NIST 41

PUBLICATIONS

Wind and Seismic Effects

Proceedings of the 21st Joint Meeting

NIST SP 776

NAT L INST. OF STAND S TECH R.I.C.

411103 373MbD

—QC

100

*ARTMENT OF COMMERCE

Institute of Standards, and Technology

J

.U57

#776

1990

C2
NATIONAL INSTITUTE OF STANDARDS &

TECHNOLOGY

Research Information Center

Gaithersburg, MD 20899

DATE DUE

Demco, Inc. 38-293
Wind and ^i

Seismic

Effects

NIST SP 776

PROCEEDINGS OF

THE 21ST JOINT

MEETING OF

THE U.S.-JAPAN

COOPERATIVE PROGRAM

IN NATURAL RESOURCES

PANEL ON WIND AND

SEISMIC EFFECTS

Issued January 1990

Noel J. Raufaste,

EDITOR

Center for Building Technology

National Engineering Laboratory

National Institute of Standards and Technology

Gaithersburg, MD 20899

U.S. DEPARTMENT OF COMMERCE

Robert A. Mosbacher, Secretary

National Institute of Standards and Technology

Raymond G. Kammer, Acting Director
j

I

I

I

I

Library of Congress Catalog Card Number: 89-600788

National Institute of Standards and Technology Special Publication 776

Natl. Inst. Stand. Technol. Spec. Publ. 776, 413 pages (Jan. 1990)

CODEN: NSPUE2

U.S. GOVERNMENT PRINTING OFFICE

WASHINGTON: 1990

For sale by the Superintendent of Documents, U.S. Government Printing Office, Washington, DC 20402-9325
PREFACE

The UNITED STATES and JAPAN, in 1961, created the U.S. -JAPAN

Cooperative Science Program. The U.S. -JAPAN Natural Resources

Development Program (UJNR) , one of three collateral programs

making up the Cooperative Science Program, was created in

January 1964. The objective of UJNR is to exchange information

on research results and exchange scientists and engineers

in the area of natural resources. The UJNR is composed of

17 Panels each focusing on specific technical subjects.

The Panel on Wind and Seismic Effects was established in

1968 to provide technologies for reducing damages from high

winds, earthquakes, storm surge, and tsunami. The Panel

provides for cooperative activities of 16 U.S. and six

Japanese agencies with participating representatives of

private sector organizations. Annual meetings alternate

between JAPAN and the U.S. (odd numbered years in JAPAN; even

numbered years in the U.S.). The Panel provides the vehicle

to exchange technical data and information on design and

construction of buildings, bridges, dams, and lifeline and

waterfront structures, and to exchange high wind and

seismic measurement records. Results from this Panel's

efforts have impacted engineering practices and improved

codes and standards in both countries.

The 21st Joint Meeting was held at the Public Works

Research Institute, Tsukuba, JAPAN, during May 16-19, 1989.

The Panel featured six themes: 26 of the 33 authored

technical papers were presented plus two oral presentations

each on the 1988 Armenia earthquake and two reports on task

committee workshops. The Panel's nine task committees held

their meetings during this period to review joint research

projects and planned for future activities.

These proceedings include the program of the 21st Joint

Meeting, Panel resolutions, technical papers, and task

committee reports.

We are grateful for the partial support for the U.S. Panel

from the National Science Foundation, Department of State,

Department of Army, Federal Highway Administration, Bureau

of Reclamation, Nuclear Regulatory Commission, Federal

Emergency Management Agency, and the National Institute of

Standards and Technology.

Noel J. Raufaste, Secretary

US-Side, Panel on Wind and Seismic Effects

• • • in
The 21st Joint Meeting of the U.S. -Japan Panel on Wind and

Seismic Effects was held at the Public Works Research

Institute, Tsukuba, JAPAN, from May 16-19, 1989. This

publication, the proceedings of the Joint Meeting, includes

the program, list of members, panel resolutions, task

committee reports, and 33 technical papers.

