information to serve 

Government, industry, and the public in a manner that promotes public understanding. Standards 

and policies are used to ensure and maximize the quality, objectivity, utility, and integrity of its 

information. FHWA periodically reviews quality issues and adjusts its programs and processes to 

ensure continuous quality improvement. 
Technical Report Documentation Page 

1. Report No.

FHWA-HRT-05-159 

2. Government Accession No. 3. Recipient’s Catalog No.

5. Report Date 

June 2006 

4. Title and Subtitle 

Design and Construction of Driven Pile Foundations— 

Lessons Learned on the Central Artery/Tunnel Project 6. Performing Organization Code

7. Author(s) 

Aaron S. Bradshaw and Christopher D.P. Baxter 

8. Performing Organization Report No. 

9. Performing Organization Name and Address 10. Work Unit No. 

University of Rhode Island 

Narragansett, RI 02882 11. Contract or Grant No. 

DTFH61-03-P-00174 

13. Type of Report and Period Covered 

Final Report 

January 2003–August 2003 

12. Sponsoring Agency Name and Address 

Office of Infrastructure Research and Development 

Federal Highway Administration 

6300 Georgetown Pike 

McLean, VA 22101-2296 

14. Sponsoring Agency Code 

15. Supplementary Notes 

Contracting Officer’s Technical Representative (COTR): Carl Ealy, HRDS-06 

16. Abstract 

Five contracts from the Central Artery/Tunnel (CA/T) project in Boston, MA, were reviewed to document issues 

related to design and construction of driven pile foundations. Given the soft and compressible marine clays in the 

Boston area, driven pile foundations were selected to support specific structures, including retaining walls, 

abutments, roadway slabs, transition structures, and ramps. This report presents the results of a study to assess the 

lessons learned from pile driving on the CA/T. This study focused on an evaluation of static and dynamic load test 

data and a case study of significant movement of an adjacent building during pile driving. The load test results 

showed that the piles have more capacity than what they were designed for. At the site of significant movement of an 

adjacent building, installation of wick drains and preaugering to mitigate additional movement proved to be 

ineffective. Detailed settlement, inclinometer, and piezometer data are presented. 

17. Key Words 

Driven piles, heave, CAPWAP, static load test, 

Boston tunnel 

18. Distribution Statement 

No restrictions. This document is available to the public 

through the National Technical Information Service, 

Springfield, VA 22161. 

19. Security Classif. (of this report) 

Unclassified 

20. Security Classif. (of this 

page) 

Unclassified 

21. No. of Pages

58 

22. Price 

Form DOT F 1700.7 (8-72) Reproduction of completed page authorized
ii 

SI* (MODERN METRIC) CONVERSION FACTORS 

APPROXIMATE CONVERSIONS TO SI UNITS

Symbol When You Know Multiply By To Find Symbol 

LENGTH 

in inches 25.4 millimeters mm 

ft feet 0.305 meters m 

yd yards 0.914 meters m 

mi miles 1.61 kilometers km 

AREA 

in2 square inches 645.2 square millimeters mm2

ft2 square feet 0.093 square meters m2

yd2 square yard 0.836 square meters m2

ac acres 0.405 hectares ha

mi2 square miles 2.59 square kilometers km2

VOLUME 

fl oz fluid ounces 29.57 milliliters mL 

gal gallons 3.785 liters L 

ft3 cubic feet 0.028 cubic meters m3 

yd3 cubic yards 0.765 cubic meters m3 

NOTE: volumes greater than 1000 L shall be shown in m3

MASS 

oz ounces 28.35 grams g

lb pounds 0.454 kilograms kg

T short tons (2000 lb) 0.907 megagrams (or "metric ton") Mg (or "t") 

TEMPERATURE (exact degrees) o

F Fahrenheit 5 (F-32)/9 Celsius oC 

or (F-32)/1.8 

ILLUMINATION 

fc foot-candles 10.76 lux lx

fl foot-Lamberts 3.426 candela/m2 cd/m2

FORCE and PRESSURE or STRESS 

lbf poundforce 4.45 newtons N 

lbf/in2 poundforce per square inch 6.89 kilopascals kPa 

APPROXIMATE CONVERSIONS FROM SI UNITS 

Symbol When You Know Multiply By To Find Symbol 

LENGTH

mm millimeters 0.039 inches in 

m meters 3.28 feet ft 

m meters 1.09 yards yd 

km kilometers 0.621 miles mi 

AREA mm2

 square millimeters 0.0016 square inches in2 

m2 square meters 10.764 square feet ft2 

m2 square meters 1.195 square yards yd2 

ha hectares 2.47 acres ac

km2 square kilometers 0.386 square miles mi2 

VOLUME 

mL milliliters 0.034 fluid ounces fl oz 

L liters 0.264 gallons gal 

m3 cubic meters 35.314 cubic feet ft3 

m3 cubic meters 1.307 cubic yards yd3 

MASS 

g grams 0.035 ounces oz

kg kilograms 2.202 pounds lb

Mg (or "t") megagrams (or "metric ton") 1.103 short tons (2000 lb) T 

TEMPERATURE (exact degrees) o

C Celsius 1.8C+32 Fahrenheit oF 

ILLUMINATION 

lx lux 0.0929 foot-candles fc 

cd/m2 candela/m2 0.2919 foot-Lamberts fl

FORCE and PRESSURE or STRESS 

N newtons 0.225 poundforce lbf 

kPa kilopascals 0.145 poundforce per square inch lbf/in2

*SI is the symbol for th International System of Units. Approp e riate rounding should be made to comply with Section 4 of ASTM E380. 

(Revised March 2003)
iii 

TABLE OF CONTENTS 

Page

CHAPTER 1. INTRODUCTION................................................................................................ 1

ROLE OF DRIVEN PILE FOUNDATIONS ON THE CA/T PROJECT ......................... 1 

OBJECTIVES..................................................................................................................... 3 

SCOPE ................................................................................................................................ 3 

CHAPTER 2. DRIVEN PILE DESIGN CRITERIA AND SPECIFICATIONS.................... 5

SUBSURFACE CONDITIONS ......................................................................................... 5 

DESIGN CRITERIA AND SPECIFICATIONS................................................................ 9 

Pile Types .................................................................................................................... 9 

Preaugering Criteria................................................................................................... 10 

Pile Driving Criteria .................................................................................................. 10 

Axial Load and Pile Load Test Criteria..................................................................... 13 

CHAPTER 3. CONSTRUCTION EQUIPMENT AND METHODS .................................... 15

EQUIPMENT AND METHODS ..................................................................................... 15 

CONSTRUCTION-RELATED ISSUES.......................................................................... 19 

Pile Heave.................................................................................................................. 19 

Soil Heave.................................................................................................................. 21 

Summary.................................................................................................................... 27 

CHAPTER 4. DYNAMIC AND STATIC PILE LOAD TEST DATA .................................. 29

LOAD TEST METHODS................................................................................................. 29 

Dynamic Load Test Methods..................................................................................... 29 

Static Load Test Methods .......................................................................................... 30 

LOAD TEST RESULTS................................................................................................... 33 

Dynamic Results and Interpretation .......................................................................... 35 

Comparison of CAPWAP Data ................................................................................. 38 

Static Load Test Data................................................................................................. 39 

Comparison of Dynamic and Static Load Test Data ................................................. 41 

CHAPTER 5. COST DATA OF DRIVEN PILES................................................................... 43

CHAPTER 6. LESSONS LEARNED ....................................................................................... 45

REFERENCES............................................................................................................................ 47
iv 

LIST OF FIGURES 

Page

Figure 1. Locations of selected contracts from the CA/T project................................................... 2 

Figure 2. Soil profile at the contract C07D1 site as encountered in Boring EB3-5........................ 6 

Figure 3. Soil profile at the contract C07D2 site as encountered in Boring EB2-149.................... 7 

Figure 4. Soil profile at the contract C08A1 site as encountered in Boring EB6-37...................... 7 

Figure 5. Soil profile at the contract C09A4 site as encountered in Boring IC10-13. ................... 8 

Figure 6. Soil profile at the contract C19B1 site as encountered in Boring AN3-101. .................. 8 

Figure 7. Typical pile details for a 30-cm-diameter PPC pile. ..................................................... 11 

Figure 8. Typical pile details for a 41-cm-diameter PPC pile with stinger. ................................. 12 

Figure 9. Single-acting diesel hammer. ........................................................................................ 16 

Figure 10. Double-acting diesel hammer...................................................................................... 17 

Figure 11. Single-acting hydraulic hammer.................................................................................. 17 

Figure 12. Typical pile driving record. ......................................................................................... 18 

Figure 13. Site plan, piling layout for the arrivals tunnel at Logan Airport. ................................ 19 

Figure 14. Site plan showing locations of piles, building footprint, and geotechnical 

instrumentation. ................................................................................................................ 22 

Figure 15. Settlement data obtained during first phase of pile driving......................................... 23 

Figure 16. Settlement data obtained during second phase of pile driving.................................... 25 

Figure 17. Multipoint heave gauge data obtained during second phase of pile driving. .............. 25 

Figure 18. Pore pressure data obtained during second phase of pile driving. .............................. 26 

Figure 19. Inclinometer data obtained during second phase of pile driving................................. 27 

Figure 20. Example of CAPWAP signal matching, test pile 16A1-1........................................... 30 

Figure 21. Typical static load test arrangement showing instrumentation. .................................. 31 

Figure 22. Load-displacement curves for pile toe, test pile 16A1-1............................................. 37 

Figure 23. CAPWAP capacities at end of initial driving (EOD) and beginning 

of restrike (BOR). ............................................................................................................. 39 

Figure 24. Deflection of pile head during static load testing of pile 12A1-1. .............................. 40 

Figure 25. Distribution of load in pile 12A1-1. ............................................................................ 40 

Figure 26. Deflection of pile head during static load testing of pile 14........................................ 40 

Figure 27. Distribution of load in pile 14. .................................................................................... 40 

Figure 28. Deflection of pile head during static load testing of pile IPW. ................................... 41 

Figure 29. Distribution of load in pile IPW. ................................................................................. 41 
v 

LIST OF TABLES 

Page

Table 1. Summary of selected contracts using driven pile foundations. ........................................ 2 

Table 2. Summary of pile types used on the selected CA/T contracts. ........................................ 10 

Table 3. Summary of pile types and axial capacity (requirements identified in the selected 

contracts). ....................................................................................................................... 13 

Table 4. Summary of pile driving equipment used on the selected contracts............................... 15 

Table 5. Summary of pile spacing from selected contracts. ......................................................... 21 

Table 6. Maximum building heave observed during pile driving................................................. 23 

Table 7. Summary of pile and preauger information.................................................................... 34 

Table 8. Summary of pile driving information. ............................................................................ 34 

Table 9. Summary of CAPWAP capacity data............................................................................. 35 

Table 10. Summary of CAPWAP soil parameters........................................................................ 38 

Table 11. Summary of static load test data................................................................................... 39 

Table 12. Summary of dynamic and static load test data. ............................................................ 42 

Table 13. Summary of contractor’s bid costs for pile driving. ..................................................... 43 

Table 14. Summary of contractor’s bid costs for preaugering...................................................... 43

1 

CHAPTER 1. INTRODUCTION 

Pile foundations are used extensively for the support of buildings, bridges, and other structures to 

safely transfer structural loads to the ground and to avoid excess settlement or lateral movement. 

