Q NBS SPECIAL PUBLICATION 444

U.S. DEPARTMENT OF COMMERCE / National Bureau of Standards

WIND AND SEISMIC EFFECTS

Proceedings of the

Sixth Joint UJNR

Panel Conference
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WIND AND SEISMIC EFFECTS

Proceedings of the Sixth Joint

Panel Conference of the U.S. - Japan

Cooperative Program in

Natural Resources

May 15-17, 1974

National Bureau of Standards

Gaithersburg, Md.

H.S. Lew, Editor

Institute for Applied Technology

National Bureau of Standards

Washington, D. C. 20234

U.S. DEPARTMENT OF COMMERCE, Elliot L Richardson, Secretary

James A. Baker, III, Under Secreiary

Dr. Betsy Ancker-Johnson, Assisfani Secrefary for Science and Technology

NATIONAL BUREAU OF STANDARDS, Ernest Ambler, Acfing Direcior

Issued April 1976
Library of Congress Catalog Card Number: 76-608055

National Bureau of Standards Special Publication 444

Nat. Bur. Stand. (U.S.), Spec. Publ. 444, 462 pages (Apr. 1976) CODEN: XNBSAV

U.S. GOVERNMENT PRINTING OFFICE

WASHINGTON: 1976

For sale by llie Superintendent of Documents, U.S. Government Frintin"; Office, Washington, D.G. 20402 (Order by SD Calalop No. CI.^.10:444). Price $530 (Add 25 percent additional for other than U.S. mailing).
PREFACE

The Sixth Joint Meeting of the U.S. - Japan Panel on Wind and Seismic Effects was held

in Washington, D.C. on May 15-17, 1974. This panel is one of the twenty panels in the U.S. -

Japan Cooperative Program in Natural Resources (UJNR) . The UJNR was established in 1964 by

the U.S. - Japan Cabinet-level Committee on Trade and Economic Affairs. The purpose of the

VJNR is to exchange scientific and technological information which will be mutually beneficial

to the economics and welfare of both countries. Accordingly, the purpose of the annual joint

r-:eeting of this panel is to exchange technical information on the latest research and develop-

-ent activities within governmental agencies of both countries in the area of wind and seismic

effects

.

The proceedings include the opening remarks, the program, the formal resolutions, and

the technical papers presented at the Joint Meeting. The papers were presented in the respective language of each country. The texts of the papers, all of which were prepared in English,

have been edited. The remarks made by the delegates during the opening session were recorded

and transcribed. The formal resolutions were drafted at the closing session of the Joint

_ng and adopted unanimously by the panels of both countries.

Pages of the technical papers are numbered with a prefix corresponding to the Theme

number. The texts are consecutively numbered in each theme.

H. S. Lew, Secretary

U.S. Panel on Wind and

Seismic Effects
SI Conversion Units

In view of present accepted practice in this technological area, U.S. customary units

of measurements have been used throughout this report. It should be noted that the U.S. is a signatory to the General Conference on Weights and Measures which gave official status to the metric SI system of units in 1960. Conversion factors for units in this report are:

Customary Unit International

(SI)

,

UNIT

Conversion

Approximate

Length inch (in)

foot (ft)

meter (m) meter (m)

1 in=0. 0254m

i 1 ft=0. 3048m

Force pound (Ibf)

kilogram (kgf)

newton (N) newton (N)

1 lbf=4.448N

1 kgf=9.807N

Pressure

Stress

pound per square

inch (psi)

Kip per square

inch (ksi)

newton/meter

newton/meter

1 psi=6895N/m

1 ksi=6895xlO^N/m^

Energy inch-pound (in-lbf)

foot-pound (ft-lbf)

joule (J) joule (J)1 in-lbf=0.1130 J

1 ft-lbf=l. 3558 J

Torque

or Bending

Moment

pound-inch (Ibf-in)

pound-foot (Ibf-ft)

newton-meter (N-m)

newton-meter (N-m)

1 lbf-in=0.1130 N-m

1 lbf-ft=l. 3558 N-m

Weight

or Mass

pound (Ibf) kilogram (kg) 1 lb=0.4536 kg

Unit Weight pound per cubic foot

(pcf)

kilogram per cubic

meter (kg/m^)

1 pcf=16. 018 kg/m~

Velocity foot per second

(ft/sec)

meter per second

(m/s)

1 fps=0. 3048 m/s

Acceleration foot per second per

2 second (ft/sec )

2 2 meter per second per 1 ft/sec =0.3048 m/s

2 second (m/s )

Meter may be subdivided. A centimeter (cm) is 1/100 m and a millimeter (mm) is 1/1000 m.

t Exact
CONTENTS

Page

PREFACE iii

SI CONVERSION iv

ABSTRACT viii

PROGRAM OF THE JOINT MEETING ix

OPENING SESSION

Remarks by Dr. Ernest Ambler, Deputy Director,

National Bureau of Standards xiii

Remarks by Mr. Isao Uchida, Counselor,

Embassy of Japan xiii

Remarks by Dr. Edward O. Pfrang, Chairman,

U.S. Panel xiv

Remarks by Mr. Mitsuru Nagao, Chairman,

Japan Panel xv

LIST OF MEMBERS xvi

PHOTOGRAPH OF PARTICIPANTS xix

FORMAL RESOLUTIONS xx

THEMES AND TECHNICAL PAPERS

THEME I: EXTREME WINDS IN STRUCTURAL DESIGN

On the Gust Response of Long Span Suspension Bridges

N. Narita and K. Yokoyama I- 1 Extreme winds in Hurricanes and Possibility of Modifying Them

R. C. Gentry I- 21 Extreme Winds in the United States

A. R. Hull and P. E. Hughes I- 34 THEME II: ASSESSMENT AND EXPERIMENTAL TECHNIQUES FOR MEASURING

WIND LOADS

Wind Tunnel Experiments for Studying a Local Wind

K. Suda, S. Soma and K. Takeuchi II- 1 A Study of Wind Pressures on a Single-Family Dwelling in Model

and Full Scale

R. D. Marshall II- 21 THEME III: DYNAMICS OF SOIL STRUCTURES AND GROUND RESPONSE IN EARTHQUAKES

Non-linear Calculations of Ground Response in Earthquakes

W. B. Joyner, A. T. F. Chew and P. C. Doherty Ill- 1 Observation and Analysis of Ground Response in Earthquakes

S. Hayashi and H. Tsuchida Ill- 13 Research Study on Liquefaction

W. F. Marcuson Ill- 24

V
CONTENTS-Continued

Page

Estimation of Liquefaction Potential by Means of Explosion Test

K. Yamamura and Y. Koga Ill- 38 Landslide Incidence and Mechanisms During Earthquakes

G. E. Erickson Ill- 52 Stress Condition Effects on Dynamic Properties of Soil '

E. Kuribayashi and T. Iwasaki Ill- 72 Prediction of Maximum Earthquake Intensities for the

San Francisco Bay Region

R. D. Borcherdt and J. F. Gibbs Ill- 84 THEME IV: STRUCTURAL RESPONSE TO WIND AND EARTHQUAKE, AND DESIGN CRITERIA

A Statistical Approach to the Loading and Failure of Structures

R. G. Merrit IV- 1 Synthetic Experimental Research on the Ductility of Reinforced

Concrete Short Columns Under Large Deflection

''

• K. Nakano and M. Hirosawa IV- 16 Non-Linear Analysis of A Guyed Tower

S. K. Takahashi and W. A. Shaw IV- 37 A Standard for the Structural Integrity of Prefabricated Dwellings

K. Nakano, M. Hirosawa and T. Murota.:. IV- 55 An Analytical Model for Determining Energy Dissipation in Dynamically

Loaded Structures

J. F. McNamara and S. K. Sharma IV- 70 Design of Pile Foundations Subjected to Lateral Loads

M. Nagao, T. Okubo, K. Komada and A. Yamakawa IV- 90 Comprehensive Seismic Design Provisions for Buildings

C. G. Culver IV-113

Wind Loading and Modern Building Codes

E. Simiu and R. D. Marshall IV-127 THEME V: DISASTER MITIGATION AGAINST NATURAL HAZARDS

Seismic Retrofitting of Existing Highway Bridges

J. D. Cooper, R. R. Robinson and A. Longinow V- 1 Dynamic Tests of Structures by Use of A Large Scale Shake Table

S. Inaba V- 25 A Methodology for Evaluation of Existing Build'ings Against

Earthquakes, Hurricanes and Tornadoes

H. S. Lew and C. G. Culver V- 35 Experimental Research on the Aseismic Characteristics of A Spherical

- Steel Tank for Liquid Petroleum Gas ,

K. Nakano and M. Watabe. V- 50

vi
CONTENTS-Continued

Page

Research on Minimizing Earthquake Structural Damage to Single-Family Dwellings

W. J. Werner V- 63

Earthquake Engineering Research Supported by the National

Science Foundation •

. C. Thiel V- 66

The Wind Engineering Program

M. P. Gaus V- 80 THEME VI: TECHNOLOGICAL ASSISTANCE TO DEVELOPING COUNTRIES

Preliminary Report on Present Status and Development Project of

Volcanological Observation and Research in Indonesia

A. Suwa VI- 1 Use of Stabilized Adobe Block and Cane in Construction of Low-Cost

Housing in Peru

S. G. Fattal VI- 16

A Comment on the Technological Aid to Developing Countries

M. Nagao and T. Okubo VI- 25

vii
ABSTRACT

The Sixth Joint Meeting of the U.S. - Japan Panel on Wind and Seismic Effects

was held in Washington, D.C. on May 15-17, 1974. The proceedings of the Joint

Meeting include the opening remarks, the program, the formal resolutions, and the

technical papers. The subject matter covered in the papers includes extreme winds

in structural design; assessment and experimental techniques for measuring wind

loads; dynamics of soil structures and ground response in earthquakes; structural

response to wind and earthquake and design criteria; disaster mitigation against

natural hazards; and technological assistance to developing countries.

Key Words: Bridges ;' buildings ; codes; disaster; dynamic analysis; earthquakes;

modeling; soils; structural response; volcanoes; and wind.
SIXTH JOINT MEETING

PROGRAM OF THE

U.S. - JAPAN PANEL ON WIND AND SEISMIC EFFECTS

May 15-17, 1974

at National Bureau of Standards

WEDNESDAY-May 15

OPENING SESSION (10th Floor Conf. Room, Admin., Bldg.)

1000 Call to order by Dr. H. S. Lew, Secretary, U.S. Panel

Remarks by Dr. Ernest Ambler, Deputy Director - National Bureau of Standards

Remarks by Mr. Isao Uchida, Counselor, Embassy of Japan

Remarks by Dr. Edward 0. Pfrang, Chairman, U.S. Panel

Remarks by Mr. Mitsuru Nagao, Chairman, Japan Panel

1030 Introduction of U.S. Panel Members by U.S. Chairman

and Japan Panel Members by Japan Chairman

1045 Election of Conference Chairman

1100 Adoption of Agenda

1115 Adjourn

1130 Group Photograph

1145 Lunch-Dining Room C

1300 Leave for Fairbanks Highway Research Laboratories

Federal Highway Administration

THURSDAY-May 16

THEME II-ASSESSMENT AND EXPERIMENTAL TECHNIQUES FOR

MEASURING WIND LOADS

Chairman: Dr. Edward O. Pfrang

0900 Wind Tunnel Experiments for Studying a Local Wind - K. Suda* , S. Soma and K. Takeuchi

0920 A Study of Wind Pressures on Single-Family Dwellings in

Model and Full Scale - R. D. Marshall

0940 Discussion

THEME I-EXTREME WINDS IN STRUCTURAL DESIGN

Chairman: Mr. Mitsuru Nagao

1000 The Gust Response of Long Span Suspension Bridges - N. Narita and K. Yokoyama (Presented by T. Okubo)

* Underline designates the person presenting the paper.

ix
1020 Extreme Winds in Hurricanes and Possibility of

Modifying Them - R. C. Gentry

1040 Break

1100 Extreme Winds in the United States - A. Hull and P. Hughes

1120 Discussion

1130 A film presentation, "The Only Decision" - J. Lefter

1215 Lunch-Dining Room C

THEME IV-STRUCTURAL RESPONSE TO WIND AND EARTHQUAKE,

AND DESIGN CRITERIA

Chairman: Dr. Edward 0. Pfrang

1340 A Statistical Approach to the Loading and Failure of

Structures - R. G. Merrit

1400 Synthetic Experimental Research on the Ductility of

: Reinforced Concrete Short Columns under Large Deflection

K. Nakano and M. Hirosawa

1420 Non-linear Analysis of a Guyed Tower - S. K. Takahashi

and W. A. Shaw

1430 A standard for the Structural Integrity of Prefabricated

Dwellings - K. Nakano , M. Hirosawa and T. Murota

1440 Break

1455 An Analytical Model for Determining Energy Dissipation in

Dynamically Loaded Structures -J. F. McNamara and S. K. Sharma

1505 Design of Pile Foundations Subjected to Lateral Load - M. Nagao, T. Okubo, K. Komada and A. Yamakawa

1515 Comprehensive Seismic Design Provisions for Buildings - C. Culver

1525 Wind Loading and Modern Building Codes - E. Simiu and

R. Marshall

1535 Discussion

THEME VI-TECHNOLOGICAL ASSISTANCE TO DEVELOPING COUNTRIES

Chairman: Mr. Mitsuru Nagao

1600 Preliminary Report on Present Status and Development Project

of Volcanological Observation and Research in Indonesia - A. Suwa

1620 Use of Stabilized Adobe Block and Cane in Construction

of Low-cost Housing in Peru - S. G. Fattal
1640

1715

FRIDAY-May 17

0900

0920

0940

0950

1000

A Comment on the Technological Aid to Developing Countries

M. Nagao and T. Okubo

Discussion

Adjourn

THEME III-DYNAMICS OF SOIL STRUCTURES AND CxOUND RESPONSE IN EARTHQUAKES

Chairman: Dr. Edward 0. Pfrang

Non-linear Calculations of Ground Response in Earthquakes - W. B. Joyner , A. T. F. Chew and P. C. Doherty

Observation and Analysis of Ground Response in Earthquakes

S. Hayashi (Presented by H. Tsuchida)

Research Study on Liquefaction - W. F. Marcuson (Presented by K. O. O'Donnell)

Estimation of Liquefaction Potential by Means of Explosion

Tests - K. Yamamura and Y. Koga (Presented by S. Inaba)

Landslide Incidence and Mechanisms during Earthquakes - G. E. Ericksen

1010

1020

1040

1050

1115

1215

1340

1400

1420

1430

Stress Condition Effects on Dynamic Properties of Soil

E. Kuribayashi and T. Iwasaki (Presented by H. Tsuchida)

Break .

•

Prediction of Maximum Earthquake Intensities for the

San Francisco Bay Region - R. Brocherdt and J. F. Gibbs

Discussion :

Tour of NHS Wind Tunnel Facilities

Lunch-Dining Room C

THEME V-DISASTER MITIGATION AGAINST NATURAL HAZARDS

Chairman: Mr. Mitsuru Nagao

Seismic Retrofitting of Existing Highway Bridges -

J. D. Cooper , R. Robinson and A. Longinow

Dynamic Tests of Structures by Use of a Large Scale Shaking

Table - S. Inaba

A Methodology for Evaluation of Existing Buildings against

Earthquakes, Hurricanes and Tornadoes - H . S . Lew and C. Culver

Experimental Research on the Aseismic Characteristics of :

Spherical Steel Tank for Liquid Petroleum Gas - K. Nakano

and M. Watabe

xi
Research on Minimizing Earthquake Structural Damage to single-Family Dwellings - W. J. Werner

Earthquake Engineering Research Supported by the National

Science Foundation - C. Thiel

The Wind Engineering Program - M. Gaus

Discussion

Break

THEME VII-Free Discussion

Chairman: Mr. Mitsuru Nagao

Discussion

CLOSING SESSION . .

Chairman: Dr. Edward O. Pfrang

Adoption of Formal Resolution

Adjourn
OPENING REMARKS BY DR. ERNEST AMBLER

DEPUTY DIRECTOR, NATIONAL BUREAU OF STANDARDS

Counselor Uchida, Mr. Nagao and distinguished guests, common interests are the bonds

between individuals and between nations. We at the National Bureau of Standards, as fellow

members of the scientific community and your hosts, welcome you and hope you find your stay

pleasant and profitable. Our laboratories will be open to you, and the staff will be happy to talk with you about any problems that interest you.

In the pursuit of scientific information on wind and seismic effects, you are travelling

like the wind to the four corners of the earth. This same pursuit of scientific knowledge has

led many Bureau scientists to become world travelers as they studied, for instance, the engineering aspects of the 1964 Alaska earthquake, the 1971 San Fernando earthquake, and the 1972

Managua disaster.

Our two nations, bordering as we do on the Pacific Ocean, the borders of which show

considerable seismic activity, are interested in developing design methods and criteria for

building structures that will better withstand wind and shock. Both of us individually

conduct research programs, but we are here today to make our programs more effective through

cooperation. We think this is a very important undertaking and we are very happy that you are here today, and again, welcome you to the National Bureau of Standards.

OPENING REMARKS BY MR. ISAO UCHIDA

COUNSELOR, EMBASSY OF JAPAN

Ladies and Gentlemen,

It is my great honor to be given the opportunity to say a few words at the opening of

the Sixth Joint Meeting of the UJNR panel on wind and seismic effects.

The United States and Japan have very close cooperative relations in every facet of our activities. Science and technology is not the exception. Now we have the U.S. - Japan

committee on scientific cooperation and the U.S. - Japan Committee on Medical Science, as well as UJNR. -

Many activities such as exchange of scientists, joint seminars and joint research are

being carried out actively and effectively under these cooperative programs. We have also

very close cooperative programs in the area of so called big sciences: nuclear development

and space development. In addition, the new cooperative agreement on energy research and

development will be signed in the near future.

xiii
The Japanese government has been making considerable advancement in research and develop0 ment of science and technology by realizing that science and technology play an important role

in the progress of the national economy and the improvement of the welfare of the people. At

the same time, we understand that the science and technology belong to all people in the

world. From this point of view, we have developed a policy to promote international cooperation in research and development of science and technology. I hope and believe that our cooperative effort in science and technology contributes not only to the prosperity of both

countries but also to the welfare of all people in the world.

UJNR has already marked its ten year history and the second five year report is presently

in progress. At present, we have twenty working panels under UJNR programs who are producing

valuable information. I believe this is the result of the efforts by many concerned people

from both countries. . . _

. .

In regard to the matter of wind and seismic effects, which is handled in this panel,

although I am a layman in this field, I know that the U.S. and Japan are the two leading

countries in the field and all members here are the experts of both countries. I believe in

the success of this panel and the hope for future developments.

I should like to express my heartfelt gratitude to each and all of the participants,

especially to Dr. Ambler, Dr. Pfrang and other American panel members. Dr. Lew and other

secretariate staff, for their great contribution toward making the arrangements and for their

wairm hospitality to the Japanese members.

Thank you.

.. . , OPENING REMARKS BY DR. EDWARD O. PFRANG

, CHAIRMAN, U.S. PANEL

Mr. Uchida and Mr. Fujisawa, I would like to thank you and the Japan Embassy for all the

assistance you have given us, not only with respect to this panel meeting but all the past

meetings. To Mr. Nagao, I would like to congratulate him on the outstanding panel which he

has brought with him to the Joint Meeting. I have had a brief opportunity to begin studying

the papers and I find them to have excellent technical contributions,

I don't know how many people in this room realize the fact that this meeting was almost

not held. On May 9th, a Richter 6.8 earthquake took place 100 kilometers from Tokyo. Had it

been on the Kanto plane, this meeting would have definitely been cancelled. This is one measure of the need for our joint study. Following the San Fernando earthquake of 1971,

xiv
Dr. Fukuoka, former chairman of the Japan Panel, led a team of distinguished Japanese engineers and scientists to the United States. I believe that you in Japan learned much from our

sad experience. We look forward to learning from you, on the occasion of the 7th Joint Meeting,

your lessons from the Izu earthquake.

Another area of our cooperation which needs to be mentioned regards our work with developing countries. We have scientifically cooperated in Peru. We appreciate the great kindness

that you have shown to us in regard to our efforts in the Phillipines. It has been very kind

of Mr. Nagao and Dr. Okubo and others to meet with our people in their efforts with regard to

the Phillipines and to provide their counsel.

We have accomplished much through our UJNR Program. However, when we consider the loss

of life in San Fernando and Izu, and when we also take into account the fact that on May 3rd

approximately 300 people lost their lives in tornadoes within the U.S., there is much that

remains to be done. Thank you very much.

OPENING REMARKS BY MR. MITSURU NAGAO

CHAIRMAN, JAPAN PANEL

Dr. Ambler, Dr. Pfrang, members of the U.S. Panel and guests, I feel very honored to have

this opportunity to express my greetings on behalf of the Japan Panel. I also would like to express my appreciation for the great efforts made by Dr. Pfrang and all other U.S. Panel

members preparing for the Sixth Joint Meeting.

As you know, in our past five joint meetings many topics concerning important wind and

seismic problems have been presented and discussed. The valuable results from these meetings

have been highly evaluated at the UJNR Conference. At this Joint Meeting, as many as thirty

three technical papers have been sutanitted. This clearly indicated continuing tireless

efforts of the members of both sides for the success of our program and I am assured that we

shall end our meeting with fruitful results. The fear of wind and earthquake will not diminish from mankind. We hope that we can continue our efforts toward our mutual goals.

In closing, I would like to express my gratitude for the cooperation and the assistance

of the U.S. Embassy in Tokyo and the Japan Embassy in Washington, D.C. , the National Bureau of

Standards, and others, expecially Dr. Pfrang and the Secretary of the U.S. Panel. Thank you

for your kind attention.

XV
U. S. Panel on Wind and Seismic Effects

Membership List

April 1974

Dr. Edward 0. Pfrang, Chairman

Chief, Structures, Materials S Safety Division

Center for Building Technology, lAT

National Bureau of Standards

Washington, D.C. 20234

Dr. S. T. Algermissen

Seismicity and Risk Analysis Branch

Building 25, Denver Federal Center

Denver, Colorado 80225

Mr. Billy Bohannan

Enginerred Wood Structures

Wood Engineering Research

U. S. Forest Products Lab.

Madison, Wisconsin 53705

Dr. George Ericksen

U.S. Geological Survey

12201 Sunrise Valley Drive

Reston, Virginia 22092

Dr. Michael P. Gaus

Section Head of Engineering Mechanics

Engineering Division

National Science Foundation

1800 G Street, N.W. Washington, D.C. 20550

Dr. R. Cecil Gentry

Director, National Hurricane Research Lab.

National Oceanic Atmospheric Administration

P.O. Box 8265, University of Miami Branch

Coral Gables, Florida 33124

Mr- Sigmund Gerber

Director for Construction Standards

and Design for Installation and Logistics

Office of the Secretary of Defense

Room 3E763, The Pentagon

Washington, D.C. 20301

Mr. Jack Healy

Chief, Material Systems & Science Division

Department of the Army

Construction Engineering Research Labs.

P.O. Box 4005

Champaign, Illinois 61820

Mr. Arnold R. Hull

Associated Director for Climatology

NOAA-EDS

Washington, D.C. 20235

Dr. William B. Joyner

U.S. Geological Survey

Office of Earthquake Research

345 Middlefield Road

Menlo Park, California 94025

Mr. John W. Kaufman

Chief, Atmospheric Dynamics Branch

Aerospace Environment Division

National Aeronautics S Space Administration

Marshall Space Flight Center

Mobile, Alabama 35812

Mr. James Lefter

Director, Civil Engineering Service

Office of Construction

Veterans Administration

Lafayette Building, Room 507

811 Vermont Avenue, N.W. Washington, D.C. 20420

Dr. R. B. Matthiesen

Director, Seismic Engineering Branch

U.S. Geological Survey

390 Main Street

San Francisco, California 94104

Mr. Keith 0. O'Donnell

Assistant Chief, Structural Branch

Engineering Division, Civil Works Directorate

Forrestal Building, ATTN: DAEN-CWE-D

Washington, D.C. 20314

\

xvi
Mr. Larry F. Rousch

Public Building Service

General Services Administration

18th S F Strett, N.W. Washington, D.C. 20405

Mr. Charles Scheffey

Acting Associate Administrator

Federal Highway Administration

Washington, D.C. 20590

Dr. Warren A. Shaw

Head, Civil Engineering Department

Civil Engineering Laboratory

Naval Construction Batallion Center

Port Hueneme, California 93043

Mr. Lawrence C. Shao

Directorate of Licensing

U.S. Atomic Energy Commission

Washington, D.C. 20545

Dr. Charles T. Thiel

Program Manager, Earthquake Engineering

Division of Advanced Technology Applications

National Science Foundation

Washington, D.C. 20550

Mr. William J. Werner

Division of Building Technology and Safety

Department of Housing and Urban Development

Room 4120

451 7th Street, S.W. Washington, D.C. 20410

Dr. Roger D. Brocherdt, Alternate

U.S. Geological Survey

National Center for Earthquake Research

345 Middlefield Road

Menlo Park, California 94025

Dr. H. S. Lew, Secretary and Alternate

Structures, Materials & Safety Division

Center for Building Technology, lAT

National Bureau of Standards

Washington, D.C. 20234

Dr. John B. Scalzi, Alternate

Division of Advanced Technology & Applications

National Science Foundation

Washington, D.C. 20550

Mr. F. J. Tamanini, Alternate

Chief, Structures and Applied Mechanics Di\'.

HRS-10

U.S. Department of Transportation

Federal Highway Administration

Washington, D.C. 20550

xvii
Japan Panel on Wind and Seismic Effects

Membership List

April 1974

Mr. Mitsuru Nagao, Chairman

Director, Public Works Research Institute

Ministry of Construction

2-28-32- Honkomagome, Bunkyo-ku, Tokyo

Dr. Satoshi Hayashi

Head, Structure Division

Port and Harbor Research Institute, Ministry

of Transport

3-1-1 Nagase, Yokosuka-shi , Kanagawa-ken

Mr. Seiichi Inaba

Chief, Earthquake Engineering Laboratory

National Research Center for Disaster

Prevention

Science and Technology Agency

4489-1 Kurihara Aza, Sakuramura, Niiharigun,

Ibaragin-ken

Dr. Yasunori Koizxjmi

Director, Building Research Institute

Ministry of Construction

3- 28-2 Hyakunin-cho , Shin Juku-ku, Tokyo

Mr. Eiichi Kuribayashi

Chief, Earthquake Engineering Section

Structure and Bridge Division, Chiba Branch

Public Works Research Institute, Ministry

of Construction

4- 12-52 Anagawa, Chiba-shi, Chiba-ken

Dr. Kiyoshi Nakano

Head, Structural Division, Building

Research Institute

Ministry of Construction

3- 28-8 Hyakunin-cho, Shin Juku-ku, Tokyo

Mr . Nobuyuki Narita

Chief, Structure Section

Structure and Bridge Division, Chiba Branch

Public Works Research Institute, Ministry

of Construction

4- 12-52 Anagawa, Chiba-shi, Chiba-ken

Dr. Tadayoshi Okubo

Head, Structure and Bridge Division

Chiba Branch, Public Works Researcy Institute

Ministry of Construction

4-12-52 Anagawa, Chiba-shi, Chiba-ken

Mr. Tetsuo Santo

Head, International Institute of Seismology

and Earthquake Engineering

Building Research Institute, Ministry of

Construction

3- 28-8 Hyakunin-cho, Shin Juku-ku, Tokyo

Dr. Ken Suda

Director, Meteorological Research Institute

Meteorological Agency, Ministry of Transport

4- 35-8 Kita, Koenji, Suginami-ku, Tokyo

Mr. Akira Suwa

Head, Siesmological Laboratory, Meteorological

Research Institute

Meteorological Agency, Ministry of Transport

4-35-8-Kita, Koenji, Suginami-ku, Tokyo

Dr. Kiyohide Takeuchi

Head, Applied Meteorological Laboratory

Meteorological Research Institute

Meteorological Agency, Ministry of Transport

4-35-8-Kita, Koenji, Suginami-ku, Tokyo

Mr. Hajime Tsuchida

Chief, Earthquake Resistant Structures Lab.

Structure Division

Building Research Institute, Ministry

of Construction

3-28-8-Hyakunin-cho , Shin Juku-ku, Tokyo

Dr. Makoto Watabe

Chief, International Institute of Seismology

and Earthquake Engineering

Building Research Institute, Ministry

of Construction

3- 28-8-Hyakunin-cho, Shin Juku-ku, Tokyo

Vacant

Chief, Soil Dynamics Section

Construction Method and Equipment Division

Chiba Branch, Public Works Research Institute

Ministry of Construction

4- 12-52 Anagawa, Chiba-shi, Chiba-ken

Mr. Hiroshi Yoshimura, Secretary

Head, Planning Division

Public Works Research Institute

Ministry of Construction

2-28-32-Honkomagome , Bunkyo-ku, Tokyo

xviii
xix
SIXTH JOINT MEETING PROPOSED RESOLUTIONS

Both delegations proposed the following formal resolutions at the conclusion of the

Sixth Joint Meeting of the U. S. - Japan Panel.

1. That the Sixth Joint Meeting was of great value to both sides and the joint

program should continue.

2. That the next Joint Meeting should be held in Tokyo in 1975, preferably in May,

and that the technical sessions should be extended to three days.

3. That increased effort should be made in the near future to encourage joint

research programs, expecially in the area of the mutual utilization of research

facilities and the exchange of researchers.

4. That the scope of the activities of the Joint Panel should be expanded to include

areas of mutual interest in seismic risk analysis as related to earthquake

prediction

.

5. That efforts should continue, or be implemented, to publish the proceedings of

past Joint Meetings in the respective countries and make them available to the

respective engineering and scientific professions at large.

XX
ON THE GUST RESPONSE OF LONG-SPAN SUSPENSION BRIDGES

by

N. Narita

Chief, Structure Section

Public Works Research Institute

Ministry of Construction

Japan

K. Yokoyama

Research Engineer

Structure Section

Public Works Research Institute

In the design of above-ground structures it is established practice to con- sider the effects of wind. For some structures, like long-span suspension bridges,

the influence of wind may be the primary design control which will then govern

the inherent safety of the structures and the final construction cost.

This paper, therefore, will describe the required changes in design specifications to incorporate the influence of gusts on long-span suspension bridges.

The necessary numerical calculations are illustrated in addition to some long-term observations on the Kanmon Bridge. The importance and necessity of

studies on gust response are emphasized.

Key Words: Bridge; design; field data; gust response; model; specifications;

structure; theory; wind.

I-l
INTRODUCTION

Necessity of Study on Gust Response of a Structure

When one considers wind effects, its fluctuation must be relative to time and space. The time fluctuating wind tends to introduce forced random vibrations in above-ground

structures. This necessitates consideration of some dynamic magnification effects applied

to the static response of the structure. The spatial fluctuation of wind tends to decrease

the probability that the wind, with mean speed, attacks the entire structure at the same

time, thus some modification may also be necessary in order to include this effect.

The determination of the response of a structure to wind, as based on a statistical

concept relative to a stationary time series, was introduced by A. G. Davenport in 1961.

This concept has since been adopted into the design specifications for building structures

in Denmark and Canada. The importance of gust response in long-span suspension bridges

was also suggested by Davenport and the magnification effect due to gustiness was estimated as three or more times that of the static response.

The gust response in suspension bridges, relative to buildings, points out the following ;

(i) The structural configuration of a bridge is far more complicated, as compared to a building structure, i.e., the cable and the suspended

structure are composed of horizontally long elements, and the main

tower is composed of vertically long elements.

(ii) The damping capacity is small, compared to a building structure.

(iii) The geographical location of the bridge inhibits one for obtaining

aerodynamic characteristics by analytical methods.

Moreover, the accumulation of statistical data of natural wind is minimal and it may be

dangerous to utilize a statistical theory.

In order to solve this difficult problem, it is necessary to study the following'

themes

;

(i) To evaluate the statistical characteristics of wind, especially

strong wind.

(ii) To evaluate the structural and vibrational characteristics of longspan suspension bridges.

(iii) To evaluate the aerodynamic characteristics of brige sections.

(iv) ^To improve the method of statistical analysis.

(v) To accumulate test data, as obtained during the observation of the

gust response of real structures, and to compare these with the

, theoretical results. . ,

In this paper items (ii) , (iv) and (v) will be presented. .

'

Gust Response and Design Specifications ^ , . ,

'

,

A wind resistant design specification, for long-span suspension bridges, was first

presented in 1974 for use in the design of the proposed Honshu-Shikoku Bridge. This

specification was developed by a Technical Advisory Committee of the Japan Society of

Civil Engineers. However, these initial design specifications, which have been applied

to bridges of 150 meters in length or less, have not specified dynamic effects due to wind loading. Subsequent modifications to the specifications have since occurred through

1-2
consideration of recent research activities. The basis for the (1964) specification and

subsequent revision (1972) will now be presented.

(i) Wind Resistant EJesign specification (1964) for the proposed HonshuShikoku Bridge

2 In this specification the design wind load is defined as = 1/2 x pV^ ^^d'

where denotes design wind load, p air density, drag coefficient, and A the

exposed area of the structure per unit length. The design wind speed V(j was derived

relative to the basic wind speed V]_q as given by: = V^^q x y ]_ ^ Y2' where Yi is

a modification factor for V^^q and takes into account the altitude of the structure,

and it is assumed to satisfy the power law. The second modification factor Y2 is

relative to the wind gustiness and was determined by making the following assumptions.

The instantaneous wind speed pa^rallel to the mean flow, at an arbit^rary point,

is the sum of the mean wind speed (V^) and the fluctuating wind speed (V^) , averaged

over (s) seconds or;

t = V3 + V. = V + I sin (W^t +6^)

i The wave length Lj^, whicli corresponds to the circular frequency W^^ , may then be

written as L- = V 111= V^T., where T- = HL. The lateral scale of turbulence B. 1 s „^ s 1' 1 1 corresponding to W. , is assumed to be_l/k times L. and equal to the horizontal length

of the structure B. Then, s = k x b/V^

.

The relation between mean wind speed and the averaging time of the wind speed is;

Vg/V^ = (S/600)"'' , r = rQ(Z/Zo)~°'^^

where is the ten minute averaged wind speed at the height of 10 meters, and the

constant r^ = 0.090 at = 12.2 meters. From the above equations the gust response

modification factor can be obtained as;

r^ = V /V = (600 V,/kB)^/^"^

where varies according to V-|_q and the power index p, r varies according to the

altitude as noted previously. When V-^q = 40 m/s , the power index p = 1/6, which are

the values that are assumed in the calculation of the modification factor y

2-

However, as mentioned above, the (1964) design specifications did not take into

account the dynamic characteristics of the structure. These considerations will now

be presented.

(ii) Wind Resistant Design specification (1972) for the proposed HonshuShikoku Bridge

More than ten years have passed since the wind resistant design statistical con- cepts were first proposed by Davenport. Though the concept has been applied to the

building structures in Denmark and Canada, it has not been established in Japan. In developing the 1972 specification, a preliminary study was first conducted using the

stationary time series theory. The modification factor for the gust response was

then evaluated and found to equal 1.3 to 1.5, for the lateral bending moment of stiffening frames of suspension bridges with center span lengths of 500 to 1,500 meters.

However, there is no evidence of induced lateral vibration of real bridges, therefore,

the dynamic response effect the stiffening frame of the structure was neglected in the

modification factor Y_, which gave a value between 1.14 and 1.19. This revised specification, V7hich considered statistical characteristics of wind, still remained ambiguous.

1-3
GUST RESPONSE OF A SUSPENSION BRIDGE

Fundamental Equations and Some Remarks on the Numerical Calculations

The fundamental equations for evaluating the gust response of a suspension bridge will

now be introduced without presenting the details.

(i) Structural Analysis of the System

Lateral motion of the system was analyzed by use of the so-called finite element

method. The fundamental equation of the system is given by;

_2_(El£iJ^) = q^ + (u + V)

dx^ dx^ ^

2 P^

dx

k 4-V, and the mechanical admittance for r mode of vibration is;

H^(n)^ 2 ^ i/p^4[( _ (p/p^)2}2 + 4C^^(P/P^)2]

The fraction of damping was assumed as

= C„ + , where = Ml , ? "o "aero ° 2tt ae^o 32'Tmn^

(ii) Spectral Density of Wind Force

The mean wind force P corresponding to the mean wind speed is - - 2 P = 1/2 PV^ C A

The spectral density of the u-component was evaluated by Hino's formula, namely,

6KV 2 10 2 S„(n) = 0.476 (1 + B-) 5/6

6 32

6 = 1-169 x 10-3 (z/10) ^^"'eTM " D

The cross-wind correlation spectrum was evaluated by Davenport's expression;

z Then, the spectrum of the fluctuating wind force is expressed as;

S^_,(n) S (n) = 4 P|x (n) P —

P u ^

^ • z The aerodynamic admittance equation is given by Vickery's formula and is;

Ix (n)|2 = {1 + 2(1^)4/3] -1 .

(iii) Aerodynamic Response

The response spectrum of the lateral bending moment of the stiffening frames

is;

S^{A, n) = [ M^2(^^) |H^(n) |2|j^(n) |2 Sp(n) . r where the joint acceptance Jj,(n) is given in the following form;

1-4
'2 The variajice and effective frequency for the bending moment are given as;

00 00 j^O^) =/ S_CH, n)dn = [ M^^(H')/ I (n) | ^1 (n) | ^S^ (n) dn m ' m ' j.vwj,^

o r o

2

(-A) = / mh^Cii, n)dn/a

The expected value of the maximum response is '3j-^(^'l'^^(^)> where

g„ (H) = [21n{Y^0'JT}]^/2 ^ "'^^^^

[21n{Y 00 T}]^/' m

Finally, the modification factor for the wind speed relative to bending moment is

expressed as; 1/2

The above equations were incorporated into a computer program. The accuracy of the

numerical procedure was examined relative to the following terms and expressions;

(i) Terms Vj,(^0, u^())), M^U), Q^(il), Ma'), andQW)

(ii) Double integral in the joint acceptance

(iii) / (. . .n)dJV and l{. . .n) o r

ILLUSTRATIVE EXAMPLES

Symmetric Suspension Bridges Consisting of Three Suspended Spans with Hinged

Stiffening Frames

Numerical calculations were made relative to six model designs, the essential properties of which are given in Table 1. The maximum bending moment occurs near x/L = 0.2

for the L1500 model, with the peak moment moving gradually toward the span center as the

span lengths decrease. This means that the influence of the cable against the wind

loading increa'=;es proportionally with the span length. The modification factor for

the various bridges is shown in Figure 1. In general, does not vary, except for the

L1500 model, and has an average value of about 1.35.

Symmetric Suspension Bridges Consisting of Three Suspended Spans with Continuous Stiffening Frames

The suspension bridges which have continuous stiffening frames, that were analyzed,

are detailed in Table 2. The resulting modification factor Y^ for these bridges is given

in Figure 1-1. The characteristic feature of the continuous suspension bridge is that Y2

increases around the point x/L = 0.2, where the bending moment changes its sign. It also

appears that Y2 increases gradually along the end of the side span.

It should be noted that the modification factor analyzed under the assumption of

hinged suspension bridges is nearly equal to that of continuous bridges, except at the

peak location.

Effects of Wind Characteristics on Gust Response

Table 3 shows the effect of the wind characteristics on the modification factor. In

this table the power index p, the surface drag coefficient K and the cross-wind correlation factor k are chosen as variables. The functions K and k appear to be the governing

factors

.

1-5
FIELD OBSERVATION OF THE GUST RESPONSE AT THE KANMON BRIDGE

Observation System at the Kanmon Bridge

The instrumentation and data-recording system, designed to monitor and record wind

speeds and direction as well as earthquakes, and to indicate their effects on the superand sub-structures of the bridge, has been provided by Japan Highway Corporation, The

location of instruments and elements are shown in Figure 2 and Table 4. The system was placed into operation last winter, just after the opening of the bridge.

Some Examples of the Observed Data

The original trace of the wind-induced vibration of the super- structure is given in

Figure 3, with the wind record at that time tabulated in Table 5. Table 6 shows the re- sults of a preliminary analysis. The natural frequencies obtained from the wind induced

vibration record coincides with those obtained from field forced vibrations induced by a large scale excitor during tests conducted last September. However, the damping capacity

could not be obtained from the power spectral density given in Figure 4. The damping

capacity, as shown in Table 6, was derived from the forced vibration test. The observed

data will be analyzed and compared with the theoretical values in the near future.

' CONCLUSIONS

The authors have outlined the present status of a study on the gust response problems

of long-span suspension bridges. In order to enhance the wind resistant design technique,

vigorous research activities are necessary relative to the wind characteristics and aero- dynamic response of a structure.

The Honshu-Shikoku Bridge Authority is now conducting a field observation test by using

a large scale section model of the beach of Tateyama, located at the south apex of the Boso

Peninsula. These tests whould provide us with useful information about the gust response

as well as the flutter of a long-span suspension bridge.

ACKNOWLEDGMENT

The authors are indebted to Mr. Y. Kono and Mr. K. Nomura for considerable help in

doing numerical calculation.

REFERENCES

1. Davenport, A.G. , Application of Statistical Concepts to Wind Loading of Structures,

Paper No. 6480, ICE.

2. Davenport, A.G., Buffetting a Suspension Bridge by Storm Winds, ST3, ASCE, June 1962.

3. Hino, M. , Relationships Between the Instantaneous Peak Values and the Evaluation Time,

N. 117, Trans, of JSCE, May 1965.
Table-1 Hinged suspension bridge model

item

mod e 1 L1500 LllOO LIOOO L800 L650 L5 00 unit Ny sp an m 1 J UU liUU J.UUU cOO J UU

chord distance m 7 A J o J 0 33 33 33 33

truss height m 14 14 11 11 8 8

sag m 150 130 100 73 59 45

lateral bend,

rigid i ty

2

t X m 1 .266 1.249 0.923 1.012 0,731 0.781

-9

(x 10 )

dead weight

of truss

t/m 20.28 20, 25 19.32 19. 66 19.01 19.01

dead weight

of cables t/m 11.40 9. 35 7.95 5.57 4.51 3.47

Table-2 Continuous suspension bridge model

i t em model

LllOO L890 unit \

center span m 1100 890

side span m 260 326

chord distance m 32 34

sag m 100 82

lateral bend,

rigidity t x m

2 1.637 2. 174

(x lO"*^)

dead weight

of trusses

t/m 31.82 44.51

dead weight

of cables

t/m 15 .91 14.23

1-7
o

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CM

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CM

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<}

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in00

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m

CM

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CN

ON

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csl CO

00

COm

CO

00

CN

ON

NO

csl

egO

CO

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CN oin

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ON

ONmCO ON

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1-8
Table - 4 Instruments at Kanmon Bridge

Observation

item

Measuring

point

Maximum

measuring

range

Instruments Number

of

instruments

Number of

elements Rate of

sampling

(f req. /sec) per

M.P

Total

Acceleration

Ground ±500 gal Submerged

accelerometer 3 3 9 100

Abutment ±500

Electromagnetic

accelerometer

4 1 4

50 Pier -5UU QO 1 8

Tower ±500 4 X1 4 Stif fenin

frame

' ±200 Servo6 1 6 25

Displacement

Stiffening

frame

Horizontal

±2 m

Vertical

±2 m

XY-Analyser 1 3 3 25

Longitudinal

±25 cm

Differential

transducer

1 1 1 25

Wind

speed

Stiffening

frame

Horizontal

60 m/ sec Vertical

±10 m/sec

Supersonic

anemometer

1 3 3 25

Wind

direction

Horizontal

60 m/sec

540°

Propellar type anemomer

3 2 6 25

Wind

speed

Top of

Shimonoseki

tower

Horizontal

60 m/sec

1 2 2 25

Wind

direction

540°

1-9
Table ~ 5 Wind record at Kanmon Bridge

Wind speed

M.P.

Order of observation

(m/sec)

1 2 3 4

A 8.59 8.95 9.36 9.02

Mean B 9.80 9.99 10.58 10.23

C 8.88 9.49 9.65 9.99

A 10.52 10.75 11.45 11.97

Maximum B 11.93 12.24 12.79 13.63

C 11.22 11.39 11.38 14.88

A 6.21 6.09 7.26 4.25

Minimum B 6.44 5.88 8.39 4.11

C 6.74 6.97 7.85 5.13

A 0.85 1.22 1.48 1.97

Standard

deviat ion B 0.89 1.36 1.66 2.26

C 1.06 1.42 1.59 2.41

the center of the center span

40 m apart from M.P. ~.A to Shimonoseki

the quarter point of Shimonoseki side

I-IO

Note M.P. - A

M.P. - B

M.P. - C
Table - 6 Natural frequencies and damping capacity

of Kanmon Bridge

Mode of

vibration

Symmetry Order

Natural frequency (C/S) Logarithmic

decrement Theoretical Observed (1) Observed (2)

Vertical

bending

Symmetric

1 n 91 T 0. 0313

2 0.270 0.298 0.296 0.0180

3 0.429 0.450

4 0.528 0.570 0.579 0.0090

5 0.914 0.918 0.919 0.0087

Antisymmetric

1 0.152 0.180 0.178 0.0501

2 0.375 0.414

3 0.705 0.740 0.739 0.0125

Torsional

Symmetric

1 0.384 0.387 0.388 0.0129

2 0.766 0.717 0.721 0.0126

Antisymmetric 1 0.492 0.472 0.468 0.0162

(Note) Theoretical natural frequencies were calculated by Bleich's method.

Observed (1): field experiment with excitors

Observed (2) : wind induced vibration

I-ll
L 1500m

L ISOO'"

15 I 5

10

J I \ I I

.1 .2 3 4 5

J I I L

2 3 .4 .5 X/^

15-

L 1000 m L800'

10 1.0-

J L

I 2 3 4 5 ^/^

_i I i

I 2 3 4 .5 X/.

15-

L 650'

1.5- L500^

10

^1326 + 870 +326^

^

.—.—. h

1.0 1-

_l I i I 2 3 4 5

Hinged bridge

30

h

1-20

J I I I

I

.1 2 3 4 5 X/.

L260 + II00+260

0

300 200 100

Side span

0 100

Continuous bridge

200 300

Mom span

400 500 m

Fig. 1.1 Gust response factor (V2)

1-12
1-13
-1.0

I I \ I I I I 1 1 300 200 100 0 100 200 300 400 500 m

Side span 'Mam span

Fig. 1.3 Gust response factor (V2)

1-14
c

f—

t

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ca •H c c

0 to •H 11 CJ

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1-17
38S/ 3A0/^( |D6 ) Q S d

1-18
1-19
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1-20
EXTREME WINDS IN HURRICANES AND POSSIBILITY OF MODIFYING THEM

by

R. Cecil Gentry

National Hurricane Research Laboratory

Weather Modification Program Office

Environmental Research Laboratories

National Oceanic and Atmospheric Administration

Miami, Florida

A hurricane (or similar storm called by other names) is the most destructive of nature's phenomena. This is partly because of the extreme winds asso- ciated with the storms (ranging up to 320 kilometers per hour) , but also because

the winds may continue blowing for several hours, and they are accompanied by

rising ocean water, strong along-shore currents, and torrential rains.

Discussions will be presented of the frequency of hurricanes of various

intensities, the rate at which the wind speeds decrease after the storm crosses

the coast and moves inland, the effect of the winds on the storm surge, and the

variation of the wind speed with height.

For several years, members of the United States Government have been experimenting to reduce the maximum intensity of the winds of hurricanes. Summaries

will be presented of the progress and future prospects for this work.

Key Words: Cloud seeding; frequency distribution; hurricanes; typhoon; wind;

wind intensities.

1-21
INTRODUCTION

Hurricanes are the most destructive of natural phenomena. While severe tornadoes may

have stronger winds, those in hurricanes last longer and affect much larger areas. In addition, the storm surge, the associated strong coastal currents, and rain-induced floods,

all combine with the winds to make the mature tropical cyclone the most dangerous weapon

in nature's arsenal.

Because of the great intensity of hurricanes and similar tropical cyclones that go by

different names in other parts of the world, the extreme winds experienced at many locations are those that occur in hurricanes. This paper will discuss the maximum winds in

hurricanes, gusts associated with these winds, frequency of extreme winds, rates at which

the maximum winds decrease as the storm moves inland, variation of wind speed with height,

and possibility of moderating the hurricane by seeding the storm.

MAXIMUM WIND SPEEDS IN HURRICANES

It is rare that the maximum wind in a severe hurricane is measured. Rarely is an anemometer installed in the area of maximum hurricane winds, and most of them fail before

the winds reach their peak values in the more severe storms. As a result most of our knowledge of the extreme winds is based either on interpretation of wind force from the

damage or from knowledge of the pressure gradients in the storms which have a reasonably

close relationship to the wind speeds since the two quantities are related physically.

A small but very severe hurricane with minimum pressure at sea level of 25.85 inches

(909 mb) crossed the Florida Keys in 1935. Maximum winds were estimated at about 200 mph

from the types of damage. Hurricane Camille struck the Mississippi coast in August 1969.

The minimum central pressure was 905 mb , or less, shortly before the storm reached the

coast. A reconnaissance aircraft measured 140 knots while still several miles from the

maximum winds while flying at 3 km elevation before the storm reached land. Again the

maximum winds were not measured at a land station, but were estimated at about 200 mph.

Hurricane Janet (1955) crossed Swan Island in the Western Caribbean and the winds were estimated at 200 mph. It later struck Chetumal in Mexico where the minimum sea level

pressure was 27.00 inches (914 mb) . The winds were measured at 152 knots (175 mph) before

the anemometer collapsed.

U.S. Air Force Reconnaissance aircraft have measured flight level (3 km) winds at 170

knots (197 mph) in at least three different storms in the Pacific: Typhoon Opal, December

1964, Typhoon Kit, June 1966, and Typhoon Nancy, 1961. The reconnaissance aircraft have

also measured minimum sea level pressures of 877 mb in Typhoon Ida, 1958, and Typhoon Nora,

1973. In the Atlantic, the maximum winds measured by the Research Flight Facility were

157 knots (181 mph) in Hurricane Inez, 1966. Comparing the minimum pressure readings

recorded in the Atlantic and Pacific would suggest that the extreme maximum winds in the

Pacific from typhoons have probably exceeded those in hurricanes in the United States, but

the writer has no knowledge of measured values at land stations in the Pacific exceeding

the maximum values quoted above. . .

. ,,

.

GUSTS IN HURRICANE WINDS

For certain types of structures the damage caused by winds is more a function of the

gusts than of the sustained wind. If our knowledge of the maximum sustained wind (variously defined as the mean for an hour, mean for five minutes, fastest minute or fastest mile)

is inadequate, our knowledge of the maximum gusts is even less. Many weather stations are

not equipped to measure the short period gusts, and usually estimate the gusts by observing the wind speed dial over a five to ten second period.

The Royal Observatory in Hong Kong has used their recorded data including those mea- sured by a Dines Instrument to derive ratios of winds for various periods to the mean wind

for an hour. The data in Table 1 was copied from a report by Faber and Bell (1963)

.

1-22
Hebert and Neumann (1973) have used data reported by weather stations in the United

Ste.tes to also study the gust ratios. They concluded that the ratio varied not only with

the period of the wind, but with the speed of the wind. At lower wind speeds they found

the ratio of the gust to the sustained wind to be in general agreement with the Hong Kong

data, but at the higher wind speeds it was of the order of 1.2. Their data are summarized

in Figure 1. The regression line presents an empirical relationship between the steady

wind and peak gust based on the available data. The major weakness in their study other

than not having a lot of cases is that the term gust and sustained wind are not rigorously

defined. Hebert examined the weather reports received from stations experiencing tropical

cyclones during the years 1963-1970. In most cases the sustained wind would be either the

fastest minute or the fastest mile. The gust in a few cases may have been measured with

some special instrument, but in most cases would have been estimated by the observer after

examining a recording of the anemometer output or after observing the wind speed dial readings for a few seconds.

EXTREME WIND FREQUENCIES IN HURRICANES

The frequency of extreme winds such as those described in the earlier paragraphs is,

of course, very low. The data in Figure 2 give an idea how often winds of certain broad

categories may occur in the United States. The number in the "A", "B" and "C" belts tell

how many years during the period 1901-1965 a storm of the respective categories affected

the various sectors of the United States Coastline. Hurricanes with a minimum central

pressure of 29.00 inches (982 mb) include the storms whose maximum winds usually exceed,

85 mph (38 mps) and whose maximum storm surge usually exceed 6 feet (1.8 m) in height.

Hurricanes with a minimum central pressure less than 28.25 inches (957 mb) include the very

intense hurricanes in which the maximum height of the storm surge will be 10 to 15 feet

(3 to 4.7 m) or more above mean sea level in the sectors illustrated, and in which the maxim-urn wind velocities will exceed 120 mph (54 mps)

.

The general subj^ect of recurrence of extreme winds has been studied by Thom (1960) . Figure 3 is reproduced from his report and shows the annual extreme-mile 30 feet (9.1 m)

above ground, 50-year mean recurrence interval. His report includes similar charts for

varying mean recurrence intervals. In his data the extreme winds along the Gulf of Mexico

and Atlantic Coasts are in most cases from hurricanes.

REDUCTION IN HURRICANE WIND SPEEDS AS STORM MOVES INLAND

The hurricane wind speeds start decreasing soon after the storm crosses the coastline.

The phenomena has been studied extensively by Malkin (1959) and by Goldman and Ushijima

(1974) . Figure 4 is copied from the later report and it includes the summary gust factors

developed by Malkin. The Goldman and Ushijima report includes graphs similar to Figure 4 for two other storms. There is, as should be expected, some variation from storm to storm,

and from one locale to another. The heavy broken arrow in Figure 4 shows the track of

Hurricane Camille, 1969. The numbers written on the horizontal lines show the percent of

peak gusts at landfall at the respective locations. The "Malkin 's Factors" listed at the

left end of each line are from Malkin 's study.

Goldman and Ushijima summarize the results by writing, "The decrease in maximum hurricane winds after landfall has been shown to be both appreciable and variable. When com- pared with the earlier work, the maximijims in the peak gusts of the three significant

storms are in strong agreement with the general function of Malkin. However, the change

in the two-dimensional distributions of peak gusts inland reveals that a more complicated

function than Malkin's is necessary ...." The Goldman and Ushijima study showed considerable variation in gust factors between storms, so more studies are needed.

* Prepared by Gentry (1966) and adapted from an earlier chart prepared by U.S. Weather

Bureau (1957) . 1-23
VARIATION OF HURRICANE WIND SPEEDS WITH HEIGHT

The wind speeds in hurricanes vary with height. Within a millimeter of the ground the

wind is blowing very slowly--even in a hurricane. The speeds increase with height up to some unknown but variable level which is believed to be between 300 and 1500 feet (90-460

m) . Above that level the winds in the belt of maximum winds for the storm decrease very

slowly up to six kilometers where the sustained winds are normally only 10 percent less

than at low levels according to a study by Hawkins (1962)

.

Very little well documented data are available showing how the winds in hurricanes

vary from 10 meters up to 100 meters. Various assumptions have been used by engineers and

it is well known that the variation is a function of the roughness of the terrain.

POSSIBILITY OF MODIFYING HURRICANES

During the past ten years the average annual damage caused by hurricanes in the United*

States has been about $450,000,000 per year not including the damage from Hurricane Agnes,

1972, which was largely from rain-induced floods. Because of this and because research

over the last decade has suggested means by which hurricanes might be modified, there is

extensive research and experiments to modify hurricanes.

The principal damages from hurricanes can be attributed to three causes: (1) destructive force of the wind, (2) the storm surge, and (3) floods caused by hurricane rains.

Results from the modification experiments conducted thus far and the supporting research

suggest that the maximum winds in hurricanes may be reduced 10 to 15 percent if the hurricane clouds radially outward from the maximum winds are seeded with freezing nuclei. Since]

the force of the winds varies with the square of the wind speed, this could mean a reduction of 20 to 30 percent in the force of the wind and in damages caused by the winds. The

storm surge is the result of complex interactions of a number of forces including the wind.

More research is needed in this area, but there are reasons to believe that a reduction of

maximum winds will also often result in a reduction of the storm surge. The rain-inducted

floods will probably be little affected by the experiments.

Table 2 summarizes the results of previous experiments to modify hurricanes.

The only multiple seedings of the clouds near the eyewall made so far were in Hurricane Debbie. Gentry (1970) has reported the results.

The storm was seeded 18 August 1969 five times during a period of 8 hours. Before

the first seeding the maximum winds were 98 knots (Figure 5) . After the third seeding

the maximum winds were considerably reduced, and by about 5 hours after the last seeding

the maximum winds were down to 68 knots, which is a reduction of about 31 percent.

The crews rested on 19 August, and by 20 August Debbie had regained intensity. At the;

beginning of operations on 20 August, the maxim\im winds were 99 knots, but the storm structure had changed considerably. The inner maximum was about 10 nautical miles from the

center and the outer maximum was approximately 20 nautical miles (Figure 6) . The hypothesis requires that seeding be radially outward from the maximum winds. In this case it

would mean starting the seeding runs at about 12 nautical miles and proceeding outward for ' about 20 more nautical miles. If the hypothesis is correct, this should result in a re- duction in the maximum winds in the inner maximum, but an increase in the outer maximum.

This is precisely what happened. The inner maximum decreased radically during the day and

the outer maximum increased slightly. The fifth and last seeding started at about 20 nautical mile radius and proceeded outward; that is, it was radially outward from the outer

wind maximum. The winds in the outer maximum then decreased and by about 6 hours after

the fifth seeding the maximim winds were 85 knots, which is a reduction of 15 percent for

the day

.

The changes in the storm following the seedings on 18 and 20 August are very encourag- I ing and are highly suggestive that the storm was modified. The natural variability of I,

1-24
hurricanes is such, however, that we cannot be sure that the changes observed were caused

by the seeding.

Figure 7 shows the frequency distribution of 12-hour changes in maximum wind speeds in

hurricanes (see Sheets, 1970). The changes are expressed in percent. See the horizontal

axis. The data have been stratified according to the original intensity of the storm. Notice that even a reduction of 15 percent is a very rare event for any of the categories.

We have examined the weather maps for 18 and 20 August and have sought other explanations for the reduction in the maximum winds. On 18 August there was a trough in the upper

troposphere moving eastv;ard at more northernly latitudes which probably did affect Debbie

and may have caused part of its weakening. This would suggest, therefore, that the 30 percent reduction should not be attributed entirely to the affects of the seeding.

If one considers the full sequence of events for the Hurricane Debbie case, however,

one cannot help but be impressed by the results . On 18 August the storm was seeded five

times, and the maximum winds were reduced about 30 percent. On 19 August there was no seeding and the storm reintensified. On 20 August the storm was seeded 4 times in such a fashion as to cause reduced winds at the inner maximum and it essentially disappeared.

The fifth seeding was in such a fashion as to cause reduced winds in the outer maximum and

a net reduction of 15 percent resulted. In view of the probabilities expressed in the frequency diagram, it is clear that such a sequence of natural events would be very rare.

The data from the experiments, and especially when backed up by calculations with the

theoretical models, are highly suggestive that we do have an excellent chance of achieving

beneficial modification of hurricanes to the extent of reducing the maximum winds by 10 to

15 percent. Since the force of the wind varies with the square of the wind speed, a re- duction of 10 to 15 percent of maximum wind speed would mean a reduction of 20 to 30 percent

in the maximum force exerted by the hurricane. If damages can be reduced by this much, it means savings of about $100 million per year for the United States alone.

REFERENCES

1. Faber, S.E. and G.J. Bell, 1963: Typhoons in Hong Kong and Building Design , The Engineering Society of Hong Kong, 29 pp.

2. Gentry, R. Cecil, 1966: Nature and Scope of Hurricane Damage. Hurricane Symposium , October 10-11, 1966 , American Society for Oceanographers , Houston, Texas, pp. 229-253.

3. Gentry, R. Cecil, 1970: Hurricane Debbie Modification Experiments. Science , 168 , pp. 473-475.

4. Goldman, Joseph L. and Toshimitsu Ushijima, 1974: Decrease in Hurricane Winds after

Landfall. Journal of the Structural Division , Proceedings of the American Society of

Civil Engineers , 100 , No. STl, January 1974, pp. 129-141.

5. Hawkins, H.F., 1962: Vertical Wind Profiles in Hurricanes. National Hurricane Research

Project Report No. 55 , U.S. Weather Bureau, Washington, D.C., 16 pp.

6. Hebert, Paul and Charles Nexamann, 1973: A Comparison of Gusts and Sustained Winds in

Hurricanes. Unpublished Manuscript of National Hurricane Center, Miami, Florida, 3 pp.

7. Malkin, William, 1959: Filling and Intensity Changes in Hurricanes Over Land. National

Hurricane Research Project Report No. 34 , U.S. Weather Bureau, Washington, D.C. , 18 pp.

8. Sheets, Robert C. , 1970: Application of Bayesian Statistics for Stormfury Results.

Project STORMFURY Annual Report, 1969, Appendix H . U.S. Department of the Navy and

U.S. Department of Commerce, pp. H 1-16.

1-25
9. Thorn, H.C.S., 1950: Distributions of Exteme Winds in the United States. Journal of

the Structural Division , Proceedings of the American Society of Civil Engineers , 86, No. ST4, April 1960, pp. 11-24.

10. U.S. Weather Bureau, 1957: Survey of Meteorological Factors Pertinent to Reduction

of Loss of Life and Property in Hurricane Situations. National Hurricane Research

Report No. 5. U.S. Weather Bureau, p. 21.

1-26
TABLE 1

RATIO OF WINDS OF VARIOUS PERIODS TO HOURLY MEAN WIND

(Data from Royal Observatory in Hong Kong)

Period of 600 sec 300 sec 60 sec 30 sec 10 sec 5 sec 3 sec Dines Gust

Wind

Ratio to Hourly Mean

1.05 1.09 1.28 1.42 1.64 1.72 1.81 2.05

TABLE 2

RESULTS OF EXPERIMENTS IN SEEDING HURRICANE CLOUDS

NEAR THE EYEWALL

Name Date Number of Seedings

Approximate Maximum Wind

Speed Chcinge (Percent)

Hurricane Esther

Hurricane Esther

Hurricane Beulah

Hurricane Beulah

Hurricane Debbie

Hurricane Debbie

16 September 1961

17 September 1961

23 August 1963

24 August 1963

18 August 1969

20 August 1969

-10

* 0

0* -14

-30

-15

Seeding material dropped outside "seedable" clouds.

(In addition, a hurricane was seeded 13 October 1947 and Hurricane Ginger was seeded

26 and 28 September 1971. The clouds seeded in these storms were far removed from the

central core of the storm and had only weak or no vertical currents. Their seeding, therefore, should not have caused much change in the storm's intensity or structure.)

1-27
Firure 1. Empirical Relationships Between Steady Wind and Peak GUST m Atlantic

Area Trooical Cyclones. Dots arc data points. Solid line is regression

line drawn to fit the data. (Prepared by Hebert and Neumann, 197,).,)

1-28
Figure 2. Frequency of hurricanes and tropical storms

penetrating the Uaited Staces coasts along

the Atlantic and the GuJ f of Mexico.

(Gentry, 1966)

1-29
Figure 3. Isotach 0.02 Quantiles, in Miles Per Hour. Annual Extreme-Mile 30 Ft. Above Ground, 50~Yr. Mean Recurrence Interval. (Thorn, 1960)

Figure 4. Percent of Peak Gusts at Landfall for Camille at

3fr-iQile

Intervals after Landfall

(10-mile (16.1 km)

increments from maximum

are computed at each inland

line. Circled values are maximum at each time.)

(Goldman and Ushijima,

197A)

1-30
o o o o o o o

O C» CO N CD \D ^

^ (S10N>i) G33dS QNl/a

1-31
1-32
1-33
EXTREME WINDS IN THE UNITED STATES

by

Arnold R. Hull

and

Patrick E. Hughes

Environmental Data Service

National Oceanic and Atmospheric Administration

U.S. Department of Commerce

The highest winds reported in the United States have been associated

with tornadoes. Fujita (1971) has developed a tornado windspeed/damage classification system which permits extreme wind estimations without follow-up surveys. Fujita classifications of tornadoes for the 1965, 1971, and 1972 seasons,

as well as extreme wind values associated with thunderstorms and extratropical

cyclones during 1973 are reviewed as they relate to ANSI building code require- ments for design loads. Next, a new observational tool, an acoustic, dopplershift "sonar" capable of profiling low-level wind regimes at the actual building site, is briefly described. Finally, the proposed NOAA Severe Environmental

Storms and Mesoscale Experiment (SESAME) is examined. SESAME is an observational/

research effort to identify the processes and controlling parameters of extremewind generating severe weather systems such as squall lines, thunderstorms, and

possibly tornadoes, and to aid in the development of conceptual and numerical

models of these phenomena.

Key Words: Building code; damage classification; extreme wind; tornado; wind loads

1-34
TORNADOES AND EXTREME WINDS

The highest winds and the greatest wind-related structural damage that occur in the

United States are associated with tornadoes. Since anemometers are seldom located in the

immediate area of a tornado and could not survive the highest winds if they were, windspeeds are estimated from structural damage and by other subjective methods. Engineering

estimates reveal, for example, that winds up to 350 mph lasting a few seconds are likely

to produce the most tornado damage in the Midwest (Fujita, 1971) . Based on estimated windspeeds and photographs of actual tornado damage, Fujita developed the following windspeed/

damage relationships, designed to permit extreme wind estimations without special damage

surveys

.

FO 40-72 mph, LIGHT DAMAGE

Some damage to chimneys and TV antennae; breaks twigs off trees; pushes over shallow rooted trees.

Fl 73-112 mph, MODERATE DAMAGE

Peels surface off roofs; windows broken; light trailer houses pushed

or overturned; some trees uprooted or snapped; moving automobiles

pushed off the road. 73 mph is the beginning of hurricane wind speed.

F2 113-157 mph, CONSIDERABLE DAMAGE

Roofs torn off frame houses leaving strong upright walls; weak buildings in rural areas demolished; trailer houses destroyed; large trees snapped or uprooted; railroad boxcars pushed over; light object missiles generated; cars blown off highway.

F3 158-206 mph, SEVERE DAMAGE

Roofs and some walls torn off frame houses; some rural buildings com- pletely demolished; trains overturned; steel-framed hangar-warehouse

type structures torn; cars lifted off the ground; most trees in a forest uprooted, snapped, or leveled.

F4 207-260 mph, DEVESTATING DAMAGE

Whole frame houses leveled, leaving piles of debris; steel structures

badly damaged; trees debarked by small flying debris; cars and trains

thrown some distances or rolled considerable distances; large missiles

generated.

F5 261-318 mph, INCREDIBLE DAMAGE , , .

Whole frame houses tossed off foundations; steel-reinforced concrete

structures badly damaged; automobile-sized missiles generated; incredible phenomena can occur.

F6 319 mph to sonic speed, INCONCEIVABLE DAMAGE

Should a tornado with the maximum windspeed in excess of F6 occur, the

extent and types of damage may not be conceived. A number of missiles

such as ice boxes, water heaters, storage tanks, automobiles, etc. will

create serious secondary damage on structures.

1-35
1973 TORNADO SEASON

Last year 1109 tornadoes occurred throughout the United States on 208 days, shattering the old records of 929 tornadoes (1967) and 194 annual tornado days (1972) . Tornadoes

were reported in 46 States, skipping only Alaska, Rhode Island, Utah, and Washington.

The tornado season began on January 18 and ended on New Year's Eve. During the interval, 87 people were killed, 2481 injured, and property losses exceeded $500 million. Texas led the Nation in tornado fatalities (14) , while Georgia recorded the most property

damage, with estimates exceeding $150 million. Most of the Georgia losses were the result

of a single storm (March 31) , which moved east-northeastward through north-central Georgia

causing extremely heavy and almost continuous damage along a 75-mile path. A State survey

team estimated total damage at more than $113 million —the largest loss for any natural

disaster in the State's history.

More recently, during the afternoon and evening of April 3, 1974, an outbreak of tor- nadoes killed more than 350 people in the United States and Canada in eight hours —three

times as many people as had been killed by tornadoes in the entire previous three-year

period. The area hardest hit extended from northern Alabarra and Georgia, across Tennessee

and Kentucky into Indiana and western Ohio. In Xenia, Ohio a swath of almost complete devastation was cut through the center of town.

As you know, new distributions of extreme wind in the United States were incorporated

into the revised American National Standards Building Code Requirements for Minimum Design

Loads in Buildings and Other Structures, published in 1972. Based on the work of Thom

(1968) of NOAA's Environmental Data Service, the new distributions increased the average

period of data records used from 15 to 21 years, and the accuracy of the values by about

15%.

Although the ANSI Code does not consider tornadoes in its extreme wind distributions,

many engineers must consider tornado winds in the design of such vulnerable structures as hospitals and nuclear powerplants. Indeed, at this moment, Texas Tech University in Lubbock is conducting a four-day short course on "Engineering for Extreme Winds and Tornadoes.'

The course includes "a balanced risk approach to structural design which may be used to provide occupant protection and economic tornado resistant designs...."

In a recent, related paper, "Residential Buildings Engineered to Resist Tornadoes,"

Sherman (1973) adopts two maximum loading conditions to design and render a typical ranchstyle, wood- frame house tornado-resistant. For the maximum loading conditions recommended,

the increase in construction costs amounts to only 19%.

Given the current engineering needs and interest, the advent of the Fujita classification system seems particularly timely. Since 1971, experimental Fujita estimates of tor- nado intensities have been made for practially all tornadoes reported. Estimates are also

available for 1965 tornadoes from Tecson (1972) . The table below is reproduced from

Fujita and Pearson (1973) and shows the percentages of tornadoes reported by Fujita cate- gories .

EXTREME WINDS AND ANSI DESIGN LOADS

F Scale 1965 1971 1972

0

1

2

34

5

33%

44

17

5

10.1

893

20%

4226

82.5

0.2

888

.23% 46

24

60.8

0.0

Total 740

1-36
As you can see by comparing the Fujita classifications presented earlier and the ANSI

100-year recurrence values for basic windspeed (Figure 1) , all tornadoes in categories F3

and above exceed the ANSI standard values and, except in a few hurricane-prone coastal areas of the Gulf and Atlantic States, so do the tornadoes in category F2. These categories

total approximately 23%, 37%, and 31% of all the tornadoes reported in 1965, 1971 and 1972,

respectively. In numbers, this works out to approximately 205, 329, and 229 tornadoes for

these years. I might add parenthetically that the minimum design values on the 100-year

recurrence chart is centered over southern Arkansas, in the heart of tornado alley.

Leaving tornadoes aside for the moment, and considering only extreme winds associated

with thunderstorms and extratropical cyclones during 1973, 74 reported values exceeded the

100-year recurrence values. This in no way casts doubt on the validity of the standard

values, but it does reflect the fact that the ANSI mean recurrence charts are based on only

21 years of climatological record. It would appear desirable to update these calculations

when say 25 years, then 30 years of data become available. In each instance, the increased

data sample should increase the accuracy of the standard extreme values.

ONSITE WIND MEASUREMENT

A promising new remote wind-measuring instrument, an acoustic echo sounder potentially

capable of profiling low-level wind structures at actual building sites —as well as those

associated with tornadoes —is now being tested by NOAA Environmental Research Laboratories

(ERL) scientists. Developed by the ERL Wave Propagation Laboratory under an agreement with

the Federal Aviation Administration (FAA) , the sounder operates like sonar; a sound pulse

is transmitted into the atmosphere and scattered by variations in the acoustic refractive

index caused by temperature and wind fluctuations. Specifically, wind fluctuations produce

doppler frequency shifts from which both vertical and horizontal wind components may be

determined

.

Work to date indicates that it is possible to determine wind profiles to altitudes of

1 km with the sounder (Beran, et al. , 1973). In addition, the simplicity and relatively

low cost of acoustic sounders as compared to radar or lidar makes it probable that such

sounders will play an increasingly important role in boundary layer wind studies.

MODELING EXTREME-WIND GENERATING SYSTEMS

The atmospheric mechanisms which determine the type, severity, and variability of

local weather events which generate extreme winds are mesoscale systems. Until recently,

however, most atmospheric studies have concentrated on micro- and macroscale phenomena.

Interest in the mesoscale has quickened in the past few years due largely to the development of doppler radar and acoustic sounders, which appear capable of defining the threedimensional motion field of convection storms. This, in turn, has quickened interest in

the three-dimensional numerical simulation of severe storms.

The proposed NOAA Severe Environmental Storms and Mesoscale Experiment (SESAME) is an observational/research effort designed to identify the processes and controlling parameters

of extreme-v/ind-generating systems such as squall lines, thunderstorms, and possibly tornadoes, and to aid in the development of conceptual and numerical models of these phenomena.

Satellite, acoustic echo sounder, radar (doppler, continuous-wave, and pulsed), radiosonde,

and surface and aircraft observations —coordinated in a time and space framework pertinent

to the phenomena studied—will be fed into computers to develop and test mesoscale numerical models.

A preliminary shakedown period is scheduled for the Fall of 1976 to provide for initial calibration and evaluation of the data system, and to allow a short period for minor

alterations prior to the first three-month observational program, planned for the Spring

of 1977. A second observational period is planned for the Spring of 1979. The 21-month

gap will allow time for analysis of the 1977 observational data and permit revisions of

any inadequate data acquisition procedures.

1-37
ACKNOWLEDGMENT

The authors would like to thank Allen Pearson, Director of the National Severe Storms

Forecast Center, and Frank Quinlan and Vincent Cinquemani of the National Climatic Center

for their considerable contributions to this article.

REFERENCES ' - " '

1. Fujita, T.T. , "A Proposed Characterization of Tornadoes and Hurricanes by Area and

Intensity," SMRP Paper 91, University of Chicago, 42 pp., 1971.

2. Sherman, Z. (Dept. of Aerospace Engr. , Pennsylvania State University, University Park),

"Residential Buildings Engineered to Resist Tornadoes," American Society of Civil Engineers, N.Y. Structural Div., Journal, 99 (4) : 704-714 , April 1973. Refs. DAS

3. Tecson, J.J., "Characterization of 1975 Tornadoes by Their Area and Intensity," SMRP

Paper 94, University of Chicago, 22 pp., 1972.

4. Fujita, T.T. and A.D. Pearson, "Results of FPP Classification of 1971 and 1972 Tornadoes

Eighth Conference on Severe Local Storms, AMS Conference Proceedings, Denver, Colorado,

pp. 142-145, October 1973.

5. Beran, D.W., B.C. Willmarth, F.C. Carsey, and F.F. Hall, Jr., "An Acoustic Doppler Wind

Measuring System," J. Acoust. Soc. Am., Vol. 55, No. 2, pp. 334-338, February 1974.

1-38
1-39

WIND TUNNEL EXPERIMENTS FOR STUDYING A LOCAL WIND

by

Ken Suda

Director, Meteorological Research Institute (MRI)

Seiji Soma

Physical Meteorology Laboratory, MRI

Kiyohide Takeuchi

Applied Meteorology Laboratory, MRI

In general, the study of the effects of local winds on structures is primarily based on data obtained from wind tunnel experiments. As an example of

such studies, an investigation of air flow around Nakatojima Island relative

to the wind resistant design of a long suspension bridge has been conducted

and will be presented.

However, it is not reasonable to conduct such a study entirely related to wind tunnel experiments, as the similitude rule has not been established. Therefore, field observations have been made in parallel with the wind tunnel experiments and the data obtained by both sources are then compared. Thus, the re- sults obtained from the wind tunnel e:)^eriment are more reliable.

Key Words: Bridge; gust; metorological data; topographical model; wind load; wind

profile; wind tunnel.
INTRODUCTION

In studying the effects of local winds, the collection and analysis of field data is,

needless to say, the most desirable. In practice, however, the acquisition of field data

is laborious and requires a vast amount of time and expense, therefore, wind tunnel tests

are conducted despite various difficulties notably the principle of similitude.

Because of these difficulties, wind tunnel tests were conducted in conjunction with

field observations, and then comparisons of the data obtained from both sources was made.

As will be noted, the field data does not always agree with the wind tunnel test data.

In collecting the data a reference point should be fixed so there is similarity be- tween the tunnel and the natural winds and they are optimum. Therefore, the location of

the reference points and the physical elements to be observed are usually determined during

the planning of wind tunnel experiments. The number of the observation points, the elements to be observed, and the time required are generally greater than would be allowed

because of budget restrictions. Also, the principles of similitude, which are most reliable in the area of effective Reynolds number, must be employed.

Based on these conditions, the effects of local winds on structures have been studied

by using a wind tunnel. An example of the type of wind tunnel test study that can be con- ducted will be presented in relation to an investigation of the wind regime around the

Nakatojima Island and the wind-resistant design of a long suspension bridge. Numerous

other studies have been made in our laboratory, as detailed in Table 1.

INVESTIGATION OF THE WIND REGIME AT THE BRIDGE BUILDING SITE

NEAR NAKATOJIMA ISLAND

In designing a suspension bridge for wind loading, the upward attack angle of the wind

against the bridge is an important matter, and must be carefully considered. The wind effects, due to the land topography, were initially evaluated during a preliminary survey

prior to the building of the Second Giant Bridge of Kurushima, which connects Honshu and

Shikoku. The primary area where wind conditions are important is Nakatojima, a little

island located four kilometers to the north of Imabari in the Shikoku District. In the

construction of the bridge, the bridge axis is planned to pass north of and parallel with

the ridge line of the island. Therefore, a considerably steep up-draft air flow can be ex- pected during a northerly wind. Although the air current may not necessarily affect the

entire span of the bridge, its vertical inclination can be exceedingly large and the

region in which the air current appears from the different wind directions is not known.

In case of a southerly wind, the bridge axis will be located on the leeward side of

Nakatojima. Although the level of the bridge girder is a few meters higher than the island

top, a violent turbulence is expected to dominate the leeward side air current. Therefore,

the study included both the northerly and southerly wind directions.

The primary investigation was directed toward the wind tunnel experiments, however,

in order to compare these results with those in the field, actual observations of the wind

direction and speed were made at two points on Nakatojima Island during a three-month

period. Location of the observation sites and the level of the instruments are given in

Figure 1. Stations are distributed in such a way that the difference in wind speeds at two points in the island and the existence of separation phenomenon expected in the leeward side of the island can be examined.

TOPOGRAPHIC CONDITIONS

Nakatojima Island is located to the north of Imabari, approximately in the middle of

Kurushima Straits. It is a small island, spanning a distance of 390 m. east to west and j 180 m. north to south with a height of 63 m. above sea level. On its northern and souther'

shores, the slopes have an inclination of 40°. With this background, it can be easily

understood that the peculiar topography of the island may provide strong air currents on
the island in addition to the fact that there is no nearby island which may affect the

air current.

METHOD OF INVESTIGATION

(i) Wind Tunnel Experiment

For the wind tunnel experiment, three topographic models with reduced scales of 1/500,

1/1,000, and 1/1,500 were prepared. The first model which was used in the experiments,

in conjunction with the Gottingen-type wind tunnel (1.5 m. in diameter), distributions of

the wind angle and wind speed around Nakatojima, were measured in detail.

In order to study the behavior of the air current across the island, in case of stable

stratification of the atmosphere, a second model was prepared.

The last model was used to study the topographic effect of a wider area, including

Nakatojima and other adjacent islands, relative to the air flow.