The papers were presented under six themes: (I) - Wind

Engineering, (II) - Earthquake Engineering, (III) - Storm

Surge and Tsunami, (IV) •- U.S. -Japan Cooperative Research

Program, (V) - Code Development Process and Code Enforcement

Responsibility, and (VI) - Armenia Earthquake (oral

presentation)

KEYWORDS: accelerograph ; Armenia; bridges; codes; concrete;

design criteria; disaster; earthquakes; geotechnical

engineering; ground failures; inelastic; lifelines;

liquefaction; masonry; repair and retrofit; risk assessment;

seismicity; soils; standards; storm surge; structural

engineering; tsunami; and wind loads.

iv
CONTENTS

PAGE

PREFACE iii

ABSTRACT iv

AGENDA FOR 21st JOINT UJNR MEETING ix

LIST OF PANEL MEMBERS xiv

LIST OF TASK COMMITTEE MEMBERS xxi

RESOLUTIONS xxiii

THEME I: Wind Engineering

A Summary of the Federal Highway Administration's

Wind Engineering Research Program 3

Harold R. Bosch

Wind Resistant Design Manual for

Highway Bridges in Japan 13

Nobuyuki Narita, Koichi Yokoyama, and Hiroshi Sato

Field Measurements of Wind Pressures 20

Kishor C. Mehta

Transfer of Research Results in

Wind Engineering 26

J. Eleonora Sabadell

Wind Resistance in Manufactured Housing in

Hurricane Zones 29

G. Robert Fuller

THEME II: Earthquake Engineering

Behavior of the Nagara Dam During the

Earthquake of December 17, 1987 43

Yasushi Sasaki, Tetsuro Kuwabara, Masao Nishimura, Ikuji Seko,

Jun Kashiwagi, and Yoshiki Aoki

Analysis of Structural Response Records During the Whittier Narrows,

California, Earthquake, October 1, 1987 55

A. Gerald Brady

v
CONTENTS

Behaviours of Dams

During Chiba off Earthquake

Norihisa Matsumoto and Nario Yasuda

Surrey for Estimation of Seismic Damage in

South Kanto Area

Hideaki Oda and Shigeru Tamura

Seismic Design Considerations for Offshore

Oil and Gas Structures

Charles E. Smith

Centrifuge Dynamic Model Tests in Phri

Takamasa Inatomi, Motoki Kazama, Setsuo Noda, and Hajime Tsuchida

Modeling of Earthquake Induced

Pore Pressure Behavior

Richard H. Ledbetter

Seismic Design Method of Earth Embankment With

Seismic Coefficient

YasuyuM Koga and Osamu Matsuo

Seismic Monitoring of Buildings: Analyses of Seismic Data

M. Celebi

Development of Advanced Reinforced Concrete Buildings

Using High-Strength Concrete and Reinforcement

Tatsuo Murota and Masaya Hirosawa

Guidelines for Identification and Mitigation of Seismically

Hazardous Existing Federal Buildings

H. S. Lew

U.S. Army Corps of Engineers

Seismic Research Program

William E. Roper

The History and Status of Strong-Motion

Earthquake in Japan

Keiichi Ohtani and Hirosbi Takahashi

Summary of Lifeline Earthquake Engineering

James D. Cooper

PAGE

61

73

87

96

109

121

134

150

155

163

172

188

vi
CONTENTS

PAGE

A Strong Motion Data Catalog for

Personal Computers 192

Lee Wesley Row, III and Herbert Meyers

Nonlinear Seismic Analysis of

Concrete Gravity Dams 199

Lucian G, Guthrie

Seismic Mlcrozonation 208

Walter W. Hays

Full Scale Test on a Three-Storied

Wood-Framed Building 213

Tatsuo Murota and Motoi Yasumura

Damage by the Chlba-ken Toho-Oki Earthquake 227

Yasushi Sasaki

THEME IIL Storm Surge and Tsunami

Observations of Tsunamis on the West Coast of the

United States 247

James F. Lander

Present State of Tsunami Numerical Simulation 255

Chiaki Goto and Katsutosbi Tanimoto

THEME IV: U.S.-japan Cooperative Research Program

U.S.-Japan Coordinated Earthquake Research Program on Masonry Buildings -Seismic Test of Five Story Full Scale

