NBSIR 73-222

Investigation of the Skyline Plaza

Collapse in Fairfax County, Virginia

E. V. Leyendecker and S. G. Fattal

Center for Building Technology

Institute for Applied Technology

National Bureau of Standards

Washington, D. C. 20234

June 1973

Final Report

Prepared for Occupational Safety and Health Administration

Department of Labor

Washington, D. C. 20210
r
NBSIR 73-222

ReF. 9¥

INVESTIGATION OF THE SKYLINE PLAZA

COLLAPSE IN FAIRFAX COUNTY, VIRGINIA

E. V. Leyendecker and S. G. Fattal

Center for Building Technology

Institute for Applied Technology

National Bureau of Standards

Washington, D. C. 20234

June 1973

Final Report

Prepared for Occupational Safety and Health Administration

Department of Labor

Washington, D. C. 20210

U. S. DEPARTMENT OF COMMERCE, Frederick B. Dent. Secretary

NATIONAL BUREAU OF STANDARDS. Richard W. Roberts. Director

ABSTRACT

The collapse of the Skyline Plaza apartment building

A-4 has been studied by using information contained in case

records of the Occupational Safety and Health Administration

(OSHA) , U.S. Department of Labor and obtained from on-site

inspections by investigators from the National Bureau of

Standards

.

Non-compliance with OSHA construction standards has been

identified with regard to formwork, field-cured concrete

specimens and crane installation. Specifically, the construction procedures did not comply with standards for the

removal of supporting forms. It is concluded that premature

removal of forms was a contributing factor to the collapse in

building A-4.

An analysis of the 23rd-floor slab indicates that its

most likely mode of failure was in shear around one or more

columns in section 3 of the floor slab. The strength of the

23rd-floor slab on the day of collapse has been estimated

to be at a level that removal of shoring could have produced

shear failure in the slab.

TABLi; OP CO.'JTIiNTS

1^ a g c

1. Introduction ^

1 . 1 Backirround ^

1.2 Objective and Scope ^

2. Description of the Structures and Their Collapse ^

2.1 Introduction ^

2.2 Buildinp, A- 4

^

2.2.1 General ^

2.2.2 Pormwork and Interior Views ^

2.2.3 lirection Cranes

2.3 Parkin?^ Garage

2.4 Crane .No. 2 2 5

3. Concrete and Reinforcement in Building A- 4 27

3.1 Concrete 27

3.1.1 Standard Cylinder Tests 27

3.1.2 Field Cores 27

3.1.3 Pstiinate of Quality of Concrete as Delivered 23

3.1.4 Pstiiiiates of Concrete Strengths on the Day of 30

Collapse

3.2 Reinforcing Steel . 33

4. Structural Investigation of Failure Conditions in 35

Building A4

4.1 Introduction 35

4.2 Controlling Regulations and Basis of Design 36

4.3 Analytical Procedure 36

4.4 Discussion of Results 40

4.5 Probable Mode of Collapse in Building A- 4 30
5. Summary of Findings 52

5.1 Mode of Failure 52

5.2 Non-Compliance with OSIIA Regulations 52

6. References 56

4
THE SKYLINE PLAZA COLLAPSE

IN FAIRFAX COUNTY, VIRGINIA

by

Edgar V. Leyendecker and S. George Fattal*

1. INTRODUCTION

1.1 Background

The Skyline Center Complex located near Bailey's

Crossroads, Fairfax County, Virginia is a development planned

to contain eight apartment buildings, six office buildings,

a hotel, and a shopping center [19], Two apartment buildings

which have been completed are shown in figure 1.1. A pair of

apartment buildings (similar in appearance to those in figure

1.1) and an adjoining parking and lobby structure were under

construction and included the structures which collapsed on

March 2, 19 73.

The apartment buildings under construction are shown in

figure 1.2, an aerial photograph taken at about 11:00 a.m.

on Friday, March 2, 1973, from an altitude of 5000 ft [18].

Several hours later, at about 2:30 p.m., a portion of the

building shown in the top of figure 1.2 collapsed. The

collapsed portion of the building was located approximately

* Dr . Edgar V. Leyendecker and Dr. S. George Fattal are

structural research engineers with the Structures Section;

Structures, Materials, and Life Safety Division; Center for

Building Technology; Institute of Applied Technology;

National Bureau of Standards.

^ Numbers in brackets refer to references.

1
under the slab area being cast (the tall building in figure 1.2)

stories plus four basement stories. The collapse progressed j hori - , zontally from the tall building to include the entire parking

garage area and stopped at the building partially shown at the

bottom of figure 1.2. The full extent of the collapse is shown

in figure 1.5. It has been reported by OSHA that fourteen

construction workers were known to have been killed, four in

the garage and ten in the tower, and another 34 injured in

the incident. ; -.J n

Witliin a few hours of the incident an inspection team

from the Occupational Safety and Health Administration (OStiA)

,

Department of Labor, began arriving at the site to begin an

investigation into the collapse.

1.2 Objective and Scope : ^.|

' '

On Monday, March 5, OSlly\ requested the technical assistance of the Center for Building Technology of the National

Bureau of Standards (NBS) with respect to the collapse.

Tiie National Bureau of Standards was requested to ascertain,

if possible, the cause of the incident, to assist OSHA in

determining whether tliere had been non-compliance with

OSHA standards ("Safety and Health Regulations for Construction"

[10 j) and whether such non-conpliance contributed to the

collapse.

OSHA compliance officers were on tire site from March 2

througli March. 16 collecting m.aterial for their case records

and extended vertically for the

2
[18]. During this time, personnel from the NBS made numerous

site inspections.

The NBS investigators used data gathered during on-site

inspections, OSHA case records, structural and architectural

drawings, shop drawings, and structural computations in

preparing this report. Where reference is made in this report

to employee statements, such statements are part of OSHA case

records examined by NBS.

3
2. DESCRIPTION OP THE STRUCTURES AND THEIR COLLAPSE

2.1 Introduction

Three structures may be identified in figure 1.2;

building A- 4 at the top, building A- 5 partially shown at tlie

bottom, and a parking garage in between the two buildings. The

collapse started in building A-4 and progressed vertically to

the ground and horizontally to include the entire parking garage.

The collapse stopped at building A-5 which was structurally

isolated from tlie parking garage. The two affected structures

are shown in a plan view in figure 2.1 and are discussed in

subsequent sections.

The three structures under construction were designed

under the Fairfax County Building Code Ordinance [5] which

incorporates by reference the provisions of the American

Concrete Institute's Building Code Requirements for Reinforced

Concrete (ACI 318-63) [6]. The building design is discussed

further in section 4, Structural Investigation of Failure

Conditions in Building A-4. Applicable Federal safety regulations are described in the Safety and Healtli Regulations

for Construction [10] which incorporates by reference the

American National Standards Institute's Safety Requirements

for Concrete Construction and Masonry Work, ANSI AlO. 9-1970 [2].

2 . 2 Building A4

2.2.1 General

Building A-4 was of reinforced concrete flat plate

construction supported on a 4-ft thick foundation mat. The

4
completed structure was to have 26 stories of apartments,

plus a penthouse and a four-story basement (designated B-1,

B-2, B-3, and B-4). The typical story height from the first

story up was 9 ft 0 in from top of slab to top of slab.

Floor slabs \siere 8 inches thick.

The basement story heights varied in order to suit

mechanical equipment layout with a total basement height of

about 40 ft as measured from the top of the foundation mat

(level B-4) to the top of the first floor slab. The basement

floors were about the same elevation as the corresponding

floors in the garage, although there was no actual floor B-3

in building A-4. That is, the lowest "story" in the basement

consisted of stories B-4 and B-3. The first floor slab was

at the same elevation as the roof of the garage which was to

become a landscaped area.

The plan view shown in figure 2.1 is that of the

22nd story and is typical for the 1st through 26th stories

in the column layout. The column layout remained essentially

the same through the height of the building (1st story and up)

varying only in size, reinforcing steel, and concrete strength.

There are eight shear walls in the structure. These are

designated as A through H in the drawing. The configuration of

the shear walls above the 20th story is shown in figure 2.1.

For the 20th story and below, the portion shown dashed became

part of the shear wall. The floor slab thickness of 8 in was

constant through the height of the building. A 1/2-in expansion

joint separated- the building into two parts at grid line H.

5
Normal weight aggregate concrete v;as used in the columns.

The specified column concrete strength varied; it was 5000 psi

from the foundation mat to the 7th floor, 4000 psi from the 7th

to the 17th floor, and 3000 psi above the 17th floor. The slabs

used lightweiglit aggregate (coarse aggregate only) concrete with

a specified strength of 3000 psi. Inspection of two floors,

the 24th and 10th, indicated that the lightweight aggregate

concrete floor slab passed through the columns at these floors.

A typical floor of the building was poured in four sections;

the progress of construction at the time of the collapse is

shown in figure 2.2 [18]. The actual sizes of the pour sections

are sliown in figure 2.1. Note that the sections are not equal

in size and configuration. According to statements provided

to OSIIA, a normal construction rate was one section a day.

This rate of construction v;ould permit the completion of one

floor per week and allow an extra day for weather variation.

In actual practice, this rate was not always maintained as

indicated in figure 2.3.

Figure 2.3 is a plot of daily temperature versus calendar

date for January 28 through March 2, the day of the collapse.

The maximum, minimum, and average temperatures are those

recorded at Washington National Airport, the official weather

bureau recording station nearest the construction site. Where

available, the daily temperature range recorded at tlie construction

site (from job records) is also plotted. The dates for pouring

the various sections of the 21st through 24tli floors are indicated

6
on the drawing. Note t]iat the casting dates shown tended to

fall on tlie days with the higlier temperatures.

The general appearance of the building A- 4 as viewed from

the southeast after the collapse is shown in figure 2.4. A

number of the floors and one of the columns is identified for

later reference. Note the absence of a floor slab at level B-3

(in accordance with the plans). Note also the column corbels at

floor levels B-3, B - 2 , B-1, and 1. These corbels supported one

edge of the parking garage slabs. Level B-4 of the garage was

on grade.

A partial view of the north face of building A- 4 is

shown in figure 2.5. A number of columns and floors are

identified. The collapse extended between shear wall II and

column 33 on the south face, a distance of about 65 ft (refer

to figure 2.1). On the north face the collapse extended

between columns 12 and 17, a distance of about 104 ft. Note

that for most floors the slab between columns 16 and 17 did

not sever from the main building but is sagging for a one-story

lieight. Below the 20th floor, the horizontal extent of the

collapse is as discussed above. Above the 20th floor, the

failure zone extended slightly to the west and a greater

amount to the east.