(ii) Field Observation

In conducting the field tests, aerovanes were placed at two points on the island, one

at the top of the island (Point A) and the other on the northern slope (Point B) in order

to obtain data on the wind direction and speed. The levels of the aerovanes are 76.0 m. and 29.5 m. , respectively. Point B, which is located at about half of the height of the

island and susceptible to topographic effects, was chosen as an observation point to check

similitude.

WIND TUNNEL EXPERIMENT

(i) Measurement and Analysis of the Vertical Inclination Angle

The vertical inclination angle of the air current over the northern slope of Nakatojima Island was the principle object of the present investigation. In the wind tunnel ex- periment, the vertical inclination was generally determined from the measurement of the U and W coit^onents of the wind, making use of the X-proble of the hot-wire anemometer. However, in order to improve the precision of measurement, a continuous record of the vertical inclination was made. This requires a modification to the electric circuit of the

instrument, the details of which are not given herein. The tunnel wind speed that was

used in the experiment was 7.0 m./s. for both the vertical inclination and wind speed distribution. Results of the measurement for each wind direction will be given as follows:

Case of Northerly Wind • '

In this case, the wind blows perpendicular to the bridge axis. As seen from

the distribution curve of the vertical inclination given in Figure 2-A, there is a very strong up-draft current which has a maximum of 25° relative to the northern

slope. This maximum inclination naturally corresponds to the top of the island.

The up-draft of the wind is maintained for a considerable distance, even when the

point of measurement is over the waters off the western edge of the island, as shown

in Figure 2-A.

Case of Southerly Wind

Figure 2-B shows the result of the measurement for the case of the ciirrent blowing perpendicularly to the bridge with a southerly wind direction. In this case, the

bridge girder seems to be in the wake of the island and thus, the vertical inclination

changes so strongly that its upward or downward sense cannot be determined from the

distribution curve.

II-3
(ii) The Measurement and Analysis of Wind Speed Distribution

The experiments were performed with the same topographic model, as described previously, with the same measuring line and points as stated above.

Case of Northerly Wind

The wind distribution along the bridge axis, as induced by the northerly wind,

is shown in Figure 3-A. As seen by the distribution curve at that level where the

bridge girder is located (second curve from the top in Figure 3-A) , there is a ten- dency for the wind to decrease in the middle of the island and increase on both ends,

relative to the tunnel wind speed. This tendency is especially prominent in the

western part of the island.

Case of Southerly Wind

In this case, the distribution curve of the wind speed shows unusual characteristics (see Figure 3-B) . Namely, on the leeward side at the middle of the island,

the wind is exceedingly weak, being 1/7 or less of the wind tunnel speed. The dis- tance where the wind speed decreases substantially is 200 to 250 m. , after conversion

to actual scale. On the eastern and western ends, the wind speeds conversely in- crease as the distance increases.

The decrease of the wind speed due to the southerly wind can be understood if we consider that the measuring line along the bridge axis is in the wake of the island.

However, since the decrease of the wind speed was drastic, the level of the boundary

surface between the general air current and the reverse air current due to separation

was measured. The results of these measurements are shown in Figure 4. Although the

level of the boundary surface changes strongly and is difficult to determine precisely,

a general trend can be established. As seen in Figure 4, the level of the boundary is

highest over the northern slope of the island and its height is nearly the same as that of the bridge girder.

In principle, this boundary surface represents the discontinuity between the general and reverse currents and is theoretically the surface on which the wind speed is

reduced to zero. Accordingly, it is clear that the rapid decrease of the wind speed,

as seen in the distribution curves given in Figure 3-B, is due to the coincidence in

height of the bridge girder and the boundary surface.

(iii) Experiments in the Stratified Wind Tunnel

In order to visually obtain a pattern of the air flow passing over Nakatojima Island,

and to establish the effect of the stratified atmosphere on the pattern, experiments were | made with a model built to a scale of 1/1,000, and tested in the stratified wind tunnel.

The first investigation was conducted using a stable air flow pattern. Plate 2 shows : the pattern when the wind speed is 0.4 m./s. and the vertical density gradient is 0.006 | gr./cm.^. A stagnant vortex, almost round, was found in the windward side of the island.

j Note that the wind passing over the top of the island tends to descend on the leeward

j side. In such a case the boundary surface also decreases on the leeward side. Separation i does not completely disappear and the reverse flow still exists. [

The Froude number, which represents the stability in the atmosphere, is approximately j: 1.79 for the present experiments, which can be found in the real atmosphere. Accordingly, j the flow pattern obtained in the experiments can exist in the atmosphere when the wind is I light and the stratification is stable. i

The next experiments that were conducted were under the condition that the wind speed I was 1.5 m./s., the stratification was neutral, and a grid was installed in order to make i the air flow turbulent. This pattern is shown in Plate 3. The streamline, in contrast

with Plate 2, still ascended after passing over the top of the island. j:

II-4
(iv) Experiments with a Regional Topographic Model and Their Analysis

Experiments on a regional topographic model, built to a scale of 1/1,500, were made

so as to include not only Nakatojima Island but adjacent islands. These islands can make

the flow pattern much more complicated and should be considered in the bridge design.

Measurements of the vertical inclination of the air flow in the northerly direction

shows that the wind blows up on the side of Nakatojima Island and Umajima Island (see Figure 5). A horizontal air flow can only be found at the central part of the straits.

The wind speed distribution has no marked characteristics, when the wind comes from

the north, but the southerly winds show characteristics of a strait wind: i.e., the flow

coverages and the speed increases all over the strait. Its increase rate is high at the

coast of both islands and low at the central part of the strait (see Figure 6)

.

ANALYSIS OF FIELD METEOROLOGICAL DATA

In order to evaluate the reliability of the wind tunnel experiments, meteorological

data was obtained from Nakatojima Island and then analyzed. The aerovane at the top of

the island (Point A) was installed on the pole 15 m. above the ground, thus the wind speed

can be considered to be general air flow. The wind at the north slope of the island

(Point B) is cort5)ared with that obtained at Point A.

The results are depicted in Figure 7, where the wind is represented as a vector and

the figure in the parentheses shows the frequency. Only the following wind directions

were studied, because these have the main influence on the bridge:

Southerly Wind: SW, SSW, S, SSE, SE Northerly Wind: NW, NNW, N, NNE , NE

The winds were divided into three groups: weak (0.5-5.0 m./s.), moderate (5.1-10.0

m./s.), and strong (higher than 10.1 m./s.), because separation depends upon the wind

speed.

Case of Southerly VJind

When the general wind (namely, the wind at Point A) is southerly, the local wind

at Point B has almost an opposite direction (see Figure 7-A) , with separation taking

place on the north slope of the island. Separation does exist even during a light

wind. The wind speed at Point B is low, but increases almost proportionally with an increase of the general flow.

Case of Northerly Wind

When the general flow is northerly, the local wind at Point B has a rather large

fluctuation of the direction, but has almost the same direction (see Figure 7-B)

.

FLOW IN THE WIND TUNNEL AND NATURAL WIND

In order to evaluate the results obtained from the wind tunnel, meteorological observations were made at two points on Nakatojima Island. Relationships between the air flow

in the wind tunnel and the natural wind was then studied. The following will describe

these relationships:

(i) Separation

During a southerly wind, separation was found over the north slope for both cases.

(ii) Ratio of Wind Speeds

II-5
The ratio of the wind speeds, between the two points, is shown in Table 2. In the northerly wind, the wind speed in the wind tunnel is a little lower than that observed at the actual sites. These values, however, seem to be in good agreement, con- ) sidering the complicated characteristics of the winds.

In the northerly wind, when the flow pattern is much more complicated, the ratios || show that the flow seems to agree fairly well between the model experiments and field

observation.

In Table 2, the ratio of wind speeds as observed at Points C and D (see Figure

1) that were obtained a few years ago are presented. These values were obtained at

a tower (110 m. high above the ground) which was used to support electric power lines,

and at another tower (51 m.) used as a signal station for tidal current.

(iii) Similarity of Turbulence

In the present experiment, a grid was placed in the wind tunnel as a turbulence

generator (see Plate 4). The grid consists of 40 mm. of square mesh, constructed of " 10 mm. square bars. The turbulence thus generated should be similar to that developed

by a natural wind. The similarity has not been completely studied, however, the grid

seems to generate effective patterns which are similar for both cases (for example,

separation and stagnant vortex)

.

(iv) Effective Reynolds Number

A complete solution has not been obtained of the similitude between the model in

the wind tunnel experiment and the field observation. However, when the effective

Reynolds number has the same value for both cases, the similarity seems to hold with j a high degree of accuracy, at least under the condition that the strong wind passes over the isolated island such as Nakatojima. Here the effective Reynolds number is I defined as the number using the turbulent diffusivity, K, in place of the kinematic

viscosity, p. The effective Reynolds number is expressed as follows:

j Rp = HI (1)|

where U and L are the wind speed and the length, respectively.

However, from the theory of turbulence, K = iu' , where £ is the representative

length of the "eddies," and u' is the RMS of the fluctuation of the turbulence. Ac- cordingly, the following equation is obtained:

Re = — ^^'> iu' The term U/u' is the reciprocal of the intensity of the turbulence and L/l is the

ratio of the representative length of the topography to that of the eddies. Consequently, an adjustment of the effective Reynolds number can mean an adjustment to both of these two terms. Both terms can be measured during the field observation,

and they can be adjustable in the wind tunnel. From existing data obtained in the

field, Rg can be estimated as 10 to 100. In the present experiments R^ in the wind

tunnel is about 300. It means that the grid size is smaller, considering that the

model is made on the scale of 1/500. Investigation of the effect of the difference

of Rg has not been made

.

CONCLUSION

Difficulties in the wind tunnel experiments are due to the fact that the similitude

rule has not been established. However, because it is difficult to conduct intensive observations at the site and to analyze the long-term data, a combination of the wind tunnel

experiments and rather simple field observations seems appropriate. In this manner the

wind conditions at Nakatojima Island were examined and described herein.

II-6
Table 1 . Examples of Studies Made By

Wind Tunnel Experiment

Theme of Study

Clear Air Turbulence on the leeward side of Mt.

Fuj i

Point of Reference

Distinction of wind regimes

whether it is mountain wave

type or separation type

Observation Point and

Elements observed

Tracking of non-buoyant

balloon flown from summit

by a tracking radar

Turbulent flow around Distinction of turbulent Double theodolite observaan airport wind systems on the lee- tion of balloons over the

ward side slope of a hill runway and accelerometric

observation of aircraft

vibration

Strong winds around a high building

Comparison between general

wind speed in urban area and

the wind speed near the

building

Svirface wind observation

at the roof-top of a building and around it

Determination of the

anemometer level for

Mt. Fuji Weather

Station

Vertical inclination of

air flow

Observation of vertical

inclination by emitting

smoke from the summit

Strait wind at the Vertical distribution of Observation of vertical

Kammon Straits wind speed at a cape fac- distribution of wind on ing to the Straits the power transmission

tower

Strait wind at Mekari

Straits

Comparison of winds at three points on the coast of Strait and on the sea

Wind observation by aerovane and observation of

strait wind by non-buoyant

balloon flown from three

points on the sea

Wind regime around

Nakatojima Island

Ratio of wind speeds at the top and on the slope

of the island and the

separation

Observation of wind direction and speed at the top

and on the slope of the

island.

II-7
Table 2. Ratio of the wind speeds

j Wind direction NW NNW N NNE NE SSW SW S SSE SE

wind tunnel

experiment 0. 79 0.64 0.54 0.54 0.51 0.13 0.20 0.18 - 0.23

field 0.92 0.73 0.73 0.62 0.67 0.30 0.18 0.12

observation

wind tunnel

experiment 0.91 1.00 1.08 0.96 0.94 0.96 0.94 - 0.46

field 0.92 0.98 0.99 0.86 0.90 0.81 0.92 - 0.79

observation

II-3
Plate 1. Nakatojima Is. viewed from the eastern sea. Northern and southern

slopes are very steep.

Plate 2. Pattern of air flow in case of highly stable stratification and weak

wind (experiment in the stratification wind tunnel). Air flows down

after crossing the ridge of island.

II-9
Plate 3. Pattern of air flow in case of neutral stratification and strong wind.

Air continues to flow up after crossing the ridge of island.

Models of Nakatojima Island and the Second Giant Bridge of Kurushima

in the wind tunnel (scale: 1/500) and grid of turbulence generator on

the left.

11-10
11-11
Fig. 2—A. Vertical inclination along the bridge axis on the northern side of

Nakatojima obtained by wind-tunnel experiment. Wind direction

N-NNE.

11-12
HEIGHT I7CM(85M)

Fig. 2-B. Vertical inclination along the bridge axis on the northern side of

Nakatojima obtained by wind-tunnel experiment.

Wind direction SSW-S.

11-13
HEIGHT I7CM(85M)

Hi III I II Ill II I

Fig. 3 -A. Distribution of wind speed along the bridge axis on the northern side

of Nakatojima obtained by wind-tunnel experiment.

Wind direction N-NNE.

11-14
HEIGHT I7CM(85M)

—

lOOm NAKATOJIMA

Fig. 3—B. Distribution of wind speed along the bridge axis on the northern side

of Nakatojima obtained by wind-tunnel experiment.

Wind direction SSW-S.

11-15
11-16
4

Fig. 5. Vertical inclination of the wind obtained with a regional topographic

model including both Nakatojima and Umajima Islands.

Scale 1/1,500, Wind direction N-NNE

11-17

MODERATE

. 7-A. Relationship between the wind at Points A and B in Nakatojima

Island. Wind direction at Point A SSW

11-19
Fig.7-B. Same as in Fig.7-A but wind direction at Point A NNE

11-20
A STUDY OF WIND PRESSURES ON A SINGLE-FAMILY DWELLING

IN MODEL AND FULL SCALE

by

R. D. Marshall

Structures Section

Center for Building Technology, lAT

National Bureau of Standards

Washington, D.C. 20234

Wind pressures measured on a single-family dwelling are compared with results obtained

from a 1:50 scale model placed in a turbulent boundary layer. It is shown that the fluctuating components of surface pressures far exceed the mean or steady pressures and are well

correlated over sizeable roof areas. The consistently low fluctuating pressure coefficients

obtained from the wind tunnel model are attributed to improper simulation of the lower

portion of the atmospheric boundary layer. Comparisons between actual loads and specified

design loads suggest that certain current provisions are marginal for tributary areas and

excessive for localized areas such as ridges, eaves and corners. A procedure for expressing

loads on both localized and extended roof areas in terms of mean pressure coefficients and

a peak factor is described.

Key Words: Aerodynamics; boundary layers; buildings; codes and standards; wind loads;

wind tunnels.

11-21
1.0 INTRODUCTION

Research over the past ten years into the effects of wind on buildings and

other structures is significantly influencing design philosophy and practice as

is evidenced by recent major revisions of building codes and standards, both

here in the United States and abroad. Perhaps the most significant improvement

has been the recognition of wind loading as a stochastic process and the formulation of design criteria based upon acceptable levels of risk. Other improvements include provisions for various classes or categories of terrain roughness

and the wind-tunnel simulation of the atmospheric boundary layer when measuring

pressure coefficients and dynamic response factors.

In spite of these important advances, considerable work remains to be done.

This is particularly true of those code provisions covering the design of lowrise buildings which have not benefited from this research in the same proportion as tall structures. It is interesting to note that pressure coefficients

in current use are based in large part on wind tunnel studies carried out in uniform flows of low turbulence, using instrumentation capable of measuring only mean pressures. It is obvious, therefore, that a major effort must be made in

the wind-tunnel modeling of low-rise structures. Prior to this, however, the

validity of modeling techniques must be established by comparing test results

with representative measurements obtained from full-scale buildings. The investigation described in the following sections is an attempt to provide a preliminary comparison between model and full-scale test results for a single-family

dwelling.

2.0 THE TEST SITE

The site selected for this study is located at Malmstrom Air Force Base, Montana,

directly to the east of the city of Great Falls. The region is noted for its "chinook"

winds which regularly blow out of the southwest during the winter months. The terrain

surrounding the test site is markedly flat and free of significant obstructions. Although

the mean hourly speeds for the area are quite high (14-17 mph) , extreme winds seldom exceed

70 mph.

The building investigated is a single-family dwelling, one of four quite similar units

located in an area having a clear wind exposure extending from the west clockwise aroxind to the south. A cluster of two-story housing units is located approximately 300 feet southwest of the test site and extends in that direction for approximately 1800 feet (see Figure

1) . The test building and adjacent structures are shown in Figure 2. The test building

has basic plan dimensions of 22 x 75 feet with a 16 x 19 feet wing (Figure 3) . The roof

pitch is 11.5 degrees and the eaves overhang is 2.5 feet.

11-22
3.0 FULL-SCALE TEST PROCEDURE

Surface pressures were measured at nine points on the roof (P1-P9) and at one point

under the eaves (PIO) . In addition, the internal pressure was measured in the garage area

(P12) . An additional pressure tap (Pll) was installed under the eaves in the wind-tunnel

model. To avoid penetrating the roof membrane with pressure taps and at the same time

accurately measure pressures over the roof surface, the pressure transducers were mounted

under low-profile housings having a height of 1.4 inches and a diameter of 2 feet. All transducers were referenced to a vane-mounted pitot static tube located 9.5 feet above the ridge

line, i.e., 20 feet above ground level.

Wind speed and direction were obtained with a propeller-vane anemometer located 21.5

feet above ground level. A standard National Weather Service three-cup anemometer was

used to trigger the data acquisition system when wind speeds exceeded a preset level.

Positions of the pressure transducers, pitot static probe, and anemometers are shown in

Figure 3.

A fourteen-channel analog tape recorder was used to acquire data. In addition to the

eleven pressure signals, wind speed, and wind direction, a time code was recorded to identify the data runs or records. Normal operating procedure was to set the system threshold

at 40 mph and record for fifteen minutes. The system would then enter a one-hour hold,

period before checking the wind speed against the preset value. While this procedure re- sulted in a number of redundant records, it did provide data corresponding to peak winds

in winter storms passing through the region. These storms generally had a duration of

from one to three days. Barometric pressure, temperature, and other weather data were obtained from hourly observations made by the 3rd Weather Wing, USAF.

4.0 WIND TUNNEL TEST PROCEDURE

In order to increase the value of the full-scale test results, aid in their interpretation and explore the feasibility of modeling the natural wind at an unconventional scale,

a series of wind tunnel tests were conducted during the course of the study using a model

scale of 1:50.

The tunnel used for these tests is one of several operated by Colorado State University and has a 6 ft. square by 40-ft. long working section. To simulate the natural wind,

a thick shear layer was established by use of a row of spires and a sawtooth fence installed at the working section entrance. In addition, all adjacent and upwind structures

were modeled in the tunnel. This approach has been described by Melbourne (1) and Standen

(2) and when compared with results obtained using surface roughness elements alone, it substantially increased the growth rate of the boundary layer and increased the scale of tur- bulence at the position of the model, 33.7 feet downstream from the spires. Spire details

and position of the model are shown in Figure 4.

11-23
Pressures were measiared with the same transducers and recording system used for the

full-scale study. The orifice-tube combinations exhibited a satisfactory frequency re- sponse out to 100 Hz. All pressure measurements were referenced to a pi tot static tube

located four feet above the tvinnel floor, directly over the building model. Wind speed

records and static and dynamic pressures were obtained by means of a hot-wire anemometer

and pitot tube located in the same relative positions as the instruments in the fullscale study. In addition, mean velocity profiles were obtained at the position of the

model for each of the four wind directions studied.

5.0 DATA REDUCTION AND ANALYSIS

Analog tapes containing full-scale and wind tunnel data were processed using the data

system described in Ref. 3. The usual procedure was to plot the full-scale wind speed and

direction records on a stripchart recorder and to then select those records exhibiting a satisfactory degree of stationarity for detailed analysis. Analog to digital conversion

was accomplished at a rate of sixteen samples per second with a total of 12,000 samples

per record or a digital record length of 750 seconds. By using a reduced playback speed,

it was possible to digitize the wind tunnel data at the same effective rate and account for

the 50:1 change in time scale. Thus the time and frequency scales were assumed to be

identical in the subsequent data analysis. With the exception of the hot-wire records, all

records were obtained from transducers exhibiting linear characteristics.

A series of computer programs have been developed at the NBS for the analysis of ran- dom data. These include PROGRAM 2 which formats sequential channel samples into sequential samples for a given channel; CORREL which performs low-pass filtering and contains

options for computing the mean, rms, auto- and cross-correlation, spectral density and

coherence functions; PDF which computes probability densities and tabulates peak values and

associated zero-crossing rates; and SUMP which generates a new data series based on the area integration of surface pressures. In addition, subroutines exist for linearizing hotwire records and correcting fixed-direction propeller anemometer records for departures

from the cosine law.

In the study reported herein, no attempt was made to remove trends from the data series

prior to analysis, the records being visually screened before conversion. However, even with this screening, there were certain records processed which indicated significant

trends as reflected by their segmental means and auto-correlation functions. Most steps

in the analysis were preceded by low-pass filtering, each four successive samples being

averaged and resulting in a record size of 3,000 samples. Auto-correlations were calculated

for 200 lags, followed by a 0.02 Hz fixed-bandwidth spectral analysis.

6.0 MEASUREMENT RESULTS AND DISCUSSION

Field measurements at the Montana test site were obtained over a five-month period

during 1971-72, yielding approximately fifteen hours of recordings under strong wind con- ditions. Four different wind directions were selected for subsequent wind tunnel simulation on the basis of differing obstructions over the wind fetch. Due to demands placed on test equipment, only a preliminary analysis of the field data was available at the time

the wind tunnel studies were conducted. This proved to be unfortunate in that a better

simulation of the natural wind could have been achieved with slight additional effort.

In the following sections, run numbers with three digits designate full-scale data

and wind azimuth angles (3) are measured clockwise from north.

Simulation of the Atmospheric Boundary Layer

Because of the model scale used in this study and the physical size of the wind

tunnel facility, a simulation of only the lower 100 feet of the atmospheric boundary

layer was attempted. Restrictions placed on the use of the field test site did not

allow the installation of a meteorological tower of sufficient height to establish

II-2U
characteristics of the atmospheric boundary layer over this range. Field measurements

of wind speed were thus limited to one point, 21.5 feet above ground level. As will

be discussed later, this necessitated corrections to the mean reference speeds and

pressures during the data analysis.

Typical mean velocity profiles measured at the position occupied by the model

(33.7 feet downstream) are shown in Figure 5 as a power law representation. In plotting these profiles, it was assumed that the actual scale of the shear flow was 1:50

and that the thickness of the atmospheric boundary layer was 900 feet. The first profile corresponds to the case of only the spires and sawtooth fence installed and ex- hibits a satisfactory velocity distribution only up to a full-scale height of approximately thirty feet. The second profile corresponds to an azimuth angle of S = 341

degrees, the direction having the least number of structures upwind. Although the

profile agrees well with the power law up to approximately ninety feet, there is a substantial departure above this height. Only for azimuth angles of 186, 211, and

256 degrees did the mean velocity profiles correspond to the power law above 100 feet.

The departure of the profile for S> = 211 degrees from the power law for Z/Z_ < 0.025

IS due to the neighboring house directly upwind.

While the exponents obtained from the plots in Figure 5 are in good agreement

with recommended values for terrain typical of the test site, there is nothing to suggest that the scale ratio is in fact 1:50. The profiles for g = 211, 256, and

341 degrees are plotted as a log law in Figure 6. The corresponding roughness

heights (Zq) , were found to be 0.12, 0.23, and 0.007 inches (0.30, 0.58, and 0.02 cm).

Helliwell (4) determined a value of 8 cm. for open country at Heathrow and Cardington.

This would suggest a scale ratio of from 1:15 to 1:400 for the atmospheric boundary

layer simulation. Only the very lowest portion of the profile for 3 = 341 degrees

corresponds to the log law and it is suggested that the scale ratios associated with

the peaks of the turbulence spectra are more meaningful.

Spectral density fxinctions for 6 = 256 degrees (Run Nos . 311 and 22) and Z = 21.5

feet are plotted on Monin coordinates in Figure 7. The wavelengths associated with

spectral peaks are approximately 10.8 and 870 feet for model and full-scale, respectively, indicating a scale ratio of 1:80 for this wind direction. It is seen that

the slopes of the spectra below and above the peaks agree reasonably well with those

of the von Karman spectral density function, +1.0 and -2/3, respectively. The scale

ratios obtained in this manner for 6 = 211 and 341 degrees were found to be 1:90 and

1:60, respectively.

Flow properties for four groups of directions are listed in Table I . Unfortunately, turbulence measurements were obtained only at Z = 21.5 feet, these being limited

to the longitudinal component. With the exception B = 211 degrees, the turbulence

intensities measured in the tunnel are approximately fifty percent of the correspondding full-scale values. Run No. 24 for g = 211 degrees was obtained with the spires

and upwind models removed from the tiinnel, the boundary layer thickness for this

case being approximately sixteen inches. Also included in Table I are the maximum

speeds observed in the full-scale data over a time interval of 750 seconds, the

average frequency of occurrence of peak values (commonly referred to as the_ zero crossing rate) , and the me_an velocity at Z = 10 meters. Full-scale values of U-j^q are based upon the ratio ^j_q/^^ which was measured in the wind tunnel.

The use of spires and roughness elements in tunnels with short working sections

results in mean velocity profiles that are quite acceptable when compared with either

the log law or power law representations. However, it is obvious that the turbulence

characteristics are not always in satisfactory agreement with full-scale values.

While the shapes of the spectral density functions agree fairly well with the

full-scale functions, both the scale and intensity of turbulence are low. This suggests that the surface roughness elements must be modeled at a distorted scale and

that a minimum roughness must be maintained in the tunnel when m.odeling relatively

smooth prototype surface conditions

.

11-25
Pressure Measurements

It was anticipated, and later confirmed by the wind tunnel tests, that corrections would have to be applied to the full-scale dyneunic and static reference pressures due to the close proximity of the anemometer and pressure probe to the test building. The procedure was to determine a static pressure correction coefficient

C

in the tunnel and to then apply this correction as an offset to all full-scale pressure records. The dynamic pressure correction coefficient

dp - o 1/2 PUj^^

was applied in a similar manner. Both corrections were direction-dependent and values

of and C^^ ranged from -0.10 to -0.21 and from 1.02 to 1.39, respectively.

The results of pressure measurements are summarized in Tables II - V and are com- pared graphically in Figures 8 and 9. The mean pressure coefficient, C , and the

fluctuating pressure coefficient, C^^, correspond to the usual definitions and are referenced to the free-stream dynamic pressure at Z = 10 meters. The peak factor, g, is defined as the number of standard deviations included in the maximum peak departure from the mean, i.e.,

I Pmax or min ~ P

I

P rms

The zero crossing rates (Table V) are as previously defined.

The mean pressure coefficients, C , plotted in Figure 8 indicate considerable

disagreement between model and full-scale measurements. Equipment used to measure

and record the field data was somewhat prone to zero drift with the extreme range of

temperatures experienced at the test site (80°F to -20°F) . The field test equipment

has since been modified to provide transducer zero readings and recorder calibrations

prior to each data run. Corrections for this drift could not be accurately determined

and any resulting errors are directly reflected by Figure 8. With the exception of Run No. 310, the addition of a fixed value for a given full-scale run would improve

the agreement between the two sets of coefficients.

Changes in equipment sensitivity due to temperature variations were found to be

small and, therefore, greater confidence can be put in the measurements of pressure

fluctuations. Fluctuating pressure coefficients, Cp^, plotted in Figure 9 indicate

a fairly good 'correlation between model and full-scale, the former averaging approximately one-half of the latter. This discrepancy is believed to be due primarily to the low turbulence intensities observed in the model studies, although scale effects

cannot be completely ruled out. There is no obvious explanation for the good agreement between Run Nos . 207 and 21, and the poor agreement between Run Nos . 311 and 22 It may be that the pressure fluctuations are quite sensitive to wind direction and

that directions were not properly matched in the model studies. It is expected that

a new series of tunnel tests in which the turbulence characteristics are more accu- rately simulated will bring the results into much better agreement.

Spectral densities for tap positions P3 and PIO (Run Nos. 311 and 22) are com- pared in Figures 10 and 11, respectively. The shapes of the spectra compare quite

favorably and suggest a slope of -4/3 over the higher frequency range. As indicated

previously, the wind data were normalized on the assumption that the scale ratio was

11-26
1:50. Although the peaks are not well defined, they suggest a scale ratio in line

with that obtained by comparing the velocity spectra.

Spectral densities for the full-scale pressure fluctuations above the spectral

peak usually exceeded those obtained from the wind tunnel model (see Figure 11)

This is quite likely due to the fact that the analog filters used in the full-scale

studies had a cutoff frequency of 10 Hz, resulting in aliasing errors in the spectral

analysis. The wind tunnel pressure signals, on the other hand, were subjected to pronoimced attenuation above 100 Hz which would appear as a 2 Hz cutoff in the analysis.

Thus the aliasing errors can be expected to be considerably smaller for the wind tun- nel data.

Another indication of similarity between model and full-scale pressure fluctuations is the coherence function. The coherence function, or more properly (coher~

ence)-'-/^, is the normalized modulus of the cross-spectrum and is a measure of the cor- relation between fluctuations at two points over the frequency range for a given separation distance. This function for P1-P5 and Run Nos. 310 and 23 is plotted in Figure 12.

The peak factors, which are listed in Table IV, are fairly consistent and suggest

an average value of 4.6. It has been shown, both theoretically and experimentally

(5, 6), that the peak factor increases with length of record and a value of g = 4.3

has been suggested for calculating design loads on cladding elements for those situations where resonant response is insignificant. The peak factors in Table IV are based upon single records of 750 seconds and represent a worst-case excursion from

the mean, i.e. negative excursions for pressure taps 1-9 and positive excursions for

taps 10-12. It should be pointed out that the peak factors will vary from record to record for a given wind speed and direction and that several records would be required

to establish their mean and variance.

Fluctuating pressure coefficients and peak factors determined from the area-integration of pressure records are presented in Table VI. The procedure was to con- struct a new pressure time series by multiplying the samples of each record to be

summed by a weighting factor. The weighting factor was proportional to the area at- tributed to each pressure tap which implies that all pressure fluctuations over that area are perfectly correlated. The resulting record was then analyzed in the usual

manner. As expected, the fluctuating pressure coefficients show a decrease in increasing area, but there is no indication of a similar reduction in the peak factor.

Although the transducer separations used in this study do not permit a detailed assess- ment of the correlation of peak pressures acting over extended roof areas, some indication of the area reduction of peak pressures can be derived from Table VI. Comparing the product of the fluctuating pressure coefficient and peak factor for transducer

combinations 1, 2, 3 for Run No. 332 with the product of the averaged fluctuating

pressure coefficients and peak factors for the individual transducers (Tables III and

IV) , the corresponding reductions in peak pressure fluctuations are approximately 27 and 40 percent, respectively. A similar comparison for the combination 1, 2, 3, 6, 7, 10, 12 (Run No. 310) indicates a reduction of 42 percent. In averaging the pressure

records, it was assumed that the pressure fluctuations under the eaves overhang were perfectly correlated and equal to the fluctuations measured by transducer No. 10. An

increase of 42 percent was obtained from the combination 1, 2, 3, 10 for Run No. 310. It should be noted that the records for transducer Nos. 10 and 12 were inverted so that positive fluctuations acted in the same sense as negative fluctuations on the

roof

7.0 COMPARISON WITH RECOMMENDED DESIGN LOADS

In addition to providing a check on wind tunnel test results, the full-scale data re- ported herein allow some direct comparisons with current recommended design wind loads.

American National Standard Building Code Requirements for Minimum Design Loads in Buildings

and Other Structures, A58. 1-1972, states a procedure for calculating wind loads on roofs

11-27
and considers both tributary (parts and portions) and total roof areas (7) . The basic wind

speeds used in this document are the fastest-mile speeds for a 50-year mean recurrence interval and flat, open country (Exposure C) at a height of 30 feet above ground.

Because they are defined differently, the pressure coefficients measured in full-scale

and those specified in the A58.1 Standard cannot be compared directly. However, the pressure can be compared for a given wind speed and category of exposure. Assuming flat, open

country and a basic wind speed of 90 mph, the corresponding effective velocity pressures

for heights less than thirty feet are as follows:

= 20 psf (A58.1 - Table 5 - Ordinary Buildings and Structures)

= 31 psf (A58.1 - Table 6 - Parts and Portions)

For buildings with a ratio of wall height to least width of less than 2.4, the A58.1

Standard specifies a general external pressure coefficient of -0.7 for roofs. For gabled

roofs with the wind direction perpendicular to the ridge and the height-width ratio and

roof slope being considered here, a pressure coefficient of -1.0 is specified for the windv/ard slope. The Standard also allows for local peak pressures which are assumed to act at ninty-degree corners and on strips running along the ridge and eaves. The width of these

strips is taken as ten percent of the least width of the building normal to the ridge.

The specified pressure coefficients are -2.4 for ridges and eaves and -3.9 for ninety-degree

corners

.

No specific provision is made for pressures on the underside of eaves, but this can be

taken as the pressure acting on the windward wall for which a coefficient of 0.8 is specified. Internal pressures are based on the fastest-mile speed at thirty feet above ground

for the appropriate terrain category and an internal pressure coefficient, , which is

related to the distribution of wall openings and the ratio of open to solid wall area. The

design pressures for the assumed wind speed and exposure are as follows:

(a) Leeward Slope, Total Area (-0. 7) (20) = - 14 psf

(b) Windward Slope, Total Area (-1. 0) (20) = - 20 psf

(c) Leeward Slope, Tributary Area (-0. 7) (31) = - 21.7 psf

(d) Windward Slope, Tributary Area (-1. 0) (31) = - 31 psf

(e) Ridges and Eaves (-2. 4) (31) = - 74 psf

(f) Ninety-Degree Corners (-3. 9) (31) = -121 psf

(g) Underside of Eaves ( 0. 8) (31) 25 psf

(h) Internal Pressure ( 0. 3) (21) 6.3 psf

At Z = 10 meters, the corresponding mean speed averaged over 750 seconds (the record length

used in this study) is approximately 75 mph (see Ref . 8) . For standard atmospheric condi- r n tions, the effective velocity pressure is 14 psf. Using the full-scale pressure coefficientji

^ listed in Tables II, III, and IV, the mean and peak pressures were calculated for ten trans- ducer locations and four wind directions. The results are presented in Figure 13 along with

the design pressures specified by the A58.1 Standard (Case f excepted.)

It is seen from Figure 13 that the negative design pressures for the total leeward and ,: windward roof areas (Lines a and b) exceed the observed mean pressures for all transducer !, ^ locations. However, the observed peak pressures dominate and exceed the design pressures • ^ for tributary areas (Lines c and d) at several locations. This may or may not be signifi- ; pj cant, depending on the degree to which pressure fluctuations are correlated over the roof j, area. The design pressures for ridges and eaves (Line e) substantially exceed the observed |: ^ peak pressures and the design pressure for ninety-degree corners is approximately 2.5 times

11-28
the maximum observed negative pressure. The positive design pressure for the underside of

the eaves (Line g) is exceeded by the observed peak pressure for two wind directions and

the internal design pressiare (Line h) is less than the measured peak internal pressure for

all four wind directions.

As discussed previously, the fluctuating pressure coefficients determined from the

area-integration of pressure records decrease with increasing area while the peak factors

remain about the same. Although reductions from twenty to forty percent in the peak pressures (peak departures from the mean) are indicated for roof areas of up to fifty square

feet, the data presented in Figure 13 suggest that the provisions of the A58.1 Standard

for tributary roof areas are marginal.

The provisions for ridges and eaves and for ninety-degree corners, on the other hand,

appear to be overly conservative. The maximum negative pressure, based on the measured

coefficients, was -50 psf for transducer No. 1 at 256° while the design pressures are -74

psf for ridges and eaves and -121 psf for ninety-degree comers. It is recognized that

extreme negative pressures are associated with vortices generated along the edges of the

roof and that these vortices are extremely sensitive to wind direction and roof geometry.

However, the wind tunnel studies described previously did not reveal any critical wind directions not covered by the full-scale data. If the effective velocity pressure of 20 psf

(q^) is used in place of 31 psf (q^) , the resulting design pressures are -48 psf for ridges • and eaves and -78 psf for ninety-degree corners.

As with the negative pressures acting over the roof area, the peak pressures under the eaves (Transducer No. 10) and in the garage airea (Transducer No. 12) far exceed the corresponding mean pressures averaged over the record length of 750 seconds. Again referring

to Table VI and Figure 13, the average maximum pressure (peak plus mean) acting upward on

the eaves is 3 - 3 + 35 = 35 psf for Run No. 310 and combination 1, 2, 3, 10. The corre- sponding design uplift pressure is 74 + 25 = 99 psf. For combination 1, 2, 3, 6, 7, 10,

the average maximum pressure acting over a roof area of 258 square feet is 3 + 13 = 16 psf

as compared with 2117 + 6.3 = 27 psf (tributary area plus internal pressure) as specified

by the A58.1 Standard.

The maximum internal pressure (based on measured coefficients) in the garage area was approximately five times the corresponding design pressure. The ratio of effective open

area to solid area is difficult to determine since all doors and windows were closed during

the recording intervals reported herein. One door and one window of approximately nineteen

and fifteen square feet, respectively, are located on the southeast wall, one window of

fifteen square feet on the southwest wall, and an overhead garage door of 55 square feet

is situated on the northwest wall. Because of the extremely cold winters in Montana, great

care is usually taken to provide doors and windows with adequate seals or weatherstripping.

It is probable, therefore, that infiltration rates for this garage area would compare with

those for living quarters in regions having a mild climate.

8,0 A PROCEDURE FOR THE CALCULATION OF DESIGN PRESSURES

The 1970 edition of the National Building Code of Canada (NBC) (9) provides for risk

of occurrence, terrain roughness, height above ground and building geometry in calculating

design wind pressures

P

= g Cg Cp (1)

In this expression, q is a reference mean velocity pressure for a given mean recurrence

interval, Cg is an exposure factor which varies with surface roughness and height above

ground, C is a gust effect factor to provide for the dynamic response characteristics of

the structure and surface pressure fluctuations caused by turbulence and localized flow

phenomena, and C is the conventional mean pressure coefficient determined from wind tunnel

tests. For the design of cladding, it is assumed that dynamic response can be neglected

and a factor of 2.5 is used for C_.

11-29
The form of Equation (1) is particularly convenient in that it allows a complex process to be treated as a combination of independent elements and provides for the separate

treatment of mean and fluctuating components of pressure. The peak design pressure at any

point on a roof area can be expressed as

P CD = p - gp^^ rms Or in terms of dimensionless pressure coefficients based upon a suitable reference pressure

such as the free-stream dynamic pressure at Z = 10 meters

,

Cg = 1 + g (2)

While Equation (2) is a practical means of expressing C , full-scale and wind tunnel mea- surements (Figures 14 and 15) suggest an empirical relationship between C^^ and C . If the

envelope of full-scale C^^ values is expressed as

Cpf = 0.3(1 + iCpl )

and it is assumed that g = 5.0, Equation (2) becomes

(1 + I C

'

, = 1 + 1.5

gC = 1 + 1.5 —; —^— (3) Cpl

As indicated earlier, the peak factor g does not appear to change with surface area. There

fore, it may be possible to determine a gust factor, Cg, for extended roof areas simply by

reducing by means of a factor, Rp, roughly analogous to the size reduction factor used

in calculating dynamic response. The gust effect factor would then be defined by

= 1 + 1.5 (4) P

where is the corresponding mean pressure coefficient for the extended roof area. P

9.0 CONCLUSIONS

The use of spires at the entrance of a wind-tunnel working section substantially in- creases the growth rate of rough wall boundary layers, thereby placing the study of building aerodynamics within the capability of many conventional tunnels. However, the use of

spires and scaled upwind roughness elements does not alone ensure the establishment of

flows with proper turbulence characteristics. A minimum degree of surface roughness is

required to establish suitable scales and intensities of txirbulence. With some modification of the roughness elements, it is believed that close simulation of the lowest 100 feet

of the atmospheric boundary layer can be achieved at a scale ratio of 1:50.

The agreement between model and full-scale spectra for both velocity and pressure

fluctuations is encouraging. Measurements of coherence suggest that the spatial extent of

siirface pressure fluctuations can be modeled to an acceptable degree.

It is believed that the consistently low values of C ^ determined from the model re- sult primarily from improper simulation of the turbulence. However, scale effects cannot

be ruled out at this time. Peak factors were found to agree quite closely with previous

measurements, the overall average for model end full-scale results being 4.6. Based on

the preliminary results reported herein, it appears that a gust effect factor can be ex- pressed in terms of a peak factor, mean pressure coefficients and a size reduction factor.

Wind pressures based on measured coefficients and an assumed wind speed suggest that

certain provisions of the current A58.1 Standard deserve additional study. For the buildir

11-30
investigated, design pressures for tributary areas and interiors appear to be marginal

while those for localized areas such as ridges, eaves, and corners appear to be excessive.

10.0 ACKNOWLEDGMENTS

Acknowledgment is made to the Tri-Services Committee, Department of Defense, for financial support of the field studies reported herein. The writer wishes to express his

thanks to Lt. Col. C.Y. Holland, Jr., USAF, for his assistance at the field test site and

to Col. R.A. Nelson and family whose home was instrumented during the course of the study.

Special thanks are extended to Dr. H.I. Laursen who developed much of the computer software

used in the data analysis.

11,0 BIBLIOGRAPHY

1. Melbourne, W.H. , "Comparison of Pressure Measurements Made on a Large Isolated Building

in Full and Model Scale." Proceedings, Third International Conference on Wind Effects

on Buildings and Structures, Tokyo, 1971, pp. 253-262.

2. Standen, N.M. , "A Spire Array for Generating Thick Tubulent Shear Layers for Natural

Wind Simulation in Wind Tunnels." National Aeronautical Establishment Report LTR-LA-94

Ottawa, May 1972.

3. Marshall, R.D. and Hsi, G. , "Techniques for Measuring Wind Loads on Full-Scale Buildings." Proceedings, Research Seminar on Wind Loads on Structures, University of Hawaii

October, 1970, pp. 133-148.

4. Helliwell, N.C. , "Wind Over London." Proceedings, Third International Conference on Wind Effects on Buildings and Structures, Tokyo, 1971, pp. 23-32.

5. Dalgliesh, W.A. , "Statistical Treatment of Peak Gusts on Cladding." Proceedings of the

Structural Division, ASCE, September, 1971, pp. 2173-2187.

6. Davenport, A.G. , "Note on the Distribution of the Largest Value of a Random Function

with Application to Gust Loading." Paper No. 6739, Proceedings, Institution of Civil

Engineers, Vol. 28, June, 1964, pp. 187-196.

7. American National Standard Building Code Requirements for Minimum Design Loads in Build

ings and Other Structures, A58. 1-1972. Issued by American National Standards Institute

New York, New York.

8. Deacon, E.L., "Wind Gust Speed; Averaging Time Relationships." Australian Met. Mag.

No. 51, 1965.

9. National Building Code of Canada 1970. Issued by the Assoc. Comm. on the NBC, National

Research Council, Ottawa, Canada.

11-31
NOTATION

The following symbols are used in this paper:

Cjjp = Dynamic pressure correction coefficient

Cg = Exposure factor

C = Gust effect factor

g

C = Mean pressure coefficient

P

= Rms pressure coefficient

C = Static pressure correction coefficient

sp •

,

I = Intensity of turbulence, in percent

U-j^Q = Mean velocity at standard ten-meter height, in feet per second

Uj^ = Reference mean velocity at 21.5 feet

Z = Height above ground, in feet

Z = Gradient height

G

^

Z^ = Surface roughness height, in inches

g = Peak factor

p = Mean pressure, in pounds per square foot

p = Rms pressure

rms

p_, = Reference dynamic pressure at 21.5 feet R

p-j^Q = Reference dynamic pressure at ten meters -—

q = Reference dynamic pressure for open country

R = Size reduction factor

P

3 = wind direction measured clockwise from north, in degrees

p = Mass density of air, in slugs per cubic foot

II- 32
LIST OF TABLES

TABLE I Flow Properties

TABLE II Mean Pressure Coefficients (0^)

TABLE III Fluctuating Pressure Coefficients (Cp^)

TABLE IV Peak Factors (g)

TABLE V Zero Crossing Rates

TABLE VI Area-Averaged Coefficients

FIGURE 1

FIGURE 2

FIGURE 3

FIGURE 4

FIGURE 5

FIGURE 6

FIGURE 7

FIGURE 8

FIGURE 9

FIGURE 10

FIGURE 11

FIGURE 12

FIGURE 13

FIGURE 14

FIGURE 15

LIST OF FIGURES

Aerial View of the Test Site

Test Site Layout

Plan and Elevation of the Test Building

Wind Tunnel Layout

Mean Velocity Profiles - Power Law Representation

Mean Velocity Profiles - Log Law Representation

Velocity Spectra (Runs Nos. 311 and 22)

Mean Pressure Coefficients

Fluctuating Pressure Coefficients

Pressure Spectra (P3, Run Nos. 311 and 22)

Pressure Spectra (PlO, Run Nos. 311 and 22)

Coherence (Pl-PS', Run Nos. 310 and 23)

Comparison of Measured and Specified Design Pressures

C vs. C^-r - Full Scale

p pf

C„ vs. C„<r - Wind Tunnel

P pf
TABLE I FLOW PROPERTIES

Run*

Numbe r

Direction

(degrees

)

^"r (m/s) (m/s)

1 (percent)

Max i mum

Peak

(m/s)

Zero

Crossing Rate

(Hz)

^o

(m/s)

207 186 I 1 .37 3.38 30 19.93 .267 12.07

21 186 1 1 .89 2 .5'» 21 - 12 .62

3 I 0 2 1 7 7.77 2.93 38 18. Al 1 "58 23 211 13.17 '.OS 31 15.51

2k 21 1 18.71 I . 10 6 18.78

311 256 8.38 2.62 31 16.25 .169 8. 1*7 22 256 U.69 2.59 18 15.06

322 309 9.91 2.35 15.6i» .166 10.36

1 1 3'»1 13.78 1 .'»3 1 0 U. U

101 3^7 S.Zk 2.50 27 1 A.8I .2 11 9. its

*Run numbers with 3 digits denote full scale data.

TABLE I 1

MEAN PRESSURE COEFFICIENTS

C P Run Numbe r

0 i rec t i on (degrees

)

Pressure Tap Number

1 2 3 k 5 6 7 8 9 1 0 1 1 12

207 186 -.38 -.'.3 -.59 -AS -.(>k -.kk -.29 -.32 -.56 -.08 -.20

21 186 -.'•9 -.59 -.76 -.68 -.62 - .22 -.2k -.06 -.\k .kS -.21

310 217 .26 .23 -.18 -.18 -.38 -.05 -.05 -.k] .22 .2k

23 21 1 -.09 -.11 -.\k -.13 -.18 -.05 -.03 -.Ok -.Ok ,08 .07

2k 21 1 -1 .00 - .60 -.53 -.55 -.63 -.22 -.12 -.16 -.02 .67 -.19

311 256 -.12 .12 -.36 -.'•5 -.61 -.8k -.21 -.70 .^16 .35

22 256 .,0 -.11 -.36 -.55 -.59 -.k] .10 -.18 .30 .37

322 309 -.52 -.05 -.22 -.20 -.2k -.78 -.k7 -.k7 -.09 .15

1 ] 3k] -.32 '.]k -.15 -.23 -.12 -.50 •.\k -.37 -.10 -.12 .30

II-3U
TABLE I I I FLUCTUATING PRESSURE COEFFICIENTS

Run

M1^ 1U1 m1 11 hIv' pC r1 Direction

(dcQrees)

Pressure Tap Number

1 2 3 k 5 6 7 8 9 1 0 1 1 1 2

20 7 186 . Ik . 28 ? 2 . 2 1 . 1 9 . 1 8 . 2 1 2 7 1 9

2 1 1 86 .25 .28 .31 .33 .38 .08 .09 . 1 0 .08 .26 .08

310 217 .38 .33 .28 .28 .32 .32 .25 . 22 .29 .29

23 21 1 . 1 k .15 . 1 5 . 1 5 . 1 7 .09 .07 .07 .Ok . 1 k . 1 1 Ik 21 1 .11 . 10 .09 .09 . 1 0 .09 .05 .07 .03 .16 .07

21 1 256 .70 .29 .32 .ko .'3 .62 .31 .30 .kk .'*3 22 256 . 1 2 . 19 .18 .25 . 20 .18 .09 .Ik .05 . 1 3 . 16

322 309 .30 . 1 5 . I 5 . 1 7 .19 .3'* . 2 1 . 1 8 .15 .18

1 1 l>k\ .13 .05 .05 .07 .07 .19 . 1 0 . 1 1 .05 .Ok . 1 2 10 1 3'»7 .31 . 10 .Ik . 1 2 .09 .38 .25 .20 . 10 .25 .'9

TABLE IV PEAK FACTORS

g

Run Numbe r

207

2 1

310

23Ik

31 1 22

322

1 1 101

Direction

(degrees)

186

1 86

217

21 1 21 1

256

256

309

3A1

Ikl

Pressure Tap Number

5.9

5.2

k.S

k.O

2.9

'.9

5.7

k.l

3.6

6.

6.8

7.0

'.9

6.5

6.2

5.5

5.

3.8

2.7

k.k

k.S

6.6

5.7

k.3

5.0

3.9

5.0

3.6

2.3

5. 1 6.0

7.1

5.8

5.7

5.0

7.0

3.7

k.l

2.8

k.S

k. 1

k.k

5.1

'.3

3.9

6.3

7.3

6.6

2.6

3.7

3.8

S.k

5.3

l.k

k.l

k.S

k.l

5.0

A.

2.6

3.9

3.8

k.l

3.7

k.l

6.5

3.7

6.

3.1

6.9

'.3

k.S

3.5

l.k

1.

1.7

6.3

2.9

1 .5 6.8

3.9

k.l

l.k

3.5

1 0

'.3

5.8

6.6

5.9

5.0

3.5

3. 1 k.l

l.k

1 1

k.S

1.

3.1

3.9

3.6

12

3.7

'».6

'.7

3.1

2.0

11-35
TABLE V

ZERO CROSSING RATES

(C ross i ngs/Sec .

)

Run Number

Direction

(deg rees

)

Pressure Tap Number

1 2 3 k 5 6 7 8 9 1 0 1 1 1 2

207 1 86 .37 .3'» .3^ .35 .36 .36 .37 .39 .31 .37 - .25

2 1 1 86 .39 .^3 .31 .35 .33 .30 .36 .34 .15 .21 .26

310 217 .23 .26 • 32 .31 .32 .29 .27 .31 .31 . 1 1 23 21 1 .29 .33 .29 .32 .35 .29 .31 .34 .16 .2 1 .16

2't 21 1 A7 .32 .30 .33 .28 .48 .45 .52 .17 .33 .31

311 255 .20 .28 .26 .2 1 ,21 .24 .34 . 20 .26 .17

22 256 .30 .33 .33 .36 .28 .24 .29 .20 . 1 2 .19 . 1 8

322 309 .^47 AS .36 .37 .32 .41 .30 .31 .21

1 1 3k] .^49 • 32 .35 .35 .25 .27 .23 . 1 1 .22 .25

101 3A7 .46 .i48 .3k .k2 .33 .38 .47 .44 .26 .22

TABLE VI AREA-AVERAGED COEFFICIENTS

Run NumberDirection

(deg rees

)

Tap

Comb I n a 1 1 on

g Area

310 217 1 .2.3 .28 4.6 ( 76) (4.27) - 3 25

II II

1 ,2,3,10 .40 6.2 ( 76) (4.27) - 3 25

II II

1 ,2,3.6,7,10.12 .2 1 4.3 (5 61) (4.27) - 23 97

322 309 1 .2,3 .12 4.3 ( 76) (4.27) = 3 25

II II 1,6 .27 5.3 { .82) (5.61) = 4 65

II II A, 5, 10 .12 8.8 ( .76) (3.05) = 2 32

11-36
11-37
11-38
11-39
2 ' • g

O LlI

in

o>-

UJz

z

n

H-a

z

II -UO
TI-Ul
0.8

0.7 -

0.6

0.5

Uz/Ug 0.4

0.3

T

1 1—I—I I I I

I 1 1 1—I—r-r-r

/3 B 341 degrees

Zq 0.2 mm

/9 • 256 degrees

Zo ' 5.8 mm

1

I I I I I ! \ I I

I I I I I I

Z (METERS)

Fig. 6 Mean Velocity Profiles - Log Law Representation

II -12
TI-h3
O Run Nos. 207, 21

310, 23

A 311, 22

322, II A A

FULL-SCALE

/

/

/ / ' /

/

- 0.2 /

/

/

/ / ° ^ \

^ ^ —^

^ 57 -0.8 -0.6 -j5.4 q -0.2 U

'• / . A /

- /

m / 0 ^ /a

/ 0

/ °

//

/

/

/

/

0.2

A

--0.2

-0.4^

UJz

z

1- --0.6 Q

Z

--0.8

Fig, 8 Mean Pressure Coefficients

ii-Ui
\ \ \ 1 7

o Run Nos. 207, 21 /

310,23 / A 311,22 /

322.11 /

• 101,11 //

/

/

/

/

/ ~

/

/

/

/

/

/

0 /

/a/

l_ I I L

0.2 0.4 0.6 0.8 1.0

FULL-SCALE

Fig. 9 Fluctuating Pressure Coefficients
n-h6
It

11-17
1.0 1

T r • FULL-SCALE

WIND TUNNEL

0.8

liJ 0.6 O

Z

UJ

01

UJ

J 04

r

0.2

0.0 Ju^ I L i

0.0 0.2 0.4 0.6

n(H2)

Fig. 12 Coherence (P1-P6, Run Nos. 310 & 23)

0.8 1.0

II-U8
IT-h9
\

\

\\

A\ + \ % \

\

d d \ d —1

\—^

\

1

1

1

1

0.2

0.4

0.6

0.8

-

0.2

-

0.4

-

0.6

i<^o dA3O0 ssawd oNiivniomj

Q. o -

u:

ii. UJO

o

o - V) - O — — £J Qj

CVi fO K) to 2

6

°- z z

c <

3 UJ (c s - o <

° \

\

?

^ ^

\

-d o 6

\\

-s ° ° <\

O \

<

00 \

~d \

\

\

<

>

Tl-50
\\

\

\\\

\

\\

\ —A\\ o

d \ + \ <

d \ <M

1 \\

\

o d \d

1 \ \

1

1

1

1

0.2

0.4

06

0.8

-0.2 -0.4 -0.6

a. o

LL ~

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11-51

NONLINEAR CALCULATIONS OF GROUND RESPONSE IN EARTHQUAKES

by

William B. Joyner

Albert T.F. Chen

and

Patrick C. Doherty

U.S. Geological Survey

345 Middlefield Road

Menlo Park, California 94025

The response of soil to strong earthquake motion involves a high degree of

nonlinearity . Because of the difficulties in solving the nonlinear problem,

most calculations of ground response are currently made by a method--variously

characterized as "equivalent linear", "quasi-linear", or "strain-compatible"

that assumes the true solution can be approximated by the response of a linear

model whose properties are chosen to accord with the average strain that occurs

in the model during excitation. The average strain level is determined by iterative calculation. To solve the nonlinear problem directly, we have developed

algorithms by which the hysteretic behavior of an individual soil element can

be efficiently modeled in a coirputer. The algorithms enable us to model any rea- sonable set of hysteresis loops of the Masing type that laboratory experiments

may dictate. We are experimenting with various numerical techniques for integrating the basic nonlinear differential equations, including the method of characteristics as described by Streeter, Wylie, and Richart. A comparison was made

between the equivalent linear and the nonlinear solution (using the method of

characteristics) for a 200-meter section of firm alluvium excited at its base by

the N21E component of the Taft accelerogram multiplied by four. This excitation

produced peak strains of several tenths of a percent. The nonlinear solution

showed substantially higher spectral levels of response at five percent damping

for periods between 0.1 and 0.6 seconds.

Key Words: Elastic medium; engineering seismology; ground layer; numerical solution;

shear wave.

III-l
INTRODUCTION

The problem we undertake to solve in this paper is basically a very simple one. We

postulate a system of horizontal soil layers bounded above by the free surface and below

by a semi- infinite elastic medium representing the bedrock. We further postulate a vertically incident shear wave in the underlying medium, and we ask the question, "How will the

overlying layers respond and in particular, what will be the motion of a point on the free

surface?"

This is a classical problem in engineering seismology. There is some disagreement

concerning the range of applicability of the solution, but no one would deny the importance

of solving this relatively simple problem correctly.

The problem was solved for the case of linear visco-elastic layers by Kanai (1952) some years ago. When we are dealing with input motion sufficiently intense to cause severe

damage to structures, however, we cannot assume simple linear behavior. To do so would

imply stresses many times greater than the strength of typical materials as measured in

the laboratory.

To circumvent this difficulty, the method in common use currently is what we shall

refer to as the "equivalent linear method" (Idriss and Seed, 1968; Schnabel , . Seed, and

Lysmer, 1972) . It is based on the assumption that the response can be approximated by the

response of a linear model whose properties are chosen in accord with the average strain

that occurs in the model during excitation. The average strain level is determined by

iterative calculation.

On an intuitive basis the equivalent linear assumption is reasonable with respect to the frequencies that are dominant in the strain history. It is less clear that the assumption is adequate with respect to higher frequencies, which may be important to the safety

of small or rigid structures (Dobry, Whitman, and Roesset, 1971) .

We undertook to obtain a solution without making any assumptions that bypass the underlying nonlinear character of the physical processes involved.

METHOD

The basic requirement for a solution to the problem is a constitutive relation—in

simple terms we need a rule that will tell each soil element how to find its way around

the stress-strain plane. For this purpose we have adopted a model (Figure 1) that was originally proposed by Iwan (1967) . It is composed of simple linear springs and Coulomb

friction elements arranged as shown. The friction elements remain locked until the stress

on them exceeds the yield stress S^. Then, they yield, and the stress across them during

yielding is equal to the yield stress. Generally, the yield stress of the first element

S^ is set to zero. By choosing the spring constants Kj appropriately, we can model a very

broad range of material behavior as dictated by laboratory experiments. The faithfulness

of the modeling depends upon the number N of elements used. We have found it possible to use large numbers economically. For our typical problem N is 50; we have gone as high as 100 without unreasonable increases in computing time. There is one model of the kind diagrammed in Figure 1 for each soil layer in the system.

The type of hysteresis loops that such a model produces is shown in Figure 2, which

illustrates the behavior of a single soil layer subject to cyclic loading of increasing

amplitude. The loops of Figure 2 are plotted in terms of reduced stress and reduced

strain--scaled in such a way that the maximum stress on the sample is one unit and the low

strain modulus has a value of one. The spring constants of the model were chosen to give

the behavior as a function of strain indicated by the experimental work of Hardin and

Drnevich (1972a, 1972b)

.

To satisfy the boundary conditions at the base of the system of soil layers, we use

a method described by Papastamatiou (1974) . That method allows us to satisfy the boundary

III-2
conditions exactly for a vertically incident wave in the underlying elastic medium, given

the density and shear velocity of the elastic medium. The approach is similar to that

described by Lysmer and Kuhlemeyer (1969)

.

Given a constitutive relation and a boundary condition at the base of the system, we need only integrate the equations of motion step by step in time and space to obtain the

motions at the surface. We have experimented with a number of methods of integration (Chen

and Joyner, 1974) . This experimentation is continuing and we are not yet prepared to discuss the relative merit of different methods, but we are satisfied that the method of

characteristics, as described by Streeter, Wylie, and Richart (1973), gives accurate results and the example that follov/s was done with that methoa.

EXAI4PLE

To demonstrate the method we chose the soil profile illustrated in Figure 3, representing a 200-meter section of firm alluvi\im. We evaluated the depth variation of maximum

stress the low-strain modulus using the methods of Hardin and Drnevich (1972a, 1972b) , with

minor modifications. We assumed a density of 2.05 gm/cm^ and converted shear modulus to shear velocity, which is plotted in Figure 3. A past consolidation vertical stress of 2.94

bars was assumed. As a result, the material was overconsolidated above a depth of 29 meters,

causing kinks at that depth in the two curves on Figure 3. The underlying medium was assximed to have a shear velocity of 2.0 km/sec and a density of 2.6 gm/cm-^.

For imput we chose the N21E component of the Taft accelerogram, recorded during the

1952 Kern County, California, earthquake. We multiplied the amplitude by a factor of four

so that the input motion, if incident on a free surface, would have a peak acceleration of

0.7 g and a peak velocity of 67 cm/sec. Figure 4 shows the input acceleration time history and compares the surface acceleration computed by the nonlinear method and the equivalent linear method. There are definite points of similarity, but it is clear that the

equivalent linear approximation does not adequately represent the short-period components

of motion present in the nonlinear solution.

Comparing the nonlinear solution with the input shows the effect of the postulated

soil profile on ground motion. The peak acceleration is sharply reduced. The longer

period components are amplified, however, and the overall effect may be a more damaging

motion as will be illustrated subsequently.

Figure 5 shows the corresponding velocity time histories for the same example. Comparing the nonlinear and equivalent linear solutions, we see much better agreement, indicating that the equivalent linear approximation is adequate with respect to the longer

period components that are dominant in the velocity time history. Comparing the nonlinear

solution with the input shows clearly the amplifying effects of the soil profile for longperiod motions. .

To illustrate the consequences of these results for structures we have computed response spectra. Figure 6 shows rhe acceleration response at five percent damping for the

three motions between zero and 1.2 seconds period. The line represents the input, hexagons

the nonlinear solution, and crosses the equivalent linear solution. It is clear that the

equivalent linear method significantly underestimates the intensity of motion for periods

between 0.1 and 0.6 seconds. This period range corresponds to structures between one and

about six stories. The importance is obvious.

Comparing the input response with the nonlinear response in Figure 6, one might be

tempted to conclude that for short-period structures, the motions on alluvium would be

less damaging than on bedrock. Considerable caution is indicated here. For one thing,

different soil materials, dense sands, for example, might given more intense motions. Posible lengthening of structural periods due to deformation beyond the linear range reeds to

be considered as well as the effects of duration not accounted for in the response spectrum.

Consideration should also be given to rhe effects of ground deformation and ground failure.

III-3
Figure 7 compares the relative velocity response spectra at five percent damping for

the range from zero to six seconds period. The results show that the equivalent linear

approximation is adequate for the longer periods and that the soil site gives large amplification for longer periods.

During the running of the nonlinear solution, we monitored the peak strain for each

depth interval. The maximum was 6 x 10"-^ for the interval from 32 to 35 meters.

The comparative costs of the two methods is difficult to evaluate in the general case because it is possible to run the equivalent linear method using fewer layers, depending

on the detail one wishes to represent in the soil profile. For the example presented here

however, using the same number of layers for both methods, the nonlinear solution required

less than half as much computer time as the equivalent linear. So, we believe that, in

general, it will be competitive, at least.

REFERENCES

1. Chen, A.T.F., and W.B. Joyner, 1974, Multi-linear Analysis for Ground Motion Studies

of Layered Systems. Manuscript in process.

2. Dobry, Ricardo, R.B. Whitman, and J.M. Roesset, 1971, Soil Properties and the OneDimensional Theory of Earthquake Amplification: School of Engineering, Massachusetts

Institute of Technology, Cambridge, Massachusetts, Research Report R71-18, Soils Publication No. 275.

3. Hardin, B.O., and V.P. Drnevich, 1972a, Shear Modulus and Damping in Soils: Measurement and Parameter Effects: Proc. Am. Soc. Civil Eng. , Journ. Soil Mech. and Found.

Div. , 98, 603-624.

4. Hardin, B.C., and V.P. Drnevich, 1972b, Shear Modulus and Damping in Soils: Design

Equations and Curves: Proc. Am. Soc. Civil Eng. , Journ. Soil Mech. and Foiind. Div. , 98, 667-692.

5. Idriss, I.M., and H.B. Seed, 1968, Seismic Response of Horizontal Soil Layers: Proc

.

Am. Soc. Civil Eng. , Journ. Soil Mech. and Found. Div. , 94 , 1103-1031.

6. Iwan, W.D., 1967, On a Class of Models for the Yielding Behavior of Continuous and

Composite Systems: Jour, of Applied Mechanics , 34 , 612-617.

7. Kanai, K. , 1952, Relation Between the Nature of Surface Layer and the Amplitudes of

Earthquake Motions: Bull. Earthq. Res. Inst. , Tokyo Univ. , 30 , 31-37.

8. Lysmer, John, and R.L. Kuhlemeyer, 1969, Finite Dynamic Model for Infinite Media:

- Proc. Am. Soc. Civil Eng. , Journ. of the Eng. Mechanics Div. , 95 , 859-877.

9. Papas tamatiou, D.J., 1974, A Method for the Evaluation of Strong Ground Motion. Submitted for publication in Geotechnigue .

10. Schnabel, P., H. Bolton Seed, and J. Lysmer, 1972, Modification of Seismograph Records

for Effects of Local Soil Conditions: Bull. Seism. Soc. Am. , 62 , 1649-1664.

11. Streeter, V.L., E.B. Wylie, and F.E. Richart, 1973, Soil Motion Computations by Characterics Method, Meeting Preprint, 1952, Am. Soc. Civil Eng., National Structural Eng.

Meeting, April 9-13, San Francisco, California.

III-4
LIST OF FIGURES

Model used for constitutive relationship. Model consists of simple elastic springs

with spring constants K-j. and Coulomb friction elements with yield stress S-j-.

Hysteresis loops for the model shown in Figure 1. Stress and strain are scaled so that the maximum stress and the low-strain modulus are unity.

Dynamic properties for soil profile used in sample problem.

Acceleration time histories for sample problem.

Velocity time histories for sample problem. •. .- •

Acceleration response at five percent damping for sample problem. The line represents

the input, hexagons the nonlinear solution, and crosses the equivalent linear solution.

Relative velocity response at five percent damping for sample problem. The line represents the input, hexagons the nonlinear solution, and crosses the equivalent linear

solution.

III-5
111-6
II I -7
III-8
III-9
TII-10
iii-n
III-12
OBSERVATION AND ANALYSIS OF GROUND RESPONSE IN EARTHQUAKES

by

Satoshi Hayashi

Head, Structures Division

Port and Harbor Research Institute

Ministry of Transport

Hajime Tsuchida

Chief, Earthquake Resistant Structures Laboratory

Structures Division

Port and Harbor Research Institute

Ministry of Transport

In the field of port and harbor engineering, the ground response due to earthquakes is generally considered. However, there are many problems at present in idealization of the surface layer, input ground motions, and nonlinear

behavior of the soil. In order to provide some design data in these areas, research has been conducted using six dovmhole seismometer arrays established

in port areas in Japan. Typical observed acceleration time histories have been

obtained and are shown and compared with those calculated by the multiple re- flection theory. In order to investigate ground response relative to structures

of large length, such as a tunnel, a two-dimensional seismometer array has been

established. Examples of the considerations of the ground response required

in practice are seen in microzonation in the earthquake proof design of a subaqueous tunnel.

Key Words: Earthquake; field data; ground response; seismic waves; seismometer;

soil-structure interaction.

III-13
INTRODUCTION

In recent earthquake engineering practice, a concept has been introduced which con- siders the earthquake response of the ground; here the ground means surface layers over- laying the base rock. The amplification of ground motions through the surface layers was

first observed and the theoretical analysis, so-called the multiple reflection theoiry, developed by Dr. Kanai (1). However,. the introduction of this concept into design practice

has taken several years. In the field of port and harbor engineering, the first step re- lated to the amplification problem was to revise the procedure in determining the design

seismic coefficient. The old design seismic coefficient was a function of only local seismicity. In tlie new procedure the design seismic coefficient is a function of subsoil con- dition underlaying the structure, as well as local seismicity and importance of the structure (2) . Even in this revision, the subsoil condition was taken as only one of the factors, and only because of past experiences obtained during destructive earthquake events.

It was noted that during these earthquakes, damage to the structures did depend on the

subsoil conditions, however, a clear understanding of the amplification of the ground

motions was not considered.

The soil structure interaction analysis, which has become one of the main topics in

earthquake engineering research during the last decade, has provided insight to researchers

and practicing engineers relative to the significance of ground motion response. Progress

through the application of computer techniques has assisted in the introduction and the

consideration of the earthquake response of surface layers to be used in practice. The

International Conference on Microzonation, held in Seattle in 1972, clearly showed that

many researchers and engineers intend to consider the earthquake response of surface layers

in design practice. In the previous joint meetings of the U.S. -Japan Panel on Wind and

Seismic Effects, UJNR, Dr. Joyner and Dr. Borchcherdt have presented reports on this topic

(3) and (4)

.

In the present paper the authors will report on the practical problems involved in

earthquake response calculations of the ground, observation of the ground response in port

areas in Japan, analysis of observed records, and design examples in which the ground re- sponse in earthquakes is taken into consideration.

PRACTICAL PROBLEMS ON GROUND RESPONSE CONSIDERATION

From the viewpoint of the seismology, the ground response in earthquakes can be ex- plained as follows: the seismic waves which enter the surface layer, travel from the

epicenter to the base rock site in consideration and propagate upward into the surface

layer, and during the propagation, the amplification and phase shift of the motion occurs. I In principle, the process is quite clear, however, in practice, there are some problems.

[ The first problem that confronts the engineers is the determination of the base rock. In } many sites the rock formation is so deep that it is not possible to idealize the surface | layer overlaying the base rock into a model to calculate the earthquake response.

The second problem involves the input base rock motions; in other words, the earthquake motions in the base rock. There are not many recorded records available of the base

rock motion that can be used by engineers

.

The third problem is the idealization of the surface layer. At present three well ,! known major idealizations are used; they are 1) multiple reflection model, 2) finite ele- ' ment model, and 3) lumped mass model. There is, however, little information on the application of these idealizations for use by engineers. Also, all of the idealizations are faced with the problem of how to incorporate the non- linear behavior of the soil.

The solution to these problems have not been found. However, to apply the techniques,

the engineer is forced to make reasonable assumptions within the present state of the art.

The following will present some of the answers relative to the problems concerning the earthquake proof design of port structures. When the base rock exists at considerable ( depth, and it is not practical to idealize the whole layer into a model to calculate the

j III-14 i
earthquake response, a layer extending widely beneath the structure is considered as an as- sumed base rock instead of the real rock formation. The assumed base rock, however, must

satisfy the following requirements: the shear wave velocity of the assumed base rock is

significantly larger than that of a layer immediately overlaying it; the standard penetration value of practical base rock is equal to or greater than fifty.

The idealization of the surface layer depends on several conditions to calculate the

earthquake response. The multiple reflection model is frequently used to estimate the

ground art5)lif ication and to calculate the base rock motions, as obtained from the records

at the ground surface. In soil structure interaction problems, the finite element model

and the lumped mass model are applied more conveniently than the multiple reflection model.

Non-linear behavior of soil is a very difficult problem to solve, therefore, the evaluation

of the ground response in earthquakes, in connection with port structures, are treated

assuming that the soil behaves linearly. However, it will be noted that in the calculation

of the lateral resistance of piles, the non-linear behavior of soil has been taken into

consideration (5) . Also a quasi-non-linear approach is introduced into the amplification

calculation, that is, the soil is considered as a linear material but the shear modulus is

assumed as modified, depending on the calculated maximum strain of the layer. This approach

essentially requires repetitive calculations.

The solution to the basic problem described herein depends largely on the findings

from the analysis of strong-motion accelerograms and base rock accelerations, as calculated

from the accelerograms obtained from the multiple reflection theory (5)

OBSERVATION OF GROUND RESPONSE IN EARTHQUAKES

Six downhole seismometer arrays have been established by the Port and Harbor Research

Institute. The first array is located in the test field of the research institute in

Yokosuka (7); the other four arrays are located in Tokyo, Funabashi, Nagoya, and Osaka

Ports to provide information for coastal structure design, and the last array is in Kawasaki Port, to provide data for the earthquake proof design of a subaqueous tunnel. This

tunnel will be constructed similarly to the procedure that was used for the BART tunnel

crossing at San Francisco Bay. The arrays in the Tokyo Port and in the test field in

Yokosuka have the lowest seismometers on rock formations.

The specification for the seismometers of the arrays, except those in Yokosuka, are

as follows

Type Moving Coil Type

Component Two/Three Components

Natural Frequency of Transducer 5 Hz

Natural Frequency of Galvanometer 110 Hz

'

Overall Sensitivity About 1 mm/gal in max.

The signal corresponding to acceleration is recorded on an electro-magnetic oscillograph.

A magnetic tape data recorder may be more sophisticated than the electro-magnetic oscillograph, however, the mechanism of the data recorder is very delicate and the oscillograph

that was chosen has a higher reliability in recording operation. A disadvantage of the

oscillograph is the time required to digitize the oscillograms. To overcome this disadvantage, an oscillogram digitizer has been directly connected to a small scale hydrid computer, which has been introduced into the authors' laboratory. With the digitizer, a trace of about seventy centimeters in length is digitized within a few minutes.

III-15
RECORDS AND ANALYSIS

As an example of the data that has been obtained, the records from the seismometer

array in Kawasaki Port and the results of preliminary analysis will be described herein.

As mentioned previously, the array has been established to obtain data for use in the

earthquake proof design of the 1380 meter long Kawasaki Port subaqueous tunnel.

The array consists of four downhole seismometers, located at depths of 1, 15, 40, and

63 meters below the ground surface. The lowest seismometer rests in a sand layer, which

has a standard penetration value exceeding fifty. This layer is considered as the assumed

base rock in the earthquake response calculation. The subaqueous tunnel to be constructed

will exist entirely above this layer. The boring log and the shear wave velocity are shown in Figure 1.

During the last year, eight earthquakes were recorded. Three of the records show

relatively large amplitudes of the vertical distribution of maximum accelerations, as shown in Figure 2. It was observed that the maximum accelerations decrease with increase

in the depth.

In Figure 3, reproductions of the acceleration time histories of the EW component re- corded on April 6, 1973 are shown. The acceleration time histories at the boundaries between soil layers were calculated from the ground surface acceleration time history based

on the multiple reflection theory. These results are shown in Figure 4, together with the

recorded time history at 63 meters below the ground surface.

The results from the observation and the analysis, such as a frequency transfer function of the site, are reflected in the response calculation and the dynamic model test of

the tunnel.

TWO DIMENSIONAL SEISMOMETER ARRAY

The seismometer arrays described previously are all vertically uni-directional . For I many types of structures, such as a tunnel and a pipeline, relative ground movement along

| the structure is very important for their earthquake proof design. The Port and Harbor [ Research Institute has, therefore, established a two-dimensional seismometer array. This

array is located at Tokyo International Airport along a straight line of 2500 meters, parallel with the C airport runway. Seven seismometers, each of which contains two horizontal

transducers, were installed at an equal spacing of 500 meters along the 2500 meters. At

one end of the line and at 500 meters inside from the other end the downhole seismometers were installed at about 50- and 67 meters below the ground surface. Therefore, the seismometer system can record a two-dimensional ground response in the vertical plane. The

arrangement of the seismometers is illustrated in Figure 5.

Setting of the seismometers was completed in March of this year 1974.

EXAMPLES OF GROUND RESPONSE CONSIDERATION IN PRACTICE

Practical examples, in which the ground response to earthquakes should be considered,

will now be explained. As described previously in this paper, in the earthquake proof design of the subaqueous tunnel in Kawasaki Port, the ground response was given in detail.

The actual tvinnel was modeled, together with the surface layer, using a lumped mass system

in which the surface layer and the tunnel was connected by springs which represent the

stiffness of the soil. The input earthquake accelerations that were used were obtained

from the base rock accelerations calculated from the accelerograms at the ground surfaces

given by the multiple reflection theory (8)

.

Another example was the microzonation in the coastal area along Tokyo Bay. At present, effort is being directed toward the examination of the seismicity of the coastal and

port structures in the area, and the microzonation of the amplification of the earthquake

III-16
motion. In the zoning work, the amplification of the earthquake motions through the surface layer were calculated by the multiple reflection theory. Also, the quasi-non-linear

approach was used (9) (10)

.

REFERENCES

1. K. Kanai: Relation Between the Nature of the Surface Layer and the Amplitude of

Earthquake Motions, Bull. Earthquake Research Institute, Vol. 30, 1952, pp. 31-37.

2. The Japan Port and Harbor Association: Design Manual of Harbor Structures in Japan,

1967.

3. William B. Joyner and Richard E. Warrick: The Influence of Geology on Earthquake

Ground Motion—Preliminary Results from a Downhole Seismometer Array on the San Francisco Bay Shore, Proceedings, Third Joint Meeting of U.S. -Japan Panel on Wind and

Seismic Effects, UJNR, 1971.

4. Roger D. Borcherdt: A Preliminary Strategy for Zoning the San Francisco Bay Region

on the Basis of Relative Ground Motion Prediction, Proceedings, Fifth Joint Meeting,

U.S. -Japan Panel on Wind and Seismic Effects, UJNR, 1973.

5. Satoshi Hayashi, Hajime Tsuchida, and Takamasa Inatomi: Inelastic Response of Steel

Pile Structure and Its Seismic Stability, Proceedings, Fourth Joint Meeting, U.S. -Japan

Panel on Wind and Seismic Effects, UJNR, 1972.

6. Hajime Tsuchida and Tatsuo Uwabe : Characteristics of Base-Rock Motions Calculated

from Strong-Motion Accelerograms at Ground Surface, Report, Port and Harbor Research

Institute, Vol. 11, No. 4, 1972, pp. 191-270.

7. Hideo Arai and Moriyoshi Iwabuchi : Studies on the Vertical Distributions of the Earthquake Motions in the Superficial Soil Layers (1st Report) - Observation and Analysis

of the Earthquake Motions in a Clay Layer, Report, Port and Harbor Research Institute,

Vol. 12, No. 2, 1973, pp. 63-83. .

8. Takuji Nakano and Kenji Mori: Earthquake Resistant Calculation and Dynamic Model Test

on Trench Type Tunnel, Technical Note, Port and Harbor Research Institute, No. 172,

1973, 87 p.

9. Satoshi Hayashi, Hajime Tsuchida and Setuo Noda : Estimation of Aseismicity of Port

Structures, Proceedings, Fifth Joint Meeting, U.S. -Japan Panel on Wind and Seismic

Effects, UJNR, 1973.

10. Setuo Noda and Tatsuo Uwabe: Estimation of Maximum Ground Acceleration Around the

Tokyo Bay Area During Great Earthquake, Technical Note, Port and Harbor Research In- stitute, No. 177, 1974.

III-17
Q.

0)

iO)

4-> q;E

o

E

(/)

OJ TT

P

0

0

—r

' L

-Aa—

7/:

Soi 1 s

Sand fill

Sand

S i 1 ty c 1 ay

Silty sand

Sandy clay

Sand

Sandy gravel

Shear wave

vel oci ty

(m/sec]

120

200

150

220

220

310

630

Fig. 1 Soil condition at seismometer array site (Kawasaki)

III-18
Maximum acceleration

10 (ga ls) 0 in

I "I

Earthquake

April 6, 1973

T 1

Feb. 22, 1974

Fig. 2 Vertical distribution of maximum

accelerations, EW component

III-19
I

Fig. 3 Acceleration time histories

Earthquake of April 6, 1973, EW component

III-20
III-21
III-22
III-23
RESEARCH STUDY ON LIQUEFACTION

by

W.F. Marcuson, III

U.S. Army Engineer Waterways Experiment Station

Vicksburg, Mississippi

A research program is being undertaken by the U.S. Army Corps of Engineers

to evaluate the liquefaction phenomena relative to earthquake response of earthfilled dams. Present and future studies are summarized.