Reinforced Masonry Building, Part 2 265

Shin Okamoto, Yutaka Yamazaki,

Takashi Kaminosono, and Masaomi Teshigawara

Influence of Horizontal Reinforcement on the

Shear Resistance of Masonry Walls 286

Charles W. C. Yancey

Inelastic Behavior of Reinforced Concrete Bridge Pier— Effect of

Bilateral Loading and Circular Hollow Section 298

Toshio Iwaskai, Osamu Ueda, Kazuhiko Kawashima,

Kinji Hasegawa, and Takeshi Yoshida

U.S. Coordinated Program for Masonry Building

Research— A Fourth Year Status 310

James L. Noland

vii
CONTENTS

PAGE

THEME V: Code Development Process and Code Enforcement Responsibility

Minister-of-Construction's Agreement System in the

Building Standard Law of Japan 329

Tatsuo Murota

History of Structural Regulations in Japanese Building Codes and

Outline of Wind-Resistant and Seismic Regulations 335

Yuji Ishiyama and Yuji Ohashi

Development and Enforcement of U.S. Building Regulations 348

James G. Gross and Richard N. Wright

Appendix: Task Committees A-J Reports 367

• • • Vlll
Tuesday May 16

Opening Session 10:30-12:00

Lunch 12:00-13:00

Technical Session 13:00-16:50

* Speaker

Session 1 (Theme I WIND ENGINEERING)

Chairman :N. Narita

13:00-13:20 A Summary of Federal Highway Administration's Wind

Engineering Research Program

Harold R. Bosch*

13:20-13:40 Wind Resistant Design Manual for Highway Bridges

in Japan

N. Narita, K.Yokoyama*, H.Sato

13:40-14:00 Field Measurements of Wind Pressures

K. C. Mehta*

14:00-14:20 Discussion

Session 2 (Theme II EARTHQUAKE ENGINEERING)

Chairman :N. Narita

14:20-14:40 Behavior of the Nagara Dam during the Earthquake

of December 17, 1987

Y. Sasaki*, T. Kuwabara, M. Nishimura, I. Seko,

J. Kashiwagi , Y. Aoki

14:40-15:00 Analysis of Structural Response Records during the

Whittier Earthquake ,. 1 October 1987

A. G. Brady*

15:00-15:20 Behaviors of Dams during Chiba Off Earthquake

N. Matsumoto*, N. Yasuda

15:20-15:40 Discussion

15:40-16:00 Break

Guest Lecture (Theme II EARTHQUAKE ENGINEERING) 16:00-16:50

16:00-16:40 Survey for Estimation of Seismic Damage in South

Kanto Area

H. Oda*

16:40-17:00 Discussion

Adjourn

ix
Wednesday May 17

Technical Session 9:00-12:00

Session 3 (Theme III STORM SURGE AND TSUNAMI)

Chairman :R. N. Wright

9:00- 9:20 Observations of Tsunamis on the West Coast of the

United States

J. F. Lander*

9:20- 9:40 Present State of Tsunami Numerical Simulation

C. Goto*, K. Tanimoto

9:40-10:00 Discussion

10:00-10:20 Break

Session 4 (Theme II EARTHQUAKE ENGINEERING)

Chairman: R. N. Wright

10:20-10:40 Seismic Design Considerations for Offshore Oil

and Gas Structures

C. E. Smith*

10:40-11:00 Centrifuge Dynamic Model Tests in PHRI

T. Inatomi, M. Kazarna, S. Noda* , H. Tsuchida

11:00-11:20 Modeling of Earthquake Induced Pore Pressure

Behavior

R. H. Ledbetter, (A. G. Franklin*)

11:20-11:40 Seismic Design Method of Earth Embankment with

Seismic Coefficient

Y. Koga*, 0. Matsuo

11:40-12:00 Discussion

Lunch 12:00-13:00

Technical Session 13:00-14:00

Session 5 (Theme II EARTHQUAKE ENGINEERING)