Figures 2.6 through 2.8 are closeup viev-zs of the east

end of the failure zone. These figures require little comment

at this time; however, it should be noted that the formwork

is clearly visible in figure 2.6.

7
A general view of the west end of the failure zone is

shown in figure 2.9 along with closeup views in figures 2.10

through 2.12. These figures also require little comment at

this time; however, it should be noted that the formwork is

clearly visible in figure 2.10.

On the night of March 4 the remaining portion of the

building to the east of the failure zone was completelydemolished, with the result shown in figures 2.13 and 2.14.

2.2.2 Formwork and Interior Views

Access was not gained to building A- 4 by OSHA personnel

until March 5 when the building appeared as in figure 2.14.

Most of the photographs used in this discussion of the building

interior were taken on or after that date. Between March 2 and

March 5, additional shoring was placed in some parts of the

building

.

Figure 2.15 shows the locations of the formwork on

March 2, shortly after the collapse. Full formwork was in

place on the 24th and 23rd stories in sections 1 and 2 (refer

to figure 2.1 for location). Some formwork may be seen on

the 22nd story in sections 2 and 3. In section 2 this is

primarily the area around the material elevator shown toward

the western end of the south elevation in figure 2.15.

Investigation by OSHA personnel on March 5 indicates that the

only formwork not stripped on the 22nd story in section 2

was around that elevator. Employee statements indicate that

this area was not stripped in order to prevent lumber from

8
falling on bricklayers working below. Formwork may also be

seen in section 3 of the 22nd story in figure 2.15. However,

closer examination of the 22nd story in figure 2.10 indicates

that formwork was stripped in the center portion of the 22nd

story.

Full formwork was also in place under section 4 on the

22nd and 23rd stories. No reshoring can be seen under section 4

below the 22nd floor in figure 2.15 or in figures 2.4, 2.5,

and 2.6. Some employee statements indicate, however, tliat

reshoring was present below pour 4 on the 21st story. Other

statements contradict this.

There is conflict also among employee statements as to the

formwork which was in place in section 3. Employees interviewed

by OSMA agreed that the 23rd-story forms supporting the recently

cast 24th floor were in place. However, employee statements

indicate that the 22nd-story forms were: (1) entirely removed,

(2) partially removed, or (3) not removed. The same contradictions were indicated for reshoring on the 21st story. The

angle of the aerial photograph in figure 1.1 prevents examination

of conditions of formwork in section 3. The location of formwork

will be discussed further after a description of the interior

of the building.

The formwork on the 24th story is shown in figure 2.16.

Several columns liave been identified for reference (see figure

2.1 for location). These slab forms had not been completely

erected. A schematic of this formwork is shown in figure 2.17

as derived from OSHA case records [18].

9
The formwork in figure 2.17 is typical for the floor

slabs with design noted by the concrete contractor as being

based on the publication, Formwork for Concrete , SP-4 [11].

Formwork sheets submitted to Fairfax County [18] contained

no mention of lateral bracing for the form system. However,

a limited amount of bracing may be seen in figure 2.16 and

some subsequent figures. OSIIA regulations (ANSI-A10.9,

2 Sections 6.3.2 and 8.1.5) [2] require a lateral bracing system

capable of resisting a lateral force of at least 2 percent of

the dead load of the slab.

As may be seen in figure 2.16, the stringers were placed

east-west and the joists were placed north-south. Employee

statements indicate that stringers and joists were usually

16 ft in length although some were as short as 6 to 8 ft.

The erection procedure is described by reference to figure

2.16. Stringers were erected with a shore under each end

(shores 1 and 5). These shores were attached to the stringers

by a metal or wooden plate. Shores 2, 3, and 4 were then

inserted in basket-like sockets which were fastened to the

stringers. Employee statements indicate that diagonal bracing

was installed in the north-south direction on every 4th shore

at approximately 16-ft intervals (also see figure 2.20). The

joists were placed over the stringers at about 16-in intervals

(figure 2.17 indicates the spacing could be 20 in). Joists

were toenailed to the stringers at about 8 -ft intervals. Plywood

1

For relationship between OSHA regulations and ANSI-A10.9

see section 2.1.

10
sheathing was then placed on the joists with the long dimension

parallel to t'ne stringers.

Forms were removed by knocking out shores 1, 2, 4, and

5 in figure 2.16. Shore no. 3 would then be knocked down

whic]\ would allow the stringer to fall. The joists and plywood

would then be pulled down.

Usually, the center portion of the slab would be stripped

first. Approximately 10 stringers would be pulled down

before reshoring began. The area stripped prior to reshoring

was approximately 20 ft wide by 35 ft long.

During the .\'BS site inspection, a considerable portion

of the lumber used for the remaining forms in tlie 24th story

was found to be in poor condition or out of plumb. For instance,

the top pliotograph in figure 2.18 shows a stringer with a battered

end directly over the shore. T]ie second photograph illustrates

a shore which is out of plumb and has a vertical crack near tlie

top. OSIiA regulations (ANSI-A10.9, sections 8.1.24 and 8.1.25)

[2] require correction of both of these conditions prior to

placing concrete (which was not yet placed on these forms).

The portion of section 3 of the 24th-story slab which

remained standing is shown in figure 2.19. This portion of

the slab was several hours old at the time of the collapse.

Apparently, the only activity on the slab at the time of collapse

was work being done by concrete finishers. One concrete

finisher working near the stair well (around columns 65 and

66) indicated that a large deflection was seen in the middle

of section 3. The statements of numerous employees agree

11
on this point. ViO statements provided an exact location of the

sag. However, the middle of section 3 is about midway between

columns 6 7 and 35 (refer to figure 2.1 for location)..

The formwork on the 23rd story is shown in figure 2.20.

Tliese forms supported tlic 24th floor which had been cast on

February 28 (section 1) and March 1 (section 2). Note the

number of sliores out of plumb and the location of lateral

bracing. The method of attaching the brace to the stringer

may be seen in figure 2.20. The method of attaching t]ie brace

to the floor is shown in figure 2.21. A portion of a nominal

3x4 was nailed to a piece of plywood which was in turn nailed

to the floor. A plywood plate was occasionally used to nail

the brace to the nominal 5 x 4. Note tiie absence of such a plate

in figure 2.21. Occasionally, the brace was installed with the

plate, in other ijistanccs, tliere was no positive attachment

of the brace to tjie floor. In one case, the plywood stop was

not nailed to the floor, providing no lateral resistance.

As in the case of the 2 4th story, tiie 2 3rd story lumber

which vsras apparently damaged was reused. Examples of battered

shores are sliown in figure 2.22. ^'ote the large reduction in

bearing area. As in the case of tlie 24th story, OSIIA regulations

(A;\'SI - AlO . 9 , sections 8.1.24 and 3.1.25) [2] required correction

of these conditions prior to placing concrete.

Reshoring in section 2 of tlie 2 2nd story as of !Iarch 6 is sliown

in figure 2.23. OSIIA investigators liavc indicated that the reshoring

on March 5 was not as extensive as shovni in figure 2.23.

12
Apparently on March S the reshores were Drincipally in place

on the balconies, as they were on March 3, the day of the

collapse.

Careful examination of figure 2.23 reveals that the

center ceiling area of section 3 (area beyond column 59) was

bare, indicating that some stripping had occurred in section 3

(see also figure 2.10). Although there is conflict in their

statements, a number of workmen that escaped the building by way

of the stairs in section 4 (see figure 2.1) have indicated

that tlie 22nd story was either partially or entirely stripped

3 in section 3. Notes on the engineer's structural drawings

call for two full stories of shoring and one story of reshoring

under a slab being cast.

A view of a portion of section 2 of the 21st story is

shown in figure 2.24. Column 59 is at the expansion joint

(Grid l! in figure 2.1). Note the cables on the floor which

were installed after the collapse to anchor to section 2 the

portion of section 3 which was still standing. No reshoring

was present on this floor in sections 1 or 2.

A number of workmen have stated that at least some

reshoring v;as present in sections 3 and 4 of the 21st story.

The statement of one employee ini.licates tliat he was worlcing in

section 3 installing reshores. At about 2:00 p.m. he could hear

the stripping crew working on the next floor up (22nd floor).

He could hear the noise caused by the falling lumber. At

about tills same time, the workman's reshores began falling so

3

Skyline Center Structural Drawings, iVcihe, Black and Jeffries,

Architects, dated Marcli 16 , 1972 , with revisions tlirougli

October 2, 1972 - Drawing S5.

13
that just before the accident there were no reshores in section

3 (this point is discussed later).

As shown in figure 2.25, the 20th floor was completelybare of reshoring. This was reportedly the case for sections

1 through 4. Note the presence of shear walls A and B.

.

These walls were not present (the design did not require them)

above the 20th story. Note also the erection crane v;hich is barely

visible to the left of the designation for shear v;all A.

A view of the failure section at the end of the 20th floor

is shown in figure 2.26. ^

:

=

It has been indicated in this report that there is conflict

in the various employees' statements as to what formwork was in

place at the time of the incident. However, it should be noted

that conditions changed throughout the day as construction

proceeded; that is, what may have existed at one time may not

have existed a short while later. Nevertheless, it is necessary

to determine the location of forms and reshores at the time of

the incident. \

The estimated location of forms and reshores is given

in figure 2.27 and discussed below. Physical evidence examined by OSHA and NBS investigators as well as employees' statements indicate that all formwork was in place for all poured

sections on the 23rd and 24th stories. Based on employees'

statements and careful observation of photographs taken on the

day of the collapse and over the following days, it is concluded

that all formwork had been removed from sections 1 and 2 of the

22nd story on the day of collapse. The one exception is in

14
the limited area near the materials elevator on the soutii face

of tlie building. It appears that very little reshoring

had been done. Lxainination of photographs indicates tliat

fornivsork was in place in section 4 of the 22nd story.

Section 3 is discussed later. On the 21st story there was

no resliorina in sections 1 and 2. One statement indicates that

the reshoring had been removed from section 2 on the day of the

accident. Although some statements indicate the presence

of reshores in section 4 of the 21st floor, this is not supported

by photographic evidence available (figures 2.4, 2.5, 2.6, and

2.15). The photographs at the angle at which they were taken

would not show reshoring, if any, located at some distance away

from the edge of the slab. It is considered possible that there

was some reshoring present although probably a sinall amount.