They are very effective in transferring structural loads through weak or compressible soil layers 

into the more competent soils and rocks below. A “driven pile foundation” is a specific type of 

pile foundation where structural elements are driven into the ground using a large hammer. They 

are commonly constructed of timber, precast prestressed concrete (PPC), and steel (H-sections 

and pipes). 

Historically, piles have been used extensively for the support of structures in Boston, MA. This 

is mostly a result of the need to transfer loads through the loose fill and compressible marine 

clays that are common in the Boston area. Driven piles, in particular, have been a preferred 

foundation system because of their relative ease of installation and low cost. They have played 

an important role in the Central Artery/Tunnel (CA/T) project. 

ROLE OF DRIVEN PILE FOUNDATIONS ON THE CA/T PROJECT 

The CA/T project is recognized as one of the largest and most complex highway projects in the 

United States. The project involved the replacement of Boston’s deteriorating six-lane, elevated 

central artery (Interstate (I) 93) with an underground highway; construction of two new bridges 

over the Charles River (the Leverett Circle Connector Bridge and the Leonard P. Zakim Bunker 

Hill Bridge); and the extension of I–90 to Boston’s Logan International Airport and Route 1A. 

The project has been under construction since late 1991 and is scheduled to be completed in 

2005.(1) 

Driven pile foundations were used on the CA/T for the support of road and tunnel slabs, bridge 

abutments, egress ramps, retaining walls, and utilities. Because of the large scale of the project, 

the construction of the CA/T project was actually bid under 73 separate contracts. Five of these 

contracts were selected for this study, where a large number of piles were installed, and 15 pile 

load tests were performed. The locations of the individual contracts are shown in figure 1 and 

summarized in table 1. A description of the five contracts and associated pile-supported 

structures is also given below. 

1. Contract C07D1 is located adjacent to Logan Airport in East Boston and included 

construction of a part of the I–90 Logan Airport Interchange roadway network. New 

roadways, an egress ramp, retained fill sections, a viaduct structure, and retaining walls were 

all constructed as part of the contract.(2) Driven piles were used primarily to support the 

egress ramp superstructure, abutments, roadway slabs, and retaining walls.

2. Contract C07D2 is located adjacent to Logan Airport in East Boston and included 

construction of a portion of the I–90 Logan Airport Interchange. Major new structures 

included highway sections, a viaduct structure, a reinforced concrete open depressed 

roadway (boat section), and at-grade approach roadways.(2) Driven piles were used to support 

the boat section, walls and abutments, and portions of the viaduct. 
2 

Figure 1. Locations of selected contracts from the CA/T project.(3) 

Table 1. Summary of selected contracts using driven pile foundations. 

Contract Location Description 

C07D1 Logan Airport I–90 Logan Airport Interchange 

C07D2 Logan Airport I–90 Logan Airport Interchange 

C08A1 Logan Airport I–90 and Route 1A Interchange 

C09A4 Downtown I–93/I–90 Interchange, I-93 Northbound 

C19B1 Charlestown I–93 Viaducts and Ramps North of the Charles River 

3. Contract C08A1 is located just north of Logan Airport in East Boston and included 

construction of the I–90 and Route 1A interchange. This contract involved new roadways, 

retained fill structures, a viaduct, a boat section, and a new subway station.(2) Both vertical 

and inclined piles were used to support retaining walls and abutments. 

4. Contract C09A4 is located just west of the Fort Point Channel in downtown Boston. The 

contract encompassed construction of the I–90 and I–93 interchange, and the northbound 

section of I–93. Major new structures included surface roads, boat sections, tunnel sections, 

viaducts, and a bridge.(2) Piles were used to support five approach structures that provide a 

transition from on-grade roadways to the viaduct sections. Piles were also used to support 

utility pipelines.

5. Contract C19B1 is located just north of the Charles River in Charlestown. The contract 

included the construction of viaduct and ramp structures forming an interchange connecting 

Route 1, Storrow Drive, and I–93 roadways. Major new structures included roadway 

transition structures, boat sections, retaining walls, and a stormwater pump station.(2) Piles 

I-93 

I-90 

C19B1 

C9A4 

C7D1/D2 

C8A1 
3 

were used to support the ramp structures that transition from on-grade roadways to the 

viaduct or boat sections.

OBJECTIVES 

The overall objective of this report is to document the lessons learned from the installation of 

driven piles on the CA/T project. This includes review and analysis of pile design criteria and 

specifications, pile driving equipment and methods, issues encountered during construction, 

dynamic and static load test data, and cost data for different pile types and site conditions.

SCOPE 

This report consists of six chapters, the first of which presents introductory and background 

information about the contracts where significant pile driving occurred. The second chapter 

discusses the criteria and specifications used for pile design and construction on the CA/T 

project. The third chapter documents the equipment and methods used for pile driving. Major 

construction issues encountered during driving, such as pile and soil heave, are also discussed. 

The fourth chapter presents the results of pile load tests performed on test piles using static and 

dynamic test methods, including a discussion of axial capacity, dynamic soil parameters, and pile 

driving criteria. The fifth chapter presents the unit costs for pile driving and preaugering for the 

different pile types used, as identified in the original construction bids. Finally, the sixth chapter 

summarizes the important findings of this study.

5 

CHAPTER 2. DRIVEN PILE DESIGN CRITERIA 

AND SPECIFICATIONS 

This chapter presents the pile design criteria and specifications used on the CA/T project in 

contracts C07D1, C07D2, C08A1, C09A4, and C19B1. These include information on the types 

of piles used, capacity requirements, minimum preaugering depths, and testing requirements. 

The subsurface conditions on which the design criteria were based are also discussed. 

SUBSURFACE CONDITIONS 

Representative soil profiles from each of the contract sites are shown in figures 2 through 6 

based on the interpretation of geotechnical borings. (See references 4, 5, 6, 7, 8, 9, and 10) 

As shown in figures 2 through 5, the conditions encountered at sites in East Boston (C07D1, 

C07D2, and C08A1) and in downtown Boston (C09A4) are similar. The subsurface conditions at 

these locations typically consisted of fill overlying layers of organic silt, inorganic sand or silt, 

marine clay, glacial soils, and bedrock. The subsurface conditions shown in figure 6 for the 

C19B1 site in Charlestown, however, were different from the other four sites. Organic soils and 

marine clays were only encountered to a limited extent at the site. Also, the thickness of the fill 

layer was greater relative to the other sites. 

The physical properties and geological origin of the soils encountered at the contract sites are 

described below.(11-12)

Bedrock: The bedrock in the area consists of argillite from the Cambridge formation. The 

condition of the bedrock varies considerably with location, even within a given site. Evaluation 

of rock core samples indicates that the rock is typically in a soft and weathered condition and 

contains a significant amount of fracturing. However, hard and sound bedrock was found at some 

locations. 

Glacial Soils: The glacial soils were deposited during the last glaciation approximately 12,000 

years ago. These deposits include glacial till, and glaciomarine, glaciolacustrine, and 

glaciofluvial soils. Till is characterized by a mass of unsorted debris that contains angular 

particles composed of a wide variety of grain sizes, ranging from clay-sized particles to large 

boulders. Glaciomarine or glaciolacustrine deposits generally consist of clay, silt, and sand, 

whereas glaciofluvial deposits contain coarser grained sand and gravel. The glacial soils are 

typically dense in nature as indicated by high standard penetration test (SPT) resistance, and the 

piles were typically terminated in these deposits. 

Marine Soils: Marine soils were deposited over the glacial soils during glacial retreat in a 

quiescent deepwater environment. The marine clay layer, as shown in figures 2 through 5, is the 

thickest unit in the profile, but was encountered only to a limited extent at the Charlestown site. 

The clay is generally overconsolidated in the upper portions of the layer and is characterized by 

relatively higher strengths. The overconsolidation is a result of past desiccation that occurred 

during a period of low sea level. By comparison, the deeper portions of the clay layer are much 
6 

softer and penetration of the SPT split spoon can sometimes occur with just the weight of the 

drilling rods alone. 

Inorganic Soils: Inorganic silts and sands are typically encountered overlying the marine soils. 

These soils were deposited by alluvial processes. 

Organic Soils: The organic soils that are encountered below the fill generally consist of organic 

silt and may contain layers of peat or fine sand. These soils are the result of former tidal marshes 

that existed along the coastal areas. 

Fill Soils: Fill material was placed in the more recent past to raise the grade for urban 

development. The fill layer is highly variable in its thickness and composition, ranging from silts 

and clays to sands and gravels. The consistency or density is also variable as indicated by the 

SPT blow counts. The variability in the fill is attributed to the characteristics of the particular 

borrow source material and the methods of placement. 

Figure 2. Soil profile at the contract C07D1 site as encountered in Boring EB3-5. 

0

5

10

15

20

25

30

35

40

45

50

55

1 10 100

SPT N Value

Depth (m)

Fill

Sand

Marine Clay

Sand (Glaciofluvial)

Glacial Till

Bedrock

Organic Silt

ET2-C2

EB3-5
7 

Figure 3. Soil profile at the contract C07D2 site as encountered in Boring EB2-149. 

Figure 4. Soil profile at the contract C08A1 site as encountered in Boring EB6-37. 

0

5

10

15

20

25

30

35

40

1 10 100

SPT N Value

Depth (m)

Fill

Sand

Marine Clay

Silt(Glaciomarine)

Bedrock

Organic Silt

923

EB2-149

0

10

20

30

40

50

60

70

1 10 100

SPT N Value

Depth (m)

Fill

Silt and Sand

Marine Clay

Sand and Gravel 

(Glaciofluvial)

Bedrock

ET2-C2

EB6-37
8 

Figure 5. Soil profile at the contract C09A4 site as encountered in Boring IC10-13. 