Chairman :R. N. Wright

13:00-13:20 Seismic Monitoring of Buildings : Analysis of

Seismic Data

M. Celebi, (A. G. Brady*)

13:20-13:40 Development of New Reinforced Concrete Buildings

with High Strength and High Quality Concrete and

Steel

T. Murota*, M. Hirosawa

13:40-14:00 Guidelines for Identification and Mitigation of

Seismically Hazardous Existing Federal Buildings

H. S. Lew*

14:00-14:20 Discussion

14:20-14:30 Break

Task Committee Meetings 14:30-17:00

B : Large-Scale Testing Programs

E: Natural Hazard Assessment and Mitigation through Land Use

Programs

F: Disaster Prevention Methods for Lifeline Systems

H:Soil Behavior and Stability During Earthquakes

I: Storm Surge and Tsunamis
Thursday May 18

Technical Session 9:00-12:00

Session 6 (Theme II EARTHQUAKE ENGINEERING)

Chairman :N. Narita

9:00- 9:20 U.S. Army Corps of Engineers Seismic Research

Program

W. E. Roper*

9:20- 9:40 The History and Status of Strong-Motion Earthquak

Observation in Japan

K. Ohtani*, H. Takahashi

9:40-10:00 Summary of Lifeline Earthquake Engineering

J. D. Cooper, (H. R. Bosch*)

10:00-10:20 Discussion

10:20-10:40 Break

Session 7 (Theme V Code Development Process and Code Enforcement

Responsibility)

Chairman :R. N. Wright

10:40-11:00 Minister-of-Construction 1 s Agreement System in th

Building Standard Law of Japan

T. Murota*

11:00-11:20 History of Structural Regulations in Japanese

Building Code and Outline of Wind-Resistant and

Seismic Regulations

Y.Ishiyama, Y. Ohashi*

11:20-11:40 Development and Enforcement of US Building

Regulations

J. G. Gross, R. N. Wright*

11:40-12:00 Discussion

Lunch 12:00-13:00

Technical Session 13:00-14:20

Session 8 (Reports of Workshops)

13:00-13:30 Task Committee

Y. Sasaki*

13:30-14:00 Task Committee

H. R. Bosch*

14:00-14:20 Discussion

14:20-14:30 Break

Task Committee Meetings 14:30-17:00

A : Strong-Motion Instrumentation Arrays and Data

C:Repair and Retrofit of Existing Structures

Devaluation of Performance of Structures

J: Wind and Earthquake Engineering for Transportation Systems

Chairman :R. N. Wright

(F)

(J)

xi
Friday May 19

Technical Session 9:00-12:00

Session 9 (Theme IV U.S. -Japan Cooperative Research Program)

Chairman :R. N. Wright

9:00- 9:20 U.S. -Japan Coordinated Earthquake Research Program

on Masonry Buildings -Seismic Test of Five

Story Full Scale Reinforced Masonry Building, Part

2-

S. Okamoto, Y. Yamazaki*, T. Kaminosono,

M. Teshigawara

9:20- 9:40 Shear Strength of Masonry Wall Structural Systems

H. S. Lew*

9:40-10:00 Inelastic Behavior of Reinforced Concrete Bridge

Pier -Effect of Bilateral Loading and Circular

Hollow SectionT. Iwasaki , 0. Ueda, K. Kawashima* , K. Hasegawa,

T. Yoshida

10:00-10:20 U.S. Coordinated Program for Masonry Building

Research -A Fourth Year StatusJ. L. Noland, (G. R. Fuller*)

10:20-10:40 Discussion

10:40-11:00 Break

Session 10 (Theme VI Reports of the Armenia Earthquake)