Key Words: Earthfill dams; earthquakes; ground shaking; liquefaction; soil

density; stability.

III-24
INTRODUCTION

A major cause of damage of catastrophic failure associated with earthquakes is often

attributed to liquefaction and/or slumping of cohesionless soils as a result of dynamic

excitation. The importance of liquefaction was recently highlighted by the near failure

of the Lower San Fernando Dam during the 9 February 1971, San Fernando Earthquake. Many

other liquefaction failures have occurred and are documented in the literature (1, 2, 3,

4).

As a result of the near failure of the Lower San Fernando Dam, the Corps of Engineers

(CE) initiated a research program on liquefaction phenomenon. A number of these investigations are being conducted at the U.S. Army Corps of Engineer Waterways Experiment Station (WES), Vicksberg, Mississippi. The objectives of these studies are (a) to evaluate

the seismic stability of the CE hydraulic-fill dams, which fortunately are located in

areas of low to moderate seismicity, (b) to evaluate the dynamic response of proposed Corps

structures which are to be located in areas that are known to be seismically active and

(c) to improve the Corps ability to determine in situ density and relative density or to develop an index which can be reliably related to liquefaction potential. This paper will

discuss these activities.

CYCLIC TRIAXIAL COMPRESSION TESTS

Soil behavior under conditions of strong ground shaking is determined by cyclic tri-

. axial compression tests. In such tests, the deviator stress is uniformly increased and

1 1 decreased while maintaining a constant chamber pressure. The test equipment utilized by

WES is shown in Figure 1. The pneumatic control unit consists of regulators and solenoid

valves that are actuated by a cam-operated microswitch. The values provide alternating

air pulses to a double acting air cylinder (loading) piston such that a cyclic load is

transmitted from the air cylinder through the piston to the sample. Obviously, careful

calibration was required, considering friction in the system and uplift pressures generated by the chamber pressure, to regulate the pulsating air pressures so that the desired loads are imposed on the specimens.

Electrical pressure transducers, a deformation transformer, and a load cell are used

to measure pore pressure and chamber pressure, axial deformation, and axial load, respectively. Because of the rapid change in sample behavior at liquefaction, a high speed recorder is required to provide a continuous record of events during the tests. Figure 2 is a photograph of this equipment. The triaxial cell is on the right with a 2.8 inch diameter soil sample inside. The large rectangular unit in the center is the pneumatic conI trol unit which houses the air pressure regulators and solenoid values. The unit on the

' left is a high speed recorder which provides a continuous record of the test data.

The majority of tests performed to date have been conducted on remolded samples of

' cohesionless material consolidated to isotropic or anisotropic conditions as required. A

back pressure is applied to the specimen to ensure saturation. Saturation for these testing programs is defined in terms of Skempton's B-parameter. The B-parameter is equal to

the ratio of the change in pore pressure to an induced change in chamber pressure. The

value is checked in these tests by closing the drainage line and increasing the chamber

pressure a desired amount, and observing the increase in pore water pressure. Typical B-

( values are 0.98 or greater for these testing programs. The specimens are subjected to a cyclic vertical load at a frequency of 2 Hz. Liquefaction consists of large deformation

and a partial to total loss of strength. At the start of cyclic loading little deformation

occurs, but as the cyclic loading progresses, the pore pressure increases thus reducing

the effective stress. The CE presently defines initial liquefaction as the stage when the

pore pressure first becomes equal to the chamber pressure giving an effective stress of

zero. When this happens with a loose specimen, large deformations soon occur. This is

shown in Figure 3, which is a plot of stress ratio a, /2o versus cycles of loading for a dc a cohesionless material at a density of 98 pcf or a relative density of approximately 60

percent, a^^ is equal to the cyclic deviator stress and is equal to the ambient confining pressure. The curves are for initial liquefaction, ten and twenty percent peak-topeak strain. Approximately three cycles of loading are required for the specimens to III-25
undergo twenty percent strain after initial liquefaction has been reached. The response

of a dense specimen is substantially different. Figure 4 is a plot of stress ratio versus ! number of cycles of loading for the same cohesionless material at a density of 109 pcf or

a relative density of approximately ninety percent. At this density approximately fifty

cycles of stress were required beyond initial liquefaction to reach ten percent peak-to- ; peak strain and twenty percent peak-to-peak strain was never reached.

WES has used the results of these tests along with dynamic finite element computer

calculations to assess the stability of both operational structures and proposed structures (5, 6, 7, 8)

.

The CE is also currently funding cooperative research on laboratory testing in this area under the direction of Professor A. Casagrande at Harvard University. Dense sand

specimens have been dynamically excited in a gyratory shear device. Specimens have been

frozen immediately after liquefaction. Samples were then cut into many increments and

soil properties of each increment are determined. Evaluation of these data is ongoing "i today and should be published within the year. Tentative results indicate that the behavior of dense sand specimens is governed by the migration of water within the specimen. ,

EVALUATION OF IN SITU DENSITY

The response of a cohesionless soil when subjected to dynamic loads is a function of

its density or relative density. The accurate determination of the in situ density or relative density of a cohesionless material below the water table is a difficult and time

consuming procedure. Work was conducted at WES in the 1950 's which showed that undisturbed sand samples could be taken below the water table for density determinations (9)

,

During this investigation, a procedure developed by WES (10) using a fixed piston sampler

and drilling mud was employed to obtain undisturbed sand samples. Briefly this procedure

is as follows: (a) As soon as the sample is removed from the drillhole, a packer is placed

in the bottom of the sample tube and the sample is allowed to drain while remaining in a ,i vertical position, (b) After drainage, a packer is also placed in the top of the tube and

the sample is placed in a rack in a horizontal position and fixed against rotation. A

riibber hammer is then used to tap the specimen 25 times from left to right and 25 times

from right to left. (c) The sample is transported to a laboratory in this condition. ' (d) Once in the laboratory, the samples orientation is never changed (i.e., no rotation

of the sample whatsoever) . It is cut into three-inch increments for density determination."

Presently WES is X-raying the sample and cutting the samples into increments as indicated

in the radiograph for density determination. Figure 5 is a plot of height from bottom of

sample tube versus correction for location of increment in tube for various relative den- ^ sities and summarizes the results found during this earlier investigation. This plot

shows that no correction is necessary near the center of the tube and the extreme density

corrections are in the order of 2 pcf.

WES is presently checking the previously developed method of undisturbed sampling of ' sands to determine its adequacy for determining the in situ density of cohesionless ma-

' terials below the water table. Indirect methods of in situ density determinations are also being evaluated. Penetration and sampling tests are being conducted inside a 4 ft. '• diameter by 6 ft. high stacked ring assembly. Figure 6 is a photograph showing a sand

sample inside the stacked ring device. The rings are approximately one inch thick and

have a tongue and grooved neoprene gasket between each ring. The purpose of this device

is to minimize sidewall friction and thus give a more realistic stress gradient in the

sample. The sand has a median size of approximately 0.2 mm and about five percent passing

the No. 200 sieve. This material is placed inside the stacked rings using a raining technique developed at WES. Samples are placed at known densities in six inch lifts. Figure

7 is a schematic showing the entire stacked ring assembly and the sand Scimple. WES pressure cells are placed at three locations in the specimen. Each location has a series of

six cells, three vertical and three horizontal. The assembly is located in a pit in the

floor of the laboratory building. A rock filter exists below the sand sample and water

is introduced at this elevation to submerge the sample. A water bag is placed on top of

the sand sample and a loading device with three hydraulic rams is used to apply pressure

III-26
to the water bag thus ensuring a uniform stress distribution on the top of the sample.

There is a five inch diameter hole in the center of the loading device and water bag and

three three-inch diameter holes located thirteen inches out from the center on 120 degree

radials. Figure 8 is a photograph of the bag before it is placed on the top of the sand

specimen.

A vertical pressure is applied to the top of the specimen and consolidation is allowed

to occur. Undisturbed fixed piston samples are taken in the center hole. Standard penetration tests (SPT) are conducted in each of the three radial holes. Figure 9 shows the

overburden device assembled and the drill rig in its operational position. Tentative results of this work indicate that an accurate determination of density can be made using

the fixed piston samples and add confidence to our earlier findings. Use of absolute instead of relative densities is desirable for seismic analysis of a specific project, but

relative densities are needed to compare analyses and experiences at different sites.

Attempts have been made to correlate SPT "N" values (number of blows required to drive

a splitspoon the last twelve inches of an eighteen-inch drive) to relative density. The

analysis of these data is inconclusive at this time.

FUTURE INVESTIGATIONS

Most of the liquefaction analysis that have been made in the U.S. have been based on

work done by Professor Seed and his coworkers at the University of California. This work

has placed great emphasis on the liquefaction which occurred during the Niigata earthquake

of 1964 (10) . Professors Casagrade, Harvard University; Peck, University of Illinois;

and Seed have all emphasized the need for direct determination of density at sites where

liquefaction has occurred. It is believed that this is especially important at the Niigata site. Professor Casagrande also believes that the geologic age of the various sand

deposits could be an important parameter which has not been studied to date. The CE feels

that a program of undisturbed sampling at the Niigata site would greatly enhance existing

knowledge on liquefaction phenomenon and recommends that such studies be made if feasible.

WES would be pleased to offer technical assistance in this matter if desired.

WES has recently purchased a sophisticated closed loop loading apparatus. This equipment along with other apparatus that are currently being developed at WES should give the

CE the capability of cyclicly testing up to fifteen-inch diameter specimens. This equipment also has the capability of using a random input (actual earthquake time history) to excite the specimen. The CE plans to utilize this equipment to gain insight into parameters such as the geometry effects, the effects of frequency, the effects of wave shape

and form, and the assumptions currently used in going from the actual field earthquake to equivalent laboratory conditions.

The CE is extremely interested in developing a mathematical model which would predict

liquefaction. Theoretical work is proposed in the near future to develop such a model.

WES is currently funding research at the University of Michigan utilizing the method of

characteristics to see if this approach can be used to predict liquefaction (11) . The objective of this investigation is to see if the method of characteristics can acciarately

predict the response of the Lopez Dam during the 1971 San Fernando Earthquake.

REFERENCES

1. Seed, H.B. , "Landslides During Earthquakes Due to Liquefaction," Journal of Soil Me- chanics and Foundations Division , American Society of Civil Engineers, Vol. 94, No. SMS, September 1968, pp. 1053-1122.

2. Coulter, H.W. and Migliaccio, R. R. , "Effects of Earthquakes of 27 March 1964, at Valdez, Alaska," Geological Survey Professional Paper 542-C, U.S. Department of In- terior, 1966.

III-27
3. Seed, H.B. and Idriss, I.M., "Analysis of Soil Liquefaction, Niigata Earthquake,"

Journal of Soil Mechanics and Foundations Division , American Society of Civil Engineers, Vol. 93, No. SM3, May, 1967, pp. 83-108.

4. Seed, H.B. and Whitman, R.V. , "Design of Earthquake Retaining Structures for Dynamic

Loads , " Proceedings of the Speciality Conference on Lateral Stresses in the Ground

and the Design of Earth Retaining Structures , Soil Mechanics and Foundations Divisioi

American Society of Civil Engineers, 1970, pp. 103-147.

5. Marcuson, W.F., III and Collins, S.A. , "Cyclic Triaxial Compression Tests, Center Hi! Dam, DeKalb County, Tennessee," MS-74-5, March 1974, U.S. Army Engineer Waterways Ex- periment Station, CE, Vicksburg, Mississippi.

6. Marcuson, W.F., III and Gilbert, P.A. , "Earthquake Liquefaction Potential at Patoka

Dam, Indiana," Miscellaneous Paper S-72-42, December 1972, U.S. Army Engineer Waterways Experiment Station, CE, Vicksburg, Mississippi.

7. Marcuson, W.F. , III, "One-Dimensional Wave Propagation Analysis, Patoka Dam, Indiana,

Miscellaneous Paper (in publication) U.S. Army Engineer Waterways Experiment Station,

CE, Vicksburg, Mississippi.

8. Marcuson, W.F. , III, "Liquefaction Analysis of the W.G. Huxtable Pumpting Plant Sitej

Mississippi River," Report (in publication) U.S. Army Engineer Waterways Experiment j Station, CE, Vicksburg, Mississippi.

I 9. Rotary Cone Penetrometer Investigations, Potomology Investigations, Report 18-1, Jun* 1962, U.S. Army Engineer Waterways Experiment Station, CE, Vicksburg, Mississippi.

10. Undisturbed Sand Sampling Below the Water Table, Bulletin 35, June 1950, U.S. Army

Engineer Waterways Experiment Station, CE. Vicksburg, Mississippi.

11. Seed, H.B. and Idriss, I.M., "Simplified Procedure for Evaluating Soil Liquefaction

Potential," Journal of Soil Mechanics and Foundation Division , American Society of

Civil Engineers, Vol. 97, No. SM9, September 1971, pp. 1249-1273.

12. Streeter, V.L. , Wylie, E.G., and Richart, F.E. , Jr., "Soil Motion Computations by

Characteristics Method, " Journal of the Geotechnical Division , American Society of

Civil Engineers, Vol. 100, No. GT3, March 1974, pp. 247-263.

III-28
CONTROL REGULATORS

FOR CHAMBER PRESSURE

BACK PRESSURE

LOADING PISTON

INSTRUMENTATION

FOR MONITORING

LOAD CELL,LVDT,

PRESSURE

TRANSDUCERS

DOUBLE-ACTING

LOADING PISTON'

n

1 note:

n

LVDT IS A LINEAR

VARIABLE DIFFERENTIAL TRANS - FORMER

L VDT

SOIL SPECIMEN

2.8- IN OR 1. 4 -IN. DIAM~

LOADING FRAME

TRIAXIAL CHAMBER

CHAMBER AND

PORE PRESSURE

TRANSDUCERS

Fig. 1. Cyclic triaxial test equipment

III-2S

III-31
oiiva ss3ais

III-32
30

MEASURED DRY DENSITY . LB/CU FT

MEASURED RELATIVE DENSITY , -3-2-1 0 1 2 CORRECTION FOR LOCATION OF

INCREMENT IN TUBE . LB/CU FT -

Fig. 3' Density correction for location in

sample tube, combined plot

III-33
Sand Sample in Stack Ring Assembly

III-34
III-35
rig. 8' Water Balloon Prior to Assembly on Sample

III-36
Fig. 9. Overburden Assembly and Drill Rig in Operational Position

III-37
ESTIMATION OF LIQUEFACTION POTENTIAL BY MEANS OF EXPLOSION TEST

by

K. Yamamura

Director

Kita-Shuto National Construction Office

Ministry of Construction, Japan

Y. Koga . . . , •i:- Research Engineer

Soil Dynamics Section

P. W.R.I.

Ministry of Construction, Japan

A series of field vibration experiments were performed in order to estimate the liquefaction potential of sandy soil during earthquakes. Bore hole

explosions were used as a vibration source. A significant relation was found

between the ground stiffness and the pore-water pressure as caused by the ex- plosion. A proposed method has been developed for estimation of the liquefaction potential. . .

Key Words: Earthquake; explosion test; ground strength; ground vibration;

liquefaction; pore-water pressure.

III-38
PREFACE

Since the Niigata and Alaska Earthquakes in 1964, and the disastrovB effects due to liquefaction of a sandy ground or a sandy embankment, there has been increased interest

and concern in the liquefaction phenomena. Extensive research and study has, therefore,

been conducted in various countries relative to this problem.

Based on the results of these studies, several simple methods have been proposed in

order to estimate the liquefaction potential (1, 2, 3, 4). These methods use the N-value,

as obtained from conducting standard penetration tests, the relative density and the grainsize distribution, etc. Other methods (5, 6) use a combination of the results obtained

from dynamic shear tests and the analytical results of the ground vibration to estimate

liquefaction potential. In addition to these methods, a technique which can estimate the

liquefaction possibility of the soil in the in-situ position, that is at the field site,

is most desirable. This is because generally the in-situ soil has the following characteristics ;

(1) There may exist a certain cementation strength among the grains in an un- disturbed state, even in a sandy soil, depending upon the age and the development of the deposit (for instance, old alluvial deposits or reclaimed

land )

.

(2) The soil may not be uniform, but may have laminations which can include silt

or gravel layers, or a surface soil may be present.

One of the most logical methods that can be employed to investigate the ground behavior under vibration is to apply an artificial vibration source at the actual test site.

The inducement of an artificial vibration can be achieved by a bore hole explosion and by

driving piles.

Explosion Method (7, 8, 9) : Florin has proposed a criterion for the estimation of

liquefaction potential, by examining the magnitude of the average settlement of the

ground surface within the radius of 5 m from the explosion point, by blasting 5 kg

of ammonite explosive at the depth of 8 m to 10 m and the ratio of settlement by

three successive explosions at one place. By this method, if the average magnitude

of the settlement of the ground surface within 5 m radius is below 8 m to 10 cm, there

is no need to provide any measures against liquefaction of the soil.

Pile Driving Method (10) : This method estimates liquefaction by reading the acceleration meters and the pore pressure gauges installed in the ground to measure the

vibration of the ground caused by the driving of a steel tube pile or a sand compaction pile.

The following will describe an outline of the liquefaction estimation experiment, us- ing the explosion technique, undertaken by the Public Works Research Institute.

PLAN OF EXPERIMENT

(1) Site of Experiment

The site where this experiment was performed was at the high water channel of the

Agano River, in Niigata Prefecture. The embankment of this river had suffered cracks,

slidings, and settlements at numerous places during the Niigata Earthquake, and was known

to have a oil which could easily liquefy. Also, in selecting the site, it was remote from

the inhabited area, thus eliminating any human complaints due to the tests.

(2) Soil Condition

The soil condition can be classified into three major zones, as shown in Figure 1 (a) to (c) . A- zone has a layer of 0.5 - 1.0 m thick humus on the surface and a sand layer

III-39
underneath of approximately uniform fine and coarse sand partially intermixed with finegrain gravels. The ground water level is approximately at the ground surface. The ground

stiffness N-values are 3-8 at a depth of 5.5 to 6.0 m. Further down the N-values suddenly

increase to a value of above twenty. The B-zone has a cross layer of fine and coarse sand

mingled with gravel. The ground water level is located approximately at the depth of 0.7

to 1.3 m from the ground surface. The N- value to the depth of about 0.5 to 7.5 m is three

to eight. Further down, the N-value shows a sudden increase to a value of thirty. The

C-zone is covered with a layer 1.0 m of surface soil with an underneath cross layer of

fine-coarse sand. The N-value, down to the depth of about 3.0 m, is three to six. At further depths, the value is ten to twenty. The ground water level is in the vicinity of

2.0 to 2.3 m from the ground surface. The grain size distribution of the sand in each

zone is shown in Figure 2. Such a distribution curve is indicative of the type of sand

which is prone to cause liquefaction.

(3) Test Method

Dynamite which was charged into a bore hole of 4-8 m deep was used for the explosion.

No sand was filled in the explosive charged hole. Strain gauge type accelerometers , pore _ pressure gauges of the strain gauge type, and differential transformers were respectively

installed. These units were placed in both the bore hole, surrounding ground, and in the

shallow ground surface. The changes or fluctuations caused by the explosion were then re- corded by a data- recorder (magnetic tape type) and an eletromagnetic oscillograph. All

of the bore holes, which were made for monitoring, were filled with sand. The magnitude

of the settlement of the wooden piles, installed in the ground surface, were measured before and after each explosion by use of a level. For some of the experiments, Swedish

weight-sounding tests were performed before and after an explosion, in the vicinity of the

explosion hole, to investigate the strength variation of the ground.

(4) Test Condition

Table 1 shows the test conditions of the experiments that were performed.

RESULTS OF EXPERIMENT , ; ,

• , .

'

(1) Measured Records i

Examples of the recorded acceleration data and the pore water pressure data are shown

in Figure 3 (a) and (b) . The initial shock part of the acceleration data, caused by the

explosions, is from the body wave. The relatively long frequency appearing thereafter i is thought to be either the reflected wave from the ground surface and the lower stratum, or by the surface wave. Even with the pore water pressure after it had shown an initial

short cycle vibratory characteristic (Ud) , it changed into a somewhat longer cycle vibration leaving a partial residual substance. This, however, rapidly dissipated generally

in two to five minutes, and then remained constant. This is probably because the ground

structure has been disturbed by the seismic motion as caused by the explosion, and there

has occurred a phenomenon of a partial liquefaction causing an excess pore water pressure

in the ground.

(2) Ground Acceleration

| Figure 4 shows a diagram of the maximum value of the surface ground acceleration

against the distance from the explosion hole to illustrate the ground acceleration characteristic. As shown by this diagram, the vertical acceleration is larger than the horizontal acceleration at the ground surface, with a greater scattering with respect to the

vertical acceleration.

(3) Pore Water Pressure '.

Figure 5(a) represents the arrangement of the pore water pressure gauges and accelerometers installed for the test to measure the pore water pressure. Figure 5 (b)-(e)

III-40
illustrate the distribution of the shock pore water pressure Ud against the 1.0 kg dynamite

explosion and the residual pore water pressure Ur. In these examples, there has occurred

in the vicinity of these explosion holes an excess pore water pressure of 10% to 70% of

the overburden pressure, and it is believed that a small scale liquefaction has taken place

in the ground. To further indicate whether a liquefaction has actually taken place, such

items as a sand blow, water spout, and ground settlement should be observed. However,

neither a sand blow or a water spout did occur although ground settlement was noted, as will be detailed later. Figure 6 illustrates the relationship of Ud and Ur to the ground

stiffness from all tests including those presented in Figure 5 (b)-(e). Ud and Ur have

been non-dimensionalized by dividing them by the effective overburden pressure ^

^'

According to this Figure 6, in C-zone where the ground is stiff, Ur is smaller than

Ud, and by contrast in the B-zone where the ground is soft, the reverse trend occurs. As will be shown, there is a relationship between ground stiffness and liquefaction potential.

(4) Ground Surface Settlement

Examples of the measured ground surface settlement caused by the explosion are shown

in Figures 7(a) and 7(b). The test data presented corresponds to the same tests presented

in Figures 5(b) and (d) . Assume now that the soil stratum above the explosive depth charge

of GL-6m has a relation with the ground settlement rate. This can be illustrated by the

relationship between the mean N-value of the soil stratum and the settlement rate at one meter from the explosion point, as shown in Figure 8(a). Further, consider the pore water

pressure at a measuring point 6 m from the explosion hole and at the depth of 6 m, as shown in Figure 8(b). This figure shows the relationship between Ur, at this point, and

the previously mentioned settlement rate. It appears from these plots that as the ground

settlement rate increases, the ground becomes softer and Ur increases and vice-versa.

Also by measuring the ground settlement rate in this manner, a greater potential of occurrence of Ur in the ground may be estimated.

(5) Variation of Ground Strength

In order to investigate the ground strength variation by explosion, a set of Swedish

weight-sounding tests were conducted surrounding the explosion hole before and after the

e}^losion, an example of which is shown in Figure 9(b) -(e). This experiment was carried

out before and after the ejcperiment No. 1-5. The position of these tests are shown in

Figure 9(a). The following trends are noted from Figure 9. In the case where the ground

strength is estimated in terms of an average number of half-turns per m (Nsw) , the preexplosion ground strength tends to increase with ground depth. However, after the explosion, its distribution range is apt to expand. This is particularly true in the vicinity

of the explosion point, where its strength drastically decreases. At the central point,

at a distance of 5 m from all the explosion holes, the strength throughout the entire

stratum has s\ibstantially increased. This may be due to the ground compaction, as a result of the explosion vibration.

SIMPLE ANALYSIS OF TEST RESULTS

At present, the liquefaction phenomenon of sandy soil at the time of an earthquake

is said to be caused by the repeated shear stress under the force of the earthquake and

the dilatancy mechanism of sand rather than the ground acceleration or its inertia force.

Therefore, even the seismic motion produced by explosion differs from that caused by an earthquake. A simple analysis was, therefore, tried, which would consider partial liquefaction to take place as caused by the dynamic shear stress.

The ground vibration by an in-bore hole explosion cannot be analyzed very simply because of the existance of bore holes surrounding the ground surface, as well as the non- uniformity of the ground strata and beddings. However, a simple approximation may be

made by considering the ground vibration caused by an explosion as that due to a spherical

wave (P wave) emitted from one point in an infinite elastic body.

III-41
Using the symbols given in Figure 10, the equation of motion in the radial direction

r may be expressed (11) by the Equation (1)

.

8ar _ 2 9r

Ofc) = P S^u

(1) 3t'

where: o^. and represent the normal stresses in the radial and tangential directions,

respectively. Taking the compression stress as a positive, the following expressions can be given by employing the modulus of elasticity E of the ground containing water, and

Poisson's ratio v:

(1+v ) (l-2v)

(1+v) (l-2v)

{ (1 - v) + 2v H }

^ 9r

(2)

From Equations (2) , the dynamic shear stress j ^ and mean principal stress are determined

as

:

2 + 2o,

2(1 + v) 3^ ^ ___E___(3U + 3(1 - 2v) 3^ ^

(3)j:

In order to satisfy Equation (1) , the following equations are assumed:

. : u = = 1 f (r Ct)

3r ^ Substituting Equations (4) into Equations (3) gives:

(4)

The ratio of Equations (5) are:

a m

2(1 + v) rE 3(1 - 2v)

= 3(1 - 2v)

2(1 + v)

.(!_ f - i_ f ' +1 f")

1 f"

(3_

(5)

r2 f"

f_ _ 3_r il^ 1) (6)

It is now desirable to evaluate the stress ratio at the time the P wave arrives at each point. Eit^jloying the following expression from the continuity conditions of displacement and velocity gives:

f = 0 f = 0 (7)

Then from Equation (5), the stress ratio at the time of P wave's arrival, namely, at the time when the initial shock wave motion is propagated, is;

Li = 3(1 - 2v)

a

(6'1 m 2 (1 + V ) Assuming the ground is in a saturated state, the shock pore pressure Ud, produced by

the arrival of this P wave, is considered to create an undrained condition because of the

rapid occurrence of the phenomenon. Because of this, shock pore pressure is equal to the mean principal stress a , or; m

III-42
Ud = am

(8)

Also, Poisson's ratio v in an undrained condition of the saturated soil may be given

by the following expression (12)

,

V = y (1 - nGCw) (9)

where n = Porosity; G = Modulus of rigidity of the ground; Cw = Coefficient of compressibility of water and v = 1/2.

Substituting Equations (8) and (9) into (6') gives;

= nGCw • Ud (10) d

This equation states that the shock pore water pressure Ud, created immediately after

explosion, is proportional to the dynamic shear stress which affects the soil.

Figvire 11 summarizes the results of a dynamic triaxial compression test under a sat- urated undrained condition performed on sand similar to the sand at the test site. This

diagram illustrates the effect of residual pore water pressure in the test specimens of

different densities subjected to different dynamic shear stresses. These results indicate

that at smaller densities, the residual water pressure increases. Thus, the greater the

initial dynamic pore water pressure at the time of explosion, the softer the ground becomes

and the greater will be the residual pore water pressure. Figure 6 shows this same trend.

Figure 6 represents an approximate comparison of the classified ground for each test zone.

This shows that as the density varies in the depth direction, the liquefaction potential

may also differ. For the density at various stages of depth, a classification by use of

the actually measured or estimated N-values was considered, since no sufficient sampling

of the undisturbed test specimens was performed. For this reason, on Figure 6 there are

listed the estimated N-values at each measuring point, and a tentative classification made

by N < 10. According to this diagram, it may be said the differences between Ud and Ur

are more evident than by the mere division according to zone. Therefore, the relation between Ud and Ur is closely related to the ground density and from the relationship of both

values, it may be possible to estimate the ground density or possibly the liquefaction

potential.

CONCLUSION

(1) A reasonable method for estimations of ground liquefaction potential by means of bore

hole explosion has been obtained.

(2) As an estimated index for liquefaction potential, a relationship between the shock

water pressure, residual water pressure caused by the explosion, and the ground set- tlement rate has been developed.

(3) Despite the above developments, the following problems remain to be solved:

(a) Determine the applicability of the technique for other soil conditions (creation

process, soil property, density, etc.).

(b) Determine the applicability for deeper soil.

(c) Improve the accuracy of measuring ground acceleration and pore water pressure.

(d) Improve the accuracy of the measurement of soil densities.

(e) Examine further the characteristics of residual pore pressure as obtained by a laboratory dynamic shear test.

(f) Review of the analytical methods of ground vibration by explosion.

III-43
REFERENCES

1. Tsuchida, H. : Forecast and Measures of Sandy Ground (in Japanese) , Technical Note of

Port and Harbour Research Institute, 1970.

2. Ohsaki, Y. : Effects of Sand Compaction on Liquefaction during the Tokachioki Earthquake, Soils and Foundations, Vol. 10, No. 2, 1970.

3. Gibbs, H. J. : Earth Structure and Foundations Subjected to Seismic Loading, Preprints

of 2nd Joint Meeting of U.S. -Japan Panel on Wind and Seismic Effects, 1970.

4. Japan Road Association: Aseismic Design Manual of Road Bridge (in Japanese), 1972.

5. Seed, H.B. , Idriss, I.M. : Simplified Procedure for Evaluating Soil Liquefaction Po- tential, Proc. of ASCE, Vol. 97, No. SM9, 1971.

6. Iwasaki, T. : Study on Aseismic Design of Foundation of Structure and Ground (in Japa-

.

nese) , Technical Note of Public Works Research Institute, No. 764, 1972.

7. Florin, V.A., Ivanov, P.L. : Liquefaction of Saturated Sandy Soils, Proc. of 5th Int. Conf. on S.M. and F.E., Vol. 2, 1961.

8. Kummeneje, 0., Eide , 0.: Investigation of Loose Sand Deposits by Blasting, Proc. 5th

Int. Conf. on S.M. and F.E. , Vol. 2, 1961.

9. Prakash, S., Gupta, M.K. : Blast Tests at Tenughat Dam Site, Jour, of Southeast Asian

Soc. Soil Eng., Vol. 1, 1970.

10. Fukuoka, M. , Ishihara, K. : Measurement of Pore Pressure During Field Vibration (in

Japanese), Preprints of 27th Annual Meeting, Japan Soc. Civ. Eng., 1972,

11. Timoshenko, S.P., Goodier, J.N. : Theory of Elasticity, 3rd ed. , McGraw-Hill.

12. Ishihara, K. : Approximate Forms of Wave Equations for Water-Saturated Porous Materials

and Related Dynamic Modulus, Soils and Foundations, Vol. 10, No. 4, 1970.

I

III-44 • \
III-45
!0C s

0.5

G

J

\

1

(a)

CMP

s

kg/c o

(b) Fig. 3 Typical records (a) Acceleration (b) Pore water pressure

o - Horizontal Acc.

• - Vertical Acc.

0 Distance

5/0 (m)

Fig. 4 Accelerations as functions of distances from explosion point

amount of charge = 1 kg

depth = 6 m

111-16
Distance (m)

G.L.

56

<i) Q

6

Hor. A-1

Vert. A-4

O

P-1 o

P-2

O

P-3

A2

A5

O

P-4

O

P-5

10 —

A-3

A6

O

P-6

• Explosion point

Horizontal

A Accelerometer Vertical

o Pore pressure gage

(a)

1 4

a 6

0) Q

ud/CTv

4 6 8 10 12

EX. NO. 12

distance 6 m distance 4 m

(b)

Q

0.2

ur /av

0.4

EX. NO. 12

distance 10 m

(c)

0.6

distance 6 m

distance 4 m

-iJ 6 a

Q

EX. NO. 17

distance 10 m

(d)

0,2

ur/(7v

0.4

EX. NO. 17

(e)

distance 6 m

distance 4 m

Fig. 5 (a) Cross-section showing the location of the explosion

point and instruments

(b,c,d,e) Increase of pore pressure with depth

III-47
0.7

ur/Ov 01

N <lo

07

° B zone

• C zone figures of each point

= N-value

N .->10

5 6 7 Ud/Oy

/I 12

Fig. 6 Relation between dynamic pressure U^j and residual pressur Ur

_ 2 E

u

c

(De

CD 6

w 8

Distance (m)

4 6 8 —I ^i-— 1

EX. NO. 12

10

E 2 u

i5 4 c

0)E

(1) 6

i-H P

P

w 8

Distance (m)

2 4 6 8

(a) Fig. 7 Settlements of the ground surface

Average N-value

2 4 6 8 10

EX. NO. 17

(b)

Ur (kg/cm2)

0.1 0 7 O.i 0.4 0.5

E 2 o — 2 Eu

(a) (b) Fig. 8 Relation between settlement and (a) average N-value

and (b) residual pressure

III-U8
< S-l J* • Explosion point

y O Sounding point

•o

S-2

o •

S-5 oS-4-

(a) S-4-

depth (m) number

weight (kg)

0 3S 50 75 u

of half turns per m

a 20 *o to So loo 200 30O

0 1

/

- 2-

3«

-

5

-

6

- 7- 8-

9

-

10-

1=

(b) before expl'osion

S-5

depth number c weight (kg)<

0 25 50 75

if half turns per m

T 20 ^ 60 go 100 200 300

0

/

-

2

-

J

-

«

5

i

-

7

-

8

- 9-

/<?-

1

c

.

1—

(d) before explosion

S-4-

depth(m) number

weight (kg)

0 25 so 75 /

of half turns per m

3 10 «0 60 80 /OO 200 ioo

/

-

2

3

« •

5

6

7

8

-

9

-

10-

\

L

L 7

i

—

(c) after explosion

S-5

x;

-p ~

a e

(1) —

2

3«5

<5

78

9

nurrujer of half turns per m

we ight ( kg ) o 20 M 60 80 /oo 200 300

0 25 50 75 /OO

(e) after explosion

Fig. 9 (a) Location of the explosion point and the sounding point

(b,c,d,e) Penetration resistance with depth

III-49
012

010

S S 0.08

w

(U 0) 0 a

a

D tH T) £1

•H G

ui ra 0)M 0) ^5 0.04

rt) o

^ (D > 1+-4 CC 0)

0.02

stress condition

^ Dr = 50%

* Dr = 70%

O Dr = 80%

Dr = 50%

^ X 1

^Dr = 70%

/

/ o

/ o

' o

, Dr = 80^

- - - -K - r

0.1 0.2 0 3 0.4 0.5 Dynamic shear stress versus effective ambient pressure ady2o's

Fig. 11 Relation between dynamic shear stress and residual pore

pressure by dynamic triaxial tests on fine sand

I II -50
Table 1 Condition of experiments

EX. NO. Test

site

Amount of

^ 1 I Ci J- C Depth ofG 1o s i o

Measured quantity

1

A

0.2 4 Acceleration on ground surface

Pore water pressure

2 0.5 4

3 0.5 6

4 1.0 6

5 1.0 8

6

1. 0 X 2

(1 sec interval)

6 Acceleration on ground surface

and under ground

Pore pressure

Settlement of ground 7 0.9 6

8

B

0.2 6 Acceleration on ground surface

9 0.2 6

10 0.2 6 Acceleration on ground surface

and under ground

11 0.2 6

12 1.0 6 Acceleration on ground surface

Pore water pressure

13 1.0 6 Settlement of ground surface

14 1.0 6 Acceleration on ground surface

Pore water pressure

15 1.0 6 Settlement of ground surface

16

C

1.0 6 Acceleration on ground surface

Pore water pressure

17 1.0 6 Settlement of ground surface

18 1.0 6 Acceleration on ground surface

Pore water pressure

19 1.0 6 Settlement of ground surface

III-51
LANDSLIDE INCIDENCE AND MECHANISMS DURING EARTHQUAKES

by

George E. Ericksen

U.S. Geological Survey

Reston, Virginia 22092

Strong earthquakes affecting mountainous terrain are generally accompanied by hundreds or even thousands of large potentially destructive landslides

of certain types; earthquakes affecting areas of low relief cause fewer and generally different types of landslides, which, however, may be equally destructive

to works of man. On steep slopes, among the many types of landslides that may

occur, falls, slides, and avalanches of rock and soil are most frequent during

earthquakes. These landslides take place where slides are commonly part of

normal mass-wasting processes that affect hillslopes; the earthquake causes the

reactivation of old slides as well as the formation of new slides. Surface move- ment on faults may also cause landslides by the formation of scarps that change

slope stability although landslides of this type are relatively rare. In terrain

of low relief, failures by rotational slump, translatory sliding, and lateral

spreading are frequent causes of destruction in towns and cities that have been

constructed on unstable, generally water-saturated soil or unconsolidated sediment

Landslides, with the exception of those classified as falls, result from fail ure of earth materials under shear stress. Earthquake accelerations trigger landslides by causing a transitory increase in shear stress in earth materials at the

site of the slide, and by causing a decrease in the shear strength of certain ma- terials, such as water-saturated soil. Most slide failure takes place along one

or more planes of weakness, except for flows or landspreading, wherein a given mass fails by loss of coherence. Among the most spectacular or earthquake- triggered landslides are large high-speed debris avalanches that move over a cushion

of entrapped compressed air.

Key Words: Avalanches; earthquakes; falls; flows; land slides; mechanisms;

slides.

III-52
INTRODUCTION

Landslides, which may be widespread and frequent during large earthquakes, are among

the most destructive of earthquake-related phenomena. They have been major causes of destruction during large earthquakes (M > 7.5) in western North and South America during

the past two decades. Among the most destructive were the translatory slides affecting

seacoast towns and cities during the 1960 earthquake in southern Chile and the 1964 Alaska

earthquake, both M = 8.5, and the Huascaran rockfall avalanche that obliterated the Andean

city of Yungay during the 1970 Peru earthquake (M = 7.7). In addition, each of these

earthquakes triggered thousands of other landslides that caused extensive damage to works

of man.

Landslides are part of the mass wasting of hillsides and thus are part of weathering

processes that over time result in leveling of terrain. Most landslides are the final result of the evolution of stress conditions which develop an unstable slope that only requires a triggering mechanism to produce a landslide. Earthquake vibration is one triggering mechanism that is unique in causing great numbers of slides to occur in seconds or

minutes, slides that under normal conditions of weathering might occur over periods of

hundreds or even thousands of years . The earthquake may cause renewed movement of old

landslides as well as the formation of new slides, which may or may not fail under normal

stress conditions. Some earthquake- triggered landslides are clearly larger than they would

have been had they been triggered by some other mechanism. Causes of landslides other than

earthquakes are: lubrication or saturation of the slide mass by water, overloading by manmade structures, and changes in slope stability because of stream undercutting, surface

faulting, and manmade cuts.

The effect of earthquake- triggered landslides on terrain stability is variable. In

some places, a given slide area is more stable after the earthquake, as, for example, where

rockfalls and soilfalls on over- steepened slopes result in a more gentle stable slope; in

others, the area is less stable, as, for example, in terrain where fissuring and incipient

sliding caused by the earthquake result in increased water infiltration that causes a former stable terrain to become unstable. Examples of changes in slope stability such as these are generally readily recognizable in the field, but many times the stability of a given slope after an earthquake depends on more subtle geologic features that may be difficult to recognize.

CLASSIFICATION OF LANDSLIDES

The landslide classification used in this report follows that of Varnes (1958, p. 21)

,

wherein landslides are characterized by type of movement and type of material (Figure 1)

.

Landslides are considered to involve downward and outward movement of slope-forming materials consisting of bedrock, soil (includes all natural unconsolidated material on bedrock)

,

and manmade fill. Movement may consist of falls , slides , and flows as well as combinations

of these. In this report, translatory slides are considered to be landslides, even though

their movement is essentially horizontal and failure in some (landspreading) may be due

to incipient flow of liquefied materials or to plastic deformation.

The various types of landslides and their size and frequency are determined largely

by the slope angle at the slide area and by the physical character and structure of the

slide material. Falls of rock and soil break away from oversteepened or undercut slopes,

and movement is largely by free fall of material through the air. Slides take place by

gliding along one or more planes or zones of failure, which may be curving and concave upward (slump), inclined planar (block glide, soil slip, debris slide), or horizontal

(translatory slide) . Rotational and translatory slides (Figures 2 and 3) generally occur

in poorly drained unconsolidated sediments and weak bedrock on gentle slopes or in horizontal terrain at the margins of topographic depressions. Debris slides (Figure 4) are

most prevalent in cohesionless soil and weathered bedrock on steep slopes. Flows , which

take place on slopes ranging from steep to nearly horizontal, result from loss of shear

strength of a material so that it behaves as a viscous fluid (Figure 5) . Flows generally

form in water-saturated fine-grained sediments, which become liquefied because of external

III-53
forces such as earthquake vibration, but also occur in dry fine-grained sediments such as ,j cl sand and silt. Complex landslides (Figure 1) involve a combination of the above types of | movement and materials. In the strictest sense, nearly all landslides are complex althouc

generally one type of material predominates in the slide mass, and one type of movement

dominates in certain parts of a landslide or at a particular time during its movement. |

Figure 2 shows the form and nomenclature of the component parts of a rotational slide or slump. This terminology may be applied to landslides of all types although most land^ , slides are not as complex as the rotational slide, and consequently only part of the terms are applicable to these slides. The slide components may be defined as follows (Varnes,

1958, p. 1)

:

Main Scarp : Scarp or surface of rupture at the head of the slide, which con- ! tinues beneath the slide and is the principal plane of failure,

the surface of rupture .

Minor Scarp : Scarp in the slide formed by differential rotational movement be-' tween blocks within the slide mass; a plane of failure that may ex tend downward to the surface of rupture .

Head : Uppermost part of the slide mass. i

Top ; Uppermost part of the slide mass at the main scarp.

Foot ; Line of intersection of the lower part of the surface of rupture and the

original ground surface; may be either covered, as shown in Figure 2, or marked by an upthrust ridge.

Toe: Outermost limit of the slide material.

Flank: The side of the slide.

Crown : Relatively undisturbed material above the main scarp .

Transverse Cracks : Tensional fissures within the slide mass.

Longitudinal Fault Zone : Plane or zone of shear parallel to slide movement,

caused by differential movement within corrponent

blocks of the slide.

LANDSLIDES DURING EARTHQUAKES

Mechanisms , .

Landslides take place in a given rock or soil mass when the internal shear stress ex- ceeds the shear strength of the mass as a whole, or of planes or zones within or at its

margins. Shear stress within an earth mass on ground that is not level is due to gravity

and to seepage force.''" Shear strength, the property that allows material to remain in

equilibrium on surfaces that are not level, is a function of degree of cementation or induration of bedrock and of friction or electrostatic bonding between grains as well as cementation of soil.

Most types of rock have shear strengths so great that their only mode of failure is \ by rupture along structural planes of weakness such as faults, fractures, joints, and bed-j

ding planes. The shear strength of soil, on the other hand, is low. In the uncemented

cohesionless soils (those lacking or having only sparse clay minerals) , the shear strengthi

'"Seepage force results from flow of water through a soil, and is equivalent to loss

of head because of transformation of seepage force to effective stress by frictional drag. In an isotropic soil, seepage force acts on the direction of flow (Lambe and Whitman,

1969, p. 261-262)

.

III-54
is chiefly a function of friction between component grains and rock fragments, whereas in

clay-bearing cohesive soils it is a function of bonding between grains that results from

adsorbed ions. Cohesive soils are characteristically plastic because of the presence of

clay minerals of colloidal size, and such soils tend to deform plastically when shear stress

exceeds shear strength. Cohesionless soils, in contrast, are not plastic and fail either

by rupture or flow.

Earthquake accelerations cause transitory increases of shear stress in materials on hillsides and significant decreases in shear strength in some materials such as watersaturated soil. Most earthquake-triggered landslides are due to the temporary increase in

shear stress within a soil or rock mass that is already in a near-critical stress condition.

Alternatively, cyclic loading due to the accelerations may cause a gradual increase in pore

pressure in undrained cohesionless soil to a point of complete loss of shear strength and

bearing capacity, and the soil liquefies (Lambe and Whitman, 1969, p. 444-445) . This is

referred to as a quick condition, and soil in this state will flow freely. A quick condition develops most readily in fine sand and silt; it does not develop in cohesive soils.

Many landslides fail by rupture along distinct geologic structures, such as beds and

bedding planes in stratified rocks, and unconformities, faults, fractures, and joints in

rocks and soils of all types, whereas others fail along planes or zones that do not mark

any apparent structural discontinuity. Of these latter types of landslides, the planes

of rupture of some are at interfaces in rock and soil that mark subtle changes in moisture

content, leaching, and alteration, whereas the position of rupture in others is due to a combination of factors —slope angle, physical character of the slide mass, and intensity

of earthquake shaking. Rotational slides are among those that typically lack a controlling

structure at the plane of rupture (except for those that have been reactivated) and involve blocks that rotate about an axis parallel to the hillside. Rotation is due to moment

imparted by gravity, whereby the block tends to move downward because of its weight and

outward toward a free face. A curving surface of rupture is a resultant of these combined

movements. Degree of saturation and flow of water at the slide site, including seepage

force, may influence position and curvature of the surface of rupture. Earthquakes may

trigger rotary slides when the transitory shear stress due to seismic loading acts in a direction that increases the moment about the axis of rotation.

Translatory slides, in contrast to rotary slides, occur in materials having horizontal

planes of weakness that control failure or that deform plastically or by liquefaction.

Movement at the head of some translatory slides is rotational, so that one or more slump

blocks form; the heads of other translatory slides are marked by graben over a downdropped

wedge, as shown in Figure 3. Most known translatory block glides during earthquakes have

taken place along bluffs with extensive horizontal or near-horizontal surfaces in back and

in front of them (Idriss and Seed, 1966, p. 2) . The slide blocks involve material in and

behind the bluff, which slides forward over the land surface in front of the bluff. Lateral movement is due to the thrust of sl\imp blocks and downdropped wedges of material at

the head of the slide and within the slide, as well as failure by landspreading and flow.

Translatory slide blocks such as those at Anchorage, Alaska, triggered by the 1964 earthquake, probably move outward along a layer of sensitive clay or water-saturated finegrained sediment that had been liquefied by cyclic loading during the earthquake (Hansen,

1965, p. A65)

.

Dry flows (Figure 1) may form during an earthquake either as a direct result of the

transitory increase in shear stress, by destruction of coherence because of cyclic loading,

or by transformation of a fall or slide to high-speed flow or avalance, many of which have

air layer lubrication. Unconsolidated silt and sand, which have little shear strength,

are the materials most subject to flow because of increased stress. Cyclic loading of

fairly coherent fine-grained porous materials such as loess or dry mudflows, which are commonly stable in vertical cliffs, may destroy their coherence by causing collapse of the

skeletal framework, so that a flow will form on any slope having an inclination greater

than the angle of repose of the newly formed noncoherenc material. Falls of these same materials may be pulverized upon impact and develop into flows. High-speed avalances may

move on cushions of entrapped coit5>ressed air or contain entrapped air that keeps particles

in suspension.

III-55
Destructive Landslides During Earthquakes

Falls . - Rockfalls and soilfalls (Figure 1) are frequent and widespread during earthi

quakes affecting mountainous terrain, particularly glaciated terrain. Most falls are small, ranging from a few m. ^ to a few hundreds m.-^, but large falls containing mil- i lions of m. 3 of material have occurred. Falls generally consist of masses of rock or soil that break away from a steep slope or cliff along a plane of weakness such as a bedding plane, fault, or joint. On steep hillsides the fall may change to a debris

slide (Figure 4), or if large enough, an avalance (Figure 6). Also, rock fragments

from a fall, as well as isolated boulders dislodged during the earthquake, tumble and bounce down steep hillsides, killing people and animals in their paths, and crashing

through buildings. During the 1970 Peru earthquake, such rockfalls caused thousands

of injuries and deaths to people and livestock, and destruction of farm buildings in

the high Andes of northern Peru, an area where steep hillsides have been extensively

farmed since pre-Columbian times. In contrast, the great number of rockfalls during

the Alaska earthquake of 1964 caused comparatively little damage because they oc- curred chiefly in unpopulated areas.

During the Alaska earthquake of 1964 and the Peru earthquake of 1970 several hig

speed avalanches were generated by rockfalls. These falls were large, probably havin

volumes of more than a million m. of material, and vertical drops of at least severa hundred meters. The Huascaran debris . avalanche (Figure 6), the most destructive land

slide of historic time, originated by fall of a slab of granodiorite and ice from

the near-vertical west face of the north peak of Nevados Huascaran. This slab was estimated to contain more than 25 million m.^ of material, being approximately 800 m. wide and 1,000 m. long (Ericksen and others, 1970, p. 7-8). The average vertical

fall of the slab, taken as the vertical drop at its center, was about 600 m.

Slides . - On the basis of mechanics of movement, two types of slides can be recognize

one in which the slide mass is relatively little deformed and the other in which it

is greatly deformed (Varnes, 1958, p. 23-24). Sluitps and block glides (Figures 2 and

3) are typical of the undeformed slides, and debris slides and failure by lateral

spreading are examples of deformed slides. In this report, slides are divided into

three types on the basis of inclination and form of the plane of rupture as follows: '

1) rotational, 2) inclined planar, and 3) translatory. Inclined planar slides are by

far the most frequent of these types, cind together with rockfalls and soilfalls make

up most of the slides that are triggered by earthquakes. Comparatively few rotationa.

and translatory slides occur during any given earthquake, but because they tend to occur in unstable material in areas of low relief, which also may be sites of towns and cities in mountainous areas and bordering coastal regions, they may be major

causes of destruction.

Slumps and translatory slides tend to occur in terrain of gentle relief underlaii

by unstable bedrock or unconsolidated sediments. They tend to be large, commonly involving tens to hundreds of thousands or even millions of m.-^ of material. They may , break and move either as essentially a single unit (Figures 3 and 7) , as several seg- ments (Figure 2) , as many segments (lateral spreading) (Figure 3) , or deform by incipient flow (landspreading) . As also shown in these figures, a flow, slide, or upthrusi

ridge may form at the toe of the slide. Earthquake- triggered slides of these types

commonly occur in areas where scars of earlier slides can be seen.

Slumps and translatory slides triggered during earthquakes in the Western Hemisphere during the past two decades caused extensive damage to cities, notably to coastal cities during the 1960 Chile earthquake, the 1964 Alaska earthquake, and tlie 1970 Peru earthquake. They also caused damage to transportation routes (Chile and

Alaska earthquakes) , and dammed streams to form temporary lakes that had to be

drained (Chile and Peru earthquakes) . A subaqueous slump on a delta at Kenai Lake,

Alaska, caused destructive waves to form in the lake. The sequence of conditions at the site of the slide is shown in Figure 9. As can be seen in the figure, the slump

caused a wave to wash over the area back of the main scarp. This wave and another

that swept over the opposite lakeshore, a distance of about 100 m. , each attained a maximum height of 10 meters above normal lake level (McCulloch, 1966, p. AS)

.

III-56
Inclined planar slides that are most frequent during earthquakes are debris

slides on steep hillsides in unconsolidated regolith and in weathered rock. Such

slides originate as failure of a thin slab of surface material aiong a flat to gently

curved surface. On steep slopes, the moving material generally becomes a debris

slide that moves downward away from the source area. Depending upon the type of ma- terial involved, movement becomes less and less on more gentle slopes, to a point

where the slide remains essentially in place, being marked only by fracturing and

small soil slips.

Debris slides, which may occur by the thousands during earthquakes, along with

falls are among the most nximerous but smallest of earthquake-triggered landslides.

Most involve not more than a few tens to a few thousands of m.-^ of material, but they

may be so numerous on some hillsides that it is almost iitpossible to distinguish individual slides. As in the case of rockfalls and debris falls, debris slides cause damage to farms and isolated small communities . An unusual type of damage by debris

slides, which occurred during the 1970 Peru earthquake, was destruction of fields and

trails in steeply inclined areas by sliding of thin soil to expose underlying fresh

bedrock. Not only were farmlands destroyed but the construction of new trails across

the newly-exposed bedrock was prohibitively expensive; in some are^/ farm buildings

and lands that were not otherwise damaged had to be abandoned.

Comparatively rare debris slides are those caused by surface faulting. Figure

10 shows the development of the scarp of a reverse fault that moved during the 1964

Alaska earthquake, and caused many slides to break away from the hillside on the up- thrown block of the fault.

Flows . - Although flows of one type or another (Figure 1) are triggered by earthquakes,

most are small and cause relatively little damage. The major exceptions are debris

avalanches, which are potentially among the most destructive of all landslides. As has been noted, flows occur in both dry and wet or water-saturated material. The most

notable earthquake-triggered dry flows were the loess flows during the 1920 earthquake

of Kansu Province, China, where loess banks failed, and a fluid mass of dry powder

filled valleys and buried villages (Close and McCormick, 1922) . Flows of wet material

range from the relatively viscous earth flow that moves only short distances (Figure

5) to debris and midflows, v^ich are highly fluid masses that may travel for tens of

kilometers

.

Debris avalanches, which start as either wet or dry material, may become debris

flows if they move into a channel having a flowing stream, as happened to an avalanche near Caraz (Figure 11) during the 1970 Peru earthquake and to the Huascaran avalanche

as it flowed into the Santa River (Figure 6) . Most earthquake-generated flows originate by liquefaction of water-saturated material under cyclic loading (see p. 111-57) . Liquefaction of bogs and swamps result in earth flows and mudflows during earthquakes

.

At least two major mudflows and many small earth flows originated in this way during

the 1970 Peru earthquake.

Some submarine slides may be due to liquefaction of fine-grained material, as occurred at the port of Valdez during the 1964 Alaska earthquake, causing total destruction of the dock facilities (Figure 12). This slide, described by Coulter and

Migliaccio (1966), involved about 75 million m. of water-saturated silt and sand.

Failure was sudden, taking place shortly after the earthquake tremors were first felt.

The slide caused formation of destructive water waves as much as 10 m. high.

Subaqueous flows caused by liquefaction of water- saturated sediments are probably

much more common than generally recognized. Undoubtedly, such flows occurred at other

localities during the 1964 Alaska earthquake , and they probably took place along the

south Chile coast during the 1960 earthquake. Large siiimarine flows did not occur

during the 1970 Peru earthquake, but small submarine flows probably did occur in areas affected by landspreading, notably along the bayshore of the coastal city of Chimbote.

111-57
Avalanches are moving masses of earth materials that attain sufficient speed to move by flow rather than shear. They occur frequently during earthquakes, generally

originating as falls or slides on slopes steep enough and with sufficient vertical

drop for the moving mass to attain a velocity necessary to transform it into a flow.

Thus, avalanches may be either dry or wet and consist of rock, soil, ice, or snow, or a combination thereof.

In areas of glaciers or regions of heavy winter snows, earthquake-triggered ice

and snow avalanches may be numerous and widespread. Although such avalanches are potential earthquake hazards, they generally do not cause damage because most are con- fined to uninhabited ice and snowfields, or flow into sparsely inhabited valleys below. Such was the case of dozens to hundreds of snow and ice avalanches during the

earthquakes in Alaska (1964) and Peru (1970) , where large areas are covered with

glaciers and snowfields.

Large high-speed avalanches are among the most awe-inspiring and potentially most

destructive of all types of landslides. They are generally triggered by huge falls or slides of rock, on the order of millions of m.-^ minimum, on steep to vertical

slopes where vertical drops are on the order of hundreds of meters. They may attain

speeds of several hundred km/hr. , and move for several kilometers outward over gently

undulating to near-horizontal terrain at the base of steep mountain slopes. At high

speed the avalanche tends to entrap air which is compressed into a cushion on which

the avalanche overrides irregularities in terrain or even ridges without causing significant modification of their form. Observers have reported such avalanches as being accompanied by strong turbulent blasts of air (Plafker and others, 1971, p.

557) . . .

\ The largest and most destructive avalanche of historical times was the Huascaran,'

a rockfall avalanche that occurred during the 1970 Peru earthquake (Ericksen and

others, 1970; Plafker and others, 1971). This avalanche probably involved more than ^ 50 million m. of rock and ice; during its trajectory over a horizontal distance of

14.5 km and a vertical descent of about 3,000 m., it attained maximum velocities of

more than 400 km/hr. The avalanche obliterated the Andean city of Yungay and nearby

small settlements and farms over an area of about 22 sq. km, causing the death of

more than 18,000 people.

REFERENCES

1. Baker, R.F., and Chieruzzi, Robert, 1959, Landslide Occurrence and Analysis: National

Research Council, Highway Research Board, Bull. 216, 43 p.

2. Close, v., and McCormick, E., 1922, Where the Mountains Walked: Nat. Geog. Mag., V.

41, p. 445-464.

3. Coulter, H.W., and Migliaccio, R.R. , 1966, Effects of the Earthquake of March 27,

1964, at Valdez, Alaska: U.S. Geol. Survey, Prof. Paper 542-C, 36 p.

4. Davis, S.N., and Karzulovic, K. , Juan, 1963, Landslides at Lago Rinihue, Chile: Seismol. Soc. America, Bull., V. 53, No. 6, 1403-1414.

5. Eckel, E.B., 1970, The Alaska Earthquake, March 27, 1964: Lessons and Conclusions:

U.S. Geol. Survey, Prof. Paper 546, 57 p.

6. Ericksen, G.E., Plafker, George, and Fernandez Concha, Jaime, 1970, Preliminary Report

on the Geologic Events Associated with the May 31, 1970, Peru Earthquake: U.S. Geol.

Survey, Circ. 639, 25 p.

7. Hansen, W.R. , 1965, Effects of the Earthquake of March 27, 1964, at Anchorage, Alaska:

U.S. Geol. Survey, Prof. Paper 542-A, 68 p.

III-58
8. Idriss, I.M. , and Seed, H.B., 1966, The Response of Earth Banks During Earthquakes:

Berkeley, California, University of California, Soil Mechanics and Bituminous Materials

Research Lab., 22 p.. Appendices.

9. Karol, R.H., 1960, Soils and Soil Engineering: Englewood Cliffs, N.J., Prentice-Hall,

194 p.

10. Komarnitskii , N.I., 1968, Zones and Planes of Weakness in Rocks and Slope Stability:

New York, Consultants Bureau, 108 p. ( English Translation of Vliyaniye zon i poverkhnostey oslableniya v porodakh na ustoychivost ' otkosov: Moscow, Izd. "Nauka,"

1966.)

.

11. Lambe, T.W. , and Whitman, R.V. , 1969, Soil Mechanics: New York, John Wiley and Sons,

553 p.

12. McCulloch, D.S., 1966, Slide-induced Waves, Seiching, and Ground Fracturing Caused by

the Earthquake of March 27, 1964, at Kenai Lake, Alaska: U.S. Geol. Survey, Prof.

Paper 543-A, 41 p.

13. McLerran, J.H. , 1957, Glossary of Pedologic (Soils) and Landform Terminology for Soil

Engineers: Natl. Research Council, Highway Research Board, Spec. Rept. 25, 32 p.

14. Plafker, George, 1967, Surface Faults on Montague Island Associated with the 1964

Alaska Earthquake: U.S. Geol. Survey Prof. Paper 543-G, 42 p.

15. Plafker, George, Ericksen, G.E., and Fernandez Concha, Jaime, 1971, Geological Aspects

of the May 31, 1970, Peru Earthquake: Seismol. Soc. America, Bull,, Vol. 61, No. 3, p. 543-578.

16. Ruiz F., Carlos, and Saint-Amand, Pierre, eds., 1961, Observations Concerning the

Chilean Earthquakes of May 1960: Internat. Geol. Cong., 21st, Copenhagen, 1960, Rept.

pt. 26, p. 116-133.

17. Varnes, D.J., 1958, Landslide Types and Processes, Chap. 3 of Eckel, E.B., ed.. Landslides and Engineering Practice: Natl. Research Council, Highway Research Board,

Spec. Rept. 29, p. 20-47.

III-59
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III-60
III-61
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III-62
Figure 4. - Block diagram of a debris slide in which

failure took place along the contact between soil and

bedrock. From Varnes (1958, pi. 1).

III-63
Figure 5. - Block diagram of an earthflow in

weathered shale. From Varnes (1958, pi. 1)

.

III-64
Figure 6. - Oblique aerial view of the Santa Valley

and snow-covered Cordillera Blanca showing the

Huascaran rockfall avalanche that devestated the

Yungay-Ranrahirca area during the 1970 Peru earthquake.

Photograph by Servicio Aerof otograf ico Nacional del

Per^, June 13, 1970.

III-65

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III-67
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III-68
Figure 10. - Diagrams illustrating fault movement and

landslide development along the Patjon Bay fault, a reverse fault on which displacement took place during the

1964 Alaska earthquake. From Plafker (1967, fig. 17)

III-69
Figure 11. Huge ancient rotational landslide (R) in the Callejon

de Huaylas, east side of Santa River, 2-3 m north of Caraz. This

slide, which is 5 km long and averages 1 km wide, dammed the

Santa River to form a lake in which was deposited a thick segment

of silts and sands (SS) now exposed in river banks. Head of slide,

outlined by dashed line, shows no evidence of movement during 1970

earthquake. Also shown is large landslide (LS) , triggered by the

1970 earthquake, which developed into a fluid debris flow in small

tributary valley (V); debris slide (light-colored scars) also were triggered by the 1970 earthquake.

III-70
Figure 12. - Submarine slide that destroyed dock facilities

at Valdez, Alaska, during 1964 earthquake; water saturated

silty and sandy delta deposits liquefied under cyclic

loading due to earthquake accelerations. From Coulter

and Migliaccio (1966, fig. 4).

III-71
STRESS CONDITION EFFECTS ON DYNAMIC PROPERTIES OF SOILS

by

Eiichi Kuribayashi

Chief, Earthquake Engineering Research Section

Public Works Research Institute

Ministry of Construction

Toshio Iwasaki

Chief, Earthquake E; lineering Section

Building Research Institute

Ministry of Construction

In the engineering field, the evaluation of the dynamic characteristics of

soils and foundation subgrades has been required in the course of studying vibrational problems and especially problems associated with earthquake engineering.

The dynamic characteristics of soils have, therefore, been obtained by laboratory

tests, using a resonant column method. These tests were conducted to evaluate

the shear modulus and the damping characteristics in dry and saturated specimens

of various soils in Japan.

The hollow cylindrical samples tested were 25 cm. in height, 10 cm. outside

diameter, and 6 cm. inside diameter. This arrangement permitted a more uniform

deformation of the sample cross-section. The specimens were fixed at the bottom,

and at the top of the specimens oscillators were fastened to a rigid mass, which

supply a torsional vibrational force to the system. A confining pressure, which

was applied equally to the outside and inside of the test sample, was supplied

by air pressure. An axial load was also applied, and was independent of the con- fining pressure in order to produce an anisotropic stress condition.

After the sample was prepared and all proper alignments and forces imposed,

the frequency of the torsional excitation wss introduced and then varied \intil

the oscillator specimen system resonated. The resonant frequency varied from 40 cps to 100 cps, depending upon the dimensions and density of the sample and the

applied stress condition and the shearing strain amplitude. The strain amplitude

varied from 5 x 10~^ to 5 x 10"*^ for these tests. The shear moduli were calculated from the resonant frequencies and the other parameters, as given above.

The damping characteristics were obtained by using the amplitude-time decay re- sponse curves of the free vibrations.

III-72
In the triaxial state of stress, the mean principal stress, p, is defined

by (o ^ + 2 o^)/3 and the deviator stress, q, is given by - o are the axial

and radial stresses, respectively. The test results indicate t£e following

trends for the dynamic properties of soils;

(1) Under constant values of the other parameters, the shear moduli vary

with 1/2 power of p and decreases with an increase in the void ratio

and shearing strain aitplitude. Furthermore, the damping capacity de- creases with an increase in p and also increases with an increase of

strain amplitude. However, the damping capacity remains constant irrespective of the change in the void ratio, when the other parameters

remai n cons tant

.

(2) When the value of p is kept constant, the shear moduli are nearly con- stant irrespective of the value of q until the stress ratio, q/p,

reaches a value of aibout 1.0. However, beyond this value of q/p, the

shear moduli begin to decrease with an increase in q/p. This phenomenon is due to the anisotropic stress condition and the corresponding

anisotropy in the inner structure of specimens

.

Key Words: Damping; damping coefficients; shear modulus; soil; tests; torsional excitation.

III-73
INTRODUCTION

In the civil engineering field, the evaluation of the deformational characteristics

of soils and foundation subgrades has been required for vibrational problems, especially

for those problems associated with the earthquake response of structures. In order to obtain such information, relative to dynamics of soils, elastic wave measurements at field

locations or laboratory tests using simple shear apparatus of triaxial apparatus have been

conducted. However, the shearing strain amplitudes in the soil structures and subgrade

during strong earthquake motions have produced magnitudes of values 10"-^ to 10"^ of strain. It is, however, difficult to evaluate the deformational characteristics of soils at these

magnitudes using ordinary methods.

Therefore, in order to obtain the shear modulus and damping characteristics of soils

subjected to such magnitudes of shear strain, a special resonant column test has been designed and used to test dry and saturated specimens of various soils. This paper presents

the required instrumentation and initial use of the resonant column test apparatus.

APPARATUS

The general apparatus and procedure used to evaluate the moduli of soil specimens in

the laboratory by means of a vibrational method was conceived by lida (1). However, the

detailed equipment which can control the stress conditions and the magnitudes of the shearing strain was developed by Hardin et al (2) . The principal of the apparatus, shown in

the following figures, is the same as Hardin's apparatus. The schematic diagram and photograph of the apparatus is shown in Figures 1 and 2. Figure 3 shows the general stress condition on the test sample. The hollow cylindrical specimens that were tested were 25 cm

in height and had a 10 cm outside diameter and 6 cm inside diameter. This arrangement permitted a more uniform deformation of the cross-section. The specimens were fixed at the

bottom, and at the top of the specimens oscillators were fastened to a rigid mass, which

supply a torsional vibrational force to the system. The confining pressure, which was applied equally to the outside and inside of the samples, was supplied by aid pressure. The

axial load was introduced independent of the confining pressure in order to produce the

anisotropic stress condition similar to the --stress condition in the horizontal ground.

The quantities that were measured during a steady state vibration test were the reso- nant frequency of the oscillator-specimen system, the vibration amplitude at the top of

the specimen, the length and volume change of the specimen, the confining pressure, and

the axial load. Also, the amplitude time decay curves were recorded by shutting off the

driving power. From all of these data, the shearing strain amplitudes, the shear moduli,

and the logarithmic decrements were calculated for the various soils tested, as will be

described in the following sections.

PRINCIPLE OF TESTING

Shear Strain Amplitude '

The shear strain in a sample is denoted by y, and is equal to;

^ 3= 9u = 3 (r, 6) (rad.) (1) 3X 3x -

. •

where u(x, t) is the displacement, 6 (x, t) is the angular displacement in radians, x is

the axial coordinate, and r is the radial coordinate as shown in Figure 4. The configuration of the specimen and the large mass at the top, produce linear deformations and accordingly a uniform shear strain condition over the entire length of the specimen. Furthermore,

the shear strain in a horizontal section can be expressed in terms of the radius equal to

a value of 4 cm. Accordingly, the shearing strain in the specimen is represented as,

Y = 1 e (rad.) (2)

i - where £ is the length of specimen (cm) ....

III-74
shear Modulus

The shear-strain curves of the sand specimens, loaded axially, yield the relationships

shown in Figure 5 and where the hysteretic curve has sharp points. In the case of hysteretic dandling, such as occurs in sands, the linear approximation can be used to model its dis

sipative behavior and to siirplify the dynamic problems. The material constants of sands

will be defined as;

G = Equivalent shear modulus

n = Hysteretic damping coefficient = AW/2ttW

W = Strain energy (shown in Figure 5)

AW = Damping energy (shown in Figure 5)

For an eqioivalent linear isotropic material having the same values of G and n, as given

above, a more useful notation is in the form of a complex modulus, as shown in Figure 6. In this form t and y a^re the shear stress and strain in complex notation, and G and G' are

the complex coefficients. Then, stress-strain relation is expressed as

T = (G + iG')Y (3)

The hysteretic loop of such a linear material, excited by steady vibrational force, is

such as shown in Figure 7 and has no sharp points. Defining the strain energy W and damping energy AW, as given in Figure 7, then the hysteretic damping coefficient ri can be re- lated to the conplex modulus G and G' as

n = GV2G (4)

However, the wave propagation equation for the shear case is; p3u_ = (5) 3t2 3x

where p is the density and u is the displacement orthogonal to the propagation direction.

Then, siabstituting Equations (1) , (3) , and (4) into Equation (5) gives the wave equation

for linear isotropic material used in the resonant column tests, and is;

2 2 = G(l + i 2n)i-A (6) dt^ 3x2

The analytical solution of this equation for the model shown in Figure 4 is not as simple

as the one given by Hardin (3) and Hardin and Music (2) . Under steady state vibration,

the shear modulus G is a function of the density and the dimensions of the samples, the

apparatus constants, and the resonant frequency. Thus, in order to obtain a value of G, an interative procedure using an electronic computer was required.

Damping Coefficients

By setting the system into a steady-state forced vibration and then shutting off the

driving power, the logarithmic decrements. At was obtained. From the logarithmic decrements, the hysteretic damping coefficient n, which does not depend upon sample's dimensions and boundary conditions, was obtained. For the free vibration of the model, shown

in Figure 4, we can assume a solution of the form;

9 = A(x)ei('^n - At)

(7)

where A(x) is the mode of deformation, is the natural circular frequency of the system

and X is the attenuation factor with respect to time. Substituting Equation (7) into

Equation (5) gives;

X = — [-1 + /I + (2n)2 ] = cj n (8)

2

III-75
Then, relating the logarithmic decrement to the hysteretic damping coefficient as

A t = X ^ = 2 irn . (9) "n

Now using Equation (9) , the hysteretic damping coefficient n can be obtained by measuring

the logarithmic decrement. Furthermore, it is worthwhile to note that the logarithmic

decrement depends only on the damping characteristics of the sample and does not depend

upon sample's dimensions nor boundary conditions providing the hysteretic damping coefficient n of sartple material does not vary with the frequency of the forced vibration.

TESTING METHODS

Two kinds of sands, Toyoura-sand and Senkenyama sand, whose physical properties are shown in Table 1 were tested.

Table 1 Physical Properties of Sands

DlO Deo "c ®max. ^min.

Toyoura-sand 2 641 0 12 mm 0.145 mm 1.21 0.953 0.686

Sengenyama-sand 2 695 0 16 0.38 2.37 0.961 0.484

Toyoura-sand has a uniform gradation with round particles and is used as the standard

sand for testing in Japan, Sengenyama-sand from Sengenyama near Tokyo is considered well

graded. The resonant column tests were conducted on samples having various void ratios.

For the Toyoura-sand, air-dried and saturated samples were prepared and subjected to stress ratios (a^/oi) equal to 0.5 and 1.0. The Sengenyama-sand, in which only air-dried

specimens were tested, the stress ratio io i) was set equal to 1.0.

After applying the confining pressure and the axial load to the sample, the stress condition that will be designated as (a-j) Sind (o-|_) was kept constant and the vibratory

shear strain amplitude was then increased above 5 x 10"^ to about 2 x 10"^^ Both the

shear modulus and logarithmic decrement was measured at various values of shear strain am- plitude. Then, another stress condition was imposed and the above mentioned procedures

were repeated. Figure 8 describes the test results obtained from one of the Toyoura-sand

samples. The term p is the principal stress denoted by 1/3 (a^ + 2a ^) and e is the void

ratio, which slightly changes value during consolidation. In one sample, it would be impossible to control all the values of the shear strain amplitude, stress condition, and

void ratio. It is, therefore, necessary to convert the measured values of G into the

values of the predetermined magnitudes of p, y, and e. However, Hardin et al showed that

the experimental G equation for Ottawa-sand, for a strain y = 10"'*, is equal to;

G = 697 (2-17 - e)^ pO.5 (10)

1 + e where G is the shear modulus (kg/cm^) , p is the mean principal stress (kg/cm^) , and e is

the void ratio. The exact values of G for strain y = 10"^ and 10"^ were first obtained.

Then G was coitputed using Equation (10) , where the mean void ratio value of the tested

sample was used (e = 0.81). The change of the G values using this procedure varied by

only a few percent. The values of G and p were then plotted on a full-log graph, as shown

in Figure 9, then the values of G for p = 0.5, 1, 2, 3, 4, and 6 kg/cm^ were obtained.

DYNAMIC CHARACTERISTICS OF SANDS

Effects of Shear Strain Amplitude on Shear Modulus

Figures 10 through 13 show the effects the shear strain amplitude has on the shear

modulus. The ratios of G to {G}y = 10"^, i.e., (shear modulus at y = 10"^) and shear

III-76
strain amplitude, are plotted on a semi-log graph. It was found that G/f G^Y = 10"^ v . Y curves were not affected by changes in void ratio, principal stress ratio, and types of

sands. However, when the mean principal stress is of a greater magnitude, the G/( gIy = 10"^

ratio decreases with an increase in y • It was also found that in these tests, the variation of the shear modulus with shear strain amplitude is smaller than the values given in

a chart made by Seed and Idriss (5), even when p is as low as 0.5 kg/cm^.

Effects of the Mean Principal Stress on the Shear Modulus

In all of the tests, the shear modulus varied with the 1/2 power of the mean principal

stress, designated as p, as shown in Figures 9, 14, 15, and 16.

Effects of the Stress Ratio on Shear Modulus

Figures 17 and 18 show the effects of the stress ratio on the shear modulus when the, mean principal stress is a constant for air-dried Toyoura-sand. It can be seen that the stress ratio has little effect on the shear modulus under a constant value of p within a certain range of stress ratio, i.e. ^^'^ i = 0.5, 1.0. This is the same trend as noted by

Hardin and Block (4) . To examine the effects when the range of the stress ratio was

greater, one test was conducted where the stress ratio was increased to failure under a constant value of p = (3 kg/cm^) . The results are shown in Figure 19, where the stress

ratio is denoted by

q/p = (11)

1/3 (^^ + 2 0^)

This figure shows that shear moduli are nearly constant, irrespective of the variation

of stress ratio, q/p, until q/p reaches a value of 1.0, i.e. [a ^/a ^) = 0.4 . However, beyond this value of stress ratio, the shear modulus begins to decrease with an increase in

the stress ratio. This phenomenon is due to the anisotropic inner structure of the specimen which is caused by the anisotropic stress condition.

Effects of the Void Ratio on the Shear Modulus

All the data that was obtained is summarized in Figure 20, where the shear moduli and

void ratios are plotted on a semi-log plot for the values of p = 1 and 5 kg/cm^ for Y =

10"^. As noted in Figure 20, it was found that all the data obtained agreed with Equation

(10) for Y = lO"'*, in the range of ° 2^'^ 1 ^°

Logarithmic Decrements

As stated previously, the damping characteristics of the material tested can be re- presented by the logarithmic decrements of the system when the mechanism of damping in

soils is one of hysteretic dissipation of energy. Figure 21 shows a typical vibrationdecay curve, and by plotting each of the amplitudes against niomber of cycles on a semi-log

plot, as given in Figure 22, the logarithmic decrements were obtained. In Figures 23 and

24, the logarithmic decrements are shown for the Sengenyama and Toyoura sands, respectively.

In each case, it was found that the logarithmic decrements increased with an increase in

the shear strain amplitude and with a decrease in the confining pressure. It was also

noted that the values were independent of the void ratios. It has been shown by Hardin

(3) that the log decrement experimental equation takes the form;

At = 2^ 1_ yO.2 p-0.5 (12)

10

for the range of y = 10"^ to lO"'^, and p = 0.244 to 1.46 kg/cm^. Using Equation (12), the

log decrement At = 0.1, for values of y = 10~^ and p = 1 kg/cm^. Now examination of Figure 23 for the Sengenyama-sand At = 0.17 and in Figure 24 for Toyoura-sand, At = 0.11, thus

it may be stated that the experimental values will take the form of Equation (12)

.

III-77
CONCLUSIONS

(1) The resonant column apparatus that has been developed provides a most effective technique in evaluating the shear modulus and damping characteristics of soils.

(2) The test results on samples of Toyoura-sand and Sengenyama-sand indicate that Equation

(10) fits the experimental results for Y = 10"'^ and in the range of stress ratio

a^/a^ = 0.5 - 1.0. However, when the stress ratio ('-'3/'^]^) is smaller than 0.5, the

shear modulus decreases with a decreasing value of stress ratio ('-'3/*-'i) •

(3) The logarithmic decrements obtained were in the range of At = 0.1 - 0.2 for Y = 10"^

and p - 1 kg/cm^.

ACKNOWLEDGMENTS

This research was encouraged and supported by Dr. Tadayoshi Okubo, Head of Structure

and Bridge Division, P.W.R. I. The tests were conducted by Dr. Fumio Tatsuoka and Mr. Shyunichi Horiuchi, Earthquake Engineering Research Section, P.W.R. I. The authors gratefully acknowledge their help and cooperation.

REFERENCES

1. lida, K. (1938), "The Velocity of Elastic Waves in Sand," Bull. Earthquake Res. Inst.,

15. E.R.I. , Tokyo University.

2. Hardin, B.O. , and Music, J. (1965), "Apparatus for Vibration During the Triaxial Test,

ASTM, STP, No. 392.

3. Hardin, B.O. (1965), "The Nature of Damping in Sands," J. of S.M.F.D., Proc. ASCE,

Vol. 91, No. SMI, January.

4. Hardin and Block (1966), "Sand Stiffness Under Various Triaxial Stresses," ASCE, Vol.

92, No. SM2.

5. Seed, H.B. and Idriss, I.M. (1970), "Soil Moduli and Damping Factors for Dynamic Re- sponse Analyses." Report 'No. EERC 70-10, December, College of Engineering, University

of California, Berkeley.

6. Iwasaki, T., Tasuoka, F. , and Horiuchi, S. (1974), "A Measurement Method of Dynamic

Deformational Characteristics Sand Samples," Proc. of the 1st Annual Meeting of Kanto

Branch of JSCE, May. (in Japanese)

.

7. Iwasaki, T. , Tatusoka, F. and Horiuchi, S. (1974), "Dynamic Deformational Characteristics of Soils with Resonant Column Apparatus," Proc. of 9th Annual Meeting of JSSF,

June (in Japanese)

.