Section 3 of the 21st story is discussed below^.

One workman indicated that, at the time of the incident,

he was placing reshores in section 3 of the 21st story and that

some reshores were present when he started work. Prior to the

incident all of the reshores fell out (except those in the balcony

areas). This is consistent with what could occur if the forms had

been removed in tlie story above. For example, consider figure

2.28 (a) which shows a frame with forms in the 3rd and 4th

stories and reshores in the 2nd story. The fifth floor has

been freshly poured and carries none of its own weight which

must be distributed to the 4th, 5rd, and 2nd floors. The

exact distribution of loads depends on a number of factors, one

of which is the construction history.

15
If the forms are removed in the 3rd story as in figure

2.28 (b) , then the 4th floor carries all of its own weight

in addition to that of the fifth floor. The 2nd and 3rd

floors are now relieved of the previous load from the 4th

and 5th floors. The 3rd floor deflection will decrease due

to this reduction of load. A possible smaller contribution

to the decrease in the 3rd floor deflection is due to the

downward deflection of the 4th floor under the added load

causing the third story column to bend outward and the 3rd floor

to bend upward as schematically illustrated in figure 2.28 (c)

.

The net result of the removal of forms causing an upward

deflection of the 3rd floor could quite conceivably cause the

2nd story reshores to fall out with the final result shown in

figure 2.28 (d)

.

Based on the above simplified analogy plus previously

mentioned photographs (figures 2.10, 2.15, 2.23) it is concluded

that form stripping was in progress in the 3rd section of the

22nd story. The photographic evidence indicates that at least

the central portion of the section was being stripped heading

toward section 4.

2.2.3 Erection Cranes

Two climbing cranes were used in the construction of

building A-4. Crane no. 1 was located in section 2 and crane

no. 2 was located in section 4. The terminology used in the

discussion of the cranes is shown in figure 2.29 [22],

16
Botli cranes were initially used in a free-standing

position mounted to the foundation mat. Later, as construction

proceeded, the cranes were supported by floors within the

building. Hxcept for certain installation requirements,

crane dimensions, and capacity, the two cranes were quite

similar. Descriptive data are contained in figures 2.30

[21J and 2.31 [22] for the two cranes.

The standard climbing crane consists of the tower base

section (which contains the climbing machinery, no. 2 in

figure 2.29), four standard tower sections (no. 3 in figure

2.29), and a tower slipring section (no. 4 in figure 2.29).

The four standard tower sections are equal in length and may

be interchanged. The ring gear assembly (no. 5 in figure 2.29)

which forms the connection between the stationary tower and

the rotating portion of the crane fits on top of the slip

ring section.

A schematic of the climbing sequence is shown in figure

2.32 [22]. In picture 1 the crane is installed in its initial

position on the foundation mat. A lower support, the climbing

frame, is installed a maximum of 36 ft from the mat. The

climbing ladder is suspended from the climbing frame (the

ladder is about 38 ft long). An upper support, consisting of

four corner clamps, is attached a specified distance above the

lo\\?er support. This required distance was specified as 21* -4"

for crane no. 2 (figure 2.31) and 18 '-4" for crane no. 1 (figure 2.30). In the actual use, a two-story distance of 18' -0"

was used for crane no. 1 and apparently for crane no . 2 . In

17
picture 2, the tower climbs up the ladder to the first support

using the climbing mechanism included in the tower base section.

In picture 3 a third support is added, the climbing ladder

is moved up to a second climbing frame and the crane is ready

for climbing. In the fourth picture the crane is in position

and the lower frame can be removed.

The upper support transfers only horizontal forces

to the building, while the lower support transfers vertical

and horizontal forces. The maximum free standing height of

the crane above the upper support as measured from the bottom

of the jib is 70 ft 2 in for crane no. 1 (figure 2.30) and

81 ft for crane no. 2 (figure 2.31).

Crane no. 1 was supported by the climbing frame, the lower

support, on the 20th floor as shown in figure 2.33. The upper

support was provided on the 22nd floor as shown in figure 2.34.

This total distance is 18 ft, slightly less than the manufacturer'

specified 18 ft 4 in. A typical corner clamp detail on the

22nd floor is shown in figure 2.35. Note the cracked slab

where the clamp is attached.

Crane no. 2 was reported to be on either the 14th or

17th floor with most reports indicating the 17th. Reported

conditions of shoring under both cranes varied from no shores

to shores all the way to the foundation mat. Crane no. 1

has been positively located with its base support on the

20th floor. Shoring under that crane is less certain. OSHA

investigators have indicated that it was not shored to the

foundation until after the collapse.

18
In order to determine more accurately the location of

crane no. 2 it is necessary to look at previous construction

photos. Such photos were available from the OSHA file [18].

However, the exact dates of the photos are not known. It is

known that construction records indicate that crane no. 1 was

last raised on February 26 and crane no. 2 on February 7.

Figure 2.36 contains two construction photos reported

as taken around February 10. This date is in error as will

be shown. Since the lower floors are not visible it is not

possible to tell from figure 2.36 where the cranes are located.

Figure 2.37 shows a photograph taken on March 6 that can be

used to establish the location of the 15th floor in figure 2.36.

A certain pattern of "color" may be seen in the doors and door

frames at the elevator landings. A comparison of the patterns

between figures 2.36 and 2.37 (taken March 6) establishes the

location of the 15th floor in figure 2.36 (It will be seen

later that the same door pattern existed in a set of photos

taken between figure 2.36 and 2.37).

Examination of figure 2.38 establishes the lower crane

support on the 10th floor and establishes that the crane,

as installed, could operate through a ten-story height of

building. It should be noted that the crane had marginal

clearance to cast the 20th floor although it appears that it

could. The photos were not clear enough to determine shoring

conditions under the crane.

Close examination of figure 2.38 shows openings in section

4 of the 20th floor forms. Therefore, the 20th floor had not

19
been cast at the time of the photo. Reference to figure

2.3 indicates that this section was cast on February 1.

Construction records indicate that framing for the 20th floor

forms, which is visible in figure 2.38, was installed on

January 27. Based on this evidence it appears that figures

2.36 and 2.38 were taken between January 2 7 and February 1.

Therefore the photograph was taken prior to February 7 which

was the last date crane no. 2 was raised.

Figures 2.39 through 2.42 were reported taken on February 11 or February 18, both dates after crane no. 2 was

raised. The floor numbers can be established simply by

counting at the east end of the building. Note the location

of the 15th floor with respect to the top elevator landing.

This location is in agreement with photos taken before and

after figure 2.39. The elevator door pattern can be clearly

seen in figure 2.40. Note also that at the time of the photograph, crane no. 1 is based on the 16th floor (crane no. 1 was

raised on February 26) and crane no. 2 is based on the 14th

floor. Since crane no. 2 could work through 10 stories

(figure 2.38) it could have been used to cast section 3 of the

24th floor while based on the 14th floor.

The various tower sections of the crane are identified

in figure 2.41. Note that five standard tower sections

were used instead of the usual four (figure 2.31 and reference

22). Assuming the upper and lower supports were two stories

apart as has been stated, the crane had a free standing

height of about 99 ft which exceeds the maximum free height

20

'
oi" 81 ft shown ill figure 2.31. It is not known if any

special precautions were taken to allow the difference between

tlic actual and the recommended installation.

Examination of fij^ures 2.40 and 2.42 indicates that either

the crane shoring or the formwork for the concrete closure slab of

the crane opening was in place in the third through tliirteenth

stories on the day of the photograph. Conditions below the third

story are not visible in the photograph.

The date of the pJiotograph (reported as February 11 or

18) can be cliecked by looking at figures 2 . 39 , 2 . 42 , and

2.2. Close examination indicates that none of the 22nd floor

had been cast at the time of the photograph (openings can be seen

in the slab forms). Since figure 2.2 indicates that section

1 on the 22nd floor was cast on February 14, it is established

that the photograph was taken before that date. Due to the

construction progress shown in the figure it can be stated that

section 2 of the 21st floor which was cast February 6 (figure

2.2) was in place. Based on this evidence it appears that the

photograph was taken betv/een February 6 and February 14.

Although the forms are in place for sections 3 and 4

of the 21st floor, it cannot be positively stated that they have

been cast. Therefore, on the basis of these photographs (figures

2.39 through 2.42) it cannot be conclusively stated that the

crane which was raised on February 7 was based on the 14th floor.

However, based on figures 2.36 and 2.38 the crane no. 2 was

attached to the 10th floor between January 26 and February 1.

Since this crane was raised only once after February 1, it must

have been located on the 14th floor at the time of the collapse.

21
It has been established by physical examination that

crane no. 1 was based on the 20th floor. Examination of

photographs and contractor's records have been used to

conclude that crane no. 2 was based on the 14th floor.

Crane no. 2 was located in the failure zone. The

position of the crane after the collapse will be discussed

after consideration of the parking garage.

2.3 Parking Garage

The parking garage was a flat plate structure of posttensioned unbonded concrete construction. The completed

structure was to have four stories of parking with a landscaped

roof. The B-4, or lowest level, was a s lab-on-grade . The

parking garage slabs were at approximately the same elevations

as the corresponding floors in the building A-4 basement.

The plan view is shown in figure 2.1. A typical panel

was 28 ft by 30 ft with an 8-in slab. The columns were supported

on footings. The story height was 9 ft for stories B-4, B-3,

B-2, and varied from 10 to 14 ft for story B-1. Normal weight

aggregate concrete with a design compressive strength of

4,000 psi was specified throughout the structure.

Slabs B-4 and B-3 had been cast and slab B-2 had been

cast to the extent shown in figures 2.1 and 1.2. Note that

in the vicinity of grids lines BB to FF and grid line 1 slab

B-3 was placed on compacted fill. Compacted fill was apparently

placed around the three columns, BB-1, DD-1, and EE-1 in story

B-4. The footings for these columns were a few feet higher

than the rest of the column footings.

22
The parking garage, as seen from building A- 4 is shown

in figure 2.43. A number of columns have been identified for

reference. At the time of the collapse, slab B-2 had been

cast to the extent shown. The rest of B-2 had been formed

(some of the forms had been cleared at the time of the

photograph). A closeup of several of the columns in figure

2.43 are shown in figure 2.44.

The locations of slabs B-3 and B-2 are shown in figure 2.

on column DD-4. Note the virtual lack of shear cones at the

slab levels on most of the columns in that figure.

