Synthesis of Geosynthetic Reinforced

Soil (GRS) Design Topics

PUBLICATION NO. FHWA-HRT-14-094 FEBRUARY 2015

Research, Development, and Technology

Turner-Fairbank Highway Research Center

6300 Georgetown Pike

McLean, VA 22101-2296
FOREWORD

Geosynthetic reinforced soil (GRS) for load-bearing applications has been identified by the 

Federal Highway Administration (FHWA) as a proven, market-ready technology and is being 

actively promoted through the agency’s Every Day Counts initiative. With the publication of the 

FHWA interim design guidance for GRS abutments and integrated bridge systems, presented in 

GRS Integrated Bridge System (IBS) Interim Implementation Guide (FHWA-HRT-11-026), 

FHWA took the first steps to differentiate the design of GRS and conventional, larger-spaced 

geosynthetic mechanically stabilized earth (GMSE). This was based on considerable research 

and the change in behavior observed when reinforcement is closely spaced (i.e., less than about 

12 inches (30.48 cm)). Because of its similarities to GMSE walls, however, there are some 

misconceptions about the design and behavior of closely spaced GRS systems. This synthesis 

report outlines the background and research for some of the most pertinent changes related to the 

design differences between GRS and GMSE: embedment length, reinforcement pullout, 

eccentricity, lateral earth pressures and connection strength, the W equation for required 

reinforcement strength, and long-term reduction factors for geosynthetic materials. This report 

can be used by transportation agencies to support design changes related to closely spaced GRS 

systems.

Jorge E. Pagán-Ortiz

Director, Office of Infrastructure

Research and Development

Notice

This document is disseminated under the sponsorship of the U.S. Department of Transportation

in the interest of information exchange. The U.S. Government assumes no liability for the use of

the information contained in this document.

The U.S. Government does not endorse products or manufacturers. Trademarks or

manufacturers’ names appear in this report only because they are considered essential to the

objective of the document.

Quality Assurance Statement

The Federal Highway Administration (FHWA) provides high-quality information to serve

Government, industry, and the public in a manner that promotes public understanding. Standards

and policies are used to ensure and maximize the quality, objectivity, utility, and integrity of its

information. FHWA periodically reviews quality issues and adjusts its programs and processes to 

ensure continuous quality improvement.
TECHNICAL REPORT DOCUMENTATION PAGE

1. Report No.

FHWA-HRT-14-094

2. Government Accession No. 3. Recipient’s Catalog No.

4. Title and Subtitle

Synthesis of Geosynthetic Reinforced Soil Design Topics

5. Report Date

February 2015

6. Performing Organization Code: 

HRDI-40

7. Author(s)

Jonathan T.H. Wu and Phillip S.K. Ooi 

8. Performing Organization Report No. 

9. Performing Organization Name and Address

Department of Civil and Environmental Engineering, University 

of Hawaii at Manoa, 2540 Dole Street, Honolulu, HI 96822

Civil Engineering Department, University of Colorado—Denver, 

Denver, CO 80217

10. Work Unit No.

11. Contract or Grant No.

N/A

12. Sponsoring Agency Name and Address

Office of Infrastructure R&D

FHWA Research, Development and Technology

6300 Georgetown Pike

McLean, VA 22101

13. Type of Report and Period Covered

Technical

14. Sponsoring Agency Code

15. Supplementary Notes

The FHWA Contracting Officer’s Technical Representative was Mike Adams, HRDI-40. FHWA technical 

reviewers were Jennifer Nicks, Daniel Alzamora, and Khalid Mohamed.

16. Abstract

This report synthesizes six topics related to geosynthetic reinforced soil (GRS) design: embedment length, 

pullout check, eccentricity, lateral pressures, the W-equation for GRS capacity and required reinforcement 

strength, and geosynthetic reduction factors. The synopsis for each topic includes a summary of the relevant 

research and a review of pertinent issues. The intent is not to provide recommendations but to explain the 

methodology behind the Federal Highway Administration’s (FHWA) composite design for GRS, which is 

different than that for geosynthetic mechanically stabilized earth. The synopses support the FHWA’s GRS 

Integrated Bridge System (IBS) Interim Implementation Guide, (FHWA-HRT-11-026). 

17. Key Words

Geosynthetic reinforced soil, pullout, embedment length, 

eccentricity, overturning, connection strength, required 

reinforcement strength, abutment, integrated bridge system, 

geotextile

18. Distribution Statement

No restrictions. This document is available 

to the public through the National Technical 

Information Service, Springfield, VA 22161.

http://www.ntis.gov

19. Security Classif. (of this report)

Unclassified

20. Security Classif. (of this page)

Unclassified

21. No. of Pages

86

22. Price

Form DOT F 1700.7 (8-72) Reproduction of completed page authorized
SI* (MODERN METRIC) CONVERSION FACTORS 

APPROXIMATE CONVERSIONS TO SI UNITS

Symbol When You Know Multiply By To Find Symbol 

LENGTH

in inches 25.4 millimeters mm 

ft feet 0.305 meters m 

yd yards 0.914 meters m 

mi miles 1.61 kilometers km 

AREA 

in2 square inches 645.2 square millimeters mm2

ft2 square feet 0.093 square meters m2

yd2 square yard 0.836 square meters m2

ac acres 0.405 hectares ha

mi2 square miles 2.59 square kilometers km2

VOLUME 

fl oz fluid ounces 29.57 milliliters mL 

gal gallons 3.785 liters L 

ft3 cubic feet 0.028 cubic meters m3 

yd3 cubic yards 0.765 cubic meters m3 

NOTE: volumes greater than 1000 L shall be shown in m3

MASS oz ounces 28.35 grams g

lb pounds 0.454 kilograms kg

T short tons (2000 lb) 0.907 megagrams (or "metric ton") Mg (or "t") 

TEMPERATURE (exact degrees) o

F Fahrenheit 5 (F-32)/9 Celsius o

C 

or (F-32)/1.8 

ILLUMINATION 

fc foot-candles 10.76 lux lx

fl foot-Lamberts 3.426 candela/m2 cd/m2

FORCE and PRESSURE or STRESS 

lbf poundforce 4.45 newtons N 

lbf/in2 poundforce per square inch 6.89 kilopascals kPa 

APPROXIMATE CONVERSIONS FROM SI UNITS 

Symbol When You Know Multiply By To Find Symbol 

LENGTH mm millimeters 0.039 inches in 

m meters 3.28 feet ft 

m meters 1.09 yards yd 

km kilometers 0.621 miles mi 

AREA mm2 square millimeters 0.0016 square inches in2 

m2 square meters 10.764 square feet ft2 

m2 square meters 1.195 square yards yd2 

ha hectares 2.47 acres ac

km2 square kilometers 0.386 square miles mi2 

VOLUME 

mL milliliters 0.034 fluid ounces fl oz 

L liters 0.264 gallons gal 

m3 cubic meters 35.314 cubic feet ft3 

m3 cubic meters 1.307 cubic yards yd3 

MASS 

g grams 0.035 ounces oz

kg kilograms 2.202 pounds lb

Mg (or "t") megagrams (or "metric ton") 1.103 short tons (2000 lb) T 

TEMPERATURE (exact degrees) o

C Celsius 1.8C+32 Fahrenheit oF 

ILLUMINATION 

lx lux 0.0929 foot-candles fc 

cd/m2 candela/m2 0.2919 foot-Lamberts fl

FORCE and PRESSURE or STRESS 

N newtons 0.225 poundforce lbf 

kPa kilopascals 0.145 poundforce per square inch lbf/in2

*SI is the symbol for the International System of Units. Appropriate rounding should be made to comply with Section 4 of ASTM E380. 

(Revised March 2003)

ii
TABLE OF CONTENTS

CHAPTER 1. INTRODUCTION................................................................................................ 1

CHAPTER 2. EMBEDMENT LENGTH ................................................................................... 3

RESEARCH AND CASE HISTORIES.................................................................................. 3

Reinforced Soil Walls With Uniform Reinforcement Lengths............................................... 3

Reinforced Soil Walls With Nonuniform Reinforcement Lengths......................................... 6

DISCUSSION............................................................................................................................ 9

CHAPTER 3. PULLOUT CHECK........................................................................................... 11

RESEARCH AND CASE HISTORIES................................................................................ 12

Pullout Tests.......................................................................................................................... 12

Case Histories of Geosynthetic Pullout ................................................................................ 12

Assessment of Pullout Stability ............................................................................................ 14

DISCUSSION.......................................................................................................................... 16

CHAPTER 4. ECCENTRICITY............................................................................................... 19

ECCENTRICITY, BEARING PRESSURE DIAGRAM, AND FOOTING LIFTOFF... 19

OVERTURNING VERSUS LIFTOFF................................................................................. 23

OVERTURNING DESIGN REQUIREMENTS FOR REINFORCED SOIL WALLS .. 25

SUMMARY ............................................................................................................................. 28

CHAPTER 5. LATERAL PRESSURES................................................................................... 29

COMPACTION-INDUCED STRESSES ............................................................................. 29

MEASUREMENTS OF LATERAL PRESSURES IN REINFORCED SOIL ................. 33

MEASUREMENTS OF LATERAL PRESSURES AGAINST THE FACE OF 

REINFORCED WALL SYSTEMS................................................................................. 34

METHODS FOR ESTIMATING LATERAL EARTH PRESSURES ON GRS 

WALL FACING ............................................................................................................... 41

Wu’s Procedure (2001)......................................................................................................... 41

Soong and Koerner’s Procedure (1997)................................................................................ 43

COMPARISON OF PREDICTED AND MEASURED LATERAL PRESSURES.......... 45

FIELD CRITIQUE OF RANKINE ACTIVE THRUST THEORY .................................. 46

LATERAL PRESSURES DUE TO APPLIED LOADS ..................................................... 47

SUMMARY ............................................................................................................................. 49

CHAPTER 6. W EQUATION ................................................................................................... 51

W EQUATION........................................................................................................................ 51

RELIABILITY OF THE W EQUATION............................................................................ 54

CHAPTER 7. REDUCTION FACTORS ................................................................................. 57

RESEARCH AND CASE HISTORIES................................................................................ 57

Degradation........................................................................................................................... 57

Installation Damage .............................................................................................................. 58

Creep ..................................................................................................................................... 58

iii
DISCUSSION.......................................................................................................................... 61

Long-Term Degradation ....................................................................................................... 61

Installation Damage .............................................................................................................. 62

Creep ..................................................................................................................................... 62

Current Practice .................................................................................................................... 62

REFERENCES............................................................................................................................ 65

iv
LIST OF FIGURES

Figure 1. Drawing. Pullout failure, exaggerated for purposes of illustration ............................... 11

Figure 2. Illustration. Effects of facing connection condition on tensile force distribution in 

reinforcement .......................................................................................................................... 14

Figure 3. Illustration. Sequence of cutting of the PWRI test wall ................................................ 15

Figure 4. Chart. The resulting horizontal movement of the PWRI test wall. ............................... 15

Figure 5. Chart. Lateral stress/pressure on different sections of a GRS wall. .............................. 17

Figure 6. Drawing. Free body diagram, cantilever wall system for overturning calculations...... 19

Figure 7. Equation. Estimated maximum bearing pressure, subscript max .............................. 19

Figure 8. Equation. Estimated minimum bearing pressure, subscript min................................ 20

Figure 9. Equation. Simplified estimated maximum bearing pressure, subscript max............. 20

Figure 10. Equation. Simplified estimated minimum bearing pressure, subscript min ............ 20

Figure 11. Equation. Resultant force from the toe, subscript R................................................ 20

Figure 12. Drawing. Bearing pressure diagram—resultant force within middle third. ................ 21

Figure 13. Drawing. Bearing pressure diagram—resultant force on middle third. ...................... 21

Figure 14. Drawing. Bearing pressure diagram—resultant force outside middle third................ 22

Figure 15. Equation. Resultant ratio, RR. ..................................................................................... 22

Figure 16. Equation. Percent area of footing in compression, A subscript fc (in percent) ........... 22

Figure 17. Drawing. Forces on a rectangular rigid wall ............................................................... 23

Figure 18. Equation. Vertical component of resultant force at offset e, Rve ................................ 24

Figure 19. Equation. Force e......................................................................................................... 24

Figure 20. Equation. Safety factor, F............................................................................................ 24

Figure 21. Equation. Footing width, B.......................................................................................... 24

Figure 22. Equation. Value of footing width B for liftoff to occur............................................... 24

Figure 23. Photo. First negative batter GRS wall constructed in Colorado for CDOT. ............... 26

Figure 24. Photo. GRS wall in New Zealand with a negative. ..................................................... 27

Figure 25. Photo. Research on negative batter GRS walls in Japan............................................. 27

Figure 26. Chart. Residual lateral stress induced by vibratory plate compaction......................... 30

Figure 27. Chart. Vertical stress distribution at 6-kN (1.35-kips) vertical load on the GRS 

(a) without and (b) with reinforcement................................................................................... 31

Figure 28. Chart. Horizontal stress distribution at 6-kN (1.35-kips) vertical load on the GRS 

(a) without and (b) with reinforcement................................................................................... 32

Figure 29. Chart. Shear stress distribution at 6-kN (1.35-kips) vertical load on the GRS 

(a) without and (b) with reinforcement................................................................................... 33

σ

σ

σ

σ

χ

v
Figure 30. Drawing. Founders/Meadows bridge abutment cross-section and instrumentation.... 36

Figure 31. Drawing. GRS mini-pier elevation plan...................................................................... 37

Figure 32. Drawing. GRS mini-pier elevation plan...................................................................... 38

Figure 33. Chart. Measured increase in lateral pressures on the facing of GRS mini-piers 

during load test in kPa............................................................................................................. 39

Figure 34. Charts. Measured increase in lateral pressures on the facing of GRS mini-piers 

during load test in kN/m2 ........................................................................................................ 40

Figure 35. Diagram. Bin pressure diagram................................................................................... 42

Figure 36. Equation. Total lateral thrust, F subscript bin, as a function of the spacing 

between reinforcements, S subscript v, and peak bin pressure subscript h. ........................ 42

Figure 37. Equation. Total lateral thrust, F subscript bin, in terms of the coefficient of active 

earth pressure. ......................................................................................................................... 42

Figure 38. Drawing. GMSE wall configuration showing nonreinforced backfill zones.............. 43

Figure 39. Drawing. Lateral earth pressure distribution against wall facing exerted by the 

nonreinforced backfill............................................................................................................. 44

Figure 40. Equation. Connection force F...................................................................................... 44

Figure 41. Photo. Grand County, Colorado, wall with a maximum height of 55 ft (16.8 m)....... 47

Figure 42. Equation. Increase in lateral pressure subscript h. ................................................ 47

Figure 43. Drawing. Definition of angles and ........................................................................ 48

Figure 44. Chart. Increase in lateral stress due to 4,000-psf (191.5-kPa) vertical stress on a 

4-ft- (1.2-m-) wide footing with a 1-ft (.3-m) setback............................................................ 48

Figure 45. Drawing. Mohr circles for unreinforced and reinforced soil....................................... 52

Figure 46. Equation. Major principal stress subscript 1............................................................ 52

Figure 47. Equation. Confining stress due presence of reinforcement subscript 3R ................ 52

Figure 48. Equation. Capacity of the GRS subscript 1R........................................................... 53

Figure 49. Equation. Additional confining stress as function of reinforcement tensile strength 

divided by the reinforcement spacing subscript 3. ........................................................... 53

Figure 50. Equation. Modified version of the equation in figure 49 for additional confining 

stress, subscript 3 imposed with the W factor. ................................................................. 53

Figure 51. Equation. W factor....................................................................................................... 53

Figure 52. Equation. Capacity of the GRS, subscript 1R. ......................................................... 54

Figure 53. Equation. Expansion of the equation in figure 52 for the capacity of the GRS, 

subscript 1R, to include cohesion........................................................................................ 54

Figure 54. Graph. Predicted versus measured capacity of large-scale GRS tests using the 

equation in figure 52. .............................................................................................................. 56