III-78
t{Q) pressure formation gauge

TT < transducer /^^^

loading piston i f ((/)) (C/)\

axial force | A |—"jampj

axial deformation

| g |—

»

jamp.|

pore pressure I C 1—Jamp.l r

photocorder

^scillator

j " j D | driver

oscilloscope

driving force

I velocity

air compressure

vacuum pump

I

Fig. 1 Schematic diagram of the apparatus

I'ig. 2 Photograph of the apparatus

driving power

Msindit

Fig. 4 niodel for the specimen- apparatus system

intee]—jiintegl

, jatorllratorj l deformation

indicator

acceleration indicator

frequency counter

axial force

a2=KoCri

Fig. 3 Stress condition of sample

III -79
Fig. 5 Struss-strain curve of sands

I 271 W

tare' &

/ \\ ^

/

/wt \ ,

r= a sin cut

Pig 6 Complex modulus

p = -3-107 + 203)

Toyoura-sand (Saturated)

0\ /or =10

P = 4 066 k9/cm' e= 0 797

p = 2 064kq/cni2 e = oeo6

^« e=oeii '^^^ P = 0564*9/cmi ^e=0820

2 5 IQ-S 2 5 10-' 2 5

shear strain amplitude 7

Fig. 8 Shear modulus versus shear strain amplitude

Fig 7 StresS'SIrain curve of Voigt maLenals

2.000

.000

500

200

100

Goc p

Toyoura-sand (Saturated)

e = 0.81

a"3/cri = io

05 I 2 5 2, P ( kg /cm )

Fig. 9 Shear modulus and mean principal stress

Seed ond Idriss (19701

I Toyoura-sand (d ry)jc = 0 66

Oi/0"3 =10

0"' 2 i 10'* 2 5

10"'

shear strain amplitude 7

' Fig. 10 Shear modulus versus shear strain amplitude

III -80
S<ad and Idnss 1 19701

-P=6 kg/cm'

-I mirf

I Toyoura-sand (dry)[

' e = 0.65

e = 0 75

0;/0i=2.0 a e = 085

10"* 2 5 C' 2 5 10* 2 5 Shear strain amplitude 7

Fig. 11. Shear modulus versus shear strain amplitude

1 kg /erf

S«ed and ldfiss(l970l^^\^

1 Sengenyama-sand (dry)| „ ^ e = 0 69

Oi/aj =1.0 • a e = 0 85

10"* 2 5 lO"' 2 5 10"' 2 5 I shear strain amplitude 7

Fig. 12" Shear modulus versus shear strain amplitude

(P = 05-6kg/tnf

Toyoura-sand and ^^'^^»iA^\\

Sengenyama-sand /

N

Seed and Idriss (1970)

—/ \\\\

V 2 5 lO-' 2 5

10'' 2

shear strain amplitude 7

FIl 13 Shear modulus versus shear strain amplitude

Sengenyama-sand (dry)

-S 200

100

e-0.85

cr3/Oi" = i.o

2 5 10 mean principal stress p (kg/cm^

)

Fig. 15 Shear modulus versus confining pressure

200

100

Sengenyama-sand (dry)

I

e = 0.69

(Ji/or=i.o

12 5 10 mean principal stress p (kg/cm^

)

Fig. 14 Shear modulus versus confining pressure

M 2D0Ovoid ratio o 85 (loose)

stress ratio k o 5

Sengenyama-sand (dry)

02 05 10 20 50 100 mean principal stress p

(kg/ cm'

)

Fig. 16 Shear modulus versus mean principal stress

III-81
void ratio e = 0B5 (dense) mean principal stress p = — (o, + 2a,)

3

P = 60 kj/cm"

-_:^-^-.ZZZTZzr.:^ } " = 20 kg/cm" -

"ZZ.":^ }

P = I 0 kg /CI

r:

10

'

lO"*

toyoura-sand (dry)

stress ratio K =

Fig. 17 Shear modulus versus stress ratio

void ratio e ^ 0 65 (loose)) mean principal stress P = -y lOi <• 2(jj I

'

> ''60 kg/cm'

Toyoura-sand (dry)

r=io''

^=10"*

/• = 5 1 10'*

10

Fig. 18 Shear modulus versus stress ratio

1500 -o — o

. Sengenyama- Sa Cdry)

Hard in and Black (W66; P = Skg/caf shear failure

0.0 OA 12 16 2.0

Fig. 19 Shear modulus versus stress ratio in case of p =constanl = 3^g /(J

shutting off the

driving power

amplitude

Sengenyama-sand (dry)

e = 0 854

/\/V\AAA'

P = 2 0 kg/on*

(Tj/OT =10

ir = l4 kIO-* At=0 170

Fig. 21 Amplitude-time decay curves

5000

- 2000

o 1000

Z 500

200

100

^ 03/01^ = 1.0 Toyoura-sand (saturated)

o oyo?=i.o I _ , ? Toyoura-sand (dry)- ^ 0"3/01=0.5j

• 03/01 = 1.0 Sengenyama-sand (dry)

050 0 60 070 080 090 1.00

void ratio e

Fig. 20 Shear modulus versus void ratio

10 -l 1 1 1 Sengenyama-sand (dry) 6=0.854

P =2.0 kg/fcm« 1- 1.4x10"*

At =4- in 44^ = 0.170

2.90

_L12345_1_ -L.

cycle number

Fig. 22 Amplitude versus cycle number

III-82
o e =0.85

• e =0 69

S:ngenyama-sand (dry)

0.5 10 2.0 5.0 100

mean principal stress p (kg/cm'

)

Fig. 23 (a) logarithmic decrement versus mean

principal stress

050

020

i 0,10

I 0.05

«

o

0 02

0.0

P = I.O

p

- 2 0

kg/cm*

A o e =085

A . e =0.69

Sengenyama-sand (dry) 03/0?= 10

-I 1 1 I 10"^ 10-" 5x 10"

shear strain amplitude 7

Fig. 23 (b) Logarithmic decrement versus shear starin amplitude

0.50

B

0.02

° e = 074

A e = 0.81 X'lO"* Toyoura-sand

(saturated) (Jt/U, =10

05 20

002

00

1

e=Q74

Toyoura-sand

(saturated) ov/oiMO

5.0 10.0 ^lio^' siw' I0-* 2nlO"^ StKT* mean principal stress ip (kg/cm^ ) shear strain amplitude 7

Fig. 24 (a) Fig. 24 (b)

Logarithmic decrement Logarithmic decrement

versus mean principal stress versus shear strain amplitude

III-83
PREDICTION OF MAXIMUM EARTHQUAKE INTENSITIES

FOR THE SAN FRANCISCO BAY REGION

by

Roger D. Borcherdt

and

James F. Gibbs

Office of Earthquake Studies

U.S. Geological Survey

The intensity data for the California earthquake of April 18, 1905, are strongly dependent on distance from the zone of surface faulting and the geological character of the ground. Considering only those sites (approximately

one square city block in size) for which there is good evidence for the degree

of ascribed intensity, the empirical relation derived between 1906 intensities

and distance for 761 sites underlain by rocks of the Franciscan Formation is

Intensity = 2.30 - 1.90 log (Distance).

For sites on other geologic units intensity increments, derived with respect

to this empirical relation, correlate strongly with the Average Horizontal Spec

tral Amplifications (AHSA) determined from 99 three-component recordings of

ground motion generated by nuclear explosions in Nevada. The resulting empirical relation is

Intensity Increment = 0.27 + 2.70 log (AHSA).

Resulting average intensity increments for various geologic units are -0.29 for

granite, 0.19 for Franciscan Formation, 0.64 for other pre-Tertiary , Tertiary

bedrock, 0.82 for Santa Clara Formation, 1.34 for Older Bay sediments, 2.43 for

Younger Bay mud. These empirical relations have been used to delineate areas in the San Francisco Bay region of potentially high intensity from future earth

quakes on either the San Andreas fault or the Hayward fault.

Key Words: Earthquake; empirical relation; Franciscan Formation; geological

character; ground shaking; intensity.

III-84
INTRODUCTION

The amounts of damage resulting from the great California earthquake of April 18, 1906,

varied greatly for different areas in the San Francisco Bay region. In some areas the

damage was WEAK with "occasional fall of chimneys and damage to plaster, partitions, plumbing, and the like," in other nearby areas the damage was VIOLENT with "... fairly general

collapse of brick and frame structures when not unusually strong..." (Wood, 1908). These

large variations were due in part to the geological character of the ground and to the

distance from the zone of surface faulting (compare the intensity map for the city of San

Francisco (Figure 1) with the geologic map (Figure 2) )

.

Comparative ground motion measurements made for 99 sites in the San Francisco Bay

region (Gibbs and Borcherdt, 1974) show that a strong correlation exists between amplitude

levels of observed ground shaking and the type of geologic unit. The purposes of this

paper are

:

1) To quantify the relationship between distance and the observed 1906 earthquake

intensities for a particular bedrock unit, and

2) To show the existence of a relationship between intensity and the measured ground

motion artpli fi cations .

These relationships permit a quantitative estimate of the dependence of the observed 1906

intensities on the geological character of the ground. Such estimates are useful for delineating areas of potentially high intensity from future earthquakes on either the San

Andreas fault or the Hayward fault.

INTENSITY VS. DISTANCE

The 1906 earthquake intensities ascribed sites on the same geologic unit generally

decrease with increasing distance from the zone of surface faulting (Lawson, 1908) . To quantify this apparent relationship, the 1906 intensity data for the San Francisco Bay

region were reconsidered on a site-by-site basis. The intensity data from only those sites

(approximately one square city block in size) for which there was good evidence for the degree

of ascribed intensity were considered. For each site underlain by rocks of the Franciscan

Formation, the distance to the zone of surface faulting was measured and plotted as a function of the ascribed 1906 earthquake intensity (Figure 3) . The resulting empirical relation.

Intensity = 2.30 + 1.90 log (Distance),

determined by the method of least squares, suggests that the ascribed intensity values for

sites on the Franciscan Formation generally decrease as the logarithm of increasing distance.

The empirical relation shows that the intensity values decrease very rapidly with distance,

with sites 3 km from the fault having observed intensities more than two intensity units

smaller than those at the fault.

The sites with the highest ascribed intensities ("A", 1906 San Francisco scale) are located within 0.7 km of the center of the zone of surface rupture. For most of these

sites, the unit of intensity was assigned on the basis of evidence for some form of ground

failure most of which was associated with surface faulting. The degree of intensity as- signed to most of the other sites at greater distances from the fault was based on damage

resulting from ground shaking or ground shaking induced ground failures. To quantify the

dependence of the intensities due only to shaking on distance, another empirical relation

was determined with the intensity data near the fault omitted. The resulting empirical

relation is essentially the same as the one determined from the complete data set. (Intensities predicted by either relation differ by less than 0.02.) This similarity suggests

that the dependence of intensity on distance is not influenced by the intensity data near

the fault due to surface faulting. For explicitness only the relation determined from the

complete data set will be referred to hereafter.

III-85
INTENSITY VS. MEASURED LOW-STRAIN AMPLIFICATIONS

Recordings of three con^jonents of around motion generated by distant nuclear explosions have been made at 99 sites in the San Francisco Bay region (Figure 2) (Borcherdt,

1970; Gibbs and Borcherdt, 1974) . Spectral ait^slification curves computed from these re- cordings show that low-strain ground motions of certain frequencies are an^lified considerably by thick sections of unconsolidated alluvial deposits. Averages of the spectral

amplification curves over the frequency band for which there is a good signal to noise

ratio correlate strongly with the t\'pe of geologic deposit.

To coirpare the low-strain amplifications with the obser\-ed 1906 intensity data, intensity increments \vere defined for each of the recording sites as the difference between

the obser\'ed intensity at tlie site and that predicted by the empirical relation for sites

at the same distance on the Franciscan Formation. These intensity increments are plotted

as a function of the corresponding Average Horizontal Spectral Airrolification (AHSA) values

(Figure 4) . (The AHSA values plotted have been normalized by the average AHS.*^ value de- - termined for sites on tl:e Franciscan Formation.) Earoirical relations were determined using

the method of least squares from the con^plete data set (".") and from only the data (" ") for sites in the city of San Francisco for which there was "unequivocal evidence" for the

degree of ascribed 1906 intensity. The t^vo empirical relations are similar with intensity

increments predicted by either relation differing by less than t^'.'o-tenths . The eii^irical

relation (5 1 = 0.27 + 2.70 log (AHSA)) based on only the good intensity data in the city

of San Francisco is preferred.

The correlation coefficient of 0.95 computed for the preferred enpirical relation

(i?I = .27 + 2.70 log (AHSA)) shows that a strong correlation exists beto^een the computed

intensity increments and the amplifications obser\"ed at low-strain levels. The physical

meaning of this empirical correlation is conK>lex and does not necessarily mean that amplifications observed at low-strain levels can be extrapolated directly to high-strain levels.

However, there are t\v-o possible reasons for this correlation:

1) For levels of ground shaking that did not cause ground failure, the higher amplifications indicate those sites that experienced the higher levels of ground shaking, and

2^ For levels of ground shaking that did induce ground failure, the higher amplifications indicate those sites that were most susceptible to ground failure.

In either case, the sites of higher amplification would have experienced greater amounts

of damage and be assigned higher degrees of intensity.

PREDICTION OF >LAXIMUM EARTHQUAKE INTENSITIES

Fault studies (Wesson et al., 1974) indicate a high potential for large (magnitude,

7.6-8.3) earthquakes on both the San Andreas fault and the Haward fault in the San Francisco Bay region. Historically, large earthquakes have occurred along both faults. Due

to a lack of cultural development at the times of the earthquakes, the intensity data for

an earthquake on the Ha\^v-ard fault is very scanty' and there are several presently developed

areas ^vith no 1906 intensity- data. However, the empirical relation between the 1906 intensity increments and the AHSA \'3.1ues measured from, the nuclear data provide a means for

Using the eirpirical relation bet^ceen intensity increments and the AHSA values, intensity increments were predicted for each of the 99 sites at which amplification values have

been measured from the nuclear ejclosions. The predicted intensity increm.ents were grouped

according to the tN-pe of underlying geologic unit (see Borcherdt, 1970, for a description

of the units) . The means and standard de^riations for the various samples are tabulated

(Table 1). The miean increm.ents for the various units range from -0.29 for granite to 2.43

for Vounger Bay mud. These mean intensity increments provide a quantitative estimate of

the dependence of the 1906 earthquake intensities on the geological character of the

ground

.
utilizing the computed average intensity increments for the various geologic units,

the empirical relation between intensity and distance, and a geologic map (compiled by K.R. Lajoie, personal communitcation) , Borcherdt and Gibbs (1974) predicted absolute intensities

on a regional basis for the San Francisco Bay region. The resulting predicted intensity

map (Borcherdt and Gibbs, 1974) shows the maximum intensity predicted for a site that might

result from an earthquake in the San Francisco Bay region on either the San Andreas fault

or the Hayward fault. Such a map is useful for delineating general earthquake problem

areas in the SSn Francisco Bay region and for evaluating the earthquake hazard in areas not

developed at the time of the 1906 earthquake. In addition, the map is useful for evaluating the hazard due to another large earthquake on the Hayward fault. The map shows that

earthquake hazards are not uniformly distributed throughout the San Francisco Bay region.

REFERENCES CITED

1. Borcherdt, R.D. (1970) . Effects of Local Geology on Ground Motion Near San Francisco

Bay, Bull. Seism. Soc. Am. , 60 , 29-61.

2. Borcherdt, R.D. , and J.F. Gibbs (1974). Maximum Predicted Earthquake Intensity Map

for the San Francisco Bay Region, in press.

3. Gibbs, J.F., and R.D. Borcherdt (1974). Effects of Local Geology on Ground Motion

Near San Francisco Bay—A Continuation, in press.

4. Lawson, A.D. (Chairman) (1908) . The California Earthquake of April 18, 1906, Report

of the State Earthquake Commission , Carnegie Institute of Washington, Publ, 87, Washington , D . C

.

5. Wesson, R.L., CM. Wentworth, R.D. Brown, Jr., E.J. Helley, and K.R. Lajoie. A Pre- liminary Analysis of Seism.ic Zonation in the San Francisco Bay Region- -Chap. I Faults

and Future Earthquakes, U.S. Geological Survey Circular, in press.

6. Wood, H.O. (1908) . Distribution of Apparent Intensity in San Francisco in The California Earthquake of April 18, 1906 , Report of the State Earthquake Investigation Commission , Carnegie Institute of Washington, Publ. 87, Washington, D.C.

III-87
TABLE 1

STATISTICS FOR SAMPLES OF LOW-STRAIN AMPLIFICATIONS AND INTENSITY INCREMENTS

WITH RESPECT TO FRANCISCAN FORMATION FOR VARIOUS GEOLOGIC UNITS

Geologic Unit

Average Horizontal

Spectral Amplification

Intensity Increment

(1906 San Francisco Scale)

Granite

Standard

Mean Deviation

0.63 0.11

Mean

-0.29

Standard

Deviation

0.21

Franciscan

Formation 1.00 0.38 0.19 0.47

Pre -Tertiary,

Tertiary

Bedrock*

1.42 0.45 0.64 0.34

Santa Clara

Formation 1.70 0.64 0.82 0.48

Older Bay

Sediments 2. 76 1.16 1. 34 0.58

Younger

Bay Mud 7.06 3.78 2.43 0.58

III-88
ILLUSTRATIONS

Figure 1. Map showing distribution of apparent 1906 intensities for the city of San Fran

Cisco, California (after Wood, 1908)

.

Figure 2. Map showing distribution of geologic units for the city of San Francisco, Cali

fomia (compiled by K.R. Lajoie from data of Schlocker and others).

Figiire 3. Observed 1906 intensities for sites (one square city block in size) underlain

by rocks of the Franciscan Formation versus perpendicular distance from the zone of surface rupture during the 1906 earthquake. For sites with "unequivocal evidence" in the city of San Francisco (Map No. 19, Lawson, 1908) , the

number of observed intensity values is shown below the corresponding distance

interval. For sites intersected by an "examined route" south of the city of

San Francisco (Maps No. 21 and 22, Lawson, 1908), the number of observed inten

sities is shown above the correponding distance interval. The observed 1906

intensities are expressed in terms of the 1906 San Francisco intensity scale

with the letters A-E corresponding respectively to the numbers 4-0 (see Appendix 1 for detailed description of intensity scale)

.

Figure 4 Increments in 1906 intensities versus average spectral amplification computed

at corresponding sites from recordings of nuclear explosions. Both the intensity increment values and the average spectral amplification values were com- puted with respect to the corresponding average value determined for sites

underlain by rocks of the Franciscan formation.

III-89
Figure 1

!

III-90
ALLUVIUM + OTHER SEDIMENT { 0 - 100 FEET THICK )

BEDROCK 0 3 KM

Figure 2

III-91

II
<^i- oj — O

{31V3S OOSIDK'Vyj NVS 9061)

NOIlVildOJ NVOSIONVdJ NO A1ISN31NI

III-92
-2 I I I I 1 I I I—! I I I 1 I '

I

I 0.2 0.5 I 5 10 20 30

AVERAGE HORIZONTAL SPECTRAL AMPLIFICATION (AHSA)

Figure 4

T 11-93
APPENDIX 1 SAN FRANCISCO APPARENT INTENSITY SCALE

The following grades of apparent intensity were ascribed by H.O. Wood (1908, pp, 224-

25) in the city of San Francisco to describe damage which resulted from the California

arthquake of April 18, 1906.

Grade A. Very Violent - Comprises the rending and shearing of rock masses, earth,

turf, and all structures along the line of faulting; the fall of rock from

mountain sides; numerous landslips of great magnitude; consistent, deep,

and extended fissuring in natural earth; some structures totally destroyed.

Grade B. Violent - Comprises fairly general collapse of brick and frame buildings

when not unusually strong; serious cracking of brick work and masonry in

excellent structures; the formation of fissures, step faults, sharp com- pression anticlines, and broad, wave-like folds in paved and asphalt-coated'

streets, accompanied by the ragged fissuring of asphalt; the destruction of

foundation walls and underpinning structures by the undulation of the

ground; the breaking of sewers and water-mains; the lateral displacement of streets; and the compression, distension, and lateral waving or displacement of well-ballasted streetcar tracks.

Grade C.

Grade D.

Grade E.

Very Strong - Comprises brick work and masonry badly cracked, with occasional

collapse; some brick and masonry gables thrown down; frame buildings lurched

or listed on fair or weak underpinning structures, with occasional falling

from underpinning or collapse; general destruction of chimneys and of masonry, brick or cement veneers; considerable cracking or crushing of foundation

walls

.

Strong - Comprises general but not universal fall of chimneys; cracks in

masonry and brick work; cracks in foundation walls, retaining walls, and

curbing; a few isolated cases of lurching or listing of frame buildings

built upon weak underpinning structures.

Weak - Comprises occasional fall of chimneys and damage to plaster, partitions, plumbing, and the like.

I

III-94
A STATISTICAL APPROACH TO LOADING AND FAILURE OF STRUCTURES

by

Ronald G. Merritt

U.S. Army Construction Engineering Research Laboratory

A fundamental problem of structural engineering is the examination and

selection of loading criteria. It is imperative that any solution to the problem center around a rationale that relates information available on loading to selected criteria. Such available information is generally in the form of data.

It is the purpose of this brief paper to abstract the problem and outline preliminary work on a rationale for addressing the problem.

The paper begins by defining the general nature of the problem. Solution

to the problem is related to consideration of available information in the form

of data. The next three sections of the paper discuss the initial stages of a rationale for consistent examination and selection of loading criteria. The

first of the sections examines available information on structural load and the

second examines available information on instances of structural failure. Classes

of statistical methods are discussed in the third section. This section also includes discussion of a proposed method for assessing the overall information con- tent of the available data. Finally, several illustrative examples of application

of statistical methods to load and failure data are presented and the paper con- cludes with a discussion of future extension to this preliminary work.

Key Words: Failure; probability theory; random process; safety; statistical

analysis; structural engineering.

IV-1
GENERAL NATURE OF THE PROBLEM

Examination and selection of loading criteria involves the consideration of a state- ment S with quantifiers that relate variables useful in description of the load. This may

be written generally as

(1) S:C = C(t, r, a)

for

S = Loading criteria statement

C = Load function

t = Time vector

r = Space vector

a = Parameter vector

The statement S has a quantifier the load function C which is expressible in terms of time

space and a finite set of parameters. The expression is general enough to allow for sev- eral components of time, space, and parameters, as denoted by the vector notation. The

problem may now be stated in terms of examination of the validity of S.

Validity of S is usually established through some subjective and objective evaluation

of available information related to S. In order to be consistent in this evaluation of

information, a rationale for carrying out this evaluation must be set forth. The preliminary outline of the rationale proposed in this paper centers upon a means of assessing

available information related to S by use of statistical techniques, correlating this information and obtaining quantitative factors upon which the validity or invalidity of S may be established. In a real sense this rationale, in part, already exists in that sta- tistical interpretation of collected data is commonplace in examination of load data. The

discussion to follow extends this rationale. However, it is important to note here that

"all" available information is to be examined in evaluating the validity or invalidity of

S. This includes consideration of load information for one. But in addition, since the

invalidity of S tacitly implies possibility of structural failure because of load, one must

also consider structural failure information. It is the general nature of these sets of

information that provides the basis for this preliminary work on development of the ra- tionale .

LOAD INFORMATION

Load information is obtained in a quantity termed a datum. Such datum may be in a raw form or in a summary form. The raw form consists of the most basic unit and results

from direct quantization of the phenomenon under observation. The summary form results

from a transformation of raw data.

Upon collection of data on loading, e.g., wind loading, it became apparent that some way of classifying individual pieces of datum would need to be developed. Once classified,

then groups of datum within any one designated category could be examined for consistency

and their relationship to the proposed loading criteria. The discussion to follow defines

the datum classification system and the example in Section VI illustrates application of

a statistical technique to a piece of datum within the system.

The requirements on a load data classification system are very basic. First, a single

piece of datum must be recognized as such in the system and second, a piece of datum must

be classifiable within the system. In order to facilitate this a "generalized random process", L, is defined whose "sample functions" consist of pieces of datum described by a

IV2
set of parameters related to the load phenomenon (1) . This is most easily expressed as

L(t, r, a) = {l{t, r, a) : t£T, reR, a eA

}

where

L(t, r, a) = Generalized random process

l(t, r, a) = A piece of datum

t = Time related description of the datum

T = Time indexing set

r = Space related description of the datum

R = Space indexing set

a = Parameter related description of the datum

A = Parameter indexing set.

It is assumed that every piece of datum related to a load phenomenon belongs to L{t, r, a)

ajid that^ each piece of datum is uniquely defined through an ordered triple of vectors

(t, r, a).

The advantages of such a means of classifying data by evaluation of t, r, and a are readily apparent. First, in evaluation of t, r, and a dattam sets are established within

L(t, r, a) that relate similar information about the load_phenomenori under investigation.

Second, correlation of information through evaluation of t, r, and a allows one to assess

the overall information content of the available data. Third, ready evaluation of data

within a given datum set is possible and links amongst datum sets provide a key to links

amongst data within different datum sets. Finally, this approach lends itself well to either the synthesis approach or the analytical approach to criteria selection. In the

synthesis approach, all data is structured into a description of the load phenomenon and

criteria are selected from this description. In the analytical approach, the datum is

checked against the proposed criteria for consistency and selection of criteria is based

upon this check. In either case, the pertinent datum is easily identified.

A total of twenty-two parameters that must be evaluated for each piece of datum were

selected. The twenty-two parameters may be divided into five groups. Brief mention of

these five groups will suffice for the present discussion. The first group consisting of

two parameters uniquely identifies the piece of datum. The second group consisting of

five parameters identifies the datum by defining the overall load phenomenon properties,

e.g. static, deterministic, stationarity , its source, its spatial extent. The third group

consisting of eight parameters describes the piece of datum in terms of the time history

information available. The fourth group consisting of five pcirameters describes the datum

in terms of the spectral information available. Finally, the fifth group of parameters

consisting of two parameters gives a brief narrative description of the piece of datum

along with a source reference.

This general scheme of datum referencing permits a consistent examination of structural load data. Extensions to this general scheme will be discussed in the last section

of this paper.

*Nuinbers in parentheses denote references.

IV-3
STRUCTURAL FAILURE INFORMATION

The considerati6n of data on instances of structural failure in the preliminary i stages of development has provided for some most interesting ideas on structuring of data

from diverse and complex phenomenon. In the case of structural load data, the content of

individual pieces of datum was described in terms of a set of parameters and the raw or summary form data lay within each of the random process sample functionSj_ Data analysis

was assumed to take place on a "level below" the datum structures, L(t, r, a) . For the case of structural failure, the nature of the available data and information desired from

the data requires that the description of the datum, i.e., instance of structural failure,

be complete enough for data analysis. That is, the analogous structural failure "generalized random process" should contain the available information concerning the structural

failure. This approach to structural failure datum is the product of several considerations. First, quantitative structural failure data is difficult to obtain since few instances of structural failure are instrumented. Second, unless the failure is controlled

in some manner, quantitative data tends to be meaningless because of the complex loadresponse path that usually describes the failure. Third, any one case of structural failure is but one of many possible structural failures and it may or may not share properties j in common with other cases of structural failure. Fourth, detailed quantitative data from

instr\imentation of a structural failure would present a prohibitively high collection and

reduction cost to information ratio. Finally, detailed reduction of quantitative data ob- ' tained during or after structural failure would tend to de-emphasize the overall characteristics of the structural failure. Structural failure data was considered in the following way.

It was hypothesized that structural failure may be considered a "generalized random

process" (1). Thus, it can be represented by an expression

S(t, r, a) = {s(t, r, a); tcT, reR, a^A)

Structural failure generalized random process

Structural failure sample function

Time vector -i- •

Time indexing set •. •- ..•

Spatial vector

Spatial indexing set ' •

Parameter vector

Parameter indexing set. '

• .

All instances of structural failure belong to S(t, r, a) and every failure is in S (t, r, a) either explicitly through collected data and parameter evaluation or implicitly in cases where the structural failure is unrecorded but the indexing sets are broad enough for description. The problem of structural_failur£ data structuring now becomes a matter of defining T, R, and A and evaluating t, r, and a from collected data on structural failure.

Ultimately, forty-five parameters were considered adequate to define the structural

failure random process, i.e., forty-five parameters were considered sufficient to describe

any instance of structural failure. Obviously, only the overall gross characteristics of

an instance of structural failure were considered appropriate for description and most

pertinent to the overall rationale.

The forty-five parameters fall into nine major categories. For the sake of brevity,

these nine major categories v;ill be listed with a few comments regarding the parameters

within each category.

S(t, r, a) =

s(t, r, a) =

t =

T =

r =

R =

a =

A =

IV-4
1. Identification - This category includes information on the source and information

content of the structural failure data available.

2. Structure Characteristic Information - This category includes all information re- lated to the structure that experienced the failure. The dates of construction

and failure are recorded along with general structural, material, and functional

characteristics of the structure. The geometrical dimensions of the structure

along with those of the failed portion of the structure are also recorded.

3. General Failure Description - This category describes the cause of the failure,

the extent of the failure both in qualitative and quantitative terms, the nature

of the failure in terms of progressive or nonprogressive characteristics, horizontal or vertical characteristics, the total time of the failure, and the stages

of failure.

4. Global Failure Description - For failures in which a major portion of the overall

structure has failed, the failure takes on a global nature. This is subsequently

described by three parameters naming elements of the structures that failed, modes

of failure, and material composing the failed elements of the structure.

5. Local Failure Description - A failure of a structure may include a small portion

of the overall structure in which case the failure takes on a local nature. The

same three parameters as for the global failure description provide for the local

failure description.

6. Global Load Description - Loading on a structure that is over a large portion of

the structure may be termed a global load. It is described in terms of four

parameters including identification,- general dimensions, a general statement, and

estimated value if this is available or able to be deduced.

7. Local Load Description - Loading on a structure that is over a small portion of

the structure may be termed a local load. The same four parameters as in the case of global load description describe the local load.

8. Load-Failure Relationship - In most instances of structural failure, there exist

a general spatial relationship between load and failure. This relationship may

be expressed in terms of local load-local failure, global load-local failure,

local load-global failure, global load-global failure. This parameter provides

insight into the nature of the extent of the loading and the corresponding fail- ure .

9. General Statement - This final parameter group consisting of one parameter is a general statement about the failure and its cause.

Here again it is well to remind the reader that structural failure does not relate well to phenomenological description because of its complexity. The categories of parameters and

the parameters themselves provide for an overall view of the_structural failure process.

Given_da_ta on structural failure, the parameters of S(t, r, a) can be evaluated and

S(t, r, a) better defined. The statistical t^echnig^ues to be discussed in the next section

are applied directly to the parameters of S{t, r, a).

Extensions of this scheme for structuring failure data will be discussed in the last

section of this paper.

BASIC CONSIDERATIONS FOR STATISTICAL ANALYSIS

The nature of the problem under consideration and breadth of the field of statistics

make it possible to consider only a few topics in relating statistical methods to the

datum within the framework of the load and failure generalized random processes discussed

above

.

IV-5
One of the first considerations in applying statistical methods to data of the processes above is an examination of the way in which data is measured. There exist four ac- ceptable statistical data measures by which the measure of data is defined (2) (3) . Listed

in order from least to most powerful they are as follows: nominal, ordinal, interval, and

ratio. A brief description of each is in order. The nominal measure applied to data implies the data may be categorized according to a set of mutually exclusive conditions.

The ordinal data measure applied to data implies there exists an order relationship amongst

pieces of the datum. The interval data measure applied to data implies a relationship of

the foirm x-y> 0 , x-y = 0 , orx-y< 0

exists between any two pieces of datum. The data is in some way commensurable. Finally,

the ratio data measure applied to data implies numerical relationships for the datum are available and for y 7^ 0, x/y is a meaningful expression between any two pieces of datum.

The data is numerically commensurable.

Although there are a number of ways of dividing statistical methods into categories

for purposes of this discussion perhaps the categories distribution and distribution free

will suffice. Distribution related statistical methods, in general, correlate with instances in which finite parameter distribution functions may be utilized in the statistical

analysis of the data. Distribution free related statistical methods, in general, correlate

with instances in which lesser restrictions are imposed upon conditions that must be sat- isfied for application of the method to a given set of data. These statistical methods

may be further subdivided into methods concerned with point estimates of parameters, con- fidence regions for parameters, or significance tests for parameters.

In the illustrative examples to follow, distribution free statistical methods are applied to both load data and failure data. In general, distribution related methods apply

well to load data because of its tendency to be describable in terms of the ratio data measure and distribution free related statistical methods apply well to structural failure

data because of its tendency to be describable in terms of data measures less powerful than

the ratio measure

.

In work to date emphasis has been placed on consideration of structural failure data.

It has become important to consider categorical distribution free statistical techniques

for use on parameters of the structural failure random process. Categorical techniques

are most applicable because structural failure data is for the most part of a categorical

nature. Distribution free techniques are most applicable because of the difficulty in

determining the distributions and their related parameters because of lack of large amounts

of data.

It is found useful when considering the structural failure generalized random process | to construct a statistical method-process parameter matrix whereby statistical methods ap- plicable to given process parameters are correlated one to another. Table 1 below provides

a segment of this matrix.

The construction of the matrix in Table 1 leads naturally to an assessment of the

overall information content of a set of data based upon an evaluation of factors useful in

defining tlie overall characteristics of a statistical method (4) . Table 2 lists factors

useful in evaluating the effectiveness of a statistical method along with proposed weights

for these factors. The overall information content of a set of data is determined by as- sociating a set of statistical techniques with the data and proceeding to tabulate weight

values for the various factors. A relative measure of information content amongst sets of data is obtained.

There exist several major weaknesses in the approach. First, not all statistical

methods may be accurately evaluated in terms of these factors. Second, it presumes that

one has selected an optimal set of statistical methods to operate on a given set of sta- tistical data. Third, it presumes that data information content is related to abstract measures on the statistical method independent of the data. Finally, it assumes the

IV-6
weighting factors are accurate and constant over the ranges of statistical methods. Even

though these weaknesses exist, a matrix relating statistical method versus weighting factor

provides a crude measure of the relative information content of a set of data to which the

statistical rrethod may be applied.

ILLUSTRATIVE EXAMPLES

The examples in this section of the paper are illustrative in the sense that (a) they

are not based upon all the data that is available and (b) they present a rather new approach

in the reduction of civil engineering data. The first point is a result of the preliminary

nature of this work and ability to reduce only a portion of the data available. The second

point refers to the use of distribution free statistical techniques on categorical data.

In general, measurement distribution oriented statistical techniques are used on numerical

data result ing from a well controlled experiment. The results of the statistical analysis

are then presented in some concise form. Categorical distribution free statistical techniques require data to be categorized and are often times related to a statistical hypothesis test which may or may not be related to a parameter describing the data, e.g., trend

or randomness of data may be under investigation.

It is also well to point out that the conclusions drawn from the illustrative example

may seem trivial, however, each example conclusion presents only a minute piece of information extending that which is already known about the case under investigation. That is to say, the effectiveness in use of techniques in this way comes by way of construction

of an overall view of the case by means of statistics. This implies application of many

statistical techniques in many different ways to the data available. Fortunately, once

a data base has been constructed and the statistical techniques selected, this becomes a rather simple and automatic procedure.

Example 1. The first illustrative example concerns correlation of external wind

pressure coefficient with structure configuration. Wind tunnel data given in Table

3 (5) relates the external wind pressure coefficient on four sides of a structure

(from wind blowing at two different angles of incidence to a reference face of the structure. Figure 1) to the relative dimensions of the structure. A measure of the

structure configuration is taken as the "relative volume" of the structure defined by

V = hbL/(max{h, b, l})^.

Utilizing a variation of Spearman's Rho test for a measure of correlation between V and the external pressure coefficient, the results of the rank correlation coefficient

computation are shown in Table 4.

A few observations that may be made on the basis of this single analysis are as follows

:

(1) Positive correlation, i.e., large V associated with large Cpe , on walls

denoted A, Ba = 0° and A', B', C' a = 45° and negative correlation, i.e.,

large V associated with small Cpe, on walls denoted C, D a = 0° and D' a =

45° ,

(2) A lower degree of correlation between V and Cpe when Cpe is negative,

(3) In the case of wall C, C, there is a change in sign of correlation coefficient with change in angle of incidence of wind,

(4) The correlation between V and Cpe is in no instance very strong.

Example 2. From some data available on failure of structures (6), the cause of failure of structures has been placed into one of sixteen categories, as presented in

Table 5. Table 5 also presents the empirical distribution of causes for some 88 cases

of failure. In order to establish a distribution fmiction for cause, attempts are

IV7
made to fit a uniform distribution curve and an arbitrary distribution curve to the

data of Table 5. The chi-square goodness of fit test is utilized in both instances.

A formal statement of the problem for the case of uniform distribution fitting

is as follows

:

"o F(x) = F^(x) All X

where

: F(x) ?^ F-'-(x) At least one x

F-*- (x) is a uniform distribution over all values of x.

Utilizing the chi-square goodness of fit test, is rejected at all meaningful levels

of significance. This is not surprising from an examination of the data. In the case of an empirical distribution fit to the data of Table 5 again, a formal statement of

the case is as follows:

: F(x) = F-'-(x) All X

H-j_ : F(x) f F-'"(x) At least one x

where

F"''(x) is an arbitrary distribution function over all values of x given in Table 6. F (x) is similar to the distribution function

for the normal distribution.

Utilizing the statistic

I

(0. - E.)2

3=1

and after combining expected frequencies, causes, and observations consistent with

the statistical method and computing T on the basis of data in Table 6, one finds

T = 5.6.

The arbitrary empirical distribution fits the data well and is accepted at the

1001 level. It should be noted that combining of failure of structure causes and

observations implies a transformation of the original data. Care must be exercised

in interpreting the empirical distribution curve fit to the data.

Example 3. The final illustrative example examines the relationship between load and

failure in terms of the qualitative terms local and global. A 2 x 2 continuing table

made up of 86 pieces of datum is shown in Table 7 with a key for identification of

the variables and their values. The hypothesis for McNemar test for significance of

change applied to the contingency table in Table 7 may be stated as follows:

"o • = 0) = P^^i = 0) All i

"l • = 0) ^ ^("^1 = 0) All 1

: P(X^ = 1) = P(Y^ = 1) All i

"l : PCXj^ = 1) 7^ P(y^ = 1) All i

IV-8
In a physical sense in the first set of hypotheses, states that the probability of a local load is equal to the probability of a local failure and H-j^ the negation of this for all observations of load and failure. In the second set of hypotheses H states that the probability of a global load is equal to the probability of

a global failure and H-j_ the negation of this for all observations of load and failure. Application of the McNemar test statistic rejects both hypotheses at the .001

level. It should be noted that in the McNemar test for significance, an inner con- sistency in the data must be assumed. This is difficult to verify for the data available.

The use of contingency tables for categorical data is an important key to a con- sistent approach to examination of load and failure data.

CONCLUSION

The above represents a very preliminary basis for a statistical examination of the

load and failure of structures and a rational approach to examining available information

related to loading criteria. The next stage in the development will consider construction

of a data base of available data along with establishing a broader group of statistical

techniques. This should lead to the consideration of mathematical pattern language in the

correlation of collected data and in the utilization of appropriate statistical techniques

on the collected data. In addition, it is anticipated that more advanced mathematical

techniques, e.g., in the area of combinatorial methods will be used for investigating general relationships amongst the diverse pieces of datum.

The ideas expressed above form a basis for a rationale for the examination and selection of load criteria. The rationale is based upon a consistent and thorough statistical

analysis of available load data aind failure data. Given the statement S representative

of a statement of load criteria, the validity of S is deduced from the consistent and

thorough statistical analysis of all available data. Work to date described above is a first step in the rationale development.

ACKNOWLEDGMENT

The author would like to express his appreciation to Dr. W.E. Fisher of the Structural

Mechanics Branch at CERL and CERL for their support during this study.

REFERENCES

1. Merritt, R.G., "A Statistical Approach to Loading and Failure of Structures," CERL

Technical Report to be published.

2. Conover, W.J., Practical Nonparametric Statistics , John Wiley and Sons, Inc., 1971.

3. Stevens, S.S., "On the Theory of Scales of Measurements," Science, 103, pp. 677-580.

4. Walsh, J.E., Handbook of Nonparametric Statistics , Vol. I, D, VanNostrand Company, Inc.,

1962.

5. "Wind Forces on Structures," Final Report, Task Committee on Wind Forces Committee on Loads and Stresses, ASCE Paper No. 3269, Transactions of ASCE, Vol. 126, Part II, 1961,

pp. 1124-1198.

6. McKaig, T.H., Building Failures , Case Studies on Construction and Design, McGraw-Hill,

1962.

IV-9
TABLE 1

STATISTICAL METHOD - FAILURE PROCESS PARAMETER MATRIX

\ SM

FP \^

1 2 3 4 5

6 X X X

7 X X X

8 X X X

9

Key:

Failure Parameter (FP) statistical Method (SM)

6 Descriptive Name

7 Construction Date

8 Failure Date

9 Structural Characteristics

1 Binominal Test

2 Chi-Square Test for Goodness of Fit

3 Wald-Wofowitz Runs Test

4 Quantile Test

IV- 10
TABLE 2

FACTORS FOR EVALUATING THE EFFECTIVENESS OF A STATISTICAL METHOD WITH WEIGHTS

Statistical Data Measure ' (10)

Nominal 2

. Ordinal 4 Interval 6 Ratio 10

SanjJle size 10

Data Transformation and Restrictions on Data Parameters 2

Level of Computational Effort 2

Extent of Use of Symmetry 2

Sensitivity of Procedure to Assumptions 4

Precision Level (10)

Exact 10

Theoretical Approximate 7 Judgment Empirical 4

Efficiency of Method 10

Consistency of Method 10

Sensitivity of Procedure to Assumptions and Difficulty

in Verifying Assumptions 10

Population Properties and Importance Amongst Other Data Groups 5

IV-11
TABLE 3

STRUCTURAL CONFIGURATION AND EXTERNAL PRESSURE COEFFICIENT Cpe

AT ANGLES OF INCIDENCE OF 0° AND 45°

h:b:L

1:1:1

2.5:2.5

2. :2.5

2.5:2.5

1:4:4

1:8:16

2:1:2

1:2.4:12

1:1:5

V

1.00000

0.20000

0. 20000

0.20000

0.25000

0.03125

2.5:1:1 p 0.16000

0.50000

0.10667

0.04000

a = 0'

B C

-.5 -.6

-.5 -.7

-.5 -.7

-.5 -.8

-.3

-.5

-.6

-.5

-.5

-.5

-.4

-.5

-.7

-.8

-.6

D

-.6

-.7

-.7

-.8

-.4

-.5

-.7

-.8

-.6

-.6

A'

.5

.6

.6

.6

.5

,5

.5

.6

,5

.5

a = 45°

B' C

-.5

-.5

-.5

-.5

-.4

-.5

-.5

-.5

-.6

.5

.4

.4

.4

.5

.4

.5

.4

.4

.4

D'

-.5

-.5

-.4

-.4

-.4

-.3

-.5

-.4

-.4

-.5

TABLE 4

RANK CORRELATION COEFFICIENT OF STRUCTURAL CONFIGURATION

AND EXTERNAL PRESSURE COEFFICIENT

Variables Correlation Coefficient

V-A .41

V-B . .32

V-C -.25

V-D -.25 .

V-A' .36 .

V-B' .61

V-C .42'

V-D' • -.20

IV-12
TABLE 5

FAILURE OF STRUCTURES CAUSE WITH EMPIRICAL DISTRIBUTION

Failure of Stmacture Cause 12345578

Observations

1 21 0 2 7 1 5 15 0 7 15

9 10 11 12 13 14 15 16

Key: Failure of Structure Cause

1 Unknown

2 Structural

3 Material

4 Functional

5 Workmanship

6 Structural and Material

7 Structural and Functional

8 Structural and Workmanship

9 Material and Functional

10 Material and Workmanship

11 Functional and Workmanship

12 Structural, Material and Workmanship

13 Structural, Functional and Workmanship

14 Material, Functional and Workmanship

15 Structural, Material and Workmanship

16 Structural, Material, Functional and

Workmanship

IV-13
TABLE 6

E^4PIRICAL DISTRIBUTION FUNCTION WITH DENSITY FUNCTION AND

EXPECTED FREQUENCY OF OCCURRENCE FOR N = 88 CASES

F^(x)

0.0005

0.0010

0.0100

0.0300

0.0800

0.1800

0. 3200

0.5000

0.6800

0.8200

0.9200

0.9700

0.9900

0.9990

0.9995

1.0000

f-^(x)

0.0005

0.0005

0.0090

0.0200

0.0500

0.1000

0.1400

0.1800

0.1800

0.1400

0.1000

0.0500

0.0200

0.0090

0.0005

0.0005

Ej = f-^(x)N

.004

.004

.792 7.040

1.760

4.400

8.800

12,320

15.840

15.850

12. 320

8.800

4.400

1.760

.792 7.040

.044

,044

Failure of

Structure Cause

3

9

1

4

7

5

10

8

2

11

15

13

14

6

12

16

Observations

0

0

1

2

5

7 .

7

15

21

15

7

5

2

1

0

0

TABLE 7 2x2 CONTINGENCY TABLE FOR LOAD-FAILURE RELATIONSHIP

Y = 0 Y = 1

4 33

4 45

Key: Load and Failure Type

Local Load X = 0 Global Load X

Local Failure Y = 0 Global Failure

IV- 14
T

h

1

FIGURE I STRUCTURAL CONFIGURATION FOR

WIND PRESSURE COEFFICIENTS
SYNTHETIC EXPERIMENTAL RESEARCH ON THE DUCTILITY OF SHORT REINFORCED

CONCRETE COLUMNS UNDER LARGE DEFLECTION

by

Kiyoshi Nakano

Head of the Third Division

Building Research Institute

Ministry of Construction, Japan

Masaya Hirosawa

Chief Researcher of the Third Division

Building Research Institute

Ministry of Construction, Japan

In general, short reinforced concrete columns will fail in a brittle manner.

In order to create and establish better ductility in such columns, a synthetic

research experimental program has been conducted. This program consisted of the

testing of 125 short column specimens, subjected to multi-cycles of flexure-shear

loadings. The result from these tests indicate the following:

1) The ductility of structural members is influenced by shear, bond, and

buckling of the compression bars.

2) To prevent buckling of compression reinforcement, under small curvature,

the spacing of the web reinforcement must be controlled.

3) In order to prevent a shear failure of structural members within reasonable ductility, an effective set of restrictions on the combination of

axial force, tensile reinforcement ratio, and shear span ratio are re- quired.

4) The bond failures which were observed in the test members, where deformed

bars were used as axial reinforcement, consisted of bond-splitting of the cover concrete. The conventional method, which uses bond strength as an index to verify bond failure, is not effective for the bond-splitting

failure mode. It is, therefore, necessary to restrict the tensile reinforcement ratio in order to prevent this type of bond failure.

Key Words: Column; ductility; earthquake; reinforced concrete; shear tests; structural engineering; web reinforcement.

IV- 16
INTRODUCTION

In a country such as Japan, where severe earthquakes occur often, the structural

safety of buildings is governed largely by earthquake design criteria.

Due to the results obtained from many strong motion earthquakes and relative to re- sponse analyses and earthquake damage observations, it has become gradually clear that the

actual influence of earthquakes on structures is more severe than the simulated earthquake

load which is designated in many countries as a design seismic force coefficient. In fact,

during the 1958 Tokachi-Oki earthquake, several buildings, which had yield shear coefficients greater than 0.5, were destroyed.

In general, it is not practical to construct all buildings strong enough so they can

resist severe shocks as controlled by their load-carrying capacities. However, it is possible to construct ductile buildings with appropriate load-carrying capacity and rigidity,

which can survive severe shock.

Reinforced concrete buildings constructed in Japan are designed with a seismic shear

coefficient of 0.2 or greater, however, the resulting ductility factors required to resist severe earthquake shocks does not seem excessively large. Therefore, a synthetic ex- perimental research project was initiated to determine how to make reinforced concrete

short columns ductile.

This research was sponsored by the Ministry of Construction and a committee was thus

organized to execute this project. This committee consists of members selected from various research organizations belonging to government, universities, and private firms.

In this report, the failure modes and the factors which affect the ductility are disdussed based on results of tests of 125 specimens.

OUTLINE OF EXPERIMENT

Objective and Master Plan of Experiment

The objective of this synthetic experimental research was to obtain quantitative design criteria to insure that reinforced concrete columns are ductile.

The mean unit axial compressive stresses in the first story^columns of reinforced concrete buildings in Japan due to permanent load is about 40 kg/cm . Assuming a value of

0.4 as the yield shear coefficient, the mean unit shearing stresses during a severe shock

are generally less than 20 kg/cm^. The quantity of web reinforcement required to meet the

above stress conditions will not cause construction difficulties during fabrication of the

columns.

However, as columns are the most important of the various structural members, it is

preferable to keep them ductile even under the maximum shear force expected during severe

earthquake shocks. The maximum shear force corresponds to the shear force that is present

v/hen the column yields at both ends.

Therefore, the combinations of the ratio of tensile reinforcement and shear span

ratio were first selected so that their maximum shear capacities did not become excessively

large under constant axial load, which corresponded to a value of 40 kg/cm^ . Next, the

reinforcing details such as quantity, spacing shape of web reinforcement, and axial bar

arrangement were examined experimentally in order to induce a ductile condition.

The standard shear span ratio that was adopted was intentially small and thus contradicts the characteristics relative to a ductile condition.

IV-17
Outline of Test Series and Specimens

A. Variable Factors

Based on the past experimental and research results, the committee selected ten variables as those factors which effect ductility. These factors are shown in Table

1 and are listed as Fl to FIO. The values and characteristics of some of the variables shown in Table 1 were determined from previous investigations which were con- ducted on various existing reinforced concrete buildings.

The selection of the standard size of cross-section, web reinforcement ratio,

loading hysteresis and loading apparatus was then determined, as will now be described

1) Scale Effect

Reliable data relative to the scale effects on concrete columns is scarce.

Also, due to limitation in budget and facilities, it is often impractical to conduct full-scale tests. Therefore, the size of the main test cross-section

that was selected was 25 square cm. Also, a series of 50 square cm sections was tested to investigate the scale effect.

2) Web Reinforcement Ratio .,!!

The method to be used for calculating a reasonable web reinforcement ratio,

in order to obtain ductile columns, has to date not been established. Therefore,

it was tentatively decided to use Arakawa's minimum equation to set the standard

web reinforcement ratio. This experimental equation determines the minimum re- quired shear strength of reinforced concrete members when not subjected to an axial force. This equation is used as a basis for the regulations of A.I.J,

code for shear reinforcement. That is, one of the standard values of the web

reinforcement ratio was tentatively set as Pw]_ which can be obtained by substituting the flexure capacity, c(^\j , equal to the shear force, given in Arakawa's

minimum formula. Half of the ^-^-^ value was also adopted as part of the standard,

3) Loading Excursion ' :|

Until a rational dynamic method has been established for estimating the

seismic properties of structural members, it is difficult to evalute these properties from the results of special vibration tests. However, some standard

static cyclic loading method must be adopted in the test program. Therefore, a multi-cycles alternate loading method, as shown in Figure 1, was adopted as a standard loading excursion. This loading pattern was selected for the following

reasons

:

a) The ductility factors determined from a response analyses of medium

height buildings during severe shocks is approximately three or four

providing their strength and rigidity is not excessively small.

b) The number of acceleration responses which correspond to approximately 80% of the maximum has been reported to be of the order of

ten.

c) The structural characteristics of members under alternate cyclic

| loading at constant deflection is usually determined within ten I cycles of the loading.

4) Loading System

In most past experiments, simple beam systems were used as the loading sys-

>

tern for the structural members. However, this method is not acceptable for the \. investigation of the seismic behavior of reinforced concrete columns because of

the following reasons: i IV-18

CJassi

B

totati

I
a) When discussing shear and bond problems in members, a restrained beam

type is better than simple beam type in simulating the real condition

of columns in actual buildings during an earthquake.

b) Both ends of the column sections should be kept parallel and without

inclination with respect to each other to simulate the actual conditions .

c) In discussing the characteristics and effects of large deformations,

it is preferable that the influence of additional moment due to ec- centric axial load be easily estimated.

d) It is preferable to iirpose many cycles of load reversals and that the

developing cracks be easily observed and recorded.

A new loading apparatus, as shown in Figure 2, was developed and was used

to test the series of columns. In order to discuss the influence of the loading systems, two test series were also conducted on continuous beams.

B. Typical Series and Specimen Details

On the basis of the previous discussions, it was decided that a typical series

should consist of a total of fifteen specimens. Each series was subdivided into

types, dependent on the variables to be investigated. For example, three specimens

were a function of P^, two specimens were dependent on o^, two specimens relative to

M/QD, and two a function of P^. A list of the specimens relative to a typical series

is shown in Table 2.

In the other series, each common factor, for example, C or size of section,

etc. , was systematically varied. Figure 3 shows an example of the details of the

specimens.

To date, ten series, a total of 165 specimens, have been tested, as detailed in

Table 3. The results on the first seven series of these tests will be discussed in

this report. In Figures 4a to 4d, the frequency distributions of P^, o^, M/QD, and

P of the 125 specimens are shown with their failure modes and estimated grade of

ductilities

.

OUTLINE OF TEST RESULTS

Failure Mode of Tested Specimens

The following will describe the typical failure modes that were observed during the

testing of these specimens:

1) Shear failure prior to or after flexure yielding (Notation of mode; S'SC, F-SC,

S'DT, S-ST, F-ST) . This mode can be divided into shear tension failure including

diagonal shear tension failure and shear compression failure.

2) Bond split failure prior to or after flexure yielding (B-B, F-B)

.

3) Compression failure of concrete after flexure tensile yielding with or without

corrpression steel buckling (F, F-C, F'C*B^).

Classification of Ductilities

When discussing the deflection abilities of structural members, the critical angle of

rotation of the members or the critical deflection is called the ductility and is used as an index.

IV-19
In this report, the test results were classified by the ductilities, as defined in

Table 4.

Relationship Between Failure Modes and Estimated Ductilities

The test results of 125 specimens were classified by their failure modes, estimated

ductilities, and their interrelationships and frequencies, as shown in Figures 5a and 5b. As observed in these figures, the most ductile failure mode is type F-C followed by type

SC.

Many specimens failed in the type ST mode and type B mode, which indicated poor ductilities. Accordingly, one of the most important subjects then is to determine how to prevent such brittle failures as revealed by failure mode types SC, ST, and B.

The resulting shear force-deflection relationships, cracking patterns, and deteriorations of the load-carrying capacity of typical specimens are shown in Figures 6, 7, and 8.

DISCUSSION ON TEST RESULTS

Steel Buckling

Buckling of the compressive reinforcement was observed during testing of 32 specimens

and failed in type F-C mode. Ten of the specimens buckled with a ductility factor equal

to 2^4, and thus caused deterioration of the load-bearing capacity. The test results of

these specimens indicate the following:

1) The length of steel which buckled, 5,^, was approximately l'V'2 times the spacing

of the web reinforcement, S. In the case of spiral hoops and welded square

hoops, l-^ was nearly equal to S.

2) The slenderness ratio, X, was calculated for each specimen assuming ly^ = S. The relationship between X and the angle of rotation of the member at the

buckling load R was determined relative to those specimens which did not

fail in bond or shear. These results indicate that when A = 34, the buckling load Rp,,, ^

1 g. _1, however, when A is less than 34, steel buckling did ^ 100 50

not commence until fairly large deformations occurred.

3) Thus, when A = 34, the web reinforcement spacing S <_ 8^ , where (jj is the diameter of the compression bar.

4) In conclusion then, the spacing of web reinforcement at a column end should be

less than eight times the diameter of the axial bar reinforcement.

Discussion on Shear Failure Mode

During the testing of the specimens, 41 failed in shear prior to or after flexure

yielding. Nineteen of the 41 specimens failed in shear compression, ten of the specimens

failed due to diagonal shear tension and twelve failed in shear tension.

Consideration of the Use of Arakawa's Formula

As there are no quantitative equations which relate the ductility of members to the web reinforcement ratio, the failure mode considering Arakawa's formula was used. This equation was adopted as the basis for calculating the web reinforcement in the

specimens

.

The relationship between the observed ductility and the ratio of calculated

shear strength, cQ^j^^, was studied in relationship to the calculated flexure strength,

cQgjj. The formulas that were used to calculate cQaoa and cQ_ are given as Equations

IV- 20
(1) and (2). All of the test results, with the exception of 52 specimens which

failed due to bond-splitting, are shown in Figure 9. As observed from this figure,

the estimated ductility in which cQj^jy^/cQgy is greater than one are for cases A or

B. When the ratios are less than one, the estimated ductility results are referenced

to cases C or D, It should also be noted that as increases, the ductility tends

to become poor.

/ 0.0754 P^°-23k^(c''B + 180)

1 cQ^ = \ + 2.7 v^P„ • sV J X Z bd (1) M/Q +0.12 8

where

= 0.80(d = 21.5 cm) , 0.72(d = 43 cm)

^BU = ^^t.s.y.g + - ) >A (2)

Diagonal Tension Failtire

Initially, the diagonal tension cracking load, Qpr^Q, was investigated. Thus,

theoretical load value (cQj-^pq) was obtained by assuming the principal tensile stress equal to 1 . This stress was calculated at the center of the section, neglecting the appearance of other cracks, and the influence of the axial bars. This re- sults in the following:

cQDTC

= £^ bD /TT^TT^t (3) 1.5

The ratio of test variable (TQj-jt^q) , to the calculated values (cQprpQ) for 23 spe- cimens, in which diagonal tension cracks appeared, varied from 0.66 to 1.13 with an average value of 0.87;

Next, the critical web reinforcement, (P„j^^^) , was calculated by assuming the

web reinforcement carries the diagonal tension, which was initially carried by the

concrete before the appearance of the crack. This results in the following:

P = .

. (4) WDTC 2 a

cos o • s wy

where

e = i tan-l(2 /c^t/a^ + (c°t 70^)2) .

, (4a)

The values of (P^/P^j^^) vs. (TQ^ax/^SoTC^ excluding those test specimens which

failed due to bond-splitting have been plotted, as shown in Figure 10. The terms

'^2max ^w represent the tested maximum shear strengths and the web reinforcement

ratio, respectively.

It can be observed from this figure that the specimens, which have a value of

'^^max 9^^^ter than 0.8 times (cQj^^^) and (P^) and less than ^'^^^^^^J ' ^ diagonal

tension failure mode.

Shear Tension Failure

The initial shear tension cracking load, (Q51PQ) , was first examined. The theoretical value (cQgrp^) was calculated by using the following formula, which was derived assuming the following:

IV-21
when A, = A/D < 0.75 + 0.283 k

1 o

cQg^C " • c t . bD {5a)

when A-j^ = 0.75 + 0.283 k^.

cQgTc = {0.78 - 1.04 A-^ + /(0.78 - 1.04 A-j^) 2 + 0.694 k^^ }c°^ bD (5b)

-2

where k = cP t + o /eft

o o The ratios of (TQ to (cQ^^^q) for 119 of the test specimens in which shear

tension cracks appeared varied from 0.62 to 1.59 with an average value of 0.91.

The second critical web reinforcement ratio, (P^ ) , which is required to have

the specimens reach flexure strength after the appear§Sce of a shear tension crack, was determined from the following:

1.6 jxj^ • • s^^y PwcT

=

where

0^ + Pt • s^v • A.-O-^S

= I di - ~ -T (6a) D 0.85 c^B

The values of (P /P,, ) vs. {TQ~=v/cQnrn^) » excluding those test values which

^ itting, are plotted in Figure 11.

As shown in this figure, there are many specimens which have values of (TQjj^^j^) greater than (cQqTM,-.) . It is also observed that the specimens with a value of (TQ„^ ) max

greater than 0.8 cQg^^ demonstrated a shear-tension failure mode when (P,^) was less

than (P„ )

.

Shear Compression Failure

Because shear compression failure is primarily caused by crushing of the concrete

in the compression zone, it is reasonable to assume that there is a limiting degree

of ductility that can be expected, even if the quantity of web reinforcement is increased. The determination of this limitation is difficult to evaluate and was not

possible in this report. However, certain trends have been established as will now

be described.

To achieve reasonable ductility, it is necessary to maintain the intensity of

the axial stresses at the appropriate level. In general, the influence of the shear

stress on ductility is not important. Therefore, the distance from the neutral axis

to the extreme compressive fiber, X^^, can be determined by neglecting the compression

reinforcement in the compression zone and assuming the ratio of tensile principal

stress to concrete tensile strength, (a^/c'^t) is calculated assuming that all of shear|

force is carried only by the concrete in the compression zone at the critical section,"

thus

,

N + at • s'^ t

"l

0.425 c'^B + /(0.425 c"" ) ^"TT^ (7b)

Xn = -ii = (7a) ^ ° 0.85 • c^B . bD

where

^

TQ^^,

bXn

(7c)

IV- 2 2
The third critical web reinforcement ratio, (P^ ) , was calculated by the so-called

trus s -analogy , which is used as one of the shear strength theories for beams.

TO^max

"wz

=

. — b • J

• s wy

where

j = 0.875 d

04. The values of (.o-^/c '-) vs. Xn, excluding those 52 test specimens exhibiting bondsplit failures, were then plotted as shown in Figure 12.

As observed in this figure, many of the specimens which satisfy the following

limitation showed good ductility even though may be less than or greater than P^^^*

°1

a + 6 -ii < 3 (8) t D -

Using these results, the value corresponding to the left side of Equation (8) was calculated for each specimen. Relationship between the above value and (P^/P^ ) are shown in Figure 13. Clearly shown on this figure are the zones of the shear-tension

failure modes, the shear-compression failure modes, and the flexure-compression failure modes

.

If a design control, as given by Equation (8) , can be refined through more rig- orous tests and study, it may be possible to provide a limitation on the effective

web reinforcement ratio and a combination of P.to, a , and M/QD within which the web

reinforcement is not effective.

Bond-Splitting Failure

A distinguishing feature of the experimental results was the presence of bond-splitting failures of various test specimens. This type of failure mode occurred in 52 specimens and was observed in those specimens which had high tensile reinforcement ratios. The

tensile reinforcement ratio (P^) was 0.95% for 47 specimens and 46 specimens showed this

type of failure mode.

For those specimens which failed in this mode, a small inclined crack first appeared

at the position of the tensile bar close to the end of the so-called shear crack. As the

number of cyclic loadings increased or as the horizontal deflection increased, similar

cracks progressed in number near the center of the specimen. Spalling of the concrete

cover then commenced and the shear capacity of the specimen decreased. Thus, this initial

small inclined crack, called a bond-splitting crack, was considered to be the triggering

mechanism for this type of failure mode

.

The shear force and bond stress at a point on the tensile bar at a distance "d" of

the member, where the shear force corresponds to this initial crack, ^'-^Q-^q^ • "^^^ tained by setting the principal tensile stress equal to (c t) ^ as caused by the bending

moment; _

cQBO

B + B^ + 4C'^(c + c

,ot

2C^

°o)

(kg) (9)

where

C = 1.75 n"

^ 6.1(D-dt - dt ) d le b-D^

B = (M/Q - d) 21e

IV-23
b'

n

'

I, le = Moment of inertia with and without the effect of the bars,

respectively

Comparison of the calculated to the test results, (TQ^^) , are shown in Figure 14. As shown in this figure, (TQ^q) generally is greater than (cQgQ) as (M/QD) and (P^) decrease

and as (o^) increases. However, the influence of the web reinforcement ratio cannot be

perceived.

Next, the ratio of (cQ^^/cQ-qq) vs. the ratio of (^max^^al^ ^'^^ °^ test re- sults are plotted in Figure 15 in accordance to their failure mode and ductility. The

terms, Fj^^^^^ and F^^.' represent the maximum test results of the mean bond stress and the

allowable bond stress in accordance with the A.I.J. (Architectural Institute of Japan)

code. As shown in this figure, the value of {cQ^^/cQ^qO has a greater influence on the

bond-splitting failure mode than does Fmax/^al*

Figure 16 shows the relationship between the value of (P^/P^^) and the estimated ductility of the specimens, which had failed due to bond-splitting.

i As shown in this figure, their ductilities cannot be improved, even if the web reinforcement is increased. However, in those specimens where (c ^) is greater than 270 kg/cm^

and when (cQg^) is less than 1.4 times (cQ-qq) and (P^) and is approximately equal to (P^^ ), good ductility can be observed. It is further presumed from their cracking patterns tha?

the confinement of the concrete within the core of the spiral hoops is effective. ;

CONCLUSION

Multi-cycle flexure-shear tests were conducted on 125 short column specimens; the

following items relative to their ductility were observed:

a) The factors which control ductility of a structural column include not only

shear, but also bond and buckling of the compression bar.

b) To prevent compression reinforcement from buckling at small curvature, the

spacing of the web reinforcement should be less than eight times the diameter of the axial reinforcement.

c) Examination of the bond stress is not an effective means for preventing mem- bers from developing a bond-split failure. However, it is effective in keeping the flexure capacity within 1.4 times the initial bond-splitting cracking

load, as given by Equation (8) . Increasing the number of rectangular- type

hoops is not as effective in preventing bond-splitting failure as is using

more spiral hoops.

d) When the flexure capacity is more than 0.8 times the diagonal tension cracking

load, as given by Equation (3) , an effective means in preventing members from

developing a diagonal tension failure is to place more web reinforcement than

is indicated by Equation (4)

.

e) When the flexure capacity is more than about 0.8 times the shear tension cracking load, calculated by Equation (5) , it is effective to place more web reinforcement that is indicated by Equation (6)

.

f) There is a limiting combination of (P^) , (o^) , and (M/QD) where the shear com- pression failure cannot be avoided, even with an increase in web reinforcement.

b - l<i,^ Minimum value as given by (2/2 dt - (|) ) and

o n

'

Number of tensile reinforcement bars

Diameter of the tensile reinforcement and the summation,

respectively

IV-24
This limiting combination is not apparent, however, Equation (8) can be con- sidered to be one such control. When a combination of (P^) , (a^) , and (M/QD)

of the specimens satisfy Equation (8) and when (P^) is greater than CP^^) i as shown in Equation (7d) , good ductility should be achieved. However, there is

a possibility that the value of (Pw2^ decrease the ductity.

ACKNOWLEDGMENT

The authors wish to thank the follow'ing members of the Reinforced Concrete Structures

Committee in B.R.I, for conducting this experimental research and to Mr. M. Ohkubo, Assistant, Tokyo Metropolitan University, Mr. K. Yamamoto, Research Member, and to Fujita Kogyo

for helping in the preparations of this report.

Chairman ; Y. Higashi, Tokyo Metropolitan University

Members : H. Aoyama, University of Tokyo; Y. Kanoh, Meiji University; M. Fukuhara,

Tokyo Institute of Technology; S. Ohmori , Kajima Construction Company,

Ltd; S. Yoshizaki, Taisei Construction Company, Ltd.; T. Takeda, Obayashigumi Construction Company, Ltd.; Y. Fukada, Shimizu Construction Company,

Ltd.; O. Chiba, Toda Construction Company, Ltd.; K. Yamamoto, Fujita Kogyo;

I. Yamaguchi , Takenaka Komuten; N. Yanagisawa, Secretary of the Committee

REFERENCES

1. M. Ozaki, M. Wakabayashi, and M. Hirosawa, "Experimental Study on Large Models of Re- inforced Concrete Columns," Proceedings of the Fifth World Conference on Earthquake

Engineering in Roma, 1973.

2. M. Hirosawa, "Synthetic Research on Preventing Reinforced Concrete Columns from Total

Collapse (in Japanese)," Building Research Institute BRI Transactions of Main Research

in 1972, 1972.

3. M. Hirosawa, "A List of Past Experimental Results of Reinforced Concrete Columns,"

Kenkyu Siryo No. 2, Building Research Institute, 1973.

4. The 1971 AIJ Standard for Structural Design of Reinforced Concrete Construction.

IV-25
NOTATION

b, D \

d

gPtPw

a

s y

o

s wy

a a

c B, c t QN

aoM

A

width and overall depth of specimen, respectively (cm)

distance from extreme tension fiber to centroid of tension reinforcement (cm)

distance from extreme compression fiber to centroid of tension reinforcement (cm = d - dt (cm)

tensile reinforcement ratio

web reinforcement ratio

2 yield stress of tensile reinforcement (kg/cm ) 2 yield stress of web reinforcement (kg/cm ) 2 compressive and tensile strength of concrete, respectively (kg/cm ) shear force (kg)

constant axial force (kg)

2 : = N/bD (kg/cm ) bending moment at critical section (kg* cm) = M/Q (cm) -.

ADDITIONAL REMARKS

The following analytical studies will be conducted by the committee and authors, the results

of which will be presented in the near future;

1. Influence of scale effect, loading system and loading excursion on the ductility

of structural members.

2. Quantification of the rational web reinforcement ratio to cause structural members

to be ductile.

3. Pursuit of a structural device in preventing the members having split-bond failure.

4. Estimation of hysteresis damping of the test results.

5. A method for estimating the seismic properties of the members by using the results

of certain static cyclic loading tests.

IV-26
TABLE 1 LIST OF FACTORS COMMON TO SPECIMENS

COiJjCN FACTORS NOTE

Fl Section: b x D = 25 x 25 ca, dt = d(.= ?.5 c:ti SE Series: b x D = 50 x 50 era, d^ = A(,= l ,0 era P2 Concrete: Normal concrete, ?c=-iO kfv'cra^ {Desien} FC Series; ?^ = 350 k/'^cra^ C'^esipn)

F3 Shape of wet reinforcement: .-.ectan^lar hoop -vith a standard hook at each end

I'lLOT Series; Aelded rectancjlar hoops used in the half of specimens

F4 eb -ar: SK24 Round bar Specified yield point of 2400 kG/ctn'^ P5 Aiial bar: SD35 deforced bar Specified yield point of 3500 krycn*^

P6 Tensile reinforcement ratio: D^=0.34b (3-DlO), C.Dli (3-D13), 0.95' (3-JlC)

PILOT Series: p^= 0.41* (2-D13), 1.24? (2-D22) AP Series: Pt = i.38' C-Clg)

P7 Shear span ratio 'V^D: iVQD = 1 and 3 IILCT and LS Series: ;i/QD = 1.5 and 3.0

FB Axial stress; K/b3 - 210/4 and 210/5 >^/c3i- 30 ke/cra2 (ULCT S.), 0 and -210/10 kf/cm? (AF S.

)

P9 let reinforcerient ratio; by .^akawa's xin. Equation riLCT Series: by A.I.J, "quatlon

FIO Loadinc r^ethod: Restrained bea:^ type, reversal loadin- of aulti cycles.iAKA3AYASHI-type (riI/)T S.), Inverse-Symraetricaltype (U.;l, SE 3.), D.H.I. -type (other soricr)

0.7Py ~

I ^ tlOcyc les(^-^y)

tlOcycles(6'2Sy) t10cycles(S = 3&y) ^

noi,io2S-88j.

Prr Pao S = 65,^

ilOcycles (S=4&y) n

103

5 '21,'

0

^1,92

^nn! Load at teat equals shear force Q of specimen, (mn; loading number).

Tji Load when tensile reinforcement yields at test, ( Py P21 )•

i Measured horizonteil displacement at yielding.

FIG. 1 LOADING PROCEDURE USED IN ALL SPECIMENS

Q125

FIG. 2 B.R.I. -TYPE LOADING APPARATUS

IV-27
TABLE 2 LIST OF SPECIMENS BELONG TO A TYPICAL SERIES

Mark Tensile

Rei nforcement

Tensile

Rei nforcement

Ratio

Shear Span Ratio Web

Rei nforcement

Web Reinforcement

Ratio

Axial Unit

Stress

Number-Size

(«)

M

Q D Number-SizeSpacing (mm)

(JO 311/ ou (kg/cm2)

IB 3-010 0.34 1 16- 9 - 33.3 1 .53 210/4

2A 3-DlO 0.34 2 15- 9 - 71.5 0.71 210/4

2B 3-DlO 0.34 2 17- 6 - 62.5 0.36 210/4

3A 3-DlO 0.34 1 10- 9 - 55.5 0.92 210/8

38 3-010 0.34 1 11- 6 - 50.0 0.45 210/8

4A 3-010 0.34 2 10- 4 - 55.5 0.18 210/8

48 3-DlO 0.34 2 10- 4 - 111.1 0.09 210/8

5A 3-013 0.61 12-13 - 45.5 2.33 210/8

3-013 0.61 12- 9 - 45.5 1.12 ' 210/8

1 6A 3-D13 0.61 24- 6 - 43.5 0.51 210/8

6B 3-013 0.61 2 28- 4 - 37.0 0.27 210/8

3-D16 0.95

' 2

' 24- 4 - 43.5 2.44 210/4

3-015 0.95 2 25- 9 - 41.7 1.22 210/4

8A . 3-016 0.95 2 26- 9 - 40.0 1.27 210/8

88 3-D16 0.95 2 28- 6 - 37.0 • 0.61 210/8

1500-

500 —za- 500 500

2-9'Ka)6e3

o

I 250 —

-

35 35 -|— 180—j5> u —a

1 —a

FIG. 3 AN EXAMPLE OF SPECIMEN DETAILS (in mm)

IV-28
TABLE 3 LIST OF TEST SERIES

Name of

Series

Section of Specimen

(cm X cm)

Shear Span Ratio

(M/QD)

Load ing 1

Sys t em

Institution

•2

in Charge

Year Main

Objective

Jumbcr of

Spec imens

Pilot 25 X 25 1.5, 3.0 W - Type B.R.I. 1971 Welded Hoop 36

LMl 25 X 25 1.0, 2.0 R.B - Type T.I. of Takenaka

Komuten

1972 Loading System

Scale Effect

15

LM2 25 X 25 1.0, 2.0 BRI - Type Tokyo Instltue 1973 Loading System 15

SE 50 X 50 1.0, 2.0 R,B - Type .Meijl Univ. and B.R.I. 1972 Scale Effect 15

FC 25 X 25 1.0, 2.0 BRI - Type T.I. of Talsel

Const Co • , Ltd. 1972 Concrete

Strength

15

WS 25 X 25 1.0, 2.0 BRI - Type T.I. of ObayashlGual Co .

, Ltd

.

1973 Welded Hoop of Deformed Bar

15

Af 25 X 25 1.0, 2.0 BRI - Type T.I. of Fujita Kogyo 1973

Axial Force 15

CW 25 X 25 1.0, 2.0 BRI - Type Tokyo Metropolitan L'nlv. 1973 Columns with

Side Wall

10

WS2 25 X 25 1.0, 2.0 BRI - Type T.I. of ItaJ Ima Const. Co., Ltd. 1973 Spiral Hoop 15

LS 25 X 25 1.5, 3.0 BRI - Type T.D.C. of Toda Const . Co. , Ltd

.

1973

Shear Span Ratio15

*1 W - Type : Loading System devised by Dr. Wakabayashl

R.B - Type : Restrained Beam Type

EPI - Type : Loading System newly developed in B.R.I.

•2 T.I. : Technical Institute

T.D.C. : Technical Development Center

(•/.) 40.

30.

20.

10

a) (N/bD)

36.0 I

DT 0ST

QSC

Q BO

FC

^28.8 i

3.2

5.6

20.'5' — ("/») ,100

(•/.) AO

30

5.6

80 20

J 6040 10

20

-21 30 50 70 0

b) (M/QD) t

n46.4

16.0x'

12.8

1.5 — M/QD

(•/.) 20.

10_.

(Pt)

20.8

0l_

d) (Pu

.19.2- 16.0 I 13.6 I

7.2

0.5

J12_

1.0

2.4

(°/.) 100

- 806040

20

15 2.0 -»P^(7.)

FIG. 4 FREQUENCY DISTRIBUTION OF ^o, H/QD, Pt and Pi

IV-29
TABLE 4 CLASSIFIED DUCTILITY

DUCTILITY CHARACTERISTICS

A

Very ductile columns which failed by shear or by buckling of

compression bars at horizontal large deflectiont

fPrt^ Pot > O.ft^OTarr ^Owt* linftxiil»Al tl'fin^^'fch oli^Alnut) "hv A.I.J, formula)

B

Ductile columns, whose deterioration of shear capacity wer*

small untill /^-4, <^/<f/ ), but which failed by ehear

or by bond or by buckling of bar before ^ • 6.

(P21 ~ P71 Z 0.75cQpu)

. C

Columns yielded by flexure at first, but deteriorated remarkably due to shear or bond failure or buckling of bars before

they reached to large deflection.

(P2i'^ P3I ^ 0.75cQpu)

D

Columns failed by shear or by bond before flezural yielding.

(Others than A, B and C)

g) DUCTILITY b) MODE

FIG. 5 CLASSIFICATION OF TEST RESULTS BY DUCTILITY AND FAILURE MODE

IV-30
IV-31
LM2-1B Type CRACKING PATTERNS

LOAD Pii Q/A5 = 19.2('"9/Jm') Ry =5.0 ('lO^'Radi.) r

LM2-6AType CRACKING PATTERNS

LOAD P)i I Q/Ag=ia3('>?^m')

Ry =8.0

'

(«IO''Ra(Ji-) '

LOAD P30 Q/Ag=20.A8 Rv =9,2

LOAD P30

Q/Ag=13.3

Ry =a0

LOAD P40 Q/Ag=19.2

Ry =18.4

LOAD Pdti(72) Q/Ag =21.4 Ry =36B

LOAD Pio Q/Ag=U,7

Ry =16.0

LOAD Peo Q/Ag =144 Ry =32.0

LOAD Puiti Q/Ag = 15.7 Ry =640

LM2 7BType CRACKING PATTERNS

LOAD Pii I Q/Ag=20.5(i'a^n<)-^'"'

Ry =aq ni:

LM2-8AType CRACKING PATTERNS

LOAD Pii I Q/Ag = 1 6.8 ( '-JJ/

Ry =78 ' \

fxlo'Radi) '

LOAD P30 Q/Ag=17.9

Ry =95

LOAD P30 Q/Ag = 144 Ry =7.5

J

LOAD P40 Q/Ag = 212 Ry =19.0

LOAD P/.0 Q/Ag = 19.

2

Ry =14.0

LOAD P80 Q/Ag = 12.5 Ry =3aO

LOAD Puiti Q/Ag =22.5

LOAD Peo Q/Ag =18.1 Ry =28.0

LOAD Puiti Q/Ag =19.4 Ry =66.0

FIG. 7 CRACKING PATTERNS OF TYPICAL SPECIMENS

IV-32
91 101

Number of Cyclic Loac DETERIORATION OF SHEAR CAPACITY BY CYCLIC LOADING ON TYPICAL SPECIMEN

<

<

Of

1.20.

1.10.

1.00.

0.9.

0.8.

0.7.

0

I Ductility A B C D

Mark o X

D -

a

CD A

cn o

o

D

A

A O

X X

-i— —

10 20 30 Am. /kg / .\

FIG. 9 DISCUSSION OF DUCTILITY BY ARAKAWA'S EQUATION

iv-33
Ductility A B C D DT + X *

others o 4> •

O0

o /r

^-(F^DT)

° of E

Ox

0.5 1.0

FIG. 10

tQmax

cQdtc

1.5

p

I 'WST 4

• Ml » 4- <)D

Ductility a|b c D

others o • ST . + X +

4-

0

o

4-

-e 4-

_^^4 J J (F-^ST)

0 0.5 1.0

FIG. n

1.5 tQn CWST

0

\

\

\

V—

\

\

\

\

3.0^ I I

A

2.0

1.0

\

\

0.25

FIG. 12

XnO.5 0

D

DUCTILITY A B C D FC *

SC o •

4 ST A

\

\

A

',v 1 ^

\

\

\

0.25 Xn 0.5

D

IV- 3U
1.5

Pwz

10

Ductility A B C D

FC 0 tp SC O • ST A

0.5

n a

o

(F-SC) . O 1 J

(F-^C)^

o

I

'—

* o

4

rri-n

Vn I

4^

0 1.0

FIG. 13

2.0 3.0 5, ^.XnfO

15 _

omO

10

1/

AA/

5aXx

* « *

^ o «

I 90 I

M

Q-D

Pt (°/°)

0.61 2 0.95

1,1.5 o A X 2,3 • A *

5

FIG. 14

10 15 cQBo(t)20

IV- 35
3.0

Fal

2.0

1.0

, ° 1 ^ , X ^

1 <?K A —-—

o

X A

O

A

A AA

1 ^

° i

J*

^s. -

A

A -AA«

Ductility A B c D

Bond •

others j O A X

05

FIG. 15

1.0 1.5 2.0 cQbo

FIG. 16

2.0 cQsu 3.0 cObo

iv-36
NONLINEAR ANALYSIS OF A GUYED TOWER

by

S.K. Takahashi

Research Structural Engineer

W.A. Shaw, Ph.D.