The garage slabs generally came "straight" down except

near the slab edges. In the latter cases the columns usuallyfailed in bending. Figure 2.45 shows a view of slab B-2 with

the edge between columns JJ-5 and JJ-8 visible on the left.

The view of this edge from the south is shown in figure 2.46.

Note that the columns failed at the B-3 level. Formwork can

be seen lying on top of the slab B-3. These forms apparently

were in place under slab B-2.

Column KK-6 may be seen in figure 2.47. Once again the

shear cone usually associated with a slab punching failure

is not visible in the photograph. Note the ruptured reinforcing bars at the base of the column. These are shown more

clearly in figure 2.48 (but not for column KK-6).

23
The east edge of the garage was supported by columns as

shown in figure 2.49. The northern edge was simply supported by

corbels on the columns. These corbels are visible in figures 2.49

and 2.50. The pool deck level (corresponding with the first

floor of building A-4) was to rest on corbels on columns 25,

26, 27 and 33-37. Slab B-1 was to rest on corbels on columns

25, 26, 27, 29, 31-38, and shear wall H. Slab B-2 was supported

by corbels on columns 25 , 26 , 27 , 29 , 31-38 , and shear wall II. Slab B-3 was supported by corbels on columns 25, 26, 27, 29,

31-36, and shear wall II. The loss of columns between shear wall H and column 33

meant the loss of two columns (31 and 32) which provided support

for slabs B-2 and B-3. The total span between column 33 and

shear wall M was about 65 ft.

The west edge of the garage was supported by bearing on

a wall and framing into a ramp as shown in figure 2.51. Slab

B-2 may be seen resting on tlie portion of B-3 cast on a slab-ongrade (vicinity of column Eii-l). More detail of the ramp framing

can be seen in figure .2.52. The ramp was of reinforced concrete

construction. The west wall south of the ramp can be seen in

figure 2.53 and with more detail in figure 2.54. The formwork

and strand which had been i)ositioned for slab B-2 can be seen

in figure 2.53. y\ portion of a sliear cone is visible on column

KK-3 in the same figure.

The soijith edge of the garage was supported by columns which

were independent of building A- 5. This was in contrast to the

corbel support system on the north edge.

24
The progress of formwork and stressing in the parking ^araqe

lias not been established at the time of this writing. Tlicse

factors must be determined prior to analysis of the garage since

they liave a direct bearing on the behavior of the structure.

This unbonded pos t- tensioned concrete construction sucli as

used in the parking garage, is a common form of construction,

however, consideration of its failure was beyond the scope of

this report.

2.4 Crane No. 2

At the time of the collapse, crane No. 2 was reported idle

with the jib pointed towards the garage. The exact orientation

is not known. A distant view of the tower sections of the crane

after the collapse may be seen in figures 2.49 and 2.51 (also

2.5). The tower top and hoist machinery (see figure 2.29) may

be seen near column DD-2 in figure 2.55 (Note the partial shear

cones on column DD-2). Based on an exam.ination of the photographs

it appears (refer to figure 2.1) that the crane toxver lines up

with tJie original position of the crane base (in plan view) and

passes almost over columns 32 and DD-3, ending up with the top

as shown in figures 2.55 and 2.56. This is a total distance of

125 to 130 ft. This distance compares to a total tower height

(base section, 5 standard sections, and 1 slipring section)

of about 125 ft. The implication here is that the tow^er fell

while rotating about its original plan position. IIov;ever,

careful inspection of figures 2.54, 2.55, and 2.5 shows the

jib is lying alongside the tower sections. The sequence of

25
falling which would permit a jib initially pointing over the

garage (as reported by employees) to assume its final position

has not been determined.

26
3. CONCRETE AND REINFORCEMENT IN BUILDING A4

3.1 Concrete

3.1.1 Standard Cylinder Tests

The standard (ASTM C39) cylinder test results for compressive

strengths available at the time of this writing, are shown in

figure 3.1 for the 17th and 21st through 24th floors [18]. Both

7-day and 28-day results are shown for section 3, the failure area.

The results for section 3 indicate concrete meeting the strength

requirements called for by the structural engineer (3000 psi for

the slabs and columns). The results of standard cylinder tests,

however, do not reflect the effect of field curing conditions

or the strength of the concrete in place. Field-cured cylinders

were not made as called for by OSIIA regulations (ANSI A10.9,

section 6.4.7) [2]. It should also be noted that the test results

for the 22nd floor, section 2, and the 23rd floor, section 1, had

strengths less than those normally obtained for other floors

at 7 days. Both of these sections were outside the failure zone.

3.1.2 Field Cores

In order to evaluate the strength of the concrete in the

building the portion of the slabs in section 3 between grid lines

H and I (figure 2.1) were cored. Four-inch diameter cores were

obtained from the structure in the 17th and 21st through 23rd

floors. The results of compressive and split cylinder tests

on the cores are shown in figure 3.2 [17]. The 17th floor was

sampled because crane no. 2 was originally thought to be

27
supported on tliis floor, although in section 2.2.3 it was

concluded that the tower was based on the 14tli floor. The other

floors were sampled because of their proxinity to where the

failure started.

3.1.3 Estimates of Quality of Concrete as Delivered

The concrete quality as indicated by the standard 6-in

dianeter cylinders was satisfactory for the 22nd and 23rd

floors in section 3 (figure 3.1). The field cores were used

to checl; on tlic quality of concrete as delivered. This

comparison is made by converting tlie average values of 4-in

diameter core strength to 6-in standard cylinder values.

There has been conflicting in forinat ion on whether strengtlis

derived from cores are greater or lower than those from standard

cylinders [3, 4, 8, and 16]. The strengtli of cores obtained

from mass concrete (such as dams) frequently are greater than

strengths derived from standard cylinders. However, Hloen [4],

using 6-in thick slabs with normal v;eight aggregate concrete,

has shown that 4-i]i diameter cores, such as obtained from building

A- 4, have lower strengtlis tlian standard 6-in diameter cylinders.

Earlier Campbell and Tobin [8] obtained similar results using

lightweight aggregate concrete and 4-in cores obtained from

12 -in thick slabs. The data froivi Bloem is reproduced in figure

3.3 to show tiie relationships obtained. Campbell and Tobin data

are similar but arc not shown because of insufficient data at

ear ly ages

.

lUoem tested cores from botli well -cured and poorly -cured

slabs. His well-cured slab was sprayed with a curing compound

28
as soon as the water sheen had disappeared. Later the slab

was covered with wet burlap and a plastic sheet. The burlap

was kept wet and in place for 14 days. At that time it was

removed and the slab raised to permit air circulation. The

poorly-cured slab was left uncovered after pouring. Three

days later the forms were stripped and the slab was raised from

the floor. The environmental conditions in the period between

placement and coring of the slabs ranged in temperature from

about 60 to 90° F and in relative humidity from about 25 to

90 percent. The field curing conditions for each floor in

building A- 4 are not known.

Estimates were obtained for a range of standard 6-in

diameter cylinders by converting from the age at time of test

to 28-day 6-in cylinder strength using the curves of figure 3.3.

The calculations rounded to the nearest 5 psi follow:

17th floor: well-cured slab 2780/0.98 (fraction @ 66 days)

= 2835 psi

poorly-cured slab 2780/0.72 (fraction @ 66 days)

= 3860 psi

21st floor: v/ell-cured slab 2470/0 .90 (fraction @ 38 days)

= 2745 psi

poorly-cured slab 2470/0.72 (fraction @ 38 days)

= 3430 psi

22nd floor: well-cured slab 1960/0.82 (fraction @ 25 days)

= 2390 psi

poorly-cured slab 1960/0.68 (fraction @ 25 days)

= 2880 psi

29
23rd floor: well-cured slab 2295/0.73 (fraction @ 16 days)

= 3145 psi

poorly-cured slab 2295/0.62 (fraction § 16 days)

= 3700 psi

Within the limitation of the scatter observed in the core

test results (see figure 3.2), the above data, when compared

to the standard 6-in cylinders (figure 3.1) indicate that

the concrete quality as delivered was generally acceptable

although concrete strength of the 22nd floor slab appears to

be low. The foregoing is not a measure of the strength of

in-situ concrete which is discussed in the next section.

3.1,4 Estimates of Concrete Strengths on the Day of Collapse

The many factors influencing concrete strength have been

discussed elsewhere [3, 14, 16, and 23] and are not repeated

here. However, it should be pointed out that for the same

concrete:

(1) Strength will decrease with a deficiency in curing

moisture after the initial set of the concrete.

(2) Strength potential will increase with sustained low

temperatures (above freezing) , although strength will

be less at the usual test periods (7 and 28 days).

(3) Strength development will accelerate with high

temperatures, although the eventual maximum strength

will be less than in (2).

. (4) Calcium chloride will accelerate the rate of strength

development although it will not prevent freezing.

30
A set of curves have been drawn in figure 3.4 from data

obtained by Klieger [14]. These curves, although obtained from

normal weight aggregate concrete, indicate the effect of

curing temperatures and 2 percent calcium chloride on concrete

compressive strength development. The 73°F curing curve without

calcium chloride represents strength gain under standard

conditions. Note the accelerating effect on strength development of 2 percent calcium chloride. For example, a concrete

cured at 55°F with 2 percent calcium chloride is equivalent

in strength (during the first 28 days) to the same concrete

cured at 73°F. Although not used in all floors, 2 percent

calcium chloride was used for the 17th and 21st through 24th

floor slabs

.

The cores are used as a measure of in-situ strength. The

Klieger data require the use of 28-day strengths. The core

strengths for the 22nd and 23rd floors are converted to

28-day strengths by using figure 3.3. The basic core data

are multiplied by the ratio of the 28-day fraction to the

fraction at the date tested. These calculations rounded to the

nearest 10 psi follow:

22nd floor: well-cured slab 1960x0.85/0.82 (fraction @ 25 days)

= 2030 psi

poorly-cured slab 1960x0.69/0.68 (fraction § 25 days)

= 1990 psi

23rd floor: well-cured slab 2 295x0.85/0.73 (fraction @ 16 days)

= 2670 psi

poorly-cured slab 2295x0.69/0.62 (fraction @ 16 days)

= 2550 psi

31
In order to convert these strengths into estimates at

the time of collapse the curing temperature should be considered.