σ

∆σ

α δ

σ

σ

σ

∆ σ

∆ σ

σ

σ

vi
Figure 55. Graph. Predicted versus measured capacity of large-scale GRS tests using the 

equation in figure 52 without the W term. .............................................................................. 56

Figure 56. Equation. Allowable reinforcement design strength, T subscript a............................. 57

Figure 57. Illustration. Schematic diagram of a SGIP test ........................................................... 61

vii
LIST OF TABLES

Table 1. Research: walls and abutments with uniform “short” reinforcement lengths. ................. 6

Table 2. Research: walls and abutments with “truncated base” reinforcement lengths. ................ 8

Table 3. Resultant force location and impact on safety against overturning, base area in 

compression, and bearing pressure diagram. .......................................................................... 23

Table 4. Key points relating to the need to limit the eccentricity in the design of GRS walls..... 28

Table 5. Comparison of Wu’s versus Soong and Koerner’s calculated lateral thrusts as a 

function of reinforcement spacing. ......................................................................................... 45

Table 6. Predicted versus measured lateral pressures on reinforced soil wall facing................... 46

Table 7. Prediction data for large-scale tests. ............................................................................... 55

viii
ABBREVIATIONS

AASHTO American Association of State Highway and Transportation Officials

ASD Allowable Stress Design

CDOT Colorado Department of Transportation

CIS Compaction Induced Stresses

CMU Concrete Masonry Unit

DACSAR Deformation Analysis Considering Stress Anisotropy and Reorientation

FEM Finite Element Model

FHWA Federal Highway Administration

GMSE Geosynthetic Mechanically Stabilized Earth

GRS Geosynthetic Reinforced Soil

IBS Integrated Bridge System

LRFD Load and Resistance Factor Design

MARV Minimum Average Roll Value

MSE Mechanically Stabilized Earth

NCHRP National Cooperative Highway Research Program

NCMA National Concrete Masonry Association

NGI Norwegian Geotechnical Institute 

NHI National Highway Institute

PWRI Public Works Research Institute (Japan)

RRR Reinforced Railroad/Road with Rigid facing

SGIP Soil-Geosynthetic Interactive Performance

USACE U.S. Army Corps of Engineers

ix

CHAPTER 1. INTRODUCTION

This report provides technical synopses on six topics related to geosynthetic reinforced soil 

(GRS) systems: embedment length, pullout check, eccentricity, lateral pressures, the W equation, 

and reduction factors. The synopsis for each topic includes a summary of the relevant research 

and a review of pertinent issues. GRS refers to closely spaced layers of geosynthetic 

reinforcement and compacted granular fill material. The GRS systems examined included, but 

were not limited to, walls, abutments, and the integrated bridge system (IBS).

Embedment length refers to the length of reinforcement for GRS walls or load-bearing walls. 

The synopsis on this subject addresses the issues of structures with short reinforcement lengths—

lengths less than 0.7 of a wall height at its face. The synopsis includes a review of research and 

case histories on short reinforcement. GRS walls involving a “constrained fill” zone (that is, 

where rock, heavily overconsolidated soil, or a nail wall is present behind a wall) are also 

discussed.

The pullout check refers to the implications of performing, or not performing, the check for 

geosynthetic pullout of a GRS wall or abutment. The pullout check is performed to determine 

required reinforcement length, but there is some uncertainty about the check’s efficacy. The 

synopsis examines the research and performance data, discusses the history of this particular 

design component in traditional geosynthetic mechanically stabilized earth (GMSE) theory, and 

considers the appropriateness of extending the pullout check to GRS technology.

Eccentricity is another name for overturning. As part of the design of conventional cantilever and 

gravity retaining walls with relatively rigid footings, overturning failure is commonly checked to 

determine whether the design meets the required margin of safety. This check typically consists 

of taking moments of all the stabilizing forces acting on the free body of the wall system about 

the toe of the wall and comparing them with the moments for the destabilizing forces about the 

same point. An alternative approach involves basing overturning criteria on a minimum base area 

in compression. This synopsis examines the purpose and theoretical justification of the 

eccentricity design component.

Lateral pressures at a GRS facing differ from those in the soil portion. The internal lateral stress 

in the GRS soil is governed by compaction-induced stresses and by supplementary confining 

effects that the reinforcement adds to the soil. The pressure at the facing is governed by those 

two factors as well as the movement of the facing, especially in facings with little or no 

connection strength. The synopsis discusses the effects of compaction-induced stresses on the 

internal lateral stress of both the unreinforced and reinforced soil. Measured thrusts against 

facing elements with and without purely frictional connection are presented. Methods for 

estimating this thrust are compared with the measured values.

The W equation is used to estimate the vertical capacity and required reinforcement strength of 

GRS walls and abutments. The W refers to a term or factor that attempts to explain why an 

increase in the reinforcement strength does not have the same effect as a proportional decrease in 

the reinforcement spacing. Proportional effects between strength and spacing have been assumed 

in simplified GMSE design. The synopsis discusses the various terms in the W equation and 

1
previous research and theory leading to its development. Based on load test results from the 

literature, the reliability of the equation is considered.

Lastly, the topic of reduction factors is examined. The synopsis includes a summary of the 

research and associated theory behind the reduction factors of creep, degradation—or 

durability—and installation damage. Deficiencies in current practices are also discussed. The 

reviewed literature includes reports on the impact of the use of various polymer types.

2
CHAPTER 2. EMBEDMENT LENGTH

Uniform reinforcement lengths of 0.6 to 0.7H (H is the wall height at face) have commonly been 

used for reinforced soil walls and abutments. This is a result of three major design guidelines: the 

American Association of State Highway and Transportation Officials (AASHTO) Standard 

Specifications for Highway Bridges, the Federal Highway Administration (FHWA) National 

Highway Institute (NHI) manual, and the National Concrete Masonry Association (NCMA) 

manual.

(1,2,3) The NCMA design manual, widely used in the private sector, requires a minimum 

reinforcement length of 0.6H. A minimum length of 0.7H, on the other hand, has been specified 

in the FHWA and AASHTO guidelines that are used routinely in the public transportation sector.

The standard of practice in Europe and Asia uses a similar criterion for minimum reinforcement 

length: 0.7H for routine applications and 0.6H for low lateral load applications with a minimum 

length of 9.84 ft (3 m).

(4,5) Brazilian guidelines, perhaps the most conservative of all, require a 

minimum length of 0.8H.

Despite the popularity of the 0.6 to 0.7H rule, uniform reinforcement lengths as small as 0.3 to 

0.4H have been employed with good success for construction of GRS walls.

(6)

The subject of 

short reinforcement lengths is important because such lengths can result in significant cost 

savings, especially in situations where construction involves excavation of existing ground.

Another important design issue with reinforcement length is nonuniform reinforcement lengths 

(truncated base walls). In situations where excavation into rock or stiff deposits to allow for 

uniform reinforcement lengths involves significant costs, the use of nonuniform reinforcement 

lengths in the lower part of the wall may prove beneficial. There are, however, design concerns 

with this practice.

RESEARCH AND CASE HISTORIES

This synopsis addresses the issues of short reinforcement length and design implications as they 

relate to GRS walls and abutments with short reinforcement lengths. The synopsis includes a 

review of research and case histories on short reinforcement and a discussion of the issues. Both 

uniform and nonuniform reinforcement lengths are addressed. GRS walls involving a 

“constrained fill” zone (i.e., where rock, heavily overconsolidated soil, or a nail wall is present

behind a wall) are also discussed.

Reinforced Soil Walls With Uniform Reinforcement Lengths

When considering the subject of short reinforcement lengths, the case histories and research by 

Tatsuoka and his associates in collaboration with Japan Railwayare perhaps the most 

noteworthy.

(7,8,9) Beginning in the mid-1980s, Tatsuoka and his associates developed a GRS wall 

system with full-height rigid facing, referred to as the reinforced railroad/road with rigid (RRR)

facing system. Because of its superior performance during earthquakes and heavy rainfalls

compared with conventional cantilever walls and metallic-reinforced soil walls, the RRR system

has become the “default” retaining wall type in Japan railway construction. To date, more than

74.5 mi (120 km) of RRR walls have been built for railway embankments in Japan with great 

success.

3
The RRR system involves a two-stage construction procedure. First, a GRS wall with 

geosynthetic reinforcement of approximately 0.3H in length, at typical spacing of .98 ft (0.3 m),

is constructed using gravel gabions as facing. The wall is allowed to deform under its self-weight 

before the second-stage construction is begun, which involves installing full-height rigid 

concrete facing (cast-in-place) over the gabion face. The rigid concrete facing is attached to the 

reinforced soil mass via extruded steel bars that are embedded in the soil mass. In spite of the 

very short reinforcement lengths of 0.3 to 0.4H, no failure has ever been observed during 

construction. It is important to note that both granular and cohesive backfills have been used in 

the construction of the RRR system. (Nonwoven–woven composite geotextiles were used for 

cohesive backfills to facilitate drainage while maintaining sufficiently high tensile capacity.)

(7)

Tatsuoka et al. suggest that overturning may be the most critical mode of failure for an GMSE 

wall with short reinforcement lengths of 0.3 to 0.4H, while sliding typically governs for 

conventional GMSE walls with reinforcement lengths of 0.6 to 0.7H.

(7)

Segrestin of Terre Armée International, in a rebuttal to Tatsuoka’s many comments regarding 

RRR walls versus reinforced earth walls, stated that a number of reinforced earth walls with 

reinforcement lengths as low as 0.45H had been constructed successfully.

(10)

Segrestin also cited 

a 34.4-ft- (10.5-m-) high instrumented reinforced earth wall with reinforcement length as short as 

0.48H. Segrestin maintained that the basic mechanism and behavior of reinforced soil structures 

with reinforcement lengths between 0.4H and 0.7H are identical, as has been suggested by a 

finite element model (FEM) parametric study conducted by Terre Armée.

(10)

Bastick conducted FEM analysis of MSE walls undergoing changes due to reduced 

reinforcement length.

(11) It was found, similar to Terre Armée’s study, that the performance of an 

MSE wall would remain quite similar as long as the reinforcement lengths were kept above 

0.4 to 0.5H and the reinforcement spacing stayed the same.

(10) The finding was confirmed by a 

full-scale experiment with reinforcement length of 0.48H loaded to an average surcharge of 

121.83 psi (840 kPa).

Other cases of using short reinforcement lengths have typically been associated with (a) walls 

with “constrained fill zone,” where there is the presence of a rock or heavily over-consolidated 

soil outcrop, or an existing nailed wall, and the space constraint makes commonly used 

reinforcement lengths of 0.6 to 0.7H impractical, and (b) walls with reinforcement anchored by 

metal plates or geosynthetic loops.

(12,13,14) Case (a) is in line with the scope of this synopsis and 

will be elaborated in the following discussion.

Lawson and Yee proposed a design-and-analysis method for reinforced soil retaining walls 

involving a constrained fill zone.

(12) Within the constrained reinforced fill zone, the full active 

failure wedge is unable to develop because of the relative close proximity of the rigid zone 

behind the reinforced fill. In the method, the magnitude of horizontal thrust acting on the wall 

face, Ph, is evaluated by the following equation: Ph = 0.5 K H2

. For walls with reinforcement

lengths greater than 0.5H, the Rankine active wedge can fully develop within the granular fill 

zone; hence, K is equal to Ka. However, for reinforcement lengths less than 0.5H, the full active 

wedge cannot develop fully, and the magnitude of K decreases with decreasing reinforcement 

lengths.

γ

4
An extreme example of reduced lateral thrust for walls with a constrained fill zone is a very tall 

wall constructed by Lin et al. of Taiwan, a region with very heavy seasonal rainfall.

(15) The wall 

was 129.6 ft (39.5 m) high, constructed in six tiers (with three 26.2-ft- (8-m-) high tiers and three 

shorter tiers). The geogrid reinforcement length was 4.9 ft (1.5 m), or 0.19H. All tiers, despite 

having to carry the soil-weight from upper tiers, performed satisfactorily even with the very short

reinforcement length.

It is interesting to note that the reinforcement length of 0.19H happens to agree with the finding 

of a numerical study for a reinforcement spacing of 1.3 ft (0.4 m).

(16) Vulova’s numerical study

also agrees with Tatsuoka’s assertion that overturning will be the controlling failure mode for an

GMSE wall with short reinforcement lengths.

Morrison et al. performed centrifuge tests on shored GMSE walls.

(17) The results were the 

following: (a) reinforcement lengths in the range of 0.25 to 0.6H generally produced stable wall 

systems; (b) reinforcement lengths of 0.25H or shorter generally produced outward deformation 

followed by an overturning collapse of the GMSE mass under increasing gravitational levels; 

(c) at reinforcement lengths less than 0.6H, deformation produced a “trench” at the shoring 

interface, interpreted to be the result of tension because the trench was observed with 

reinforcement lengths of 0.6H or longer; and (d) a conventional GMSE wall with retained fill 

and a reinforcement length of 0.3H was stable up to an acceleration level of 80 g, which 

represented a prototype height of approximately 88.6 ft (27 m).

It has been suggested that a shorter reinforcement length may result in larger lateral 

displacements and likely more settlement as well. A FEM study conducted by Chew, et al. 

showed that shortening reinforcement length from 0.7 to 0.5H caused about a 50-percent 

increase in lateral deformation.

(18) Ling and Leshchinsky reported that, with a reinforcement 

length of 0.5H, a reinforced soil wall would give satisfactory performance considering the

maximum displacement mobilized in the reinforcement layers.

(19) A recent study by Liu suggests 

that the larger lateral displacement is due to larger lateral deformation of the soil behind the 

reinforced zone.

(20)

It is of interest to note that a soil mass reinforced by closely spaced reinforcement (i.e., spacing 

not more than 0.82 to 0.98 ft (0.25 to 0.30 m)) will likely behave as a coherent mass. This 

behavior is clearly evidenced in two loading tests of full-scale segmental facing GRS bridge 

abutment walls, referred to as the National Cooperative Highway Research Program (NCHRP)

GRS test abutments.

(21,22) Two different woven geotextiles were used as reinforcement, each 

10.33 ft (3.15 m) long and at 0.66-ft (0.2-m) spacing. The backfill was a nonplastic silty sand, 

and the abutment was loaded by applying increasing vertical loads via a strip footing near the 

wall face. A tension crack was observed on the wall crest in both tests. The tension crack was 

first detected exactly where the reinforced zone ended (i.e., 10.33 ft (3.15 m) from the back face 

of the facing) under an applied pressure of 21.8 to 29.0 psi (150 to 200 kPa). The location of the 

tension cracks suggests that the reinforced soil mass behaves as a coherent mass. The coherent 

soil mass behavior is particularly prevalent for GRS walls with closely spaced reinforcement.

This also explains why overturning has been the most critical failure mode with short 

reinforcement lengths of 0.3 to 0.4H, especially with closely spaced reinforcement. The 

exception is when a constrained fill zone is involved.

5
Table 1 summarizes highlights of the case histories and case studies with uniform short 

reinforcement lengths (reinforcement lengths ≤ 0.5H).

Table 1. Research: walls and abutments with uniform “short” reinforcement lengths.

L/H

(length/height 

ratio)

H

(wall 

height)

Sv

(reinforcement

spacing) Key Findings Notes References

0.3

(typical)

16.4–

39.4 ft

(5–12 m)

(typical)

1 ft 

(0.3 m)

(typical)

Two-stage construction 

(wrapped-face wall 

with gravel-gabion in 

first stage, and fullheight panel in second

stage). Wall 

movement in the first 

stage is not of concern;

no failure has ever 

been observed.

Both granular and 

cohesive backfills have 

been used; more than

74.5 mi (120 km) of 

railroad railway walls 

and abutments have 

been built with excellent 

performance under static 

and seismic loads and 

heavy rainfalls. 

7, 8, and 9

0.48 34.4 ft

(10.5 m) not reported

Performed 

satisfactorily under an 

average surcharge of 

121.8 psi (840 kPa).

A FEM study conducted 

by Terre Armée 

indicated reinforcement 

lengths between 0.4H 

and 0.7H would perform 

approximately the same.