Figure 6. Soil profile at the contract C19B1 site as encountered in Boring AN3-101. 

0

5

10

15

20

25

1 10 100

SPT N Value

Depth (m)

Granular Fill

Sand

Gravel 

(Glaciomarine)

Glacial Till

Bedrock

Organic Silt

IPW

AN3-101

0

5

10

15

20

25

30

35

40

45

50

1 10 100

SPT N Value

Depth (m)

Fill

Organic Silt

Marine Clay

Sand and Silt 

(Glaciofluvial)

Bedrock

12A1-1

IC10-13
9 

DESIGN CRITERIA AND SPECIFICATIONS 

The variable fill and compressible clay soils encountered at depth necessitated the use of deep 

foundations. Driven piles were selected, and design criteria and specifications were developed 

for their installation, ultimate capacity, and testing. Because the CA/T project was located in 

Massachusetts, the design criteria were required to satisfy the regulations given in the 

Massachusetts State building code.(13) The technical content of the State code is based on the 

1993 edition of the Building Officials and Code Administrators (BOCA) national building code. 

The specifications that were used for each CA/T contract are contained in two documents of the 

Massachusetts Highway Department (MHD). The first document includes the general 

requirements for all CA/T contracts and is entitled Supplemental Specifications and CA/T 

Supplemental Specifications to Construction Details of the Standard Specifications for Highways 

and Bridges (Division II) for Central Artery (I-93)/Tunnel (I-90) Project in the City of Boston.

(14) 

The specifications pertaining to individual contracts are covered in a second document 

concerning special provisions.(15) The special provisions are necessary given the uniqueness of 

the environmental conditions, soil conditions, and structure types found in each contract. The 

special provisions present specific details regarding the pile types, pile capacity requirements, 

and minimum preaugering depths. 

Information selected from the specification regarding pile types, preaugering criteria, pile driving 

criteria, and axial load and test criteria is highlighted below. 

Pile Types 

Two types of piles were specified on the selected contracts of the CA/T: (1) PPC piles, and 

(2) concrete-filled steel pipe piles. The PPC piles were fabricated using 34.5- to 41.3-megapascal 

(MPa) (28-day strength) concrete and were prestressed to 5.2 to 8.3 MPa. The design drawings 

of typical 30-centimeter (cm)- and 41-cm-diameter square PPC piles are shown in figures 7 and 

8, respectively. 

To prevent damage to the pile tips during driving in very dense materials, the PPC piles were 

also fitted with 1.5-meter (m)-long steel H-pile “stingers.” In the 41-cm-diameter PPC piles, an 

HP14x89 section was used as the stinger. The stingers were welded to a steel plate that was cast 

into the pile toe, as shown in figure 8. Stingers were used intermittently on the 30-m-diameter 

PPC piles, consisting of HP10 by 42 sections. 

The concrete-filled steel pipe piles were 31 to 61 cm in diameter, with wall thicknesses ranging 

from 0.95 to 1.3 cm. The piles were driven closed-ended by welding a steel cone or flat plate 

onto the pile tip prior to driving. Once the pile was driven to the required depth, the pile was 

filled with concrete. 

A summary of the pile types used on the CA/T is given in table 2, along with the estimated 

quantities driven. The quantities are based on the contractor’s bid quantities that were obtained 

directly from Bechtel/Parsons Brinckerhoff. As shown in table 2, the 41-cm-diameter PPC piles 

were the dominant pile type used, accounting for more than 70 percent of the total length of pile 

driven. 
10 

Table 2. Summary of pile types used on the selected CA/T contracts. 

Estimated Length of Pile Driven (m) Pile Type C07D1 C07D2 C08A1 C09A4 C19B1 Total 

32-cm pipe - - - - 5,550 5,550 

41-cm pipe - - - 5,578 - 5,578 

61-cm pipe - - - - 296 296 

30-cm square PPC 7,969 3,981 792 3,658 2,177 18,577 

41-cm square PPC 32,918 19,879 8,406 14,326 6,279 81,808 

Preaugering Criteria 

Preaugering was specified for all piles that were installed in embankments or within the specified 

limits of adjacent structures. Settlement problems observed at the Hilton hotel (contract C07D1) 

initiated the use of preaugering to reduce the potential for soil heave caused by pile installation. 

Soil heave is discussed further in chapter 3. The required depth of preaugering varied depending 

on the contract and pile location, but ranged from 7.6 to 32.0 m below the ground surface. 

Pile Driving Criteria 

The specifications required that a Wave Equation Analysis of Piles (WEAP) be used to select the 

pile driving equipment. The WEAP model estimates hammer performance, driving stresses, and 

driving resistance for an assumed hammer configuration, pile type, and soil profile. The 

acceptability of the hammer system was based on the successful demonstration that the pile 

could be driven to the required capacity or tip elevation without damage to the pile, within a 

penetration resistance of 3 to 15 blows per 2.5 cm. 

The pile driving resistance criteria estimated from the WEAP analysis was also used as the initial 

driving criteria for the installation of the test piles. Additional WEAP analyses were required for 

changes in the hammer type, pile type or size, or for significant variations in the soil profile. It 

was also specified that the WEAP analyses be rerun with modifications to the input parameters 

to match the results obtained from the dynamic or static load test results. Modifications to the 

driving criteria could be made as appropriate, based on the results of the pile load tests. 
11

1 foot = 0.30 m 1 inch = 25.4 mm 

Figure 7. Typical pile details for a 30-cm-diameter PPC pile. 
12

1 foot = 0.30 m 1 inch = 25.4 mm 

Figure 8. Typical pile details for a 41-cm-diameter PPC pile with stinger. 
13 

Axial Load and Pile Load Test Criteria 

The required allowable axial capacities that were identified in the special provisions are 

summarized in table 3. Allowable axial load capacities ranged from 311 to 1,583 kilonewtons 

(kN). Lateral load criteria were not identified in the selected contracts. 

Table 3. Summary of pile types and axial capacity 

(requirements identified in the selected contracts). 

Pile Type Required Allowable Axial 

Capacity (kN) 

32-cm pipe 890 

41-cm pipe 1,583 

61-cm pipe 311 

30-cm square PPC 356–756 

41-cm square PPC 534–1,379 

The axial capacity of the piles was verified using pile load tests, which were specified in section 

940.62 of the general specifications.(14) The required ultimate capacities for the load tests were 

specified by applying a minimum factor of safety of 2.0 to the required allowable values. A 

factor of safety of 2.25 was specified in contract C19B1, which is consistent with the 

recommended American Association of State Highway and Transportation Officials (AASHTO) 

criteria for piles designed and evaluated based only on a subsurface exploration, static analysis, 

WEAP analysis, and dynamic pile testing.(16)

Dynamic load testing was required for test piles and for a portion of the production piles to 

monitor driving-induced stresses in the piles, evaluate hammer efficiency and performance, 

estimate the soil-resistance distribution, and evaluate the pile capacity during initial installation 

driving and restrikes. A waiting period of 12 to 36 hours (h) was required after pile installation 

before restrike tests could be performed. 

Static load tests were required for test piles to confirm that the minimum specified allowable 

capacity was achieved and to better estimate or establish higher allowable design capacities. 

Section 1817.4.1 of the Massachusetts State building code says that the load reaching the top of 

the bearing stratum under maximum test load for a single pile or pile group must not be less than 

100 percent of the allowable design load for end-bearing piles. Therefore, the specifications 

required that the static load test demonstrate that 100 percent of the design load was transferred 

to the bearing layer. If any of the test criteria were not met, the contractor was required to 

perform additional static load test(s). 

15 

CHAPTER 3. CONSTRUCTION EQUIPMENT AND METHODS 

This chapter presents a description of the equipment and methods used during pile driving 

operations at the CA/T project in the selected contracts. This includes a general overview of 

impact hammers, how a pile is installed, and how to tell when a pile has reached the desired 

capacity. Construction issues associated with pile driving during this project are also presented. 

Pile heave was identified as an issue during construction of the arrivals tunnel at Logan Airport, 

which required a significant number of piles to be redriven. At another site at the airport, soil 

heave resulting from pile driving caused significant movement of an adjacent building and 

required changes to the installation process, including preaugering the piles to a depth of 26 m. 

EQUIPMENT AND METHODS 

Impact hammers were used to drive all of the piles for the CA/T project. An impact hammer 

consists of a heavy ram weight that is raised mechanically or hydraulically to some height 

(termed “stroke”) and dropped onto the head of the pile. During impact, the kinetic energy of the 

falling ram is transferred to the pile, causing the pile to penetrate the ground. 

Many different pile driving hammers are commercially available, and the major distinction 

between hammers is how the ram is raised and how it impacts the pile. The size of the hammer is 

characterized by its maximum potential energy, referred to as the “rated energy.” The rated 

energy can be expressed as the product of the hammer weight and the maximum stroke. 

However, the actual energy transferred to the pile is much less a result of energy losses within 

the driving system and pile. The average transferred energies range from 25 percent for a diesel 

hammer on a concrete pile to 50 percent for an air hammer on a steel pile.(17)

Three types of hammers were used on the selected contracts: (1) a single-acting diesel, (2) a 

double-acting diesel, and (3) a single-acting hydraulic. The manufacturers and characteristics of 

the hammers used in these contracts are summarized in table 4, along with the pile types driven. 

Schematics of the three types of hammers are shown in figures 9 through 11. 

Table 4. Summary of pile driving equipment used on the selected contracts.

Make and Model Type Action 

Rated 

Energy 

(kN-m) 

Pile Types Driven Contracts Designation 

DelmagTM 

D 46-32 Diesel Double 153.5 41-cm PPC C07D1 I 

HPSI 2000 Hydraulic Single 108.5 41-cm PPC C07D1, 

C07D2 II 

ICE 1070 Diesel Double 98.5 31-cm PPC, 41-cm 

PPC, 41-cm pipe 

C08A1, 

C09A4 III 

HPSI 1000 Hydraulic Single 67.8 41-cm PPC C19B1 IV 

Delmag D 19-42 Diesel Single 58.0 32-cm pipe C19B1 V 

Delmag D 30-32 Diesel Single 99.9 32-cm pipe C19B1 VI 

A single-acting diesel hammer (figure 9) works by initially raising the hammer with a cable and 

then releasing the ram. As the ram free-falls within the cylinder, fuel is injected into the 
16 

combustion chamber beneath the ram and the fuel/air mixture becomes pressurized. Once the 

ram strikes the anvil at the bottom of the cylinder, the fuel/air mixture ignites, pushing the ram 

back to the top of the stroke. This process will continue as long as fuel is injected into the 

combustion chamber and the stroke is sufficient to ignite the fuel. 