The average temperature history has been shown in figure 3.5

(enlarged from figure 2.3). The average air temperature was

42°F for the 22nd floor and 45° for the 23rd floor (temperatures

as recorded at National Airport). Estimates of strength using

the Klieger data for 2 percent calcium chloride in figure 3.4

(and interpolating between temperature curves) yields the

following results rounded to the nearest 10 psi:

22nd floor: well-cured slab 2030x0.66 (fraction @ 10 days)

= 1340 psi

;- poorly-cured slab 1990x0.66 (fraction @ 10 days)

= 1310 psi

23rd floor: well-cured slab 2670x0.45 (fraction @ 4 days)

> = 1200 psi

poorly-cured slab 2550x0.45 (fraction @ 4 days)

i-,:^^"'

' = 1150 psi

In the prediction of concrete strength it is recognized

that the average in-situ strength of the floor slabs in

question may be different because of non-uniform curing temperatures, effects of coring and size of core, scatter in test

results in seemingly identical core specimens (see figure 3.2),

chilling factor of the wind and approximations introduced by

application of test results from independent sources of study.

Consideration of these factors, including quantitative

32
assessment whenever possible, indicates a ran^e in the order of

+ 20 percent on tlie average concrete strength estimate.

The structural engineer's calculations for evaluation

of shear strengtli for the structure requires the use of a

splitting ratio [6] for the lightweight concrete. The

structural engineer's design calculations (page 26A, floor

slab design for building A-4) provided data indicating that

a splitting ratio of 6.0 could be used. The splitting

ratio is equal to the ratio of splitting tensile strength

to the square root of the compressive cylinder strength

(figure 5.2). The core data in figure 3.2 indicate a splitting

factor of 6.5 to 7.0.

3.2 Reinforcing Steel

Examination of the reinforcing steel in the area bounded

by columns 66-68-83-86, indicates that the steel called for by the

structural engineer was included in the reinforcing steel shop

drawings. The structural engineer generally called for no .

4

bars in the slab. Slabs were not available for examination

to check if steel was actually placed in accordance with shop

drawings . A total of 12 reinforcing bars were tested, 6 of

these were removed from floor slabs of the collapsed section

of the building and 6 were tested from unused reinforcement

steel at the construction site. The results of tliese tests

are shown in figure 3.6 [17].

The structural engineer generally called for reinforcing

steel with a 60,000 psi yield point and meeting the requirements

33
of ASTM A-615. The steel tested satisfies the requirements

of that specification.
4. STRUCTURAL INVESTIGATION OF FAILURE CONDITIONS

IN BUILDING A4

4.1 Introduction

A structural investigation of the 22nd and 23rd floors of

the Skyline Plaza apartment building A-4 in the region where

the collapse occurred was conducted. The investigation consisted

of:

(1) determination of internal forces in slab,

beam, and column elements of the structural

assembly in accordance with the principles

of elastic analysis ;

(2) evaluation of the ability of structural

elements to resist previously determined

forces

.

The three-dimensional elastic analysis was performed using

the finite element analysis program known by the acronym SAP

[22A] and the computer facilities at the National Bureau of

Standards. The simulated model of the structural assembly

shown in figure 4.2 consists of three-dimensional beam elements

simulating beams and columns of the system and quadrilateral

plate bending elements simulating the floor slab. The beam

element properties are discussed in the SAP user manual while

the plate element properties are described in reference [9],

The evaluation of the capacity of structural elements

was based on the provisions of the ACI 318-71 Code [7] and on

procedures from published analytical and experimental research

[12, 13, 13A, 15].

35
4.2 Control liiig Regulations and Basis of Design

The basic national standard governing the design of

reinforced concrete buildings is the Building Code Requirements

for Reinforced Concrete by the American Concrete Institute,

ACI 518 [6,7], New editions of the ACI Code are issued periodically to provide for tlie advancement in the state of the art

resulting from research and professional experience. The

latest edition of ACI 318 was issued in 1971 [7] and the one

prior to that, in 1963 [6].

In building design the applicable building code is the

latest edition available at the time of tlie design. In the

case of tlie Skyline Plaza apartment building, the design

sliould comply with tlie requirements of the 1956 edition of

Building Code Ordinance (amended in 1971) of Fairfax County,

Virginia [5] , which incorporates the design provisions of

ACI 318-63.

It is noted that for certain areas of the floor the

spacing of columns is sucli that in order to comply with

section 2101(e) 2 of ACI 318-63, the slab thickness would

have to be greater than 8 in. Section 3 of the floor slab,

as identified in figure 2.1 contains such areas.

4.3 Analytical Procedure

Whereas the design of the structure was governed by

ACI 318-63, the analysis that follows is based upon ACI 318-71

v/hich represents the present state of the art. The calculations

36
of shearing stresses were made in accordance with the procedure

described in section 11.13.2 of the Commentary to the ACI 318-71.

The latter is essentially based on the results of experimental

studies by Hanson and Hanson [13] of slabs supported by square

columns. In addition, consideration was given to the effect

of column rectangulari ty on shear capacity as reported by

Hawkins, Fallsen and Hinojosa [13A] . For distribution of

loads between shored slabs consideration was given to findings

by Grundy and Kabaila [12], Nielsen [15], and observations of

field measurements of high-rise flat slab buildings related by

Agarwal and Gardner in an unpublished report.

Three finite element analyses were made for the slab

assembly in the region of collapse. The approximate area

of collapse is shown in figure 4.1. All three analyses were

based on the grid model shown in figure 4.2.

The accuracy of a finite element analysis is generally

improved as the number of elements is increased and as the

elements become more square. In constructing the grid, a finer mesh was used in the region west of the crane tower

opening, where failure was assumed to have been initiated.

The boundaries of the rectangular grid are approximately

defined by the column line 18-73-72-79-28 on the east and

column line 11-63-64-90-91-35 on the west. This area encompasses

the failure region and extends about one column line beyond

its boundaries on each side.

The slab element thickness was 8 inches as called for in

the contract drawings. The supporting column elements were

37
rigidly attached to tlie slab elements at tlie top and were

fixed at their base. The columns above the floor being analyzed

were also rigidly attached to this slab at their base with

the tops hinged. Columns designated by C. and 32 in the contract

drawings were each simulated by a rigid frame consisting of

two columns and a stiff beam.

The grid layout in figure 4.2 is different from the slab

in the contract drawings in certain aspects. The balcony slabs

were assumed extending to the centerlines of the end columns.

Likewise the boundary of the slab at north and south fascia's

between balconies was assumed extending to the centerlines of

the flanking columns, eliminating, in effect, the stepped

portion of the slab between a balcony and an adjacent fascia.

Wherever possible, rectangular or square plate elements v/ere

used to construct the grid. In order to make column centroids

coincident with nodal points of the grid a few trapezoidal

plate elements were also introduced. By using a finer mesh

one could develop a model which would more precisely represent

the actual assembly. However, such a refinement is not

justified because it would not significantly affect the

results [ 7A]

.

Each of the three finite element analyses had a particular

purpose : .' - -

Case 1 : Some of the employee statements indicated

that the 22nd story forms were entirely removed at

the time of collapse (see section 2.2.2). This analysis

of tlie 23rd floor system is based on the premise that

38
the 23rd floor slab was not shored. The loading consisted

of the weight of the newly poured slab of the 24th floor

(80 pounds per square foot) , the weight of the formwork

on the 23rd floor (5 pounds per square foot) and the

weight of the 23rd floor (80 pounds per square foot).

Thus, a total uniform load of 165 pounds per square

foot was placed on the portion of the slab directly

below the poured section of the 24th floor (section 3)

and a load of 85 pounds per square foot was placed

elsewhere. The modulus of elasticity used for the

concrete slab of the 23rd floor was calculated in accordance

with ACT 318-71 using the estimated average concrete

strength of 1200 psi (see section 3.1.4), and an

approximate density of 120 pounds per cubic foot (see

table 3.2). A similar calculation was made for the

elastic moduli of the columns. For stiffness calculations, the moment of inertia was assumed to be that

of the gross section of the concrete.

Case 2 : To check the influence of estimated concrete

strength on the analytical results, Case 2 assumed the

concrete to have attained its full 28-day design strength

of 3000 psi. The 23rd-floor system was analyzed using

elastic moduli calculated as in Case 1 except that design

concrete strength was used.

Case 3 : Some of the employee statements indicated that

removal of shoring under the 23rd-floor slab in the central

corridor area of the building was in progress on tlie day of

39
the collapse (see section 2.2.2). A loading simulating the

'^condition of partial removal of shoring was used. This

Case is similar to Case 1 except the loads on the 23rd

' floor slab were reduced outside the region bounded by

grid lines 7-320, 320-330, 330-16, and 16-7 as indicated

in figure 4.2. To obtain these reduced loads, the loads

from case 1 were multiplied by a factor which was based

on the assumption that 22nd and 23rd floor slabs shared

the loads on the 23rd floor in proportion to their

respective elastic moduli. The modulus of elasticity of

the 23rd floor slab was the same as was used in Case 1.

The modulus of elasticity of the 22nd floor slab was

calculated on the same basis as described in Case 1

using the average estimated concrete strength at the time

of collapse of 1340 psi for the 22nd floor slab (see

section 3.1.4). '

4.4 Discussion of Results

Figures 4.3 through 4.5 show the analytical results for

cases 1 through 3. Moments M (rotation about x-axis) and

My (rotation about y-axis) are designated by horizontal and

vertical lines, respectively. They are shown plotted in pairs

(and approximately to scale) at the centroids of slab elements.

The numerical entries are moment magnitudes at the centroids

and are expressed in units of foot-kips per linear foot of

slab. Moments producing compression in the top of the slab

are positive

.
FLsures 4.6 througli 4.9 show moment diagrams developed

for column strips taken from the grid of figure 4.2. These

strips were investigated for their adequacy to resist moments

obtained by analysis. Negative moments were investigated

at tlie face of the columns according to ACT 513-71 . The slab was also investigated at column locations for

its adequacy in shear. Shearing stresses were determined

following the procedure described in section 11.13.2 of

the commentary to the ACI 318-71 code. The applicable

equations are given belovv for convenience. Figure 4.10 shows the

critical section for shear around the periphery of a rectangular

column and other parameters used in the following equations:

y X

J = ^(b + d)-^ ^ d-^(b->-d) ^ d(h-^d) (b + d)^

X 6 6 2

J

- d(h-^d)-^ ^ d-^(h-Hd) ^ d(b-Kd) (h-^d)^

- g

- ^ 2

3 vHTa:

^ ^ ^ 2,|b+d"

a = 1

- y 1 + 3 ^b+d

A = 2d (b+h+2d)

41
where

V = maximum shear stress in the slab

max

\\

' X = bending moment about x-axis transmitted to column

My. = bending moment about y-axis transmitted to column

b,h = sectional dimensions of rectangular column

d = distance from centroid of tensile reinforcement

to bottom face of slab

a a = moment reduction factors defined above

J^,Jy,= properties of slab section as defined above

V = vertical load transmitted to column

A = area of critical slab section peripheral to column

as defined above

According to ACI 318-71, the limiting value for shear stress

is 4 ^

f

^' for normal weight concrete (section 11.13.2) and 85

percent of that value or 3.4^ f^' for "sand-lightweight"

concrete (11.3.2) such as used in the floor slabs. Critical

shear in the slab will occur when the maximum shear stress

determined by equation (4.1) exceeds 3.4^ f^' or,

w^-^i/fT • •

• ^'-'^

Alternately, expression (4.2) may be solved to give the

minimum concrete strength required to prevent shear failure,

f = 0.086 v^ (4.3)

c max ^

^

Therefore, the shear capacity of the slab is considered to

be exceeded when the required compressive strength

42
determined in accordance with equation (4.3) is greater than

the estimated strength at the time of collapse.