10 and 11

0.19

129.6 ft

(39.5 m)

in 6 tiers 

(tallest 

tier was 

26.2 ft 

(8 m))

6.6 ft

(2.0 m)

Performed 

satisfactorily, even in a 

region with very heavy 

seasonal rainfall.

Walls constructed in a 

“constrained fill”

condition (constructed 

over a nailed wall); L = 

0.19H happens to match 

FDM analysis results of 

minimum reinforcement 

length with Sv = 1.31 ft 

(0.4 m).

15 and 16

0.3

88.6 ft

(27 m)

(equiv.)

.78 inches

(20 mm)

(typical)

Stable up to an 

acceleration level of 

80 g.

Centrifuge modeling; 

L/H = 0.25 to 0.6 

generally produces a 

stable wall system in a 

“constrained fill”

condition.

17

FEM = Finite Element Model

FDM = Finite Difference Model

Reinforced Soil Walls With Nonuniform Reinforcement Lengths

Nonuniform reinforcement lengths in a reinforced soil wall, also known as a truncated base wall, 

typically have reduced reinforcement lengths in the lower part of the wall. A truncated base wall 

is employed when costs of excavation to allow for uniform reinforcement lengths are significant

and construction may be difficult. The reduction in length takes the following two forms: stepped 

wall (reducing lengths in groups of two to four layers) and trapezoidal wall (reducing lengths at 

almost every layer).

Nonuniform reinforcement lengths for GMSE walls are allowed in the FHWA NHI manual.

(2)

The manual provides general design guidelines for a truncated-base wall and states that this 

provision should only be considered if the base of the GMSE wall is founded on rock or 

6
competent soil; competent soil is soil that will exhibit minimal post-construction settlement. For 

foundation soil that is less than competent, ground improvement techniques may be used prior to 

GMSE construction.

The British standard BS 8006 design manual for GMSE walls also allows the use of a truncated 

base.(5) It states that a trapezoidal wall should only be considered where foundations are formed 

by excavation into rock or where other competent foundation conditions exist. The manual

prescribes a minimum reinforcement length of 0.4H for the lower portion of the wall. In Asia, 

Hong Kong’s Geoguide 6 follows the FHWA NHI manual, except it also stipulates that soil 

arching needs to be accounted for in design.

(4)

Japan Railway has employed reinforcement with truncated lengths in the lower part of a GRS 

wall when the costs of excavation to allow for full length reinforcement are high.

(7,23,24) The 

walls constructed with truncated base have performed satisfactorily during heavy rainfalls and 

severe earthquake events.

Segrestin reported applications of truncated base walls with steel-strip reinforcement in 

constrained fill situations.

(10) It was reported that wider metal strips, or a more closely spaced, 

increased number of strips, have been employed in truncated base walls. 

The Colorado Department of Transportation (CDOT) constructed a 24.9-ft- (7.6-m-) high GRS 

wall with a truncated base in DeBeque Canyon along Interstate 70.

(25) For comparison purposes, 

a 32.8-ft (10-m) section was constructed with full-length reinforcement. A road base material 

was used for backfill, and the wall was situated over a firm foundation. Measurements of lateral

displacements along a full-length reinforcement section (reinforcement length = 16.4 ft (5 m))

and along a truncated-base section (reinforcement length at base = 3.61 ft (1.1 m), or 0.14H), 

taken 6 months after construction, were very similar—both on the order of 0.12 to 0.23 inches 

(3 to 6 mm), with a maximum displacement of about 0.31 inches (8 mm). Adams et al. of the 

FHWA reported a number of GRS abutments with a truncated base.

(6) The GRS abutments have

performed satisfactorily.

Thomas and Wu conducted FEM analysis on the behavior of GRS walls with a truncated base.(26)

Major findings of the study were threefold. First, when designing a GRS wall with a truncated 

base, external stability should be thoroughly checked. Truncated-base walls are more likely to 

experience sliding failure, and the length of reinforcement at the lowest level should be at least 

0.35H or 0.9 m. Second, soil type and compaction of the backfill play a significant role in the 

performance of a GRS wall with a truncated base. The use of cohesive backfill should be 

avoided for a truncated-base wall. Third, the foundation soil needs to be sufficiently stiff for a 

truncated-base wall. Other numerical studies have also indicated that truncated base walls can 

perform equally well as full-length walls under certain conditions. (See references 11, 16, 26, 27, 

28, and 29.)

Lee et al. conducted a forensic study on a series of failed walls founded on rock with rock 

forming the back-slope for the lower reinforcements, and concluded that the resistance against 

sliding failure is reduced by a truncated base due to a smaller base area.

(30) They noted that soil 

arching due to the rock behind the fill would reduce vertical stress above the back of lower 

reinforcements and hence lead to overestimation of resistance to pullout failure.

7
Table 2 summarizes highlights of the case histories and case studies with “truncated base”

reinforcement lengths (reinforcement lengths ≤ 0.5H).

Table 2. Research: walls and abutments with “truncated base” reinforcement lengths.

LB

(reinforcement

length at base)

H

(wall 

height)

Sv

(reinforcement

spacing) Key Findings Notes References

4.6 ft

(1.4 m)

(0.27H)

17.1 ft

(5.2 m)

1 ft

(0.3 m)

The truncated base 

wall has performed 

satisfactorily.

This is one of the many 

truncated base walls 

constructed by Japan 

Railway. Of the 74.5 mi 

(120 km) of GRS walls 

constructed by Japan 

Railway, only a handful of 

truncated base walls are 

reported in the literature, 

all with base cut into stiff 

sloping ground. 

7

3.6 ft

(1.1 m)

(0.14H)

24.9 ft

(7.6 m)

0.7 ft

(0.2 m)

The difference in 

performance 

between truncated 

base wall and fulllength wall was 

found to be 

insignificant. 

Lateral 

displacements 

measured 6 months 

after construction 

for both walls were 

typically on the 

order of 0.12 to 

0.23 inches (3 to 

6 mm), with 

maximum 

displacement of 

0.31 inches (8 mm).

This is the only full-scale 

field tests on truncated 

base walls. A 32.8-ft 

(10-m) section of the wall 

was constructed with fulllength reinforcement for 

comparisons with truncated 

base wall.

25

4.6 ft

(1.4 m)

(0.38H)

12.1 ft

(3.7 m) 0.4 m

A truncated base 

wall deforms 

slightly more than 

full-length wall in 

the lower half of 

the wall, with 5- to 

10-percent larger 

lateral 

displacement.

FEM analysis results using 

a sophisticated FEM code 

Deformation Analysis 

Considering Stress 

Anisotropy and 

Reorientation (DACSAR).

29

8.2 ft

(2.5 m)

(0.4H)

19.7 ft

(6 m)

(approx.)

Not reported

The truncated base 

wall has performed 

satisfactorily.

When using a truncated 

base, wider metal strip 

reinforcement or, more 

often, increased number of 

strips have been employed.

10

8
LB

(reinforcement

length at base)

H

(wall 

height)

Sv

(reinforcement

spacing) Key Findings Notes References

4.9 ft

(1.5 m)

(0.11H)

44.3 ft

(13.5 m)

1 ft (0.3 m)

(within 4.9 ft 

(1.5 m) of the

wall face)

The truncated base 

wall has performed 

satisfactorily.

Reinforcement spacing was 

0.98 ft (0.3 m) for shorter 

(4.92 ft (1.5 m)) 

reinforcement; longer 

reinforcement (with a 

truncated base) was used at 

29.5 ft (0.9 m) spacing. 

Cement mixed gravel was 

used as backfill in the 

reinforced soil zone.

23

4.92 ft

1.5 m

(0.11H)

43.3 ft

(13.2 m)

1 ft

(0.3 m)

The truncated base 

wall has performed 

satisfactorily. 

The truncated base was 

installed into stiff sloping 

ground in steps.

17

GRS = Geosynthetic Reinforced Soil

FEM = Finite Element Model

DACSAR = Deformation Analysis Considering Stress Anisotropy and Reorientation

DISCUSSION

A minimum reinforcement length of 0.6 to 0.7H (H = height of wall at wall face) has been used 

in most designs of GRS/GMSE walls. However, reinforcement lengths as short as 0.3 to 0.4H

have been shown to be stable under certain conditions, and studies have suggested that GRS 

walls with reinforcement lengths between 0.4 and 0.7H behave approximately the same. For 

situations where a uniform reinforcement length is to be employed, a minimum reinforcement 

length of 0.6H is well justified. Care must be exercised to prevent tension crack on the wall crest 

if the wall is to carry significant vertical loads near the wall face (e.g., bridge abutments).

Extending the top one or two reinforcement layers well beyond the assumed failure plane based 

on a design analysis should help reduce the tension cracks.

Use of uniformly shorter reinforcement, with a reinforcement length of 0.35 to 0.5H, may result 

in larger lateral movement and is acceptable only if (a) the wall will not be subject to heavy edge 

loads such as a bridge abutment, (b) a granular backfill is employed and well compacted, (c) the 

foundation is competent (to minimize post-construction settlement), and (d) external stability, 

especially against overturning, is satisfied.

Use of a truncated base wall is a viable approach when costs of excavation to allow for uniform 

reinforcement lengths are significant and the foundation material is competent. There is strong 

evidence offered by case histories and research that a truncated base wall will perform 

satisfactorily as long as the base of a reinforced soil wall is founded on a competent foundation, 

and external stability, especially against sliding, is assured. The reinforcement length at the 

lowest level should generally be at least 0.3H.

9

CHAPTER 3. PULLOUT CHECK

Pullout and rupture of reinforcement are two modes of failure that have been identified for 

checking the internal stability (stability within reinforced zone) in the design of a reinforced soil 

wall or abutment. The pullout check is performed to determine required reinforcement length, 

whereas the rupture check is conducted to determine required reinforcement strength. The need 

to check pullout failure stems from the fundamental design concept of GMSE systems in which

the reinforcement is considered as tiebacks acting as tension members to “hold” an assumed 

failure wedge in place (see figure 1), thus preventing it from sliding down and away from the rest 

of the reinforced soil mass.

 

Figure 1. Drawing. Pullout failure, exaggerated for purposes of illustration.

In the three major design guidelines for reinforced soil walls and abutments—AASHTO

Standard Specifications for Highway Bridges, FHWA NHI manual, and NCMA design 

manual—the check of reinforcement pullout is performed by requiring the reinforcement be long 

enough that the “resisting force” (resulting from soil-reinforcement interface friction/interlocking 

resistance) beyond an assumed slip plane be sufficiently greater than the “driving force” (earth 

thrust within the tributary zone) for “any” level of a reinforcement layer.

(1,2,3) A minimum safety 

factor of 1.5 and a minimum anchored length of 2.95 ft (0.9 m) have typically been specified to 

assure pullout failure will not occur.

The driving force at a given reinforcement layer is evaluated by multiplying the reinforcement 

spacing by the lateral stress in the tributary zone along an assumed critical slip plane. On the 

other hand, the resisting force of a geosynthetic reinforcement layer is determined by multiplying 

the unit interface shear resistance by the length of reinforcement behind the assumed slip plane.

Both forces require a realistic assumption of the location of the most critical slip plane, which 

may be a function of, among other factors, soil friction angle and facing rigidity.

To develop a sound design concept, checking reinforcement pullout of a GMSE system is 

needed. This is because if a tieback system does not have a sufficient anchor length, the system 

11
is destined to fail. However, for a GRS system, the same is not true. Reinforcement length in a 

GRS system is determined by external stability. In a GRS system, geosynthetic reinforcement is

considered to be tensile inclusion, which serves to improve soil stiffness and strength (through 

such reinforcing mechanisms as inducing apparent cohesion, increasing confinement, 

suppressing dilation, and reducing lateral deformation). Because geosynthetic reinforcement 

layers are not considered tieback members, the necessity to check geosynthetic pullout for a GRS 

system has become a debatable issue.

Those who insist that geosynthetic pullout failure should not be checked for a GRS system have 

stated that if the design protocol of a GRS system involves checking pullout failure, the design 

has resorted to the tieback concept, which violates the fundamental design concept of GRS 

systems—it is like “mixing apples and oranges.” In addition, the pullout check is confusing and 

redundant for GRS systems.

On the other hand, those who maintain that geosynthetic pullout should be checked have argued 

that checking geosynthetic pullout failure is a simple step in a design, and therefore, it would not 

be a drawback to perform that extra step to ensure pullout stability. Although this argument 

sounds reasonable, it is perhaps not so simple.

RESEARCH AND CASE HISTORIES 

Pullout Tests

A great deal of research effort has been expended on the determination of reinforcement length 

to ensure stability against pullout failure. Since the 1980s, hundreds of studies have been 

conducted on the subject of a simple test called the “pullout test.” Huang collected 478 sets of 

pullout tests for a statistical analysis.(31) In a pullout test, a reinforcement element (metallic strip, 

metallic mesh, geosynthetic strip, geosynthetic sheet, etc.) is confined in soil under a constant 

surcharge, inside a test bin, and is subject to increasing tensile loads. Pullout tests that have been 

performed range from small- to large- to field-scale tests, from short- to long-term tests, from 

static to cyclic tests, from laboratory to field tests, from drained to undrained tests, from 

unsaturated to saturated tests, from finite element/difference analysis to close-form solution, and 

include centrifuge model tests.(See references 32 through 37.)

It is to be noted that pullout tests in fact only address the interface shear resistance that occurred 

beyond an assumed slip surface. In a pullout test, the location of the failure surface is not of 

concern; i.e., it does not address the question of the available resisting force in its entirety. For a 

GRS wall, it has been tacitly assumed that the Rankine or Coulomb active failure surface for a 

“uniform” soil mass is applicable to a reinforced soil mass that contains layers of geosynthetic 

reinforcement (e.g., AASHTO, Berg et al., and NCMA).(1,2,3)

Case Histories of Geosynthetic Pullout

After an exhaustive search of the literature, a single case of “in-service” reinforced soil wall, 

including both GMSE walls and GRS walls, was not found in which failure can be clearly

attributed to pullout of reinforcement. Part of the reason pullout failure is difficult to identify is 

because, unless a distinct failure surface is fully developed, pullout of reinforcement is “buried”

and not visible without careful reconnaissance work that involves removing the covering soil. It 

12
is conceivable that pullout failure in a reinforced soil wall or abutment may occur under certain 

combinations of the following conditions: (a) reinforcement is very stiff and has very high 

strength; (b) backfill is loosely compacted, or cohesive and wet; (c) reinforcement spacing is 

large (say, greater than 1.97 ft (0.6 m)); and (d) reinforcement is relatively long (say, ≥ 0.6H).

Suah and Goodings performed 27 centrifuge model tests of wrapped-faced geotextile walls, all 

with 100-percent clay backfill.(38) The reinforcement lengths varied from 0.5 to 1.5H. All the 

tests were brought to failure. Wall failure in all the tests occurred by tensile breakage of the 

geotextile, and never by pullout. This is contrary to the popular perception that clayey backfill is 

highly susceptible to pullout failure, and often cited as the primary disadvantage of employing 

clayey backfill in construction of a reinforced soil wall. Suah and Goodings also indicated that 

when reinforcement was short (0.5H), prefailure differential settlement at the front of the wall 

was pronounced, and failure by overturning was likely; when reinforcement was longer (1.0 to 

1.5H), differential settlement was less pronounced, and sliding along an inclined failure surface 

was more likely.

Woodruff performed a series of centrifuge model tests to examine potential failure surfaces of 

shored GMSE walls.(27) Yang et al. investigated the same subject using a limit equilibrium 

approach.(39) The two studies came to a similar conclusion that the critical failure plane for 

reinforcement length less than 0.7H was bilinear. The inclination angle of the critical failure 

plane within a shored GMSE wall is 10- to 20-percent less than that of the Rankine theory and 

extends along the contact surface between the back of the reinforced wall and the shored wall.