Chairman :R. N. Wright

11:00-11:30 Reports of the Armenia Earthquake -Discussion on

Damages to Buildings-

(Oral Presentation)

M. Hirosawa*

11:30-12:00 US Experiences from Post-earthquake Fact Finding

Team Visit to Armenia during 19-28 December

(Oral Presentation)

H. S. Lew*

12:00-12:10 Discussion

Lunch 12:10-13:00

Reports of Task Committees 13:00-15:00

Chairman :N. Narita

A : Strong-motion Instrumentation Arrays and Data

B : Large-Scale Testing Programs

C:Repair and Retrofit of Existing Structures

D: Evaluation of Performance of Structures

E: Natural Hazard Assessment and Mitigation through Land Use

Programs

F:Disaster Prevention Methods for Lifeline Systems

H:Soil Behavior and Stability during Earthquakes

I: Storm Surge and Tsunamis

JrWind and Earthquake Engineering for Transportation Systems

Xll
15:00-15:20 Break

Adoption of Final Resolution 15:20-16:10

Chairman :N. Narita

16:10-16:20 Break

Closing Session 16:20-16:40

Press Conference 16:50-17:20

• • • Xlll
LIST OF PANEL MEMBERS

UNITED STATES PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

1989

Dr. Richard N. Wright

Chairman

Director, Center for Building

Technology

National Engineering Laboratory

National Institute of Standards

and Technology

Gaithersburg, MD 20899

(301) 975-5900

Mr. Noel J. Raufaste

Secretary-General

Head, Cooperative Research Programs

Structures Division

Center for Building Technology

National Engineering Laboratory

National Institute of Standards

and Technology

Gaithersburg, MD 20899

(301) 975-5905 FAX 301-948-0228

Dr. S. T. Algermissen

Research Geophysicist

Branch of Engineering and Geology

Office of Earthquakes, Volcanos

and Engineering Technology

US Geological Survey

Denver, CO 80225

(303) 236-1611 FTS 776-1611

Dr. Celso S. Barrientos

Supervisory Physical Scientist

National Environmental Satellite,

Data, and Information Service (NESDIS)