In addition to flexure or shear type failure in the slab,

the collapse could have been attributed to other causes such

as excessive creep in concrete or failure of one or more of

the supporting columns. A limited investigation indicated

that the columns had sufficient capacity to resist the applied

loads. The possibility of creep being a significant contributory factor in the collapse was ruled out because of the

short duration of peak loads on the 23rd and 22nd floor slabs

prior to the collapse (about 4 hours) and the presence of reinforcement in the compressive regions of the slabs.

The scope of the following investigation is based on the

hypothesis that failure conditions were governed either by

flexure or by shear.

A. Investigation of Flexural Failure

A comparison of results shown in figures 4.3 and 4.4

reveals the internal moments in the slab from the analyses

of Cases 1 and 2 to be virtually identical. This indicates

that the accuracy of predicted concrete strength will not

have a significant influence on the results of flexural

investigations.

Altogether four column strips were investigated.

Of these, the strip shown in figure 4.6 was found to be

the one most critical in flexure. Case 1 negative

moment at the critical section near column 31 (node 258)

was about 9.2 ft-kips/ft. The moment capacity of the

43
slab at this section, determined on the basis of reinforcement details in the contract drawings, was found to be

about 8.4 ft -kips/ft, indicating that the flexural capacity

of the section was exceeded by about 10 percent. In

this case, as in all other cases investigated, the flexural

capacity of the section was found to be governed by

yielding of the tensile reinforcement rather than by

crushing of concrete in compression. Thus it is

estimated that yielding occurred in the slab at column

31 and yielding propagated toward column 83 (see figure 4.1).

However, this local yielding is not sufficient to cause

a collapse mechanism of the 23rd story floor slab.

In fact, a flat slab construction of the type used in

Skyline Plaza building is generally recognized to have

considerably greater moment capacity than the capacity

at which tlie first local yielding occurs because of its

inherent ability to redistribute peak moments to neighboring

regions through the mechanism of yielding.

The next most severe condition occurred in the column

strip shown in figure 4.9. Maximum negative moments in elements

adjacent to column 84 (node 165) were in the order of

7 to 8 ft-kip/ft at the critical section. This compares

with a 10 ft-kip/ft ultimate moment capacity in the slab.

The difference is not large enough to preclude yielding

or even some yield line propagation from column 84 in

the east-west direction; however, if such yielding were

to occur the situation would be very similar to the

• 44
localized condition around column 31. Again, it is concluded

that this was not a significant contributory factor to tlie

collapse of the 23rd floor.

To further check the ultimate flexural capacity of the

floor slab a selected critical area was investigated using

yield line analysis. The yield line pattern assumed for the

balcony panel bounded by columns 32, 84, 83 and 31 is shown

in figure 4.11. Top and bottom reinforcement taken from the

contract documents is indicated in the same figure. This

particular analysis gave an ultimate load capacity of about

300 pounds per square foot indicating 80 percent excess

capacity over construction loads. Other yield line analyses

performed at selected interior panels gave consistently

similar results. It should be noted that unless the assumed

yield line pattern happens to be the same as would actually

develop in the slab, the approach used will overestimate the

slab capacity. However, the margin of approximation involved

would be much less than the excess capacity indicated by

these analyses.

The possibility that the collapse might be attributed

to flexural failure in the 22nd story slab was also considered.

Tlic most critical situation would occur when the 23rd story

slab is fully shored and no reshores are installed under the

22nd floor slab (some employee statements indicate that the

23rd floor slab could have been fully shored, as discussed in

section 2.2.2). This condition will be designated hereafter

as Case 4. In this case, tlie loads acting upon the 22nd

floor would consist of its own weight, the weight of

of the 22nd story shoring and the loads transmitted from the

45
23rd floor. Assuming the loads on the 23rd floor to be

distributed to both slabs in proportion to their respective

elastic moduli, and shoring to be perfectly rigid [12], tlie

maximum total load tliat could act on the 2 2nd floor would be

about 4 percent greater than those assuriied acting on the

unshored 23rd floor (Case 1). Since both slabs are of identical

design, the internal forces (Siiears , moments and torques) in

tiie 22nd floor slab would likewise be about 4 percent greater.

However, the flexibility of timber shores would nodify the

load distribution between slabs significantly [15], with the net

effect of reducing the loads transm.itted to the 22na floor (viz.

a 100 percent flexible sliore transmits no loads). Further

reduction in transmitted loads would occur as a result of poor

shoring conditions whicii appears to be a reasonable assumption

to make (see section 2.2.2). Taking all these factors into

account, it is concluded that flexure in tiie 22nd floor slab did

not constitute the initial mode of failure. This conclusion is

furtlier corroborated by observations made, elsewhere in this

report (see section 2.2.2 and figure 2.28). It should be

noted that since the 23rd floor slab, when fully shored, is

less critical in flexure tlian either in Case 1 (no shoring) or

in Case 3 (partial shoring) , it requires no consideration with

respect to Case 4.

Upon examination of all the probable conditions prior to

tlie collapse, it is concluded that the initial mode of failure

of cither the 22nd or the 23rd floor slab was not flexural.

46
B. I nvestigation of Shear Failure

Trom the analytical results, axial loads and moments

transmitted to column elements from the floor system were

assessed in relation to peripheral column dimensions in order

to establish which of the columns would be expected to create

a severe shear condition in the slab. Figure 4.12 shows the

four most critical regions identified at columns 67, 68, 83

and 84.

Figure 4.13 tabulates the results of the analysis of shear

stresses in the 23rd-story floor slab for cases 1 through 3,

based on the critical sections shown in figure 4.12.

Maximum shear stresses calculated in accordance with eq . (4.1)

are given in column 3. Column 4 gives the required compressive

strength to resist this shear as determined from equation (4.3)

and column 5 gives the probable range of in-situ concrete

strength at the time of collapse.

According to recent studies [13A] the maximum shear capacity

of a slab is significantly reduced when the supporting column

has a narrow rectangular cross section as compared to a square

column. In the case of column 68, the reduction in strength

would be in the order of 20 percent. The bracketed values in

column 4 of the table reflect the effect of column rectangular ity

.

Examination of the tabulated results reveals that the shear

capacity of the 23rd floor slab is exceeded at columns 67, 68,

83, and 84 under unshored and partially shored conditions (Cases

1 and 3, respectively). It can be seen by comparing Case 1 values in column 4 with the probable range of compressive

47
strength in column 5 that eacli of tlio four rep,ioi^s of the slab

at tlie columns is critical in shear. For Case 3, conditions are

critical at columns 84 and 68 and marginal at columns 83 and

67. Shear stresses at these columns are comparable to stresses

in Case 1, indicating that partial removal of slioring creates

conditions nearly as critical as v>;ould develop in the unshored

slab. Note that none of the strength requirements exceed the

3000 psi design strength of the concrete considered in Case 2.

The loss of support as a result of shear failure at any one of

these columns would cause a progressive propagation of failure

at neighboring columns and eventual rupture of the slab.

The cases discussed above represent conditions under vvliich

the 23rd story slab is the one most critically stressed in shear.

The possibility of shear failure having first been initiated in

the 22nd story slab was also examined. The most critical

condition in this slab would occur under Case 4 loading

(23rd floor fully shored and 22nd floor not reshored)

.

As previously noted, with infinitely rigid shoring, the

load acting on the 22n.d floor in Case 4 would be about 4 percent

greater tlian the Case 3 load acting on the 23rd floor. However,

the actual load distribution between the two floors would be

significantly modified by the flexibility of the timber shoring

[15] with the net effect of reducing tlie load transmitted to

the 22nd floor. The load transmitted would be decreased further

by the poor condition of shoring such as described in section

2.2.2.

48
The effect of tliese two factors on the load distribution

cannot be quantified because of insufficient information.

However, the load distribution wliicli v;ould simultaneously

mobilize both floors to\;ards their respective shear capacities

can be examined; a condition under which the combined resistance

of both floors is a maximum. In tliis case, the total weiglit

of tliree floors plus shoring in tv;o stories would be distributed

between the 22nd and 23rd floors in proportion to the square

roots of their respective compressive strengths. The results

indicated required compressive strengths of 1030 psi and lllS

psi for tlie 23rd and 22nd floors, respectively. These compare

with 1200 psi and 1340 psi respective average values estimated

for these floors. Consequently, failure under Case 4 loading

cannot be precluded. However, the following evidence is taken

as an indication that the collapse did not initiate in the 22nd

floor blab: (a) employee statements did not indicate sagging

of the 22nd floor slab at tlic time of collapse and (b) one

op.ployee statem.ent indicating the loss of the 21st story reshore

prior to the incident v;as interpreted as a decrease in the

22nd floor deflection (section 2.2.2 and figure 2.23).

The contract documents (structural drawings) specify tb.e

follov;ing: "Slab being poured to be shored for two floors and

backpropped at center of S]:)an each v:ay and at center of bay on

next floor dov.-n." Uncertainty about tlic effectiveness

of backpropping or its presence prior to the collapse (section

2.2.2 and figure 2.27) makes it virtually impossible to make a

49
quantitative assessment of loading distribution between 23rd

through 21st floor slabs for this shoring configuration.

Upon examination of all the probable conditions prior to the

collapse, it is concluded that the initial shear mode of failure

of the 23rd story floor slab resulting from partial or complete

removal of shoring prior to the incident was a major contributing

factor to the collapse.