It is interesting to note that Leshchinsky and Vulova conducted a parametric study, using a finite 

difference program, Fast Lagrangian Analysis of Continua, or FLAC, to examine the effects of 

various parameters on GMSE walls with segmental facing.(40) Factors such as backfill strength, 

foundation strength, reinforcement stiffness, soil-reinforcement interface strength, reinforcement 

spacing, and wall height were examined. The study indicates that for closely spaced 

reinforcement (spacing less than 1.31 ft (0.4 m) in their definition), the critical slip plane did not 

pass through the reinforced soil zone. They suggest that a method different from Rankine or 

Coulomb analysis is needed to assess internal stability of a closely spaced GMSE wall. They also 

argue that the reason many experiments were able to bring about internal failure was because 

those walls do not typically represent realistic prototypes. They offered the following analogous 

example:

Consider the case of a rigid, non-yielding (unreinforced) retaining wall. Correct design 

would use at-rest lateral earth pressures reflecting the impossibility of an active wedge 

developing within the retained soil (a unrealistic failure mechanism). However, one can 

conduct an experiment on such a structure and impose the active case by exerting very 

large surcharges (i.e., cause yielding of the wall). Such an unrealistic experiment may 

lead to the incorrect conclusion that, for non-yielding walls, active soil pressures should 

be used. That is, the observed failure mechanism in the experiment is not necessarily 

relevant to the prototype. (page 363)(40)

13
Assessment of Pullout Stability

As noted in earlier in this synopsis, evaluation of pullout failure involves calculations of driving 

forces and resisting forces, and both involve the knowledge of the location of the critical slip 

plane (often taken as the plane of maximum reinforcement loads). The following presents studies

on the location of the critical slip plane and on the lateral thrust for the determination of the 

driving forces.

Critical Slip Plane

Tatsuoka pointed out that facing rigidity has a significant effect on the distribution of tensile 

forces induced in the reinforcement (see figure 2).(41) For reinforcement securely connected to 

rigid facing, the largest tensile force occurs near the wall face; whereas it occurs away from the 

wall face if the connection force is small. The distribution of reinforcement loads, hence the 

location of the critical slip plane, is strongly affected by the rigidity of the facing. 

 

©F. Tatsuoka, O. Murata, and M. Tateyama

Figure 2. Illustration. Effects of facing connection condition on tensile force distribution in 

reinforcement.

(7)

Yang et al. reported the behavior of a railway embankment wall constructed with a procedure 

similar to Japan Railway’s RRR system. (See reference 42 and 7 through 9.) From measured 

maximum reinforcement strains in different layers, they conclude that the critical slip plane is 

likely curved, with the failure surface in the upper portion of the wall not deviating significantly 

from the 0.3H-distribution suggested by Schlosser and Long, and the slip plane in the lower 

portion of the wall being quite different from those of Rankine or Schlosser and Long.(43)

Perhaps the most extreme example to illustrate the deficiency concerning the critical slip planes 

is a full-scale experiment conducted by the Public Works Research Institute (PWRI), Japan. The 

experiment involves a 19.7-ft- (6-m-) high GRS wall with a segmental concrete blocks wall face, 

and the backfill was a sandy soil reinforced with six layers of 11.5-ft- (3.5-m-) long polymer 

grid. The reinforcement inside the wall was sequentially severed at pre-selected sections after 

14
construction of the wall was completed. Figure 3 shows the sequence of cutting the 

reinforcements, and figure 4 shows the maximum horizontal movement corresponding to cutting 

of the reinforcements. It is seen that there was little movement due to the cutting of the 

reinforcements until Cut No. 55. Any current design methods of reinforced soil walls would have 

predicted a failure condition to occur long before Cut No. 55. In a tieback design method, cutting 

reinforcement is equivalent to shortening the tieback lengths. The cuts before Cut No. 55 would 

have rendered the tiebacks far shorter than the anchorage needed in any tieback design method

(e.g., AASHTO, FHWA, NCMA).

 

1 m = 3.28 ft

Figure 3. Illustration. Sequence of cutting of the PWRI test wall.

 

1 m = 3.28 ft

Figure 4. Chart. The resulting horizontal movement of the PWRI test wall.

From the standpoint of an internally stabilized retaining wall, the observed behavior is not at all 

surprising. A sheet of reinforcement, whether continuous or not, offers a similar restraining 

0

0

56 49 48 42 36 30 24 18 12 6

60

55

48 42 36

43 37 31 25 19 13 7 1

1 2 3 4

Horizontal distance (m)

1

2

3

4

5

6

Height (m)

0

1

2

3

4

5

6

Height (m)

0

0

56 49 48 42 36 30 24 18 12 6

60

55

48 42 36

43 37 31 25 19 13 7 1

56 49 30 24 18 12 6

60

55

48 42 36

43 37 31 25 19 13 7 1

1 2 3 4

Horizontal distance (m)

1

2

3

4

5

6

Height (m)

0

1

2

3

4

5

6

Height (m)

15
effect to lateral movement of soil, and hence will be able to achieve a stable composite. Note that 

the resulting stress distribution in the reinforcement will be rather different for continuous and

discontinuous reinforcements. It should be noted that the observation of the PWRI experiment 

was supported by similar experiments by John in the United Kingdom and by the Amoco test 

wall in the United States.

(44,25)

Lateral Earth Thrust Along Critical Slip Plane

Calculation of the driving force in pullout stability analysis involves determination of lateral 

thrusts along the critical slip plane. The FHWA NHI manual and AASHTO guidelines use 

Rankine active earth pressure for calculation of the driving force for GMSE walls with extensible 

reinforcement. The NCMA design manual uses Coulomb earth pressure for the calculation of the 

driving force.

Similar to AASHTO guideline, the British standard BS 8006 includes two methods for internal 

design of GMSE walls: the tieback wedge method and the coherent gravity method.(5) The 

former uses active earth pressure Ka for design; whereas the latter uses a K-value varying from 

the at-rest coefficient (Ko) to the active coefficient (Ka) in the upper 19.68 ft (6 m) of the wall, 

and Ka below 19.68 ft (6 m). The tieback wedge method is recommended when the short-term 

tensile strain exceeds 1 percent (e.g., polymeric reinforcement) while the coherent gravity 

method is recommended when the short-term tensile strain is less than 1 percent (e.g., steel 

reinforcement).

Chou and Wu conducted a study to investigate the performance of GRS walls with reinforcement 

at 0.98 ft (0.3 m) spacing by a sophisticated finite element code, DACSAR.

(29) The lateral 

pressures and stresses along three separate sections were examined: earth pressure against wall 

face, earth pressure against the back of the reinforced soil mass, and lateral stresses along the 

plane of maximum tensile force in reinforcement, as shown in figure 5. The study showed that 

the lateral thrusts in the three sections are rather different. The lateral earth pressure, the smallest 

of the three, is nearly constant with depth except near the base of wall (where there is greater 

constraint to deformation). The lateral stresses along the plane of maximum reinforcement 

tensile force is almost parallel to the lateral earth pressure profile except the magnitude is 

somewhat larger. The study also showed that the earth pressure exerted on reinforced soil mass 

by the retained fill, i.e., the earth pressure commonly used for evaluation of external stability of 

reinforced soil walls, is close to the Rankine active earth pressure.

DISCUSSION

Pullout of reinforcement stems from the design concept of GMSE walls in which reinforcement 

serves as tiebacks. The reinforcement needs to be long enough to provide sufficient anchorage 

length to ensure internal stability of the reinforced zone of a GMSE wall or abutment. GRS 

walls/abutments, however, are designed based on a different concept. The reinforcement in a 

GRS structure serves to improve the stiffness and strength of the soil-reinforcement composite. 

Reinforcement length of a GRS wall is dictated by external stability. In a GRS wall, pullout of 

geosynthetic reinforcement is theoretically not a design issue.

16
 

©N.N. S. Chou and J.T.H. Wu

1 ft = .305 m

1 psi = 6.89 kPa

Figure 5. Chart. Lateral stress/pressure on different sections of a GRS wall.

(29)

A very large number of pullout tests have been conducted by many researchers and designers to 

help assess pullout capacity of geosynthetic reinforcement under various conditions. It is 

important to point out that a great majority of the failures of reinforced soil walls involve 

prolonged rainfall (see references 35 and 45 through 50). In view of these findings, for design 

purposes, it will be prudent to conduct pullout tests under saturated undrained condition.

Observation of field performance of GRS walls and abutments with closely spaced reinforcement 

has strongly suggested that pullout is not a probable failure mode for closely spaced GRS walls.

Even though pullout failure is not impossible in a reinforced soil wall, a single case of in-service 

reinforced soil wall where failure can be clearly attributed to pullout of reinforcement has not yet 

been identified. Nonetheless, the reinforced soil system does not know whether the 

reinforcement is designed as tiebacks (as in a GMSE) or a tensile inclusion to improve soil 

stiffness and strength (as in a GRS). In other words, the reinforcement in a GMSE wall, even 

though designed as tiebacks, also serves to improve the stiffness and strength of the sounding 

soil (although not as effectively because of the large reinforcement spacing); and conversely, the 

reinforcement in a GRS wall, even though designed as reinforcing inclusion, also serves as 

tiebacks. As a result, one can argue that it would not be a drawback to also check for pullout 

 

 

0.0

0.5

1.0

1.5

2.0

2.5

3.0

3.5

4.0

0 10 20 30 40 50

Lateral Earth Pressure (psi)

Height (ft)

@ Wall face

@ Reinf. zone

@ Tmax plane

Ka

Ko

17
stability in the design of a GRS wall or abutment. Longer reinforcement length will likely lead to 

smaller lateral movement of a wall, unless the wall is in a “constrained fill” condition (where a 

firm sloping ground is behind the wall).(20)

The question then becomes: how to check for pullout failure if it is deemed warranted? The 

answer involves the knowledge of (a) the location of the critical slip plane in a reinforced soil 

mass and (b) the lateral thrust along the critical slip plane. The latter is less of an issue because 

the thrust can probably be assumed to be a combination of a Ka- or Ko-condition, depending on 

the desired degree of conservatism. The location of the critical slip plane, however, is a more 

troubling issue.

The location of the critical slip plane has traditionally been assumed to follow the Rankine 

theory. Based on studies by Tatsuoka and his associates, the location of a potential slip surface is 

a function of facing rigidity. The stiffer the facing, the closer the slip plane will be to the wall 

face.(41) Also, the location of the slip plane has been known to be affected by the loading 

condition on the wall crest. When a GRS wall is subject to large edge loads (as in an abutment), 

the critical slip plane is likely to be quite different from the Rankine active slip plane. The 

experiment performed by PWRI of Japan, as described previously under the subheading Critical 

Slip Plane, has further cast doubts on the applicability of the Rankine active slip plane. Also, it is 

possible that that such a slip plane does not exist for typical configurations, material properties, 

and loading conditions of GRS walls, as argued by Leshchinsky and Vulova.(40) As a final note, 

the NCHRP GRS test abutments with reinforcement spacing of 0.66 ft (0.2 m) and reinforcement 

length of 0.67H, and with loads being applied near the wall face, did not show any sign of 

geosynthetic pullout when loaded to failure. The reinforced soil mass clearly behaved as a 

coherent mass as it was subject to front edge loads all the way to failure. This further suggests 

that geosynthetic pullout is likely not a probable failure mode for closely spaced GRS abutments.

The inability to compute the driving force and resisting force with confidence may cast doubts 

on the usefulness for checking geosynthetic pullout of a GRS wall or abutment. Moreover, the 

method for checking pullout failure as adopted in current design guidelines is unrealistically 

conservative. The current design guidelines require that the pullout safety factor at each

reinforcement layer be greater than a prescribed value (typically 1.5).(1,2,3) The fact is, pullout 

failure only occurs if a group of reinforcement layers experiences instability related to pullout.

Pullout of any single reinforcement layer will never occur.

In summary, even though the extra design step of checking pullout failure will not harm the 

overall design of a GRS wall or abutment, two critical reasons—(a) pullout does not appear to be 

a probable failure mode, and (b) there is no reliable and realistic procedure for checking 

pullout—have rendered the reinforcement pullout check not particularly meaningful.

18
CHAPTER 4. ECCENTRICITY

As part of the design of conventional cantilever and gravity retaining walls with relatively rigid 

footings, overturning failure is commonly checked to determine whether the design meets the 

required margin of safety. This check typically consists of taking moments of all the stabilizing 

forces acting on the free body of the wall system about the toe (e.g., Point O for a cantilever wall 

shown in figure 6) and comparing them with the moments for the destabilizing forces about the 

same point. The required margin of safety for overturning is typically specified in the relevant 

code of practice.

 

Figure 6. Drawing. Free body diagram, cantilever wall system for overturning calculations.

An alternative means of assessing the overturning potential of retaining walls has been proposed 

by the U.S. Army Corps of Engineers, (USACE).

(51) Rather than comparing the stabilizing and 

destabilizing moments, the overturning criteria are based on a minimum base area in 

compression, i.e., limiting the amount of footing “liftoff.” This technical synopsis examines the

purpose and theoretical justification of this design component.

ECCENTRICITY, BEARING PRESSURE DIAGRAM, AND FOOTING LIFTOFF

For a wall supported on a rigid concrete footing, the amount of footing liftoff is determined from 

its bearing pressure diagram. If R and Rv are the resultant force and vertical component of the 

resultant force, respectively, acting on the free body of the wall system per unit length, the 

bearing pressure diagram can be determined for several possible resultant force locations. Key to 

this evaluation is the estimation of the maximum ( max) and minimum ( min) bearing pressures, 

calculated using the equations shown in figure 7 and figure 8:

 

Figure 7. Equation. Estimated maximum bearing pressure, subscript max.

σ σ

I

Mc

A

Rv σ max = +

σ

19
 

Figure 8. Equation. Estimated minimum bearing pressure, subscript min.

A equals the footing area, which is the footing width, B, per unit length for a plane strain 

problem. The distance from the centroid of footing to extreme fiber is c, which equals B/2. I is 

the moment of inertia of footing, and equals B3

/12 per unit length. M is the moment about the 

footing centroid, and equals Rv times e, e being the eccentricity of the resultant force from the 

footing centroid. After simplifying, the equations in figure 7 and figure 8 become the equations 

in figure 9 and figure 10:

Figure 9. Equation. Simplified estimated maximum bearing pressure, subscript max.

Figure 10. Equation. Simplified estimated minimum bearing pressure, subscript min.

For no footing liftoff, min must be ≥ 0. By setting min equal to 0 in the equation in figure 10, it 

can be seen that the resultant force acting on the free body of the wall system must lie within the 

middle third of the footing; i.e., e ≤ B/6.

As shown in the equation in figure 11, the point of action of the resultant force from the toe ( R) 

is calculated by simply summing the moments about the toe of all the forces acting on the free 

body of the wall system and dividing by the vector sum of the vertical force components acting 

on the footing (Rv).

 

Figure 11. Equation. Resultant force from the toe, subscript R.

If the resultant force acts within the middle third, the bearing pressure diagram is trapezoidal and 

100 percent of the footing is in compression; i.e., no liftoff (see figure 12). If the resultant force 

acts on the middle third point, the bearing pressure diagram is triangular and spans the entire 

footing width (figure 13). There is still no footing liftoff in this case. If the resultant force acts 

outside the middle third but within the footing, the bearing pressure diagram is also triangular but 

the span of the triangle is less than the footing width. The remaining portion of the bearing 

pressure diagram is negative; i.e., min < 0 (figure 14). The portion of the footing that is outside 

the triangular bearing pressure diagram (i.e., when < 0) is in a state of tension or liftoff.

Clearly, it is undesirable to allow any footing liftoff except perhaps during temporary live 

loading or during extreme events such as earthquakes.

I

Mc

A

Rv σ min = −

σ





 



 = +

B

e

B

Rv 6

σ max 1

σ





 



 = − B

e

B

Rv 6

σ min 1

σ

σ σ

χ

χχRR = ∑ Moments about the toe

RRvv

χ

σ

σ

20
 

Figure 12. Drawing. Bearing pressure diagram—resultant force within middle third.

 

Figure 13. Drawing. Bearing pressure diagram—resultant force on middle third.

21
 

Figure 14. Drawing. Bearing pressure diagram—resultant force outside middle third.

The area of the footing in compression can be calculated by first computing the resultant ratio, 

RR, as shown in the equation in figure 15:

Figure 15. Equation. Resultant ratio, RR.