Figure 9. Single-acting diesel hammer.(17)

A double-acting diesel hammer (figure 10) works like the single-acting diesel hammer except 

that the system is closed at the top of the ram. As the ram rebounds to the top of the stroke, 

gasses are compressed in the bounce chamber at the top of the hammer. The bounce chamber 

temporarily stores and redirects energy to the top of the ram, allowing the stroke height to be 

reduced and the blow rate to be increased. Bounce chamber pressure is monitored during pile 

driving because it is correlated with hammer energy. The stroke of the hammer, and thus the 

energy, is controlled using the fuel pump. This is effective for avoiding bouncing of the hammer 

during the upstroke, which can lead to unstable driving conditions and damage to the hammer.(17)

A single-acting hydraulic hammer (figure 11) uses a hydraulic actuator and pump to retract the 

ram to the top of the stroke. Once the ram is at the top of the stroke, the ram is released and free-
17 

falls under gravity, striking the anvil. An advantage of hydraulic hammers is that the free-fall 

height, and thus the energy delivered to the pile, can be controlled more accurately. 

Figure 10. Double-acting diesel hammer.(17) Figure 11. Single-acting 

hydraulic hammer.(17)

In preparation for driving, a pile is first hoisted to an upright position using the crane and is 

placed into the leads of the pile driver. The leads are braces that help position the piles in place 

and maintain alignment of the hammer-pile system so that a concentric blow is delivered to the 

pile for each impact. Once the pile is positioned at the desired location, the hammer is lowered 

onto the pile butt. A pile cushion consisting of wood, metal, or composite material is placed 

between the pile and the hammer prior to driving to reduce stresses within the pile during 

driving. 

Once the pile is in position, pile driving is initiated and the number of hammer blows per 0.3 m 

of penetration is recorded. Toward the end of driving, blows are recorded for every 2.5 cm of 

penetration. Pile driving is terminated when a set of driving criteria is met. Pile driving criteria 

are generally based on the following: (1) the minimum required embedment depth, (2) the 

minimum number of blows required to achieve capacity, and (3) the maximum number of blows 

to avoid damage to the pile. All information that is associated with pile driving activities (e.g., 

hammer types, pile types, pile lengths, blow counts, etc.) is recorded on a pile driving log. 
18 

A typical pile driving log is shown in figure 12. This particular record is for the installation of a 

24-m-long, 41-cm-diameter PPC pile installed at the airport as part of contract C07D2. A 

hydraulic hammer with an 89-kN ram and a 1.2-m stroke was used. The number of blows per 0.3 

m of driving was recorded from an embedment depth of 9.5 m to a final depth of 16.5 m. At a 

depth of 16.5 m, the hammer blows required to drive the pile 2.5 cm were recorded in the righthand column of the record. Driving was stopped after a final blow count of 39 blows per 2.5 cm 

was recorded. 

Once a pile has been installed, the hammer may be used to drive the pile again at a later time. 

Additional driving that is performed after initial installation is referred to as a redrive or restrike. 

A redrive may be necessary for two reasons: (1) to evaluate the long-term capacity of the pile 

(i.e., pile setup or pile relaxation), or (2) to reestablish elevations and capacity in piles that have 

been subject to heave. Both of these issues were significant for the CA/T project, and they are 

discussed in the next section. 

Figure 12. Typical pile driving record. 
19 

CONSTRUCTION-RELATED ISSUES 

Pile Heave 

Pile heave is a phenomenon where displacement of soil from pile penetration causes vertical or 

horizontal movement in nearby, previously driven piles. Pile heave generally occurs in 

insensitive clays that behave as incompressible materials during pile driving.(17) In these soils, 

the elevation of adjacent piles is often continuously monitored during driving to look for heave. 

If a pile moves in excess of some predetermined criterion, the pile is redriven to redevelop the 

required penetration and capacity. From a cost perspective, pile heave is important because 

redriving piles can require significant additional time and effort. 

Pile Layout and Soil Conditions 

Of the contracts reviewed, pile heave was an issue during construction of the arrivals tunnel at 

Logan Airport (contract C07D2). The location of the C07D2 site is shown in figure 1. A plan 

view of the arrivals tunnel structure showing the pile locations is shown in figure 13. The tunnel 

structure is approximately 159 m in length and is located where ramp 1A-A splits from the 

arrivals road. The tunnel was constructed using the cut-and-cover method, and thus a portion of 

the overburden soil was excavated prior to pile driving. 

Figure 13. Site plan, piling layout for the arrivals tunnel at Logan Airport.(18)

Approximately 576 piles were driven beneath the alignment of the tunnel structure. The piles, 

consisting of 41-cm-diameter PPC piles, were designed to support a concrete mat foundation in 

addition to a viaduct located above the tunnel. They were generally installed in a grid-like 

pattern, with a spacing of approximately 1.2 m by 1.8 m center to center (figure 13). 

The general subsurface conditions based on borings advanced in the area prior to excavation 

consist of approximately 3 to 6.1 m of cohesive and/or granular fill, overlying 1.5 to 3 m of 

organic silt and sand, overlying 12.2 to 42.7 m of soft marine clay, overlying 0.9 to 2.8 m of 

glacial silts and sands, underlain by bedrock.(6) Excavation was accomplished into the clay layer, 
20 

resulting in a clay layer thickness of about 6.1 m at the southeastern end of the structure to 

around 3.7 m at the northwestern end.(19)

The piles were designed for end bearing in the dense glacial silts and sands, and were preaugered 

to about the bottom of the marine clay layer to minimize heave and displacement of these soils. 

The preauger depths were approximately 30 to 70 percent of the final embedment depths of the 

piles. Preaugering was done using a 46-cm-diameter auger, which is the equivalent circular 

diameter of the 41-cm square pile. The piles were driven using an HPSI 2000 hydraulic hammer. 

Field Observations 

Pile heave was monitored during construction by field engineers. As described in the 

Massachusetts State building code and project specifications, piles identified with vertical 

displacement exceeding 1.3 cm required redriving. According to field records, 391 of the 576 

piles (68 percent) installed required redriving. Of those 391 piles, 337 piles (86 percent) were 

driven in one redrive event, 53 piles (14 percent) required a second redrive event, and 1 pile 

required a third redrive event. The impact on the construction schedule or costs was not 

identified. Despite the use of partial preaugering, a significant portion of the piles showed 

excessive heave and required substantial redrive efforts. Heave is attributed to the displacement 

of the underlying glacial soils that were not preaugered. 

Pile heave issues were not identified on the other CA/T contracts. Since partial preaugering was 

used on the majority of these contracts, the difference may be related to the spacing between 

piles. Table 5 summarizes the pile spacing used on the selected contracts. As shown in table 5, 

the pile spacing of 1.2 m used at the arrivals tunnel structure is significantly less than the spacing 

used for structures of comparable size. Therefore, it is anticipated that a pile spacing of greater 

than about 1.8 m may limit pile heave to within the 1.3-cm criterion. 
21 

Table 5. Summary of pile spacing from selected contracts.

Contract Structure Foundation Bent Spacing (m) Pile Spacing (m) 

Slab 2.7 2.7 Ramp ET C07D1 Pile cap 1.4 1.4 

Egress Ramps Pile cap 1.8 1.8 

Pile cap 1.8 1.2 

C07D2 Arrivals Tunnel Pile cap 1.8 1.2 

Pile cap 1.4 1.2 

South Abutment Pile cap 3.05 1.82.4 

C08A1 East Abutment Pile cap 1.1–2.7 1.4–2.6 

West Abutment Pile cap 1.1–2.1 1.4–2.7 

Utilities Pile cap 2.0–2.7 1.8 

Slab 3.7 5.6 

Pile cap 1.4 2.6 

Pile cap NA 1.4 Approach No. 1 

Pile cap NA 1.5 

Approach No. 2 Slab 4.57 3.1–4.6 

C09A4 

Approach No. 5 Slab 3.7–4.9 2.1–4.3 

NS-SN Slab 3.7 4.9 

C19B1 Ramp CT Slab 3.1 4.6 

Ramp LT Slab 2.9–3.2 2.4–3.1 

NA = not applicable or available 

Soil Heave 

Soil heave caused by pile driving was primarily responsible for the significant movement 

observed at a building adjacent to the construction of the east abutment and east approach to 

ramp ET at Logan Airport (contract C07D1). Shortly after the start of pile driving, settlement in 

excess of 2.5 cm was measured at the perimeter of the building and cracking was observed on the 

structure itself. These observations prompted the installation of additional geotechnical 

instrumentation, installation of wick drains to dissipate excess pore pressure generated during 

pile driving, and preaugering of the piles to reduce soil displacement. Despite these efforts, 

heave continued to a maximum vertical displacement of 8.8 cm. (See references 20, 21, 22, and 

23.) 

Pile Layout and Soil Conditions 

The location of the project in relation to the building is shown in figure 14. The portion of the 

east approach that is adjacent to the building consists of two major structures, including an 

abutment and a pile-supported slab. Both structures are supported by 41-cm-diameter PPC piles. 

The layout of the pile foundation system is also shown in figure 14. The piles for the slab are 

arranged in a grid-like pattern with a spacing of about 2.7 m center to center. A total of 353 piles 

support the structures. 
22 

Figure 14. Site plan showing locations of piles, building footprint, 

and geotechnical instrumentation. 

Prior to construction activities, five deformation monitoring points (DMPs) were installed along 

the front perimeter of the building closest to the work area. The DMPs consisted of 13-cm-long 

hex bolts fixed to the building. These points, designated DMP-101 through DMP-105, were 

monitored for vertical movement. The DMPs were monitored initially by the contractor and 

subsequently monitored by an independent consultant. 

The subsurface conditions based on borings advanced in the area consist of approximately 3 to 

4.6 m of fill, overlying 3 to 6.1 m of organic silt and sand, overlying 27.4 to 33.5 m of soft 

marine clay, overlying 6.1 to 12.2 m of glacial silt and sand, underlain by bedrock. The piles 

were designed as end bearing piles to be driven into the dense underlying glacial materials. The 

glacial soils were encountered at depths of approximately 39.6 to 45.7 m below the ground 

surface and bedrock was encountered at a depth of approximately 48.8 m. 

Field Observations (Phase I Pile Driving) 

Pile driving for the east approach was executed in two phases. The first phase began on April 5, 

1995, and concluded on June 10, 1995. The second phase began on July 13, 1995, and concluded 

on August 17, 1995. The piles were driven using a Delmag D46-32 single-acting diesel hammer. 