4.5 Probable Mode of Collapse in Building A-4

The most likely mode of collapse has been determined to be

a shear failure around columns 67, 68, 83, or 84. The premature

removal of forms supporting the 23rd story slab wlien the concrete

of that slab had a relatively low strength pr*oduced shear stresses

that were in excess of the concrete capacity at the time of the

incident .

"

The three-dimensional finite element analyses have shown

the slab to be overstressed in flexure in only a few local regions.

The capacity of flat slabs to redistribute moments is well known

and thus local flexural yielding should not have led to failure.

Approximate ultimate flexural capacities were computed

by the yield line analysis method. These ultimate flexural

loads indicated that, even with the forms removed, the 23rd

floor slab should not fail in flexure, thus confirming the

interpretation of the results of the elastic analyses.

Hawkins test results [13A] indicate that shear cracking

can cause concentrated slab rotations near the column that

are quite large. Based on this observation it is felt quite

50
possible that large deflections could have occurred, even witli

a shear type of failure that is estimated to have occurred.

Most of the eyewitness reports indicated deflection in the

23rd and 24th story slabs (varying from 6 in to 2 ft) which

increased over a 15 or 20 minute time period before failure.

.'Vii increasing deflection of this type is usually associated

with a flexural failure; however, this type of deflection could

also be associated with a shear failure to the layman observer.

The collapse is believed to have started with shear

around columns 67, 68, 83, or 84. The loss of support from

any one of these columns would then lead to overs tress ing

at the remaining columns.

The accumulation and impact of debris from the 2 3rd and

24th floor slabs would have overloaded the 22nd floor slab

and induced progressive collapse of successive floors to the

ground

.

There is no indication that the crane was a contributing

factor to the beginning of collapse in building A-4. No

witness statements indicated that the crane moved prior to or

during the initial sagging of the 23rd and 24th floors. The

crane supports on the 14th and 16th floors are far enough away

from the initiation of failure to preclude the crane as a cause. However, the crane probably became a driving force in

the collapse once its support was lost.

51
5. SUMI'lARY OP FINDINGS

The findings given in this section are based on site

investigations, OSHA case records, structural and architectural

drawings, shop drawings, and structural calculations. The

applicable Federal regulations are the Safety and Health

Regulations for Construction [10] which incorporates the

American National Standard A 10.9, Safety Requirements for

Concrete Construction and Masonry Work [2].

5.1 Mode of Failure

On the basis of evidence as well as analysis, it appears

that the collapse was initiated at tlie 23rd floor level.

An analysis of the 23rd-floor slab indicates that its

most likely mode of failure was in shear around one or more

columns in section 3 of the floor slab. The strength of the

23rd-floor slab on the day of collapse has been found to be

of a magnitude that complete or partial removal of shoring

underneath the slab would have produced a shear failure in

the slab. The weight of debris resulted in failure in the

slabs below and carried through the heiglit of the building.

5.2 Non-Compliance with OSHA Regulations

Non-compliance with OSHA regulations was found in a number

of instances. These are listed below along with a discussion

of each item.

52
(a) Shoring in section 3 of the 22nd story

Lxamination of physical evidence and employees' statements

indicate that tlie 22nd story forms were being removed on the

day of the incident. OSHA regulations (AMSI - AlO . 9 , section

5.4.7) require adherence to engineer's specifications and local

building codes in determining lengtli of time for forms to

remain in place. The engineer's requirements were expressed

in the fern of a note on the structural drawings (section 2.2.2).

Tliis note required the ''slab being poured to be shored for two

floors and backpropped at center of span each way and at center of

bay on next floor down." The architect's specifications [19

j

required "In all cases, two floors shall be fully shored."

The removal of the 2 3rd story forms left only one story of

formwork in place under tlie recently poured 24th floor.

(b) Premature removal of 22nd-story forms

The length of time forms were required to be left in place

was not explicitly stated by the engineer, architect or local

code. In such instances OSHA regulations (A:n;SI - AlO . 9 , section

G.4.C) provide minimum curing times. V'hcn the design live

load is less tlian the dead load, 4 days are required for

spans less than 10 ft, 7 days for 10 ft to 20 ft spans, and

10 days for spans exceeding 20 ft. The time periods are for

cumulative numbers of days in which the air temperature surrounding

the concrete exceeds 50°F. The 4-day old 23rd floor slab had

spans exceeding 10 ft. The forms removed on tlie 22nd story

were in an area with spans exceeding 20 ft and therefore,

53
Silould liavc been in place for 10 days of ter.iperatures excecdinji

50°F.

(c) rield-cured concrete specimens

OSIIA regulations (AMSI-A10.9, section 6.4.7) require the

use of field-cured concrete specimens in order to insure that

concrete lias obtained sufficient strengtli to safely support

the load prior to removal of forms . No evidence has been found

which indicates that field-cured specimens were prepared or

used.

(d) Lateral bracing

OSIIA regulations (ANSI-A10.9, section 6.3.2 and 8.1.5)

require the design of braces and shores to resist all foreseeable lateral loads. Minimum values of 100 pounds per foot

of floor edge or 2 percent of the total dead load of the floor,

vsrhichever is greater, is required. No evidence has been found

which indicates that lateral load was considered in the design

of forms. The lateral bracing provided (about 2 nominal 3x4 's

per 16 ft) would not provide this resistance.

(e) Shoring out of plumb

OSHA regulations (.^\'SI - AlO . 9 , section 8.1.24) allow a maximum deviation of 1/8 in per 3 ft out of plumb. Deviations

of shoring exceeding these limits were found on the 23rd

(figure 2.20) and 24th (figure 2.16) stories.

(f) Damaged shoring

OSHA regulations (ANSI-A10.9, section 8.1.25) require

removal of damaged or weakened shoring. On-site inspection

after the incident indicates this was not done on the 23rd

(figure 2.22) and 24th (figure 2.18) stories.

54
(g) Inspection

OSHA regulations (29 CFR 1926.700 (e) {1) (iv) ) require inspection immediately before, during, and after placing

concrete. This either was not done or deficiencies in (e)

and (f) above were not corrected.

(h) Crane Installation

OSHA regulations (29 CFR 1926.550 (a) (1) ) require the operation of cranes as prescribed by the

manufacturer. The following deviations were found:

(1) Crane no. 1 (in section 1, away from tlie collapse

area) - distance between top and bottom supports

was less than the required 18 ft 4 in (figure 2.30)

(18 ft was used)

.

(2) Crane no. 2 (in section 4, in the collapse area)-

distance between top and bottom supports was less

than the required 21 ft 4 in (18 ft was used).

(3) Crane no. 2 (in section 4, in the collapse area) -

The number of standard tower sections used was one

more than the four recommended by the manufacturer

(figure 2.41 and section 2.2.3).

(4) Crane no. 2 (in section 4, in the collapse area) -

The maximum tower height exceeded (by approximately

one standard tower section) the 81 ft-0 in recommended

(figure 2.31 and section 2.2.3).

55
6. ri-feiu:ncj.:s

1. ACI-ASCr; Committee 326 , "Shear and Diagonal Tension,"

J ournal ,of the Ame rican Concrete Institut e, Vol. S9

,

Tanuary, Feb ruary , and Mafcli 1962.

2. American National Standard Institute, "Safety Requirements

; for Concrete Construction and Masonry IVork," ANSI A10.9-

1970, American National Standard Institute, Inc., New

York, New Yorlc, 1970 .

3. Kloem, O.L. "Concrete Strength ,'Ieasurement - -Cores versus

Cylinders," rroc eedings of tli e American Society for Testing

Materials , Vol. TT, 1965": ~

4. Bloem, D.L. "Concrete Strength in Structures," Journal

of the y\merican Concrete Institute , Vo 1 . 6 5, I 'arch 1968

.

5. Building Code Ordinance, County of Fairfax, Virginia,

Adopted May 2, 19 56 with Amendments to August 4, 19 71.

^ • Building Code Requirements for Reinforced Concrete

CACI 513-63) , American Concrete Institute, De troi t

,

Michigan, June, 1963.

^* Building Code Requirements for Reinforced Concrete

"CACI 3 18- 71) , American Concrete Institute, Detroit,

Michigan , 1971

.

7 A . The Building Collapse at 2000 Commonwealth Avenue,

Boston , Massachusetts , Report of the Mayor's Investigating

Commiss ion , City of B'oston, Massachusetts, 1971.

8. Campbell, R.H. and Tobin , R.E. "Core and Cylinder Strengths

of Natural and Lightweight Concrete," Journal of the

American Concrete I nstitute , Vol. 64, April 196 7.

9. Clough, R.W. and Felippa, C.A. "A Refined Quadrilateral

Element for Analysis of Plate Bending," Proceedings , Second Conference on ^latrix Methods in StructUral

Mechanics , V/right -Patterson Air Force Base, Ohio , 1968.

10. Department of Labor, Bureau of Labor Standards, "Safety

and Health Regulations for Construction," (29 CFR Part 1926),

Washington, D.C.

11. Formwork for Concrete SP-4, American Concrete Institute,

Detroit, Michigan, 196 3.

12. Grundy, P. and Kabaila, A. "Construction Loads on Slabs

With Shored Formwork in Multistory Buildings," Journal

of the American Concrete Institute, Vol. 60, December 1963.

56
13. Hanson, M.I'/., and Hanson, J.'v. "Sliear and 'loricnt Transfer

between Concrete Slabs and Columns,'' Journal PCA Research

and Development Laboratory, Vol. 10, No. 1, January 1968.

13A. Hawkins, N.M., Fallsen, H.B. and Hinojosa, [l.C. ^'Influence

of Column Rec tangul ar i ty on the Behavior of Flat Plate

S t rue tur e s

, '

' Cracking, Deflection, and Ultimate Load o f

Conc rete Slab Systems , SP-30, American Concrete Institute,

1971.

14. Klieger, P. "Effect of Mixing and Curing Temperature on

Concrete Strength," Journal of tlie American Concrete

Institute , Vol. 54, June 1958.

15. Nielsen, Knud , L.C, "Loads on Reinforced Concrete Floor

Slabs and Their Deformations during Construction,"

Proceeding no. 15 Swedish Cement and Concrete Research

Institute, Royal Institute of Technology, Stockholm,

Sweden, 19 52.

16. Price, Vv. H. "Factors Influencing Concrete Strength,"

Journal of the American Co n crete Institute , Vol. 47,

1951.

17. "Results of Core Testing Project at Skyline Center

Building No. 4," Value Engineering Laboratory Report

No. 45-RD-73, Alexandria, Virginia, March 21, 1973.