The percent area of footing in compression (Afc) can be estimated as shown in the equation in

figure 16:

Figure 16. Equation. Percent area of footing in compression, A subscript fc (in percent).

It follows that the percent area of footing experiencing liftoff is 100 percent minus Afc.

If the resultant force acts at the toe of the footing, the wall is on the verge of overturning; i.e., the 

destabilizing moment is just equal to the stabilizing moment and there is no margin of safety 

available. If the resultant force acts outside the footing, the wall will overturn. The USACE 

requires that, in designing for earthquake protection, the resultant force must lie within the 

footing for retaining walls.(51) The various possible scenarios described above are summarized in

table 3.

RR = χR/B

Afc (in %) = 100 when 1⁄3 ≤ χR/B ≤ 1⁄2

Afc (in %) = 300 x RR = 300χR/B when χR/B < 1⁄3

22
Table 3. Resultant force location and impact on safety against overturning, base area in 

compression, and bearing pressure diagram.

Resultant Force 

Location

Safety Against 

Overturning

Base Area in 

Compression, 

Afc

Bearing Pressure 

Diagram

Within middle 

third Very safe 100 percent

Trapezoidal and positive 

over entire footing width B

On middle third 

point Safe 100 percent

Triangular and positive 

over entire footing width B

Outside middle 

third but within 

footing

Margin of safety 

diminishes as resultant 

pushes towards toe but 

stabilizing moment is still 

≥ overturning moment

0 percent < Afc < 

100 percent

Triangular and positive 

over front portion of 

footing; triangular and 

negative over back portion 

of footing

On footing toe

No margin of safety; i.e.,

stabilizing moment = 

overturning moment 0 percent

Negative over entire 

footing width B except at 

toe where max → +

Outside footing Unsafe. Will overturn 0 percent

Negative over entire 

footing width B

OVERTURNING VERSUS LIFTOFF

In this section, a simple methodology is presented to determine when liftoff will control.

Figure 17 depicts the simple case of the lateral (Ea is horizontal because the wall friction is 

assumed to be 0) and vertical (W) forces acting on a rectangular wall with a rigid base.

 

Figure 17. Drawing. Forces on a rectangular rigid wall.

σ ∞

δ

23
The two forces—lateral and vertical—can be represented by a single resultant force, R, which is 

acting at an offset e from the centroid of the wall base. This eccentricity can be estimated by 

decomposing the resultant force into the vertical and horizontal (not shown) components at the 

base and summing moments about the centroid of the wall base as shown in the equation in

figure 18:

 

Figure 18. Equation. Vertical component of resultant force at offset e, Rve.

Rv is the vertical component of the resultant force, whose moment about the centroid of the wall 

base is not zero as opposed to the horizontal component. Therefore, the equation in figure 19

follows.

Figure 19. Equation. Force e.

The factor of safety against overturning can be obtained by summing moments about the toe as 

shown in the equation in figure 20:

Figure 20 . Equation. Safety factor, F.

Solving the equation in figure 20 for B gives the equation shown in figure 21:

 

Figure 21. Equation. Footing width, B.

For liftoff to occur, the condition shown in figure 22 must be satisfied.

 

Figure 22. Equation. Value of footing width B for liftoff to occur.

By inspecting the equations in figure 21 and figure 22, it can be determined that liftoff occurs 

when the factor of safety against overturning is less than 3; i.e., liftoff is not an issue when the 

overturning factor of safety exceeds 3. This criterion is only true when no surcharge loads are 

acting on top of the wall. When surcharge loads act on top of the wall (as is the case in a GRS 

IBS), then the factor of safety against overturning for no liftoff will be less than 3. How much 

less than 3 depends on the magnitude of the load on top of the wall and its point of action.

3

H Rv e = Ea

v

a

R

E H

e

3 =

e

B

E H

R B

E H

WB F

a

v

a / 3 2

/ 2

/ 3

/ 2 = = =

B = 2eF

B < 6e

24
OVERTURNING DESIGN REQUIREMENTS FOR REINFORCED SOIL WALLS

• AASHTO.

(52) In the 1998 AASHTO LRFD Bridge Design Specifications, the practice of 

calculating the ratio of stabilizing moment to overturning moment as a check for 

overturning of bridge abutments and GMSE walls was replaced by the resultant location 

criteria coupled with a check for bearing pressure. The criteria state that the resultant 

force for walls with footings on soil and rock must lie within the middle third and middle 

half, respectively, for allowable stress design (ASD) (middle half and middle three 

quarter for load and resistance factor design (LRFD)).

• FHWA.

(53) FHWA 1997 guidelines for GMSE structures state that “the flexibility of 

MSE walls should make the potential for overturning failure highly unlikely.” Unlike 

cantilever and gravity walls with relatively rigid footings, the base of GMSE or GRS 

walls have little or no flexural rigidity. The notion that they can support bending stresses 

with tensile contact pressures or liftoff developing seems counterintuitive and has not 

been observed in instrumented GMSE structures such as those studied by Simac et al. and 

Bathurst et al.(54,55) Therefore, the FHWA 1997 guidelines omit overturning as a potential 

failure mode for GMSE structures. To control lateral deformation due to tilting, however, 

FHWA requires that the eccentricity of the resultant force be checked to determine 

whether it exceeds a maximum permissible value as prescribed by AASHTO.

• NCMA.

(3) NCMA indicates in its Design Manual for Segmental Retaining Walls (these 

are geosynthetic reinforced walls that use dry stacked interlocking modular masonry 

concrete units as a facing) that overturning about the toe should be considered as a 

potential failure mechanism, a view that differs from that of the FHWA. NCMA does not 

require consideration of the base eccentricity criteria.

• Canadian Geotechnical Society.

(56) Like the NCMA, the Canadian Foundation 

Engineering Manual (Canadian Geotechnical Society, 2006) does require checking for 

overturning of reinforced walls.

• British Standards Institution.

(5) British standard BS 8006 (1995) does not require 

overturning of strengthened/reinforced soil and other fills be assessed during design. For 

external stability, the British standard does require a check for bearing capacity, sliding, 

and external slip. In checking for bearing capacity, the maximum bearing pressure (σmax) 

is compared with the bearing capacity. Because σmax increases with increasing 

eccentricity, the implication is that the eccentricity will be automatically limited by 

imposing a limit on the bearing capacity.

• Designing With Geosynthetics.

(57) Koerner indicated that the factor of safety against 

overturning for geotextile reinforced walls is typically very high. He further points out 

that overturning is not a likely failure mechanism because of base flexibility. Therefore, 

“many engineers do not even include an overturning calculation in the design process.”

• Retaining Wall-Dialog: “A Tale of Two Walls” and “GRS—A New Era in 

Reinforced Soil Technology.”(58,59) Barrett and Ruckman presented cases of GRS walls 

with negative batter being used or tested successfully in Colorado (figure 23), New 

25
Zealand (figure 24), Japan (figure 25), and elsewhere. Barrett wrote “prevailing 

guidelines for internally supported MSE walls are based on translations from externally 

supported cantilever walls and are generally inappropriate.” GRS walls with negative 

batter that generally do not meet the overturning requirements are still performing well 

today.

 

Source: R.K. Barrett and A.C. Ruckman

Figure 23. Photo. First negative batter GRS wall constructed in Colorado for CDOT.

(59)

26
 

Source: R.K. Barrett and A.C. Ruckman

Figure 24. Photo. GRS wall in New Zealand with a negative.

(59)

 

Source: R.K. Barrett and A.C. Ruckman

Figure 25. Photo. Research on negative batter GRS walls in Japan.

(59)

27
SUMMARY

Table 4 offers key points relating to the need to limit the eccentricity in the design of GRS walls.

Table 4. Key points relating to the need to limit the eccentricity in the design of GRS walls.

Points that argue for the need for a 

limiting eccentricity criterion

Points that argue against the need for a limiting 

eccentricity criterion

It is prudent to consider a limiting 

eccentricity criterion.

The check that the bearing pressure meets the 

bearing capacity criterion requires max to be less 

than the bearing capacity with an adequate margin of 

safety. Because max increases with increasing 

eccentricity, it implies that the eccentricity will be 

automatically limited by imposing a limit on the 

bearing capacity.

GRS walls can be less wide at the base 

than GMSE walls according to different 

design methods (Adams et al., and Berg 

et al., respectively), and hence, the 

factors of safety pointed out by Koerner 

in GMSE structures may not be as high 

for GRS walls. (6,2,57)

Flexible reinforcing elements (e.g., geotextiles) do 

not reinforce the soil by any interaction involving 

bending or shear across their cross-sectional area. 

They merely interact with the soil by absorbing axial 

tension. Consequently, the base of GRS walls cannot 

be considered rigid. While overturning is a valid 

failure mechanism for walls with rigid bases, GRS 

walls cannot fail by overturning because bending 

moments cannot transfer to the toe of such a flexible 

system. Moreover, it is questionable whether the 

methodology and equations presented herein apply 

to a reinforced soil wall system with flexible bases.

Even when the equations presented herein are used, 

Koerner points out that the factor of safety against 

overturning for geotextile reinforced walls is 

typically high.(57)

FHWA 1997 guidelines require limiting the 

eccentricity of the resultant force only for the sake of 

limiting lateral deformation. It may be implied that 

lateral deflection increases with increasing

eccentricity but such correlations are not readily 

available to decide on an acceptable eccentricity 

value for design.

Barrett showed that GRS walls with negative batters 

have performed very well.(58,59)

GMSE = Geosynthetic Mechanically Stabilized Earth

GRS = Geosynthetic Reinforced Soil

Clearly, checking for limiting eccentricity is generally prudent. However, such a check is based 

on theories that really do not apply to flexible base systems. The check is intended as a failure 

mechanism that is feasibly impossible in a GRS and thus should be discontinued.

σ

σ

28
CHAPTER 5. LATERAL PRESSURES

The lateral stress at a GRS facing is different from that in the soil portion. The internal lateral 

stress in the GRS soil is governed by compaction-induced stresses (CIS) and by additional 

confining effects that the reinforcement provides to the soil. The pressure at the facing is 

governed by these two factors as well as the movement of the facing especially in facings that 

have little or no connection strength. The effects of CISs on the internal lateral stress of both the 

unreinforced and reinforced soil is discussed. Measured thrusts against facing elements with and 

without purely frictional connection are presented. Methods for estimating this thrust are used to 

compare with the measured values.

COMPACTION-INDUCED STRESSES

Based on Duncan and Seed’s work on CISs, Williams developed simple charts to estimate the 

residual lateral earth pressures on a nonyielding wall due to compaction.(60,61) An example of a 

chart for residual lateral stress due to vibratory plate compactors, usually used in GRS 

construction, is shown in figure 26. At shallow depths, the CIS is a function of the load from the 

vibratory plate (q) and it exceeds the at-rest value. The family of curves represents different 

values of q ranging from 5 to 16 psi (34.5 to 110.3 kPa). At greater depths, the lateral pressures 

revert to the at-rest value once the overburden pressure exceeds the preconsolidation pressure 

generated during compaction (by the vibratory plate compactor). The family of dashed lines 

represents the at-rest stresses for varying soil unit weight and friction angles, (Ko = 1 - sin ).

Figure 26 is applicable to compaction of unreinforced soil.

CIS in GRS is likely to be higher than in the same soil unreinforced because the reinforcementsoil-interface friction restrains soil movement thereby producing larger values of soil confining 

stress, which manifests in the form of greater CIS. Evidence of increased lateral stress has been 

demonstrated by numerical analysis performed by Ketchart and Wu for a compacted GRS mass 

loaded to 6 kN (1.35 kips).(62) The vertical stress distribution in the loaded unreinforced soil and 

GRS are similar, as shown in figure 27. However, the horizontal (figure 28) and shear (figure 29) 

stresses are higher in the GRS than in the same soil unreinforced. They are higher near the 

reinforcement and decrease with increasing distance from the reinforcement. The increase in 

lateral stress will result in increased strength of the GRS. Based on these results, Ketchart and 

Wu recommended that reinforcement spacing be limited to 12 inches (30.48 cm) for improved 

performance—hence the advent of closely spaced reinforcement, more commonly called 

GRS.(62)

Even though CIS clearly adds an important contribution to the vertical capacity of a GRS wall, 

one drawback when considering CIS in design is that the compaction equipment is seldom 

known a priori. Therefore, any quantitative analysis of the contribution of compaction to enhance 

the lateral stresses, and hence capacity in a GRS in the field, is speculative at best.

φ φ

29
 

Source: Francis Group LLC Books

1 ft = .305 m

1 psf = .05 kPa

1 psi = 6.89 kPa

1 pcf = 16.02 kg/m3

Figure 26. Chart. Residual lateral stress induced by vibratory plate compaction.

(63)

30
©K. Ketchart and J.T.J.H. Wu

1 kPa = 20.89 psf

1 mm = .039 inches

1 kN = 0.23 kips

Figure 27. Chart. Vertical stress distribution at 6-kN (1.35-kips) vertical load on the GRS 

(a) without and (b) with reinforcement.

(62)

31
 

©K. Ketchart and J.T.J.H. Wu

1 kPa = 20.89 psf

1 mm = .039 inches

1 kN = 0.23 kips

Figure 28. Chart. Horizontal stress distribution at 6-kN (1.35-kips) vertical load on the 

GRS (a) without and (b) with reinforcement.

(62)

32
 

©K. Ketchart and J.T.J.H. Wu

1 kPa = 20.89 psf

1 mm = .039 inches

1 kN = 0.23 kips

Figure 29. Chart. Shear stress distribution at 6-kN (1.35-kips) vertical load on the GRS 

(a) without and (b) with reinforcement.

(62)

MEASUREMENTS OF LATERAL PRESSURES IN REINFORCED SOIL

Lateral stresses in a reinforced soil mass were measured by Yang.

(64) He conducted footing 

model tests on both reinforced and unreinforced soil using a rigid 8-inch- (20.3-cm-) diameter 

stainless steel plate as a footing. The soil consisted of a dry sand with an average density of 

92 pcf (1,474 kg/m3) (corresponding to a relative density of 70 percent ( d min = 81.5 pcf

(1,306 kg/m3, and d max = 96.5 pcf (1,546 kg/m3)). This density was achieved by pluviation,

raining the sand from a height of 9 inches (22.9 cm). Compaction-induced stresses were 

nonexistent in this study. Therefore, any difference in lateral stress between the reinforced and 

γ

γ

33
unreinforced sand was purely a result of the effect of the reinforcement. The reinforcement 

consisted of six 20-inch-(50-cm-) diameter fiberglass nets having average tensile strengths of 

202 lb/inch (35.4 kN/m) in one direction and 178.5 lb/inch (21.3 kN/m) in the other. The first 

reinforcing was placed 2 inches (5.1 cm) below the footing. The additional reinforcings were 

spaced at 2 inches (5.1 cm).

Lateral stresses were measured at several depths along the centerline of the footing using 

pressure cells that were 1.4 inches (3.6 cm) in diameter and 0.2 inches (0.5 cm) thick. To prevent 

damage to the pressure cells that were originally designed by the U.S. Bureau of Mines, the 

maximum applied stress on the footing was limited to 75 percent of the ultimate bearing capacity 

for the unreinforced sand. For the reinforced sand, the applied stress was kept below a level that 

corresponded to a maximum lateral pressure of 40 psi (276 kPa).

Based on the normalized measured lateral stress (lateral stress divided by the applied stress) 

versus the normalized depth (depth divided by the footing radius) in both the reinforced and 

unreinforced soil. Yang offered the following results.

(64)

• The unreinforced soil at two different stress levels clearly indicate that the radial stress to 

applied stress ratio increased with increasing stress level. For the reinforced soil, Yang 

indicated that the lateral stress ratio was independent of the applied stress level. He did 

caution that this might not be true as the applied stress approached the ultimate bearing 

capacity of the footing on the reinforced soil. However, in the interest of preserving the 

pressure cells, the tests did not reach such high loads.