The extent of the first phase of pile driving is shown in figure 15. This first phase of work was 

performed no closer than 27.4 m from the building. The majority of the piles for the slab were 

installed from the west side of the site working toward the east during the periods of April 5 to 

April 23, and May 15 to June 2. The majority of the piles for the abutment were installed at the 

west end of the site during the period of April 23 to May 15. 
23 

Settlement data obtained by the contractor during the first phase of pile driving are shown in 

figure 15. On April 21, 1995, after approximately 2 weeks of pile driving on the west side of the 

site, initial heave displacements of 0.9 and 0.7 cm were measured in DMP-102 and DMP-103, 

respectively. Notable heave was observed at DMP-101 and DMP-104 on May 1, which 

registered displacements of 1.3 and 0.8 cm, respectively. An initial heave displacement of 0.4 cm 

was measured in DMP-105 on May 9. The heave increased steadily to maximum values as pile 

driving commenced toward the east side of the site. 

Figure 15. Settlement data obtained during first phase of pile driving. 

A summary of the maximum heave values attributed to the first phase of driving is given in 

table 6. The greatest amount of heave occurred in DMP-103, which was centrally located relative 

to the pile grid. On June 2, 1995, 1 week before completion of construction, the heave measured 

in DMPs 101 through 103 began to level off and subside. 

Table 6. Maximum building heave (in cm) observed during pile driving. 

Construction 

Phase DMP 101 DMP 102 DMP 103 DMP 104 DMP 105 

Phase I 2.5 3.5 4.3 3.8 1.6 

Phase II 3.6 4.8 5.3 3.7 1.3 

As a result of the excessive heave (greater than 2.5 cm) observed in the first phase of pile 

driving, mitigation measures were implemented for the second phase of work. This was critical 

considering that the second phase involved driving piles even closer to the building. The 

geotechnical consultant recommended three approaches to limiting heave based on schedule and 

cost constraints.(24) These included: (1) installation and monitoring of pore pressures in the clay 

during driving and adjusting mitigating measures as appropriate; (2) installation of wick drains 

between the Hilton and the work area to intercept and aid in the reduction of pore pressures 

beneath the Hilton that may be generated from pile driving; and (3) based on the performance of 

the wick drains, preauger phase II piles to limit soil displacement. 

-2

-1

0

1

2

3

4

5

6

1/25 2/8 2/22 3/8 3/22 4/5 4/19 5/3 5/17 5/31 6/14 6/28

Date

Vertical Heave (cm)

DMP 101

DMP 102

DMP 103

DMP 104

DMP 105

Phase I
24 

Field Observations (Phase II Pile Driving) 

Prior to the start of the second phase of pile driving, three double-nested vibrating wire 

piezometers (VWPZ) were installed to measure pore pressures. These piezometers were installed 

in close proximity to three of the existing deformation monitoring points (DMP-102 through 

DMP-104). Additional instrumentation was also installed following the start of the second phase 

of work, including a multipoint heave gauge (MPHG) to measure vertical movement with depth 

and an inclinometer to measure lateral movement. The locations of the additional geotechnical 

instrumentation are shown in figure 14. 

The second phase of pile driving began on July 13, 1995, and concluded on August 17, 1995. 

The extent of the work area is also shown in figure 14. Pile driving generally progressed from the 

west side of the site toward the east. The location of the second phase of work was no closer than 

15.2 m from the existing building. 

Shortly after the start of driving, 200 wick drains were installed from July 20 to July 28, 1995, 

around the western and northern perimeters of the work area. The drains were installed through 

the clay layer at a spacing of 1.2 m center to center. 

Settlement data for the second phase of work, shown in figure 16, demonstrate that heave began 

to increase at DMP-101 through DMP-104 approximately 1 week after the start of pile driving. 

Based on the review of initial settlement data, preaugering was implemented from August 4, 

1995, through the completion of construction. Preaugering was accomplished using a 41-cmdiameter auger to a depth of 26 m, which is approximately 50 to 60 percent of the pile’s final 

embedment depth. The auger diameter is 11 percent less than the 46-cm equivalent circular 

diameter for a 41-cm square pile. 

As shown in figure 16, heave continued to increase even after preaugering was initiated. Net 

heave values of 3.3 to 13.5 cm (table 6) were observed from the start of preaugering to the 

completion of pile driving, resulting in total heave values ranging from 2.6 to 8.8 cm. 
25 

Figure 16. Settlement data obtained during second phase of pile driving. 

Data from the multipoint heave gauge showed that the magnitude of the heave was relatively 

constant within the upper 30 m, as shown in figure 17. However, vertical displacement decreases 

dramatically below this depth to approximately zero at the bedrock depth of approximately 50 m. 

The maximum heave of approximately 5.1 cm at a depth of 3 m below the ground surface is also 

consistent with the maximum value of 5.3 cm recorded at DMP-103. 

Figure 17. Multipoint heave gauge data obtained during second phase 

of pile driving. 

0

1

2

3

4

5

6

7

8

9

10

4/5 4/19 5/3 5/17 5/31 6/14 6/28 7/12 7/26 8/9 8/23

Date

Vertical heave (cm)

DMP 101

DMP 102

DMP 103

DMP 104

DMP 105

Phase I Phase II

Install wick drains

Begin preaugering

0

10

20

30

40

50

60

-1 0 1 2 3 4 5 6 7

Vertical Heave (cm)

Initial Depth (m)

8-Aug

11-Aug

18-Aug

Fill 

Silt and Sand 

Marine Clay 

Glacial Soils 

Bedrock 
26 

The excess pore pressures recorded during the second phase of pile driving are presented in 

figure 17. The six gauges shown in figure 18 correspond to three pairs (55894–55895, 

55896–55897, and 55898–55899) located adjacent to DMP-102, DMP-103, and DMP-104, 

respectively. There was an increase in the excess pore pressures throughout the pile driving, with 

maximum values ranging from 0.6 to 12.8 m of head with an average of 5.9 m. The greatest head 

was measured in VWPZ-55896 at a location nearest DMP-103. These data suggest that the wick 

drains were not effective in dissipating all excess pore pressures generated during pile driving. 

Figure 18. Pore pressure data obtained during second phase of pile driving. 

The inclinometer data that were obtained adjacent to the building are shown in figure 19. These 

data showed increasing lateral movement in the direction of the building during pile driving. The 

maximum net lateral deformations were relatively constant with depth within the upper 30 m of 

the profile. A maximum deformation of approximately 6 cm was recorded at a depth of 

approximately 34 m. Similar to the vertical deformations, the lateral deformations decreased 

sharply below this depth to zero at the bedrock depth. These data suggest that the lateral 

deformations are of the same magnitude and behavior as the vertical deformations. 

-2

0

2

4

6

8

10

12

14

16

6/28 7/5 7/12 7/19 7/26 8/2 8/9 8/16 8/23

Date

Excess Pore Pressure Head (m)

55894

55895

55896

55897

55898

55899
27 

Figure 19. Inclinometer data obtained during second phase of pile driving. 

Summary 

Soil heave was recognized early on as a potential problem and following phase I driving in 

contract C07D1, several mitigation efforts were initiated. These included installing wick drains 

to promote rapid dissipation of excess pore pressures and preaugering piles through a portion of 

the soft clay layer to a depth of 26 m. Additional instrumentation was installed, including 

piezometers, MPHG, and an inclinometer. Despite these efforts, heave during phase II pile 

driving continued to increase to a maximum displacement of 8.8 cm. The piezometer data 

indicate that the wick drains were not effective in rapidly dissipating pore pressures generated 

during pile driving. The deformation data indicated that soil heave can still occur in piles that are 

preaugered over a portion of their embedded depth. 

0

10

20

30

40

50

60

01234567

Lateral Deformation (cm)

Depth (m)

12-Aug

16-Aug

18-Aug

Fill 

Silt and Sand 

Marine Clay 

Glacial Soils 

Bedrock 

29 

CHAPTER 4. DYNAMIC AND STATIC PILE LOAD TEST DATA 

This chapter presents the methodology and results of dynamic and static pile load test data for 

the selected contracts. At least two static load tests were performed per contract, and the results 

of 15 tests are presented herein. The Pile Driving Analyzer® (PDA) was also used on these piles 

for comparison, and analyses were performed periodically during production pile installation. 

Issues related to design loads and load test criteria are discussed, including factors of safety and 

load transfer requirements. A comparison is made between the results of the static load tests and 

the CAse Pile Wave Analysis Program (CAPWAP®) analyses. The CAPWAP data suggest that 

the quake values generally exceed the values typically recommended in wave equation analyses. 

A review of the literature is presented to evaluate the significance of this finding. High blow 

counts recorded during the end of driving also suggest that the majority of the estimated pile 

capacities from CAPWAP are conservative. 

LOAD TEST METHODS 

Dynamic Load Test Methods 

Approximately 160 dynamic pile load tests were performed to evaluate pile capacity, driving 

stresses, and hammer performance during the installation of test piles and production piles. The 

data presented in this report were obtained from project files. (See references 25, 26, 27, 28, 29, 

30, 31, 32, 33, 34.) 

The PDA was used to record, digitize, and processes the force and acceleration signals measured 

at the pile head. These signals were used to estimate static capacity using the Case Method, a 

simplified field procedure for estimating pile capacity, as well as the more rigorous CAPWAP. 

The dynamic load test results discussed in this report are primarily from the CAPWAP analyses. 

A description of the fundamentals of dynamic testing, including CAPWAP, is presented in 

Design and Construction of Driven Pile Foundations (Federal Highway Administration (FHWA) 

report no. FHWA-HI-97-013).(17) The dynamic testing was carried out in general accordance 

with project specifications section 940.62.C,(14) Dynamic Load Tests, and D4945-89 of the 

American Society for Testing and Materials (ASTM). D4945-89 is entitled “Standard Method 

for High Strain Testing of Piles.”(35) 

CAPWAP is an iterative curve-fitting technique where the pile response determined in a wave 

equation model is matched to the measured response of the actual pile for a single hammer blow. 

The pile model consists of a series of continuous segments and the total resistance of the 

embedded portion of the pile is represented by a series of springs (static resistance) and dashpots 

(dynamic resistance). Static resistance is formulated from an idealized elastoplastic soil model, 

where the quake parameter defines the displacement at which the soil changes from elastic to 

plastic behavior. The dynamic resistance is formulated using a viscous damping model that is a 

function of a damping parameter and the velocity. 