Head, Civil Engineering Department

Civil Engineering Laboratory

Naval Construction Battalion Center

Port Hueneme, California

The Civil Engineering Laboratory has performed an analysis of a tower at

a Naval communication facility. The tower is about 600 feet high and is guyed

at three levels. The upper guy level contains twelve wires; the middle and lower

guy levels have three wires each. The guy wires had numerous large electrical

insulators attached, each weighing 510 pounds.

A nonlinear finite element analysis of the guyed tower was conducted. Sep- arate finite elements were used for portions of the tower and the guy wires were modeled with a truss element. Deflections, forces, and stresses in the tower and

guy wires were determined from dead load and an equivalent static wind load corre- sponding to 90 miles per hour.

Eigenvalue solutions were obtained for the first five mode shapes and natural

frequencies of one of the top guy wires; the guy wire had concentrated masses at

five different locations and was initially prestressed.

Key Words: Finite element; guyed tower; structural analysis; structural

engineering, vibration analysis, wind load.

IV- 37
OBJECTIVE

The objective of this paper is to present results of a structural analysis of a 600-

foot tall radio tower which is guyed at three levels; the analysis was conducted using a three-dimensional nonlinear computer program. The tower was idealized as a vertical con- tinuous beam supported at three elevations by guy wires. In the analysis, the tower and

guys were subjected to gravity loads and equivalent wind loads corresponding to a horizontal v;ind speed of 90 miles per hour.

The free vibration analysis of one of the top guy wires with concentrated vertical

loads (insulators) and initial tension in the wire was also perfoirmed.

BACKGROUND - ^

The U.S. Navy has numerous tall guyed towers for radio communication located through- . out the world. Many of these towers were designed and constructed over ten years ago. These towers are often s\abjected to high winds; the accompanying vibrations of the towers

and guy wires have, at times, resulted in damage to non-structural elements such as beacon

lights and copper straps which attach the electrical wires to the tower.

Concern has been expressed about the structural integrity of these towers from the

standpoints of injuries to personnel, damage to equipment, and inability to carry out the

assigned defense mission.

There have been several significant failures of tall towers in the United States during

1973. A 1600-foot television tower collapsed and killed two men near Tampa, Florida in June (1)*; the failure of this tower was attributed to loosening the wrong bolts during

modification of the tower. In October, a 2000-foot television tower fell in Cedar Rapids, Iowa and killed five men (2) ; this failure may have been caused by deterioration of a guy

wire at the 500-foot level. A third tower collapsed near Des Moines, Iowa in December;

the failure of this 1882-foot television tower was caused by severe ice and wind loadings

(3) .

The U.S. Navy believed it desirable to re-analyze some of the towers in its communication system. The Civil Engineering Laboratory, Naval Construction Battalion Center, Port

Hueneme, California has devoted some effort to this task. A nonlinear computer program

has been used in the analyses. This paper discusses the analysis of one of these towers.

The analytical investigation was sponsored by the Naval Facilities Engineering Command .

DESCRIPTION OF TOWER

Two 600-foot guyed towers have been in operation at one of the Naval communication

facilities for more than ten years. A photograph of one of these towers is shown in Figure 1. Some of the components of this tower are shown in Figure 2. This tower has three

levels of guy wires; the two lower levels have three guy wires each and the top level has

twelve guy wires. An elevation sketch of this tower is shown in Figure 3. Insulators are located in the guy wires as shown in this figure; each circle represents one insulator

which weighs 510 pounds. Insulators are on each wire although they are shown only on one

side in Figure 3. Figure 4 shows the numbering system of the guy wires.

The guy cables are pre-stretched, multiple strand, zinc-coated Bridge rope with wire

strand cores. Table 1 provides information on the diameter, minimum breaking strength,

and initial tension for the guy wires. (The last two columns in this table will be discussed later in this paper.) Figure 5 shows the location of attachment of the guy wires

and defines terms used in this paper. The modulus of elasticity of the guy wires is 19,000

ksi and the allowable stress is 64 ksi.

Numbers refer to references given at the end of the paper.

IV- 38
The main legs of the tower are extra strong round steel pipes. From the base up to node 3, 8-inch round pipe was used; 6-inch round pipe was placed between nodes 3 and 5; the top portion of the tower had 4-inch round pipe. The cross-sectional area and moment

of inertia of these pipe sections are tabulated in Table 2. The main tower legs are spaced

nine feet' apart and are braced with horizontal angle sections and diagonal tension rods,

as shown in Figure 6.

FORCES ON THE TOWER

Loads on a guyed tower result from dead load of the tower members and guy wires

,

initial tension in the guy wires, wind load, and ice load. The tower under consideration

is located in an area where ice loads are not a problem.

Dead Load of Tower Members and Guv Wires

The total gravity weight of the angle irons, tensile rods, gusset plates, bolts,

ladder, and platforms is approximately 100 lb. /ft. of height. The total weight of the

three legs and the other members is calculated as

:

Diameter of Pipe Weight Weight of Three Legs Total Weight

Section (in.) (lb/ft) (lb/ft) (lb/ft)

Top 5 20.78 62 162

Middle 6 28.57 86 186

Bottom 8 43.39 130 230

As shown in Figures 1 and 3, the guy wires carry numerous insulators that weigh 510

pounds each. To simplify the analysis, it was decided to represent the actual guy wire

and the insulators by a guy wire equivalent weight and diameter so that the length of this

guy wire is equal to the length of the actual guy wire (4) . This equivalent guy wire procedure is useful where only cable end reactions, spring constants, and maximum cable ten- sions are of interest. A computer program based on Reference 4 has been used and the re- sulting equivalent properties of the guy wires are as follows:

Guy Level

Equivalent

Diameter

(in.)

Equivalent

Weight

(lb/ft)

Actual Metallic

Area (in.*')

Top 1.531 7.760 0.361

Middle 1.921 7.510 1.270

Bottom 1.879 7.710 ' 1.080

Initial Tension in Guy Wires

The initial tension at ambient temperature for the top, middle, and lower levels of

guy wires is 7.52 kips, 23.20 kips, and 31.04 kips, respectively.

Wind Load

The wind is assumed to be blowing perpendicular to one of the flat faces of the tower. For a typical 14-foot height of panel section, the wind surface area of the flat face of

the angle irons and tension rods is approximately 0.445 sq. ft. per foot of height. The

wind surface areas of the pipe legs are as follows:

w-39
Section

Outside

Diameter

(in.)

Surface Areas

of Two Pipes

(sq ft/ft)

Total Area Times

(sq ft/ft)

Top 5.563 0.927 2.425

Middle 6.625 1.104 2.695

Bottom 8.625 1.438 3.205

shape Factor for Flat Members =2.28

Shape Factor for Cylindrical Members = 2/3 x 2.28 = 1.52

The uniformly distributed horizontal load, W, on the tower segments is obtained from

the formula:

W = A X q

where A is the projected area (including the shape factor) per unit height and q is the

velocity wind pressure.

The velocity wind pressure is obtained from ' '

q = 0.00256 V^C^^

where V ^is the wind velocity of 90 mph and is the correction coefficient for height or

(h/30)^2/7

Using the mid-height elevations of 540.2 ft., 376.6 ft., and 139.3 ft. for the sec- tions with the 5-, 6-, and 8-inch diameter pipes, respectively, the height correction fac- tors and resulting uniformly distributed horizontal load on the tower sections for a 90

mph wind are as follows:

Height Correction

Section Factor, Cy,

Top 2.28

Middle 2.06

Bottom 1.55

Velocity Pressure

q (ksi)

0.0473

0.0427

0.0321

Uniform Horizontal

Load, W (kips/ft)

0.115

0.115

0.103

NONLINEAR ANALYSIS OF GUYED TOWER

The information in the previous section was used as input data to a computer program

which idealizes a tower as a vertical continuous beam supported at the specified elevations

by guy wires (5) . The assumptions used in the program are the following:

1. Each tower segment is prismatic (i.e., the section properties defining re- sistance to bending about either principal axis and axial, shear, and tor-

' sional deformations are constant between tower nodes).

2. The cable profile is a parabola for all loading conditions.

3. Anchorage cable tensions and total cable weight are known.

4. Wind loads on the tower shaft are known and are uniform between tower nodes.

IV-40
5. Concentrated forces can be applied at tower nodes only.

6. The guy wires are uniformly loaded by the wind.

7. The average wind velocity is the same for all guy wires.

8. The wind is blowing parallel to the ground.

9. Lift and drag coefficients for each guy wire are computed automatically using

the formulas given in Reference 6.

The printed output of the computer program contains the following:

1. All input data.

2. Data checking parameters such as guy chord lengths, unstressed cable lengths,

initial attach point cable tensions, etc.

3. Global translations and rotations of all tower nodes for each load case.

4. Global translations of all cable attach points.

5. Shear, moments, axial forces, and torques acting on the ends of each tower

segment.

6. Attach point and anchor point cable tensions and global components of the

cable resultant forces.

7. Tower base reactions.

The computer program required six iterations before the solution converged. A sum- mary of the displacements at various locations of the tower is given in Table 3. The maxi- mum displacement at the top of the tower was found to be 7.39 feet; this is more than

twice the deflection' at the middle level of the guy wires which indicates that the top portion of the tower is quite flexible compared to the bottom two portions of the tower.

A tabulation of shear, axial force, and bending moment in the tower is presented in

Table 4. The base shear of 6.0 kips is well below the 15 kips shear to which the base of

the tower was tested, however, the axial force of 433 kips at the base of the tower is

slightly greater than the 400 kips axial test load. The maximum moment of 2,154-foot kips

occurred just above the middle guy level. Figure 7 graphically shows the deflection,

shear, axial force, and bending moment in the tower.

The axial coirpressive stresses in the main columns of the tower were computed from

the axial loads and bending moments given in Table 4. These stresses in the leeward

column are summarized in Table 5. The highest stress of 47.8 ksi occurs at the intersection of the 5- and 6-inch diameter pipe sections and is caused by the large deflection at

the top of the tower. High stresses of approximately 40 ksi also occur at the connecting

point of the middle level of guy wires.

Table 6 summarizes the final tensions at the anchor end and the attachment point to

the column for all guy wires. Figure 4 shows the designation of the guy wire numbers.

The allowable stress of 64 ksi was slightly exceeded at the attach point for guy wires

8, 9, 17, and 18. The design working load for the guy insulators was 78 kips; the maximum

tension of 65.4 kips occurred in guy wires 4 and 6.

FREQUENCY ANALYSIS OF GUY WIRE .

A general nonlinear finite element computer program called NONSAP (Reference 7) was

used to determine the natural frequencies and mode shapes of one of the top level guy

IV-41
wires. This wire is pinned at both ends. The computer model is represented by twelve

truss elements and thirteen nodal joints. Concentrated loads of 510 pounds each were placec at joints 2, 4, and 6; a cluster of six 510-pound loads was placed at joint 8. Initially,

the deformed shape of the wire was determined considering only the gravity weight of the

wire and insulators. This deformed shape and initial tension in the guy wire was used in

the computer program to determine the natural frequencies of the wire; the first five mode

shapes and natural frequencies are shown in Figure 8. Except for the first mode shape,

the cluster of six concentrated loads at Node 8 forces the cable to cross the initial position at this node for the remaining four mode shapes; it is as if two separate cables were oscillating independently.

SUMMARY i

. . I A 600-foot high guyed tower which was designed in 1960 was re-analyzed using a non- linear finite element computer program. The tower shaft was represented by a series of

co-linear beam segments interconnected at tower nodes. The tower was analyzed for a wind

velocity of 90 miles per hour imposed on one of the flat faces of the triangular guyed

tower. The analysis showed that a maximum deflection of 7.39 feet occurred at the top of

the tower. The base shear force of 6.0 kips was considerably below the proof load of 15

kips; the axial force of 433 kips at the base of the insulator, however, was in excess of

the 400-kip proof load. The allowable guy wire stress of 64 ksi was exceeded by four of

the twelve top guy wires by less than six percent and should not be of great concern.

A frequency analysis of one of the top guy wires shows that mode shapes 2 through 5 had a node at the location of the cluster of six insulators.

REFERENCES

1. Engineering News Record, "Collapse of a Guyed 1,600-ft TV Tower Kills Two Workers,"

June 14, 1973.

2. Engineering News Record, "2,000-ft. Tower Falls, Killing Five," October 11, 1973.

3. Engineering News Record, "Ice-laden, 1 , 882-ft-high TV Tower Collapses," December 13,

1973.

4. R.C. Morales, "Shear-Volume Method of Solving Tensions in Cables," ASCE, J. Structural

Division Proceedings, 5718 STl, January 1968.

5. F.E. Peterson, "Tower Analysis Program," Engineering Analysis Corporation, June 1973.

6. W.S. Diehl, "Engineering Aerodynamics", Revised Edition, New York, 1936, p. 280.

7. K.J. Bathe and F.E. Peterson, "Theoretical Basis for NONSAP, A Nonlinear Structural

Analysis Program for Static or Dynamic Analysis of Nonlinear Structural Systems

Modeled with Finite Elements," by Engineering Analysis Corporation for Civil Engineering Laboratory, October 1973.

IV-42
r— «C CO

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CVJ L_. •> LO C_J <D -1— 1 T—' ^ -O E

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IV-43
Table 3. Summary of Tower Displacements

Tnwpy* 1 UWC 1 NnHp mUUC

No.

Hi c n 1 a ^ omo

n U "f*

1 op 1 dCcMlcM L (ft)

6 7.39

5 3.37

'

4 3.04

3 1.20

^

2 0.95

1 0

Table 4. Forces and Moments in Tower

Segment

No.

Segment

End

Shear

(kips)

Axial Force

(kips)

Bending Moment

(ft-kips)

5 Upper 2.1 160 -160

Lower 17.4 182 1,807

4 Upper 17.4 182 1,807

Lower 19.2 185 2,154

3 Upper -19.0 279 1,922

Lower 2.1 313 946

2 Upper 2.1 313 946

Lower 6.7 323 1,219

1 Upper -18.9 378 1,142

Lower 6.0 433 0

IV-44
Segment

No.

Segment

End

Axial Stress

(ksi)

Bending Stress

(ksi)

1 Total Stress

(ksi) 1

1 5 Upper 8.78 - 0.3 8.5 i

5 Lower 9.9 37.9 47.8

\

4 Upper 7.2 27.6 34.8

4 Lower 7.3 32.9 40.2

3 Upper 11.1 29.3 40.4 ;

3 Lower 12.4 14.4 26.8 ;

2 Upper 8.2 9.5 17.7 !

2 Lower 8.4 12.3 20.7

1 Upper 9.9 11.5 21.4

1 Lower 11.3 0 11.3

Q stresses indicated above for this column

^

Direction of wind

IV-45
Table 6. Guy Wire Forces and Stresses

Attach Point to Column Anchor Point

Guy

Number

Tension

Force

(kips

)

Stress

(ksi)

Percent of

Design Stress

Tension

Force

(kips)

Stress

(ksi)

Percent of

Design Stress

1 49 .

1

A f C 45.5 71 .1A A 47 .4/I o o 43.9 ^ o c 68.6

2 16.9 15.7 24.5 14.9 TO O 13.8 O T ^ 21 .6

3

ill O T 49.

1

45.5 71 .1 47 .4

/I o o 43.9 68.6

4 65.4 51 .5 80.5 62.3 49.0 76.6

5 2.8 2.2 3.4 1 .0 0.8 T O

1 .3

6 65.4 51 .5 80.5 62 .

3

yi o A 49.0 76 .6

7 22 .

5

62.4 97.5 18.3 ET A O 50.8 79 .4

8 23.3 64.3 100.5 19.1 CO A 52 .9 no "7 82.7

9 24.4 67.6 105.6 20.3 C £Z O 56 .

2

87 .8

10 22.7 63.0 98.4 TO O 18.3 C A "7 50 .

7

79 .

2

11 16.8 46.5 72.7 TO T 12.1 33.4 CO o 52.2

12 9.0 25.0 39.1 4.3 1 1 .8 18.4

13 4.1 11.5 18.0 T T

1 .

1

o 1 3.1 A n 4.8

14 9.0 25.0 39.1 4.3 11.8 18.4 ,

15 16.8 46.5 72.7 12.1 33.4 52.2

16 22.7 63.0 98.4 18.3 50.7 79.2

17 24.4 67.6 105.6 20.3 56.2 87.8

18 23.3 64.3 100.5 19.1 52.9 82.7

IV- 4 6
Figure 1. 600-ft high guyed tower at Dixon, CA.

IV-47
!

IV-48
IV-49

ELEVATION

(FT.)

605.9

SEGMENT

No.

474.6

458.9

278.6

234.9

0

NODE LOCATION OF NOM. PIPE

No. GUY WIRES DIA. OF

COLUMN LEG

6 6

5

4

3

2

1

5 INCHES

6 INCHES

8 INCHES

pipe sizes

Figure 5. Definition of terms and location of guy wires and

i

IV- 51
IV-52
IV- 5 3
IV- 5 4
A STANDARD FOR THE STRUCTURAL INTEGRITY OF PREFABRICATED DWELLINGS

by

Kiyoshi Nakano

Head, Dr. Eng., Third Division

Building Research Institute

Ministry of Construction, Japan

Masaya Hirosawa

Chief Researcher, Third Division

Building Research Institute

Ministry of Construction, Japan

Tatsuo Murota

Research Associate, Third Division

Building Research Institute

Ministry of Construction, Japan

In 1973, the Ministry of Constuction of Japan presented a standard for the

performance of prefabricated housing. The purpose of this standard was to provide consumers with an index for selecting their dwellings. The standards are related to fire, heat, sound, durability, and structural safety against earthquakes, winds, snow, etc.

The criteria for evaluating structural properties are outlined in this

paper, as well as the history and present status of prefabricated dwellings in

Japan. A typical prefabricated structural system is shown in the last section.

Key Words: Earthquake; housing; performance specification; prefabricated dwelli

standards; structural design.

IV-55
A HISTORICAL SKETCH OF THE PREFABRI CATION OF DWELLINGS IN JAPAN AND THE PRESENT STATUS OF THEIR CONSTRUCTION

In 1945, just after the Second World War, the number of houses required in Japan was 4,200 thousand units, which represented 20% of the total number of households at that time

In order to efficiently solve the housing shortage, the Ministry of Construction recognizee

the necessity to industrialize dwelling construction. Therefore, the Building Research In- stitute initiated studies to develop a new system of dwelling construction that would reduc<

the construction cost, the time required for construction, and the labor force. The re- sults of these studies were reflected in the construction of dwellings supervised under

public management.

Since that period, restoration of buildings and structures have been required due to damage caused by the Isewan Typhoon in 1959, and a subsequent building boom due to an in- crease in business prosperity in 1960. This has led to a shortage of construction mater- . ials and skilled laborers. At this time, the Ministry of Construction had decided to ac- celerate the industrialization of dwelling construction, and since 1961, introduced mass- produced systems into the construction of publicly managed dwellings. In 1962, in respons« to this policy, private enterprise has established the Prefabricated Building Association.!

At the end of 1962, 835 low-rise (one or two stories), mass-produced dwellings were | constructed under public management and in 1965, the quantities of low-rise mass-produced

dwellings accounted for one- third of all publicly managed dwellings constructed that year.

In the latter half of the 1960's, construction of prefabricated dwellings began to increase rapidly. This was because the demand for a greater number of dwellings had oc- curred with a corresponding increase in the standard of living. This demand, therefore, was characterized not by the shortage of the number of dwellings (the shortage in numbers

was solved in 1963), but for an increase in a higher standard of house. Figure 1(a) shows

the increase in the number of prefabricated dwellings, constructed each year, during 1966

through 1972. Figure 1(b) shows the percentage of the total number of dwellings construct(

in the same period. According to these figures, the number of prefabricated dwellings con^ structed in 1972 exceeded 200 thousand, which was more than 10% of the total dwellings con- structed .

The details of the prefabricated dwellings constructed in the same period is shown

in Figure 2. The ratio of the low-rise dwellings to those of more than three stories was

78 to 22 in 1966; this ratio then became 61 to 39 in 1972. The increase in the construction of high-rise dwellings is due to the rising value of land, which is estimated to con- tinue in the future.

Single- or two-story prefabricated dwellings are divided into three types, according

to the material used in the main parts of the structures, i.e., wooden type, steel type,

and concrete type. The low-rise prefabricated dwellings constructed in 1972 consisted of

the follwoing percentages: 52.4% were of the steel type, 36.4% were of the wooden type,

and 11.2% of the concrete type.

As many as 100 firms are presently producing these low-rise dwellings of more than

1,000 plan types, however, some of these have notable defects.

In 1973, the Ministry of Construction developed a performance standard for low-rise

prefabricated dwellings in order to provide consumers with a criteria in selecting their

dwellings. According to this regulation, the Minister of Construction stipulates the performances relative to fire, heat, sound, durability, and structural safety, and notifies

the consumers of these performances with the exception of structural safety.

In the following section, an outline of the criteria, relative to how the structural

performance was evaluated, will be described.

IV-56
CRITERIA FOR THE EVALUATION OF THE STRUCTURAL PERFORMANCE

OF PREFABRICATED DWELLINGS

The establishment of a criteria for prefabricated dwelling systems first requires one

to satisfy the structural performances specified by the Building Code of Japan and the

Specifications of the Architectural Institute of Japan. In addition to satisfying these

basic requirements, the criteria must include the specifications peculiar to dwellings

such as flexibility of partitions, rigidity of floors, etc. Also the evaluation of the

aseismicity of reinforced-concrete dwellings, utilizing a new method based on the earthquake

energy absorpotion capacity of the structure, has been adopted.

The criteria will be referenced to the following two areas:

The structural design philosophy relative to dead and live loads, earthquakes, winds,

and snow must be clear and reasonable. Complex and/or unreasonable structural plans and

elevations will not be governed by these design criteria. If the structural elements are arranged to resist horizontal loads, such as bracings and shear walls, they must be inspected into detail. If the structural system has wall panels which resist horizontal

loads (this system will be called a "load-bearing wall panel system") , the maximum distance

between two adjacent lines of walls positioned in the same direction shall be less than

7.5 m and at least one load-bearing wall panel, which has a minimum width of 80 cm, shall

be placed at every corner of the exterior wall lines. If the housing structure is made

of precast concrete panels, in addition to the above-mentioned regulations, the following

requirements relative to seismic characteristics shall be verified by calculation or ex- periment .

a) In case of verification by calculation

The coefficient of the horizontal load-carrying capacity, S, shown in Equation

(1) shall not be less than 0.5.

1) Structural Design Philosophy

2) Detail of Structure

STRUCTURAL DESIGN PHILOSOPHY

S - k^ • k2 • k3 • Sg (1)

where

k

1 Coefficient based on the arrangement of the wall panels in

plan and elevation, (0.7 1.0)

Coefficient of the horizontal rigidity of the floor,

(0.7 1.0)

k

3 Coefficient for the difficulty of construction, (0.8 1.0)

The standard seismic strength coefficient, the smallest of

the values obtained from Sg in Equation (2) or Sgg in Equation (3).

^

The shear strength coefficient, Sg , shall be calculated by using Equation (2)

" '^fc * '^Pw '^hj '^vj ^w ^(2)

where

kj^ = Coefficient associated with the wall panel strength of concrete, (0.2 ^ 1.1)

IV-57
k = Coefficient of the shear reinforcement ratio of the wall

panel, (0.9 1.4)

kj^j = Coefficient corresponding to the shear strength of the

horizontal joint, (0.6 1.0)

k^. = Coefficient corresponding to shear strength of the vertical

joint, (0.6 1.0)

= Coefficient of the standard horizontal load-carrying capacity

defined by 10 A^/W

= Total cross-sectional area of the shear walls in one direction

per unit floor area of the first story, (cm. ^/cm. ^)

W = Total weight of the house per unit floor area at the first

story (kg./cm.^)

10 = Assumed value of the standard unit of strength of the shear

wall, (kg./cm.^)

The flexural strength coefficient, S , shall be calculated by using Equation (3), EB

where

Seb = ^ ' 2

'^pt

• '^d

• S„ (3)

a = Coefficient corresponding to the reinforcement effect of the

walls perpendicular to the direction under examination, (1.0 1.4)

B = Coefficient corresponding to the flexural strength of the

collar beams (1.0 1.5)

kp^ = Coefficient corresponding to the bending reinforcement and

anchorage of the shear walls

k^ = Coefficient corresponding to the ductility, (1.0 or 1.5)

b) In case of verification by experiment

i) The test specimen shall be a full-scale two- or three-dimensional structure

fixed rigidly to the basement.

ii) The loading shall be cyclic and shall be controlled by horizontal deflection.

There will be four loading stages with three cycles of loading for every

stage. The various stages will consist of 1, 2, 3, and 4 times the deflection that occurs due to the design horizontal load and the deflections which

correspond to the rotation angles of 1/22, 1/100, and 1/50. After the cyclic

loads, this specimen shall be loaded to collapse.

iii) The following requirement shall also be satisfied.

Critical Condition Required Value of "S"

A. Condition when severe structural

damage is observed Not less than 0.4

B. Ultimate Condition Not less than 0.6

The coefficient "S" may be calculated by using the following equation.

IV-58
S = • k2 • Cg • /2P - 1

where k-]_, ^2 are the variables as defined for Equation (1) ,

Cg ; PA?

P is as defined in the following table (see Figure 3)

Critical

Condition A B

P 80% of P^, which is the

maximum load observed

before the specimen

reaches Condition A

80% of P„,^ max

W is the total weight of the specimen, including live load

y is 6/6

DETAILS OF STRUCTURE

A. Load-bearing Wall Panels

The height of load-bearing wall panels shall be less than three times the width. The

structural performance of the wall panels shall be checked by the following experiments

specified in the Japanese Industrial Standard JIS 1414,

i) In-plane Shearing Test

ii) Eccentric Loading Test

iii) Bending Test

iv) Impact Test

In the case of steel structures, the design shearing force of the load-bearing wall

panels shall not exceed 2/3 of the maximum shearing force, as obtained from the above sheartest. In addition, the shearing test results shall satisfy the following:

i) The horizontal displacement at the top of the panel, when the design shearing

force is applied, shall not exceed 1/150 of the panel height.

ii) When the horizontal displacement at the top of the panel becomes three times the

displacement due to the design shearing force, the panel shall be able to resist

a shearing force equal to 5/4 times the design shearing force.

iii) Failure or large local deformations shall not occur on the surface of the cladding at a minimum horizontal displacement of 1/150 of the panel height.

B. Other Wall Panels

The same type of tests, as described

In the in-plane shearing test, failure or

the horizontal displacement at the top of

above, shall be performed for other wall panels,

large local deformations shall not occur until

the panel equals 1/100 of the panel height.

IV- 5 9
C. Floors and Beams

Floors and beams shall maintain sufficient safety against dead and live loads. Vertical deflection shall not exceed 10 mm. nor 1/300 of the span. Failure or large local deformations shall not occur on the surface of the floors when a locally concentrated load

of 150 kg magnitude is applied at any point.

D. Roofs

The vertical displacement of roofs, due to snow loads, shall not exceed 1/200 of the

span.

Steel or wooden roofs shall be safe against wind loads. Deflection of the eaves when

siibjected to a distributed load of 240 kg./m. ^ shall not exceed 1/100 of the spray length

and the eaves shall not fail under a distributed load of 360 kg./m.^.

E. Horizontal Plane Structures

Horizontal plane structures such as roofs and floors shall be rigid enough to transmit

shearing forces to columns or walls. It is, therefore, necessary to provide horizontal

bracings or to rigidly connect the roof or floor panels.

F . Columns

The column slenderness ratio of steel structures shall not exceed 200. The buckling

length of a column shall be taken as its total length. Columns which consist of vertical

ribs of wall panels and are connected to each other by bolts, the slenderness ratio shall

be estimated from the loading test, as specified in JIS A 3304. The bearing capacity shall

also be confirmed by full-scale tests.

G. Foundations

Foundations shall transmit external forces safely to the ground and resist overturning

and differential settlement of the dwellings.

AN EXAMPLE OF INDUSTRIALIZED HOUSING STRUCTURE

(A Medium-sized Concrete Panel System)

As described in the first section, mass-production of industrialized housing in Japan

started in the 1960 's. During this period, the industrialized house was primarily con- structed of prefabricated reinforced concrete panels. They were called "mass-produced publicly managed dwelling type", because they were adopted as publicly managed houses. Classification of the current industrialized houses of reinforced concrete structures are shown

in Table 1. The most prevalent system at present is a medium-sized concrete panel system

listed as Type "A" in the table. This sytem was developed by improving on the above-mentioned "mass-produced publicly managed dwelling type", and is now one of the most typical

industrialized houses in Japan. In this section an outline of this type of medium-sized

concrete panel system will be described relative to the structure.

OUTLINE OF THE STRUCTURAL SYSTEM

Medium-sized concrete panel systems consist of precast wall panels and floor panels

which are connected to each other by bolts. The wall panels are surrounded by framing

ribs and are about 250 m in height, about 90 cm. in width, and 4 cm. in thickness. The

floor panels are approximately 270 to 360 cm. in length and 90 cm, in width. Some of the

systems have collar beams between the wall and floor panels.

An outline of this structural system and typical construction procedures are shown in

Figure 4. Arrangement of the reinforcement in a typical wall panel and the details of the

wall base are shown in Figures 5 and 6, respectively.

IV-60
FUNDAMENTAL PRINCIPLES FOR THE STRUCTURAL DESIGN

The fundamental principle for the structural design of this system is to set up rigid

boxes by properly arranging wall panels and floor panels. That is;

a) To arrange seismic wall panels uniformly and effectively.

b) To connect the wall panels tightly to the collar beams or to the floor panels.

c) To connect the floor panels to each other and thus increase the rigidity of the

horizontal plane.

d) To construct a continuous footing and to integrate the whole structure effectively.

e) To arrange the connecting bolts at the connections between the structural members.

In addition to these controls, the following must be taken into account;

f) Prevent occurrence of harmful cracks in members and joints under permanent load.

g) Prevent excessive deflections and unpleasant vibration in the floor.

h) Provide ductility to the structure and non-structural members and thus insure

development of large horizontal displacements

.

i) Prevent rusting of joining metals and reinforcing bars in the precast panels.

DETAILS OF EARTHQUAKE RESISTANT DESIGN

The procedure of the eathquake resistant design is as follows;

a) Determine the minimum values of wall length that are necessary for each

floor and evaluate the maximum allowable distance between these walled

panels

b) Arrange the wall panels and floor panels uniformly using the values determined above for each direction and at each story.

c) Calculate the stresses in each member and joint and check the appropriateness of the above arrangement.

Usually design procedures are dependent on the controls given in specific design

manuals. The design earthquake load that has been adopted is the value which corresponds

to a coefficient of 0.2, as given in the Building Code of Japan. The structural behavior

of shear wall panels can be determined by a simplified method, as shown in Figure 7. In this method the height of the inflection points of the walls are determined by tests. Therefore, the effect of the bolt connection at the vertical joints, between wall panels,

is neglected (see Figure 8) . At the wall base shown in Figure 9, the cross-sectional

areas of the reinforcements, a^, and the anchor bolts, B^^, are determined from the tensile forces due to bending. The shear resistance at the wall base is determined by tests.

ESTIMATED STRUCTURAL PERFORMANCE AND PROBLEMS

TO BE STUDIED IN THE FUTURE

From investigations on the structural characteristics of existing houses and test

resiilts on structural members, the following structural performance criteria are estimated.

IV-61
i) Compressive Strength of Concrete Approximately 300 kg/cm.

ii) Shear Strength Design Value (k = 0.2) t D

3 '^4 kg/cm^

Maximum Value by Test, x 7 13 kg/cm^

u

u

2'^ 3.3

In the case of a two-story house of this system, the total weight per unit area is

1.0 1.5 t/m , and with the minimum wall length per unit area equal to 15 20 cm/m . The yield base shear coefficient is estimated as 0.4 0.8.

The following additional problems need to be studied for this type of structural

system;

a) Although the elastic rigidity of this kind of structure is high, the plastic

rigidity is remarkably low due to the additional deformation of the many joints.

Further studies are, therefore, required to clarify the effects of these joints

on earthquake response.

b) The durability of thin panel concrete shells and of exposed steel bolts should

be improved.

IV-62
0)

-l-> E

1/1 OJ >> +-»

c/) >i

,

s-

(/) o o o o o o •o -a XJ x>

rs OI

»-> o r—

zs CO

s-

+-> 1/1

+J te +-> o re c c

-o XI Q. o OJE slet

he

snoi Of tens )0lt

orce ter

bar

otte

cn (-> 4- U

ion nlt

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L

(-> din C O

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)nnecti

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u

s-o or

wel t

or h

frict

i

ored

r

shear

We1

d

i

n

r

mort

w

<_> t-> 4-) o

ch o

c o

03 Bo •a:

OI XJ

OI

lht

weight

late

concret

OJ

ncrete oncret oncret

ight

weight

aggregate

oncret

laved

foam

concrete d

Co o

re

u

re

o

X3

OJ

OI E Lie

aggrec E E u

Ik on or _i oa to u. 3

tU (/) C Jli re •<- u

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of

Membe

zed

thin

rounding do sized

pai

do do

.C.

panel

-a

•r- SOI _j c 1 ISI cn

Medium with

Lar

CO <_)

IV-63
•66 67 68 69 70 71 72

YEAR

(a) NUMBER OF PREFABRICATED

DWELLING CONSTRUCTION

i/

V i /

sy i /

/

/

/

I /

"7 /

7

/

/ /

1—

1

>^

\^

i

1

1 '66 67 68 69 70 71 72

YEAR

(b) RATIO OF PREFABRICATER

DWELLING CONSTRUCTION

TO THE TOTAL DWELLING

CONSTRUCTION

FIG. 1 YEARLY CHANGES OF PREFABRICATED DWELLING

CONSTRUCTION IN JAPAN FROfI 1966 TO 1972

IV-64
CD

CO CD

o*

0)

->> i- OJ o s_ Ol +J

=3

i/> U

O M 1- oo 3 U 4-> S- 3 1/1 e *-> S_ o I/) +-> Ol

t-> l/l +->

1/1 C Ol Ol s_ (M Ol -o a> o

I. o OJ c o o +-> o

E 3 1/1

a:

o

o

>-

] a;

<i:

<:

<_) z:

t-i o cc: 1—1 CO h-

•=x. o uLlJ DC

ttL. I/) u. o o

il CD o z:

S3snoH 000 L Ni NoiionaisNoo do b3awnN

IV-6 5
FIG. 4 MEDIUM-SIZED PRECAST PANEL CONSTRUCTION WITH COLLER BEAMS

Assembling Procedure

J L

Construction of wall panels at the second floor

J L

I

bolt

09

Construction of collor beams and floor slabs

J L

bolt

Construction of wall panels at the first floor

I r^t—± t t t t t t t t , ' I " ° I Construction of footing with anchor bolts

IV-66
FIG. 5 DETAILS OF SEISMIC WALL PANELS

IV-67

y77777.

FIG. 8 NEGLECT OF COMMECTIOM BOLTS AT VERTICAL JOINT

FIG. 9 RESISTING REINFORCEMENTS AGAINST MOMENT

IV-69
AN ANALYTICAL MODEL FOR DETERMINING ENERGY DISSIPATION

IN DYNAMICALLY LOADED STRUCTURES

by

John F. McNamara

Assistant Professor

Department of Civil Engineering

Sushi 1 K. Sharma

Research Assistant

Department of Theoretical and Applied Mechanics

University of Illinois

and

Construction Engineering Research Laboratory

Champaign, Illinois

An oinalytical procedure is developed which predicts nonlinear cyclic structural response under large reversals of plastic strains. The structure is discretized by means of the finite element approximation, and the material behavior

is simulated by a refined analytical model which describes the realistic hysteretic stress-strain curves of A36 steel under arbitrary cycles of load. In order

to test the validity of this material model, some comparisons are made with ex- perimental values of the inelastic response of a simply supported beam under cy- clic bending. The model is subsequently used in tlie dynamic analysis of a portal

frame subjected to a selected portion of the El Centro NS earthquake acceleration

record. The improved cyclic response with the current approach is illustrated by

comparing results with those obtained using a simple bilinear kinematic hardening

material approximation. Comparisons are also made with values obtained using a commercially available nonlinear frame analysis computer program. Some final com- ments are made regarding the rate of solution convergence with integration time

step size for two different temporal integration operators used in this analysis.

Key Words: Analytical model; dynamic analysis; dynamic loading; earthquake;

energy dissipation; finite element; seismic response; structural

engineering.

IV-70
INTRODUCTION

The dynamic analysis of framed steel structures under earthquake loading would appear

to be well in hand, and many studies and applicable computer programs are readily available

(1, 2, 3) . In terms of building configuration, structural elements, and applied loading,

the programs are general in nature, but all assume that material behavior can be approximated in terms of moment-rotation joint relationships. A simple bilinear or Ramberg-Osgood

(2, 4, 5) curve fit is usually adapted to model this material resistance function and appears adequate for most engineering applications. Currently accepted design criteria for

steel structures with respect to lateral displacements, ductility ratios, and lateral forces

have evolved from studies using the above computer programs. However, the complexity of

the load-displacement hysteresis loops for frame type structures, where buckling, plastic

straining, large displacements, and joint slip are active, can be judged from the survey

given in (6) of computations and experiments on model structures.

A current study presents some results obtained during the course of a continuing re- search project on the magnitude and distribution of energy dissipation in dynamically

loaded structures. A significant point of this research is that the behavior of the structure is found by means of a material model with properties derived from the exact hysteretic stress-strain behavior of A35 steel. Measurements of the cyclic behavior of A36 steel

were carried out in (8) , and the analytical model was developed and validated in (7) for

uniaxial and multiaxial stress states. By means of this model, one can incorporate the

results of advanced research in material behavior into engineering analysis. In all cases known to the writers, cyclic material behavior has been deduced from monotonic test curves

while, in fact, highly accurate results on cyclic behavior are readily available (9, 10,

11) . The development of this higher order material model supplements the current activity

in the cyclic testing of structures through large strain reversals (12)

.

Since the material resistance is synthesized from knowledge of a pointwise stressstrain relation rather than a moment-rotation relation at a joint, the finite element

method rather than a stiffness method is used to model the structure. The element used

is a simple beam-column with the stiffness matrix obtained by numerical integration along

the length and through the depth of a member element. A finite element plane stress analysis of a simply supported beam, cycled through three load reversals with large plastic

strains, was used in (7) to check the material model under discussion. Further analysis

of. the same problem, using a beam-column element and idealized material properties, is

given in (13) . The purpose of the study in (13) was an assessment of the accuracy of nu- merical solution techniques with a particular reference to cyclic loading. The refined

material model has been incorporated in a general purpose finite element computer program

(14) capable of analyzing the nonlinear static and dynamic responses of engineering structures.

The objective of this research is to formulate an analytical procedure for determining

overall structural energy dissipation properties on the basis of experimental cyclic material behavior. A projected use of the numerical results is an assessment of the effect

of nonlinear material behavior on the damping properties of structures. The preliminary

results given here are aimed at verifying the overall analytical model and the accuracy

of the numerical solution procedures. For the case of a simple portal frame under earthquake excitation, it is shown that considerable differences in displacement values occur

between the current results and those obtained using a bilinear kinmatic hardening stressstrain relation.

METHOD OF ANALYSIS

(i) Finite Element Formulation

The general equation of motion for a body, derived from the principle of virtual work,

is written in terms of initial geometry as M{q^} = - (B)^{s} dV + {P^.} (1)

IV-71
where (M) is a consistent or diagonal mass matrix, iq^, 'i-'q) are the generalized displacement and acceleration vectors. { P} is a vector of equivalent nodal forces. { S} is a vector of generalized stresses. The matrix (B) transforms generalized displacement increments at the nodes to generalized strain increments at any point in an element. It is defined by the equation

{AE} = (B){Aq} (2)

and is a nonlinear function of displacement. The expressions used in forming (B) in the

current study are given in the Appendix.

A modified, or corrected, linear incremental form (14) of Equation (1) is obtained as

(M){Aqt+^tl = / (^B)T {S} dV - / (B)'^{AS} dV + {AP^+At^ + (-(M){qJ

' V V

- /

(B)'^ {S} dV + {P^}) (3) V

The incremental stress vector {AS} can be expressed in terms of {Aq} by using Equation (2) and the following linear incremental stress-strain relation

{AS} = (Dgp){AE} (4)

The construction of the elastic-plastic stress-strain transformation matrix (Dg^) with

reference to the new material model follows the approach outlined in (15) . By siibstituting

Equation (4) in Equation (3) and following procedures described in (15) , Equation (3) can

be written as

(M){Aqt+At> = - (K^HAq^^^t} + {AP^^^J + { I^} (5)

where the tangential stiffness matrix (K^) is formed by combining the first two integrals

on the right hand side of Equation (3), and the "load correction term", { I^} , is the con- tribution of the terms inside the parenthesis in Equation (3). is the total unbalanced force at time t.

(ii) Time Integration

The controversy over an optimum temporal integration operator has produced a large

volume of studies in recent years. Direct methods (16) are available for estimating the

accuracy of these operators when applied to linear systems (17) , but it seems that only

numerical experiments can provide a measure of their worth in the nonlinear case. In a recent analysis (18) of the performance of several popular operators, namely; the Houbolt

(19) , the Wilson (20) , the Newmark Beta (21) , and the central difference operator, it was noted that the accuracy and stability of solutions with these operators were very different between corresponding linear and nonlinear applications. The overall conclusion

arrived at in (18) is that inaccuracies in nonlinear solutions accumulate very rapidly

when integration operators are used with time increments larger than that of the stability

limit given by the central difference operator.

In order to further investigate the result described above, both the central difference operator and the Newmark operator with g = 1/4 and y = 1/2 were used in the current

study. The latter operator is generally applied in earthquake engineering and is the

basis of the solution method in (3) . The incremental value of the displacement vector is

given by the central difference operator as

{Aq^} = {Aq^_^^} + At2{q^_^^.} (6)

The solution cycle is formed by calculating the displacement increment from Equation (6) and using this value to compute the integral on the right hand side of Equation (1) . The

acceleration for the next step is found by solving Equation (1) for { q^} . The operator

is not self-starting and suitable initial values are outlined in (18)

.

IV-72
Because of its iitplicit form, the Newmark operator is more conveniently combined with

a linearized equation of motion such as given by Equation (5) . The increment of acceleration is given by

{Aqt4At> =^«^qt+At^ -At{qt} -^qt>) ,

' (7)

Substitution of Equation (7) into Equation (5) yields \

+ (K^)) {Aqt+At> = {^Pt+At> + ^^t' + —^(At{4t} + ^

{gt>' (8) At At

In this approach {Aq^^^^} is first calculated from Equation (8) and then substituted back

into Equation (7) to find the acceleration increment. The velocity vector is given by

{A4t+At> = {^t> + |^(^^t> +{qt+At>) (9)

Danping forces of the equivalent viscous type are not included at present since they would

coirplicate the interpretation of the effect of material nonlinearity in the response.

(iii) Constitutive Relation and Refined Material Model

The new refined material model is an extension of the work of Martin (22) and Jhansale

(23) and is described in detail in (7, 8) . A mechanical analog of the basic series model

with only three elements is shown in Figure 1. Each spring of the system has a linear

stress-strain relationship. The springs in the parallel spring-slider elements are not

deformed until the applied stress reaches the yield stress of the slider in that element.

If the applied stress in an element reaches above the yield stress, then the difference

is stored in the element as a residual stress. The tangent modulus E^ on a given segment

of the curve is obtained from the relation

(10)

where the summation is extended over all those spring-slider elements which have yielded.

The stress-strain response with this model follows Massing' s hypothesis where the closed

hysteresis loops are the same form as the stablized initial branch OA of the stress-strain

curve except for an enlargement by a factor of two (10) . The hypothesis also implies that

after a small hysteresis loop EFGH, the loading branch will follow the path EI instead of

EJ. This is an exaitple of the "memory" effect which is created by the distribution of re- sidual stress in the elastic springs. . ....

This basic model cannot, however, accurately trace the hysteretic stress-strain curve

of A36 steel for two reasons. Firstly, A36 steel in its "stable" condition does not obey

Massing 's hypothesis. This is shown in Figure 2 where the cyclic stress-strain curve, increased by a factor of two, does not accurately describe the hysteresis loop shapes, and

the upper hysteresis loop tips do not fall along the upper trace of the largest loop.

Secondly, A36 steel cyclically hardens or softens before becoming stable. It was observed

in (23) and (7) that even during the hardening and softening process each upper branch of

stress-strain curve (except for the initial flat top branch) could be fitted to the double

skeleton curve for the "stable" material by translating the loop along the elastic line.

The loop is translated along the elastic line until the upper loop tip is approximately

tangent to the double skeleton curve. An outer trace of a given loop is then described in terms of this skeleton curve and an appropriate stress offset, S^g, which defines the

amount that the loop must be displaced along the elastic line. In Figure 3 the individual

branches of hysteresis loops obtained from constant amplitude strain controlled tests have

been fitted to the double skeleton curve. As shown in the figure, the stress offset, S^^,

depends on the number of reversals and the total strain range. In this study S^g is approximated by a family of linear functions of the number of reversals and the total strain

range.

IV-73
The analytical material model also incorporates cycle-dependent mean stress as described in (7) , but this feature is not activated in the present study. An added feature

is that the initial response of the model is determined by a set of constants derived from

the "flat-top" stress-strain curve obtained by testing a virgin specimen of A36 steel. The

spring constants and slider stress values for subsequent reversals after the first loading

are found separately from the skeleton curve obtained from constant amplitude strain con- trolled cyclic tests of steel specimens. The procedure, used to compute the spring con- stants and slider stress values, is described in detail in Appendix A of (7). Agreement

between the refined material model and a complicated cyclic stress-strain history for a virgin steel is shown to be highly satisfactory by the results depicted in Figure 4,

NUMERICAL RESULTS

(i) Verification of Material Model in Structural Application

In order to test the validity of the material model in structural applications, the

beam problem, described in Figure 5, was developed. Ten spring-slider elements were used

in the material simulation, and the dimensions of the beam were chosen in order to obtain

large plastic strains without introducing nonlinear geometric effects. The beam was loaded

and reloaded through three cycles, as shown in Figure 5. In the earlier investigation of

Plummer (7) , the beam half-span was modeled by fifteen plane stress eight-node isoparametric

finite elements (24) . With a view to later, more general frame-type structural applications, the half-span was modeled with ten simple beam-column elements in the present study.

The generalized strain-displacement relations on which the beam-column element is based

are outlined in the Appendix.

The numerical and experimental results for the beam are compared in Figure 6. The , geometric shapes of both curves are so clearly similar that it appears that a slight ex- 1 tension of the elastic range would allow perfect agreement. The dimensions of the beam

span, cross-section, and concentrated loading also make it likely that shear deformations

contribute in part to the difference in results. This is investigated in Figure 7 where i the results of the plane stress analysis from (7) are compared with the current beamj column results. The continual redistribution of the two-dimensional stress state in the

plane stress results means that, even for zero hardening, the load will continue to in- crease after the critical mid-section has become totally plastic. In a similar example,

Felippa (25) found a 35 percent increase in load capacity beyond the limit load in simple

bending at a displacement of approximately five times the elastic limit value. The effect in this case is to give the plane stress curve a greater slope on the initial branch

which leads to high residual stresses in the spring elements before the first reversal.

The residual stresses for the beam-column case are practically zero except for one element,

and this means a flatter response will occur on the subsequent reversal. It is felt that

this effect rather than shear displacement causes most of the difference in results in

Figure 7 at the end of the second reversal. .

These preliminary results indicate that further tuning of the analytical material

model may be in order. Current suggestions in this respect are to drop the initial monotonic "flat-top" behavior and use the cyclic skeleton curve for calculating one set of

spring and slider values for all cases. Omission of the "flat-top" first cycle may result

in an overall improved response. It was suggested in Appendix A of (7) that fictitious

residual stresses be built in at the end of the "flat- top" cycle, but this may overly com- plicate the parameters of the model. Another modification under study concerns the fact

that in the current model, the effect of the increase in elastic range, described as a stress offset in Figure 3, is not added to the first slider, or first segment of the stress- strain curve. It was commented on earlier that the experimental values appear to indicate

that an extended elastic range is necessary for improved agreement, and the relaxation of

this restriction on the first slider may be helpful.

IV-74
(ii) Response of a Portal Frame to Seismic Loading

The portal frame described in (26) was modeled with three finite elements per member

and subjected to a selected four seconds (1.5 sec. - 5.5 sec.) of the El Centre NS earthquake acceleration record magnified by a factor of 1.5. The equations of motion must be

adjusted to reflect this type of loading and Equation (8) is replaced by

(iJML + (K^)) {Au.^.} = (AP.^. .} + {I,} + i(M).(At {uj + ^'u J - ^

{'qJ )

(11)

where is now the generalized displacement relative to base of the structure. The

vector { Q^} is formed by the components of base acceleration input. The equilibrium cor- rection term is also changed to

{I^.} = - (M)({ut> + {Qt>) - / (B)'^{S} -:v + {P^} (12)

V

For the solution with the central difference operator, Equation (1) is solved as it is

written for the absolute acceleration, and the relative structural acceleration is obtained

from the relation

{ut> =

^qt^ -

{"Qt> (13)

Displacements relative to the base, for the purposes of calculating internal forces, are found in the usual manner as

{Au^} = Uu^_^^} + At2{u^_^^} (14)

The stiffness matrix (K^) is generated numerically using three Gaussian points along

the length, and the cross-sections of the I-beams are modeled by three specially weighted

integration points, one placed in each flange and one at the center. The number of points

in all cases can be varied, and an option also exists which allows stresses to be computed

at one or all Gaussian integration points. A diagonal mass matrix is formed in the program by collapsing all rows of the consistently formed matrix onto the diagonal. The diagonal mass matrix is used with Equation (1) and the central difference operator, and the

consistent mass matrix is used in Equation (11) since there is no apparent computational

advantage in using the diagonal form.

The dotted line in Figure 8 indicates the response of the frame with the new material

model and the central difference operator. In the same figure, the solid line passing

through the triangles represents the response with kinematic hardening idealized by a simple bilinear stress-strain curve given by E = 30 x 10^ psk, a = 36 x lO-^ psi, and E^/E = 0.1. A comparison of these two curves indicate the influence^of the added cyclic hardening

effect in. the refined material model over the idealized behavior. The results clearly

demonstrate a cumulative effect as the divergence between the results increases with each

reversal of the response. It is eitphasized that the model structure is a simple portal

frame loaded with only a fraction of a typical earthquake input, and greater differences

can be expected in the analysis of more complex framed structures under an extended earthquake input.

The solid line passing through the circles in Figure 8 shows the values given by

DRAIN-2D which is a standard stiffness-type program developed by Kanaan and Powell (3)

.

The material response is based on a bilinear moment-rotation joint, or node, relationship,

and the moment-axial force interaction diagram with shape code 2 (3) was adopted. Each

frame member was divided into three equal elements, and the elastic constants and hardening modulus mentioned above were employed in the analysis. Inspection of Figure 8 shows

that the results with DRAIN-2D agree in general with those found using bilinear kinematic

hardening. The divergence of these curves at the final peak values in the response may

be attributed to the different yield criteria and to the manner in which the structural

stiffnesses are computed in the respective computer programs. The agreement between

IV-75
DRAIN-2D and the results of the refined material model at the final peak displacement ap- pears to be fortuitous since the displacement histories are considerbly different up to this point. i

(iii) Selection of Time Increment for the Temporal Integration Operators

For linear problems, the choice of an implicit versus explicit integration operator,

such as the Newmark over the central difference operator, is often governed by the fact

that the latter is only conditionally stable with respect to time increment size. Also,

accurate solutions with comparatively large time increments can be obtained from Newmark 's operator with B = 1/4 since the amplitude of each mode is conserved (21) . However, the

results obtained in (18) for a geometrically nonlinear simple beam problem showed that

all implicit methods suffered serious inaccuracies for time increments greater than that

found to give stable results with the central difference operator, and that the Newmark

method was the worst in this respect.

In the current study the portal frame, described in Figure 8, was subjected to a sinusoidal base excitation in order to find suitable integration time-steps for the central difference and the Newmark operator. The response was obtained with the refined

material model as well as with DRAIN-2D which also uses the Newmark operator, and the re- sults are shown in Figure 9. The limiting stable time increment size for the central

difference operator was found to be 0.0025 seconds, and an apparently stable response was obtained with the Newmark operator for a time step of 0.00625. Figure 9 indicates clearly

the problem of identifying between stable but not necessarily accurate solutions. In this case all results were in agreement up until a time of 0.45seoonds when both DRAIN-2D and

the finite element solution with Newmark ' s method begin to diverge from the central difference solution. Therefore, it was decided to assess this phenomenon more accurately by

running the problem described in Figure 8 with decreasing time increments until convergence! was achieved for the response over the time of the earthquake input. The results obtained i with DRAIN-2D for increments of 0.005 and 0.0025 are shown in Figure 10. No change oc- curred with a reduction in step size to .00124 and so convergence is assumed for At equal

to .0025. The problem is now evident in that apparent convergence occurs over the first

cycle or so with subsequent slow, but finally appreciably divergence after approximately

two seconds of the response. The recommendation is made to use the stable central difference time increment in that the seemingly apparent stability and delayed appearance of

gross inaccuracies with implicit operators could lead one to believe that one had a con- vergent solution at larger time steps.

The results under discussion in this section, with the exception of the central difference case, were all obtained using the incremental form of the equations of motion with

an elementary equilibrium, or load, correction term. It can be argued that an iterative

solution at each step would improve the accuracy of the implicit solution schemes. This was attempted in (18) , and for the problem studied, the iterated scheme made a difference

only at those time increments which had already caused large errors, and no overall significant improvement was apparent. Further study in this respect is necessary, but the

economic aspect of an explicit central difference versus an iterated implicit solution

may be the deciding factor in favor of the former method. An added consideration is the

simplicity of central differnce solution scheme and its considerable advantage with re- spect to computer storage demand over the other methods.

CONCLUSIONS

An analytical procedure has been described which incorporates constitutive relations

based on a higher order material model. A comparison of analytical and experimental re- sults for a virgin A36 steel coupon and for a simply supported beam has shown the capability of the refined model to accurately predict the cyclic hysteretic material behavior.

The responses of a portal frame subjected to the El Centre NS earthquake acceleration

record have been compared for the refined material model, a bilinear kinematic hardening

model, and the DRAIN-2D model. The significant differences in the computed results have

IV-76
clearly demonstrated the effect of cyclic hardening on seismic structural response, and

warrant a further study of this phenomenon with respect to energy dissipation and fatigue

life of steel structures. The preliminary dynamic results suggest that for comparable

accuracy, the time increments for implicit integration operators and for the central difference operator must be of the same order in the case of nonlinear problems with complex

time-motion histories.

ACKNOWLEDGMENT

The writers express their thanks to Dr. F.B. Plummer, Esso Production Research Company,

Houston, Texas, for his assistance in inserting his computer model of material behavior in- to their program.

REFERENCES

1. Rea, D. , Bouwkamp, J.G. , and Clough, R.W. , "The Dynamic Behavior of Steel Frame and

Truss Buildings," American Iron and Steel Institute , Bull. No. 9, April 1968.

2. Goel, S.C., "Response of Multistory Steel Frames to Earthquake Forces," American Iron

and Steel Institute , Bull. No. 12, November 1968.

3. Kanaan, A.E., and Powell, G.H. , "DRAIN-2D. .. Inelastic Dynamic Response of Plane

Structures," NISEE/Computer Applications , University of California, Berkeley, California, 197 3.

4. Romberg, W. , and Osgood, W.R., "Description of Stress-Strain Curves by Three Parameters," Technical Note 902, NACA, July 1943.

5. Jennings, P.C., "Periodic Response of a General Yielding Structure," Journal of the

Engineering Mechanics Division , ASCE, Vol. 90, No. EM2, April 1964, pp. 131-166.

6. Wakabayashi, M. , "Frames Under Strong Impulsive, Wind or Seismic Loading," Proceedings of the International Conference on Planning and Design of Tall Buildings , Lehigh

University, Bethlehem, Pennsylvania, 21-26 August 1972.

7. Plummer, F.B., "A New Look at Structural Energy Dissipation," Technical Report M-82,

Construction Engineering Research Laboratory, Champaign, Illinois, 1974.

8. Martin, John F., "Cyclic Mechanical Tests and an Appropriate Analytical Stress-Strain

Model for A36 Steel," Technical Report M-86, Construction Engineering Research Laboratory, Champaign, Illinois, 1974.

9. Landgraf, R.W. , Morrow, J., and Endo, T., "Determination of the Cyclic Stress-Strain

Curve," Journal of Materials , Vol. 4, No. 1, March 1969, pp. 176-188.

10. Morrow, J., "Cyclic Plastic Strain Energy and Fatigue of Metals," Internal Friction , Damping, and Cyclic Plasticity , ASTM STP 378, American Society for Testing and Materials, 1965, pp. 45-87.

11. Iwan, W.D., "On a Class of Models for the Yielding Behavior of Continuous and Composite Systems," Journal of Applied Mechanics , Vol. 34, No. 3, September 1967, pp. 612-617.

12. Popov, E.P., and Benter, V.V. , "Cyclic Loading of Steel Beams and Connections," Journal of the Structural Division , ASCE, Vol. 99, No. ST6, June 1973, pp. 1189-1204.

13. McNamara, J.F., "Numerical Solution Techniques for Highly Nonlinear Static Structural

Behavior," Proceedings of the 1974 Army Numerical Analysis Conference , Philadelphia,

Pennsylvania, 12-14 February 1974.

iy-77
14. McNamara, J.F., and Marcal, P.V., "Incremental Stiffness Method for Finite Element

Analysis of the Nonlinear Dynamic Problem, " Int. Symposium on Numerical and Computer

Methods in Structural Mechanics , Urbana, Illinois, September 1971.

15. Marcal, P.V., "Finite Element Analysis of Combined Problems of Nonlinear Material and

Geometric Behavior," Division of Engineering, Brown University, Report N00014-007/2

,

June 1969.

16. Nickel, R.E. , "Direct Integration Methods in Structural Dynamics," Journal of Engineering Mechanics Division , ASCE, V. 99, No. EM2 , April 1973, pp. 303-317.

17. Goudreau, G.L. , and Taylor, R.L., "Evaluation of Numerical Integration Methods in

Elastodynamics , " Computer Methods in Applied Mechanics and Engineering , Vol. 2, 1972,

pp. 69-97.

18. McNamara, J.F. , "Solution Schemes for Problems of Nonlinear Dynamics," to be published.

Journal of Pressure Vessel Technology , ASME

.

19. Houbolt, J.C., "A Recurrence Matrix Solution for the Dynamic Response of Elastic Aircraft," Journal of Aeronautical Sciences , Vol. 17, 1950, pp. 540-550.

20. Bathe, J.E., and Wilson, E.L. , "Stability and Accuracy Analysis of Direct Integration

Methqds , " International Journal of Earthquake Engineering and Structural Dynamics , Vol. 1, 1973, pp. 283-291.

21. Newmark , N.M. , "A Method of Computation for Structural Dynamics," Journal of the Engineering Mechanics Division , ASCE, Vol. 85, No. EM3, July 1959, pp. 67-94.

22. Martin, J.F. , Topper, T.H., and Sinclair, G.M., "Computer Based Simulation of Cyclic

Stress-Strain Behavior with Applications to Fatigue," Materials Research and Standards , ASTM, Vol. 11, No. 2, February 1971, pp. 23-29.

23. Jhansale, H.R. and Topper, T.H. , "An Engineering Analysis of the Inelastic Stress Re- sponse of a Structural Metal Under Variable Cyclic Strains," Cyclic Stress-Strain

Behavior—Analysis, Experimentation and Failure , ASTM STP 519, 1973, pp. 246-270.

24. Zienhievicz, O.C., The Finite Element Method in Engineering Science , McGraw-Hill Publishing Company Limited, London, 1971.

25. Felippa, C.A. , "Refined Finite Element Analysis of Linear and Nonlinear Two-Dimensional

Structures," Ph.D. Thesis, University of California, Berkeley, California, 1966.

26. Toridis, T.G. , and Khozeimch, K., "In elastic Response of Frames to Dynamic Loads,"

Journal of the Engineering Mechanics Division , ASCE, Vol. 97, No. EM3, June 1971,

pp. 847-863.

IV- 7 8
APPENDIX

Beam-Column Finite Element

The axial and normal displacments U and V at a point (X, Y) of the beam are approximated by a linear and a cubic interpolation function of X, respectively.

U = aQ + a]_X

V = bg + h-^X + bjX^ + b^X-^

where a^ , a^, bQ,...., b^ are generalized displacement coefficients.

It is convenient to separate the strain E at (X, Y) into its membrane and bending

coiiponents E^^^ and E^, where

(A-1)

(A-2)

E„ = ^ + 1(^)2 + 1(^)2

dX dX dX

^b

= d^V

dz2

(A-3)

From Equations (A-1) and (A-2)

dU/dX = a-j^

dV/dX = bj^ + 2b2X + 3b ^X^

An increment of strain is obtained from Equation (A-3) as

Ae = AdU + dU (AdU) + dV (AdV)

dX dX dX dX dX

2 AEj^ = - Y(A2_Z)

dX^

(A-4)

(A-5)

Substituting Equation (A-4) into (A-5) , Ae^ and AEj^ can be expressed in terms of in- crements of generalized displacements as

AE„

AE^

1 + dU

dX

0 ^

dX

0 0

2X^

dX

- 2Y

3X2 dV

dX

6XY

Ab,

Ab,

(A-6)

Ab,

Ab,

From Equation (A-6) , matrix (B) which relates generalized incremental strains with

generalized incremental displacements is readily identified as

CB) =

1 + ^

dX

dV 2 dX

0

fdV dX

2Y

3X2 dV

dX

- 6XY

(A-7)

IV- 7 9

IV-81
100

BO

CO

DOUBLED SKELETON CURVr

4 -20

1

note: NUh':nF.RS DESiGKATE REVERSALS;

G //jDICATES BlCINNING or h'EVERSALS

20 -I

FROM REFERENCE 7

3 —<p

(I OS

I 101

J. 0.010 0.020

ENGINEERING STRAIN

0.030

FIGURE 3. HYSTERESIS LOOP BRANCHES FITTED TO

THE SKELETON CURVE

IV-82
IV-83

FIGURE 6. Cpjyl PARI SON.__0r BEAM FINITE E LEMENT RESULTS

WIT H _ EXP E R ! h'. E N T A L VA L U E S

IV-85
IV-186
IV-87
IV-88
IV-89
DESIGN OF PILE FOUNDATIONS SUBJECTED TO LATERAL LOADS

by

Mitswru Nagao

Tadayoshi Okubo

Keiichi Komada

and

Asao Yamakawa

Public Works Research Institute

Ministry of Construction

Japan

This report presents a number of basic items relative to the design of pile

foundations subjected to lateral loads, and deals with the present design status of highway bridges in Japan. Also presented are items to be further studied for

the design standardization of pile foundations subjected to lateral loads. Such

problems as the deformation mechanism of group-pile structures are examined.

This report presents the method by reinforcement of the pile-head.

Key Words: Design; earthquake; highway bridges; lateral loads; piles; pile head;

structural engineering.

IV-90
INTRODUCTION

Because Japan is subjected to frequent earthquakes, it is necessary to consider the

effects of these earthquakes in designing structures. Such design specifications are stipulated under Paragraph 1-8-1, "Kinds of Load" in the Volume I of the "Common Rules" of

Japan's Highway Bridge Specification (Technical Standards of Ministry of Construction),

and also in Volume V of the specification given under the title "Specification for Earthquake Resistant Design."

The design of sub-structures of highway bridges is performed in compliance with the

Specification's Volume IV, "Specifications for Sub-Structure Design," The provisions relating to the "Design of Pile Foundations" in this bridge specification were established

in 1964 and have subsequently become rather obsolete. Jtoendment and revision to these specifications are now underway in order to incorporate the latest technology and field and

laboratory data.

This report will present various ideas and methods outlined in several studies, which

relate to the design of pile foundations subjected to lateral or seismic loads. Although

these studies are still in progress, various parts are being incorporated into the provisions of the "Highway Bridge Specification." Consequently, the text of this report will

indicate the new provisions of the Japan Highway Bridge Design Code.

The general contents of this report will be presented as follows.

In the section "Design of Piles," there will be introduced a conventional method and

a rigorous method of design. The former method evaluates the pile's head reaction and

displacement by balancing the three forces, while the latter considers the displacements.

For most foundations, the simple conventional method is sufficient, however, this method

may cause, in some cases, some serious errors. Therefore, in these instances, a suggestion

for applying the rigorous method is given. In general, the use of these methods considers

the piles to be long. In the case where the piles are short, the displacement method must

be revised, a description of which will be given.

For the section "Calculation of the Lateral Coefficient of the Ground Reaction," there

is introduced a method as deduced from the results of soil tests, the lateral coefficient

of ground reaction which can be applied to the analysis of the stress and displacement of

the pile, with the aid of the theory of beams on an elastic foundation.

This method was proposed on the basis of experiments performed on 150 lateral load

tests and their correlation between the soil properties.

In the "Design of Pile Head" section, a design method for connecting the pile's head

with footing is described. There is no practicable design method to rigidly connect the

head to the footing. But the Public Works Research Institute of the Ministry of Construction, desiring a strong and rigid connection, issued a design standard. In order to test

and verify the appropriateness of this standard, a series of nearly full-size scale load

tests were conducted. A description is given of the basic concept of the design process

of the connection of a pile to the footing and the results of the experiment.

The section entitled "Behavior of Pile Foundation with Free Length" describes the

lateral resistance of a pile foundation having a free length and the results obtained from

a vibration test.

There have been various vibration tests conducted on pile foundations, in which the

footings are buried in the ground. From these vibration tests and through experience obtained during actual earthquakes, the earthquake resistance of such units have been verified. However, there have been minimal experiments londertaken with respect to pile foundations of free length since it is entirely a new type of foundation concept. This sec- tion will, therefore, present the results of an experiment on nearly full-sized scale

foundation, as performed by the Public Works Research Institute.

IV-91
DESIGN OF PILE

Calculation Method of Pile Head Reaction and Displacement

There are two methods designated as the "Rigorous Method" and the "Conventional

Method", which can be used to determine the reaction and displacement which may occur at the head section of each pile due to external forces (V^, H^, M^) , as shown in Figure 1. It will be assumed that the footing is a rigid body and the ground is elastic, both in

the axial and the lateral directions of the pile. The calculation method presented is ap- plicable only in the case where the piles are sufficiently long and are deemed to have an infinite length.

The applicable extent of this calculation method and the calculation method for short

piles are presented in the following paragraphs.

(1) Rigorous Method

The following equations, when solved simultaneously, will give the deformations

of the footing-pile unit, where the lateral displacement of the footing is 6^, ver- tical displacement is 6^, and the rotary angle is a.

^xx^x + ^ V" = "o A 6 + A 6 + A a = V

'

' (1) yx X yy y ya o Afi^ + A6+Aa = M ax X oty y aa o , where

;

A = y(K, cos^ 6. + K„ sin^ 9.) XX 1 IV 1

1

A = A. = y (K - K, ) sin 9 . cos e . xy yx V 1 1 1

1 V = \x = ^^v -

'^l) ^1 ^^"^ \ ^i - ^2 ^i^

i V = Si ^1 ^i'^^ 9i)

A = A , = y{ (K cos^ 9. + K, sin^ 9,.) x. + sin 9.} ya ay^' v il ii.^ i

1

A = y{ (K cos^ e. + K, sin^ 9-) x. ^ + (K„ + K ) x. sin 9. + K.

1

K = 12 EI e"^ (1 + gh)3 + 2

1 + 6h

^2 = ^3 = ^1

26

K = 4 EI g(l + Bh)-^ + 0.5

^ 1 + ghd + Bh)3 + 2

3 = /KfjD/4 EI

D = Pile diameter (m)
3 EI = Pile's bending rigidity (kg/cm )

= Lateral coefficient of the ground reaction for a pile (kg/cm"^)

K = Axial coefficient of the ground reaction of a pile (kg/cm)

When 6^, 6^, and a are evaluated from the Equation (1)/ the pile's axial force

Pn. working upon each pile head, lateral force 'Pn^, and the pile head moment Mj_ can

be calculated from the following equations;

=

®i + (^y + ox^) cos e^lj "...

'

Hi " V'^x ""^^ ®i " ^*y ^ sin e^"j - K^a '

' P " (2)

= - K,{5 cos 6. - (5„ + a^.) sin 6-1 + K.a 1 3 x 1 yxi' 1)4

(2) Conventional Method

As an expedient design method, it is possible to perform an approximate analysis

by assuming the following;

i) In the case of only vertical piles:

p„ = — +

i Ho

"i n

(3)

6 X

where

;

^o + ^"o

Kv I ^i^

i

A = i(h + 1/e)

2 ii) In the case of skew piles

In reference to Figure 1, the following relationships are obtained;

^o "o + >^Ho V.In= + X

V .2

I -i^

1

= tan 9^ + (Hq - Jv^ tan 9^)

^sec 9^ i

IV-93
= COS °i + sin 9^

= - ^P„, (4)

6 X

6 y

I (K, cos2 9 . + K , sin^ 6 .

)

_ ± 1 V 1

o

M + ^H^

o o

2

1

(3) Limiting Application of the Conventional Method

When the pile foundation is constructed with only vertical piles and has a syiti' metrical arrangement, both the rigorous and conventional methods agree for both the

pile-head reaction and lateral displacement when = . In the case when = 0,

the behavior of each pile is the same as that which occurs in a single pile. An actual pile foundation exists between these two states, therefore, consider first a non-dimensional parameter as defined by the following expression, which relates the

rigidity ratio in the vertical direction to the horizontal direction, and is expressed

as;

2 x.

K I X.2 nEIB_

i 1+3 h

(0 <_ $ ± 1) , (5)

where, n = Number of piles.

Using this parameter the solution (defined as f) by the rigorous method for

both the pile-head reaction and displacement can be simply expressed by the following;

(K^ =0) (K^ = CO) (K^ = 0) where f^^^ =0) solution when = 0, and is the calculated value of a single

pile. f ,„ = oo) solution when IC^ = °°, and is the calculated value by the conV

°° ventional method. Namely, the parameter $ approaches 1 as the vertical direction

rigidity of an entire pile foundation increases relative to the lateral direction

rigidity. Consequently, the gap between the rigorous method and the conventional

method narrows. From the results of a preliminary calculation of various types of

foundations, it has been ascertained that if $ > 0.95, the gap between the two calculation methods is less than twenty percent.

In the case where skew piles are involved, it is difficult to obtain a sinple

parameter, as mentioned above, but in general when the vertical direction rigidity

is small and the oblique angle is substantially great, the error arising from the j conventional method will be large.

Design of Short Piles

In general, in the design of piles, the pile is considered as a beam on an elastic | foundation. The pile's displacement y in the model is presented in Figure 2 and can be

IV-94
expressed by the following fourth order differential equation,

,4 EI ^+ KjjDy =0 (6) dx

Solving this equation including the boundary condition, the pile head displacement

rate at the time when the lateral force effects the pile head can be expressed as;

2 6Hq

6 = 1 • (7) a %D

where 2 BH^/K^D is the pile head displacement rate for the case when the pile length is

infinite. The coefficient a is a correction coefficient for the pile because it has a limited length, while &l alone is esqjressed by a function. The pile head displacement

rate becomes 6 = 4H^/KjjD£ because the coefficient a in the Equation (7) is a = 3S-/2, when

the pile is regarded as a rigid body.

Figure 3 represents the relation between a and &l , and it can be seen from this plot

that if 6£ > 3, a ^ 1, which is close to the solution of an infinite log beam. If 3£ < 1, a = &1/2, which is close to the solution of a rigid body.

The above description relates to only piles displaced under the influence of a lateral

force. However, similar trends will occur for other loadings. It is, therefore, possible

to make the following classification based on the design principle: ^ 3 Elastic body, infinite length

1 < ex. < 3 Elastic body, limited length

e£ 4 1 Rigid body

A pile falls in-between the "rigid body" region and the "elastic limited length" re- gion and is, therefore, a "short pile," and, therefore, the boundary conditions at the

end-point of the pile must be taken into consideration.

Figure 4 shows the details of a lateral load test on a 'I'2m, Jll7m PC well, which was performed. In this case, with S>Z = 2, the pile belongs to the region of "short piles,"

and thus, its end-point is driven 5 m deep into the gravel soil.

Figure 5 represents the comparative values as derived by calculating the measured

oblique angle distribution and three possible boundary conditions at the pile's end-point

(free, hinged and fixed). As shown in this case, when the end-point is driven into the

gravel layer, the calculated values and the measured values agree fairly well when the

end-point is assumed as a hinge.

Figure 6 illustrates the influence of the support conditions, at the end-point of

the pile, upon the flex distribution of the pile and the bending moment distribution.

LATERAL COEFFICIENT OF GROUND REACTION (K^^)

The most important factor to control the pile's lateral resistance is the lateral re- sistance of the ground itself. The load-deformation characteristics of the ground are generally nonlinear, and moreover, when a load is applied by or through such a flexible

body as a pile, it is extremely difficult to determine a definite yield-point. This is

because there is a breakdown which takes place at the upper section of the pile as the

load is applied. Because of this, in designing, the ground is assumed to be elastic by

taking into consideration the precise data of the ground survey, and the pile is deemed

as a beam supported by this elastic body. Thus, the design load is determined from the

IV-95
allowable displacement rate and the pile's permissible stress. In this case, as the ground

is nonlinear, the standard ground surface displacement rate is determined and a virtual

free spring constant rate is enployed. From this comes the lateral coefficient of the

ground reaction Kj^. In order to determine the reliability of this term, a number of experi

ments have been undertaken.

First, in order to investigate the effect the loading plate size has upon the lateral

coefficient of the ground reaction, test pits were made in sandy and loam soil. The plates were circular and had 30 cm., 60 cm., 90 cm. and 120 cm. diameters, respectively. The

series of horizontal plate loading tests that were conducted are shown in Figure 7. This

figure is expressed as a ratio of Kj^ to K^q, where K3Q is the value when D = 30 cm. It was found that decreased in an inverse proportion to the loading plane diameter raised

to 3/4 power, regardless of the ground depth or pile buried depth, as given by;

K30 30 = ..^ . . ,,. (8)

The virtual deformation coefficient was obtained from the plate loading tests by using

the 30 cm loading plate and assuming the K-jq value as the control. In order to examine the

inter-relation in the various ground investigation methods, an experiment was conducted

for against the deformation coefficient Ep, as obtained by a boring hole circular loading test. The deformation coefficient E was obtained by either mono-axial or tri-axial

tests by using a sample of the materials and the standard penetration test value N. The

results are shown in Figures 8 and 9, resulting in an approximate experimental expression

as shown below;

E^Q = 4Ep = 4E^ = 28N (9)

By adding a proportional constant which is experimentally determined from a pile loading

test, a relationship between Equations (8) and (9) can be obtained, as given in the following expression; this proposed expression gives an estimate of the lateral coefficient

Kjj of the ground reaction for a pile.

^30 Ep ^c

= 7N

= = il— (10)

4.8d3/4 i.2d3/4 1. 20^/4 1. 20^/4

Figure 10 shows the conparison between the results obtained from a single pile loading

test performed on loam soil and the expression (8) . Figure 11 illustrates the relation

between 7N/K and the pile's diameter (D) as based on the expression (10). The data that

has been plotted was obtained by collecting the information on 150 horizontal loading pile

tests throughout Japan. The value of that was used was the ground surface displacement

at 10 mm. For the standard penetration test value N, the mean value at a depth from the

ground surface equal to 1/3, where (3 = EI) was enployed.

DESIGN OF PILE HEAD

Since the load applied to the superstructure or the substructure of a pile must be

transferred without failure to the foundation structure, the connection between the pile

body and the footing structure is extremely iirportant. In the case of a pile foundation

where the cross-section has a portion which will suddenly change, a smooth transfer of

load is imperative.

At present, the substructure design of a highway bridge requires the connection between the piles and the footing be considered as a rigid unit by using either of the following two methods

:

IV-96
Method A : Piles which are buried in a footing at a certain depth, the combined struc

ture shall resist the pile head confining moment using the buried portion

of the pile. This method is applicable to steel pipe piles, PC, and RC piles. The details of the connection are shown in Figures 12(1) and 12(2)

Method B : The pile head confining moment shall be resisted mainly by reinforcing of

the piles, even if the length of the buried pile into the footing is short

This method is applicable to any steel pipe pile, PC pile, RC pile, or cast in-place reinforced concrete pile. The methods shown in Figure 12(3) to 12(5) pertain to this method.

The design method is specified by each type of pile and by the connection method A

or B. For example, the design of a steel pipe pile, which is connected by using Method

B, is outlined as follows:

(1) The connection of the pile and the footing is designed as a rigid structure,

and the pile's head should be designed to resist all punching, pull-out, and

lateral forces and the moment.

(2) The length of pile head to be buried into the footing shall be 10 cm.

(3) The vertical bearing stress of the footing concrete shall be (refer to Figure

13) :

o = —— <_ o

IL D

4

(4) The lateral bearing stress of the footing concrete shall be:

0 ^ = IL < a • • ' - •

' ch = ca

(5) The pull-out shear stress of the footing concrete shall be:

TT (D + h;i^) h-]_

(6) The coverplate ajid the cross reinforced plate shall be:

tj^ = t2 = 22 mm h2 = 30 mm ,

which are applicable to piles of not more than 1.0 m. in diameter,

where: .

t-j^ = Thickness of the coverplate (cm) ,"

t2 = Thickness of the cross reinforced plate (cm)

h2 = Height of the cross reinforced plate (cm)

(7) Hypothetical reinforced concrete cross section stress. A punching force P, the

moment M, or the pull-out force P^ is assumed to act on a cross-section of the

reinforced concrete and the stress of concrete and the reinforcement is then ex' amined.

IV-97
(8) Anchoring of reinforcement

i) The shearing stress at the welded section shall be;

OAsa"st ^ ^ 1.4 XI "

o where

= Allowable tensile stress of the reinforcement (kg/cm )

Ag^ = Cross-section of the reinforcement (cm.^)

X = Leg length of the fillet weld (cm)

= Weld length (cm) -

ii) The anchor length of the reinforcement shall be;

"^sa^st

where

^o

=

T^^ = Allowable bond stress of concrete (kg/cm^)

u = Hoop length of the reinforcement (cm)

(9) The lateral pull-out shearing stress at the end of the footing will be studied.

The results of model tests on steel pipe piles which are connected by Method B will

now be given; the details of the test model are given in Figure 14.

The detail of the pile head connection is shown in Figure 15, which has a 10 cm. built-in and is reinforced with eight steel bars of 16 mm. diameter. On the pile's head

there is provided a steel coverplate of 15 mm. thick and 16 mm. reinforcement ring. This

ring is provided around the steel pipe pile to resist the circumferential tensile stress.

The maximum applied vertical load was 90 tons and the resulting stress distribution

on the concrete footing and the pile head coverplate under this maximum load is shown in

Figure 16. The comparison between these measured stresses and the calculated values are given in Figure 17. The calculated coverplate stress can be obtained by assuming the en- tire bearing stress acts on the coverplate or;

a = 15- 0 (D/t)2

320

where : . = Bearing stress (kg/cm )

D = Pile diameter (cm)

t = Pile thicki.ess (cm)

It should be noted that the transfer of the vertical load acting upon the pile actually

is concentrated along the wall thickness, and, therefore, the actual measured value of

stress on the coverplate is smaller than the calculated value.