• The lateral stress in the reinforced soil was fairly constant from a depth of 0.5 to 

2.25 times the footing radius. The corresponding reduction in lateral stress in the 

unreinforced soil with increasing depth was more significant.

• The lateral stresses in the reinforced sand were in general higher than those in the 

unreinforced sand. A GRS is expected to be stronger than the same soil unreinforced. For 

this to occur, the confining stress in the soil must effectively increase due to the presence 

of the reinforcement, as shown numerically by Ketchart and Wu.(62) These data provide 

compelling evidence of this lateral stress increase.

MEASUREMENTS OF LATERAL PRESSURES AGAINST THE FACE OF 

REINFORCED WALL SYSTEMS

Lateral stresses have been measured in many reinforced soil wall systems. They include systems 

with reinforcing elements that are both connected and not connected to the wall face. Among the 

those not connected, the wall systems consisted of a concrete wall at the end of a bridge (Warren 

and LeGrand), an aluminum plate (Ahmadi and Hajialilue-Bonab), brick elements mortared 

together (Dalton, Walsh, and Pinto and Cousens), and gravity/rigid retaining walls (Saran et al.,

Garg and Saran, Mittal et al., and Garg et al.). (See references 65 through 73.) Among those that 

were connected, the wall systems consisted of a full-height plywood facing panel connected to 

geogrids (Shinde and Mandal), gabion walls connected to geogrids, woven and nonwoven 

geotextiles (Won and Kim), modular block facing frictionally connected to geogrids (Abu-Hejleh 

34
et al. and Mitchell), and full-height-panel and incremental panel walls connected to geogrids 

(Yogarajah and Saad). (See references 74 through 78.)

Abu-Hejleh et al. reported lateral pressures on the facing of a 19.4-ft- (5.9-m-) high segmental 

retaining wall reinforced with 16-inch- (40.6-cm-) spaced Tensar® UX 6 geogrids at the 

Founders/Meadows bridge in Denver, Colorado (figure 30).(76) These pressures were only 

reported at the end of construction with the bridge superstructure loading the abutment. This is 

not a true GRS because the spacing was greater than 12 inches (30.48 cm). The abutment fill, 

described as a gravelly CDOT Class 1 backfill, had a measured unit weight of 141 pcf 

(2,259 kg/m3

) with peak shear strength parameters from large-scale triaxial tests of φ = 

39 degrees and c = 1,440 psf (68.0 kPa).

Earth pressure cells were mounted above the 7th, 9th, 11th, and 12th geogrid layer from the 

bottom. Abu-Hejleh et al. indicated that the measured lateral pressures were significantly less 

(6 to 35 percent) than the Rankine active pressures plus the footing-induced pressure assumed a 

2V:1H load spread (see table 6). Lateral pressures before placing the bridge footing were not 

reported.

35
 

Source: N.M. Abu-Hejleh et al.

Figure 30. Drawing. Founders/Meadows bridge abutment cross-section and 

instrumentation.

(76)

Mitchell measured lateral pressures on the concrete masonry unit (CMU) facing of four GRS 

mini-piers (figure 31 and figure 32).(77) The reinforcement (Amoco 2006—wide-width tensile 

strength = 2,100 lb/ft(30.6 kN/m)) spacings were 24 inches (60.9 cm), 16 inches (40.6 cm), 

8 inches (.2 m), and 32 inches (81 cm) in piers 1, 2, 3, and 4, respectively. The facing consisted 

of 99-lb (45.0-kg) split-face keystone retaining wall blocks that were 18 inches (46 cm) long, 

5 inches (12.7 cm) wide, and 8 inches (20.3 cm) deep with a flange protruding to the inside of 

the piers. Corner blocks with slightly different weight and dimensions were used.

Known as trap rock gravel (32.1 percent gravel, 56.5 percent sand, and 11.5 percent fines; 

Cu = 40, Cc = 3.2, Gs = 2.93; Standard Proctor d max = 144 pcf (2,307 kg/m3

); 

wopt = 8.75 percent; Modified Proctor d max = 150 pcf (2,403 kg/m3); wopt = 6.5 percent), the 

backfill had shear strength parameters of = 36 degrees and c = 637 psf (30.5 kPa) based on a 

failure relative displacement of 10 percent of the shear box length as measured in a 12- by

12-inch (30.5- by 30.5-cm) direct shear box. The shear stress-displacement curves appear to be

still strain hardening at the interpreted failure relative displacement.

γ

γ

φ

36
 

Source: J.W. Mitchell

1 m = 3.28 ft

Figure 31. Drawing. GRS mini-pier elevation plan.

(77)

37
 

Source: J.W. Mitchell

Figure 32. Drawing. GRS mini-pier elevation plan.

(77)

GeokonTM vibrating wire pressure transducers were placed at five elevations (1 ft (0.3 m), 4.3 ft

(1.3 m), 7 ft (2.1 m), 9.7 ft (3.0 m), and 13 ft (4.0 m) above base) along the GRS wall height. The 

measured lateral pressures are shown in figure 33 and figure 34. These pressure readings were 

zeroed at zero load and represent only the increase in the lateral pressures during the load test.

Mitchell noted that with a decrease in reinforcement spacing, the lateral pressures increased and 

became more equally distributed over the wall height.(77)

The actual measured lateral pressures at zero load were provided in appendix C of Mitchell’s 

report and are shown in table 6. Unfortunately, they are mostly negative, with a few exceptions.

When and how the zero readings were obtained could not be discerned from Mitchell’s report.

38
 

Source: J.W. Mitchell

1 kPa = 20.89 psf

1 m = 3.28 ft

1 cm = 0.39 inches

Figure 33. Chart. Measured increase in lateral pressures on the facing of GRS mini-piers 

during load test in kPa.

(77)

39
 

Source: J.W. Mitchell

1 kN/m2 = 20.9 lbs/ft2

Figure 34. Charts. Measured increase in lateral pressures on the facing of GRS mini-piers 

during load test in kN/m2.

(77)

40
Yogarajah and Saad experimentally measured the lateral earth pressures on incremental panel 

walls backfilled with reinforced Leighton Buzzard sand.(78) The dry unit weight of the compacted 

sand was 104 pcf (16.4 kN/m3

) corresponding to a relative density of 82 percent (peak and 

constant volume friction angles of 47 degrees and 34 degrees, respectively). The test walls were 

5.9 ft (1.8 m) high and reinforced with three layers of 2.1-ft (0.63-m)-spaced Tensar® SR80 

geogrid. Two wall configurations were used with the aid of three 4.5 in-(0.114-m)-thick wall 

panels. Known as single and multisegment, these two configurations include two wall panels that 

were 1.8 ft (0.55 m) wide and 5.9 ft (1.8 m) long. The third was 2.3 ft (0.7 m) wide and 5.9 ft 

(1.8 m) long. The reinforcement was attached to the facing panels through elongated holes to 

accommodate any vertical movement of the reinforcements during construction.

Each wall panel for the single configuration was propped vertical into position with struts at the 

1 ft (0.3 m), 3.3 ft (1.0 m), and 5.3 ft (1.6 m) heights from the base. The struts were removed 

after construction. With the multisegment configuration, struts were used at five locations on 

each panel (four near the corners and one at the center) and removed once the backfill height 

reached the top of the panel. Hence, no more than one panel was propped at any one time during 

construction of the multisegment wall.

Lateral pressures were measured using six load cells placed flush with the facing. No surcharge 

loading was applied after strut removal. Consistent with Yang’s observations on a reinforced soil

wall, the lateral pressures on Yogarajah and Saad’s geosynthetic reinforced wall appear to be 

fairly constant with depth below the top 2 ft (0.6 m).(64,78) Also below the top 2 ft (0.6 m), the 

lateral pressures were significantly smaller than the predicted Rankine active values.

METHODS FOR ESTIMATING LATERAL EARTH PRESSURES ON GRS WALL 

FACING

Two methods involving the use of simple equations to estimate the lateral thrust on GRS wall 

facing are presented. They include the procedures developed by Wu, and Soong and 

Koerner.(25,79) Ehrlich and Mitchell developed charts that are quite elegant for estimating the 

maximum tension in a reinforced soil.

(80) This study is not presented further herein for four 

reasons. First, although the tension in the reinforcement can be related to the lateral stress, this 

lateral stress represents the value within a GRS mass rather than on the wall facing. Second, the 

study assumes that each reinforcement layer is responsible for equilibrium with the lateral stress, 

meaning the tensile strength divided by the vertical spacing is equal to the average lateral earth 

pressure within the tributary area of the reinforcement, which is not true for closely spaced GRS.

Third, this method will predict a lateral pressure that increases with depth. Finally, the

compaction equipment must be known a priori.

Wu’s Procedure (2001)

For closely spaced systems, Wu indicated that the primary function of the facing is to prevent 

soil sloughing.(25) It also serves as a construction aid but is not a major load carrying element.

Wu further postulated based on evidence from numerous case studies that the lateral thrust 

against the facing is independent of the wall height.(81) Because the reinforcement effectively 

restrains lateral deformation of the soil, the lateral pressure on the facing is quite small in 

comparison to the lateral stress predicted from Rankine earth pressure theory. Its magnitude 

41
depends on the reinforcement spacing, the soil shear strength parameters, and the rigidity of the 

facing. Calling the magnitude “bin pressure,” Wu proposed that the bin pressure diagram is near 

zero at the reinforcement elevation. It increases with depth below the reinforcement before 

decreasing to near zero at the reinforcement layer below (i.e., at the base of the facing).

However, because the reinforcement may deform slightly and there may be imperfect bonding 

between the soil and reinforcement at the wall face, Wu proposed the bin pressure diagram as 

shown in figure 35.

 

©J.T.H. Wu

1 pcf = 16.02 kg/m3

1 psf = 0.05 kPa

Figure 35. Diagram. Bin pressure diagram.

(25)

The total lateral thrust (Fbin) can be calculated as the area under the bin pressure diagram as 

shown in the equations in figure 36 or figure 37(25):

 

Figure 36. Equation. Total lateral thrust, F subscript bin, as a function of the spacing 

between reinforcements, S subscript v, and peak bin pressure subscript h.

or

 

Figure 37. Equation. Total lateral thrust, F subscript bin, in terms of the coefficient of 

active earth pressure.

FFbbbbbb = 0.72SSvvσσh

σ

FFbbbbbb = 0.72γγKKaa SSvv

2

42
Soong and Koerner’s Procedure (1997)

A similar concept was proposed by Soong and Koerner for GMSE walls where the reinforcement 

is connected to the middle of the back face of the wall facing as shown in figure 38.

(79) While the 

reinforcements stabilize most of the soil mass through interface friction, Soong and Koerner 

postulated that there is a small zone of soil bearing against the wall facing that is not restrained 

by the reinforcement mobilized friction. Assuming that each layer acts independently from those 

above and below, they proposed a lateral earth pressure distribution as shown in figure 39.

 

Source: Elsevier Ltd.

Figure 38. Drawing. GMSE wall configuration showing nonreinforced backfill zones.

(79)

43
 

Source: Elsevier Ltd.

Figure 39. Drawing. Lateral earth pressure distribution against wall facing exerted by the

nonreinforced backfill.

(79)

Soong and Koerner estimated the magnitude of the connection force using the equation in

figure 40:

 

Figure 40. Equation. Connection force F.

This equation produces a thrust that is 31-percent smaller than the equation in figure 37. 

Table 5 shows the lateral thrust on the facing as a function of reinforcement spacing for both 

Wu’s and Soong and Koerner’s methods for a soil with = 125 pcf (2,002 kg/m3

), = 34

degrees and c = 0.

FF = 0.5γγKKaa SSvv

2

γ φ

44
Table 5. Comparison of Wu’s versus Soong and Koerner’s calculated lateral thrusts as a 

function of reinforcement spacing.

(25,79)

Reinforcement Spacing, Sv

(inches)

Lateral Thrust

(lb/ft)

Wu(25) Soong and Koerner(79)

4 2.83 1.96

8 11.3 7.85

12 25.4 17.7

16 45.2 31.4

24 102 70.7

36 229 159

Note: Assumes a backfill with = 125 pcf (2,002 kg/m3), = 34 degrees and c = 0.

1 inch = 2.54 cm

1 lb/ft = 14.6 N/m

COMPARISON OF PREDICTED AND MEASURED LATERAL PRESSURES

When an earth pressure cell is placed at the mid-point between two reinforcement layers on the 

wall facing, the measured pressures should coincide with the pressure at the middle of the 

pressure distribution diagrams.(82) It is estimated that the pressures mid-way between 

reinforcements are about 0.8 h and 0.5 h for Wu’s bin pressure and Soong and Koerner’s 

triangular pressure distributions, respectively.(25,79) A comparison of the calculated and measured 

lateral pressures is provided in table 6 for the case histories described previously and assuming 

no surcharge loading. The following should be noted:

• The measured lateral pressures from Abu-Hejleh include the effects of loading from the 

bridge superstructure.(76) Based on elastic theory (footing width = 12.5 ft (3.8 m) and 

setback = 4.43 ft (1.35 m), it is estimated that the increase in lateral pressures are 

13 percent, 21 percent, 25 percent, and 21 percent of the applied footing stress for the 

12th, 11th, 9th, and 7th geogrid layers, respectively, from the bottom. The applied stress 

on the Founders/Meadow bridge abutment footing was not reported by Abu-Hejleh to 

facilitate a proper comparison.

• Only Mitchell’s lateral pressures obtained at the end of construction and prior to load 

testing (i.e., zero surcharge load) are shown in table 6.

(77) Mitchell’s measured lateral 

pressures at zero load were mostly negative. When and how the zero readings were 

obtained could not be discerned from the report.

• The earth pressure cells for the Yogarajah and Andrawes study were not mid-way 

between reinforcements.(82) However, the pressure distribution does appear constant with 

depth below the top 2 ft (0.6 m). The single segment wall is less representative of a GRS 

type wall with modular block facing elements than the multisegment wall. Nevertheless, 

the comparison between predicted and measured values agrees best in this study.

γ φ

σ σ

45
Table 6. Predicted versus measured lateral pressures on reinforced soil wall facing.

Reference (pcf) (degrees)

Sv

(ft)

0.8 h

(psf)

0.5 h

(psf)

Measured

(psf)

Abu-Hejleh et al. (2003)—12th geogrid 141 39 1.33 34 21 31a

Abu-Hejleh et al. (2003)—11th geogrid North 141 39 1.33 34 21 177a

Abu-Hejleh et al. (2003)—11th geogrid South 141 39 1.33 34 21 163a

Abu-Hejleh et al. (2003)—9th geogrid 141 39 1.33 34 21 188a

Abu-Hejleh et al. (2003)—7th geogrid 141 39 1.33 34 21 334a

Mitchell (2002)—Pier 1—1 ft above base 120 36 24 50 31 -121b

Mitchell (2002)—Pier 1—4.3 ft above base 120 36 24 50 31 -402b

Mitchell (2002)—Pier 1—7 ft above base 120 36 24 50 31 -228b

Mitchell (2002)—Pier 1—9.7 ft above base 120 36 24 50 31 -96b

Mitchell (2002)—Pier 1—13 ft above base 120 36 24 50 31 -152b

Mitchell (2002)—Pier 2—1 ft above base 124 36 16 34 21 -127b

Mitchell (2002)—Pier 2—4.3 ft above base 124 36 16 34 21 -51b

Mitchell (2002)—Pier 2—7 ft above base 124 36 16 34 21 -89b

Mitchell (2002)—Pier 2—9.7 ft above base 124 36 16 34 21 -4b

Mitchell (2002)—Pier 2—13 ft above base 124 36 16 34 21 60

Mitchell (2002)—Pier 3—1 ft above base 121 36 8 17 10 541

Mitchell (2002)—Pier 3—4.3 ft above base 121 36 8 17 10 33

Mitchell (2002)—Pier 3—7 ft above base 121 36 8 17 10 343

Mitchell (2002)—Pier 3—9.7 ft above base 121 36 8 17 10 113

Mitchell (2002)—Pier 3—13 ft above base 121 36 8 17 10 -25b

Mitchell (2002)—Pier 4—1 ft above base 122 36 32 67 42 -99b

Mitchell (2002)—Pier 4—4.3 ft above base 122 36 32 67 42 -222b

Mitchell (2002)—Pier 4—7 ft above base 122 36 32 67 42 -19b

Mitchell (2002)—Pier 4—9.7 ft above base 122 36 32 67 42 -653b

Mitchell (2002)—Pier 4—13 ft above base 122 36 32 67 42 -125b

Yogarajah and Andrawes (1994)—single seg. 104 47 2.07 27 17 13

Yogarajah and Andrawes (1994)—multi-seg. 104 47 2.07 27 17 38 a

Measured lateral pressures from Abu-Hejleh include the effects of loading from the bridge superstructure.