First, the forces and accelerations acting on the actual pile during initial impact are recorded with 

a strain gauge and accelerometer mounted at the pile head. The measured acceleration is used as 

input to the pile model along with reasonable estimates of soil resistance, quake, and damping 
30 

parameters. The force-time signal at the pile head is calculated using the model and is compared 

to the measured force-time signal. The soil-resistance distribution, quake, and damping 

parameters are subsequently modified until agreement is reached between the measured and 

calculated signals. An example of a comparison between a measured and calculated force signal 

from one of the test piles is shown in figure 20. Once an acceptable match is achieved, the 

solution yields an estimate of ultimate static capacity, the distribution of soil resistance along the 

pile, and the quake and damping parameters. 

0

500

1000

1500

2000

2500

0 20 40 60 80

Time (ms)

Force (kN)

Measured

Calculated

Figure 20. Example of CAPWAP signal matching, test pile 16A1-1.(33)

Static Load Test Methods 

Static load tests were performed during the test phase of each contract to verify the design 

assumptions and load-carrying capacity of the piles. Telltale rods installed at various depths 

within the piles were used to evaluate the load transfer behavior of the piles with regard to the 

surrounding soil and bearing stratum. The static tests were carried out in general accordance with 

project specifications section 940.62.B.4,(14) Short Duration Test, and the ASTM’s D1143-81, 

which is entitled “Standard Test Method for Piles Under Static Axial Compression Load.”(36) The 

static load test data presented in this report were obtained from the project files. (See references 

37 through 50.) 

Static loads were applied and maintained using a hydraulic jack and were measured with a load 

cell. A typical load test arrangement is shown in figure 21. Reaction to the jack load is provided 

by a steel frame that is attached to an array of steel H-piles located at least 3 m away from the 

test pile. Pile head deflections were measured relative to a fixed reference beam using dial 

gauges. Telltale measurements were made in reference to the pile head or the reference beam 

using dial gauges. Pile head and telltale deflection data were recorded for each loading 

increment. 
31 

Figure 21. Typical static load test arrangement 

showing instrumentation.(51) 

An excerpt from the loading procedures for short-duration load test section 940.62 is given 

below(14): 

a) Apply 25 percent of the allowable design load every one-half hour up to the 

greater of the following: [two alternatives are described; the most general is 

200 percent of the design load]. Longer time increments may be used, but each 

time increment should be the same. At 100 percent of the design load, unload 

to zero and hold for one-half hour; then reload to 100 percent and continue 25 

percent incremental loads. At 150 percent, unload to zero and hold for one-half 

hour; then reload to 150 percent and continue 25 percent incremental loads. In 

no case shall the load be changed if the rate of settlement is not decreasing 

with time. 

b) At the maximum applied load, maintain the load for a minimum of one hour 

and until the settlement (measured at the lowest point of the pile at which 

measurements are made) over a one-hour period is not greater than 0.254 mm 

(0.01 inch). 

c) Remove 25 percent of the load every 15 minutes until zero load is reached. 

Longer time increments may be used, but each shall be the same. 

d) Measure rebound at zero load for a minimum of one hour. 
32 

e) After 200 percent of the load has been applied and removed, and the test has 

shown that the pile has additional capacity, i.e., it has not reached ultimate 

capacity, continue testing as follows. Reload the test pile to the 200 percent 

design load level in increments of 50 percent of the allowable design load, 

allowing 20 minutes between increments. Then increase the load in increments 

of 10 percent until either the pile or the frame reach their allowable structural 

capacity, or the pile can no longer support the added load. If failure at 

maximum load does not occur, hold load for one hour. At maximum achieved 

load, remove the load in four equal decrements, allowing 15 minutes between 

decrements. 

The capacity of the test piles was selected as the greater capacity defined by two failure criteria. 

The first criteria establishes the allowable design capacity as “50 percent of the applied test load 

which results in a net settlement of the top of the pile of up to 1.3 cm, after rebound, for a 

minimum of one hour at zero load.” The second criterion uses Davisson’s criteria as described 

below. 

The Davisson offset limit load criterion was used on the project to define the ultimate capacity, 

or failure, of the test piles.(52) The ultimate load is interpreted as the point at which the 

displacement of the pile head meets a limit that is offset to the elastic compression line of the 

pile. For piles less than 61 cm in diameter, the limit is defined by the following linear 

relationship: 

S S ( D) f = e + 0.38 + 0.008 (1) 

where, 

Sf = Movement of pile top (cm). 

D = Pile diameter or width (cm). 

Se = Elastic compression of total pile length (cm). 

The elastic compression in this case refers to the pile deflection that would occur if 100 percent 

of the applied load was transferred to the toe of the pile (i.e., zero shaft friction), and is given by 

the following equation: 

AE

QL Se =

 (2) 

where, 

Q = Applied load. 

L = Total length of pile. 

A = Cross-sectional area of the pile. 

E = Modulus of elasticity of the pile. 

The average load in the pile at the midpoint between two telltale locations was estimated from 

the elastic shortening of the pile using the following equation: 
33 

L

D D

Qavg AE

Δ

− = 1 2

 (3) 

where, 

A = Area of pile. 

E = Modulus of elasticity of the pile. 

D1 = Deflection at upper telltale location. 

D2 = Deflection at lower telltale location. 

ΔL = Distance between the upper and lower telltale locations. 

Both equations 2 and 3 require the modulus of elasticity of the pile. The specifications require 

that the elastic modulus be determined via compression tests performed on representative 

concrete samples (ASTM C 469-87a). However, this method is not really applicable to the 

concrete-filled steel pipe piles. It was common practice on the CA/T project to use the upper 

telltale and pile head deflections to calculate the modulus of the pile using equation 3. This 

approach was justified by assuming that any preaugering that was performed prior to pile 

installation would reduce the shaft friction, especially near the pile head. In some cases, the 

elastic modulus of the PPC piles was determined from a combination of telltale and compression 

test data using engineering judgment. 

LOAD TEST RESULTS 

More than 160 dynamic tests were performed on the selected contracts to evaluate pile capacity 

during both the testing and production phases. Of these 160 tests, the results of 28 tests are 

presented in this report because they correspond to static load tests on 15 piles. Information 

about each pile tested is shown in table 7, and pile driving information is presented in table 8. 
34 

Table 7. Summary of pile and preauger information.

Test Pile 

Name Contract Pile Type Preauger Depth (m) 

Preauger 

Diameter (cm) 

ET2-C2 C07D1 41-cm PPC 0 NA1

ET4-3B C07D1 41-cm PPC 0 NA 

375 C07D2 41-cm PPC 9.1 45.7 

923 C07D2 41-cm PPC 24.1 45.7 

I90 EB SA C08A1 41-cm PPC NI2 40.6 

14 C08A1 41-cm PPC 27.4 40.6 

12A1-1 C09A4 31-cm PPC 30.5 45.7 

12A2-1 C09A4 31-cm PPC 32.0 45.7 

16A1-1 C09A4 41-cm PPC 30.5 45.7 

I2 C09A4 41-cm PPC 30.5 40.6 

3 C09A4 41-cm pipe 24.4 40.6 

7 C09A4 41-cm pipe 24.4 40.6 

IPE C19B1 32-cm pipe 7.6 30.5 

IPW C19B1 32-cm pipe 12.2 30.5 

NS-SN C19B1 41-cm PPC 8.2 40.6 

Notes: 

1. NA = Not applicable. 

2. NI = Data not identified. 

Table 8. Summary of pile driving information. 

Test Pile 

Name Test Type1 Hammer 

Type2

Embedment 

Depth (m) 

Minimum 

Transferred 

Energy (kN-m)

Recorded 

Penetration 

Resistance 

(blows/2.5 cm) 

Permanent 

Set (cm) 

ET2-C2 EOD I 47.5 NI3 7,7,7 0.36 

 34DR – – 58.0 11 0.23 

ET4-3B EOD II 41.1 NI 8,7,10 0.25 

 NI – – 50.8 14 0.18 

375 EOD II 16.8 50.2 12,13,39 0.08 

 7DR – – 54.2 > 12 < 0.20 

923 EOD II 32.9 46.1 7,7,7 0.36 

 7DR – – 51.5 > 8 0.33 

I90 EB SA EOD III 46.6 25.8 12,10,10 0.25 

 1DR – – 25.8 13 0.20 

14 EOD III 45.4 25.8 10,10,16 0.15 

 1DR – – 23.1 21 0.13 

12A1-1 EOD III 41.8 20.7 4,4,5 0.51 

 1DR – – 28.6 > 7 > 0.36 

12A2-1 EOD III 38.7 15.3 3,4,4 0.64 

 1DR – – 18.6 8 0.33 

16A1-1 EOD III 43.3 24.4 6,7,7 0.36 

 3DR – – 17.1 11 0.23 

I2 EOD III 37.2 27.1 4,4,4 0.64 

 1DR – – 19.0 5 0.51 
35 

Table 8. Summary of pile driving information (continued). 

Test Pile 

Name Test Type1 Hammer 

Type2

Embedment 

Depth (m) 

Minimum 

Transferred 

Energy (kN-m)

Recorded 

Penetration 

Resistance 

(blows/2.5 cm) 

Permanent 

Set (cm) 

3 EOD III 39.6 57.1 11,12,14 0.18 

 1DR – – 49.9 30 0.08 

7 EOD III 38.1 49.8 11,11,11 0.23 

 3DR – – 50.2 > 16 < 0.15 

IPE EOD V 19.5 39.6 5,5,5 0.51 

 1DR – – 53.0 7 0.36 

IPW EOD VI 22.6 43.3 5,5,5 0.51 

 1DR – – 59.7 8 0.33 

NS-SN EOD IV 13.4 27.1 8,15,16 0.15 

 7DR – – 24.4 26 0.10 

Notes: 

1. EOD = End of initial driving, #DR = # days before restrike. 

2. Hammer types: I = Delmag D 46-32, II = HPSI 2000, III = ICE 1070, IV = HPSI 1000, V = Delmag D 

19-42, VI = Delamag D 30-32. 

3. NI = Data not identified. 

Dynamic Results and Interpretation 

Dynamic tests were performed both at the end of initial driving of the pile (EOD) and at the 

beginning of restrike (BOR), typically 1 to 7 days (1DR, 7DR, etc.) after installation. In most 

cases, the dynamic tests were performed before the static load tests. Test piles ET2-C2 and ET4-

3B, however, were dynamically tested during a restrike after a static load test was performed. 