The vertical load was then removed and a maximum lateral load of 70 tons was then

applied. Figure 18 shows the comparison between the actual measured stresses and the

IV-98
calculated values with respect to piles and P^. These results indicate that there exI ists a definite gap in the behavior of P-j^ on the punching side and P^ on the pull-out side.

It is known from the bearing stress distribution that the entire area of P-^ is compression,

I while on P2 there occ\irs some rotation at the built-in section. This clearly means that

I the confinement on the part of P. is rather weak and the pile head is not perfectly fixed,

j ^ In general, the established pile head connecting method has a sufficient design

i strength, and the stress which occurs inside the footing is safe.

BEHAVIOR OF PILE FOUNDATION WITH FREE LENGTH

I _ The multiple-pillar foundation of a rigid structure can be formed by groups of large

i diameter columns or pillars which penetrate and are embedded into a designated bearing

i strata with their head section rigidly held together by a top coverplate. This type of

foundation excels when working in water and can be constructed very rapidly. Even in

1 large-scale projects, such as those highway bridges which link the mainland and Shikoku

Island, this type of foundation structure was used. For the purpose of examining the lateral resistance characteristics and behaviors of the pile groups with free lengths, a series

of experimental large-size models have been undertaken. The results are given as follows.

As shown in Figure 19, two models having different top coverplate thicknesses were

fabricated. Various types of tests were conducted using these models and included a lateral loading test, a free behavior test, and a forced vibration test (the top coverplate)

.

I The symbols A, B, cind C indicate the location of the applied load and the loading direction.

Figure 20 illustrates the relation between the average lateral coefficient of the

ground reaction (Kj^) and the ground surface displacment, as obtained from the lateral loading test. Due to the group-pile effect, declined more than in the case of a single

\ pile, and it also fluctuated with the loading position. Figure 21 shows the fluctuation

of as a ratio of the central pile, which was set as the standard. The front line of

j the piles took 50% more of this ratio than that of other piles. Figure 22 represents the

torsion-spring constant for the case of an eccentric loading (loading postion B) . In this

,, case, the pile's torsional resistance took 80% of the entire applied torque, while only

',' 20% was absorbed by the lateral ground reaction. With the multiple-pillar type foundation

I having a lower rigidity, it -becomes necessary to check its torsion rate.

Figure 23 shows the acceleration mode, free vibration at the tower-end postion (C)

.

Table 1 and 2 present the results of the inherent vibration frequency and the damping con- stant obtained from the free vibration test and the forced vibration test.

The inherent period of the multiple-pillar foundation can be determined by replacing

the foundation with a model of one degree of freedom. The total mass of the ground surface

is taken as the mass which relates to the vibration. is then determined as follows;

1'

I Top-cover axial free vibration (A) K = 1.23 kg/cm

3 Tower axial free vibration (C) ^ K = 1.24 kg/cm

t 3

I Forced vibration (top-cover position) -> K = 1.08 kg/cm

These values are comparable to or slightly lower than value obtained from the static

load

.

IV-99
IV-100
1.000

0.500

Fig. 3 Relationship between a and /Si

IV-101
+ I r

Fig 5 Distribution of angle of inclination

(Comparison with colculated value)

krCTl^q/cm^ pi--2.0\ E:4.0x 10^ '*q/cm2

Bending momeni ilO^ kg' cm I

-O."^ 0 +0.5 tl

Deflection (cm)

Fig 6 Difference due. to supporting condition at pile tip

IV-102
D30 I

" Sandy ground

, • Loam

- ( -M

(In cose of elostic body

)

30 6C 90

Width of loading plate D icm

)

120

Fig. 7 Relationship between loading plate and coefficient of ground reoction

IV-103
Ep

30 eo 90

Width of loadm g plole D I f )

120

Fig, 8 Relolionship between Ep and E30

IV-104
9

-

Pile diomelef i cm i

Fig. 10 Relationship between pile diomeler and K

IV-105
Upper surtoce ot tooling Upper surfoce ol fooling

Cover plole

fl2

I Cross - shaped

.

^lifteninq plate Moin reinforcement

Lower surfoce ol (oolinq Poo; concrete

Grovel rrY'i

Steei pile

(1 )

hi Hi.

(:30

Fill - In concrete

Ivloin reintorcemeni Lower surface ol tooling

Poor concrete

Grovel RC RC

pile pile

:20

Jpper surface of footing

V Moin reintorcemeni

Reintorcement -I

DI5 pitch 150

Cover plate Lower surfoce of fooling Poor concrete

Gravel/'

Cross- shaped [ stiffening plote

Reinforcemeni -

PC bar

lUaIn reinforcement

Lower surface of fooling

Poor ':oncrel e

Grovel 'Tj

DI3-

pilch 150

Fill-in concrete -

RC RC

pile pile

_ 0. (4)

Ol3 or more — -

pitch 150 or more

Ivloin reinforcement

Lower surface of footing

Poor concrete

Grovel f

DIJ or more -

pilch 150 or more

Cost -in -place pile

Fig. 12 Connection between pile and footing

IV-106
Footing '

( —

T

I

Reinforce menI -

j \ 1 / /

\

'

N mil y

p

Pile

Shco' surlocc

Fiq 13 Design model to method B

J

Verlicol loodin;

Hofizonlal —u—

'•'1

P4 o

Pi

Pz

Concrete

^pT-Shoped steei|^ - 25O»250»9»ll"r

1.060 . , 960 L.480 _j J.500 ;

Fig 14 Shape and size of model

0 « i6

Hoop reinfofctmertl 9

Cover plate R. 16

'

Ring reinfocement »i6

3185

3 5^^is^- J.

50

103

Fig 15 Detoil of connection

IV-107
• : 4 3 Ion o 29

- 57 2

.i 82

•300 -X10 -;00 0

Stress (kg/ cm^ I

0

100 "

Ikg/crtiM

0 -lOO -200 -300

Stress I kg/cm^ I

Fig 15 Distribution of stress

IV-108
0 20 40 60 80 100

Horizontal load ( t )

(a) Bearing stress of footing concrete

Pile body —° Pi reinforcement —" P4 reinforcement

20 40 60 80 100 Horizontal lead ( t )

(b) Stress of stiffening reinforcement

0 20 40 60 80 100

Horizontal load (t )

(c) Confining moment at pile

Fig. 18 Comparison between calculated value ond measured value

IV-109
2.750 ni , ,680

I I NO g I

DQ

"4

o

Im)

P e 55 > OS

Loom 3

2 Gm

Cloy and

tine sand

5

1

13

10

Fine sond

IS

20

395 980 980 "^5

Fig. 19 Structure and dimensions of models Ground condition

Load on the lop-slob 30 tOIl

Load on the tower-top 60 ton

Load on the tower-top 100 ton

Fig 21 Ldteral ground reoction coefficient rotio of eacti pile on the bosis of central pile ( = 1)

IV-110
No I No I

Displocement of looding point

(direction of action of force) (mm)

0 12 3 4

Fig 22 Eccentric horizontal load on Itie top-slab, displacement vs torsionol spring constant

on the lop-end of the top-slob

K, ( '"" "^/Ixio" Rod )

Units gol

T

in neatly agreement with the fust

fjxed point at the lime of slotic loading

23 Acceleration mode of natural vibration at the tower-top position

IV-111
Table 1. Natural frequency f(c/s) , damping factor h (natural vibration)

Set position of Accelerometer

Axis-centered

natural vibration

test at the top-slab

Eccentric natural vibration test at the top-slab

Axis-centered

natural vibration

test at the tower-top

f h f h f h

No. I

Axial direction on the center of the top-slab

8.5 0.074 8.5

9.5

0.088 7.8 0.068

Perpendicular-to-axis direction

on the center of the top-slab

- - 0.058 - -

Axial direction on the upper comer of the top-slab

8.5 0.071 8.0 0.058 7.8 0.062

Perpendicular-to-axis direction

on the upper comer of the top-slab 15.2 0.064

Axial direction at 2m above the ground of the central pile 8.5 0.075 8.3 0.086 7.8 0.075

Axial direction at 0.3m above

the ground of the central pile

8.5 0.088 8.3 0.097 7.8 0.071

;io. II

Axial direction on the .:enter of

the top-slab

9.5 0. 128 10.0 0. 153 S.3 0.058

Perpendicular-to-axis direction

on the center of the top-slab 18.0 0.077

Table J Resonance frequency

f'(C/S]

Damping factor h

Natural frequency f (C/S)

f = f / l-2h2

(No. 1. central pile, from the measured strain, at the point

1 . 7m above the ground)

Vibrational

power f(C/S)

1 h f

0.3 7.3 0.073 7.85

0.3 7.85 0.078 7.9

0.6 7.85 0.078 7.9

0.6 7.85 0.080 7.9

1.8 7.65 0.087 7.7

1.8 7.50 0.066 7.55

IV-112
COMPREHENSIVE SEISMIC DESIGN PROVISIONS FOR BUILDINGS

by

Charles G. Culver

Disaster Research Coordinator

Office of Federal Building Technology

Center for Building Technology

National Bureau of Standards

Washington, D.C.

A review of the development of earthquake design provisions for U.S. building codes is presented. Suggested revisions to the current provisions are noted. A cooperative project directed toward developing comprehensive seismic design provisions is described. The organizational structure for the project including a breoikdown of the Task Committees required to develop the provisions and work state- ments for each Task Committee are included.

Key Words: Buildings; building codes; design; earthquakes; structural engineering.

IV-113
INTRODUCTION

Annual property losses from earthquakes in the United States average approximately

$14 million. Approximately eight lives are also lost annually. On a comparative basis,

losses from flooding and extreme winds such as hurricanes and tornadoes are much greater

(1) . The sudden loss potential for earthquakes, however, is great. It is estimated, for

example, that if the New Madrid earthquakes of 1811-1812 occurred today, losses would

probably total well over 50 billion dollars and countless lives (2)

.

The population growth and increasing urbanization in the United States have resulted

in a large number of structures built in seismically hazardous areas. Building codes and

regulations setting minimum standards for construction methods and materials are one means

for mitigating losses from earthquakes. The purpose of this paper is to examine activitie

in the United States directed toward the development of improved seismic design provision

for buildings.

Following is a brief review of the history of U.S. earthquake codes, proposed changes

these regulations and a comprehensive project directed toward updating these provisions

will be considered.

HISTORY OF EARTHQUAKE CODES IN THE UNITED STATES

Although earthquakes have occurred in many portions of the United States, the most

seismically active area is west of the Rocky Mountains. Most of the activity involved in

developing seismic design requirements, therefore, has been concentrated in California.

Following the 1906 San Francisco earthquake, the city was rebuilt under a code which provided 30 psf wind force to effect both wind and earthquake resistance. The Newtonian con- cept in which lateral earthquake forces are proportional to the mass of the structure was employed by structural engineers, but was not incorporated in the early building codes.

The requirements in the 1927 edition of the Uniform Building Code were among the first

earthquake provisions to be written into any widely used building code in this country.

Following the Long Beach earthquake in March, 1933 and passage of the Field Act in California, public school buildings were required to be designed for lateral forces ranging

from 2% to 10% g. The Riley Act of 1933 required all California buildings, except certain

types of dwellings and farm buildings, to be designed for lateral forces.

Since the time of these early code requirements, earthquake provisions have continual]

evolved. A significant step in this process involved the formation in 1948 of a Joint Com

mittee on Lateral Forces of the San Francisco Section, ASCE, and the Structural Engineers

Association of Northern California. The lateral force provisions advocated by the committee were based on Biot's work on earthquake spectra (3). Following formation of the Seismology Committee of the Structural Engineers Association of California (SEAOC) in 1957,

the first edition of a comprehensive set of requirements was published as "Recommended

Lateral Force Requirements" in 1959 (4) . These have been continually updated and the

latest edition published in 1973.

Following the San Fernando Earthquake in 1971, revision of existing seismic code re- quirements received considerable attention. A dynamic type code utilizing in an explicit

fashion the response spectra concept was proposed for the City of Los Angeles. New re- quirements were developed for hospitals and other critical facilities in California. Also

various Federal agencies including the Veterans Administration and the Department of

Health, Education, and Welfare developed new seismic design criteria.

Although many of the proposed new seismic provisions are similar, considerable differences exist in many cases. As a result, it became apparent that a need existed to up- date U.S. seismic design provisions. The next section describes a comprehensive project

in this area undertaken by the National Bureau of Standards (NBS) and the National Science

Foundation (NSF) .

'

.1

IV-114
IMPROVED SEISMIC DESIGN PROVISIONS

The last major revision in seismic code provisions in the U.S. occurred in the 1960 's Since that time, there have been significant advances in knowledge of the response of

building systems to seismic ground motion and correspondingly in analytical techniques and

design procedures. One of the recommendations developed at a workshop on disaster mitigation sponsored by NBS and NSF in 1972 (5) indicated that model code provisions for seismic

design should be prepared on a top priority basis to bring the minimum level of practice

into line with the current state of knowledge. Similarly, the Joint SEAOC-ASCE Committee

on Seismic Forces recommended strongly that existing seismic design requirements be re- vised (6) .

As part of the Cooperative Federal Program in Building Practices for Disaster Mitigation, NBS and NSF initiated a project to develop coitprehensive nationally applicable seismic design provisions. The overall project involved two phases, as outlined in the follow

ing discussion.

Phase I - Evaluation of Response Spectrum Approach

Many of the proposed revisions to seismic design provisions advocated a response spec

tirum approach and the Joint SEAOC-ASCE Committee felt that if dynamic design criteria were

established, a spectrum approach should be used (6). Recognizing this, NBS initiated a project in May 1973 to evaluate such an approach. The scope of this effort was confined

to a study of the practicality, additional costs, and related effects. Working with design professionals and researchers, a set of provisions for the design of the structural

systems of regular buildings based on a response spectrum approach was developed by a five man Engineering Panel from the Applied Technology Council (STC) . ATC functions as the

research arm of the Structural Engineers Association of California.

The Newmark-Hall response spectrum method was used in developing the provisions to- gether with ground motion values for acceleration, velocity, and displacement representative of a given site in Southern California. Two sets of spectra were used. The first,

the Damage Threshold Spectrum, was assumed representative of moderate earthquake motions

at the site having a moderately low probability (about 50%) of being exceeded during the

life of the structure (70 years) . The design philosophy adopted was assumption of the ac- ceptability of the prevention of significant structural damage during such an earthquake.

For the second set of spectra, the Collapse Threshold Spectrum representative of major intensity earthquake motions at the site having a very low prability (about 10%) of being

exceeded during the life of the structure (70years) , it was assumed acceptable to prevent

collapse of the structural system.

The design provisions were based on a linear elastic modal analysis using the two

sets of spectra. Ductility modification factors were employed to account for the effects

of inelastic action. Structural members were designed using available ultimate strength

provisions for concrete and steel and factored allowable stress provisions for masonry

and wood.

Following development of the design provisions, they were applied in the redesign of

eleven existing California buildings. The buildings selected for this phase of the work

are indicated in Table 1. Note that the selection includes a sampling of basic construction material, height of structure, and type of lateral force resisting system. The original design firms for the buildings were commissioned to do the redesign in accordance

with the new provisions. Since the buildings were originally designed using varying code

requirements in existence at the time, they were all first redesigned to the provisions

of the 1973 Uniform Building Code (UBC) to provide a common baseline for comparison.

The cost involved in applying the new design provisions was evaluated. The difference in construction costs resulting from the new provisions as compared to the designs

based on the 1973 UBC were determined. In addition, the extra engineering costs associated with the new provisions as compared to the engineering costs required to apply the

equivalent static force provisions of the 1973 UBC were determined.

IV-115
A final report including the design provisions and commentary, the results of the re- design study, analysis of the cost noted above, and recommended modifications to improve

the design provisions are contained in a final report currently under review. It is an- ticipated that this report will be published by July 1974, following review by NBS, the

SEAOC Seismology Committee, and structural engineers from California.

Phase II - Comprehensive Nationally Applicable Provisions

The study described under Phase I was of limited scope. First, it only considered

seismic design provisions for structural systems. Second, the general problem of risk

analysis and determination of seismic forces was restricted to one site in Southern California. Numerous other factors must be considered in establishing comprehensive provisions applicable to seismic design throughout the U.S. (6). To develop these provisions,

NBS and NSF working with ATC initiated a second project in December 1973. The provisions

will be based on the current state of the art incorporating the latest research results

in earthquake engineering. The organizational phase of tliis project was recently com- pleted and it is anticipated that initiation of the work to develop the provisions will

begin in July 1974.

Since it is intended that the provisions will be national in scope and adaptable for

implementation by existing model code groups and consensus standards organizations, broad

based participation in the project was necessary. For this reason, two groups were established to participate in both organizing the work and serving in an advisory capacity

throughout the course of the project—a Seismic Design Review Group (SDRG) and a Building

Code Consultant Group (BCCG) . These two groups together with ATC, NBS, and NSF determined

the organizational committee structure and work statements for the committees to be

charged with developing the provisions.

The SDRG is comprised of national authorities in earthquake engineering representing

both the design profession (architects, engineers, geologists) and the research community.

They will participate in the technical areas of the program. The BCCG consists of code

enforcement officials from throughout the U.S. representing model codes and state and local

regulatory groups. This group is charged with advising on the structuring of the format

and content of the new provisions as related to implementation.

The organizational structure established by these groups to develop the new provisions

is shown in Figure 1. A total of five task groups each with a number of committees has

been formed. An objective and brief work statement for each group is given in Table 2. Detailed lists of topics to be considered by each committee and specific questions to be

resolved have also been developed.

Referring to Figure 1, the numbers in parenthesis following each committee indicate

the number of committee members. Approximately 90 individuals will be involved in the

project. In selecting the committee members, broad based participation from the standpoint of technical disciplines (geology, seismology, engineering, architecture, etc.)

and geography will be considered. Balanced participation by design professional and re- searchers will be provided.

Recognizing that the various areas considered by each committee are inter-related,

provisions have been made for appropriate interactions in developing the various portions

of the provisions. Similarly, the provisions developed by each committee will be reviewed

by the two advisory groups, NBS, and the SEAOC Seismology Committee.

In carrying out the pro ject, provisions have been made for maintaining liaison with

the design professions, public officials, and other groups to inform them of the development of the provisions and solicit their comments.

A time period of eighteen months has been established for the work with an anticipated completion date of December 1975.

IV-116
CONCLUSION

The activities described in this paper represent the first step in the development

of improved building practices in the United States to mitigate losses due to earthquakes.

After completing the design provisions, a continuing effort will be required to keep them

updated incorporating new knowledge as it becomes available. In addition, a siabstantial

educational effort should be initiated to present the provisions to design professionals,

discuss their interpretation, and facilitate appropriate implementation.

REFERENCES

1. Sav, G. Thomas, "Losses Due to Natural Disaster," National Bureau of Standards Internal

Report , Washington, D.C., 1974.

2. Wiggins, John H., "Toward a Coherent Natural Hazards Policy," Civil Engineering Magazine , April, 1974, pp. 74-76.

3. Rinne, John E. , "Building Code Provisions for Aseismic Design," Proceedings , EERI Symposium, June 1952.

4. Recommended Lateral Force Requirements and Commentary , Seismology Committee, Structural

Engineers Association of California, 171 Second Street, San Francisco, California,

1973.

5. Building Practices for Disaster Mitigation , Building Science Series 46, National Bureau

of Standards, Washington, D.C. , February, 1973.

6, Degenkolb, Henry J., Chmn., "Basic Design Criteria of the Recommended Lateral Force

Requirements and Commentary," Journal of the Structural Division , ASCE, Vol. 98, No.

ST9, Proc. Paper 9182, September, 1972, pp. 1913-1922.

7. Uniform Building Code Standards , 1973 Edition, International Conference of Building

Officials, 5360 South Workman Mill Road, Whittier, California.

IV-117
TABLE 1

TEST BUILDINGS FOR RESPONSE SPECTRUM PROVISIONS

1 5 Stories

Ductile reinforced concrete frame

2 19 Stories

Steel frame; moment resistant in one direction

Braced in the other direction

3 10 Stories

Reinforced concrete frame in one direction

Shear wall in the other direction

4 14 Stories

Shear wall

5 7 Stories

Shear wall

6 2 Stories

Steel frame with vertical bracing system

7 3 Stories

Masonry shear wall with concrete-metal decking diaphragm

8 9 Stories

Shear wall with vertical load-carrying concrete frame

9 1 Story

Tilt-up, plywood diaphragm

10 2 Stories

Moment resistant steel frame

11 1 Story

Masonry plywood diaphragm

IV-118
TABLE 2

COMMITTEE WORK STATEMENTS FOR DEVELOPMENT

OF COMPREHENSIVE SEISMIC DESIGN PROVISIONS

Task Gro\;5> No. 1 - Definition of Seismic Input

Engineering Seismology Committee

Objective: To develop the concepts involved and to describe the procedures

for selecting and describing the earthquake ground motions to be

used in design, and for laying out the Design Maps.

Statement of Work: With due consideration of the regional geology and tectonic

structure, distance from active or potentially active faults,

historical records and geological evidence of the location

and intensity of regional earthquakes, and the recurrence interval between major earthquakes, develop means of characterizing the earthquake ground motions that should be used

for both dynamic and pseudo-static design.

Site and Foundation Effects Committee

Objective: To develop the concepts involved and to describe procedures for

evaluating the factors that affect ground motions and stability

at a site.

Statement of Work: Consideration should be given to all possible site and foundation factors that affect motion intensity, motion frequency

distribution, and soil-structure interaction including:

Distance from seismic source

Travel path from source

Type of seismic source

Site geology both vertically and laterally

Other factors as listed in the attached Guidelines

to Geologic/Seismic Reports

Risk Analysis Committee

Objective: To develop the concepts involved and to describe procedures for

evaluating the seismic exposure and risk potential for a given

site, area, or region.

Statement of Work: Evaluation of seismic risk should take into account the

available seismological and geological information pertinent to the occurrence of earthquakes. The probability of

earthquakes of various magnitudes occurring near a building

site, and the probability of ground shaking of various intensities at the site should be considered. (A structure

at a site where strong shaking is to be expected once in

500 years should not have the same seismic design criteria

as a structure at a site where such ground shaking is ex- pected once in 50 years) . The evaluation of seismic risk

should include consideration of special geological conditions such as proximity to a major active fault, extreme

softness of foundation soil (fill, bay mud, etc.), potential

landslides, or potential soil liquefaction.

IV-119

1
A performance criteria adaptable for a building code

should logically be based on the notion of acceptable probability of experiencing damage more severe than some specified degree of damage. Seismic Design Zoning Maps should,

therefore, reflect not only the probability of ground shaking of various intensities, but should also reflect the ac- ceptable probability of exceeding a specified level of damage.

Level of damage should reflect life safety considerations, ' and for critical facilities should also reflect maintenance

of function, and should be based on typical structures.

The analysis of seismic risk encompasses two separate

aspects: the relative risk exposure for different seismic

areas and the level of seismic risk to structures which will

be considered acceptable. Considering all available information such as historical records, geologic information, and

regional tectonics, develop procedures for estimating the

probability of occurrence of different levels of earthquake

motion at a site or area. In conjunction with Task Group

No. 2 and utilizing available data, develop methods for eval- ,: uating the effects on structure design and construction of i the different motion levels.

Task Group No. 2 - Structural Behavior

Struct\jral Design and Details Committee

Objective: To develop detailed seismic design provisions with appropriate com- mentary, suitable for adoption by building codas in all areas of the

United States, for the structural design and required details for

earthquake resistant design in concrete, metals, timber, and masonry ;| considering material properties, construction capabilities, and mem- i ber stresses and ductilities. i' Statement of Work: It is assumed that other task group committees will define j the appropriate design ground motions, effects of soil-struc- : ture interaction, level of risk, importance considerations,

and the structural analysis methods required to determine ' member and element forces and deformations for design pur- I poses, including the effects of setbacks, sudden changes of

'

' stiffness or mass, torsion, and three-dxmensional loadings.

' Determine and define the appropriate material design

stress or deformation levels required to achieve suitably

uniform factors of safety and the required level of ductility achievable with various materials. Define the details

of construction necessary to achieve the required ductility.

National and regional material specifications may be used,

j but detailed supplements must be added to make the result

suitable for use in seismic regions.

Typical requirements may be quantitative design requirej ments for collector bars, trim bars, degree of tying elements

of a structure together, panel zones, P-A effects if large

ductility factors are permitted, specific requirements re- lated to anchoring parts of a building together (parapets,

tilt-up walls), diaphragm action, and accompanying details,

necessity for complete force path, etc. Criteria are also

required for shear wall design, interaction of walls and

frames, differences in performance of various framing methods,

j

IV-120
basement wall pressures, necessity for or design' methods for

transferring ground motion to structure (pile bending) , etc.

Structural Analysis Committee

Objective: To develop methods for determining the design forces and/or deformations in the members and joints of structures caused by the action

of the specified earthquake input.

Statement of Work: Methods shall be developed to define values of earthquake

response quantities included in the structural system which

are suitable for use in design. Since the design earthquake

may produce structural stresses or strains which are significantly greater than the elastic limit of the materials, the

methods shall be capable of recognizing these conditions.

Emphasis shall be placed upon approximate analysis procedures

because the approximate nature of the design earthquake precludes any exact analysis, but different methods shall be

specified for various classes of structures depending on the

corrplexity of their dynamic response

.

In order to expedite the work of other Task Committees,

the Structural Analysis Committee should as first priority,

decide on the general types of analyses and conditions under

which they are appropriate. Those results should then be

disseminated to the others. The Committee should then turn

to developing details of the analytical procedures.

Soil-Structure Interaction Committee

Objective: To develop procedures for determining when soil-structure interaction

effects are significant and design provisions to provide for same.

Statement of Work: Under certain conditions, there is significant interaction

between a building and the soil on which it bears. In effect,

the soil and the structure together constitute a system that

responds to the free field seismic motion. Where the natural

periods of the structure, when assumed to be on a rigid base,

tend to coincide with one or more natural periods of the

supporting soil, there may be greatly increased amplification

of the building's response. The dynamic characteristics of

of the building may be significantly affected by the compliance of the soil. In some cases, the building may affect

the soil as, for exaitple, the weight of the building or the

rocking of the building may cause a decrease in ultimate

soil bearing pressure when there is excess pore pressure in

the material from liquefaction at lower levels.

Quality Assurance Committee - Structural

Objective: To prepare a state-of-the-art report and quality assurance provisions

regarding

:

1. Quality control of structural materials at the manufacturer

and in the field.

2. Acceptable tolerances in fabrication (shop and field) and

on-site construction.

3. Inspection--qualifications , duties, authority, and responsibility of inspectors and of design engineer.

IV-121
statement of Work: The effectiveness of a design is only as good as the materials used and quality of the actual construction. The Committee shall study present materials and procedures, develop

preferred specifications and procedures, and present same in

format compatible with those used in building codes.

k Group No. 3 - Non-Structural Components

Architectural Systems Committee

Objective: To study the effects of earthquake-induced structural deformations

on architectural systems and to develop appropriate design guidelines

and provisions to minimize life-safety related damage to architectural systems

.

Statement of Work: The Committee should consider in their deliberations seismic

design requirements affecting architectural considerations

including the following:

1. Size - Should there be limitations on undivided

floor areas; should sub-buildings be con- sidered; are volume restrictions applicable?

2. Shape - Should there be limitations on proportions

of areas and volume; should sub-buildings

be considered horizontally and vertically?

3. Dimension - Should there be any ultimate limitation

on dimension; should there be conditions

that would reduce such limitations?

4. Circulation - Should egress requirements be restated

as to protection offered—distance,

number of routes?

^

5. Materials - Should type of materials used restrict

other limitations? Should chemical or physical properties restrict material

selection?

6. Detailing - Should finish materials be connected

to structure? Should strength of con- nections be controlled? Should provisions be made so that structure elements

can move independently (within limits)

of cladding, partitions, ceilings, and

other non-structural elements? Should

secondary attachment systems be required

or is a single anchorage system adequate!

7. Equipment - Should moveable equipment be restricted;

should equipment design be controlled?

8. Damage Control Systems - Should self-contained alarm

and/or support and/or ex- tinguisher systems be re- quired?

9. Interrelations - should options be given the designer and/or client through se- lection of systems affecting other
requirements? What about effects

on adjacent buildings or people

outside the building?

Mechanical-Electrical Systems and Equipment Committee

Objective: To evaluate problems involved with life-safety as related to mechanical and electrical systems and equipment, to determine which systems

and equipment should be subject to seismic resistant design, and to develop design provisions consistent with design levels (or quality)

specified for the building structure.

Statement of Work: The Committee should make a detailed review of damage data,

existing provisions, if any, now listed for seismic-resistant

design and develop seismic design guidelines and provisions

for life-safety systems.

Present efforts should be limited to mechanical and

electrical systems that are supportive to the building function, i.e., heating, ventilating and air conditioning systems

elevators, emergency power (for essential systems) , light

fixtures, etc., and operative systems for critical facilities

Considerations should be given to structure-system (or

equipment) interaction or coupling, equipment or system response to dynamic motions, interaction between equipment,

and better detailing.

Quality Assurance Committee - Architectural, Mechanical, and Electrical

Objective: To develop the concepts involved and to describe procedures for en- suring effective quality assurance for Architectural-Mechanical and

Electrical systems and/or equipment.

Statement of Work: The Committee should study present quality assurance procedures to determine whether or how they should be modified or expanded to ensure that design, fabrication, and construction

or architectural-mechanical and electrical systems and/or

equipment will function as contemplated by the design provisions developed by the other Task Group No. 3 Committees.

Task Group No. 4 - Liaison and Format

Liaison and Information Dissemination Committee

Objective: To assist the Task Committees in developing liaison with the design

professions and public officials so as to inform them of the development of the design provisions and to solicit their comments.

Statement of Work: The Committee should work with the other Task Groups and

Committees to ascertain their progress and to develop appropriate means of informing the design professions, public

officials, and the public of the development of these design

provisions and the potential future benefit from using them.

The Committee should consider the need for informing

the many interested groups on a timely basis.

IV-12 3
Design Provisions Format Committee

Objective: To edit design provisions prepared by other Task Groups and Committees

into editorial format compatible with those used by building codes.

Statement of Work: The Committee shall work with the Building Code Consultants

Group to develop a format outline and distribute same to the

Task Groups and their Committees. The Committee should as- sist Committees as required and edit draft material before

issuance to outside groups.

Task Group No. 5 - Existing Buildings

Inspection and Evaluation of Damage Committee

Objective: To develop procedures and criteria for assessing the structural safetyof buildings subjected to earthquake ground motions.

Statement of Work: It is often difficult to quickly inspect and fully evaluate

the structural safety of a building and its components after

it has been subjected to significant seismic ground motion.

It is necessary for the protection of the public safety that

orderly and efficient procedures be developed for inspecting

damaged structures and determining whether or not they are safe for human occupancy. There is also a need for developing procedures for evaluating the damage in detail to determine whether repair work is required. i

Procedures and criteria shall be developed by which

buildings can be inspected and damage evaluated so as to determine whether they are safe for immediate occupancy.

Procedures and criteria shall also be formulated for assess- ing the extent of the damage and the amount of structural

repair required.

Repair of Earthquake Damage Committee

Objective: To develop procedures for determining the extent, type, and adequacy

of required repairs of earthquake-damaged buildings and components

so as to ensure life safety.

Statement of Work: Procedures and criteria shall be developed for determining

the extent, type, and adequacy of repairs required to ensure

the structural stability of earthquake-damaged buildings and

components. Repairs to both structural and nonstructural

elements shall be considered where their failure or instability under normal or earthquake conditions could endanger

life safety of occupants or passer-bys.

Evaluation of Existing Buildings Committee

Objective: To develop procedures and criteria for inspecting and evaluating

existing buildings and structures for possible structural inadequacies that could endanger life safety in the event of an earthquake .

Statement of Work: A large proportion of existing buildings were built before

adequate seismic design standards were developed and connections, ties, mortar, etc. have deteriorated. Many of these ;. buildings have conditions that are potential safety hazards

in the event of an earthquake. Such conditions may vary

from minor appendages to parapet walls to the entire structure

IV-124
Provisions shall be developed delineating procedures

and criteria to be followed in evaluating such potential

hazards and determining the extent of strengthening

required.

Strengthening of Hazardous Bxiildings Committee

Objective: To develop minimum standards for strengthening of existing buildings,

outline priorities for strengthening based on potential hazard and

formulate procedures for evaluating the adequacy of proposed strengthening measures.

Statement of Work: Other committees m Task Group No. 5 have the responsibility

for developing procedures and criteria for inspecting and

evaluating earthquake damage, repair of such damage, and inspection and evaluation of existing structures for potential

hazards. This Committee shall develop minimum standards for

strengthening structures which have been determined to be

either hazardous or inadequate for earthquake resistance.

Procedures for establishing possible priorities for strengthening based on degree of potential hazard or life exposure

shall be developed. Provisions for evaluating the adequacy

of proposed strengthening measures shall also be formulated.

IV-125
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IV-126
WIND LOADING AND MODERN BUILDING CODES

by

Emil Simiu

Structural Research Engineer

R. D. Marshall

Structural Research Engineer

Center for Building Technology

National Bureau of Standards

Washington, D.C.

The differences between the dynamic alongwind response, the gust factors,

and the total alongwind response obtained using various current procedures may

in certain cases be as high as 200%, 100%, and 60%, respectively. The purpose

of this paper is to investigate the causes of such differences. To provide a framework for this investigation, the paper presents an overview of the questions

involved in determining alongwind structural response, and a critical description

of the basic features of procedures currently in use. A comparison is made be- tween alongwind deflections of typical buildings selected as case studies, calculated by both new and traditional procedures , some of which are described in

various building codes. The reasons for the differences between the respective

results are pointed out. The procedures were evaluated on the basis of a recently

developed method which utilizes a logarithmic variation of wind speed with height

above ground, a height-dependent expression for the spectrum of the longitudinal

wind speed fluctuations. The method also allows for realistic cross-correlations

between pressures on the windward and leeward building faces.

Key Words: Building codes; buildings; deflections; dynamic response; gust factors;

structural engineering; wind loads.

IV-127
INTRODUCTION

Following an increasing recognition of the importance in tall building design of the

dynamic effects due to the gustiness of wind, several procedures for computing alongwind

response have been proposed in the last decade (6, 19, 24, 25) . Two of these procedures

are described in Ref. 1, herein referred to as the A58.1 Standard, and in Refs. 5, 11,

herein referred to as the NBC. The purpose of these procedures is to calculate equivalent

static loads whose effect upon the structure is the same as that of the gusty wind.

For a given structure in terrain of given exposure and in a given wind climate, it

might be expected that roughly comparable equivalent static loads (and, therefore, comparable values of the calculated response to wind) will be obtained regardless of what method

of computation is used. This, however, is not the case; as has been reported before (25)

,

depending upon whether the procedure described in the A58.1 Standard or in the NBC is used,

values of the response that may differ considerably from each other are obtained. Such

differences may be as high as 50% or more. An investigation into causes of such discrep- . ancies is thus believed to be in order.

Such an investigation is the intent of this paper. To provide a framework for the

investigation, an overview of the question of alongwind structural response will be presented first. This will be followed by a comparative analysis of the various procedures

presently in use. A comparison will also be presented between response obtained by these

procedures and that obtained by assuming wind loads traditionally used in structural design and specified by various building codes. On the basis of these comparisons, an as- sessment will be made of the procedures analyzed.

^

ALONGWIND RESPONSE: AN OVERVIEW

As a framework for the evaluation of current procedures for computing alongwind re- sponse, an overview of the questions involved in determining such response is presented

in the following. These questions include the relation between alongwind response and

wind pressures, the relation between wind pressure and wind speeds, and the definition of

wind speeds as functions of height above ground, roughness of terrain, climate and desired level of safety of the structure. A summary of relations defining the alongwind

response in terms of mechanical and environmental parameters will also be presented.

Relation Between Alongwind Response and Fluctuating Wind Pressures

It is known from elementary calculus that a periodic function may be expressed as a sum of fluctuating harmonic components with discrete frequencies. Its mean square value

can then be expressed as the sum of the square of the amplitudes of the harmonic components. Similarly, a stationary random function F(t) may be viewed as a superposition of

fluctuations of frequencies n, with n covering the entire interval from zero to infinity,

while its mean square value may be expressed as a sum of contributions associated with

these fluctuations, i.e.; — °° F2(t) = / Sp(n)dn (1) o The quantity Sp(n) is a measure of the magnitude of the fluctuating component of frequency

n and is known as the spectral density function of F(t) . ...

Consider a linearly elastic structure subjected to the action of a stationary random

force Fp(t) of known spectral density Sp (n) and applied at a point P. The spectral density of the fluctuating deflection a' (t)^at some point M of the structure can be shown

to be (16)

S^(M, n) = H*(M, P, n)H(M, P, n)Spp(n) (2)

in which H{M, P, n) = mechanical admittance = response of the structure at point M due to action at point P of a unit complex harmonic force, and H*(M, P, n) = complex conjugate of

H(M, P, n)

.

IV-128
If the structure is acted upon by two stationary random forces Fp^(t) and Fp (t) ap- plied at points P^/ ^respectively , the spectral density of the response a' (M, t) can be

shown to be (16)

S^(M, n) = H*(M, P-L n)H(M, P]_, n)Sp (n) + H*(M, P2, n)H(M, P2, n) Sp (n) ^1 ^2 + H*(M, Pj_, n)H(M, P2, n)Sp p

(n) + H*(M, P2 n)H(M, P^.' "^^p p

(n)

1 2 ^2 ^1

(3)

in which S_ ^ (n) = cross-spectral density function of (t) and Fr, (t) . If F„ (t) = Pl'P2 ^ 2 ^1

F„ (t) , then S„ „ (n) = S„ (n) . Methods for computing S„ „ (n) are described in Refs. rp rp rp ^ irp rp 2. P-L P2 ^^1 ^1 ^2

4, 16. The imaginary part of the cross-spectral density function is known as the quadrature

spectrum and has been found by measurement to be negligibly small for most wind engineering

purposes (6, 8, 24, 26)

.

If the quadrature spectrum is negligible, it is convenient to express the cross-spectral density of two functions F_ (t) , F„ (t) in the form

S_ „ (n) = S^/^ (n) S^/i (n) RpCP., P^, n) (4) Fp/P2 ^P2 ^P2

in which Rp(Pj^, V is the square root of the coherence function and can usually be determined only on the basis of experiment. RpCPj^, P2/ n) is a measure of the extent to which the functions Fp (t) , Fp (t) are correlated. If these functions are perfectly cor- related (e.g., if Fp^(f) E Fp^^t)), then RpCPi, P2' n) = 1. Otherwise, Rp(P]^, P2, n) < 1.

Equation (3) ccin easily be seen to reflect the obvious fact that the response due to two

forces increases as their correlation increases. For example, assuming Sp (n) = (n)

,

Poorly

work at cross- purposes ,

S^(M, n) will be twice as large if P2' = 1 than if Rp(P-|^, V ^, n) H 0, correlated random forces may thus be thought of as forces which, in a sense, wc

If a distributed stationary random loading is applied to an area A, Equation (3) may

be generalized (16)

:

S^(M, n) =

/ / H (M, V^, n)H(M, V ^, n)S (P;^' ^2' n)dA]^dA2 A A

/ / H*(M, P , n)H(M, V^, n)sl/2(p^, r).)S^^'^ {V ^, n)R (P^, P2, n)dAj^dA2

A A

(5)

in which P^, P^ = centers of elemental areas dA^, dA2; Sp(P-|^, n) , Sp(P2f n) , Sp(P^, P2/ n) = spectral and cross-spectral densities of pressures at points P^^, P2; Rp^^i' ^2' ~ correlation coefficient = square root of coherence function. The mechanical admittance

function incorporates the parameters describing the mechanical properties of the structure

and may be expressed as

:

lJj.(M)yg(M)

H*(M, P^, n)H(M, P2, r.) =11 " —

r s 16n M M n^n^

(1 - n2/n2) (1 - nVn^) + 46^83 (n/n^.) (n/n^)

[(1 - nVn^)^ + 432n2/n^][(l - n^/n^) ^ + ^^'^x? /r?-\ r r r s s s

(6)

in which p^(M), M^, n^, 8^. = modal shape, generalized mass, natural frequency and damping

IV-129
ratio, respectively, in the r*-" mode. For a building of height H

H

= / y ^(z) m(z)dz (7)

where z = height above ground and m(z) = mass of structure per unit height. It follows

from Equations (1) , (6) , and (7) that the mean square value of the fluctuating response

of a linearly elastic structure can be obtained if, in addition to the geometry and the

mechanical properties of the structure (mass distribution, modal shapes, natural frequencies, and damping ratios, including aerodynamic damping which can be determined as shown

in Ref . 26) , the spectral density function and the cross-correlation coefficient of the

pressure fluctuations are known. As will be shown subsequently, these quantities can be

determined using basic aerodynamic considerations and relevant information on the characteristics of the atmosphere boundary layer.

Aerodynamic Considerations: Relation Between Wind Pressures and Wind Speeds

The pressure acting at a point P of elevation z on the surface of a building immersed

in a flow which has a steady mean velocity v(z) may be expressed as

F(P) = 1/2 PCp(P) ^(z) (8)

In this expression v(z) = undisturbed mean velocity of the flow, i.e., its velocity at a sufficiently large distance upwind from the building, and C is a dimensionless mean pressure coefficient. For blunt bodies in turbulent flow, Cp (P? can usually be determined

only by escperiment.

For the windward and leeward sides of a rectangular tall building, the mean pressure

coefficients specified by the A58.1 Standard are = 0.8 and C^^^ = - 0.6, respectively.

Wind tunnel measurements reported by Haines (2) suggest that these values are, on the

average, conservative. Also, measurements indicate that mean pressure coefficients for

full-scale buildings are smaller than those obtained in wind tunnel tests, i.e., are slightly decreasing functions of Reynolds nximber (11) .

An extension of Equation (8) to include the effects of velocity fluctuations takes

the form

p(P) = p(P) +p'(P) = 1/2 pC (P) [v(z) + v' (z) ]^ + L PC (P)B(z)v' (z, t) (9)

ir 4 in which p' (P) and v' (z) = pressure and velocity fluctuations, C„(P) = added mass coefficient and B(z) = width of structure. The last term in Equation (9) is known as the added

mass term. The question of its relative importance in Equation (9) has been examined by

Vickery and Kao (26) . On the basis of wind tunnel pressure measurements (8) , these

writers showed that in determining pressures on buildings the added mass term may be ne- glected. The contribution to this term of its low frequency components is insignificant

because the accelerations of these components are very small (26) . To explain the experimental results of Refs. 8 and 26, according to which the contribution of the high frequency components is also negligible, it may be argued that, while for the small eddies,

the accelerations are large, the corresponding added mass coefficients are very small.

Indeed, while in the case of an infinite mass of fluid being accelerated with respect to

a body Cj^ Cp, in the case of a turbulent eddy the accelerated mass is of the order of

the eddy size and, therefore, decreases as the frequency increases. Thus, neglecting the

added mass term, it follows from Equations (8) and (9)

,

p"(P) = PC (P)v(z) v' (z) [1 + 1/2 v' (z)/v(z) ] (10) P

If v'(z)/v(z) is small (i.e., v^Tz) ^/^/v (z) < 0.1) and the linear dimensions of the

body are small compared with the length characteristic of the turbulence, the assumption

that the non-linear term in Equation (10) may also be neglected is confirmed by_indirect

experimental evidence (3). In the atmosphere it may be assumed that v' ^ (z) l/2/v(z) -

lV-130
2.54 u^/2.5 n(z - z^)/z ], in which u^, z^, z^ = frictional velocity, zero plane displacement, roughness length, respectively (13, 18, 19) . Thus, for tall buildings this

ratio is of the order of 0.05 to 0.3, depending upon building height and roughness of terrain. Also, except in the case of slender, line-like structures, the ratio of building

dimension to scale of turbulence is not necessarily very small. Questions may thus arise

as to whether the non-linear term may be neglected in the case of buildings with typical

widths in atmospheric flow. Practical difficulties^ have prevented so far the carrying-out

of simultaneous full-scale measurements of p'(P), v(z), and v' (z) . However, wind tunnel

measurements have been performed (7, 26) and appear to confirm the assumption that Equation (10) may be linearized. Also, the effect of the non-linear term was analyzed in

Ref. 23, according to which the contribution of this term to the fluctuating alongwind re- sponse of a 300 m. tall structure appears to be of the order of 5% (i.e., a contribution

to the total alongwind response of about 3%) . It thus appears that the linearization of

Equation (10) is acceptable and that

p' (P) = Cp v(z) V" (z) (11)

Research aimed at developing an improved model of the relation between alongwind

fluctuating pressures and velocities is clearly desirable. It appears, however, that

Equation (11) may represent a reasonably adequate engineering model for this relation. If

Equation (11) is used, it follows

Sp(P^, P^, n) = CpP2v(z^)v(z2) S^C^i/

in which the cross-spectral density of the velocity fluctuations can be written as

^v^^l' ^2 "^^ " ^^^l^^l' S^/J(P2, n) R^(Pi, P^, n) (13)

It follows from Equations (5), (12), and (13) that RpCP^/ ^2' " \^^1' ^2' "

If Equation (11) is applied to express the pressures on the windward and leeward sides

of the building:

P' (P) = C^v(z) V' (z) (15)

PjJ (P) = Cj^v(z) v- (z) (16)

As noted in Ref. 17, in reality the flow in the separation bubble is quite dissimilar from

that in the oncoming flow. Measurements suggest that the fluctuating pressures on the

leeward side are small compared to those on the windward side and, therefore, that Equation

(16) may be slightly conservative.

Planetary Boundary Layer Flow: Mean Wind Speeds

The mean wind speeds used in design depend upon roughness of terrain, wind climate,

and desired level of safety of the structure.

The roughness of terrain determines the shape of the mean velocity profile which can be shown to be given by the relation

z - z^

v(z) = 2.5 u. In (17) "o in which u^^ = 0.4 v(z^) /in [ (z^ - Zj)/z^] and z^ is some reference height. Equation (17)

is the well-known logarithmic law (14, 15, 18) and values of z^, z^ are given in Ref. 18. It the terrain is uniform over a sufficiently large fetch. Equation (18) is valid up to

a height

z„ = 100 v(z)/ln[(z - z,)/z ] (18) p a o in which v(z) is expressed in units of length pe^r second (18). For example, if z^ = 0.07 m,

= 0 (open terrain) and z = 9.15 m (30 ft.), v(z) = 35 ms"^ (~ 80 mph) , then z - 720 m.

IV-131
For a given angle of latitude, the geostropic wind speed (the wind speed in the free

atmosphere, flowing under barotropic conditions along straight isobars) is determined

solely by the magnitude of the pressure gradient force. Near the ground, however, the wind

speed also depends upon the roughness of the terrain. To any given geostrophic wind there

will thus correspond mean velocity profiles denoted by v(z, z^, z^) which are related to each other as shown in Figure 1. For example^, if over open terrain with z^ = 0.08 m, = 0, the mean^wind speed at z-|^ = 10 m is v(z-^^, z^, z^) = 20 ms"-'- (45 mph) , over builtup terrain with z^ = 0.7^8 m, z^ = 21 m the corresponding mean wind at Z2 = 43 m can be

found (Figure 1) to be v(Z2, z^, z^) = 15.8 ms"-*- (35.5 mph).

The description of the mean wind structure incorporated in Equation (17) and Figure

1, which will be referred to herein as the similarity model, is based upon recent results

of atmospheric boundary layer studies for which a satisfactory theoretical, as well as experimental foundation, exists. In the A58.1 Standard and the NBC, an empirical power

law model is employed. The power law exponents are assumed to be constant up to the

gradient height and are specified for three standard terrain roughness conditions, classified as open, siiburban, and urban. The power law model has shortcomings which may lead

to results that are not consistent with actual measurements. An analysis of these shortcomings and comparisons between the sindlarity and the power law models can be found in

Ref. 18.

In the Uniform Building Code (22) , the United States is divided into seven wind intensity zones. In the A58.1 Standard and the NBC, the design wind speeds are based upon

formal statistical analyses of wind records, i.e., they are defined in probabilistic terms on the basis of assumed probability distribution functions and of specified mean recurrence intervals

.

Planetary Boundary Layer Flow: Longitudinal Wind Speed Fluctuations

It can be shown that the following ejcpression is a correct representation of the spectrum of the longitudinal wind speed fluctuations in the higher frequency range and is con- sistent with the requirements that (1) the spectral density must approach a finite value

as n 0, (2) the product nS^(P, n) reaches its maximum at some value of the frequency

below the inertial subrange, and (3) the mean square value of the longitudinal wind speed

fluctuations is v'2(p) = 6.0 u^

:

nS^(P, n)/u^ = 200 f/(l + 50 f ) (19)

in which f = n(z - z^)/v(z) is known as the similarity (or Monin) coordinate. In Equation

(19) the peak similarity coordinate (the value of f for which nS^(n) is a maximum) is

fpj, = 0.033. As noted, for example, in Refs. 15, 19, in reality fpj^ varies rather erratically between sites and between atmosphere and laboratory, as well as with height above

ground. Orders of magnitude of f^j, are suggested in Refs. 15 and 19. The effect upon

structural response of the variation of fpj^ can be verified as will be subsequently done

in this paper by using alternative expressions for the spectra in which the peak similarity coordinate appears as a parameter. Such expressions can be found in Ref. 19 (Equations 5, 9, and 10)

.

In the A58.1 Standard and in the NBC, as well as in Vickery's procedure (25), an ex- pression for the spectrum of the longitudinal wind speed fluctuations is assiimed which is

invariant with height (6) and thus violates the similarity requirements which prevail in

the higher frequency range (14, 15, 19) . From a structural engineering point of view,

the drawback of this expression is that it over-estimates the spectral densities at frequencies equal, or nearly equal, to the natural frequencies of tall structures and that

it thus results in over-estimates of the resonant part of their dynamic response (19)

.

It is convenient to express the cross-correlation coefficient R^iP-^' ^2' ^®

form

R^(P-|^, n) = R(yi, z-|_, y^, Z2 ^1 ' (2°^

IV-132
in which y-j^, Zj^ are the coordinates of point P^(i = 1, 2) in a plane perpendicular to the

direction of the mean wind, R(y]^, z-^, ^2' ~ acrosswind, and N(P^, n) = alongwind correlation coefficient.

The following expression for the acrosswind correlation coefficient has been proposed

by Davenport (7)

:

R(yi/ z^/ Y2' ^2' = exp(- 2nv^c2(zi - z^)'^ + (y^^ - y2)^/[v(zi) + v(z2)]}(21)

in which - 10/ ~ 16 (25) . By definition, if the points P-|^, P2 are in the same plane

perpendicular to the mean wind direction, N(Pj, V n) E 1. Otherwise, N(P-]^, P2, n) < 1; its magnitude decreases as the alongwind separation between P-|^, P^ increases ana as the

frequency n increases. For the purposes of Equation (5), useful information regarding the

magnitude of N(P^, P2, n) is provided by results of measurements of cross-spectra of pressures on the windward and leeward sides of structures. Such measurements, carried out on

a full-scale building, have been reported by Lam Put (9), who found that N(P-]^^, P2 , n) < 0.2 (the subscripts w, 1 indicate that the points P-j^, P2 are on the windward and leeward

sides, respectively; points Pi / Pjj, and P2 , ^2^ have coordinates y]^, and y2, Z2, respectively) . Form a comparison of the results of wind tunnel measurements reported by Kao

(Figures 5.1, 5.19, 5.21 of Ref. 8), it also follws that, except for extremely low frequencies corresponding to eddies of negligible energy, N(P]^^, ' -"-^ nearly zero.

Alongwind Response

From Equations (5), (12), (13), (15), and (16), it follows that for a building rec- tangular in plan, of height H and width B, the spectral density of the fluctuating response

at a point M is

B B H H S^(M, n) = p2/ J / / (c2 + 2C^C^N(n) + C^) H*(M, P-^, n)H(M, P2, n) 0000

v(Zj^)v(z2)S-'-/^(P3^, n)sy^(P2, n)R(y^, z-^, ^2' ^2' n) dy3^dy2dz^dz2 (22)

In Equation (22), the identities N(Pi , P, , n) = N(P-[^ , P2 , n) e 1, and the notation

N(Pi^, ^2i>

= ^^^ll' ^2^' = N(ny are^used. ^

^

The mean square value of the dynamic response is

cc a^TzT =

/ S^(M, n)dn (23) o

and the maximum probable fluctuating response can be written as 1/2

(24) max

'

in which the peak factor is (6)

g(z) = [21nv,(z)T]l/2 + 0.577/[21nv (z)T]l/2 (25)

T = duration of v;ind loading and

v2(z) = / n2s^(M, n)dn/Z^T^ (26)

^ J a

o The mean response may be written in terms of generalized masses and forces as (19) H

Vjj,(z) / v2(z)y^(z)dz

o

a(z) = i(C + C )p B ^ H (27)

2 r 47T^n2 / m{z)\i^(z) dz

r o r

IV-133
If the contribution of the higher modes of vibration can be neglected, it is convenient to write the mean square value of the response as, approximately, the sum of the "back

ground" and a "resonant" response (6, 24, 25) . It follows then from Equations (22) and

(23) ,

a" (z) 2 p 2p2(2) (3 + j^)/j_g tt 'i^^^^

(28)

in which the quantities B and K, which are proportional to the background and resonant re- sponse, respectively, have the expressions

B = / [/ / / / ^a^Yi' y2' ^1' ^2' dy-|^dy2dz-j^dz2]dn (29) o o o o o

TTnj B B H H K = T~/ / / / la^^l' ^2' ^1' ^2' "l^ dyj_dy2dzj^dz2 (30)

1 o o o o

"a^^l' ^2' ^1' ^2' " ^^w 2C^S'^^'^^ C;^]p^(z^)pj^(z2) v(z^)v(z2)

sy^{P^, n)sj/2(p^, n)R(y^, y2, z^, Z2) (31)

Numerical integrations carried out using the computer program briefly described in Ref. 19 show that the approximation involved in Equation (28) is of the order of 1%.

The total, maximum deflection in the alongwind direction is •

a(z) = [1 + g(z) P^2Yi')l/2/-(2) ] ^(z) (32)

The quantity between brackets in Equation (32) is known as the gust response factor. The

equivalent static load is then the product of the mean load and the gust response factor.

BASIC FEATURES OF PROCEDURES FOR COMPUTING ALONGWIND RESPONSE

The framework described in the preceding section will now be used to discuss certain

procedures that have been proposed for the calculation of alongwind response.

Procedure Described in the A58.1 Standard (1)

The mean wind speeds specified for design purposes by the A58.1 Standard correspond

to 50- and 100- year mean recurrence intervals based on Type II probability distributions

of the largest values. In the calculations, hourly means are used, which can be easily

obtained if the fastest-mile speeds are known (24) . The mean pressures on both the windward and the leeward sides are proportional to the square of the mean speeds, which are assumed to obey the power law. The pressure coefficients are = 0.8 and = -0.6 or

Cj^ = -0.5 (for H/B ^2.5 or H/B < 2.5, respectively). The wind spectra are assumed to be

invariant with height. As pointed out previously, this assumption results in overestimates

of the dynamic part of the response. The acrosswind correlation coefficient is essentially

similar to that given by Equation (21) . The alongwind correlation coefficient is assumed

to be

N(Pi , Pp, n) = i - JL(1 - e-2^) - (33)

^w 2 ^ ^^2 _ _ H in which C = 3.85 nAx/v, Ax is the smaller of the distances 4B or 4H and v = / v2(z) dz/H.

o Although, by definition, N(P-l , P2 , n) h N(P2 , P2j^, n) = 1, it is implicit in this procedure that these quantities Sre h&th equal to'w(Pi^, P2£' and, as shown in Ref. 17,

this results in underestimates of the dynamic part of the response. A value of the peak

factor g(z) = 3.0 is used, rather than the more correct value given by Equation (25) which,

IV-134
for T = 3600 s, is usually g(z) = 3.4 - 4.0. The contribution of the higher modes of vibration is not taken into account. As will be shown subsequently, in most cases of practical interest this contribution may be neglected.

Procedure Described in the National Building Code of Canada (5, 11)

For design purposes, the NBC specifies mean speeds averaged over one hour and corre- sponding to 30- and 100-year mean recurrence intervals based on Type I probability distributions of the largest values (7) . A statistical analysis carried out by the writers of

30 extreme wind records taken over periods of 20 to 45 years showed that these speeds are approximately 6% lower than those specified by the A58.1 Standard. Since the mean response

is proportional to the square of the mean speeds, the corresponding difference between the

respective mean responses is about 12%. The mean pressures on the windward side are proportional to the square of the mean speeds, which are assumed to obey the power law, except

that in the lowest 30 ft. (9.15 m) in open terrain and 100 ft. (30,5 m) in urban terrain

the mean pressures are assumed to be uniformly distributed (11) . A pressure coefficient

C., = 0.8 is used. The mean pressures on the leeward side are assumed to be distributed

uniformly and are proportional to the square of the mean speed at an elevation H/2, where

H is the building height, with C^ = - 0.5. (This assumption results in a further increase

of the difference between mean responses calculated by Refs. 1 and 11). The same expression for the wind spectrum and essentially the same expression for the acrosswind correlation coefficient as in the A58.1 Standard are used. Thus, in this procedure, the variation of the spectrum with height is not taken into account and, correspondingly, overestimates of the dynamic response result. The alleviating effect of the alongwind correlation

is also not taken into account, i.e., the assumption N(P]^ , P2£' n) = 1 is implied, which

results in further overestimates of the dynamic response Yl7 ) . Equation (25) is used to obtain the peak factor and the higher vibration modes are not taken into account.

Vicker's procedure is essentially similar to that of P.ef. 11, over which it has the

advantage of being more flexible in that it permits the variation of certain parameters

(decay coefficients C^, in Equation (21) and an exponent defining the modal shape)

The peak factor is assumed to be g(z) = 3.5.

Procedure Based on Refs. 18 and 19

In this procedure the mean wind speeds are described by the logarithmic law (Equation

(17)), and the variation of the wind spectrum with height is taken into account as shown

in Ref. 19. Between the elevation (.z^ + 10) meters and the top of the building the mean

pressures are assumed to be proportional to the square of the mean speeds. In the lowest

(z^ + 10) meters, the pressures are assumed to be urjiformly distributed and to have the same value as at elevation {z^ + 10) . The pressure coefficients assumed in the calculations presented in this paper are C^ = 0.8 and C^ = - 0.5. In view of the uncertainty

regarding the actual probability distributions of extreme wind speed (7) , as well as the

optimum recurrence intervals to be used in structural design, calculations were carried

out using mean speeds specified by both the A58.1 Standard and the NBC. The peak factor was assumed to be given by Equation (25) , and the acrosswind correlation coefficient given

by Equation (21) was used. In computing the background response it was assumed conservatively that N(P„» n) E 1. In computing the resonant response, the assumption N(P^, P^,

n-^) =0. , in which n-j^ = natural frequency in the fundamental mode was used. Alternatively,

in computing the resonant response it was assumed that Equation (33) is valid. The differences between total response calculated using these two assumptions turned out to be

insignificant (0-3%)

In the expression for the wind profiles, the values z^ = 0.07 m, z^ = 0, and z^ = 0.8 m, z^ = 21m for open and urban terrain, respectively, were used in calculating the gust factors

and the deflections presented in the following. To verify the extent to which the response

might be sensitive to variations of the roughness parameters, the values z = 1.2 m, z^ = 10m were also used for urban terrain as noted by Pasquill (15) , experimental results

suggest that in urban terrain, as wind speed increases, z increases and decreases,

possibly because the air stream "penetrates" more deeply Between buildings. The results

obtained using these values differed in all cases by less than 3% from the results

presented herein.

IV-135
The effect upon the structural response of the possible variation of the peak simi- [ larity coordinate f^-^ was also verified. The dynamic response was calculated using Equation (19) (to which there corresponds f , = 0.033), as well as alternative expressions for

the spectra (Equations (5), (9), (10) of Ref. 19) in which f , appears as a parameter. The ratios of the root mean square values of the fluctuating response [a ' (h) 2] ^/2/ [a ' (h) ^]q'^' (where the numerator is based on Equations (5), (9), (10]_ of Ref. 19 and_the denominator

is based on Equation (19)), were calculated in the case Vj^ = 26.8 ms""*- (v^ = wind speed

at 30 ft. above ground in open terrain) for the typical buildings described in the following section of this paper. For f^j^ = 0.01, this ratio was found to be approximately one in all cases. In the case of the 365 m tall building, the ratio was about .95 for fpj, = 0.1 and .93 for f = 0,19. In the case of the 150 m tall building, the ratio was about

.93 for f^y. = O.ID and .92 for f = 0.19. In the case of the 45 m tall building, the

ratio was about .85 for f , = 0.1. It is noted that, for H < 45 m, on average, f^j^ < 0.1.

Equation (19), which is shown by these results to be moderately conservative, was, therefore, used in calculating the values of the dynamic response presented in this paper.

For the three typical buildings described subsequently, the contribution to the dynamic

response of the second and third modes of vibration was calculated. It was found that if, as was assumed in the calculations presented herein, the natural frequencies in the second

and third modes are approximately three and six times higher than the fundamental frequency,,

respectively, the contribution of the higher modes to the response is insignificant.

ANALYSIS OF NUMERICAL RESULTS

For the purpose of comparing the procedures examined in this paper, the gust response

factor and the maximum alongwind deflections were calculated for three typical tall buildings previously analyzed by Vickery (25) . The geometric, as well as the mechanical, properties of the buildings are shown in Table 1. In all cases, an average building weight

of 1,500 N/m-^ was assumed. For steel structures, both the A58.1 Standard and the NBC suggest a damping ratio 3^ = 0.01. This ratio was used in calculating both gust response

factors and maximum alongwind deflection. In addition, gust factors were calculated as- suming higher damping ratios (including mechanical and aerodynamic damping) . For v^ =

T — —1

26.8 ms"-^, these ratios. were assumed to be the same as used m Ref. 25. For Vj^ = 40 ms

the damping ratios were increased with respect to those of Ref. 25 by amplifying their

aerodynamic part in proportion with the wind speed, as shown in Ref. 26. The results of

all the calculations are given in Tables 1 and 2.

It can be seen from Table 1 that the gust factors computed in accordance with the

A58.1 Standard are in all cases lower than those detemined on the basis of the NBC, the

differences ranging from about 15% or more in the case of the 45 m tall building in open

terrain to almost 100% in the case of the 365 m tall building in urban terrain. The

values of Column (4) were calculated on the basis of improved models of the mean profiles,

of the wind spectra and of the alongwind pressure correlations (17, 18, 19) . If these

values are regarded as being correct (even though, as shown previously in this paper and

in Ref. 19, they are likely to be slightly conservative) , then it may be stated that (1) the values of the gust factors (Equation (32)) based on the A58.1 Standard are low, by

amounts ranging from 1% in the case of the 365 m tall buildings in open terrain to 20% in

the case of the 150 m tall building in urban terrain, and that (2) the values of the gust

factors based on the NBC are high, by amounts ranging from 1% in the case of the 45 m

tall building in open terrain to 63% in the case of the tall building in urban terrain.

The differences between the gust factors obtained by the various procedures described

in the preceding section are due primarily to the respective assumptions regarding the

alongwind pressure correlations and the expression for the wind spectrum. The alongwind

IV-136
correlation model used in the NBC results in ein overestimate, whereas the model used in

the A58.1 Standard results in an underestimate of the dynamic part of the response. The

assumption used in both Refs. 1 and 11 that the spectrum is invariant with height results

in overestimates of the response, which are negligible in the case of lower structures

but become substantial in the case of very tall structures. In the case of the NBC, these

overestimates are added to those due to the assun^itions regarding the alongwind correlation effect. In the case of the A58.1 Standard, the deviation from the actual value of

the response due to the assuit$)tion that the spectra are invariant with height and the deviation due to the assunption N(P]^^, P2^' ~ ^(^Ij^' P2 ' -'- °^ opposite signs.

Their sum is in most cases negative, as follows from the comparison between Columns (1) and (4) .

It can be seen from Table 2 that the mean response determined in accordance with NBC

is smaller than that based on the A58.1 Standard. This is the case because, first, for

a given wind climate the mean wind speeds specified by the NBC are lower than those specified by the A58.1 Standard; secondly, the leeward pressure coefficient given in the NBC

is = - 0.5, whereas that specified by the A58.1 Standard for buildings of ratio

B/H >_ 2.5 is C^ = - 0.6; thirdly, the contribution to the mean response of the leeward

pressures is lower in the case of the NBC (in which these pressures are assumed to be uniformly distributed) than in the case of the A58.1 Standard. It is also noted that, assum- ing equal wind speed conditions, the values of the mean response determined in accordance

with the procedures based on Refs. 18, 19, are in most cases slightly larger than those

obtained using the A58.1 Standard. This shows that the power law model of the mean wind

speeds is slightly unconservative

.

For equal wind speed conditions, the difference between values of the total response

determined in accordance with Refs. 18, 19 and with the NBC is relatively small in the case of the 45 m and 150 m tall buildings (2%-13%) . In the case of the 365 m building in

urban terrain, however, this difference is almost 50%. On the other hand, the differences

between total response determined in accordance with Refs. 18, 19 and with the A58.1

Standard are larger for the 45 m and 150 m tall buildings (20%-33%) and are smaller for

the 365 m tall buildings (6%-18%)

.

The maximum deflections calculated in accordance with the provisions of the New York

City Code (13) , the Uniform Building Code (UBC) (22) , and the South Florida Code (21) are also included in Table 2. • . . , -

For a basic design speed of 75 mph and urban terrain, the New York City Code is quite

conservative for the 45 m building and somewhat unconservative for the taller buildings

(assuming that 50-year mean recurrence interval is appropriate) . The Uniform Building

Code, v^ich assumes an open country exposure, is extremely conservative for the 45 m building and slightly conservative in the case of the 365 m building. For a basic design speed

of 110 mph and open country terrain, the South Florida Code underestimates the response

of all three buildings. The Uniform Building Code slightly underestimates the response

of the 45 m building and seriously underestimates the response of the 365 m building.

CONCLUSIONS

A con53arative analysis of current procedures for evaluating alongwind structural re- sponse has been presented. The analysis confirms that considerable discrepancies exist

between results obtained using these procedures. For example, in the case of a 365 m tall

building in urban terrain the procedure described in the NBC (5, 11) yields values of the

dynamic part of the response, of the gust factor, and of the total maximum deflection

which are, respectively, three times, twice, and 1.6 times larger than those based on the

A58.1 Standard (1) . The basic assuiiptions used in the various procedures have been exam- ined and their shortcomings have been pointed out. An improved procedure, based on Refs.

18, 19, has been described. Numerical calculations were carried out, based on this, as well as on the other procedures investigated. Typical buildings of 365 m, 150 m, and

45 m height, previously analyzed by Vickery, were used as case studies. For purposes of

comparison, the results obtained by the procedure based on Refs. 18, 19 were calculated

IV-137
using the probabilistic definitions of design wind speeds specified in the A58.1 Standard

and in the NBC. The question of these definitions is, however, beyond the scope of this

paper. Assuming that these definitions are acceptable, the results of the calculations

showed that the procedure described in the A58.1 Standard was unconservative in most cases.

On the other hand, the procedure described in the NBC was slightly unconservative for

the 45 m buildings and was over conservative for buildings of 150 m and 365 m height. Thus,

while in certain cases the procedures described in both the A58.1 Standard and the NBC may

yield reasonable results, their fundamental shortcomings may be the cause of considerable

underestimates or overestimates of the alongwind structural response. The writers believe,

therefore, that the use of the procedure based on Refs. 18, 19, which incorporates recently

developed, improved models of the mean wind profiles, the wind spectrum and the pressure

correlation in the alongwind direction, and which is described in the body of the paper, is

justified in the case of structures in the design of which the wind loading is a major

consideration

.

REFERENCES '

•

'

'

1. "American National Standard A58.1 - Building Code Requirements for Minimum Design

Loads in Buildings and Other Structures," American National Standards Institute, New

York, New York, 1972.

2. Baines, W.D., "Effects of Velocity Distribution on Wind Loads and Flow Patterns on Buildings," Proceedings , Wind Effects on Buildings and Structures, Teddington, 1963, Her Majesty's Stationery Office, London, 1965.

3. Batchelor, G.K. , "The Theory of Homogeneous Turbulence," Cambridge University Press,

Cambridge, England, 1953.

4. Blackman, R.B., and Tukey, J.W. , "The Measurement of Power Spectra from the Point of

View of Communication Engineering," Dover Publications, Inc., New York.

5. "Canadian Structural Design Manual," Supplement No. 4 to the National Building Code

of Canada, National Research Council, Ottawa, 1970.

6. Davenport, A.G., "Gust Loading Factors," Journal of the Structural Division , ASCE,

Vol. 93, No. ST3, Proc. Paper 5255, June 1967, pp. 11-34.

7. Davenport, A.G. , "The Dependence of Wind Loads on Meteorological Parameters," Proceedings , Wind Flffects on Buildings and structures. Vol. I, Ottawa, 1967, University

of Toronto Press, Toronto, 1968.

8. Kao, K.H., "Meas\arements of Pressure/Velocity Correlation on a Rectangular Prism in

Turbulent Flow," Engineering Science Research Report BLWT-2-20, University of Western

Ontario, London, Canada, 1971.

9. Lam Put, R. , "Dynamic Response of a Tall Building to Random Wind Loads," Proceedings , International Conference on Wind Effects on Buildings and Structures, Tokyo, Japan,

1971, pp. Ill - 4.1 - 11.

10. Lumley, J.L. , and Panofsky, H.A., "The Structure of Atmospheric Turbulence," John

Wiley and Sons, 1964.

11. National Building Code of Canada, National Research Council, Ottawa, 1970.

12. Newberry, C.W. , Eaton, K. J. , and Mayne, J.R. , "Wind Loading of a Tall Building in an Urban Environment," Building Research Station, Garston, Watford, England, 1968.

13. "New York City Building Code," Local Law No. 76 of the City of New York, New York,

New York, 1969.

IV-138
14. Owen, P.R. , "Buildings on the Wind," Quarterly Journal of the Royal Meteorological

Society, Vol. 97, 1971, pp. 396-413.

15. Pasquill, F., "Wind Structure in the Atmospheric Boundary Layer," Philosophical Transactions of the Royal Society of London, A., 269, 1971, pp. 439-456.

16. Robson, J.D., "An Introduction to Random Vibration," Elsevier Publishing Company,

1964, p. 91.

17. Simiu, E., "Gust Factors and Alongwind Pressure Correlations," Journal of the Structural Division , ASCE, Vol. 99, No. ST4, Proc. Paper 9686, April, 1973, pp. 773-783.

18. Simiu, E., "Logarithmic Profiles and Design Wind Speeds," Journal of the Engineering

Mechanics Division , ASCE, Vol. 99, No. EMS, Proc. Paper 10100, October, 1973, pp. 1073-1083.

19. Simiu, E., "Wind Spectra and Alongwind Structural Response," Meeting Preprint 2086,

ASCE Annual and National Environmental Engineering Meeting, October 1973, New York.

20. Simiu, E., "Improved Methods for Determining Wind Profiles and Dynamic Structural

Response to Wind," Proceedings , ASCE-IABSE Regional Conference on Tall Buildings,

January 1974, Bangkok.

21. "South Florida Building Code," Dade County, Florida, 1970.

22. "Uniform Building Code," International Conference of Building Officials, Whittier,

California, 1973.

23. Vaicaitis, R. , Shinozuka, M. , and Takeno, M. , "Response Analysis of Tall Buildings

to Wind Loading," Technical Report No. 1, Dept. of Civil Engineering and Engineering

Mechanics, Columbia University, July 1973, New York, p. 74.

24. Vellozzi, J., and Cohen, E., "Gust Response Factors," Journal of the Structural Division , ASCE, Vol. 94, No. ST6, Proc. Paper 5980, June 1968, pp. 1295-1313.

25. Vickery, B.J., "On the Reliability of Gust Loading Factors," Proceedings , Technical

Meeting Concerning Wind Loads on Buildings and Structures, National Bureau of Standards,

Building Science Series 30, Washington, D.C., 1970.

26. Vickery, B. J. , and Kao, K.H. , "Drag of Alongwind Response of Slender Structures,"

Journal of the Structural Division , ASCE, Vol. 98, No. STl, Proc. Paper 8635, January

1972, pp. 21-36.

IV-139
NOTATION

The following symbols are used in this paper:

A = Area

B = Width of structure

Cw = Added mass coefficient

C = Mean pressure coefficient

= Mean pressure coefficient - windward face

= Mean pressure coefficient - leeward face

C , C„ = Dimensionless coefficients (defined in text)

F(t) = Stationary random function

H = Height of structure

H(M, P, n) = Mechanical admittance function

H (M, P, n) = Complex conjugate of H(M, P, n)

M = Denotes a point on the structure

M = Generalized mass term

rN = Alongwind correlation coefficient

P = Denotes a point on the structure

R = Square root of coherence function

S (n) = Spectral density function

T = Duration time

a = Structural response

f, fp]^ = Similarity coordinate, peak similarity coordinate

g = Peak factor

m = Mass of structure per unit height

n = Frequency

p = Mean pressure

p' = Fluctuating pressure "

r = Subscript indicating r^^ mode

s = Subscript indicating s^'^ mode

t = Time

u^ = Friction velocity

IV-140
v(z) = Mean wind velocity at height z

v'(z) = Velocity fluctuation at height z

z = Height above ground

z^ = Zero plane displacement

Zq = Surface roughness length

6 = Daitping ratio

Ax = Horizontal separation in the alongwind direction

y = Modal shape

V = Effective fluctuation rate

C = 3.85 nAx/v

1, 2 = Subscripts denoting values of a quantity at points 1, 2

IV-141
0.2 0.5 1.0 1.5 2.0 2.5 3.0 3.5

ROUGHNESS LENGTH 'zq , IN METERS

FIGe 1 - Ratios

IV-142
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IV-144
SEISMIC RETROFITTING OF EXISTING HIGHWAY BRIDGES

by

James D. Cooper

Structural Research Engineer

Office of Research

Federal Highway Administration

Washington, D.C.

R. Richard Robinson

Research Engineer

Anatole Longinow

Manager of Structural Analysis

IIT Research Institute

Chicago, Illinois

The retrofitting of existing highway bridges, to provide an added measure of

protection against collapse due to earthquake ground motion, is of great importance

This interest heightened in the United States following the San Fernando earthquake

of 1971, which caused extensive damage to a number of modern freeway structures.

Some of the specific concepts for retrofitting to be explored include: (1) widening of bearing supports, (2) motion restrainers across hinges, (3) ties across expansion joints, (4) the elimination of expansion joints, and (5) adding ties or reinforcing to existing columns.

The monetary savings, resulting from an effective retrofit program in preventing collapse of structures, would far exceed the cost of the research involved in

generating feasible and practical retrofit details.

This is a progress report on research which will result in mathematical techniques to identify the seismically vulnerable bridge details and a catalog of retro

fit techniques. Such techniques will permit strengthening of such weak links, in

the total structure integrity.

Key Words: Design; earthquakes; highway bridges; retrofitting; soil-structure

interaction; structural engineering.

V-1
INTRODUCTION

In the San Fernando earthquake of February 9, 1971, numerous bridges were destroyed

even though this earthquake was only of moderate magnitude, approximately 6.5 on the

Richter Scale. Damages to highways have been estimated at $22.5 million to the State

system and $5 million to city and county roads. The chief damage to the State system was

the collapse of five overpasses and damage to an additional seven.

The western part of the United States experiences a severe earthquake on the average

of every four years and it is reasonable to assume that any future earthquake of moderate

to large magnitude can pose a threat to existing bridges unless steps are taken to in- crease their structural resistance to seismic loading. Consideration is, therefore, being

given to developing economically feasible methods for increasing the seismic resistance

of existing bridges through retrofitting measures in order to minimize possible future

damage

.

A report (1) * to the Federal Highway Administration on a reconnaissance survey made

on February 12-13, 1971, of the damage to bridges in the San Fernando earthquake contains

the following excerpts:

C. Causes of Collapse

The two principal causes of collapse of the high overcrossings are considered

to be: (1) large vibratory motions induced in the superstructures by the high intensity vertical and horizontal ground accelerations, and (2) relative ground displacements which may have occurred between the abutment and column supports. It

is the opinion of the authors that the former was the major cause of collapse in

these particular cases; there is no evidence that relative ground displacements

were a contributing factor.

D. Design Considerations

1. Expansion Joints - Collapse of the high overcrossings was initiated by

bridge spans falling off their supports at abutments

and expansion joints due to excessive displacements

of the spans relative to their supports. This type

of behavior should be carefully examined and corrective measures should be taken as soon as possible.

Full consideration should be given to eliminating ex- pansion joints wherever feasible, to widening bearing

supports, and to providing more effective ties across expansion joints.

2. Columns - The failures in the central portion of the shorter stiffer columns were caused by the transverse shear forces, while the end failures

in the larger more flexible columns were caused by flexural forces

In each case, there was a noticeable lack of transverse column

ties which contributed to these failures. Clearly, the design

details of columns should be carefully examined, particularly with

regard to size and placement of reinforcing bars and ties, and

corrective measures should be taken to improve their performance

under ultimate loading conditions involving reverse deformation

cycles such as occur during major earthquakes.

3. Column Caps - There appears to be a serious lack of reinforcing bars tying

column caps into their respective box girder bridge decks.

Corrective measures should be taken to improve this design

detail.

I

Numbers in parenthesis refer to references at the end of the paper.

V-2
4. Column Foundations - Failures at the base of columns for both types of supports, i.e., single cast-in-place pile and spread foot

ings with driven piles, showed inadequate anchorage of

the main reinforcing bars. Corrective measures should

be taken in each case so that sufficient anchorage is

provided to develop the full strength of the main rein

forcing bars

5. Abutment and Wing Walls - Failures in abutments and wing walls were caused

by large dynamic forces transmitted by backfill

earth pressures and by seismic forces developed

in the bridge decks. The design details of these

structures should be re-examined and appropriate

corrective measures should be taken to improve

their performance characteristics.

The California Division of Highways has already revised some of its designs for new

construction. For exaitple, column spiral splices must now be anchored by extending the

spiral wire through the column and anchoring it around vertical bars on the opposite side

of the column. Another example consists of revisions to the hinge restrainer to increase

the pull-apart resistance of the joint.

Following the San Fernando earthquake, California Division of Highway personnel iden

tified 860 vulnerable existing bridges requiring a $20 million program for correcting the

most obviously vulnerable structural detail through installation of horizontal motion re- strainers at bearing seats and expainsion joints. A $5 million progrcim was authorized for

the 120 bridges deemed most critical. To date this is the only known attention given to retrofitting existing highway bridges to provide increased seismic resistance.

The objective of the retrofit program is to identify and define through the use of

structural analyses and supporting laboratory tests practical techniques and criteria for

retrofitting existing bridges to increase their resistance to seismic forces. The follow

ing ten concepts for retrofitting based on post-event seismic damage observations on bridges in Alaska, California, and Japan are currently being evaluated.

1. Techniques to wide bearing supports and irrprove bearing stability.

2. More effective motion restrainers across hinges or more effective ties across expansion joints.

3. Practical means to eliminate existing expansion joints, where feasible, in

existing bridges.

4. Methods of restraining bridge slabs from collapsing such as by cabling on to

the supporting or an adjacent foundation of the bridge.