(76)

Lateral pressures before construction of the superstructure were not reported. b Mitchell’s measured lateral pressures at zero load were mostly negative.(77) When and how the zero readings were 

obtained could not be discerned from the report.

1 pcf = 16.02 kg/m3

1 ft = 0.3 m

1 psf = 0.048 kPa

FIELD CRITIQUE OF RANKINE ACTIVE THRUST THEORY

Wu’s and Soong and Koerner’s calculated lateral thrusts are much smaller than the Rankine 

active thrust because they are a function of the reinforcement spacing and not the wall 

height.(25,79) If it were true that the bin pressure were not a function of wall height, GRS walls can 

be safely constructed to very tall heights if desired (e.g., figure 41). If the Grand County wall in 

Colorado illustrated in figure 41 were an externally stabilized wall designed using strictly 

Rankine active earth pressure theory, as in GMSE design, it would have required an enormous 

footing embedment and connection requirement to resist the assumed loads, which is not the case 

in this GRS wall.(2,81) This wall offers the most compelling field evidence that the lateral pressure 

exerted on the GRS facing is much smaller than that derived from classical earth pressure theory.

γ φ σ σ

46
 

Source: R. Barrett

Figure 41. Photo. Grand County, Colorado, wall with a maximum height of 55 ft

(16.8 m).

(83)

LATERAL PRESSURES DUE TO APPLIED LOADS

The lateral pressures on the facing of a GRS wall due to applied loading behind the facing (e.g., 

bridge footing on a GRS abutment) will increase with increasing footing load as seen in 

Mitchell’s study.(77) Using linear elastic theory, the increase in lateral pressure can be 

approximated using Carother’s equation in figure 42.

(84) Assuming an allowable maximum 

applied bearing stress on a GRS abutment of p = 4,000 psf (191.5 kPa), the increase in lateral

pressure ( h) on the CMU blocks can be calculated as shown in figure 42:

 

Figure 42. Equation. Increase in lateral pressure subscript h.

Where and are angles defined as shown in figure 43. Assuming a 4-ft- (1.2-m-) wide footing 

with a 1-ft (.3-m) setback, the increase in lateral pressures on the CMU blocks are highest within 

the top 3 ft (1 m) from the facing as shown in figure 44. The maximum lateral stress was only 

1,150 psf (55.1 kPa), which is 29 percent of the applied load.

∆σ

∆σσh = pp⁄ππ [αα − sin αα cos(αα + 2δδ)]

∆σ

α δ

47
 

1 psf = 0.048 kPa

Figure 43. Drawing. Definition of angles and .

 

1 ft = 0.3 m

1 psf = 0.048 kPa

Figure 44. Chart. Increase in lateral stress due to 4,000-psf (191.5-kPa) vertical stress on a 

4-ft- (1.2-m-) wide footing with a 1-ft (.3-m) setback.

Concrete Pad

α

δ

p = 4000 psf

α δ

0

5

10

15

20

25

30

35

0 500 1000 1500

Depth (ft)

Horizontal Stress Increase (psf)

48
Figure 42 and figure 44 assume a flexible wall that is free to displace laterally. If the wall were 

rigid (i.e., unable to displace laterally), then the pressures in figure 42 and figure 44 should be 

doubled.

SUMMARY

• Lateral stresses in a reinforced soil are higher than those in the same soil unreinforced.

• Just below a critical depth, measured lateral stresses in a reinforced soil or on the facing 

of a reinforced soil wall appear to be constant with depth. However, it should be 

cautioned that this observation is based on very limited reliable data.

• Wu and Soong and Koerner postulated theories that estimate the lateral pressure on the 

face of reinforced soil walls.(25,79) Their magnitudes depend on the reinforcement spacing, 

the soil shear strength parameters, and the soil unit weight only. A constant reinforcement 

spacing will lead to a constant lateral pressure distribution with depth.

• Lateral stresses on the GRS facing will increase if the GRS is loaded vertically, as in the 

case of a bridge footing on a GRS abutment, even with a frictionally connected facing as 

seen in Mitchell’s experiments.(77) For the example above, assuming linear elastic theory 

applies, the lateral pressure was not more than 30 percent of the applied vertical stress.

49

CHAPTER 6. W EQUATION

The W equation is used to estimate the vertical capacity of footings on a GRS wall and was first 

published by Pham in his Ph.D. thesis entitled Investigating Composite Behavior of Geosynthetic 

Reinforced Soil (GRS) Mass.

(85) In this equation, the W term accounts for the fact that an increase 

in the reinforcement strength does not have the same effect as a proportional decrease in the 

reinforcement spacing, as has been traditionally assumed in simplified GMSE design. This is a 

significant difference between GMSE and GRS design. Results of mini-pier experiments 

performed by Adams, Adams et al., and Elton and Patawaran, and plane strain tests on GRS 

performed by Pham, showed that the contributions of the reinforcement spacing to the strength 

of the GRS is much greater than that of the reinforcement strength. (See references 86, 6, 87 

through 91, and 85.) The various terms in the W equation and previous research and theory 

leading to its development are discussed. The reliability of this equation is demonstrated based 

on load test results from the literature.

W EQUATION

In reality, the capacity of a GRS is related to the effects of confinement (due to CISs and the 

lateral restraint offered by the reinforcement internally, and confining stress at the facing), 

reinforcement spacing, strength and stiffness, shape and location of the failure surface, stressstrain behavior of the soil, and degree of mobilization of shear resistance along an assumed 

failure plane. The W equation was derived using limit equilibrium analysis. This type of analysis 

assumes a linear failure surface, assumes a rigid-plastic soil behavior, neglects soilreinforcement stiffness, and assumes full mobilization of shear strength along the failure surface.

Despite the many assumptions that may not reflect reality, limit equilibrium analysis is popular 

among practitioners because it provides a solution that is simple and easy to use.

The W equation can be derived by first considering the Mohr-Coulomb failure envelope and the 

Mohr circle at failure for an unreinforced cohesionless soil (figure 45). 

51
 

Figure 45. Drawing. Mohr circles for unreinforced and reinforced soil.

If the soil friction angle is ,the equation in figure 46 shows that at failure:

Figure 46 . Equation. Major principal stress subscript 1.

σ1 and 3 are the major and minor principal stresses respectively, and Kp equals the Rankine 

passive earth pressure coefficient. If the soil were reinforced, Yang indicated that the friction 

angle of the reinforced soil is approximately the same as the unreinforced soil if no slippage 

occurs between the soil and reinforcement.(64) A GRS is stronger than the unreinforced soil, and 

hence the major principal stress at failure is higher than 1, say 1R in figure 45.

For this to occur, the confining stress in the soil must effectively increase due to the presence of 

the reinforcement as shown in the equation in figure 47:

 

Figure 47. Equation. Confining stress due presence of reinforcement subscript 3R.

The reinforcement restrains lateral movement of the soil. Thus, 3 represents the additional 

confining stress that is imposed due to the reinforcement. Therefore, the capacity of the GRS can 

be expressed as the equation in figure 48:

φ

1 3 3K p 1 sin

1 sin

σ

φ

φ σ σ = −

+ =

σ

σ

σ σ

σ3R = σ3 + ∆σ3

σ

∆ σ

52
Figure 48 . Equation. Capacity of the GRS subscript 1R.

∆σ3 is analogous to a prestress in the reinforced soil equal to the frictional force developed 

between the soil and the reinforcement with a maximum value determined by the tensile strength 

of the reinforcing material. At failure, this prestress will be at its maximum and is related to the 

reinforcement tensile strength per unit wall length (Tf). If it is assumed that one-dimensional 

expansion in the horizontal direction occurs over a tributary area of reinforcement equal to 

spacing Sv per unit wall length, then ∆σ3 can be expressed as shown in figure 49:

(92)

Figure 49. Equation. Additional confining stress as function of reinforcement tensile 

strength divided by the reinforcement spacing subscript 3.

Implied in the equation in figure 49 is that a GRS with reinforcement strength Tf at spacing Sv

will have the same capacity as a GRS with reinforcement strength 2Tf at spacing 2Sv, which has 

been shown to be untrue by Adams, Adams et al., Elton and Patawaran, Ziegler et al., and Pham.

(See References 86, 6, 87 through 91, 93, and 85.) Instead, Sv has a bigger influence on the 

capacity than Tf. This led Pham to propose a modified version of the equation in figure 49 as 

shown in figure 50:

Figure 50. Equation. Modified version of the equation in figure 49 for additional confining 

stress, subscript 3 imposed with the W factor.

W is the factor that amplifies the contribution of Sv to the GRS capacity. The W factor was semiempirically derived and is calculated using the equation in figure 51:

Figure 51. Equation. W factor.

Dmax is the maximum particle size of the GRS backfill used as a normalizing parameter to make 

the exponent dimensionless. Note that the 0.7 factor in figure 51 was theoretically derived while 

the exponent was empirically derived. Using the concept of “average stresses” proposed by 

Ketchart and Wu to estimate the average and maximum forces in a reinforcement, Pham showed,

based on a load-transfer analysis, that the average reinforcement force is about 70 percent of the 

maximum reinforcement tensile strength; hence the 0.7 factor in figure 51.

(62,85) For details on 

this derivation, refer to Pham.

(85)

The exponent of figure 51 is a function of Sv and Dmax. Because the base term is less than unity, 

W, and hence the capacity, increases with decreasing Sv and increasing Dmax. This is logical 

R RK p K p ( ) σ 1 = σ 3 = σ 3 + ∆σ 3

σ

v

f

S

T

∆σ 3 =

∆ σ

v

f

S

T

∆σ 3 = W

∆ σ

6 max 0.7 D

Sv

W =

53
because one would expect a GRS with closer reinforcement spacing and larger maximum 

aggregate size to have a higher capacity.

Therefore, combining the equations in figure 48, figure 50, and figure 51, the capacity of the 

GRS can be expressed as the equation in figure 52:

Figure 52 . Equation. Capacity of the GRS, subscript 1R.

If the soil has cohesion, c, then the equation in figure 52 can be expanded to the equation in 

figure 53:

Figure 53 . Equation. Expansion of the equation in figure 52 for the capacity of the GRS,

subscript 1R, to include cohesion.

According to Pham, for a GRS wall with a dry stacked modular block facing, σ3 equals the 

lateral stress exerted by the facing on the GRS mass, which is bl times D times tan .

(85) bl

equals the bulk unit weight of the facing block, which is the weight of block/volume of the block 

assuming it is not hollow. (For a 7.625- by 7.625- by 15.625-inch (19.368- by 19.368- by 

39.688-cm) CMU block weighing 42 lbs (19.1 kg), bl equals 80 pcf). D equals the depth of the 

facing block perpendicular to the wall face. equals the friction angle between geosynthetic 

reinforcement and the top or bottom surface of the facing block. The equation in figure 53 can 

also be rearranged and solved for the required reinforcement strength (Tf,req) given the loads on 

the particular GRS composite for design.

RELIABILITY OF THE W EQUATION

A database consisting of 19 load tests on GRS from the literature (table 7) was used to examine 

the reliability of the W equation or the equation in figure 52. In addition to the references already 

cited in this synopsis, the table contains tests from Adams et al.(6) A plot of predicted versus 

measured capacities is shown in figure 54. It can be seen that overall, the coefficient of 

determination was 0.853 with a slope of 0.913 for this dataset when the regression line was 

forced through the origin. To illustrate the importance of the W term, a plot of predicted versus 

measured capacities is shown in figure 55 with the W term eliminated from the equation in 

figure 53. It is apparent that the capacities are severely overpredicted with a less than desirable 

fit. This validates the necessity of the W term in the equation in figure 53.

p

v

D f

S

R K

S

T v

( 0.7 ) 6 max σ 1 = σ 3 +

σ

p p

v

D f

S

R K c K

S

T v

( 0.7 ) 2 6 max σ 1 = σ 3 + +

σ

γ δ γ

γ

δ

54
Table 7. Prediction data for large-scale tests.

No. Test

Test Results Facing Reinforcement Aggregate

Geometry Reference

qult,emp

(kPa)

qult calc

(kPa) c(kPa)

Block 

Type

Sv

(m)

Tf

(kN/m)

Dmax

(m)

c 

(kPa) test(degrees) Method Kp

1 GSGC 2 3,400 2,182 34 N/A 0.2 70 0.0330 70 50 TX 7.55 PS

85

(Pham (2009))

2 GSGC 3 2,040 1,615 34 N/A 0.4 140 0.0330 70 50 TX 7.55 PS

3 GSGC 4 1,785 936 34 N/A 0.4 70 0.0330 70 50 TX 7.55 PS

4 GSGC 5 2,034 1,925 0 None 0.2 70 0.0330 70 50 TX 7.55 PS

5 Elton 1 230 133 0 None 0.15 9 0.0127 28 40 DS 4.60 Cylindrical

91

(Elton and 

Patawaran 

(2005))

6 Elton 2 129 33 0 None 0.3 9 0.0127 28 40 DS 4.60 Cylindrical

7 Elton 3 306 207 0 None 0.2 14 0.0127 28 40 DS 4.60 Cylindrical

8 Elton 4 292 222 0 None 0.2 15 0.0127 28 40 DS 4.60 Cylindrical

9 Elton 5 402 0 None 0.2 19 0.0127 28 40 DS 4.60 Cylindrical

10 Elton 6 397 296 0 None 0.2 20 0.0127 28 40 DS 4.60 Cylindrical

11 Elton 7 459 370 0 None 0.2 25 0.0127 28 40 DS 4.60 Cylindrical

12 NCHRP 1 420 275 0.97 CMU 0.2 21 0.0254 0 36.5 LSDS 3.94 PS 94

(Wu et al. (2006)) 13 NCHRP 2 850 908 0.97 CMU 0.2 70 0.0254 0 36.5 LSDS 3.94 PS

14 Defiance 1 542 580 0.97 CMU 0.2 35 0.0127 0 50.7 LSDS 7.84 Column 89

(Adams et al.

15 Defiance 2 1,213 1,153 0.97 CMU 0.2 70 0.0127 0 50.7 LSDS 7.84 Column (2007))

16 Vegas 1,008 1,231 2.35 SRW 0.2 35 0.0254 70 50 TX 7.55 Column

88

(Adams et al. 

(2002))

17 MP A 225 284 0 None 0.6 70 0.0254 0 53.5 LSDS 9.20 Column 89

(Adams et al. 

(2007))

18 MP B 170 671 0 None 0.4 70 0.0254 0 53.5 LSDS 9.20 Column

19 MP C 460 634 0 None 0.2 21 0.0254 0 53.5 LSDS 9.20 Column

GSGC =Generic Geosynthetic Soil Composite 

NCHRP = National Cooperative Highway Research Program

MP = Mini-Pier 

TX = Triaxial 

DS = Direct Shear

LSDS = Large-Scale Direct Shear 

PS = Plane Strain

1 kPA = 20.89 psf

1 m = 3.28 ft

1 kN/m = 68.6 lb/ft

σ φ φ

55
 

1 kPA = 20.89 psf

Figure 54. Graph. Predicted versus measured capacity of large-scale GRS tests using the 

equation in figure 52.

 

1 kPA = 20.89 psf

Figure 55. Graph. Predicted versus measured capacity of large-scale GRS tests using the 

equation in figure 52 without the W term.