The ultimate capacities of the 15 test piles as determined by CAPWAP analysis are summarized 

in table 9. The table lists when the test was performed, as well as the predicted shaft and toe 

resistance. 

Table 9. Summary of CAPWAP capacity data.

Ultimate Capacity2

 (kN) 

Test Pile 

Name Test Type1

Recorded 

Penetration 

Resistance 

(blows/2.5 cm) 

Shaft Toe Total 

ET2-C2 EOD 7,7,7 NI3 NI NI 

 34DR 11 (2,028) (1,219) (3,247) 

ET4-3B EOD 8,7,10 NI NI NI 

 NI 14 (1,744) (1,975) (3,719) 

375 EOD 12,13,39 (890) (3,336) (4,226) 

 7DR > 12 (1,245) (3,514) (4,759) 

923 EOD 7,7,7 667 1,904 2,571 

 7DR > 8 (1,664) (1,708) (3,372) 

I90 EB SA EOD 12,10,10 934 712 1,646 

 1DR 13 (1,156) (1,112) (2,268) 

14 EOD 10,10,16 (449) (2,237) (2,687) 

 1DR 21 (894) (1,926) (2,820) 

12A1-1 EOD 4,4,5 685 979 1,664 
36 

Table 9. Summary of CAPWAP capacity data (continued). 

Test Pile 

Name Test Type1

Recorded 

Penetration 

Resistance 

(blows/2.5 cm) 

Ultimate Capacity2 (kN) 

 1DR > 7 (1,103) (743) (1,846) 

12A2-1 EOD 3,4,4 316 845 1,161 

 1DR 8 1,023 431 1,454 

16A1-1 EOD 6,7,7 956 1,063 2,015 

 3DR 11 (983) (876) (1,859) 

I2 EOD 4,4,4 400 1,130 1,530 

 1DR 5 1,526 489 2,015 

3 EOD 11,12,14 (983) (2,086) (3,069) 

 1DR 30 (1,228) (1,690) (2,918) 

7 EOD 11,11,11 (80) (2,740) (2,820) 

 3DR > 16 (983) (1,984) (2,962) 

IPE EOD 5,5,5 489 1,334 1,824 

 1DR 7 645 1,535 2,180 

IPW EOD 5,5,5 778 1,223 2,002 

 1DR 8 1,290 1,468 2,758 

NS-SN EOD 8,15,16 (583) (1,806) (2,389) 

 7DR 26 (858) (1,935) (2,793) 

Notes: 

1. EOD = End of initial driving, #DR = # days before restrike. 

2. Values shown in parentheses denote conservative values. 

3. NI = Data not identified. 

Many of the capacities are listed in parentheses, which indicates that the values are most likely 

conservative (i.e., the true ultimate capacity is larger). It is recognized in the literature that 

dynamic capacities can be underestimated if the hammer energy is insufficient to completely 

mobilize the soil resistance.(53) Specifically, research has shown that blow counts in excess of 10 

blows per 2.5 cm may not cause enough displacement to fully mobilize the soil resistance.(53,54)

As shown in table 8, the majority of the piles during restrike exceeded 10 blows per 2.5 cm and 

are thus likely to be lower than the true ultimate capacity of the piles. 

The conservativeness of the CAPWAP capacities in certain piles can be illustrated by comparing 

the load versus displacement curve at the toe evaluated with CAPWAP to that obtained in a static 

load test. The toe load-displacement curves from test pile 16A1-1 are shown in figure 22. Blow 

counts of seven blows per 2.5 cm were recorded for this pile during initial driving. The static 

load test data shown in figure 22 were extrapolated from the telltale data. As shown in figure 22, 

the maximum resistance mobilized by the pile toe from CAPWAP is approximately 1060 kN. At 

least 1670 kN were mobilized in the static load test; however, the ultimate value is actually 

higher since failure was not reached. 
37 

Figure 22. Load-displacement curves for pile toe, test pile 16A1-1. 

Soil quake and damping parameters obtained from the CAPWAP analyses are summarized in 

table 10. It is often assumed that the quake values are approximately 0.25 cm in typical wave 

equation analyses. The toe quake values in this study range from 0.25 to 1.19, with an average of 

1.6 cm. Large toe quake values on the order of up to 2.5 cm have been observed in the 

literature.(55,56) However, the quake values in this study appear to be within typical values.(57)

0.0

0.5

1.0

1.5

2.0

2.5

-1000 -500 0 500 1000 1500 2000

Load at Pile Toe (kN)

Displacement (cm)

Static load test

CAPWAP (EOD)
38 

Table 10. Summary of CAPWAP soil parameters.

Test Pile Quake (cm) Damping (s/m) 

Name Test Type1

Shaft Toe Shaft Toe 

ET2-C2 EOD – – – – 

 34DR 0.43 0.84 0.72 0.23 

ET4-3B EOD – – – – 

 - 0.56 0.36 0.89 0.82 

375 EOD 0.64 1.19 0.33 0.07 

 7DR 0.51 0.86 0.23 0.20 

923 EOD 0.38 1.14 0.72 0.43 

 7DR 0.23 0.81 0.46 0.43 

I90 EB SA EOD 0.13 0.89 0.16 0.56 

 1DR 0.38 0.56 0.69 0.69 

14 EOD 0.25 0.76 0.39 0.43 

 1DR 0.25 0.41 0.59 0.43 

12A1-1 EOD – – – – 

 1DR 0.38 0.56 0.75 0.16 

12A2-1 EOD – – – – 

 1DR 0.25 0.51 0.49 0.33 

16A1-1 EOD – – – – 

 3DR 0.25 0.10 1.41 1.15 

I2 EOD 0.25 0.51 0.75 0.26 

 1DR 0.13 0.25 0.46 0.10 

3 EOD 0.48 0.64 0.13 0.10 

 1DR 0.15 0.56 0.33 0.10 

7 EOD 0.23 0.64 0.46 0.10 

 3DR 0.25 0.36 0.52 0.10 

IPE EOD 0.25 0.69 0.62 0.23 

 1DR 0.38 0.89 0.59 0.23 

IPW EOD 0.38 0.64 0.43 0.23 

 1DR 0.25 0.36 0.59 0.20 

NS-SN EOD 0.30 0.91 0.52 0.33 

 7DR 0.13 0.46 0.72 0.49 

Notes: 

1. EOD = End of initial driving, #DR = # days before restrike. 

2. s/m = seconds/meter. 

Comparison of CAPWAP Data 

A comparison between the EOD and BOR CAPWAP capacities is shown in figure 23. The line 

on the figure indicates where the EOD and BOR capacities are equal. Data points that are plotted 

to the left of the line show an increase in the capacity over time, whereas data that fall below the 

line show a decrease in capacity. In the four piles (12A2-1, I2, IPE, and IPW) where the soil 

resistance was believed to be fully mobilized for both the EOD and BOR, the data show an 

increase of 20 to 38 percent occurring over 1 day. The overall increase in capacity is attributed to 

an increase in the shaft resistance. 
39 

Figure 23. CAPWAP capacities at end of initial driving (EOD) and 

beginning of restrike (BOR). 

Static Load Test Data 

Static load tests were performed on 15 piles approximately 1 to 12 weeks after their installation. 

The test results are summarized in table 11. In general, two types of load deflection behavior 

were observed in the static load tests (figures 24 through 27). 

Table 11. Summary of static load test data. 

Test Pile Name Time After Pile 

Installation (days)

Maximum 

Applied Load 

(kN) 

Maximum Pile Head 

Displacement (cm) 

ET2-C2 13 3,122 1.7 

ET4-3B 20 3,558 2.4 

375 15 3,447 1.6 

923 33 3,447 2.4 

I90 EB SA 23 3,781 1.6 

14 6 3,105 2.2 

12A1-1 30 1,512 1.4 

12A2-1 24 1,014 0.5 

16A1-1 17 3,612 2.6 

I2 6 3,558 1.7 

3 9 3,959 2.4 

7 10 3,167 2.0 

IPE 84 2,384 1.3 

IPW 10 2,891 4.1 

NS-SN 30 2,535 1.3 

Test pile 12A1-1 (figure 24) represents a condition where the axial deflection of the pile is less 

than the theoretical elastic compression (assuming zero shaft friction). This pile was loaded to 

0

1000

2000

3000

4000

5000

6000

0 1000 2000 3000 4000 5000 6000

Capacity at EOD (kN)

Capacity at BOR (kN)

Fully Mobilized

BOR Lower Bound

EOD and BOR Lower Bound
40 

1,557 kN in five steps and at no point during the loading did the deflection exceed the estimated 

elastic compression of the pile. This behavior is attributed to shaft friction, which reduces the 

compressive forces in the pile and limits the settlement. The significant contribution of shaft 

friction is also apparent in the load distribution curve shown in figure 25, which shows the load 

in the pile decreasing with depth. This behavior is typical of test piles ET2-C2, ET4-3B, I90-EBSA, 12A1-1, 12A2-1, I2, and 3. 

Figure 24. Deflection of pile head during 

static load testing of pile 12A1-1. 

Figure 25. Distribution of load in 

pile 12A1-1. 

Test pile 14 (figure 26) represents a condition where the axial deflection is approximately equal 

to the theoretical elastic compression. This suggests that more of the applied loads are being 

distributed to the toe of the pile with less relative contribution of shaft friction. This is apparent 

in figure 27, which shows negligible changes in the load within the pile with depth. This 

behavior is typical of test piles 375, 923, 14, 16A1-1, 7, IPE, and IPW. 

Figure 26. Deflection of pile head during 

static load testing of pile 14. 

Figure 27. Distribution of load in pile 14. 

0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

0 500 1000 1500 2000

Load (kN)

Deflection (cm)

Test Data

Elastic Compression

Davisson's Line

0

5

10

15

20

25

30

35

40

45

0 500 1000 1500 2000

Load in Pile (kN)

Depth Below Ground Surface (m)

0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

0 1000 2000 3000 4000

Load (kN)

Deflection (cm)

Test Data

Elastic Compression

Davission's Line

0

5

10

15

20

25

30

35

40

0 1000 2000 3000 4000

Load in Pile (kN)

Depth Below Ground Surface (m)
41 

Of the 15 static load tests, only one test pile (IPW) was loaded to failure according to Davisson’s 

criteria. These data are shown in figures 28 and 29. This pile showed a significant increase in the 

deflection at approximately 2,580 kN, subsequently crossing the Davisson’s line at 

approximately 2,670 kN at a displacement of around 2.5 cm. The telltale data obtained near the 

toe of the pile indicated that the pile failed in plunging. 

Figure 28. Deflection of pile head during 

static load testing of pile IPW. 