18. Skyline Plaza Collapse - Case Records , Occupational

Safety and Health Administration, Region 3, Richmond,

Virginia, 19 73.

1 9 . S pecifications for Skyline Apartments A- 4, Fairfax

County, Virginia, V.'cihe, lUaclc, and Jeffries, Arch i t e c t

,

Vv ashing ton , D.C'. , September 5 , 1972 .

2 0 . Specifications for Structural Concrete for B uildings , AC I" 301-72 ( with s elected AC I and ASTM re f erences) , SP -15(72), American Concrete Institute, Detroit, Michigan,

1972.

21. Tower Crane Manual PC- 1400 , American Pecco Corporation,

Millv«ood, .'Jew York.

22. Tower Crane Manual PC- 16 0 0, American Pecco Corporation,

Mi 1 Iwood , New Yo rk

.

22A. U'ilson, F.L. "SAP, A General Structural Analysis Program,"

Report to v;alla V.'alla District U. S. Engineers Office,

September 1970 (not published).

23. V.Mnter Concreting, Theory and Practice , Proceedings of tlie

iTlLEM Symposium, Copenhagen, 1956 .

57
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1-1
Figure 2.54 Slab support on western wall of parking garage.
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a

a

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01

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(0 o

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Columns Cast

7 day cylinder strength

28 day cylinder strength

Slab Cast

7 day cylinder strength

28 day cylinder strength

Columns Cast

7 day cylinder strength

28 day cylinder strength

Slab Cast

7 day cylinder strength

28 day cylinder strength

Columns Cast

7 day cylinder strength

28 day cylinder strength

Slab Cast

7 day cylinder strength

28 day cylinder strength

Columns Cast

7 day cylinder strength

28 day cylinder strength

Slab Cast

7 day cylinder strength

28 day cylinder strength

3/ 2/73

ilisliz 3/ 1/73 3/ 2/73

2095 psi 2825 psi

-

2/27/73 2/28/73 3/ 1/73

2775 psi 2085 psi 2695 psi 2325 psi

3610 psi 3235 psi

2/23/73 2/24/73 2/26/73 2/27/73

1380 psi - 2290 psi 2575 psi

3405 psi 3750 psi

2/20/73 2/21/73 2/23/73 2/23/73

- 2740 psi 2460 psi 2460 psi

- ~ 3125 psi 3125 psi

2/14/73 2/15/73 2/20/73 2/21/73

2220 psi 1540 psi 2280 psi 2115 psi

3210 psi 3205 psi

2/ 7/73 2/13/73 2/14/73 2/15/73

3185 psi 2610 psi 2325 psi 2210 psi

4405 psi 3385 psi 3565 psi

2/ 5/73 2/ 6/73 2/ 7/73 2/13/73

2690 psi 3160 psi 2110 psi 2070 psi

3195 psi 3520 psi 3025 psi 3345 psi

Slab Cast

7 day cylinder strength

28 day cylinder strength

4

24th Floor

23rd Floor

22nd Floor

21st Floor

1/10/73

3105 psi

4270 psi

1/11/73

2185 psi

3130 psi

17th Floor

Section 1 Section 2 Section 3 Section 4

Figure 3.1 Casting dates and laboratory cylinder strengths
Floor

No.

Number

of Cores

Tested

Age at

Time of

Collapse,

Days

Age at

Time of

Test,

Days

Average

Strength at

Time of Test,

psi

Range of

Strength

psi

Average

Density,

Ib/cu ft

Diameter Core Compression Tests

23 12 4 16 2295 1690-3360 119

22 13 10 25 1960 1390-2840 118

21 12 23 38 2470 1810-3150 117

, 17 12 51 66 2780 2450-3420 124

i" Diameter Core Split Cylinder Tests

' 23 3 4 23 295 255-320

22 3 10 29 320 305 -335

Figure 3.2 Results of strength tests on 4-in

diameter cores.


70

T

\—I—

1—

60

50

ill

Z)

tr

LlI Q.

1x1

40

30

^Temperature range recorded at construction site

Average temperature recorded at National

Airport

Average curing temperature of slab 23,

section 3, 45°

Average curing temperature of slab 22,

section 3, 42°

20

10

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J L 20 21 22 23

FEBRUARY

24 26 27 28 2

MARCH

CALENDAR DATE

Figure 3.5 Temperature history for section 3 or 22nd and

23rd floor slabs.
Reinforcing Bars Removed from Floor Slabs

from Collapsed Section of Building

Sample

No. Size

Yield

Load, lbs.

Maximum

Load, lbs.

Yield

Strength, psi

Tensile

Strength, psi

-r-i — i 1 % Elongation

1 13\ \J 1 500 \J \J\J 22,800 68,700 116,000 8.2

9 12 600 22 100 64, 100 112,500 8.9

3 #4 14,000 21,600 71,300 110,000 14.1

4 #5 22,000 34,000 71,700 110,800 9.4

5 #5 21,000 33,800 68,400 110,200 9.4

6 #7 38,400 62,000 63,800 103,100 8.6

New Reinforcing Bars Obtained from

The Jobsite

Sample

No. Size

Yield

Load, lbs.

Maximum

Load, -lbs.

Yield

Strength, psi

Tensile

Strength, psi % Elongation

7 H 13,500 21,750 68,700 110,700 13.3

8 H 13,000 21,300 66,200 108,500 13.3

9 H 21,400 34,800 69,700 113,400 13.9

10 #5 21,600 35,600 70,400 116,000 13.3

11 #7 39,500 64,200 65,700 109,300 11.2

12 #7 39,400 65,000 65,500 108,900 10.9

Figure 3.6 Results of reinforcing bar tests.
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Y

1

d/2

+ d

V

d/2

htd

gure 4,10 TYPICAL RECTANGULAR COLUMN SHOWING

CRITICAL SECTION FOR SHEAR
SCALE: 1/4"= I'-O"

Figure 4.11 YIELDLINE PATTERN IN BALCONY PANEL
18.5

r

30. 5

24'

Mx=2l.7

TMy=433

16"

22.5

I I Column 67

(a)

54.5'

I I Column 68

(b)

Column 84

(c)

f— 18.5" n

12"

47.

Ik '36.5"

r

22.5"

1 C -> 30"

j

16"

46.8''^ |30.5"

21.6''*

16.2''^

>

24" 1

L J

*-x

Column 83

(d)

Figure 4.12 CRITICAL SECTION FOR SHEAR IN

THE SLAB AROUND COLUMNS 67,

68, 83, 84
Li o X uinn IN o .

1 o r* /"^ V

max

psi

c psi

£'c

C b T, Xllld UCO.

psi

1 147 1870 960-1440

f\ 7 . 7^ J. t / 1X fto 7/ nu o u u u

rift Y ? 17ft X ^ o X f X / yuU XH-fU

1 113 1104 (1724) 960-1440

fx ft 7L X X o 1 1 fid. n 77/1"^ XXUH-l^X/^fJ o u u u TAR Y 1 7

~» o lift X X Q XXOt ^^XoXD J you XH-HU

1 148 1894 960-1440

83 2 148 1894 3000

(24 X 16) 3 130 1462 960-1440

1 149 1920 960-1440

84 2 149 1920 3000

(30 X 12) 3 145 1819 960-1440

-'^The values in brackets are based on v^^^ > 0.8

(3.4//T^) or £^ = 0.135 v^^^ to account for the

effect of column rectangularity [13A]

Figure 4.13 Summary of shear stresses in

the 23rd story floor slab.
NBS-l 14A iHEv. 7-73)

U .S. OKPT. OF COMM. BIBLIOGRAPHIC DATA

SHEET

1. IM:hI,[( ATION (IK Ki:i'OK r NO. NBSIR 73-222

2. dov't Accession

No.

3. KccipuTit's At" f- SSI on Nc;,

4. i i ri.i- AND .suH'i n

JL nvco L a L 1 on or unii oKyiius Jtriaza uoiiapse

in Fairfax County^ Virginia

5. I'ul.lK-aiiun Date June 1973

6. I^crforming Organization Coiic NBSIR 73-222

7. AUTHOR(S)

Edgar V. Leyendecker and S. George Fattal

8. Performing Organ. Report No.

9. PERFORMING OKCiANIZATlON NAME AND ADDRESS

NATIONAL BUREAU OF STANDARDS

DEPARTMENT OF COMMERCE

WASHINGTON, D.C. 20234

10. Project/Task-/W'ork Unit No. 4605503

11. Contract/Grant No.

12. Sponsoring Organization Name and Complete Address (Street, City, State, ^IP)

Occupational Safety and Health Administration

U.S. Department of Labor

Washington, D.C. 20210

13. Type of Report & Peri oil Covered p^^^^

June 1973

14. Sponsoring Agency Code

15. SUPPLEMENTARY NOTES

16. ABSTRACT (A 200-word or less {actual summary almost significant information. If document includes a significant

bibliography or literature survey, mention it here.) The collapse of the Skyline Plaza apartment building A-4 has been studied by using

information contained in case records of the Occupational Safety and Health

Administration (OSHA), U. S. Department of Labor and obtained from on-site

inspections by investigators from the National Bureau of Standards.

Non-compliance with OSHA construction standards has been identified with regard

to formwork, field-cured concrete specimens and crane installation. Specifically,

the construction procedures did not comply with standards for the removal of

supporting forms. It is concluded that premature removal of forms was a contributing

factor to the collapse in building A-4.

An analysis of the 23rd-floor slab indicates that its most likely mode of failure

was in shear around one or more columns in section 3 of the floor slab. The

strength of the 23rd-floor slab on the day of collapse has been estimated to be

at a level that removal of shoring could have produced shear failure in the slab.

17. KEY WORDS (six to twelve entries; alphabetical order; capitalize only the first letter of the first key word unless a proper name; separated by semicolons

)

Apartment building; collapse; concrete; concrete strength; flexure; shear; strength

18. AVAILABILITY t^Unlimited 19. SECURITY CLASS

(THIS REPORT)

21. NO. OF PAGES

[^K For Official Distribution. Do Not Release to NTIS

UNCL ASSIFIED

136

L

' Order From Snp. of Doc, U.S. Government Printing Office

\Va.ihington, D.C. 20402, SD Cat. No. CI 3 20. SI'CURH Y CLASS

(THIS PAGE)

22. Price

[^y/^rdyr From National Technical Information Service (NTIS)

Springfield, Virginia 22151 UNCLASSIFIED

USCOMM-DC 29042-P74