56
CHAPTER 7. REDUCTION FACTORS

Current design methods for GRS walls and abutments (e.g., AASHTO guideline, FHWA NHI 

manual, and NCMA manual) typically stipulate that design strength of geosynthetic 

reinforcement should be determined by applying reduction factors for installation damage, creep, 

and durability to the ultimate strength of geosynthetic reinforcement.(1,2,3) The equation in 

figure 56 expresses the allowable reinforcement design strength, Ta:

Figure 56 . Equation. Allowable reinforcement design strength, T subscript a.

Tult is the ultimate wide-width strip tensile strength of the geosynthetic (in accordance with 

ASTM D4595) based on the minimum average roll value (MARV) for the product. (MARV is 

commonly defined as the strength that is two standard deviations below the mean tensile 

strength). FS is an overall safety factor to account for uncertainty. RF is a combined factor to 

account for geosynthetic strength loss during the wall design life and is equal to RFID times RFCR

times RFD, where RFID is a reduction factor for installation damage, RFCR is a reduction factor 

for creep, and RFD is a reduction factor for chemical and biological degradation.

In the FHWA NHI manual, RFID = 1.1 to 3.0, RFCR = 1.6 to 5.0, and RFD = 1.1 to 2.0 are 

recommended, resulting in K = 3 to 50 percent.(2) Different state transportation departments have 

adopted some variations for the reduction factors and safety factor, hence the K-value range.

CDOT, for example, has adopted the following values: the allowable design strength is on the 

order of 10 to 24 percent of MARV for preapproved products and 5 percent of MARV for not

preapproved products. This type of practice has practically excluded the use of all geotextiles in 

reinforced soil wall applications, even though many GRS walls and abutments constructed with 

geotextiles as reinforcement have performed successfully.

This synopsis addresses the issue of geosynthetic reinforcement reduction factors. The available 

research on the use of cumulative reduction factors is reviewed, and current practices are 

discussed. In addition, brief comments about the K-stiffness method, a working stress design 

method intended to address long-term behavior of geosynthetic reinforcement of reinforced soil 

walls, are provided.

RESEARCH AND CASE HISTORIES

Degradation

Aging of buried geosynthetics by processes such as hydrolysis, oxidation, and abrasion may 

result in long-term strength loss of geosynthetic reinforcement. Elias evaluated 24 geosynthetics 

from 12 retrieval sites, and confirmed that little, if any, chemical degradation had occurred in the 

geosynthetic reinforcement in those full-scale structures, some of which were up to 25 years

old.(95,96) This has been further confirmed by Allen and Bathurst.(97) They indicated that the 

greatest contributors to strength loss and reduced wall performance (i.e., increased deformation) 

ult

ID CR D

ult ult a K T

FS RF RF RF

T

FS RF

T

T = = ×

× × × = ×

57
for the geosynthetic reinforcement products in use today are installation damage and possibly 

creep.

Installation Damage

Numerous field studies regarding the installation survivability of geosynthetics have been 

performed. They have shown that the level of installation damage depends on polymer type, 

manufacturing method, and geosynthetic coating.(96,98) The level of damage has also been shown 

to be affected by the weight, type, and number of passes of the construction and compaction 

equipment, the graduation, angularity, and condition of the fill material and the lift thickness.

(See references 99 through 102.)

Allen and Bathurst show that installation damage has limited impact on the initial working stress 

performance of geosynthetic walls.(103) For most geosynthetics used as reinforcement (i.e., 

woven geotextiles and geogrids), their load-strain-time behavior is not significantly affected by 

installation damage at typical or even relatively high working strains for the levels of installation 

damage observed in full-scale walls. The data provided by Allen and Bathurst indicate that 

installation damage does not severely affect modulus, if at all, until damage levels become quite 

high for woven geotextiles and geogrids.(103)

Bathurst et al. describe how to compute bias statistics from project-specific installation damage 

trials for use in reliability-based design for the reinforcement rupture limit state, or by using data 

from multiple sources for LRFD calibration.(104) A database of results from field installation 

damage trials on 103 different geosynthetic products was collected from 20 different sources.

The computed reduction factors for installation damage (RFID) of their study confirm earlier 

recommendations by Elias that woven and nonwoven geotextiles with mass per unit area less

than 7.96 oz/yd2 (270 g/m2) should not be used in combination with Type 1 soil (with 

D50 > 19 mm (.74 inches).(95) For a certain soil type, there were detectable differences in the 

calculated RFID values depending on the geosynthetic type.

Hufenus et al. conducted a series of field installation tests and concluded that, in applications 

where only the tensile strength at relatively low elongations is relevant, the effects of moderate 

installation damage is very limited, and the factor RFID can be designated as very close to unity 

(1.0 to 1.1) based on product-specific test data.(105)

Allen and Bathurst investigated the combined effect of polymeric creep and installation damage 

using a database of constant sustained load (creep) data for both undamaged and installation 

damaged geosynthetic specimens.(106) They concluded that multiplication of creep reduction 

(RFCR) and installation damage factors (RFID) may be conservative and hence results in errors on 

the safe side for current ASD practice. Greenwood came to the same conclusion based on 

stepped isothermal creep-rupture tests performed on a polyester geosynthetic material in 

undamaged and damaged states.(107)

Creep

Major types of geosynthetics used in GMSE and GRS structures are polypropylene, 

polyethylene, and polyester. These geosynthetics, manufactured with various types of polymers,

are creep-sensitive. Stress level, polymer type, manufacturing method, and temperature have 

58
been known to affect the creep potential of a geosynthetic material. In general, polypropylene 

and polyethylene exhibit larger creep deformation than polyester and polyvinyl alcohol under 

otherwise identical conditions.

As noted above, the allowable reinforcement tensile load employed in current design guidelines

is determined by applying a safety factor and a combination of reduction factors to a limiting 

strength determined from short- and/or long-term laboratory tests. There tests are conducted by 

applying uniaxial tensile forces directly to the geosynthetic reinforcement (in a confined or 

unconfined condition) without regard to the soil-geosynthetic interaction behavior. It is important 

to point out that conducting uniaxial creep tests in the confinement of soil does not mean 

soil-geosynthetic interaction is accounted for. Wu indicates that long-term creep behavior of 

geosynthetic reinforcement in a reinforced soil structure must be determined by allowing soilgeosynthetic interaction to take place in a manner mimicking the field conditions.(108)

McGowan et al. developed a fairly sophisticated uniaxial tension test device to measure creep 

behavior of geotextiles under soil confinement.(109) Wu and Ling et al. developed a simplified 

confined creep test method, in which a constant sustained tensile force is applied to a membraneconfined geosynthetic specimen (without inducing artificial soil-geosynthetic interface 

friction).(110,111) Various geotextiles under different confining pressures have been tested. The 

results indicated that pressure confinement gave various degrees of improvement in creep 

behavior for different geotextiles. The greatest improvement was for needle-punched nonwoven 

geotextiles, while the improvement in woven geotextiles and geogrids was negligible.

Boyle manufactured a plane strain test device similar to that developed at Massachusetts Institute 

of Technology by Abramento and Whittle.

(112,113) In the test, a geosynthetic specimen was 

embedded in soil and subjected to a plane strain loading condition. The loads at both ends of the 

geosynthetic material were measured. Creep tests using sand as confining soil indicated that the 

geosynthetic material experienced stress relaxation. After creep deformation had diminished, the 

force in the geosynthetic material would reduce with time. The practical implication is that such 

GRS structures will have increasing safety margins as time progresses.

Crouse and Wu synthesized measured field behavior of seven reinforced soil walls that had been 

monitored for extended periods of time for assessment of their long-term performance 

characteristics.(114) The GRS walls represented a variety of wall types using granular backfill. 

The walls were: (1) the Glenwood Canyon wall; (2) the Tanque Verde—Wrightown—Pantano 

Roads project wall; (3) the Norwegian Geotechnical Institute (NGI) project, NGI wall; (4) the 

Japan Railway Test Embankment project, JR wall; (5) the Highbury Avenue, London Ontario 

project, Highbury wall; (6) the FHWA Algonquin wall; and (7) the Seattle Preload Fill project, 

Seattle wall. (See References 115 through 124, 7, 125, 54, 126, and 127.) The maximum creep 

strains in the reinforcements measured by strain gauges for all the walls were less than 

1.5 percent, and the creep strain rate in all cases decreased with time in that there was a linear 

relationship between log-[creep rate] and log-[time]. Crouse and Wu also proposed a creep 

equation for predicting long-term creep deformation of GRS walls.(114)

Allen and Bathurst analyzed the long-term creep data from 10 full-scale geosynthetic wall case 

histories.(97) Post-construction, long-term wall face deformation data show that geosynthetic wall 

face deformations, if the wall is properly designed, will generally be less than 0.98 to 1.17 inches 

59
(25 to 30 mm) during the first year of service and less than 1.37 inches (35 mm) during the 

design lifetime for walls lower than 42.64 ft (13 m). Allen and Bathurst et al. studied 

reinforcement strains measured in geosynthetic-reinforced walls and slopes. (128,129) The 

maximum strains were on the order of 1 to 2 percent or less.

Helwany and Wu performed finite element analysis on two 9.84 (3-m) high geosyntheticreinforced retaining walls.(130) The walls were identical in every respect except that one was with 

a clayey backfill and the other with a granular backfill. In the clay-backfill wall, the maximum 

strain in the geosynthetic reinforcement increased by 3.5 percent from the end of construction to 

15 years later. In the granular-backfill wall, the increase in maximum strain over the same time 

period was negligible. Note that the difference occurs in spite of comparable levels of load for 

the two walls. The analysis strongly suggests that the backfill played a very important role in 

creep deformation of a soil-geosynthetic composite. The finding was supported by studies 

conducted by Li and Rowe, Skinner and Rowe, Rowe and Taechakumthorn, Bergado and 

Teerawattanasuk, Liu and Won, Liu et al., and Li et al. (See references 131 through 139.)

Recognizing that the interaction between soil and geosynthetic reinforcement must be used as the 

basis for a rational design, Wu and Helwany developed a soil-geosynthetic long-term 

performance test, also referred to as the Soil-Geosynthetic Interactive Performance (SGIP)

test.(140) The test has two important features. First, the stresses applied to the soil are transferred

to the geosynthetic material in a manner similar to the typical load transfer mechanism in GRS 

structures (i.e., loads in reinforcement are transferred from soil through soil-geosynthetic 

interface bonding to geosynthetic material, rather than being applied directly to geosynthetic

material). Second, both the soil and geosynthetic material are allowed to deform in an interactive

manner under plane strain conditions. Two carefully conducted long-term performance tests, one 

using a clayey backfill and the other a granular backfill, have been reported. An element test on 

the geosynthetic material alone underestimated the maximum strain by 250 percent in the claybackfill test, and overestimated the maximum strain by 400 percent in the sand-backfill test. It is 

noteworthy that creep deformation essentially ceased within 100 min after the sand-backfill test 

began; whereas the clay-backfill test experienced creep deformation over the entire test period 

(18 days), at which time shear failure occurred in the soil.

The SGIP test subsequently evolved to accommodate taller specimens, as shown in figure 57.

(141)

A number of geosynthetic/granular road base composites have been tested, with a few under 

elevated temperatures to accelerate creep of the geosynthetic materials. For a nonwoven 

geotextile embedded in a road base material (with 20 percent of fines, prepared at 95-percent

R.C. and 2-percent wet-of-optimum moisture) subject to an average surcharge of 15 psi

(103 kPa) and in a 125 oF (52 oC) environment (note that the creep rate of the geotextile was 

about 150 times faster at 125 oF (52 oC) than at the ambient temperature, as measured in a series 

of long-term creep tests), creep deformation of the soil-geotextile composite was very small and 

decreased rapidly with time. Creep deformation ceased completely in 12 days and gave an 

accumulative average strain of 0.58 percent.

60
 

©K. Ketchart and J.T.H. Wu

1 mm = 0.039 inches

Figure 57. Illustration. Schematic diagram of a SGIP test.(141)

DISCUSSION

This discussion addresses five points regarding long-term design consideration of GRS walls and 

abutments: (1) long-term degradation, (2) construction damage, (3) creep, (4) current practice, 

and (5) the K-stiffness method. 

Long-Term Degradation

Studies conducted by Elias and Allen and Bathurst have indicated that long-term degradation of 

geosynthetics in reinforcement applications during exposure to the in-soil environment appear to 

be very small.(95,97)

A full-scale experiment conducted by the PWRI of Japan is enlightening in terms of long-term 

degradation. The experiment was to examine the failure mechanism of a 19.7-ft- (6.0-m-) high 

GRS wall. The wall face was segmental concrete blocks, and the backfill was a sandy soil 

reinforced with six layers of 11.5-ft- (3.5-m-) long polymer grid. The reinforcement inside the 

wall was severed at selected sections after the wall was constructed (for details, see chapter 3).

The maximum horizontal movement due to cutting of the reinforcements was nearly zero until 

the cut was approximately 0.2H from the wall face, at which time the movement increased from 

30 mm to 40 mm. The result of the PWRI experiment reveals the fundamental concept of GRS—

the reinforcement serves not as tiebacks but as improvements to stiffness and strength properties.

It also suggests that long-term degradation of the reinforcement is not a design issue. The cutting 

of reinforcements can be viewed as an extremely severe state of degradation in that the 

reinforcement has been degraded into small pieces and is not continuous.

61
Installation Damage

Most study on installation damage indicates that such damage does not severely affect loadstrain behavior for woven geotextiles and geogrids, if at all, until damage levels become quite 

high. In applications where only the tensile strength at relatively low elongations is relevant, the 

effects of moderate installation damage is very limited. It has been suggested that RFID should be 

designated as being close to unity, on the order of 1.0 to 1.1.

(105)

Allen and Bathurst provided strong evidence that installation damage would have little, if any, 

effect on creep strains and rates for typical levels of installation damage in full-scale 

structures.(106) They stated that “in many cases, installation damage will have a negligible effect 

on the long-term strength at working stress levels (i.e., the geosynthetic behaves as if it is not 

damaged).”

Creep

Most geosynthetic materials are susceptible to creep under sustained loads. Stress level, polymer 

type, manufacturing method, and temperature have been known to affect the creep potential of a 

geosynthetic material. However, it can be drastically misleading to evaluate the creep potential 

of geosynthetic reinforcement based on tests performed by applying a sustained tensile force to 

geosynthetic specimens, as has been done by current design guides. If the confining soil has a 

tendency to deform faster than the geosynthetic reinforcement along its direction of elongation, 

the geosynthetic material will impose a restraining effect on the time-dependent deformation of 

the soil through the interface bonding forces. Conversely, if the geosynthetic reinforcement in 

isolation tends to deform faster than the confining soil, then the confining soil will restrain creep

deformation of the reinforcement. This restraining effect is a direct result of soil-reinforcement 

interaction wherein redistribution of stresses in the confining soil and changes in tensile forces in 

the reinforcement occur over time in an interactive manner. Field-measured data reveal that 

geosynthetic creep deformation is not a design issue when well-compacted granular fill is used as 

backfill in the reinforced soil zone.

Current Practice

In an attempt to accommodate the effects of installation damage, creep, and durability, current 

practice applies a combined reduction factor to short-term reinforcement strength in design. The 

combined factor is obtained by multiplying individual reduction factors for each effect, which is 

unwarranted for closely spaced GRS. First, soil-geosynthetic interaction is critical to the 

susceptibility of geosynthetic creep; the current design approach is based on tests in which forces 

are applied directly to geosynthetic materials without regard to the soil. Second, creep is a 

deformation problem; the current design approach uses a somewhat arbitrary reduction factor for 

creep and treats it as a strength problem. Third, installation damage has been shown to have little 

effect on long-term creep, i.e., there is no compounding effect of installation damage and creep; 

also RFD for geosynthetic materials has been found to be near unity.(106, 105)

When well-compacted granular fill is employed, a single safety factor can be used to account for 

long-term effects, uncertainty, and ductility. It is an indisputable fact that creep of geosynthetic 

reinforcement in a GRS is strongly affected by the time-dependent behavior of the fill. When 

62
time-dependent properties of a given fill material are in question, a laboratory test similar to the 

SGIP test (see figure 57) may be conducted to evaluate potential creep “deformation” of a GRS 

wall or abutment.

63

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