Figure 29. Distribution of load in pile IPW. 

All test piles achieved the required ultimate capacities in the static load tests. The required 

ultimate capacities were determined by multiplying the allowable design capacity by a factor of 

safety of at least 2.0, as specified in the project specifications. A slightly higher factor of safety 

of 2.25 was used in contract C19B1. Three of the 15 static tests did not demonstrate that 

100 percent of the design load was transferred to the bearing soils. Two of the piles (12A1-1 and 

12A2-1) could not transfer the load to the bearing soils because of the high skin friction 

(figures 24 and 25). Test pile I2 could not demonstrate load transfer because the bottom telltale 

was not functioning. 

Comparison of Dynamic and Static Load Test Data 

The capacities determined by CAPWAP and from the static load tests are summarized in 

table 12, along with the required ultimate capacities. Of the 15 test piles, only one pile (IPW) 

was loaded to failure in a static load test. Likewise, only four BOR CAPWAP analyses and eight 

EOD CAPWAP analyses mobilized the full soil resistance. This means that the true ultimate 

capacity of the majority of the piles tested was not reached, and this makes a comparison of static 

load test and CAPWAP results difficult. 

Test pile IPW was brought to failure in the static load test. Coincidentally, it is anticipated that 

the CAPWAP capacities for this pile also represent the fully mobilized soil resistance because of 

the relatively low blow counts (i.e., < 10) observed during driving. Based on a comparison of all 

data for pile IPW, its capacity increased by approximately 35 percent soon after installation, 

yielding a factor of safety of approximately 3.0. Note that this pile was preaugered to a depth of 

approximately half of the embedment depth. The capacity of 2,669 kN determined in the static 

0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

4.0

4.5

0 1000 2000 3000 4000

Load (kN)

Deflection (cm)

Test Data

Elastic Compression

Davisson's Line

0

2

4

6

8

10

12

14

16

18

20

0 1000 2000 3000 4000

Load in Pile (kN)

Depth Below Ground Surface (m)
42 

load test is slightly less than the restrike capacity of 2,758 kN. However, this difference is partly 

attributed to modifications that were made to the pile after the dynamic testing, but prior to static 

testing. These modifications included removal of 0.6 m of overburden at the pile location and 

filling of the steel pipe pile with concrete, both of which would decrease the capacity of the pile 

measured in the static load test. 

Table 12. Summary of dynamic and static load test data. 

CAPWAP Ultimate 

Capacity1 Test Pile (kN) 

Name 

Required 

Allowable 

Capacity 

(kN) 

Required 

Minimum 

Factor of 

Safety 

Required 

Ultimate 

Capacity 

(kN) EOD BOR 

Ultimate Capacity 

From Static Load 

Test (kN) 

ET2-C2 1,379 2.00 2,758 NI2 (3,247) (3,122) 

ET4-3B 1,379 2.00 2,758 NI (3,719) (3,558) 

375 1,379 2.00 2,758 (4,226) (4,759) (3,447) 

923 1,379 2.00 2,758 2,571 (3,372) (3,447) 

I90 EB SA 1,379 2.00 2,758 1,646 (2,268) (3,781) 

14 1,379 2.00 2,758 (2,687) (2,820) (3,105) 

12A1-1 756 2.00 1,512 1,664 (1,846) (1,512) 

12A2-1 507 2.00 1,014 1,161 1,454 (1,014) 

16A1-1 1,245 2.00 2,491 2,015 (1,859) (3,612) 

I2 1,245 2.00 2,491 1,530 2,015 (3,558) 

3 1,583 2.00 3,167 (3,069) (2,918) (3,959) 

7 1,583 2.00 3,167 (2,820) (2,962) (3,167) 

IPE 890 2.25 2,002 1,824 2,180 (2,384) 

IPW 890 2.25 2,002 2,002 2,758 2,669 

NS-SN 1,112 2.25 2,504 (2,389) (2,793) (2,535) 

Notes: 

1. Capacities shown in parenthesis denote values that are conservative (dynamic load tests) or where failure was 

not achieved (static load tests). 

2. NI = Data not identified. 
43 

CHAPTER 5. COST DATA OF DRIVEN PILES 

This chapter presents a summary of the costs associated with pile driving operations on the CA/T 

project. The costs presented in this report were obtained directly from the contractor and 

represent the contractor’s bid estimates identified in the individual contracts. The primary 

purpose of the cost data is to document the approximate cost of pile driving on the CA/T project; 

however, the data may also be useful to design engineers for planning purposes. 

The contractor’s bid costs for pile driving are summarized in table 13 by pile type. Unless noted, 

the costs in table 13 do not include costs for preaugering or costs associated with the 

mobilization or demobilization of the contractor’s equipment. Steel pipe piles had the highest 

unit costs, ranging from $213 per meter for the 81.3-cm pile to $819 for the 154.9-cm pile. Unit 

costs for the PPC piles were lower, ranging from $72 to $197 per meter for the 30-cm PPC piles 

and $95 to $262 per meter for the 41-cm piles. As one would expect, the unit costs tended to 

decrease with the increasing size of the contract. The contractor’s bid costs for preaugering are 

summarized in table 14. Preaugering was not performed in contract C07D1, and preaugering 

costs were not identified in the contract C07D2 bid. As shown in table 14, the additional cost of 

preaugering ranged from $33 to $49 per meter. 

Table 13. Summary of contractor’s bid costs for pile driving.

Contract Pile Type 

Estimated 

Length of Pile 

Installed (m) 

Estimated Cost of 

Installation 

Cost per meter 

of Pile1

C19B1 32-cm concrete-filled steel pipe 550 $1,183,650 $213.19 

C09A4 41-cm concrete-filled steel pipe 5,578 $1,647,000 $295.27 2 

C19B1 61-cm concrete-filled steel pipe 296 $242,500 $819.26 

C08A1 30-cm square PPC with stinger 792 $156,000 $196.97 

C19B1 30-cm square PPC with stinger 2,177 $285,720 $131.24 

C09A4 30-cm square PPC 3,658 $600,000 $164.02 2 

C07D2 30-cm square PPC with stinger 3,981 $289,510 $72.72 

C07D1 30-cm square PPC with stinger 7,955 $652,500 $82.02 

C19B1 41-cm square PPC with stinger 6,279 $824,000 $131.23 

C08A1 41-cm square PPC with stinger 8,406 $2,206,400 $262.48 

C09A4 41-cm square PPC with stinger 14,326 $3,290,000 $229.65 2 

C07D2 41-cm square PPC with stinger 19,879 $2,396,800 $120.57 

C07D1 41-cm square PPC with stinger 32,918 $3,132,000 $95.15 

Notes: 

1. Unit costs include the costs of materials and labor for pile driving only. Preaugering is not included unless 

otherwise noted. See table 14 for preaugering unit costs. Mobilization and/or demobilization costs are not 

included. 

2. Unit costs include the costs of preaugering.

Table 14. Summary of contractor’s bid costs for preaugering.

Contract 

Preaugering 

Depth Range 

(m) 

Estimated Total 

Preaugering Depth 

(m) 

Estimated 

Cost of 

Preaugering 

Estimated 

Cost per meter 

C08A1 0 to 30.5 2,134 $70,000 $32.80 

C19B1 0 to 30.5 3,712 $182,655 $49.21 

45 

CHAPTER 6. LESSONS LEARNED 

This chapter presents a summary of the lessons learned from driven piles on the CA/T project. 

The conclusions presented below are based on the evaluation of field records, project 

specifications, and pile load test data compiled from the project files. Five contracts were 

evaluated, including three located in East Boston/Logan Airport, one located in downtown 

Boston, and one located in Charlestown. Significant findings are summarized below: 

• The dominant pile type used on the CA/T project was a 41-cm square PPC pile. Based on the 

contractor’s bid estimates, the PPC piles were also the most economical pile type. 

• Pile heave in excess of the 1.3-cm criteria was identified on one cut-and-cover tunnel 

structure requiring 445 restrike events for the 576 piles used in the structure. The heave 

occurred even though preaugering of the marine clay layer was performed. Pile heave issues 

were not identified at other structures where the pile spacing was greater than about 1.8 m. 

• Installation of displacement piles in contract C07D1 caused excessive movement of an 

adjacent structure. Despite the use of wick drains and partial preaugering, vertical 

displacement continued up to 8.8 cm. The wick drains were not effective in rapidly 

dissipating excess pore pressures from pile driving. 

• The heave issues observed in contract C07D1 prompted the use of preaugering on subsequent 

contracts. Preaugering was performed over a portion, generally 30 to 70 percent, of the final 

pile embedment depth. 

• Pile capacities evaluated using dynamic methods were conservative in hard driving 

conditions (i.e., penetration resistance greater than 10 blows per 2.5 cm) where the soil 

resistance may not be fully mobilized. 

• Quake values from CAPWAP analyses ranged from 0.25 to 1.19 cm, with an average value 

of 0.64 cm. These values are higher than the values typically used in wave equation analyses; 

however, they are within the range of published values. 

• Comparison of CAPWAP data evaluated at the end of initial driving and during restrike 

shows that the capacity of the piles increased over time by at least 20 percent from an 

increase in shaft resistance. 

• Only 1 out of 15 piles tested in a static load test was brought to failure according to 

Davisson’s criteria, because the specifications did not specifically require that the pile be 

brought to failure. 

• Three of the 15 piles did not successfully demonstrate that 100 percent of the design load was 

transferred to the bearing soils. Two piles did not meet the criteria because of high shaft 

friction, and the third did not meet the criteria because of a malfunctioning bottom telltale.

• Comparison of dynamic and static load test capacities was only possible on one pile (IPW), 

which reached the Davisson’s failure criteria in the static load test. CAPWAP and static 

capacities were in good agreement for this pile. 

47 

REFERENCES 

1. Massachusetts Turnpike Authority (2000), Project Summary, http://www.bigdig.com/thtml/ 

summary.htm. 

2. Massachusetts Turnpike Authority (2000), Project Contract Lists, http://www.bigdig.com/ 

thtml/contlist.htm. 

3. Massachusetts Turnpike Authority (2000), Maps and Plans, http://www.bigdig.com/thtml/ 

maps01.htm. 

4. GZA GeoEnvironmental, Inc. (1991), Central Artery (I-93)/Tunnel (I-90) Project, 

Geotechnical Data Report, South Bay Interchange, Design Sections D009B/D009C, Boston, 

MA. 

5. GZA GeoEnvironmental, Inc. (1992), Central Artery (I-93)/Tunnel (I-90) Project, 

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49 

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