5. Practical means to add additional ties or spiral to an existing column.

6. Epoxy pressure grouting of concrete hairline cracks already existing to elimi- nate discontinuities in the concrete.

7. Practical means to in^rove the ties of column caps into their respective box

girder bridge decks.

8. Develop a practical means of preventing collapse of bridge piers or columns

in the event foundation piles disintegrate during an earthquake.

9. Introduction of interface materials between abutments or wind walls and backfill such as expanded polyeurethane or other suitable frangible materials that

will cushion the seismic earth pressures.

V-3
10. Attaching elastomeric bearing pads to concrete with thioxol or other suitable

material to prevent the vibratory displacement of the pads during shaking.

At the conclusion of -'Sae retrofit investigation, design details and installation

specifications will be developed for those techniques deemed to be practical and cost effective.

SELECTION OF REPRESENTATIVE BRIDGES AND EARTHQUAKE INPUT MOTIONS ,

Seven representative types of highway bridges from several states situated in seismic

Zone 3 regions are being investigated in the study. Initially, State Highway Departments

in Alaska, California, Illinois, Massachusetts, Missouri, Montana, New York, and South

Carolina participated in the study by selecting representative bridge types for possible

use in future stages of the study. Six bridges were selected from those submitted—one each from Alaska, Illinois, Montana, and New York and two from California. Table 1 gives

bridge location and description of each of the structures. The seventh structure, an eight-span, curved concrete box girder bridge with three expansion joints located in California, has been incorporated into the study. The 607-foot long interchange connector

type bridge is typical of many of the structures located in populated metropolitan areas. This bridge was added to investigate and further develop the concept of bridge hinge re- strainer systems.

Given the life of each bridge structure and the soil boring data for each of the seven bridge sites, effective Richter magnitude and hypocentral distances were calculated.

Site spectra for ground motions in the horizontal and vertical directions were generated.

Statistical relationships between lateral and vertical spectra experienced in past historical earthquakes were utilized. An analysis earthquake including an acceleration, velocity

and displacement versus time history for the horizontal and vertical directions of each

bridge site was generated from site spectra.

Table 2 outlines structure locations, type, and region from which seismic input mo- tion has been generated for use in tho analysis and development of retrofit concepts.

... BRIDGE ANALYSIS COMPUTER PROGRAM . -v

This computer program was developed for the purpose of analyzing the dynamic struc- tural response of highway bridges when subjected to ground motions produced by an arbitrary

earthquake. Its specific purpose is to evaluate the merits of various retrofit measures

that would be employed in a given bridge for the purpose of eliminating or reducing struc- tural damage. The program is based on the finite element method of structural analysis

and models a given bridge as a three-dimensional (space) frame. As such, six degrees of

freedom are allowed per node.

The method of analysis employs a nonlinear dynamic response analysis of the structure.

An implicit integration solution scheme employing equilibrium checks is used to solve the

incremental form of the equations of motion. A tangent stiffness m.atrix for the complete

structure is reassembled at user-defined arbitrary increments of the integration step.

This feature, coupled with the equilibrium checks and the stable implicit inegration technique permits one to feasibly obtain solutions to intermediate size problems (up to 1000

degrees of freedom) subjected to long duration loading (up to 50 sec). The most out- standing feature of the computer code is its ability to model the reduction in load-carrying capacity of a member subjected to relatively small overstressing for cyclic loading.

This is a most serious behavior characteristic which is largely responsible for structural

component failures, such as reinforced concrete columns. The elastic-plastic yield surface technique usually employed to model beams subjected to overstressing does not readily

lend itself to conveiaently modeling this behavior, especially for members of complex

cross-section subjected to combined thrust and bending loading about two axes.

V-4
The computer code treats the nonlinear response of overstressed members by employing

a special elastic-plastic beam element that realistically models the behavior jLn a simple

to use, automatic manner, that is particularly adaptable to nxamerical analysis. Recalling

the formal development of a finite element beam model, the constitutive equations of the

material enter the analysis through integrals over the beam cross-section which have the

form

/AAA a (e )dA ; / a{e)ydA ,- / a (e ) zdA ; ja (e )yzdA

where a (c) is the stress at any fiber location (y, z) and £ is the associated strain. The

strain can be obtained in terms of the extension, e^, and the curvatures ^oy ^oz"

The method employed in the computer code evaluates these integrals by numerical integration over the cross-sectional area at each stage in the deformation of the structure.

The user subdivides the cross-section of each of these beam elements into a finite number

of incremental areas. Each of the subdivided areas is specified as a particular type of

material, such as concrete or steel, and as such has a certain user defined nonlinear

stress-strain behavior. A knowledge of the current strain and the previous stress-strain

path for each subdivided area provides the information necessary to deteirmine the current

stress and hence evaluate the integrals to compute the current force resultants.

This concept is ertployed to derive the incremental force-deformation relationships

at each of the becun elements two node points (six degrees of freedom per mode) . This information is then employed to integrate over the length of the besim to obtain the timevarying tangent stiffness matrix for these elements.

A soil-structure interaction finite element is provided in the program to model the

connection of the bridge structure to the ground. This element simulates the connection

by employing three translational and three rotational springs and corresponding viscous

dampers. The prescribed seismic motion is imposed at the ground node point of the soilstructure interaction finite element. At the user's option, different spring stiffnesses

can be employed for the compression and tension translational degrees of freedom. This

feature is quite useful for seismic loading, due to the cyclic nature of the imposed

motion, since it permits one to use a high stiffness for compression interaction and a low or zero stiffness for tension interaction for simulating a footing on piles or a spread footing.

A nonlinear expansion joint finite element is included to provide the analyst with

a means of accurately modeling connections between different spans of the superstructure,

superstructure-abutment (or pier) interactions, and hinges. It models the expansion

joint gap, includes coulomb friction and variable spring rates to simulate differences in

con^iression and tension resistance either horizontally or vertically. Tie bars can also

be modeled in the expansion joint element.

An elastic beam finite element is also provided in the computer code. It is used to model those components of the structure that will not be sxibjected to severe overstressing and yielding due to the dynamic loading. It is generally employed to model the majority of the superstructure. Shear deformations are optionally included in the elastic beam

element stiffness and stress-displacement matrices.

The computer program was tested on the Bahai Overcrossing in California. Plan, elevation, and cross-sectional views of the two-span, four-cell reinforced box girder bridge

are shown in Figures 1 and 2. Figure 3 depicts the reinforcing steel and concrete stressstrain relationships after Brown and Jirsa (2) . Note that the ultimate stress for the

reinforcing steel is rather low. The steel stress-strain curve for this test was selected

so that low axial tensile strength in the pier elements could be simulated.

A one-cycle sinusoidal vertical displacement waveform with period of 0.2 seconds and

double amplitude of two inches was used as a simple test input wave motion at the base

V-5
of Bent No. 2. Selected displacements, bending moments, and axial forces are shown in

Figures 4 through 7. The pier elements fail very near the 0.2 second mark, after which

time the superstructure deflects downward a very large amount. Computer runs of the seven structures identified previously are currently being carried out to identify failure mechanisms in the structures.

PRELIMINARY RETROFIT CONCEPTS

Retrofit techniques are in the process of being developed for the seven representative

bridge types described previously. It is anticipated that the final fixes selected will

be valid for use on other similar types of bridges located within seismically active areas.

Table 3 lists controlling items and methods of increasing seismic resistance of six

of the seven bridges under investigation. Some initial concepts have been proposed following the review of detail plans of each of the structures. Analysis of each of the

structures is currently being made. Analytical results should help identify additional

concepts and verify some of the proposed concepts. Standard detail drawings will be made

for each of the final retrofit concepts selected for use.

Some of the initial retrofit concepts envisioned to date are shown in Figures 8 through

16. Cost effective evaluations of the concepts described have not been made.

Figure 8 shows a longitudinal section through the end diaphragms of a concrete box

girder. Tie bars have been added to limit longitudinal joint separation. This system

has been used by the State of California in their retrofit program. A concept for limiting relative displacement at abutments is shown in Figure 9. Vertical uplift can be con- trolled through the extension of anchor bolts through the bottom flange of a steel girder

as shown in Figure 10. An expansion joint restrainer system for steel girder structures

is shown in Figure 11. Tie bars pass through end web stiffeners in this scheme. Figure

12 depicts a steel girder hinge restrainer system in which both vertical uplift and motion

normal to the plane of the web is controlled. Support bearing area or restriction of

girder motion can be achieved through use of concepts depicted in Figure 13. i

Use of fibrous polyester concrete for strengthening beams has been investigated.

Figures 14 and 15 show how the addition of high strength fibrous polyester concrete improve the strength and ductility of test beams. Figure 16 shows how the use of such a material could be used to enhance the seismic resistance for reinforced concrete bridge

columns. Additional concepts are in the development stage and will be fully evaluated

during the next year.

Careful analysis must be made following selection of final retrofit concepts. The

addition of retrofit hardward may enhance the seismic resistance of a vulnerable bridge

component while at the same time detract from the overall dynamic performance of the structure. Such concepts, if any, will be eliminated from further consideration.

SUMMARY

Retrofitting concepts and corresponding implementation methods for general types of

highway bridges are currently being investigated to enhance seismic resistance. The study

| is not being conducted to define fixes that will eliminate all structural damage resulting i from large magnitude earthquakes, but rather is intended to provide fixes which will assure

that: 1) structures remain in service for use as emergency routes following earthquakes,

2) damage is limited to the extent that the structure can be quickly repaired, and 3) total collapse of structures is avoided.

V-6
REFERENCES

1. Penzien, J., Clough, R.w. , "Observations on the Damages of Highway Bridge Structures,

San Fernando, California. Earthquake, February 9, 1971," report to the Division

Structural Engineers, Federal Highway Administration, Sacramento, California, April,

1971.

2. Brown, Russell H. and Jirsa, James 0., "Reinforced Concrete Beams Under Load Reversals

ACT Journal , May 1971 .. .

3. Fung, G.G. , LeBeau, R.J., Klein, E.D., Belvedere, J., and Goldschmidt, A.F., "Field

Investigation of Bridge Damage in the San Fernando Earthquake," Preliminary report

of the Bridge Department, Division of Highways, State of California, 1971.

4. "Bridge Details Resulting from Experience Gained from Los Angeles Earthquake," California Division of Highways, 1972.

5. The Great Alaska Earthquake of 1964 - Engineering , "Performance of Highway Bridge

Foundations," Ross, G.A. , et al., National Academy of Sciences, Washington, D.C.,

1973. . . .' • :

6. Cooper, James D. , "FHWA Research Program for Earthquake Resistant Design of Bridges,"

Prepared for U.S. -Japan Seminar on Earthquake Engineering with E^mphasis on Safety of

Reinforced Concrete Structxares, Berkeley, California, September 4-8, 1973.

7. "Techniques for Retrofitting Existing Bridges to Reduce Susceptability to Earthquake

Damage," Response to FHWA RFP 154, IIT Research Institute, Chicago, Illinois, March

1973.

8. Podolny, W. and Cooper, J.D., "Toward an Understanding of Earthquakes," Highway Focus

FHWA, U.S. DOT, Vol. 6, No. 1, March 1974.

9. Hanson, John M. , Wiss, Janney, Elstner and Associates, Inc., letter report to A. Longinow, IIT Research Institute, March 19, 1974.

10. Robinson, R.R. and Cooper, J.D., "Techniques for Retrofitting Existing Bridges to Reduce Susceptibility to Earthquake Damage," 70th Annual Convention, American Concrete

Institute, San Francisco, California, April 3, 1974.

11. First Federally Coordinated Program Progress Review, (Meeting), FCP Project 5A. "In^roved Protection Against Natural Hazards of Earthquake and Wind," FHWA, U.S. DOT,

San Francisco, California, September 1973. : .-. •

V-7
TABLE 1

BRIDGE DESCRIPTION

BRIDGE

1. Montana

(county Road

Separation

DESCRIPTION

h spans, each simply supported: 55', 100' 100' Type IV

2, California

(Bahia Overcrossing)

3. California

(Ventura

Overcrossing)

U. Illinois

(Rt. 2h over Massac Creek)

5. Alaska

(Bridge over

Lyon and Tin- can Creek)

6. New York

(1-508 over Delaware and

Hudson Railroad)

prestressed concrete girder superstructure with 20 degree skew.

Continuous deck integral with abutments. All ends of spans are

fixed. l|" high fixed steel bearings. Pier cap supported by

two 3 '-6" diam. columns} columns supported on separate spread

footings. Abutment serves as end diaphragm, supported on treated timber piles. Each span has 2 intermediate diaphragms

and also end diaphragm over pier.

2 span, U cell reinforced concrete box girder: 86', 102'

continuous straight 5 '-6" deep superstructure with no shew.

Sinple 3' X 8' column with bent cap in superstructure. Column

footing supported on concrete piles, Abutment and wingwalls

monolithic with super -structure and supported on concrete piles.

U span, 3 cell reinforced concrete box girder: 112 '-3", 129',

97', U7'-3". Articulated cuived 6 '-10" deep superstructure

with no skew. Single 6' diameter column with bent cap in

superstructure at all intermediate support points (18'-10", 21'-

U", 19'-1" ht for columns) Column footings supported in piles.

One abutment and wingwalls monolithic with superstructure and

supported on concrete piles. Superstructure on 10" high expansion bearings with no hold-doqn at other abutment (adjacent to

U7'-3" span). Abutment on concrete piles. One inch joint at hinge filled with asphalt latex, 5" wide bearing through

6 X h X 3A angles.

3 spans, continuous: U6'-5 |", 58'-l", U6'-$|" 6 rolled W33

beams with 32 degree skew. Spans are E-F-S-E, lh\'* high expansion and 13 3/8" high fixed steel bearings. Continuous non- composite deck slab-lj" open joint at ends with armor. Full

width pier on spread footing. Abutment and wingwalls on steel

piles.

3 spans, each simply supported: UO', 60', hO' , Rolled W27 and

W36 alloy steel beams, no skew. Spans are F-E, F-E, E-F. Discontinuous non-composite deck with open Joints and armored ends,

8" hiph expansion and 1 7/8" high fixed steel bearings. Full

width pier on spread footing. Abutments supported on steel

piles. Intermediate and end steel diaphragms on all spans (2 int.

in 60' span)

.

3 spans, each simply supported; 91'-8", 118', 118'. 6 welded

plate girders per span with 60 degree skew. Span are F-E, etc. Discontinuous composite deck slab-1" open joint, ends of slab

are armored. His" high fixed and expansion steel bearings. No

pier cap (girders bear directly on columns, columns transmit

load to continuous spread footing). Abutment with pedestals to

support girders (combined abutment and wingwalls on spread

footing.

V-8
TABLE 2

REPRESENTATIVE BRIDGE TYPES, LOCATIONS AND

SEISMIC ENVIRONMENTS FOR RETROFIT STUDY

Bridges Subjecced to ge^ipr.al

cartncuake m ern

ornia ern

j

ornia ornia

1

.ols

, j

a

ork Bridge

Actual

Bridge LocatnAlask North Calif South Calif Calif Illin Monta New

\ Description

Alaska * * * * * Steel girder with

Composite Deck

Northern

California * *

2 span concrete

box on column

Southern

California

* 4 span, curved

concrete box on circular piers

Mult 1—span

California

*

8 span, curved

concrete box on circular piers

Illinois * * 3 span steel

girder, noncomposite deck on

wall pier

Montana * * * Aspan skew prestress I girders

with composite

deck on col. bent.

New York * * * Prestressed I girder on Multiple

column bent

TOTAL 4 2 2 5

1 ' 2 19

V-9
TABLE 3

PROVISIONS FOR INCREASING

SEISMIC RESISTANCE

BRIDGE

1. Montana

(County Road

Seperation)

2. California

(Bahia Over

crossing)

3. California

(Ventura

Overcrossing)

U. Illinois

(Rt, 2h over

Massac Creek)

5 . . Alaska

(Bridge over

Lyon and Tin- can Creek)

New York

(1-^08 over Delaware and

Hudson Railroad)

CONTROL ITEMS

Columns and pier cap

Bearings

Abutment and footings on piles

Abutments and column

footing on piles

Hinge

Expansion Bearing

Columns

Abutments and footings on piles

Deck joints at abutments

Abutments on piles

Bearings

Deck joints and joint

at abutments

Bearings

Abutments on piles

Deck joints and joint

at abutments

Columns

Bearings

WAYS OF INCREASING SEISMIC RESISTANCE

Increase transverse shear resistance to improve torsional resistance of superstructure

Hold-down at interior pier caps

Greater longitudinal resistance

Improve anchorage of pile in cap, abutments

need greater longitudinal resistance.

Limit longitudinal mo 'ement and prevent

transverse and vertical movement.

Limit separation from abutment and prevent

transverse and vertical movement

Increase shear resistance and ductility

Improve anchorage of pile in footings j abutments need greater longitudinal resistance

Limit or cushion longitudinal movement and

possibly prevent transverse movement

Increase longitudinal resistance

Provide hold-down devices and check transverse

resistance

Limit or cushion longitudinal movement and

prevent transverse movement

Provide hold-down devices, and increase

transverse resistance

Greater longitudinal resistance

Limit or cushion longitudinal movement and

prevent transverse movement.

Increase transverse shear resistance to improve

torsional resistance of superstricture

Provide hold-down devices and check transverse

resistance or re-design bearings and possibly

increase area at abutments.

V-10
VtII
3^-0"

No. 11 LONG. BARS

7

CAST STRAIGHT

Section a-a

BAHAI OVERCROSSING

FIGURE 2

•V-12
F - 3 X lO

I

^ qc

SO -

Vtrov \r\ Cunv<z

U l+»m.a.i'<r.

Virairv Curv«

FIGURE 3

V-13

y-15
V-16
V-17
V-18
V-19
V-20
f OF BEARING

BEAM GRILLAGE

BEARING SEAT EXTENSION AND STOPPER

FIGURE 13

V-2I
i

-r-l i9 oUUUcm 0«) — ^\0 m3O 1V

k2

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Hoo O u o0

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cuB

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iJC

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<MO

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ra •H

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1^

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CM

V-22
V-23
REINFORCED CONCRETE COLUMN STRENGTHENING

FIGURE 16

V-2U
DYNAMIC TESTS OF STRUCTURES USING A LARGE SCALE SHAKE TABLE

by

Seiichi Inaba

National Research Center for Disaster Prevention

Since 1970, a large scale shake table located at the National Research

Center for Disaster Prevention (NRDP) has been widely used for the dynamic

testing of structures . This paper presents the results of some of those dynamic tests, using this jumbo earthquake simulator, and the results obtained

from some other shake tables operated by other research institutes in Japan.

Key Words: Dynamic tests; earthquake simulator; shake table; structural

engineering; tests.

V-25
LARGE SCALE SHAKE TABLE OF NRDP

The construction of a large scale shake table located in Japan was proposed in 1964

by many governmental and university researchers after the Niigata Earthquake. The planning, construction, and operation of the shaker was to be by the NRDP, but could be widely

used by other researchers throughout the country. The testing of the hydraulic actuators

and the design of the table began in 1967. In October, 1968, construction work began in

Tsukuba, Ibaraki Prefecture, and was completed in October 1970, with a total cost of

960,000,000 ($3,200,000).

The shake table is a 15 m square horizontal unit which weighs 160 tons. The table

can support a 500-ton structural test load, with a horizontal acceleration of 0.55 g. Vertical accelerations of 1.0 g can be imposed on a 200-ton test load, however, the horizontal and vertical motions cannot be applied simultaneously.

The simulator is housed in a steel framed building which is 24 m x 42.5 m in plan

and 16 m high. A prestressed concrete foundation supports the shake table, and forms an open box shape. The sides are 2 m thick with an outside dimension of 39 m x 25 m in plan

and 9 m deep. The foundation consists of a flat mat and rests on a sand layer where the

standard penetration test value is about 30. The total weight of the foundation mass is

12,000 tons. The ratio of the moving mass (table plus test structures) to the foundation

is 1:20. Four hydraulic actuators, which have a 90-ton dynamic rating, are provided to drive the table horizontally and four other 90-ton actuators are provided to drive the

table vertically. All of these eight actuators have a ± 30 mm stroke, and are electronically controlled. A moving mass up to a maximum of 660 tons is supported on four balance

cylinders during the horizontal drive. The table is constrained in one direction—horizontal or vertical—during motion in the other direction by sixteen guide rollers.

Vibration waves of various forms can also be imposed by a control system simulator,

which has a harmonic wave oscillator of sine, rectangular, and triangular wave configurations. A noise generator, using diodes, is equipped to impose random waves and programmed

transients of various forms can also be imposed. The programmed seismic input waves must

be in the range of the design spectra of the table, with the maximxim displacements less

than ± 30 mm for a single anplitude. Programmed displacement analog data are generally

used as the input signals. If a low frequency range of acceleration is used, a large displacement can be realized, however, the frequency has less influence on the magnitude of

the acceleration. The programmed data comes from motion accelerograph data, which is digitized and then the digital displacement data is obtained by double integration using a computer; the data are then filtered in the range of maximum displacements. This digitized

displacement data is then transferred electronically to analog data and a displacement

input curve is stored in the data recording system. This displacement curve is then imposed as input signals to the simulator.

The hydraulic power system is composed of a main and subsystem. The main hydraulic

system drives the four actuators, using 24,000 liters of oil at 210 kg/cm^ pressure from

thirteen pumping units. A sub-hydraulic system consisting of four pump units drives the

following units: twelve hydraulic bearings, four universal joints between the table and

actuators, and sixteen guide roller bearings by using 9,600 liters of oil at a pressure

of 75 or 140 kg/cm^. Seventeen pumping units are used in total with each pumping unit

being driven by a 150 KVA motor. The capacity of the electric power system to supply the

large scale shake table is 2,900 KVA and the entire electrical transformer station is

equipped to supply 66,000 Volts.

Figure 1 shows the maximum capacity characteristics of the shake table, in conjunction

with test results as obtained without use of a test structure. Figure 2 shows the frequency characteristics of the shake table during horizontal motion of a 100- ton structure. Figure 3 shows the frequency characteristics during vertical motion without a test load.

Figure 4 gives the El Centro earthquake displacement and acceleration of the shake table

when the El Centro earthquake input characteristics are imposed into the control system.

During the fiscal year 1973, the shake table had a total runnxng time of approximately 200

hours.

V-25
DYNAMIC TESTS CONDUCTED ON THE NRDP SHAKE TABLE

Since the construction of the large scale shake table at the National Research Center

for Disaster Prevention in 1970, a number of dynamic tests' of structures have been con- ducted. The structures that have been tested on the table are classified into the following categories: 1) full-scale structures, 2) large size model structures, 3) soil structures, and 4) ground.

The type of dynamic tests performed on these structures are categorized as follows ,-

1) Problems related to ground and soil structure, such as the river embaP-k- ment, sandy ground, and liquefaction of saturated sand layers.

2) Problems related to buildings and equipment, such as pre-fabricated

housing units and pre-fabricated room-appliance units.

3) Problems related to industrial plants, such as a graphite-pile nuclear

power reactor and a sphere tank.

4) Problems related to soil structure interaction, such as foundations of

a long span suspension bridge connecting Honshu and Shikoku, a submerged

t\innel for a Tokyo Bay highway crossing and undergound pipe lines.

All of these dynamic tests have been conducted at the Earthquake Engineering Laboratory of the National Research Center for Disaster Prevention. Most of these tests have

been performed in cooperation with other research institutes and some projects have been

supported by non-governmental organizations.

The dynamic tests that have been conducted since October, 1970 are listed in Table 1.

RESULTS OF DYNAMIC TESTS

Dynamic Characteristics of Sandy Soil

A sandy soil system was built by placing the sand in a steel box which was 12 m x 12 in plan and 1.5 m high. A sinusoidal vibration with a frequency range of 1 to 20 H^ was applied to the sand at maximum accelerations of 0.05, 0.1, 0,2, and 0.4 g in order to determine the resonance characteristics.

Figtire 5 shows the resonance curve at 0.05 g with a resonance frequency of 13 H^.

Figure 6 shows the curve for 0.4 g and a resonance of 10.5 H^. Acceleration ratios

at the resonance frequency at the center of the sand is ten and five. At the resonance

point, the acceleration ratio is greatest at the center of the soil. The greater the ac- celeration the flatter the acceleration ratio becomes throughout the sandy soil.

Dynamic Tests on a River Embankment

Tests on a river embankment constructed of sand, as detailed in Figure 7, were con- ducted in 1972 and 1973. Various ground water levels were placed in the embankment and

then sinusoidal wave vibration tests were conducted in order to find the dynamic characteristics and structural destruction criteria, such as slip failure of the embankment.

An example of a resonance curve for the embankment is shown in Figure 8. Figure 9 gives an exairple of the acceleration data that was observed. Figure 10 shows the distribution of excess pore water pressure due to the vibration. The results obtained from

these dynamic tests will be used to improve present design criteria and design procedures

on earthquake resistant embankments.

V-27
/

Liquefaction of Saturated Sandy Soil

Dynamic tests in saturated soil were conducted in 1972 and 1973 in order to determine

the characteristics of liquefaction. Figure 11 shows a resonance curve of 0.02 g sinusoidal acceleration, with a resonance frequency of about 10 H^. Figure 12 shows the change

in the excess pore water pressure during and after the vibration of 0.2 g.

These tests were performed relative to the design of a submerged tunnel and its re- claimed land.

LARGE SCALE SHAKE TABLE IN JAPAN

As shown in Table 2, several large scale shake tables for earthquake engineering have

been built in Japan. Large or full scale model tests of earth-filled dams, railway em- bankments, buildings, nuclear power reactors, and electronic equipment for tele-communication systems or computers have been conducted at these facilities.

- CONCLUSION

A brief description of earthquake engineering in Japan relative to present simulators

and dynamic tests have been presented.

However, these present large scale shake tables and other simulators cannot completely

satisfy the requirements of table size, displacement, maximum test weight, control system,

etc. needed by researchers concerned with earthquake engineering. Therefore, researchers

of the Science and Technology Agency and NRDP are planning and studying the technical , feasibility of large scale shake tables. One such simulator under consideration has a table 6.0 m x 6.0 m in size, with a horizontal and vertical motion driven simultaneously

up to ± 150 mm. Another simulator being studied has a table size of 30 m x 30 m with a maximum displacement of ± 200 mm and an acceleration of 0.5 g and a test loading of

1,000 t. The problems involved in designing such vibration test tables are difficult, to j solve, and it is, therefore, hoped that an exchange of information on earthquake simulators

and dynamic tests of structures between our two countries can meet the challenge.
TABLE 1

Test Struct\ire

Prefabricated house unit

Sand ground

Graphite-pile reactor

River embankment

Submerged tunnel

Suspension bridge, pier,

and foundation

Sphere tank

Sand ground

High-rise building

Sand ground

River embankment

Liquefaction of saturated

sand ground

Submerged tunnel

Suspension bridge, pier,

and foundation

Prefabricated room

appliances units

Sphere tank

Sand ground

Underground pipe

River embankment

Liquefaction of saturated

ground

Suspension bridge

foundation and ground

Oil tank

Underground pipe

Weight Year

40t November 1970

400t January 1971

30t July 1971

300t Septemberl971

5t November 1971

lot December 1971

30t February 1972

5t May 1972

It July 1972

400t August 1972

200t Septemberl972

200t October 1972

lot November 1972

lot December 1972

25t February 1973

30t February 1973

300t June 1973

300t June 1973

loot July 1973

200t Septemberl973

25t November 1973

3t February 1974

300t March 1974

V-29

Cooperating Institute of Sponsor*

Building Research Institute (BRI)

BRI

Public Works Research Institute

(PWRI)

PWRI

Honshi-Shikoku Connecting Bridge

Cooperation* (HSCBC)

BRI Research Institute for Pollution

and Resources (RIPR)

PWRI

PWRI

PWRI

HSCBC*

Japan Combustion Appliances In- spection Association

BRI, RIPR

PWRI

PWRI

HSCBC*

* Nippon Kokan
Test Structure

Nuclear reactor . .

Earthquake-free structure

system

Reclaimed land

Submerged tunnel

Suspension bridge

foundation and ground

Steel frame with brace

TABLE 1 (Continued)

Planned Tests

Weight Year Cooperating Institute or Sponsor*

lot June 1974 Hitachi* •

30t July 1974 Fujita Corporation*

200t Septemberl974 PWRI ;

3t November 1974 PWRI ' t

400t December 1974 HSCBC*

30t February 1975 BRI

V-30
TABLE 2

Drive Table Table Test Frequency Acceler- Displace- Actuator Foundation

Institute System Size Weight Weight Range ation ment Force Weight

National

Research

Center for Hydraulic 6mx6m 160t 500t(H) DC-50 H^^ 0.55g{H) ± 30 mm 360t

Disaster 200t(V) 1.0 g(V)

Prevention

12000t

Central

Institute

of Elec- Hydraulic 6inx6.5m 25t 120t O.I-2OH2 0.4 g trie Power

Industry

±50 mm 60t 4000t

Railway

Technical Hydraulic 10mx2mx 22t 78t 0.1-20H^ 0.4 g Research 3.2m

Institute Box

Mitsubishi

Heavy Hydraulic 6mx6m

Industry

21t loot O.l-SOH^ 1.0 g

±30 mm 40t

±50 mm loot

950t

Tokyo Hydraulic 10mx2mx

University Springs 2m 35t 135t 1.0-5 H, ±100 mm 20t 1200t

Obayashi

Corpora- Hydraulic 3inx4m tion

5t 20t DC-50 H^, 1.0 g ±100 mm 15t 225t

Kyoto Hydraulic 3mx3mx

University Electric 2.5mx

2.5m

12t O.I-3OH2, 0.5 g ±50 mm 6t

8t 1-200 0.5 g ±50 mm 4t

Telecommunication Hydraulic 3mx3in Research

Institute

5t lot 0.1-50H„ 1.0 g ±100mm(H) 15t

±120mm(V)

lOOOt

Port and

Harbor

Research

Institute

Electric 5.5mx2mx 8t 16t O.2-5OH2 0.5 g 1.5m(H)

±50 mm 12t 255t

V-31
' tiori-tDN tcil. V ijct' Ron

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3.5 1 lO 2030 50

p;g., 1 Capacity CharcictensUcs

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\ \

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i

\

\

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12 S \o ZD Freq.rti 12 5 10 20

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V-32
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v-34
A METHODOLOGY FOR EVALUATION OF EXISTING BUILDINGS

AGAINST EARTHQUAKES, HURRICANES AND TORNADOES

by

H.S. Lew

and

C.G. Culver

Center for Building Technology

Institute for Applied Technology

National Bureau of Standards

Washington, D.C. 20234

A methodology is presented for evaluation of existing buildings to determine

the risk to life safety from natural hazard conditions and to estimate the amount

of expected damage. Damage to structural building components resulting from the

extreme environments encountered in earthquakes, hurricanes, and tornadoes is considered. The methodology has the capability of treating a large class of structural

types including braced and unbraced steel frames , concrete frames with and without

shear walls, bearing wall structures, and long-span roof structures. Three independent but related sets of procedures for estimating damage for each of the natural

hazards are included in the methodology. The first set of procedures provides a means for qualitatively determining the damage level on the basis of data collected

in field surveys of the building. The second set utilizes a structural analysis

of the building to determine the damage level as a function of the behavior of critical elements. The third set is based on a computer analysis of the entire structure. All three sets of procedures are based on the current state-of-the-art. The

procedures are presented in a format which allows up-dating and refining.

Key Words: Buildings; damage; disaster; dynamic analysis; earthquakes; hurricanes;

natural hazards; structural engineering; tornadoes, wind.

V-35
INTRODUCTION

Background

Much of the loss of life and property in the United States from natural hazards such

as earthquakes, hurricanes, and tornadoes results from the inadequate performance of buildings in response to these extreme natural environments. Past observations have shown,

however, that buildings properly designed, detailed, and constructed can withstand these

environments. This involves realistic assessment of the forces produced by the environment

proper distribution of these forces to structural and nonstructural building components

and providing the necessary resistance to these components in the design process, and con- tinual inspection during construction to insure the design is correctly executed in the

field.

Improved building practices incorporating new knowledge relative to natural environments and the performance of buildings will serve to mitigate future losses. Continued

updating of building codes and standards, taking into account the latest research findings

and experiences gained from past performance of buildings, is one aspect of improved practice. Improved practices, however, apply only to future construction. They do not affect

existing buildings. The response of an existing building to an extreme natural environment will reflect the performance level inherent in the codes, standards, and construction

practices in existence at the time of design and construction. During the life of the

building, building practices continually improve reflecting the advancement of the state of knowledge. Thus, the margin of safety changes from that assumed at the time of design

as the state of knowledge and building practices advance. Deterioration during the ser- vice life of the building also affects the margin of safety. The need exists, therefore,

to continually evaluate buildings with respect to the potential hazard they pose when subjected to extreme natural environmental conditions. Following such an evaluation, appropriate rehabilitation or abatement procedures may be initiated to mitigate unacceptable

hazards

.

Following the 1971 San Fernando earthquake, several programs aimed at evaluating the

hazard posed by existing buildings in the event of an earthquake have been initiated. Most

of these programs are similar in nature, however, each uses a somewhat different method

of evaluation. Furthermore, since these methods involve evaluating buildings in accordance with the requirements reflected in current building codes, they do not provide an indication of the level of risk of explicit levels of building performance in terms of

life safety, protection of property, and maintenance of vital functions. They also do not

provide an estimate of the amount of building damage to be expected.

This report presents a methodology for survey and evaluation of existing buildings

to determine the risk of life safety under natural hazard conditions and estimate the

amount of expected damage.

Scope of the Methodology

The natural hazard loading conditions considered in this methodology are those en- countered in earthquakes, hurricanes, and tornadoes. While the source mechanism is different for each of these geophysical processes, they all impose dynamic loading. Existing historical seismic and meteorological data are used in the methodology for determining the magnitude of recurrence interval for these hazard loadings. For earthquakes,

recorded seismic data were used for determing expected ground motion. For wind loading,

annual extreme wind speeds for specific mean recurrence intervals are used.

The types of structures considered include braced and unbraced steel frames, con- crete frames, shear wall structures, combination frame and shear wall structures, bearing

wall structures, and long-span roof structures. Although no specific limitations are imposed on application of the methodology, it is intended for buildings with substantial

occupancy, i.e., fifcy or more people. One- and two-story residential buildings, therefore, are not considered.

V-36
Three sets of evaluation procedures were included in this methodology, each set representing a different level of analytical sophistication. Hereafter, these three sets are referred to as the Field Evaluation Method, the Approximate Analytical Method, and the

Detailed Analytical Method.

In the Field Evaluation Method buildings are evaluated on a qualitative basis in terms of structural characteristics, structural configuration, and the degree of deterioration

of the building. Information on these are obtained from a field survey. This method provides a rapid, inexpensive means for identifying clearly hazardous structures or potentially hazardous ones requiring a more detailed analysis to estimate damage.

In the Approximate Analytical Method buildings are evaluated in terms of the behavior

of critical structural members. The procedure requires an analysis of the structure to identify critical members and determine the stress level induced in these members by the

extreme environments. A set of building plans, specifications, and construction drawings

are needed to obtain the necessary data required to perform the analysis.

In the Detailed Analytical Method the damage level is evaluated on the basis of the

energy capacity of the structure. This procedure requires the use of a digital computer

program for the -evaluation. As in the Approximate Analytical Method, the building data

needed for input to the coitjiuter program would be obtained from a set of building plans,

specifications, and construction drawings.

DAMAGE EVALUATION METHODOLOGY

Field Evaluation Method

The Field Evaluation Method can be used where evaluation results do not need to be

refined. This method is particularly applicable if building plans are not available.

Earthquake

For earthquake, a qualitative evaluation of building is made based on a combined

effect of structural type, vertical resisting elements, and horizontal resisting elements. Based primarily on the past performance of various types of buildings, relative ratings for these three factors are developed. These are used to determine a basic structural rating which is the basis for determining the building capacity.

This is shown schematically in Figure 1.

Ratings for various structural types are to account for past performance and

the degree of \incertainty that the building would perform in a manner anticipated for

the type. These ratings are based on damage experience and judgment. For instance,

moment resisting steel frames would be rated good, whereas unreinforced masonry shear

walls would be rated poor.

The vertical resisting elements include shear walls, shear cores, vertical bracings, and colimns. In assessing the ratings for the vertical resisting systems,

"symmetry", "quantity", and "present condition" of these individual building elements

are considered.

The symmetry describes the eccentricity between the center of mass of the structure and the center of stiffness of the vertical resisting elements. Thus, in a building where the only shear walls are the exterior walls with only one opening in

the center of the opposite walls, and the building plan is rectangular, the building

would be classified as symmetrical. On the other hand, a two-story rectangular

structure with nearly solid side and rear walls, but with front wall almost entirely

glazed with only four small piers or columns, would be classified as very unsymmetrical.

V-37
The quantity refers to the number of vertical resisting elements. If there are many long shear walls, it would be rated good. In the case of moment-resisting frame

structures of steel or concrete, both the strength and number of columns are important. ) For closer spacing of columns, usually about 6 m to 10 m bays, a good rating is given.

The present condition describes wall cracks and other damage. Other damage may

be damage caused by deterioration from lactic or tannic acid (frequently found in

dairy or slaughterhouse facilities) or from severe popping of concrete caused by the use of reactive aggregates. The degree of damage in such cases can only be estimated

by visual observation.

For vertical resisting elements, "quantity" and "symmetry" are combined as one

factor. This factor is then combined with "present condition" to obtain the rating

for the vertical resisting elements, see Figure 2.

The horizontal resisting elements include diaphragms, premeter beams, and horizontal bracings. The floor or roof systems act as horizontal diaphragms to distribute

horizontal forces to the vertical resisting elements, such as shear walls, momentresisting frames, or braced frames. This diaphragm action is similar to that of a horizontal plate girder spanning between the vertical resisting elements and may be

continuous over several supports. In this analogy, the floor itself may be compared

to plate girder web, and the marginal beams or walls (chords) compared to plate girder

flanges. The floor or roof system, acting as a girder web, is primarily a shear re- sisting element. The marginal beams or girder flanges (chords) in diaphragm action

are primarily subjected to axial loads of tension or compression.

Shears are transferred by the anchorage between floor or roof and the shear walls

or frame members. The determination of the adequacy of anchorage or connections involves considerable judgment unless computations are made. For example, where the

floor or roof systems are of cast-in-place concrete and are placed integrally with

portions of the shear walls or frames, generally a good anchorage with shear transfer

capacities can be assumed dependent on the concrete strength, concrete slab thickness

and amount and anchorage of reinforcing steel. With metal deck systems, the diaphragm

values are dependent on the deck configuration, attachment between units, gage, and

attachments to supports.

The capacity of the horizontal resisting system is dependent on either the rigidity of the diaphragm, the anchorage capacity of the diaphragm or horizontal bracings to the vertical resisting system, or the effectiveness of chord members. Thus,

the lowest rating of these three factors is considered as the rating for the horizontal resisting system. The rating schemes used to rate the capacity of the horizontal resisting systems are illustrated in Figure 2.

The Basic Structural Rating which describes the capacity of a building to resist

the earthquake force is obtained by combining the rating for structural type and the

rating for either the vertical resisting system or the horizontal resisting system,

whichever provides a lower rating. The rating scheme adopted for this study is illustrated in Figure 3.

Hurricane

Damage to a building is dependent on two basic considerations. One is the effective wind force, usually in terms of pressure, positive or negative. On the other

is the resistive capacity of the structure to lateral forces and also to uplift

forces on the roof created by the structure's shape.

The same rating scheme used for structural systems in the case of earthquake

can be used for wind in the evaluation of a building's ability to resist lateral

forces. Because uplift forces are acting simultaneously with the lateral forces,

additional factors such as roof anchorage, anchorage to foundation, and internal

pressure should also be considered. .

.

.1

V~38
The rating of the building is determined by taking the lowest rating of the

"Foundation Anchorage" factor, "Roof Anchorage" factor, or the "Basic Structural

Rating" as defined in the case of earthquake since these factors affect the building's

capacity to resist wind independently.

Tornado

Damage from tornadoes has been most severe to small, light buildings; although

tall, flexible, and heavier buildings have sustained some severe structural damage.

The most extensive damage to buildings has been to roofs and exterior claddings, including glass. This includes damage from windblown debris. Because the total effects

of tornado on a building is not clearly understood at the present time, only a broad

categorical rating of buildings, depending upon their types, is possible. In this

study, a poor risk rating is given to small and light buildings. A medium risk rating to small, heavy buildings, and to large, multi-story buildings that could be

rated high in wind and earthquake resistance. A good rating can only be given to heavy vault-like buildings known to have been designed for tornadoes.

Approximate Analytical Method

The Approximate Analytical Method provides a simplified analytical procedure for evaluation of the building capability to resist natural hazards by determining stress ratios

of critical elements of structural elements. These stress ratios are the ratios of the

stresses produced by the loading to limiting stresses of the critical building elements.

For the purpose of this evaluation, elastic analyses of building response will be

cortpared with material design capacities. Design stresses will be those designated by

material specifications. In the evaluation, buildings are being analyzed, but not designed.

Stresses in structural elements will be checked for the combined effects of lateral and

vertical loads. Where lateral loads are included, the combined stresses may exeed code

working stresses by one-third, except where not permitted in the specifications, with the

provision that the stresses resulting from design vertical loads alone will not exceed

code design stresses. This method, in general, does not include the use of a dynamic analysis except in special cases.

Earthquake

Any iitportant earthquake resisting element having the highest unit stress as related to allowable design stresses is a critical element to be considered in the evaluation of the structural system. The term "important" element means an element which,

if it failed, v;ould seriously reduce the capacity of the structure as a whole to re- sist lateral forces. Some members would not be critical when deformed beyond their

yield level deformations. In other members, yielding may cause an important redistribution of loads. With a multiplicity of well-distributed similar elements, the re- distribution of loads would add only a small percentage of stress to adjoining or parallel elements.

In most buildings, the critical elements will be the vertical resisting elements

(shear walls or moment resistant frames) and the horizontal resisting elements (diaphragms) . This is for earthquake forces acting in the plane of the elements. Earthquake forces normal to a wall are a function of the weight of the wall itself. Where

a wall has a long span between floor diaphragms or vertical frame elements, it might

be a critical element if its failiire would produce collapse of the building as a whole from vertical loads or in-plane lateral forces.

The highest ratio or stress resulting from the required seismic forces (f^) to the allowable material design stress (f^) (including 1/3 increase where permitted

but deducting capacity required by gravity loads) on any critical element is termed

the critical stress ratio fg/fa' '^^^ critical stress ratio is the indicator for

evaluating the seismic resisting capability of the structure. Load factors and ulti- mate capacities are used for concrete design and for plastic design of structural steel.

V-39
Hurricane

Both internal and external pressures must be considered on portions of buildings

such as roofs and walls. Corners of walls are exposed to high negative pressures.

These corners should be checked.

In determining the response of the building as a whole, the type of diaphragm

system must be considered. A stiff diaphragm will distribute horizontal forces to vertical resisting elements in proportion to their relative rigidities. A flexible

diaphragm will distribute forces to vertical elements more nearly proportioned to the

tributary-exposed wind surfaces.

The usual critical elements for wind resistance are, as in earthquake resistance,

the vertical resisting elements, (shear walls, braced bays, or moment resistant frames)

and the horizontal resisting elements (diaphragms). There is a major difference, however, in that wind forces are applied to exposed surfaces while earthquake forces

originate at centers of mass and are proportional to mass. Thus, a lightweight ex- terior wall might have a relatively small earthquake force normal to the wall but would

be exposed to wind forces which are independent of the weight of the wall.

The calculations for the adequacy of the building to resist wind forces can be

made using any of the standard analytical procedures. The lateral loading to each

story level is determined with positive pressures on the windward side and negative

pressures on the leeward side. The path by which these forces are transmitted to the vertical resisting elements is determined and the adequacy of diaphragm or horizontal bracing system to transmit these forces should be evaluated. Lateral forces

are applied to the vertical resisting elements at each level and the stresses should

be analyzed. The overturning stresses should be checked, including uplift on foundations .

The highest ratio of stress resulting from the wind forces applicable to the

site (f^) to the allowable material design stress (f^) (including 1/3 increase where

permitted but deducting capacity required by gravity loads) on any critical element

is termed the critical stress ratio f^/f^.

Tornado

For the purpose of evaluation, it will be assumed that a free field wind velocity

of 200 miles per hour and a pressure drop of 1.2 psi. Using the formula P = .0025 V

to convert velocity to pressure gives P = 100 psf. The 1.2 psi pressure drop is, con- verting units, equal to a suction or uplift of 172 psf. If it is assumed that tor- nadoes are similar to other high winds, such as hurricanes, with respect to the re- lationship of pressures on windward and leeward sides, one finds these coefficients

of the velocity pressures to be 0.8 and either 0.5 or 0.6, respectively; depending

on the height-width ratio of the building. Thus, the total lateral force on a building subjected to tornadoes for a moderate degree of protection will be 100 x (0.8 + 0.5 or 0.6) = 130 or 140 psf.

There are probably very few buildings that will be undamaged if in the direct

path of a strong tornado. The least likely to be severely damaged will be heavy

reinforced concrete or reinforced masonry vault-like structures one or two stories

in height, with relatively heavy and solid walls. Taller buildings designed to resist hurricanes may have limited damage, but probably will not collapse unless

they are of unusual configuration or have large roof overhangs or open sides. Light

buildings of wood frame or steel frame and metal sidings have been known to have

been torn from their foundations and blown considerable distances. It is possible,

however, to provide such light buildings with some resistance to winds by proper

anchorage to foundations to the ground to resist uplift.

If a building has been evaluated for wind and given a poor rating, it will be

considered to be inadequate to resist tornadoes. The ability of roof systems, designed

V-40
only for gravity loads, to resist reversals from uplift need to be evaluated. The

wall capacities must be checked for direct horizontal positive and negative force.

The floor and roof systems must be checked for adequacy as diaphragms to transmit

lateral forces to the vertical resisting elements. Anchorage of walls to footings

should be checked for capacity to resist sliding combined with vertical uplift and

overturning. The footings themselves should be checked for sliding resistance, uplift, and overturning forces. In checking these the purpose is to determine the ratio

of stresses resulting from the imposed tornado loads to the capacity of the structure.

In each of the various resisting systems, such as roof, floors, shear walls, or moment-resisting frames, there may be one or more critical elements which will fail

before the other elements. Care must be taken not to derate a building because of

one non-important element. Where the failure of such an element would not cause failure of a system, but would only cause minor rvdistribution of loads, such a member

would not be a critical element. The procedure for determining the critical stress ratio is the same that was used in the case for hurricane.

Detailed Analytical Method

This method is based on a modular computer program with each module dealing with a particular aspect of the damageability prediction problem which includes:

1. Environmental Loads

2. Structural Characterization

3 . Response Computation

4. Estimation of Potential Damage

For seismic loads, historical and recorded data are used along with the program describing the seismic and wind activity for the entire continental United States, Alaska,

and Hawaii. Historically based tornado and hurricane activity for the continental United

States is included. With these data, the program will computer, in a probabilistic sense, the specific environment of any given building site in the country. Alternatively, the user may choose to input any of these loads directly.

Structural models of varying complexity can be generated depending on the availability of structural data and the level of effort selected for a particular task. Damage predictions are made on the basis of the building's response to the appropriate loading con- ditions. Damageability data characterizing the capacity of the building to resist fail- ure must be input by the user. Algorithms for computing damage are based on the assumption that percent damage varies continuously with key response parameters. These key parameters have been selected and are incorporated in the program. The forms of damage

distribution curves as functions of response are also built into the program. The user

may choose values of the parameters of these functions in accordance with prepared guidelines and exercise his judgment according to the application at hand. A schematic presentation of the computer program is given in Figure 4.

Natural Hazard Loading

Earthquake Loads

Site loads for an earthquake are defined by a ground site response spectrum.

This response spectrum reflects an amplification of the hardrock spectrum for the

site, which is frequency dependent. The hardroc't spectrum is generated in either of

two ways. If the risk option is selected, the usee must input seismicity data from

the Seismic Map provided with the computer program. By further specifying either a return period or building life and probability of non-occurrence, a risk earthquake

defined by its Richter magnitude is computed. If the risk option is by-passed, then

the user must input a Richter magnitude and a hypocentral distance. In either case.

V-41
the computer program computes the maximum hardrock acceleration, velocity, and displacement at the site.

Hurricane Loads

If a wind velocity is generated for the specified site, then a statistical re- gression analysis is used to compute a mean wind velocity given either a return

period or building life and probability of non-occurrence. A standard deviation is

also computed and may be used to establish some degree of confidence in the wind

velocity selected for analysis. If the user desires, he may alternately choose to input a wind velocity for analysis. As in the case of earthquake, local site conditions modify the wind velocity which is determined on the basis of historical data.

For exartple, free-field wind velocities in large cities are attenuated by the presence of buildings, whereas, on an open prairie, the attenuation is comparatively

small as one approaches the ground. Thus, in figuring the wind load on a building,

surface conditions are accounted for by specifying a site condition parameter.

Tornado Loads

The probability of being hit by a tornado can be computed from historical data.

However, the user must specify a tornado wind velocity. In this case, the static

response of the building will be computed and damage assessments made.

Structural Characterization

Three general levels are offered for structural dynamic modeling:

1. A Detailed Model

2. A Story-Stiffness Model

3. An Empirical Model

It is noted that all of the modeling options lead to the same basic dynamic characterization: building natural frequencies and modal displacements.

The three modeling options are illustrated in Figures 5, 6, and 7. If the user se- lects "Detailed Model" option to generate a detailed stiffness matrix for the structure,

then the stiffness matrix is constructed frame by frame, story by story, from top to bottom. Each frame is parallel to the direction of motion in a vertical plane. Stiffness

and mass contributions from each frame are superposed in formulating the two-dimensional

model of a building. If the user desires "Story-Stiffness Model" option, he must input

a story stiffness and story height data for each building story. A total story-stiffness

is the sum of the stiffness produced by all columns, partitions, walls, and other lateral

force resisting components at that story. If an empirical model is to be generated, then

the user must input an estimate of the building's fundamental period and story heights. A straight-line mode is assumed to compute deflections.

Response Computation

Response computations are made for earthquake loads, for tornado and hurricane loads,

and for uplift due to wind. Ponding loads are also computed in the long-span roof subroutine.

The response of a building to earthquake ground motion is evaluated by determining

the peak modal response in each of the modes (a maximum of six is considered) , and com- bining their contributions to the total response. A damping, a ductility factor, a modal

combination scheme, and a value for each story's drift-to-yield are used to compute an effective ductility for the building. An iterative procedure must be used to establish

consistent values of damping and response as damping is defined as a function of ductility. \ While this operation seems to be stable and typically converges in a few cycles (three or

four) , an upper limit (e.g. , six) on the number of interactions is put in and used to transf

V-42
control out of the iterative loop in case the computations do not converge to within five

percent accuracy. This may only occur when elasto-plastic response is considered.

Inter-stoiry drift, pseudo-velocity, and absolute acceleration are computed for each

story. In addition, story forces and shears are computed.

The response of a building to wind is treated as a static problem. The force acting

along the building is confuted by multiplying the pressure at each story by the tributary

story area. For the detailed and story-stiffness models, the inverted stiffness matrix

is used to compute deflection of the building. If the empirical modeling option is chosen

the building response is calculated considering the building to be a uniform cantilever

beam using a simplified analysis.

Evaluation of Damage

Potential damage to a building which may result from exposure to the environmental

loads computed for the building site is evaluated in a damage subroutine. Damage is ex- pressed in percent of total damage on a story-by-story basis. Damage is computed independently for earthquake, hurricane, and tornado. It is segregated into three categories:

structural, non-structural, and glass. In the case of structural damage, the damage is

further subdivided into damage to frame, walls, and diaphragms.

The key response parameters used to predict damage in each case are outlined below.

Earthquake

a. Structural: Interstory Drift

b. Non-structural: Floor Velocity or Acceleration

c. Glass: Interstory Drift

Wind, Tornado, or Hurricane

a. Structural: Interstory Drift

b. Partitions: Interstory Drift

c. Glass: Direct Pressure

SUMMARY

This report presents a methodology for evaluating the potential damage of buildings

due to earthquake and extreme wind including tornado and hurricane. Three independent,

but related, sets of procedures are developed. These ranged from a qualitative procedure

based on field surveys data to a detailed analytical procedure involving a digital com- puter program. Historical seismic and meteorological data are used as the basis for estab

lishing environmental loads. Damage estimates are based on empirical correlations between

structural response and observed damage coupled with engineering judgment.

V-43
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BMiC STRUCTU^M ^ATi^G (ESR)

- _ GR + 2 (LARGER OF SRI OR SR2)

BSR — ^

3

RATING OF TYPES OF STRUCTURE (GR)

Figure 3. Scheme for Determining Basic Structural

Rating

V-46
Q START ^

INPUT

© EnVironnenral Data

for £arthquake.

Wind and Tornado as Required for Selected

Hazards

• Structural Modeling

Data According to

Specified Modeling

Ootion

SUBROUTINE:

L0ADS

SUBROUTINE:

STRUCT

• Dcimping Type

• Ductility to Failure

o Interstory DriftTo-Yield

lAUYSI

Building Data

I^eeded to

Compute Pressure Distribution

(Dimensions, Floor

Weights

)

• Roof Configuration

Data and Analysis

Options

e Damageability Data

(Quality Factors)

YES

SUBROUTINE:

STATIC

SUBROUTINE:

DAMAGE

• Compute Site Earthquake Loads

• Define

Structural

Model

^^^^earthquakT^

^^^analysis.;.^

NO

|YES

SUBROUTINE:

DYNAMC

NO

• Evaluate Dynamic

Response to Sii^e

Ground Motion

• Evaluate Equivalent

Static Response to

Wind and Tornado

• Evaluate Equivalent

Static Response of

Long-Span Roof to

Uplift and Ponding

Evaluate Structural and

Kon Structural Damage to

Building Due to Earthquake, Wind, and/or

Tornado

Figure Schematic of Computer Program

V-47
T

MODE DEFLECTION

SHAPES AND SHEARS

Figure 6. Story Stiffness Model

Figure 7. Empirical Model

V-48
UNBRrtCED CR

MODE DEFLECTION

SHAPES AND SHEARS

v'-c DTTTT riTMr:

i

I

V-49
EXPERIMENTAL RESEARCH ON THE ASEISMIC CHARACTERISTIC OF SPHERICAL STEEL TANK FOR LIQUID PETROLEIUM GAS

by

Kiyoshi Nakano

Head of Structural Division • - Building Research Institute

Makoto Watabe

Chief of Earthquake Engineering

International Institute Of Seismology And Earthquake Engineering

The results of static and dynamic tests on a spherical steel tank are given in

detail. A theoretical technique to explain the "sloshing" effect is presented. A proposal for a safer design procedure to replace the present aseismic design practice is presented.

Key Words: Dynamic analysis; earthquake; seismic design; seismic response;

spherical tanks; structural design.

V-50
PREFACE

It has recently been pointed out that large tanks containing flammable materials in

and around densely populated cities present a dangerous situation for the safety of citizens during severe earthquakes. All people in Tokyo, for instance, would be burnt if all

petroleum and liquid petroleum gas in and around Tokyo should catch fire due to an earthquake .

Unfortunately, aseismic design of these tanks has not been adequately developed. Fortunately, research on the aseismic properties of spherical tanks used for liquid butane gas

has been conducted by various committees. Recent results on some of this research will

be presented herein.

SCOPE OF EXPERIMENTS

The test specimen that was selected consisted of a one-third scale model of a large

spherical tank. The dimensions of the test specimen are illustrated in Figure 1. The

weight of the test specimen was ten tons and can contain 150 tons of water (88 tons of

butane gas) . The weight of the test specimen was limited by the capacity of the shake

table, details of which are shown in Figure 2.

Initially, the test specimen was subjected to static lateral loading by slightly tilting the specimen, as is shown in Figure 3. This type of loading was expected to give data

on the stress distributions in various parts of the specimen and the basic lateral stiffness of the specimen. The location of the strain gages and other measuring devices are described in Figure 4.

The test specimen was then subjected to the dynamic loading. Initially, hand-powered

excitations were induced in order to obtain the approximate fundamental period and damping

characteristics of the test specimen. Mechanical input forces, using a sinusoidal wave within the range of 0 to 16 H^, were then applied. Also, forces of actual and simulated

earthquake motions, such as the El Centro Earthquake, Hidakasankei Earthquake, randomly

modified El Centrol Earthquake, were also applied to the test specimen. The quantity of

water, which represents the amount of liquid butane gas, was varied by the following percentages: 0%, 20%, 30%, 50%, 60%, 70%, 80%, 90%, and full. Through this variation in the

water content, the complete aspect of the sloshing phenomenon was made clear.

The final stage of the experiment was to induce a failure mechanism by inputing a wave of the modified El Centro Earthquake with 400 gallons of peak acceleration with 85%

of the full water capacity.

In order to measure accelerations, ten accelerometers were utilized. To measure re- lative displacements at the base, transducers were attached between the test specimen and

a rigid safety guard. This guard was installed around the test specimen to protect the

transducers in case of premature failure. Prior to the dynamic and static tests, tensile

forces were induced into the diagonal braces by means of toggle bolts. The tensile strains

that were in^osed were of the order of 250 p in. /in., the value of which is typical of the

strains induced in tanks used in the field.

DISCUSSION OF THE EXPERIMENTAL RESULTS

Static Tilting Test

Figure 5 describes the experimental results from which the lateral stiffness can be

computed. Also shown is the calculated load-deformation response. The calculated loaddeformation curve and thus, the lateral stiffness was based upon the following assiamption.

i) The spherical shell is assumed to be rigid (at Point a shown in Figure 6)

.

V-51
ii) The Point "a" in Figure 6 is assumed to be the center of the connnection

between the tubular column and the spherical shell.

iii) Deformation of the members is determined by using a conventional energy

method, excluding shear deformations. Surrounding the connection to the

tubular column, significant stress concentrations were observed at a dis- tance within three times the thickness of the spherical shell. Using the

above area, about four times the yielding strain should be expected when

30% of the total weight is applied as a lateral static loading. Outside

of this connection area, the stress concentration effect can be disregarded.

Elastic Dynamic Test

In order to determine the dynamic properties of the test specimen, the following fac- tors must be determined:

i) Characteristic Period

ii) Fraction of Critical Damping

iii) Effective Mass of the Dynamic System ? ...

iv) Response to the Earthquake Loading

a) Periods and Damping

When the test structure contains a liquid, the so-called "sloshing phenomenon"

(motion of the free surface of the liquid) becomes a common problem and thus the

simplified model illustrated in Figure 8 is useful in understanding this complicated

behavior. W^, given in Figure 8, is called the "fixed water"; is called the

"effective mass for sloshing"; k^ is the lateral stiffness of the structure; and k

is the virtual stiffness, all of these parameters characterize the shoshing phenomenon. In this experiment, Wg and are of the same order, however, k^ is much smaller

than kj. Therefore, the two resonance modes appear to be independent and thus the

"structural period, T^" and "sloshing period, T^" modes will be defined as follows:

Tf = 2Tr /Wf + W^/g • k^

Tg = 2Tr /Wg/g • k^

Through a free vibration test, excited by impulse shock and through resonance curves obtained by harmonic excitation from 0 to 16 H^, periods and fraction of critical

damping, as indicated in Table 1, were obtained. The difference in the fraction of

damping as obtained from the free vibration and the one obtained from the resonance

curve may be due to the powerless performance of the shake table. The damping of

the sloshing is extremely small, and almost impossible to state numerically (0.043%).

b) Stiffness and Effective Mass for Sloshing

The lateral stiffness of the structures, as obtained from the static tilting

test and by the theoretical calculation, coincides exactly.

The values of and were obtained from the experimental results. Consider

the weight of the structure without water to be equal to W^, the lateral stiffness

kf, the structural period without water T^, and with water T^, and the dynamic factor

of the structure y^. If all of these parameters are known from dynamic tests, then

+ = (Tf/2TT)2 g

V-52
Therefore, the fraction of fixed water f (n ) to total water is;

f(n) = Wf/lV„ = {WQ(Tf/T^)2 - i}/w^

If we define the dynamic factor of sloshing, without damping, as p^, and the

base displacement as 5 _ and 6 as 6 = T^/Tg, then;

1 - 52

The shear force or lateral spring F in the structure can be expressed as;

^s = ^s • '^x • <5o Using this value, the dynamic factor of the structure ^ can be expressed as;

T ^S^-S +

1 - ''f

where = T^/Tf

Noting that = A-n^/T^

^'^s^'^ substituting this relation into the above

equation , gives

;

1 '^f^s' 2 "s = i^f J —T ^ - ^ ^ 1 - 3f

Using the experimental results and noting that the sum of and Wg equal the total

weight W , that and are constant and are independent of the amplitude or frequency o? the input acceleration, the parameter f(ri) is obtained as shown in Figure 9.

c) Response to Earthquake Excitations

The linear response of the tank to the modified El Centro Earthquake is tabulated in Table 2. The meaning of "modified" is in relationship to a digital filtering process that is required in order to assume that the maximum displacement of the

earthq'jake is limited within the maximum stroke of the shake table. In this experiment, the maximum displacement was limited to 2.5 cm for a peak acceleration of 400

gallons. When the capacity of the shake table is not sufficient to excite a heavy

test specimen, the spectrum ratio between input and output contains a dip along the

frequency curve of the structure. Unfortunately, this condition occurred during the

testing of the tank structure. Figure 10 shows the response spectr\im of the output

wave as recorded at the shake table. The deep dip can be seen in Figure 10 during

the fundamental structural period. Modification of the earthquake excitation will

not improve the response and little can be observed about the sloshing phenomenon.

Dynamic Test of Failure Mechanism

The failure mechanism test was instituted by imposing input of the forms related to the modified El Centro Earthquake (N-S comp.) wave with water content ratios equal to 75%

and 85%. Assuming the water content ratio equal to X , the equivalent fixed "liquid"

ratio y for X percent content of liquid butane gas can then be estimated as given in the

following procedtore;

Assume a density of liquid butane gas equal to p = 0.585 and a fixed water

water ratio f(n) for ; following ejcpression,

pXy = X^y^

water ratio f(n) for a water content ratio X , then y^ is related by the

o o

V-53
From Figure 9, for instance, if is assumed to be 75%, then is 0.53 and thus Xy equal

0,68. Once again from Figure 9, the value of A and y are 0.915 and 0.74, respectively.

By this procedure, it may be understood that 75% of the water content is equivalent to 91.5% of the butane liquid content, in view of the "fixed liquid" weight.

In Figure 11, the relation between the response acceleration and response displacement, as the results of the failure mechanism dynamic test, is shown. Due to the inelasti

property of the lateral stiffness beyond the application of 260 gallons, the acceleration

and displacement is not linear, but has a multi-linear response which can be predicted

theoretically as is illustrated in Figure 11.

Fracture was observed at the connection in the area of the diagonal bracing bars.

Almost simultaneously, failure was also observed in the tubular column around the area of

the connection to the lateral bracing. No visual failure was seen in the shell sphere at the connection to the tubular column. In Figures 12, 13, and 14, the time history respons

at various locations are shown. These responses are the displacements and accelerations

at the base and at the center of the structure, the strains in the diagonal bracings, and

the strains in the shell adjacent to the tubular column connections.

CONCLUSION

Through a series of experiments the following observations have been made;

1) The sloshing behavior can be treated quite exactly by use of a simplified

mode 1

.

2) The frequency periods for the structure and sloshing are;

Tg (Sloshing period) = /D/g s(ri) '

where • D = Diameter of spherical tank

g = Gravity acceleration

n = Water content ratio

s(n) = Function of n as shown in Figure 15

In this experiment T^ is expressed

Tg = 0.134 s(n)

as shown in Figure 15. T^ can be predicted as

T^ = 2tt /Wj + W^/g • k^

where

Wf = W„ • f (n)

Wq = Weight of steel tank itself N

f(n) = Function of n as shown in Figure 15

k^ = Structural lateral stiffness as given previously

3) Fraction of Critical Damping

For structural dartping, the fraction of critical damping is 0.6%, while for

sloshing, the fraction of critical damping is 0.043%.

V-54
4) The spherical shell can be assumed to be a rigid body, as indicated by the

calculated stiffness and stresses in the diagonal braces.

5) Excitation in the Transverse Direction . .

About 20% transverse excitation was observed during the last stage of the

experiment.

6) Aseismic Coefficient

The yielding shearing force was equivalent to the base shear coefficient

(aseismic coefficient) of 1.00. Therefore, if a response dynamic factor is

assumed to equal 3.0-4.0, then this type of tank will not be safe when subjected to a peak ac/:eleration equal to or greater than 0.3 g.

7) The sloshing phenomenon was observed during the last stage of excitation.

COMMENT ON THE SAFETY OF THE STEEL SPHERICAL TANKS

In view of the explosive potential of these types of tanks, fracture cannot be permitted during severe earthquake motions. Because fracture of the diagonal bracing does

not lead to total collapse of the structure, the load capacity based on that load which

induces a yielding lateral force can be used as a design criterion.

In the case of smaller tanks, as was this test specimen, the response dynamic factor

earthquake motions will be about 3.0 to 4.0. If the size of the tank is increased, then

the fundamental period will be increased, so the response dynamic factor will be reduced

to a value of 2.0-1.0. In this sense, the base shear coefficient should be determined

according to the fundmental period. Considering that the reseirved strength of these

types of tanks is about three times the elastic design strength, the following base shear

coefficient (C ) is appropriate.

B Cg = 0.45 - 0.6 Tf 4 1.0 sec.

C^ = 0-45 - 0.6 + (Effect of Sloshing) T^ > 1.0 sec. ^ Tf

^

ACKNOWLEDGMENT

This research was supported by the National Science Fund of the Japanese Government

supplied by the Science Technology Agency. This research program was carried out by the

following committee members;

M. Watabe^, H. Akiyama^, Y. Kitagawa^, H. Murata^, H. Suzuki"^, Y. Fukazawa"^

The help extended to us by the staff of the large-scale shake table in Tsukuba is gratefully acknowledge.

International Institute of Seismology and Earthquake Engineering

'university of Tokyo

V-55
frequency (Hz)

Fig. -2 THE CAPACITY OF THE SHAKING TABLE

v-56
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V-57
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V-59
,DISP. AT BASE

^l^i^i^'il^Vv DISP. § ACC. RESPONSE

v-60
sloshing Ts Ts=0J34yjDS(n)

f(n)=m/ww

H4v : Total weight of water

Wf : Fixed water

Free water ratio : h

Ws=\!-f(T\}fA/w

0.5 10 ^ Water content ratio

Fig- 15 Function SCnjand f(T\)

V-61
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V-52
RESEARCH ON MINIMIZING EARTHQUAKE STRUCTURAL DAMAGE

TO SINGLE-FAMILY DWELLINGS

by

William J. Werner, P.E.

Office of Policy Development and Research

Department of Housing and Urban Development

This paper discusses proposed research work to be carried out by the Applied

Technology Council of the Structural Engineers Association of California under

the sponsorship of the Department of Housing and Urban Development. The objective

of the project is to develop a manual of recommended construction practice for

earthquake resistive dwellings, for use primarily by builders, building officials,

field inspectors, plan checkers, and designers.

The manual is intended to explain the structural behavior of single-family

dwellings and townhouses subjected to forces produced by earthquake shocks, illustrate the HUD Minimum Property Standards, building code earthquake requirements

and sound practical construction methods and details for the reduction of single

family dwelling damage. The paper discusses the need for this research, the var- ious tasks the contractor will perform, and the final products expected to be

achieved by the research program.

Key Words: Building codes; construction practices; damage; earthquake; houses;

residential dwelling.

V-63
Historically, there has been a great deal of interest on the part of the engineering

profession in investigating the damage to buildings and other structures as a result of

seismic activity. In the United States, as in other countries, disaster teams composed

of qualified experts in the field are among the first people at the site follov/ing an earthquake or similar disaster. This activity has even been expanded into international

cooperation whereby disasters which occur any place in the world are thoroughly investigated.

The results of these investigations can be found in detailed reports which identify

the many types of failures which have occurred and, in greater or lesser degree, offer

recommendations for prevention of the types of failure observed in the future. The incorporation of these recommendations by professional engineers into buildings for which

they have design responsibility, even though perhaps not incorporated into building codes

and standards, is by virtue of the professionalism of the engineer as a matter of good

engineering practice. This is not always the case in those types of buildings which may

not have a professional engineer or architect in responsible charge of the design. This

class of building is usually the single family house. It is because of this situation

that the Department of Housing and Urban Development has initiated a research program

into the area of the reduction and/or prevention of the types of failures observed in

single family housing in past earthquakes.

Among the many reports of the effect of past earthquakes, one of the most significant

was a study on the performance of single family dv/ellings in the San Fernando Earthquake

of February 9, 1971. This study found that the total financial losses to single family

dwellings in the San Fernando area were larger than the financial losses to any other

building category in the private sector. Wood frame dwellings generally had increasing

damage in the following order: one-story, two-story, and split level. The primary cause

of the overall damage to the buildings was attributed to the lack of adequate lateral

bracing. The types of dwelling components with the higher damage values were the exterior

wall finish, the interior wall finish on the exterior wall, and the interior partition

finish. The increase in damageability progresses from the non-brittle finished materials

(plywood) to the more brittle materials (stucco, gyspum lath, and plaster, etc.).

In order to attack this problem, the Department of Housing and Urban Development has

engaged the services of the Applied Technology Council, which is a nonprofit corporation

established in October 1971 by the Structural Engineers Association of California. ATC

will embark on a research project, the objective of which will be to develop a manual of

recommended construction practices for earthquake resistive single family residential

buildings for the use primarily of builders, building officials, field inspectors, plan

checkers, and designers. The manual will ejcplain the structural behavior of single family

dwellings subjected to forces produced by earthquake shocks, illustrate the Departmental

Minimum Property Standards and Building Code Requirements, and further illustrate sound

practical construction methods and details for the reduction of structural damage due to earthquakes. The manual will illustrate recommended construction details, architectural

layouts, types of construction recommended or to be avoided, and methods of installing

mechanical equipment to resist seismic forces. It is intended that the manual be selfexplanatory for rapid and clear understanding of the material contained therein.

Several resource documents illustrating single family house failures due to earthquakes will be used by ATC in performing this project which is anticipated to take approximately twelve months for completion. ATC will perform the project in a series of welldefined tasks including: a review of existing damage literature, a review of the HUD/FHA

MPS and earthquake code requirements, development of construction details including

supplimentary engineering analysis and an educational slide presentation to builder

organizations and related groups.

Of primary importance to the UJNR panel on wind and seismic effects is Task 3, which

is the development of construction details. ATC will develop typical engineering drawings,

illustrations, and details for residential dwellings which are required to resist forces

from earthquakes with appropriate descriptions and explanations so that the full intent

V-64
of the construction detail may be easily understood by a builder. The construction details

will be presented in such a manner that analytical calculations by the user of the manual

will not be necessary in order to fabricate the detail. If necessary, such details will

be categorized by the earthquake zones in the United States, i.e., 1, 2, or 3. However,

it will be required that the details be substantiated by calculations to the extent neces- sary for the engineering justification of the integrity of the design to resist seismic

forces. Typical details will be illustrated for the various types of construction most

commonly used in the United States for residential dwellings in seismically active areas including: wood frame with wood siding, wood frame and brick veneer, wood frame and stucco,

brick or block masonry, steel and/or aluminum frame with siding and/or masonry veneer and

other prevalent conventional combinations of framing materials and building components.

Included also will be construction details for basement and slab foundations, the structure

of the dwelling, installation of utilities and mechanical equipment, chimneys and fireplaces, attached garages and other architectural and structural components which affect

the strength, rigidity, and stability of the dwellings.

In order that this data can be effectively transmitted into the hands of the ultimate

users, a slide presentation accompanied by appropriate text material and voice tape will

be used to explain the contents of the manual to user groups, such as builders, building

officials, and designers. The presentation will be self-explanatory so that it may be

presented to the organizations without the need of experts for extensive interpretation.

Prior to extensive national exposure, presentations will be made by ATC to home builder

organizations, designers, and building officials in Los Angeles, San Francisco, Portland,

Oregon, and Seattle, Washington. After these sessions, the slide presentation will be

revised as necessary based upon the questions raised by the audience and the appropriate

answers which develop f^om the discussions. The revisions will be such that the presentation becomes self-explanatory so that it may be presented by others without further ex- planation being required.

While this paper has only dealt with our on-going research activity, the results of

which are not yet available, it was felt that it would be of interest to this body in terms

of a knowledge of the type of activity which is being sponsored by the Department of Housing and Urban Development in the area of disaster mitigation and loss control. At the

next meeting of the U.S. -Japan Panel on Wind and Seismic Effects, we will make available

the final reports of this research and present the audio visual program for your further

information.

V-6S
EARTHQUAKE ENGINEERING RESEARCH SUPPORTED

BY THE NATIONAL SCIENCE FOUNDATION

by

Charles C. Thiel

Program Manager, Earthquake Engineering

Research Applications Directorate (RANN)

National Science Foundation

A summary of earthquake engineering research work conducted by various re- searchers throughout the United States under the sponsorship of the National

Science Foundation is presented.

Key Words: Earthquake engineering; grant; RANN; sponsorship; structural engineering.

V-66
INTRODUCTION

Numerous recent piablications attest to the seriousness of the hazard presented by

earthquakes to peoples and economies, both of the United States and of the world (e.g.,

Refs. 1-6). Thus Chapter F (1) asserts, "Earthquakes, of all the natural disasters in

this country, can inflict the greatest loss of life and property. Studies have concluded

that a repetition of the 1906 San Francisco earthquake could cause billions of dollars of

damage, with the potential loss of thousands of lives. They are the most difficult

disaster phenomenon to prepare for." Reference 2 reinforces these conclusions by forecasting some $21 billion in earthquake losses for the state of California alone, the period

being 1970-2000, under the assumption of no "improvement of existing policies and practices.

The 1969 National Academy of Engineering report (3) has served as a useful basis for

recent RANN planning. Its introduction deserves to be quoted at length: "Earthquakes

devastate cities, with heavy loss of life, several times each year. A recent excimple is

the Khorasan, Iran, earthquake of September 1, 1968, with over 10,000 lives lost. The

United States has been shaken many times by large earthquakes, for example, 1964 Alaska;

1959 Hebgen Lake, Montana; 1949 Seattle, Washington; 1906 San Francisco, California; 1886

Charleston, South Carolina; 1857 Fort Tejon, California; 1811-1812 New Madrid, Missouri;

1755 Cape Ann, Massachusetts. The history of destructive shocks in the United States is

very short compared to other seismic regions such as Japan, the Middle East, and India

because the country has been inhabited by urban dwellers for only a short time. The

growth of population and the development of cities are so recent that it is only during

the past 100 years, or so, that the potential for great earthquake destruction has existed.

It is evident, however, from past occurrence of earthquakes that the highly seismic regions

of the country have a serious earthquake problem, and even the less seismic regions in the

central and eastern parts of the country have an earthquake problem which, although less

urgent, should not be ignored."

"Public welfare, in many parts of the country,, depends on facing the following questions: What intensity of ground shaking may occur? How will existing buildings respond

to the ground shaking? How should new buildings and other works of man be constructed so

as to minimize earthquake hazards?"

To illustrate the nation's life risk to earthquakes (and infer its property risk),

the following table lists the population at varying risk:

Seismic Risk U.S. Total California

Zone 0 (Low) 16.1 M (8%) 0

1 115.1 M (57%) 0

2 40.5 M (20%) 2.6 M

3 (High) 30.9 M (15%) 17.3 M

Zone 0 corresponds to no earthquake damage expected; Zone 1 - minor damage expected; Zone

2 - moderate damage expected; Zone 3 - major damage expected. Thus 35% of the U.S. population (71.4 M) are in risk zones 2 and 3, of which 28% are in California. The extent of

development in different regions of the country has led to the conclusion that there will

be as much damage to residential type structures in the next 200 years east of the Rocky

Mountains as west of the Rockies. The destructive potential posed by earthquakes is indeed a national concern.

Within the United States, Earthquake Engineering is a young field. The great San

Francisco Earthquake of 1906 has generally been credited with awakening the nation to the

disaster potential of the occurrence of quakes in its urban centers. The primary scientific consequence of the 1906 event was the identification and development of seismology

as a research area. It was not until the Santa Barbara earthquake of 1925, however, that

serious attention became focused on research into seismic design. Prior to this event,

the prevailing attitude was that if adequate information were made available to the designer about the phenomenon, he could cope with it directly. The 1933 Long Beach earthquake established earthquake engineering as a legitimate academic and professional research

V-67
pursuit and forced the inclusion of earthquake considerations in the building codes of California, and eventually elsewhere.

From NSF's beginning in 1950 to the Great Prince William Sound Alaskan Earthquake of

1964, the Fotindation ' s cominitment to the support of earthquake engineering research was limited to a few awards. The Alaskan quake caused extensive damage to modern structures.

It caused the NSF to assess its commitment to fund research in the field of engineering.

The Engineering Mechanics Program, following this assessment, identified Earthquake Engineering Research as a focused activity within its general disciplinary support of Civil

and Mechanical Engineering Research.

One of the new program's first steps was to have a coirprehensive report on the state of knowledge and research needs prepared (3) . A national conference was held to exchange

ongoing research information and establish priorities. The Universities Council on Earthquake Engineering Reserach (UCEER) was formed as a consequence of the latter activity. It acts as a forum for the exchange of ideas in the ixniversity community and as a focus for

coordination of research.

The first focused program expenditures were in FY 1966. For the next four years, the

Mechanics Program maintained an award level of approximately one million dollars annually

(see Figure 1) . During this period, a major emphasis was placed on developing a research

community capable of achieving substantial forward strides in the 1970 's. Awards were

made to a number of institutions and principal investigators.

The San Fernando Earthquake of February 9, 1971, furnished a major impetus in committing the NSF to vigorous support of Earthquake Engineering Research. About this time, the

Foundation initiated its Research Applications Directorate (RANN Program) and included the

Earthquake Engineering Program as part of its activities. From its inception to the crea- tion of RANN, Dr. M.P. Gaus was in charge of the Earthquake Engineering Program. Subsequently, the program has been vinder the direction of Dr. Charles C. Thiel.

The program has grown since becoming a part of RANN from a base level of $2.5 million

in FY 1971 to a level of $8.0 million in FY 1974. The program has assumed extensive new research responsibilities during this period. In 1972, major efforts in utilization (Data

Base, Information Dissemination, and Technology Transfer) were initiated. In 1973, the

basic program was augmented by initiating research in the area of System Response. In 1974, the areas of Implementation Studies and Policy Studies were added to make the program

more comprehensive. (See 'Section 3 for details) . Between 1966-1974, the program has made

awards to over 100 principal investigators at over 35 institutions. Not listed in these

proceedings are the details of the awards during the past three years. Currently^

about 65 grants and contracts are active. Although awards have been primarily to universities, we are beginning to use the resources of professional societies, governmental units

and profit making organizations. Subsequent sections describe the specific objectives of

the NSF program.

GENERAL PROGRAM OBJECTIVES

The occurrence of an earthquake, or any other natural event, is important only insofar

as it affects man and his works. It is the disaster potential of earthquakes that has

caused man individually and collectively to seek adjustments to his physical and social

environment that decrease his vulnerability. The measurement of vulnerability is related

to the potential or realized loss of life (and injury) , property damage, and/or disruption

of function. While each individual or group will apply a different normative combination

of these measures in its decisions, the general objective remains to control the sequences

of the event.

As adjustments, one may seek either to change the phenomenon or to alter the response

of man and his works -lo the phenomenon. Possible physical adjustments the decision maker

may seek are to:

V-68
$ 8M

66 67 69 70 71 72 73 74

Fiscal Year

Figure 1 - Earthquake ETn^ineering Prograr. Expenditures, 1966-1974

with center two year noving average

V-69
Control the event by prevention or modification of the event

Anticipate the event so that remedial actions may be taken

Identify the seismic potential of areas

Construct facilities so as to perform acceptably during and after

the event

Among the possible social adjustments, the decision maker may seek to:

Plan for the warning, response, and recovery to the event

Distribute economic risk

Generate and select alternative physical development plans

Adopt and enforce zoning, construction, and management standards

Clearly the physical and social adjustments depend critically on each other. It must

be remembered that adjustments to earthquakes are not made in vacuo; there are other con- siderations in making decisions regarding vulnerability reduction.

Noting the responsibilities of other Federal agencies to develop techniques for earth

quake prediction and control (USGS) and noting the types of adjustments given above, four

general objectives have been identified to fulfill the overall program objective. The

latter is stated first.

RANN Earthquake Program Objective

To develop methods that allow decision makers to control the consequences

of earthquake occurrences.

In furtherance thereof, the RANN Program supports research projects leading to:

A. Design Development of economically feasible design and

construction methods for building earthquake re- sistant structures of all types.

B. Land Use Development of procedures for integrating information on seismic risk with on-going land use planning processes.

C. Social Development of an improved understanding of social

and economic consequences of individual and community decisions on earthquake related issues.

D. Implementation Presentation of program results in forms usable by

the affected interest communities to control their

vulnerability to earthquakes.

SPECIFIC OBJECTIVES AND PROGRAM ELEMENTS

In order systematically to structure a program over a period of years aimed at achieving the four general objectives, nine program elements have been identified. With each

of these elements are associated several specific objectives, which serve as guidelines

for developing that element. In the context of the present report, however, particular

project descriptions are inappropriate; rather, the program elements serve as categories

into which specific research activities can be divided.

V-70
It should be noted that this method of structuring earthquake engineering and related

fields owes a debt to the NAE Committee report (3) . The detailed program breakdown and

recommendations prepared by NAE have proved very useful in structuring what follows here.

The nine elements have been framed so as to deal with specific problems in the design, analysis, and synthesis of engineering and scientific knowledge to control the consequences

of earthquake, measured by life loss, property loss, and/or function loss.

I . Ground Motion/Data Services

Destructive ground motions resulting from earthquake action are of several different

types, including heavy ground shaking, slow or rapid fault slip, subsidence, and landslides.

Fundamental to an understanding of any of these damaging phenomena, however, is an accurate

knowledge of the actual earthquake ground motion. Fundamental to the validation of structural design and analysis procedures is the actual structural response generated by the

ground motion.

The measurement of destructive ground motions in the epicentral region of large earthquakes and the associated response of structures is achieved by placing a network of instruments at a variety of sites where earthquakes are likely to occur. These instruments

include both passive and active recording devices. The strong motion accelerograph records

the time history of acceleration in three component directions. It begins recording after

an acceleration threshold has been exceeded. These instruments are the most generally

used and find applications in most types of data gathering. Other active recording systems

include pore pressure gauges to measure the liquid pressure in saturated soils, earth pressure gauges to measure the inertial effects of soils usually on a foundation wall, strain

gauges, and displacement meters. Among the passive instruments are the seismoscope, a conic pendulum that records motion on a smoked glass, scratch strain gauges and extensometers

.

The measurements obtained are directed at achieving three principal objectives:

A. To support the research program by measuring pertinent quantitities

to validate, calibrate, and/or formulate theories of earthquake response.

B. To support the designer's need for earthquake motion information at varying geological and seismological sites.

C. To obtain a comprehensive data base to perform microregionalization of

earthquake risk areas based on events in the area.

These objectives are both research and operational program related. During FY 1973,

NSF assumed responsibility for the Seismological Field Survey. SFS is the principal focus

for strong motion instrument networks. Using SFS as its principal agent, the RANN Program

will: develop criteria for placing an optimal strong motion recording network; begin the

placement of this network by new installations and adjustments to the existing network;

develop and place specialized instrument networks to answer specific research needs, e.g., down hole soil response arrays, detailed structure response; develop a qualified products

list of existent instrumentation, define characteristics for new instrumentation, and

ensure steady improvement of instrument quality, sensitivity, and reliability.

II . Soils Response and Analysis

One of the greatest potential sources of property loss is damage to structures that

rest on soils or on foundations which, although adequate to support the structures under

ordinary circumstances, might fail during an earthquake. A great potential for life loss

resides in the possibility of a dam failure. The following areas represent forms of soil

failure under consideration by the program: settlement of cohesionless soils; bearing

capacity failure; embankment failure; soil liquefaction; and waterfront bulkhead failures,

etc.

V-71
The dynamic behavior of a structure during an earthquake depends on the shaking trans- mitted to it by the surrounding soil. These motions are imparted to the structure through

the interaction between its foundation and the supporting soil and/or bedrock and include

amplification effects.

The objectives of this element are to:

A. Develop methods to evaluate and control soil failure potentials.

B. Develop analysis and design methods to evaluate and control soil amplification.

C. Develop design and analysis methods to characterize the loadings trans- mitted to structures through soil- foundation interaction.

III. Structural Response and Analysis -

.

The realization of a structure rests on two complementary activities: analysis, and

synthesis and design. Analysis forms the basis for design. The ability to analyze a hypothetical structure and determine the stresses and displacements that would be produced

by a specified loading is an essential part of the design process. The more accurately

this can be done, the more efficient and economical can be the design and the more reliable

the design factor of safety. The analysis of a structure can be exceedingly difficult,

first, since ordinary structures are exceedingly complex dynamic systems; second, because

the ground motions which the structure will be subjected to during its lifetime are probabilistic; and third, because the construction process leads to a structure which is not

precisely known. The analysis of structural response requires knowledge of the full system

including foundations, adjacent soils, and in some cases the properties of adjacent structures .

The objectives for this program element are to:

A. Determine appropriate models for element (e.g., beam, column, plate,

etc.) response to strong motion excitation from analytic and experimental studies.

B. Develop digital computer methods of analysis that predict earthquake

response of structural systems comparable in complexity to real

structures

.

C. Validate and calibrate these models by comparison with earthquake

measured motions and damage of structural system and elements.

The basic problem of earthquake design is to synthesize the structural configuration;

the size, shape, and materials of the structural elements; and the methods of fabrication,

so that the structure will safely and economically withstand the action of earthquake

ground motions. The object of design is to control the effects of an earthquake on a structure and keep damage within acceptable bounds. A further series of objectives has

been detemined for this element:

D. Determine the dynamic properties of structures, elements, and materials

under conditions of large strains, beyond the yield point and up to fail- ure .

E. Develop reliable, practical, and simplified methods of earthquake design

for widely used and special important structures: e.g., low-rise residential, low-rise commercial, school houses, high-rise buildings, dams,

bridges, and industrial structures.

F. Thoroughly study structures that failed during an earthquake, as well as those that did not fail, to refine design principles.

V-72
G. Develop design and analysis methods for nonstructural elements and

mechanical components of structures.

rv. Systems Response

Previous elements of the research program have dealt almost exclusively with localized

structiires, that is, a single building, dam, etc. The operation of the community during

the emergency and recovery periods is dependent on the functioning of utility and public

service facilities that function as a system with elements located at many sites in the

affected area. The failure of an element can cause the total system to malfunction or be

inoperative. Thus the design of system elements must consider the seismic characteristic

requirements of the extended system. Both physically connected and non-connected systems

are to be investigated. An example of a connected system is fire water distribution (stor- age, pumping stations, water mains, etc.) while a non-connected system is represented by

emergency health care facilities (hospitals, clinics, laboratories, etc.).

The objectives of the element are to:

A. Develop principles of planning and laying out facilities for minimum disruption of operations due to earthquake ground displacements.

B. Develop design procedures for special structures and equipment of each

type of utility and public service facility. Among these, ranked in

approximate priority order, are:

1. Fire fighting and emergency transportation

2. Emergency power and communications

3. Hospitals and emergency agencies

4. General communications

5. Water services and sewage disposal

6. Electrical power and natural gas or fuel supplies

7. General transportation facilities

C. Evaluate existing methods of land-use control, such as microzonation, that

will allow a community to control its earthquake vulnerability.

V. Coastal and Inland Waterways

An earthquake at sea may generate a tsunami, or tidal wave, that presents a real danger to coastal and island regions of the U.S. The risk posed by tsunamis may best be con- trolled by land use regulation in vulnerable areas. Verification of the occurrence of a tsunami permits the evacuation of potential affected areas.

If a dam fails, an inundation wave may be generated when the reservoir empties. This wave can cause serious downstream damage. Structuring downstream land use, with this

potentiality in mind, can reduce life and property loss. When a reservoir behind a dam

is filled, a series of earthquakes often occurs in the vicinity of the impoundment. The

incremental risk these quakes cause above the natural risk is at present uncertain.

The objectives of this element are to:

A. Develop methods to verify that an ocean-based earthquake has generated

a tsunami.

"Note that nuclear power plants and related facilities, which would otherwise head such a list, are a responsibility of the Atomic Energy Commission.

V-73
B. Develop methods to anticipate tsunami run-up in coastal regions.

C. Develop methods to anticipate downstream inundation levels.

D. Determine the incremental seismic risk associated with filling a reser- voir and the degree that this risk should be part of the design criteria

for the dam.

VI . Technology Transfer

Within the Earthquake Engineering Program, utilization activities have been centralized and formulated as a specific program thrust, to be conducted by a core series of

awards rather than distributed as a component in each individual award.

The objectives in this element are to:

A. Maintain a technical reference collection of piiblished reports and

papers and unpublished data available to researchers and professionals;

to provide for information dissemination.

B. Maintain a software center to archive, document, validate, and distribute computer programs developed for earthquake engineering applications

.

C. Consolidate the best knowledge regarding design and analytic methods

from current and completed research and professional experience in codes,

criteria, and standards for use by professionals and regulatory bodies.

D. Conduct regular conferences and workshops to act as foci for dialogues

between the research and professional communities.

E. Conduct post earthquake inspections to obtain engineering data and to identify unfulfilled research needs.

VII. other

This category includes conference travel, general conference support, advisory committee meetings, program planning, and other activities that do not fit conveniently under

one of the above classifications.

VIII. Implementation Studies

Advances in civil engineering methods are implemented through two types of actions:

individual professional application; and adoption by regulatory/enforcement agency. The

latter process is primarily political, while the former is educational. The structuring

of the flow of technological developments from research to implementation is a particularly arduous one, involving a vast complex of institutions, special interest groups, and

secondary agendas. The way in which this process works is not well understood. An understanding of the process is a first step in structuring a methodology to foreshorten the

period from research to practice.

The objectives of this program element are to:

A. Develop an understanding of the implementation process.

B. Develop a methodology to program the education of professionals, using

suitable aids, to the best earthquake engineering practice.

C. Develop a strategy of program results to be processed through the com- munity into regulatory and enforcement agency regulations in a timely

fashion.

V-74
D. Advocate the accomplishment by public and private groups of the developed

plan.

The accomplishment of these goals will have far reaching implications on several RANN programs. Indeed, the incorporation of earthquake engineering reserach results is but a small part of the problem of moving technical innovations into building practices. The same basic methodology is applicable to fire safety, energy conservation, building utilities, and novel energy systems such as solar heating and cooling. The program element is

beyond the single interests of the Earthquake Engineering Program and, as such, will be

considered within the context of the building sciences.

IX. Social and Behavioral Studies

The mitigation of earthquake effects may depend on a variety of actions not directly

associated with the assurance of safe structures. The possibilities of earthquake control

(or modification) and prediction present different technical methods that may lead to mitigation. Community actions during and following the event may significantly affect

the consequences of the quake. Condemnation and redevelopment attitutdes, insurance programs, and economic assistance may markedly affect the communities' vulnerability to future events. With a view to filling important gaps in our understanding of earthquakes

as they iitpact man and his social environment, research in the social and behavioral sciences, and in economics, will be undertaken to:

A. Develop policy alternatives for the community to provide emergency

rescue, recovery, and redevelopment services.

B. Develop policy alternatives for the dissermination of earthquake

warnings

.

C. Assess the consequences of technical developments and policy alternatives on the achievement of disaster mitigation.

SPECIFIC RESEARCH PROGRAMS

The National Science Foundation has supported a wide spectr\im of research on earthquake

hazard reduction. Rather than detail the specific character of these projects. Appendix

A presents the titles, performer, and institution for each project initiated in the past

three years

.

Two recent projects warrant specific highlighting.

These have as their objective the development of improved earthquake design criteria,

standards, and codes. The first of these projects is a test of the practical design feasibility of new code procedures and assumptions, plus an evaluation of the economic conse- quences of such a "new code."

The project has four chief objectives:

Draft a set of design rules based on a "design spectrum" type of code;

Apply these rules to the redesign of several types of existing buildings

in order to evaluate their applicability by example;

Make a preliminary assessment, as part of a literature survey, of methods

by which such a code could include consideration of variations in regional

seismicity, proximity of known active faults, and soil amplification and

topographic effects of local site conditions; and.

Make a preliminary evaluation of the economic and performance effects

of such proposed changes on the example buildings.

V-75
The work statement of this project divides the acconplishment of the above objectives

into two separate phases. Phase I covers the establishment of the design rules and the

effect-evaluation procedures. The application of the rules, the literature survey, and

the actual evaluation of economic and performance effects will be achieved in Phase II. Fortunately, some of the existing buildings to be studied have been subjected to special

studies as a result of the 1971 San Fernando earthquake. These special studies will prove

useful in the Phase II evaluation of the Phase I design rules.

In this study, the lateral force resistance systems of eleven existing buildings (see

Table 1) of varying structural types and heights are to be redesigned for the effects of

a dual-earthquake criterion: Specifically, a damaged threshold spectrum, and a collapse

threshold spectrum. Current applicable portions of the 1973 Uniform Building Code and

the SEAOC Code are to be employed, with the exception that earthquake loading and related

deformations are to be computed by the use of the dual spectra. Damping and ductility

values have been assigned vitiich are considered to be within certain assumed capabilities

of each type of structure being studied. Each of the eleven buildings contains a valid

lateral force resisting system within the K-factor categories of the current SEAOC Code.

Each is reasonably regular in plan and elevation, therefore, the problems of torsional

racking, setback response, and soft-story effects will not be significant in the analyses

of these particular buildings. (These problems are recognized as important, however.)

The project is nearing completion.

The second project builds upon the first. It is just beginning and will not be com- pleted until late 1975. The objective, simply stated, is to develop a new set of provisions for the SEAOC Lateral Force Requirements. The current criteria are based on the

technology of the 1950 's with minor updating in subsequent revisions. This study entails

carrying out a comprehensive program of updating and revising the present SEAOC seismic

requirements so that it will be applicable throughout the United States based upon the

latest state-of-the-art of earthquake engineering and construction practices. The format

of the seismic provisions will be augmented to include provisions for other natural mete- orologically caused phenomena such as hurricanes, tornadoes, and wind storms. The format

will follow existing building codes so that it can be codified, adopted, and promulgated

by various groups writing and implementing code provisions. The new seismic provisions

will be coitprehensive and written with a national perspective. Provisions concerning the

code will include all aspects of building and applicable geotechnical practices for mitigation of losses from earthquakes and earthquake related geologic hazards. The specifications will be based on goals and objectives developed from a performance criteria to be

clearly set forth in the provisions so that appropriate regulatory bodies can recognize

the level of risk associated with the various code provisions. It will be a balanced

statement with proper consideration of low-rise (one to three stories) , intermediate-rise

(four to six stories) , and high-rise (over six stories) as well as other nonbuilding type

structures. Provisions for nonengineered buildings and engineered buildings will be included. Recognition will be given for different construction practices and different use

of materials throughout the United States. The provisions will be stated in simple terms

to the fullest extent possible consistent with the goals of the code, the state-of-the-art

and practice to produce these objectives, and the complexities of the earthquake phenomenon. The provisions will set forth the loading criteria and performance (resistance)

criteria with recommended methods of design and analysis consistent with the seismicity

of the site, the importance of the structure, the type of construction, and height of the

structure

.

It will include where feasible, provisions for considering, within the framework of

earthquake engineering, various facets such as architectural configurations, contract

documents, construction team, quality control, supervision and inspection as well as structural analysis and design.

It will be comprehensive in scope and will be developed after consideration of many

provisions not included or adequately covered in the present seismic codes. Evaluation

of the state of the knowledge, art and practice relating to each provision will be used

to make recommendations for their inclusion in the code. Among such elements are the

following:

V-76
TABLE 1

BUILDINGS TO BE REDESIGNED IN THE SEISMIC CODE EVALUATION STUDY

Building 1

Building 2

Building 3

Building 4

Building 5

Building 6

Building 7

Building 8

Building 9

Building 10

Building 11

Five stories plus basement

Ductile reinforced concrete frame

Nineteen stories plus four levels basement parking

Steel frame, moment resistant in one direction.

Braced in the other direction

Ten stories

Semi-ductile reinforced concrete frame in one direction.

Shear wall in the other direction

Fourteen stories

Shear walls on both major axes of building

Six stories

Shear walls are non-load bearing

Two stories

Steel frame with vertical bracing system

Two stories

Concrete block masonry, bearing and shear walls

Nine stories

Reinforced concrete, shear wall

One story

Tilt-up, plywood diaphragm, re-entrant corner

Two-story school

Moment resistant steel frame

One-story school

Brick masonry, plywood diaphragm

V-77
1. Goals for structural and non-structural damage.

2. Post-earthquake factors, importance factors, and other socioeconomic considerations.

3. Various loading and analysis methodologies and when they should be

used: equivalent static force approach, elastic single degree of freedom and multidegreee of freedom modal superposition response spectral

analysis, elastic and inelastic time history response analysis and deterministic and probabilistic rock and ground motions.

4. Seismicity, probability consideration of various design earthquakes,

seismic risk, and seismic zone concepts.

5. Design earthquake criteria based on dual performance for the moderate

design earthquake and the severe design earthquake based on maximum

credible earthquake.

6. Geologic, topographic, soil and other site condition influences on ground motion.

7. Basic realistic levels of ground motion to represent the design earthquake at a site of average exposure having no unusual soil conditions.

8. Influence of soil conditions on intensity of ground shaking in terms of

displacement, velocity, and acceleration.

9. Soil structure interaction, including influence of the structure on the

ground motion and the influence of ground motion on the structural re- sponse.

10. Structural system damping, ductility and stability including P-delta

effects

.

11. Relative performance criteria of structural systems and materials with

particular attention to brittle material behavior and elasto-plastic

response

.

12. Realistic deterministic ultimate design stresses, load factors, and re- sistance factors for all materials.

13. Specific criteria to be used for dynamic analysis and design.

14. Drift limitations consistent with realistic response to strong earthquakes.

15. Vertical acceleration criteria.

16. Shear wall frame interaction provisions.

17. Hazardous and earthquake damaged buildings and their rehabilitation

and reconstruction.

18. Seismic design requirements for mechanical, electrical elevator and

other building "life-safety" systems.

REFERENCES

Office of Emergency Preparedness, Executive Office of the President, Disaster P repared- ness , 3 Vols. Report to the Congress, January 1972, U.S. Govt. Printing Office, Washington, D.C. , 1972.

V-78
2. Alfors, J.T. , J.L. Burnett, and T.E. Gay, Jr., Urban Geology, Master Plan for California , Bulletin 198, California Division of Mines and Geology, Sacramento, 1973.

3. National Academy of Engineering Committee on Earthquake Engineering Research, Earthquake

Engineering Research , published by NAS, Washington, D.C., 1969.

4. Seismology: Responsibilities and Requirements of a Growing Science , Parts I and II,

National Academy of Sciences, 1969.

5. Federal Council of Science and Technology, Office of Science and Technology, Executive

Office of the President, Proposal for a Ten-Year National Earthquake Hazards Program , U.S. Govt. Printing Office, Washington, D.C., 1966.

6. Office of Science and Technology, Executive Office of the President, Earthquake Prediction, a Proposal for a Ten-Year Program of Research , U.S. Govt. Printing Office, Washington, D.C. , 1964.

Appendix A which lists NSF earthquake engineering awards in 1972-1974 is not included

herein. The list may be obtained from the author.
THE WIND ENGINEERING PROGRAM

by

M.P. Gaus

Head, Engineering Mechanics Section

National Science Foundation

Washington, D.C. 20550

A summary of wind engineering research work conducted by various

searchers under the sponsorship of the National Science Foundation is

sented

.

Key Words: Research programs; sponsorship; wind engineering.

V-80
WIND ENGINEERING

Wind engineering is concerned with the interaction between wind and either engineered

facilities or natural systems which are located in the bovindary layer of the atmosphere or roughly from ground surface to as much as 800 meters in elevation. It is precisely in this

boundary layer where the average citizen conducts his daily affairs, where we construct

most of our facilities, and discharge the gaseous effluent of our modern technological

society. From an engineering viewpoint this boundary layer is virgin territory with a dearth of utilizable information for design or solution of problems. Unfortunately, little

assistance is obtained from meteorological studies which tend to concentrate on higher elevations .

The Wind Engineering Program presently identifies needed research studies in sixteen

categories. Some of these are oriented toward generating or collecting the basic information required for the development of wind design, analysis, and simulation methods. Others

are oriented toward providing general information of use to practitioners and for improvement of codes and regulations and in planning procedures. The remainder are oriented

toward economic and people problems which are an important component of engineering problems. At the present time, research is or has been supported which related to about twelve

of the categories. At the present level of activity, however, the amount of effort in a number of these categories is rather token.

Shortly after the wind Engineering Program was proposed, a series of intense tornadoes

converged on the city of Lubbock, Texas, causing extreme damage to the city. Within hours

after the tornado struck a combined team of engineers and meteorologists was dispatched to collect perishable information. This was followed up through grants to Texas Tech University and the Institute for Storm Research, University of St. Thomas to reconstruct from

this information details of the wind and pressure differential loads and missile effects

caused by the tornado. The resulting information, published in several reports, has provided for the first time a factual and reliable base of information regarding the actual

effects of an intense tornado from an engineering viewpoint. The results indicate that the

wind velocities which must be used in design are only about 1/2 to 1/3 of those which had

been claimed by meteorologists and make tornado resistant design feasible for many structures. Other tornado studies in progress include an experimental program at Catholic University in which special vortex tunnels are being used to measure pressure distributions,

forces, and moments on various types of buildings as on intense vortex passes; a study at

the Los Alamos Scientific Laboratory to develop numerical analysis of tornado wind loads

related to various flow variables; a study at Tulane University on prester tornadoes; and

a study at George Washington University which is looking into the relationship between

different terrain conditions and the wind loadings of ground structures inside intense

tornadoes

Another wind problem concerns the failure of glass, shingles, cladding, or isolated

components due to wind. The yearly damage due to such effects is quite widespread and substantial, but is perhaps most dramatically illustrated by the much publicized difficulties

of the new John Hancock Mutual Life Insurance Company Building in Boston, Massachusetts.

Although wind tunnel studies were made in the design stage for this building, they clearly were not adequate. Instrumentation has been installed in the John Hancock Building by a group from MIT under an NSF grant, with heavy financial participation by the Hancock In- surance Cortpany, to take advantage of this unfortunate situation. This information along

with other analytical and experimental studies will probably be instrumental in revising

the building regulations in Boston and other cities.

Clearly, v/ind tunnel testing is required to gain a better understanding of boundary

layer aerodynamics and for predicting various effects before construction takes place. A comprehensive program is being conducted at Colorado State University utilizing special

long-fetch adjustable roof wind tunnels for simulating natural winds. Problems to be ex- amined include local flow phenomena, local wind pressures and heat transfer rates on building surfaces, dynamic excitation by buffeting, wind speed and gustiness at the pedestrian

level and circulation of air pollutants. Although the CSU facilities are currently the

best available in the U.S., there are still unresolved questions regarding the scale and

V-81
distribution of turbulence, simulation of gustiness, and the validity of the extreme scaling from present wind tunnel sizes to full-scale structures. The latter problem can only

be resolved by comparing wind tunnel tests to full-scale observations. The first problem

is being studied at Notre Dame University where a feasibility study is being carried out

for a large multipurpose boundary layer wind tunnel using an active rather than passive

turbulence generation process

.

Also of great concern are the environmental wind conditions around buildings. Through

grants to Illinois Institute of Technology and Catholic University, studies are in progress

on the microclimate of buildings and urban areas. These studies directly relate to pedestrian comfort, concentration of pollutants between buildings or in depressed roadways, and

to buffeting and missile problems.

Space does not premit descriptions of other studies in progress on windwave effects,

development of new wind instrumentation for field observation, motion perception and tolerance, etc.

Studies just started or scheduled for future action include the instrumentation of

full-scale buildings, cooling towers, and other structures to obtain actual measurements

of wind effects. These full-scale measurements will be compared to very thorough wind

tunnel studies modeling the structures and terrain. Comparisons will be made to determine

the accuracy of wind tunnel determinations of wind characteristics, wind loadings, dynamic

response, interaction between wind and energy consumption in buildings, performance of

cooling towers, effect of wakes, and environmental effects. At the same time, computer

software for wind engineering analysis is under further development so that both the tools

and supporting data can be made available to practicing engineers.

Also being started are studies of storm-surge and coastal effects in hurricanes and

other strong wind storms.

Both industry and state and local governments are being encouraged to participate in

the funding of Wind Engineering research programs. Substantial progress has been made in

this respect.

In order to coordinate wind research activity and to disseminate information as rapidly as possible, two actions have been taken. The first is the provision of funding for

a Wind Engineering Research Council which will serve as a focal point to coordinate university-industry-government wind research, to organize workshops in specific research

areas , and to serve as a contact point for information on wind engineering problems . The

second is the provision of funding to prepare a wind engineering research digest which

will provide up-to-date information on wind research in progress.

The discription of the various grants allocated during July 1973 to June 197U

are not included herein, but may be obtained from the author.

V-82
PRELIMINARY REPORT ON PRESENT STATUS AND DEVELOPMENT PROJECT

OF VOLCANOLOGICAL OBSERVATION AND RESEARCH IN INDONESIA

by

Akira Suwa

Head, Seismological and Volcanological Laboratory

Meteorological Research Institute

Japan Meteorological Agency

Tokyo, Japan

Indonesia has about 130 active volcanoes and their eruptions are characterized by dangerous violent explosions, nuee ardente, and volcanic mud-flows.

The Geological Survey of Indonesia (GSI) , Ministry of Mines, has, therefore,

been carrying out observations and surveillances of volcanic activities through

out the country.

The Government of Japan, in response to a request from the Government of

Indonesia, has decided to give assistance in this field of science in the frame

work of the Colombo Plan.* The Overseas Technical Cooperation Agency (OTCA)

,

the executing agency for the Government of Japan, has, therefore, dispatched a preliminary survey mission, headed by the author of this report, Akira Suwa, of

the Japan Meteorological Agency to Indonesia in 1972.

The mission stayed in Java and Bali from November 22 to December 23, 1972,

and visited the Ministry of Mines at Djakarta, the GSI at Bandung, and eight

active volcanoes (eleven observatories) in order to study the possible scope of

cooperation in Volcanology between Japan and Indonesia. The mission recognized

two serious problems in Indonesia: deficiency of experts in volcanology, and

shortage of up-to-date volcanological instruments.

Therefore, the recommendation by the preliminary survey mission to both

governments was as follows:

1. Dispatch for several years the following Japanese experts to Indonesia

a. Instrumental seismologist

b. Volcano physicist

c. Volcanological geologist/petrologist

2. Train junior volcanologists of the GSI in Japan.

3. Provide Indonesia with the following instruments;

a. Seismographs for permanent and temporary observations

b. Instruments for petrological and mineralogical laboratory

Key Words: Field observation; Indonesia; Japan; technical aid; volcanos.

*Colombo Plan for Technical Cooperation in South and Southeast Asia

VI-1
OBJECTIVE OF THE JAPANESE PRELIMINARY SURVEY MISSION ON

VOLCANOLOGY TO INDONESIA UNDER THE COLOMBO PLAN

The Government of the Republic of Indonesia requested the Government of Japan to provide technical assistance in order to develop a program on volcanological works for the

Geological Survey of Indonesia (GSI) . (GSI) also asked that a detachment be sent on a preliminary survey mission, which was headed by Akira Suwa, writer of this report. The

Overseas T-echnical Cooperation Agency (OTCA) , which is an executing agency of the Government of Japan, sent the preliminary survey mission on volcanology to Indonesia under the

Colombo Plan in November, 1972.

The main objectives of the mission were as follows:

1. To investigate the status of Indonesian active volcanoes, the type of volcanological observations, and speak with the officer of GSI.

2. To discuss the problems of such cooperation with the Indonesian authorities.

MEMBERS OF THE PRELIMINARY SURVEY MISSION

(Chief) Mr. Akira Suwa

Head, Seismological and Volcanological Laboratory

Meteorological Research Institute (MRI)

Japan Meteorological Agency (JMA)

(Member) Mr. Masaaki Seino

Seismologist, Seismological Division .

. Observation Department, JMA

(Member) Mr. Yoshihiro Sawada

Volcanologist, Seismological Division

Observation Department, JMA

ITINERARY OF THE MISSION (NOVEMBER 22 - DECEMBER 23, 1972)

The preliminary survey mission, which stayed in Indonesia for about one month, visited

the following offices, observatories, and active volcanoes:

Ministry of Mines at Djakarta

Geological Survey of Indonesia at Bandung -

Volcano Tangkuban Prahu and Tangkuban Prahu Volcano Observatory

Volcano Papandajan

Volcano Dieng and Karang Tengah Volcano Observatory '

Merapi Central Observatory at Jogjakarta

Volcano Merapi (Central Java) and Babadan Volcano Observatory, Ngepos Volcano

Observatory, Selo Volcano Observatory, and Jrakah Volcano Observatory

Volcano Kelud (Crater Lake) and Maragomulyo Volcano Observatory

Volcano Semeru and Argosko Volcano Observatory and Vudakeling Volcano Observatory

Volcano Batur . >

. .

,

VI -2
Volcano Krakatau seen from the west coast of Java Island

Embassy of Japan at Djakarta

Djakarta Office of the OTCA

The itinerary of the mission is as follows:

November 22 (Wednesday) Tokyo-Hongkong-Singapore-Djakarta.

23 (Thursday) Djakarta-Bandung. Visit the GSI.

24 (Friday) Visit the Geological Survey of Indonesia.

25 (Saturday) Ditto.

26 (Sunday) Visit Volcano Tangkuban Prahu and its Observatory.

27 (Monday) Visit Volcano Papandajan.

28 (Tuesday) Bandung-Jogkakarta.

29 (Wednesday) Visit Merapi Central Observatory at Jogjakarta.

30 (Thursday) Visit Volcano Dieng and Karang Tehgah Observatory.

December 1 (Friday) Visit Merapi Central Observatory at Jogjakarta.

2 (Saturday) Visit Volcano Merapi. Babadan and Ngepos Observatories

3 (Sunday) Visit Volcano Merapi. Selo and Jrakah Observatories

4 (Monday) Jogjakarta-Maragomulyo.

5 (Tuesday) Visit Volcano Kelud and Maragomulyo Observatory.

6 (Wednesday) Maragomulyo-Malang

.

7 (Thursday) Malang-Volcano Semeru and Argosko Observatory - Dj ember.

8 (Friday) Djember - the Straits of Bali - Denpasar.

9 (Saturday) Visit Volcano Agung. Rendang and Budakeling Observatories.

10 (Sunday) Visit Volcano Batur.

11 (Monday) Denpasar-Jogjakarta.

12 (Tuesday) Jogjakarta-Bandung.

13 (Wednesday) Visit the Geological Survey of Indonesia.

14 (Thursday) Ditto.

15 (Friday) Ditto.

16 (Saturday) Ditto.

17 (Sunday) Bandung-Djakarta.

18 (Monday) Visit the Embassy of Japan and OTCA Office at Djakarta.

VI-3
December 19 (Tuesday) Write the tentative report of the mission.

20 (Wednesday) Visit the Embassy of Japan and OTCA Office at Djakarta.

21 (Thursday) Visit the Ministry of Mines at Djakarta.

22 (Friday) Visit the west coast of Java Island to see Volcano

Krakatau

.

23 (Saturday) Djakarta-Singapore-Hongkong-Tokyo

.

VOLCANIC ACTIVITIES IN INDONESIA

The Indonesian archipelago has about 130 active volcanoes, including solfatara fields

(ca. 15% of the total of those in the world) , as is shown in Figure 1 and Table 1. These

volcanoes have provided the people with various kinds of natural resources and beautiful

scenery. In general, Indonesia volcanic areas are high in altitude and are densely populated, because the archipelago is located in the torrid zone and the air temperature at

the seaside areas is very high all the year round.

Consequently, eruptions of these volcanoes can cause serious loss of human life and

property. To make the matters worse, eruptions of the Indonesian volcanoes are characterized by extremely dangerous phenomena, such as catastrophic explosion, nuee ardente

(glowing avalanche), and volcanic mud-flow (the so-called "lahar" in Indonesian language)

.

As an example, twelve of twenty most disastrous eruptions on the earth have taken place

in this country, as shown in Table 2. Lahar is divided into two categories: "eruption

lahar" and "rainfall lahar". Both types occur frequently at the various volcanoes in In- donesia.

Before the World War II, the Dutch Government tried to conduct volcanological observations and research and thus mitigate volcanic disasters in this territory. Since their

declaration of independence in 1945, the Government of Indonesia has been making similar

efforts to develop volcanic research capabilities. The Geological Survey of Indonesia,

which belongs to the Ministry of Mines, is in charge of the observation and svirveillance

of volcanic activities all over the countiry.

The Volcanological Division of the (GSI) at Bandung, which has sixty members, con- sists of two sections, and has two teams concerned with special research projects, as shown in Figure 2. The Volcanological Division also has a branch station "Merapi Central

Observatory" (the center for volcanological observation and research of Mt. Merapi) at Jogjakarta and 25 volcano observatories at sixteen active volcanoes on nine islands,

(see Table 3.) These observatories are manned by a total of sixty members.

Mr. D. Hadikusumo, Head of the Volcanological Division, and Mr. I. Surjo, scientist

in charge of the Project for Volcanic Debris Control, came to Japan in 1957 for one year

to study volcanology under Prof. T. Minakami of Tokyo University, and the author of this

report, at the Japan Meteorological Agency. Many of the other leading members of the

Volcanological Division also studied volcanology in Japan during recent years. However,

many members working in the volcano observatories have little schooling although they are diligent and faithful workers.

The Volcanological Observation Section not only manages all the volcano observatories,

but also conducts temporary observation at various volcanoes (periodical and urgent observations) . The Volcanological Research Section is also conducting research on the mitigation of volcanic disasters, and at present the section is primarily engaged in the zoning

THE OBSERVATION AND SURVEILLANCE OF VOLCANIC ACTIVITIES

BY THE GEOLOGICAL SURVEY OF INDONESIA

VI-4
of the dangerous area of each volcano, according to three grades of eruption, which are:

(see Figure 6 a, b, c, d.

1. Permanently Off-Limit Zones: Dangerous even in case of minor eruptions,

everyone should be outside zone.

2. First Dangerous Zone: Dangerous in case of big eruptions.

3. Second Dangerous Zone: Apt to be attacked by volcanic mud-flows.

At present, there are many inhabitants even in the Permanently Off-Limit Zones, at such active volcanoes as Merapi, Kelud, Semeru, Agung, etc. It is actually very difficult

for the authorities to evacuate this zone in the near future.

For instance. Volcano Merapi has about 30,000 inhabitants in the Permanently Off-Limit

Zone and a total of 200,000 inhabitants within the three dangerous zones.

Wami.ngs issued by volcano observatories are sent to the inhabitants through the local

governments. Warnings and information on the actual state of Volcano Merapi are usually

issued by the Merapi Central Observatory. Additional information is also obtained by observatories. However, in case there is an occurrence of an extremely dangerous phenomena,

such as nuee ardente and volcanic mud-flow, each observatory warns the public directly by

using an alarm-bell or siren. Warnings regarding the occurrence of the so-called rainfall

lahar are issued depending on the amount of precipitation (usually over 60 mm/hour)

At Volcano Kelud, the famous tunnels made in 1928 by the Dutch Government as a preventive measure against overflow of the crater-lake water were devastated by mud-flows in

1951 and 1966. They were rehabilitated by the Ministry of Pxjblic Works of the Indonesian

Government during 1966-68. The Ministry of Public Words has also constructed dams to prevent a mud-flow disaster at some volcanoes, including Mt. Kelud. The Project Team for

Volcanic Debris Control of the (GSI) is cooperating closely with the Ministry of Public

Works in the above-mentioned works.

FINDINGS AND CONCLUSIONS

All the members of the Preliminary Survey Mission were deeply impressed with the persistent efforts of the Government of Indonesia to prevent various kinds of volcanic disasters. They also admire the adequate and rational orgainzation of volcanological work at

the Geological Survey of Indonesia.

However, the Preliminary Survey Mission recognizes the following serious problems:

1) Deficiency of experts in volcanology

The (GSI) has many volcanological observers throughout the country and

has many junior volcanologists , however, there are very few senior volcanologists. Indonesia has a few well- trained specialists in the following

fields:

(a) Experts in instrumentation on seismographs.

(b) Experts in analyses of the results of volcanological observations.

(c) Experts in petrology with special reference to lahar (volcanic

mud- flow)

2) Shortage of up-to-date volcanological instruments.

In order to elucidate the mechanism of underground volcanic activities, various kinds

of instrumental observations should be performed. However, at present, even the

Vl-5
seismological observation, which is indispensable in the surveillance of volcanic activities, there are only a few in operation at several active volcanoes. These seismographs

are mostly mosaics of parts produced by various companies in various countries in different

years. Therefore, it is impossible not only to determine the parameters or characters of

volcanic earthquakes, but also to detect most of the minor shocks.

I Due to the above-mentioned findings, the Preliminary Survey Mission concludes that

the following cooperation between Japan and Indonesia should be carried out as follows:

1) Dispatch the following three Japanese experts to Indonesia for several years:

a. Instrumental seismologist

b. Volcano physicist - specialist in analyses of the results of volcanological

observation

.

c. Volcanological petrologist - specialist in petrology with special reference

to nuee ardante and lahar.

2) Conduct training of these junior volcanologists of the (GSI) at the International

Institute of Seismology and Earthquake Engineering (USEE) in Tokyo.

3) Donate the following volcanological instruments to Indonesia:

| a. Seismographs for permanent and temporary observations. ''

As the situation in volcano observatories of the (GSI) is lacking in com- mercial electric supply, mechanical recording seismographs are recommended.

The seismographs should be provided with spare parts. ,i h. Instruments for petrological laboratory. '

All of the members of the Preliminary Survey Mission will make every effort to realize

the above-mentioned cooperation between Japan and Indonesia in the near future. Item 2 is presently in progress.

VI-6
Table 1. List of the Active Volcanoes of Indonesia including

Solfatara Fields (After the Geological Svirvey of

Indonesia

)

iianic O J. VUJLt-alHJ Altitude fa) Class IN 0. Name of Vol ca,rio

Altitude W Cl ass

1 PULL WEH 34 1699 +

2 S ILAWAIH AGAM 1726 0 35 CAT A V id fii 1 0

3 2780 • 30 0 n c ft • BUR NI GEUREUDONG 2590 + 37 VAiUnA £i4 ^4 0

5 Dim XT X rr t /"vxt^ 2624 •

-)ft TJ" WAA J

T AiN AMPU WTMHTT VVXiNiJ U 0 1 ft 1

£i 10 1 0

cD UAJ UJ_.r/b 1 bi\ 1500 + 39 T A ATPW T TR A M UD AT-IT T i Ai\ IjJvUdAIN ±-^KAriU £iUO 4 9

7 SIBAJAK 2212 0 4U r'AFAIN "D A "D A "NT JJ0AA J

T AJN A M 0 A c

oO SINABUNG 246o 0 4 1 V A 1.T A U V< A XTT TV 1950

9 PUSUK BUKIT 1901 0 4

TT A 1.T A LI IT A \Af\ V\ TAMP iVAVvAri iVAJrlUUJ AJNia l640-

1730 + 10 ¥ ¥ T A rr\/Xn • a T~> T T'TrV¥ TXT/^ HE LATOBA- TARUTUNG tOO- + ^3 P T TMT'T id uUiN i UK 0 0 /i 0

11 BUAL BUALI lo 19 + 44 UAJj U JN CiuU JN u 0 -1 ^^ft

id loo • 12 SORIKMARAPI 2l45 • 45 T^TT* T A/^A f\T\ K a ibLAuA dUUAo 220

1

0 13 TALAKMAL 29 12 0 4D K'A1.7"AT4 ITADATJA i\.AWArl KAKArlA 1125-

1155

14 MERAPI 2891 • 4/ i J E/KXJMA J. JU /o •

15 T» A XT 1^ TV A T* TAND IKAT O /i T ft 24 J 0 • 40 0

Q TJ_«ArUL/ A MTTT1 34 J^i Id TALANG D 0 n A 2C59D • 4V ijRTTTAW U i AIv "DC JrE/ TAT? 1 ArvAlNuAiN A AM 9 9 9 9 9

17 KERI^lTJ I 3800 • nTU i?MP ± £1' IN *a lo SUMBING 2508 0 5

1

C T TXTnot>/^ oUJNUUKU 315 1 9

19 t/" T T XT T T KUNJI

1

2151 0

7 7 1 0

20 BLERANG BERITI -1 r\ ^ Q 0 5 3

T TMP A ID A M 9n c; n 0

21 BUKIT DAUN 24D 7 0

MT?TDR A RTT 3145 0 22 KABA 1952 • 55 XjfT? A Ti X MbKAFl 11 » 23 DEMPO 3173 •

T A TaJTT 0

2k BUKIT LUMUT BALAI 2055 57 WlJLXb 9 cr 1 +

25 MARGA BAJUR 400-

1000 +

c ft 50 1/31 9

2o SEKINTJAU BELIRANG 1719 0 p9 ArCU ADT^ J

TlUJNU— TMr^ TaTTTT WlI/ij±i\AlN TTJAMPU 33 0

27 BEMATANG BATA

ca.lOOO DU C t? "MT7> D T T JO /o 9

28 HULUBELU io4o + A -1 D 1 dKUMU 9

2Q RADJABASA 1281 0 62 LAMONGAN 1669 9

30 KRAKATAU 813 • 63 IJ ANG- ARGAPURA 3088 0

31 PULOSARI 1346 0 64 RAUNG 3332 9

32 KARANG 1.778 0 65 KAWAH IDJEN 2386 0

33 KIARABERES GAGAK 15 1

1

66 BATUR 1717 9

Classification of Active Volcanoes:

9 Volcanoes with recorded eruptions ( 16OO A. D . -pre sent

)

o Volcanoes in fumarolic stage, no eruptions known

(1600 A. D. -present)

+ Solfatara or fumarole fields

VI -7
Mo. Name of Volcano Altitude kn) Cl a s s Ino . Name of Volcano Altitude bi) Class

67 AGUNG 3142 e 99 NILA 781 c 68 RINDJANI 3726 9 100 SERUA 641 • 69 TAMBORA 2851 • 101 MANUK 282 0

70 SANGEANG API 1949 e 102 BANDA API 658 • 71 WAI SANO 903 + 103 UNA UNA 508 e 72 POTJ OK LEOK 1675 + 104 AMBANG 1795 0

73 INERI 2245 0 105 SO PUTAN 1784 « 74 INIE LIKA 1559 e 106 SEMPU 1549 0 75 AMBUROMBU 2124 « 107 BATU KOLOK 890 + f^5l'S, 12 f 39 'E) 371 108 TEMPANG 900 + 76 IJA 637 • 109 TAMPUSU 1180 + 77 SUKARIA 1500 + 110 LAHENDONG + 78 NDETU NAPU 750 + 111 SARANGSONG / DU — 770 + 79 KELT MUTU i64o • 112 LOKON-EMPUNG 1580 • 80 PALUWEH 875 e 113 MAHAWU 1331 • 81 EGON 1703 0 1 14 KLABAT 1995 0

82 ILI MUDA 1100 0 115 TONGKOKO 1149 • 83 LEWOTOBI LAKILAKI 1584 e 1 16 RUANG 714 0 84 LEWOTOBI PERAMPUAN 1703 9 117 API SIAU 1748 • 85 LEROBOLENG 1117 e 118 BANTJA WUHU 12 9 86 RIANG KOTANG 200 + 119 AWU 1320 9 87 ILI BOLENG 1659 120 SUBMARINE VOLCANO 9 88 LEWOTOLD 1319 » 121 DUKON 1087 9 89 ILI LABALEKAN i486 0 122 MALUPANG WARIRANG 1115 9 90 ILI WERUNG 1018 • 123 IBU 1340 9 91 BATU TARA 748 9 124 GAMKONORA 1635 9

92 S IRUNG 862 • 125 TODOKO 979 0

93 YERSEY 0 126 PEAK OF TERNATE 1715 9 94 EMPEROR OF CHINA - 2850 « MOTIR

(0* 27 'N, 127' 24 'E) 690 *?

95 NIEUWERKERK -2285 127 MAKI.aiN 1357 9

96 GUNUNK API 282 • 128 UMSINI 2665 9 97 DAMAR 868 0 98 TEON 655 •

VI -8
Table 2. 20 Most Disastrous Eruptions in the World

from the View-Point of the Number of Victims

VOLCA^iO (Location) Year

Round Niimber of

Vi c tims

Remarks

(1) TAMBORA

( Indonesia

)

l8l5 92 , 000 Including starvation (82,000).

(2) KRAKATAU

( Indonesia

)

l888 36 , 000 Mostly by eruption- tsunami

.

(3) PELEE

( Mc^ r "t i n i c\\i o ) 1902 28 , 000 Nuee ardente attacked

a city.

(k) VESUVIO

( Italy

)

1631 18 , 000 Violent explosion & lava-flows

(5) ETNA

( Italy

)

1169 15 , 000 Lava-flows destroyed cities.

(6) UNZEN-DAKE

( Japan

)

1792 15 ,000 Mostly by eruption- tsunami

.

(7) KELUD

( Indonesia

)

1586 10 , 000 Mostly by mud-flows (lahar)

(8) ETNA

( Italy

)

1669 10,000 Lava-flows destroyed cities.

(9) LAKI

( Iceland

)

1783 10,000 Lava- flows. Including

starvation.

(10) MERAPI (Java,

Indone s ia

)

1006 Several

thousands

Mostly by pyroclastic

material s

.

(11) KELUD

( Indonesia

)

1919 5,000 I'ostly by mud-flows (lahar).

(12) GALUNGGUNG' 1822 k , 000 Mostly by mud-flows (lahar).

(13) AWU 1711 3 , 200 Mostly by mud-flows (lahar).

(I'i) LAMINGTON

( Papu.a ) 1951 3 ,000 Mostly by nuee ardente.

(15) MERAPI (Java,

Indonesia

)

1672 3 , 000 Mostly by nuee ardente.

(l6) PAPANDAJ AN

( Indone s ia

)

1772 3,000 By pyroclastic materials.

(17) AWU

( Indone sia

)

1856 2, 800 Mostly by mud- flows (lahar).

(l8) AGUNG

( Indone sia

)

1963 2,000 Nuee ardente & mud- flows

(lahar)

.

(19) SOUFRIERE

(St. Vincent Is.)

1902 1,600 Nuee ardente & mud- flows

(lahar)

.

(20) AWU

( Indonesia

)

1892 1,500 Mud- flows (lahar) & nuee ardente

.

VI-9
Table 3- List of Volcano Observatories in Indonesia

Island (No . ) Volcano

Volcano observatory

Name of Obs.

Distance

from the

r* r ^ "i" f> T" (Km)

Seismograph

Commercial

supply

Sumatra

(l4) Merapi Kotabaru 4 Mechanical (H) No

(16) Talang Batu

ber jang jang 9 Mechanical (H) No

Java

(39) Tangkuban

Prahu

Tangkuban

Prahu

0.

1

Electromagne tic ( 32

)

No

(55) Merapi

Babadan

Ngepos

PI awangan

Deles

Selo

Jrakah

k

10.5

5

5

5

(H,Z)

No

Electroma gn etic (H,Z)

Mechanical ( 2H

)

Electromagnetic (H, Z)

(H,Z)

No

No

No

115 V/

No

No

No

'OO) Dieng Karang

Tengah

1 Mechanical (2H)

El e c tro—

magnetic (Z)

No

(58) Kelud Maragomulys 6 Mechanical (H) No

(60) Semeru G. Samir

Tawonsongo

Q7

9.5

ciianicax vn/

No

No

INO No

No

(62) Lanongan C\

9 1 i" "i a

J c* No LI 65) Kawah

Id jen Paltuding 3 No No

Bali (67) Agung

Rendang

Budakeling

Batulompeh

12

l4.5

i*iecnanicax ^.^iri;/ Mechanical (2H)

No

iNO No

No

Sangeang (70) Sangeang

Api

Tawali 9. 25 No No

Flore s (76) Ija Ende 5 Mechanical (2H) No

Banda (102) Banda

Api

Colombo 2.5 Mechanical (Z) No

Sulawe si (113) Mahawu Tomohon 3 Mechanical (2H) No

Sangir (1 1 9) Awu Bawendego 7 Mechanical (2H) No

Makian (127) Makian Ngofakiaha 3.5 Electromagnetic (H)

No

* Temporary observation. The numbers of the volcanoes in

this table are referred to those used in Fig.l & Table 1.

VI -10
r-ia) •H

(0aOC £3

•rt •H

•o

m 3 V

rH c o 0

•H cH a

rlM c

0)c0o >

In0 H

(3

mU

0) •H 0 M

fi 0H O 0 H O 0 CJ >0>H

-P

u

O o

<:0 < 0

•H

<0 H

a2

(0 Eh

•H•0

<U

01

4>

(0o

J3

+>o

+>

"0uu

0) <H

0)u

0)

Sh

cS

0.

(0

•H

0)

4)0«OH0>A

(->

(hO

01Uo

VI -11
MAPPING

DIVISION

VOLCANOLOGICAL

OBSERVATION

SECTION

GEOLOGICAL SURVEY

OF INDONESIA

DIRECTOR

EDITORIAL

BOARD

EXPLORATION

DIVISION

4

SECRETARIAT

ENGINEERING & GEOHYDOLOGY

DIVISION

1

LABORATORY & DOCUMENTATION

DIVISION

VOLCANOLOGICAL DIVISION

VOLCANOLOGICAL

RESEARCH

SECTION

MERAPI CENTRAL OBS.

& 25 OBSERVATORIES

AT l6 VOLCANOES

'PROJECTS

:

1 . GEOTHERMAD

SURVEY

2. VOLCANIC

DEBRIS

CONTROL I 1 Fig.

2

ORGANIZATION CHART OF THE GEOLOGICAL

SURVEY OF INDONESIA WITH SPECIAL

REFERENCE TO THE VOLCANOLOGICAL DIVISION.

Fig. 3 Geological Survey of Indonesia at Bandung,

VI -12
Fig.^. Members of the Vol canologi cal Observation Section,

Volcanologi cal Division, Geological Survey of

Indonesia. The left end: Mr. D. Hadikusumo , Head

of the Division. The center in the front row:

A. SUWA , Chief of the Preliminary Survey Mission,

Author of this Report.

Fig. 5.

in order to divert mud-flow along the Badak river

for the purpose of protecting Blitar City, about

23 Km SW of the summit crater-lake, from the future

mud- flows

.

VI-13
Fig. 6 -

0 1 2 Skn

Fig. 6 - b

Fig. 6 Maps showing 3 kinds of dangerous zones at representative active volcanoes in Indonesia. The original maps were made by the Geological Survey of Indonesia, but

they are reduced & simplified here.

a Volcano Merapi , b. Volcano Kelud

c Volcano Semeru, d. Volcano Agung

VI -lU
VI -15
USE OF STABILIZED ADOBE BLOCK AND CANE IN CONSTRUCTION

OF LOW-COST HOUSING IN PERU

by

S. G. Fattal

Structures Section

Center for Building Technology, lAT

National Bureau of Standards

Washington, D.C. 20234

A description of the use of adobe block and cane for construction of low-

-t housing in seismic areas of Peru is described.

Key Words: Adobe; cane; earthquake; housing; Peru; technical aid.

VI-16
INTRODUCTION

The May 1970 Peruvian earthquake affected an area of 83,000 square kilometers (7% of

Peru's land area) and an estimated population of 1,700,000 (13% of the country's total population). Casualty estimates were 50,000 persons killed and 100,000 persons injured, and

loss of material resources was over $500 million. Traditionally, adobe has been, and con- tinues to be, widely used in Peru for housing and other construction. Since a significant

number of the structures in the area affected by the earthquake were built of adobe, most of

them were destroyed or severely damaged.

In recognition of the need to provide adequate protection against natural disasters,

a cooperative program was initiated between financial and technical institutions of Peru and

the United States. The primary objective was to develop better quality shelters for lowincome occupancy through the use of improved seismic design procedures and construction

materials indigenous to Peru. The participants in the cooperative program were:

1) The United States Agency for International Development Mission in Lima, Peru (AID)

,

2) The Ministry of Housing and the Housing Bank of Peru (MOH,HBP)

,

3) National University of Engineering in Lima, Peru (NUE)

,

4) National Bureau of Standards, Washington, D.C. (NBS) and,

5) The International Institute of Housing Construction, California State University,

Fresno, California (IIHT)

.

The program was financially supported by AID and the Peruvian institutions and was implemented in two stages. Stage One consisted of research and testing activities for the development of adequate soil stabilization methods, and establishment of acceptable structural

standards for improved seismic-resistant construction. Stage Two consisted of designing

and constructing two pilot housing projects, using the improved adobe materials and techniques developed in Stage One, in two selected communities which were affected by the May 1970

Peruvian earthquake

.

The two pilot housing projects (one in the coastal town of Nepena, the other in the

town of Huaraz located in the highlands) included on-site production of oil-stablized adobe

block and the construction of prototype housing units. A major emphasis was placed on training the community in self-help construction methods.

EXPERIMENTAL PROGRAM

All program-related testing was conducted by NUE staff using locally available con- struction materials and testing specifications prepared by NBS, with IIHT supplying technical documentation, instructional material, and supervisory assistance for the preparation of

oil-stablized adobe block masonry.

The experimental work consisted of testing small specimens of masonry in axial compression, diagonal compression, and flexure. Two types of cane (Carrizo and Cana Brava) , which were tested in tension, v;ere split longitudinally and used as reinforcement in the stablized

block specimens. Load-deformation measurements were obtained by means of dial gages attached

to the specimens. The program also included full-scale racking tests on five large-scale

walls, direct shear tests on small prisms, pullout tests on cane embedded in cylindrical

blocks and miscellaneous other exploratory tests. The NUE tests results were utilized to evaluate strength and stiffness properties for use in the design of the prototype housing

units. A summary of conclusions follows:

1) The stablized block specimens developed an average compressive strength of

13 kg/cm , an average shear strength at zero axial load, of 1 to 2 kg/cm , depending on the type of mortar used, and an initial modulus of elasticity

VI-17
2 in compression of 2000 kg/cm which is about 150 times the compressive strength.

The tensile strength of unreinforced block was insignificant. These results indicate that oil stablization of adobe block does not appreciably increase or de- crease its strength properties. With these low values, it was concluded that

adobe construction should be limited to single-story units.

(2) Block stablization with rapid-curing road oil (RC-250) produced superior qualities with regard to waterproofing, abrasion resistance, durability and thermal

insulation. The proportion of stabilizer was between one and two percent depending

on the quality of soil used.

(3) The two types of c^ne tested in direct tension developed a tensile capacity of

^ 1000 to 1400 kg/cm and a constant modulus of elasticity of about 200 x 10 kg/cm

throughout the entire loading range. This was about 150 times its strength and

100 times the initial modulus of elasticity of the block specimens. Tests also

indicated that it is difficult or impossible for cane to develop this high tensile

capacity in bond with adobe mortar although a significant improvement in bond was effected by using split cane in the pullout tests. Partial loss of bond and local

cracking of mortar which was frequently observed was attributed to the expansion

of cane by moisture absorption from the surrounding mortar. Stabilization of cane by coating it with a layer of asphalt solution eliminated mortar cracking but it

also tended to reduce its bond strength. The effectiveness of split cane as a reinforcing material in stablized wall construction is predicated on its adequate

protection against prolonged exposure to adverse environmental conditions such as moisture or insect infestation, and its ability to retain initial bond characteristics.

Research and testing covered many other aspects of housing construction, including

foundations, roofing systems, floor surfaces, paints, and other finishes. For example, the

newly-developed stablized soil was tested for use as a roofing cover combined with cane, to provide an economic, light, durable, and easily reparable roof cover, offering a high degree

of thermal insulation; tiles made with stablized soil, colored or plain, offer an effective

and economic solution for floor surfacing; use of stablized soil with gravel in foundations

provides adequate protection against exposure to adverse environmental conditions.

DESIGN AND CONSTRUCTION OF PROTOTYPE HOUSING UNITS

The prototype houses were designed and detailed in accordance with well-documented principles of improved seismic construction practice and on the basis of the experimental knowledge acquired in the first stage of the program.

The basic constituents used in the construction of these housing units were stablized

soil, stablized block, soil-cement-asphalt mortar, cane (split or whole) , and wood. A typical unit consists of a four-room single-story box type structure (fig. 1) , with loadbearing walls and partitions of cane-reinforced stablized block masonry construction, and a one-way wood joist roofing system covered with a mat of closely spaced whole cane, laid across the joists, and topped with a 2- to 3-cm layer of stablized soil mixture (fig. 2).

Corrugated asbestos cement roof cover is used as a substitute material for cane in Jregions

where cane is scarce. A pair of cross-braces are fastened to the roof joists to help distribute lateral seismic forces through diaphragm action. The floor ar^a is approximately 65 m . To help partial deferment of initial costs, a two-room unit of 32 m floor area may be built

initially and expanded to full-size at a later date.

The fabrication and curing procedures for the stablized block were in accordance with

recommendations developed and documented by IIHT for this program. The blocks were produced

by pouring stablized soil into modular wooden forms on flat ground. The forms were then

removed and the units were sun-dried for a period of thirty days. Circular cores in the

units permitted construction of a single-wythe wall in running bond with the holes aligned

to accommodate vertical split-cane reinforcement in both faces and at equal intervals (fig. 3

VI-18
The size of the block, which was made 38 cm square and 9 cm thick, was governed by the compressive strength requirements of a one-story high wall.

The mortar in wall construction was one of three- types used in the tests: M-1, M-2, and

M-3. The proportioning of all mortar constituents were made by weight. M-1 mortar consists

of soil mixed with two percent Rapid Curing Road Oil, RC-250, and is identified by the symbol

S-2. M-2 mortar consists of one part of Portland Cement, ten parts of soil, and one percent

RC-250, and is identified by the symbol 1:10-1. M-3 mortar consists of one part cement, six

parts soil, four parts sand, and one percent RC-250 and is identified by the symbol 1:(6, 4)-l.

The cane used was one of two types: Carrizo had a hollow core and a hard shell; Cana

Brava had a solid fibrous core. Both canes were typically 2 to 3 cm in diameter and 3 mm in

shell thickness. They were split longitudinally and were used to reinforce the walls both

horizontally and vertically, and to provide connection between the various elements of the

system (f ig . 3)

.

The shape of the housing units, the arrangement of walls, partitions, and openings con- formed as much as possible to a balanced layout configuration in order to optimize the uniformity of seismic stress distribution (fig. 1). A rectangular plan, with as close a doublysymmetric configuration as possible, was adopted. Symmetric enclosures of horizontally curved

walls used singley or in combination with flat walls were also investigated.

Interior partitions were designed as structural loadbearing walls. This practice helped

decrease the roof weight by reducing its span length, and increased the total lateral shear

resistance of the system.

The size and spacing of openings were governed by requirements of minimum width of piers

flanking adjacent openings or an opening and end of wall, maximum permissible length of lintels

spanning across the top of opening, and uniformity of overall layout (fig. 1)

.

Footings were made continuous under the walls and vertical dowel reinforcement was provided to develop partial continuity with the walls (fig. 4) . These dowels were lapped with

vertical wall reinforcement for a minimum length of 50 cm. The foundation was cast using

specified proportions of cement, stablized soil, gravel, and rock. The width of footing was

set at 1.5 times the wall thickness. An 80 cm depth was used, with 20 cm projecting above the

ground level.

To the extent that was practical, the sizes of door and window openings were kept to a minimum and provisions were made to equalize the spaces between openings or an opening and the

vertical end of wall. Nevertheless, because critical stress conditions under later loads could

develop around these openings, it was important to use special reinforcement detailing provisions in these regions. Specially cored half and whole block units were used at both vertical

edges of the openings to accommodate vertical cane reinforcement at those locations. The cane was carried through the openings of a ladder-type double joisted lintel and this space was filled with stablized soil mixture. Horizontal cane reinforcement was also provided in the

bed joint directly below the sill and carried through beyond the openings on both sides.

Consideration was given to critical stress conditions that could develop at vertical wallto-wall joints, especially near their top junction where tensile cracking usually begins, and

then propagates downward. Special detailing provisions were made to obtain continuity between

abutting walls. Such junctions were marked by projections beyond exterior wall surfaces (fig. 3) This was done primarily for two reasons: (a) to give sufficient anchorage length to horizontal

cane reinforcement by extending it beyond the common joint area, and (b) to provide for future

wall extension. Half block units in alternate courses may be pried loose and removed to permit

running bond construction of the wall extension (fig. 3)

.

The most important seismic consideration in roof design was to provide sufficient inplane stiffness so that it can adequately transmit lateral forces to the appropriate shear

walls through diaphragm action. The roofing scheme (fig. 2) used for the prototype housing

units consisted of equally spaced wood joists of rectangular cross-section spanning in one

VI-19
direction, overlain crosswise with a closely spaced mat of cane mechanically attached to the

joists below, and topped with a 2 to 3 cm stablized soil mixture. Diaphragm rigidity was provided by wood cross-braces spanning in the two diagonal directions and were mechanically

fastened to the wood joists at all intersecting points. Joists were also provided along the

top of the walls. These joists, which formed an integral part of the roof assembly, were to be attached to the walls by mechanical fasteners in order to adequately transfer the horizontal forces to the appropriate shear walls. In addition, all other intersecting wood elements were attached to each other at their functions by using mechanical fasteners (thin nails

for attachment of cane mat to joists below were used to minimize the chance of longitudinal

splitting of cane). All roof segments, such as the two segments of the doubly-pitched roof, were attached to one another so that the entire roof of the house would act as an integral

unit

.

CONCLUSIONS

Among the more important features of the model housing units are their improved capacity

relative to the traditional adobe house to withstand the action of earthquake forces, and the

reduction of cost obtained relative to the cost of conventional construction. This economic

breakthrough was achieved by the use of inexpensive construction materials readily available

in Peru (soil, cane, lumber, oil derivative), and by the adoption of self-help construction

techniques to which the new system is particularly suitable. For instance, the cost of

building materials required for the construction of a basic housing unit of 325 sq ft of

covered area was the local currency equivalent of 350 dollars, which is about one-third as much as the construction cost of conventional building materials.

Thus far, the results of this experiment have been encouraging and have received officia

Peruvian recognition. The Peruvian Housing Ministry is presently planning to start a larger

housing project in Nepena comprising approximately fifty units. In addition, new building

sites for similar projects elsewhere in Peru are presently being selected. The Peruvian

Building Code is currently being amended to include new standards and specifications for

adobe construction. This will permit insurance companies to place insurance upon improved

adobe structures and consequently enable the Peruvian Savings and Loan industry to make mortgage loans to finance such construction.

VI-20
VI-21 I
VI-22
WALL-TO-WALL CONNECTION DETAIL

HORIZONTAL SPLIT CANE

-REINFORCEMENT IN ALTERNATE COURSES

REMOVE ALTERNATE

HALF BLOCK UNITS FOR

WALL EXTENSION

TYPICAL UNIT

SHAPE IN ALTERNATE

COURSES

FIGURE 3

VI-23
REQUIREMENT FOR WALL FOOTINGS

VI-24
A COMMENT ON" THE TECHNOLOGICAL AID TO DEVELOPING COUNTRIES

by

M. Nagao

and

Tadayoshi Okubo

Structure and Bridge Division

Public Works Research Institute

This paper describes the technological aid required in developing countries,

after a natural disaster, as observed by the writer during surveys of storm and

earthquake disasters in these developing countries.

The writer classifies the aid required after a natural disaster into three

categories, that is, emergency aid, technological aid, and economic aid. The

problems related to the scientific and technological areas in the developing

countries after a natural disaster are discussed.

The technological aid is classified into short and long term aid.

Finally, the problems related to the required technological aid are discussed

and then the necessary Governmental policy which has been prepared for implementation of a low-cost and disaster resistant housing system is described.

Key Words: Earthquake; Japan; natural disaster; storm; structural engineering;

technological aid.

VI-25
INTRODUCTION

According to the recently developed hypothesis of plate tectonics, every foci of major

earthquakes are concentrated in the peripheral zone of the plate and distributed on belts

along the peripheries. Most of the countries located near the seismic belts with the ex- ception of Japan, United States, New Zealand, South-European countries, and the U.S.S.R.,

are developing coxmtries.

Also, damage caused by the world's strongest winds, such as typhoons, cyclones, and

hurricanes, occur in the developing countries.

In the past, the writer has gone abroad five times to survey damages caused by natural

disasters; in particular, a damage survey of Typhoon "Yoling" in the Philippines in 1970 as a member of UNESCO Survey Mission, the damage survey of the San Fernando Earthquake in

the United States, and Bingol Earthquake in Turkey in 1971 as a member of the Japanese

Governmental Survey Mission, and the damage survey of the Chili Earthquake in Iran and

Managua Earthquake in Nicaragua in 1973. With the exception of the San Fernando Earthquake

in the United States, all of the natural disasters - mentioned above have taken place in the

developing countries. In this paper, the writer will discuss the required technological

aid to the developing countries in relation to the natural disasters.

In general, rehabilitation projects performed in the developing countries are apt to

take a long time when these countries suffer damages. This is because of the lack of high

technological expertese, lack of repair aind rehabilitation materials, especially fundamental materials such as steel and cement, and lack of funds and capital necessary to implement the rehabilitation work. Therefore, these countries have requested aid from

foreign countries, especially after the disaster.

This kind of aid consists of emergency aid, technological aid, and economic aid.

The emergency aid that has been requested immediately after a natural disaster con- sists of donations of relief materials such as medicine, clothes, and food. Emergency

reconstruction materials, temporary houses, and water filtration apparatus are also re- quested.

The technological aid consists of dispatching of scientists and engineers to supplement the present technological level of the disaster countries, the offering of instruments

for scientific and technological observation and research work, training and education of

people in the disaster countries, offering of restoration materials and the recommendations

and cooperation on restoration planning. In addition to the above, supplies for the con- struction industries and life-line systems are included in the technological aid.

The economic aid consists of offering and loan of restoration funds and aid for re- habilitating economic and social damages caused by the disaster.

This paper will deal primarily with the technological aid as supplied to the various

countries

.

GENERAL SITUATION OF DEVELOPING COUNTRIES

The technological level in developing countries is insufficient and of low caliber.

Among the countries the writer visited, there were countries where educational lectures

on earthquake engineering and structural dynamics at universities were inadequate. Also,

lectures on wind engineering were non-existent. Similar research work in these fields of

engineering and design code and standards of wind and earthquake resistant structures to- gether with standards of testing and materials are usually poorly established in the developing countries.

In the developing coiintries, students who intend to take advanced courses are generally educated in other (developed) countries. However, domestic construction materials.

VI-26
such as steel and concrete, are meager and most of the materials are imported from the developed countries. Therefore, engineers, who are educated in a developed country where

natural disasters such as heavy storms and earthquakes take place, design structures which

are based on the wind and earthquake resistant design codes and standards and materials

and testing standards of that country and, therefore, prefer to use materials which are imported. Engineers, who are educated in a developed country where natural disasters

scarcely occur, design structures based on the design code and standards in the country

without considering the effects of natural disasters and prefer to use materials imported

from the country where they were educated.

Therefore, in developing countries it is very difficult to establish their own materials and design standards, and together with the economic scale, the existing design system

is kept.

When an earthquake disaster takes place in one of these countries, the following difficulties occur:

(1) Because seismic observation stations are not existant or sparcely located, the

location of epicenter, focal depth, and magnitude of earthquake cannot be estimated.

(2) Because there is almost no observation and research on the tectonic movement,

the cause and mechnism of the earthquake cannot be determined.

(3) There are few skilled engineers who can judge seismic intensities and estimate

isoseismals. Especially in the case of shallow earthquakes directly under a populated city, distribution of isoseismals is complicated and difficult to estimate.

(4) In most cases, there are no strong motion accelerograms and the characteristics

of ground acceleration and time duration are not clarified.

(5) Data on ground conditions is scarce and the estimations of the amplification of

the seismic motion of the ground condition, the relationship between damage,

and the ground condition is difficult.

(6) Because there are no standards for materials and construction methods and because

laborers are not skilled, the quality of the structures is not \iniform and the

estimation of structural resistivity is difficult.

(7) The cause of structural damages is not accurately analyzed.

(8) The estimation of the bearing capacity, safety of damaged structures, and the

selection of the restoring method are difficult.

(9) Lack of risk analysis of earthquake causes trouble in the establishment of

restoration planning and of earthquake resistant design standards.

(10) Materials for emergency temporary structures and resoration are difficult to obtain, therefore, the reconstruction work is difficult to initiate.

However, storms are frequently accompanied by torrential rains, therefore, the following difficulties take place:

(1) The meterorological observatories- are sparcely distributed, therefore, the path

of the tropical cyclone, maximum wind velocity, and regional distribution of

wind velocity are not known.

(2) There are few engineers skilled in the analysis of storm damage of structures,

therefore, the estimation of the actual wind loads based on the mode of damage

is difficult to determine.

VI-27
(3) Anemometers are not installed, therefore, the vertical profile and time variation

of the wind speed is not known.

(4) Because many of the rivers become wild, the damages to structures are due to both

storms and floods.

(5) There is a lack of estimation on the return of strong winds, thus causing troubles on the establishment of the restoration planning of wind resistant design standards.

In addition to the above, difficulties similar to Items (6) , (7) , (8) , and (10) , relative to earthquake hazards, also take place.

TECHNOLOGICAL AID

The technological aid that is carried out for supplementing the difficulties mentioned

in the previous section and is required for saving life and property of people in the developing coxintry will now be described.

The technological aid is classified into a short-term aid, which is carried out immediately after the disaster, and a long-term aid, which is carried out during a restoration

period and consists of supplying specialists and materials.

The first of the short-term technological aids is the dispatch of specialists consisting of;

A. Seimologist, geologists, geophysicists , soil and foundation engineers, structural

engineers (building, bridge, dam) , and members of a fire brigade in case of earthquake damage;

B. Meteorologist, climatologist, structural engineers (building, bridge) and power

transmission system, soil and foundation engineers, and river engineers in case of wind and flood damage.

These specialists are supposed to carry out the following tasks in cooperation with their

counterparts in the developing country.

1. In the field of earthquake hazards.

Observation and measurement of crustal alteration and ground rupture caused

by the earthquake.

Analysis of the mechanism of earthquakes and dynamic characteristics of ground

layers by observing aftershocks.

Estimation of the location, magnitude, and geotectonical cause of the main

shock and the estimation of the seismicity of the area in the future.

Observation of the ground accelerations of strong aftershocks by strong motion

accelerographs

.

Survey of the soil conditions in the damaged area by collecting soil data,

sampling and testing of the soil at the sites by simplified sampler, and

measuring save velocities.

Survey of the distribution of seismic intensities in the damaged area.

Evaluating the local characteristics of structural building types and materials

and determining the cause of the structural damage.

Observation of the dynamic characteristics of the damaged and undamaged structures and ground.

VI-28
Evaluation of the seismic risk in the area.

2. In the field of storm and flood hazards.

Estimation of the path, maximum wind velocity, traveling velocity of the

tropicl cyclone, and the vertical profile of the wind velocity.

Evaluating the local characteristics of the structural building types and

materials and determining the a:;tual wind loads and cause of damage.

Estimation of the distribution of rainfall, maximum flood discharge, and

other hydrological data.

Proposes a plan for river improvement and flood control.

3. In the field common to earthquake and storm hazards.

Draft a code, standards, and system for insuring structural safety against

natural disasters.

Draft guidelines on estimation of the safety of damaged structures, methods

of repair, and reconstruction work, and a principle for restoring the damaged

cities

.

Establishment of future technolgoical aids necessary for the disaster country.

In order to implement these surveys and works, it is necessary to send specialists

with the necessary instruments after the disaster; also to offer some of the instruments

which can "be left to continue effective observations and measurements.

The long-term technological aid consists of (1) recommendation and cooperation in the area of restoration planning, re-education of the engineers in the developing country and

guidance on the establishment of standards and systems for securing structural safety

against natural disasters by sending specialists for a long period, (2) acceptance and

education of trainees from the disaster country, (3) offering of contruction materials

necessary for restoration, and (4) technological cooperation on the industrial facilities

of fundamental construction materials.

PROBLEMS OF TECHNOLGOICAL AID

The technological aid from Japan to the developing countries is made possible by

Governmental grants. The problems involved in implementing such grants are as follows.

Problems in the Process of Determining Technologic Aid

(1) Generally social disorder and confusion take place in the damaged area of the

developing countries, thus sending foreign teams to survey the damage may be

rejected. Thus, before sending a survey mission, an outline of the technological aid other than the damage survey should be determined so as to permit a broader activity of mission, as outlined.

(2) The effects of the recommendations of the emergency survey mission are increased

when the dispatch of the mission is immediate and the amount of the aid is large. It is, therefore, desirable to dispatch the mission as soon as possible.

In the 1972 Managua Earthquake, survey teams from the United States and Mexico visited the city within three days after the earthquake and started the sur- vey work. In the case of our country, it takes at least five days for the clearance of passport and visa, which may result in delaying the start of the survey

work.

VI-29
I

(3) When the dispatch of the survey mission is accompanied by other technological aid,

the mission is welcomed by the disaster country. Because of the confusion in the

disaster country, immediate communication will be difficult, but acceptance of

the mission will commence.

After the earthquake, Managua was declared an open city, and as a result,

immigration clearance was omitted. There was a complaint as to why Japan hadn't

sent emergency relief teams, including medical teams, immediately after the earthquake. Also, when our mission visited General A. Somoza, President of the National

Emergency Committee, he showed his deep interest in aid from Japan.

(4) Instruments and materials necessary for the survey work should be previously prepared and carried by the survey mission.

After the Managua Earthquake, survey teams from the U.S. and Maxico carried

their instriiments to the site and carried out their observations and testing.

These instruments should be supplied by the organizations to which the specialists

belong and should be carried by them to the disaster area. It is also desirable

to send emergency relief materials, such as medicine, medical instruments, clothes,

food, and emergency temporary houses required by the victims. These emergency

relief materials should also be reserved in case of domestic disasters.

j

(5) The survey mission should be furnished with sufficient funds to pay for necessary

expenses to conduct the survey, such as purchasing data, hiring cabs, and interpreters, etc.

Problems on the Damage Survey I

(1) Even in case of domestic disasters, survey teams must bear some inconveniences

such as lack of food, potable water, and hotels in the devastated area. In the case of natural disasters in the developing countries, these inconveniences will

be doubled due to differences in living habits. The writer has had experience

in lodging in a tent and a room of an office building in the course of conducting

a damage survey in developing countries. In the damaged area, the survey team

must bear the living conditions the residents bear.

(2) The experts should understand at least one foreign language common in the damaged

area. In the developing country to find people who can understand a foreign

language, especially in the area apart from the center of administration and culture, is difficult. Without understanding the foreign language, no communication

can be carried out between the people of the victim country and the technological

aid. Because the objective of each expert differs in carrying out the survey

work, understanding of a foreign language by each expert is indispensable.

(3) Experts must be patient if necessary materials cannot be obtained. In the developing country under military administration, it happens that even a map cannot

be obtained, much less drawings of damaged structures. In such a case, the ability to communicate with the people of the victim country will cause a difference

in the results of the survey and the effectiveness of aid.

Problems of Long-Term Dispatch of Experts

A system of long-term dispatch of experts should be arranged. Usually, a long-term

dispatch of experts is requested from developing countries at the occasion of the emergency survey. However, in our country, quite a few experts want to stay longer in a foreign

country. On the other hand, there is no sabbatical leave system at the universities.

Therefore, the technological aid is sometimes interrupted, although the long-term dispatch

of experts is eagerly requested by the developing countries. Reconsideration by both

governmental organization and experts is requested.

VI-30
CONCLUDING REMARKS

Most of the technological aid to the developing countries is carried out as part of

a Governmental policy in Japan. However, natural disasters take place without warning;

therefore, a system of technological aid on natural disasters in the developing countries

must be prepared and established previously by the Government.

According to the writer's experience, damages in the developing countries are characterized by the following.

Damage of buildings, especially residential houses, is dominate; damage of

public works is not as severe.

Collapse of residential houses causes a great loss of life.

Most buildings are built of local materials and by local methods of construction, and therefore, lack the resistivity against disasters and are apt to collapse.

Residential house construction is hard to change because of ample labor power and

retention of their traditions. These houses are constructed of materials easy to obtain,

are easily built and have good insulation capacity, but possess low resistivity against

natural disasters. Therefore, in these countries, the development of low-cost houses of

ample resistivity and heat insulation capacity is needed. A study and development of such

a low-cost house is also necessary in Japan.

VI-31 ftU.S. GOVERNMENT PRINTING OFFICE:!''''^ 210-801/200 1-3
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BIBLIOGRAPHIC DATA

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!• KUnLICA 1 lUIM UK KhKUK! NU. NBS SP-4442. Gov't Accession

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J. Kecipient s Accession No.

4. TITLE AND SUBTITLt

The Proceedings of the Sixth Joint Meeting of the U.S. - Japan Panel on Wind and Seismic Effects

5. Publication Date April 1976

6. Performing Organization Code

7. AUTHOR(S)

H. S. Lew, Editor

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NATIONAL BUREAU OF STANDARDS

DEPARTMENT OF COMMERCE

WASHINGTON, DX. 20234

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Library of Congress Catalog Card Number: 76-608055

16. ABSTRACT (A 200-word or leas (actual summary of moat sigpHicant miormation. I( document includes a significant

bibliography or literature survey, mention it here.)

The Sixth Joint Meeting of the U.S. - Japan Panel on Wind and Seismic

Effects was held in Washington, D.C, on May 15-17, 1974. The proceedings

of the Hoint Meeting include the opening remarks, the program, the formal

resolutions, and the technical papers. The svibject matter covered in the

papers includes extreme winds in structural design; assessment and experimental techniques for measuring wind loads i dynconics of soil structures

and ground response in earthquakes; structural response to wind and earthquakes and design criteria; disaster mitigation against natural hazards;

and technological assistance to developing countries.

17. KEY WORDS (six to twelve entries; alphabetical order; capitalize only Ae first letter of &ie first key word unlesa a proper name; separated by semicolons)

Bridges; buildings; codes; disaster; dynamic analysis; earthquakes; modeling;

soils; structural response; volcanoes; and wind.

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