National Oceanic and Atmospheric

Administration

1825 Connecticut Ave., N.W.

Universal Bldg., Rm. 518, E/AI3

Washington, DC 20009

(202)673-5394

Dr. Eddie N. Bernard

Director, Pacific Marine

Environmental Laboratory

Environmental Research Laboratories

National Oceanic and Atmospheric

Administration

7600 San Point Way, NE

Seattle, WA 98115

(206) 526-6239 FTS 392-6800

XIV

Dr. Roger D. Borcherdt

Chief, Branch of Engineering

Seismology and Geology

Office of Earthquake Studies

US Geological Survey

Department of the Interior

345 Middlefield Road, M/S 977

Menlo Park, CA 94025

(415) 323-8111 X2755

Dr. A. Gerald Brady

Physical Scientist

Branch of Engineering Seismology

and Geology

US Geological Survey

Department of the Interior

345 Middlefield Road, M/S 977

Menlo Park, CA 94025

(415) 329-5664 Fax:415-329-5163

Mr. H. Lee Butler

Chief, Research Division

Coastal Engineering Research Center

Waterways Experiment Station

US Army Corps of Engineers

Office CEWES-CR, P.O. Box 631

Vicksburg, MS 39180-0631

(601) 634-2405

Mr. James D. Cooper

Deputy Chief, Structures Division

Office of Engineering and Highway

Operations Research and Development

Federal Highway Administration

6300 Georgetown Pike

McLean, VA 22101

(703) 285-2447

Dr. A. J. Eggenberger

Program Director

Earthquake Hazard Mitigation

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

(202) 357-9500

Dr. A. G. Franklin

Chief, Earthquake Engineering

and Geophysics Division

CEWES-GH Geotechnical Laboratory

Waterways Experiment Station

US Army Corps of Engineers

Vicksburg, MS 39180-0631

(601) 634-2658
Mr. G. Robert Fuller

Chief of Compliance Branch

Manufactured Housing and Construction

Standards Division

Office of Housing, Room 9156

US Department of Housing and Urban

Development

Washington, DC 20410-8000

(202) 755-6920

Mr. James H. Gates y

Structural Mechanics Engineer

Supervising Bridge Engineer

Office of Structures Design

Division of Structures

Department of Transportation

State of California

1120 N Street

Sacramento, CA 95807

(916) 445-1439

Mr. Peter E. Gurvin

Chief of Civil/Structural Engineering

Office of Foreign Buildings

Department of State

A/FB0, SA-6, Room 327

P.O. Box 12238

Rosslyn Station

Arlington, VA 22209

(703) 875-6117

Dr. Robert D. Hanson

Director, Biological and Critical

Systems Division

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

(202) 357-9545

Dr. Walter Hays

Deputy Chief for Research Applications

Office of Earthquake, Volcanoes and

Engineering

US Geological Survey

Department of the Interior

905 National Center

Res ton, VA 22043

(703) 648-6711

Dr. William B. Joyner

Geophysicist

Office of Earthquake Studies

Branch of Ground Motion and Faulting

US Geological Survey

Department of the Interior

345 Middlefield Road, M/S 977

Menlo Park, CA 94025

(415) 323-8111 FTS 467-2754

Mr. Roger M. Kenneally

Structural Engineer

Engineering Branch

Seismic Section

Office of Nuclear Regulatory Research

U.S. Nuclear Regulatory Commission

Washington, DC 20555

(301) 492-3893

Mr. Richard W. Kramer

Technical Review Staff

Engineering and Research Center

Bureau of Reclamation

Department of the Interior

P.O. Box 25007

Denver, CO 80225

(303) 236-8539 FTS 776-8539

Mr. Richard W. Krimm

Assistant Associate Director

Office of Natural and Technological

Hazards

Federal Emergency Management Agency

Washington, DC 20472

(202) 646-2871

Mr. Paul F. Krumpe

Physical Scientist

Office of US Foreign Disaster Assistance

Agency for International Development

Department of State

Washington, DC 20523

(202) 647-9758

Dr. H. S. Lew

Chief, Structures Division

Center for Building Technology

National Engineering Laboratory

National Institute of Standards

and Technology

Gaithersburg, MD 20899

(301) 975-6061

Mr. Tingley K. Lew

Research Structural Engineer

Structures Division

Naval Civil Engineering Laboratory

L51

Port Hueneme, CA 93043

(805) 982-5785

Dr. Shih-Chi Liu

Program Director

Earthquake Hazard Mitigation

National Science Foundation

1800 G Street, N.W., Room 1132

Washington, DC 20550

(202) 357-9780

XV
Mr. Robert B. MacDonald

Chief, Geologic Services Branch

Engineering and Research

Bureau of Reclamation

Department of the Interior

Building 67, Denver Federal Center

Denver, CO 80225

(303) 236-4195 FTS 776-4195

Dr. Richard D. Marshall

Structural Research Engineer

Earthquake Engineering Group

Structures Division

Center for Building Technology

National Institute of Standards

and Technology

Gaithersburg, MD 20899

(301) 975-6071

Dr. Richard D. McConnell

Director, Civil Engineering Service

Veterans Administration

811 Vermont Avenue, N.W.

Washington, DC 20420

(202) 233-2864

Dr. Francis G. McLean

Chief, Geotechnical Engineering and

Embankment Dams Branch

Engineering and Research Center

US Bureau of Reclamation

P.O. Box 25007

Denver, CO 80225

(303) 236-3854 FTS 776-3854

Mr. Herbert Meyers

Chief, Solid Earth Geophysics Division

National Geophysical Data Center

National Environmental Satellite,

Data, and Information Service

National Oceanic and Atmospheric

Administration

325 Broadway

Boulder, CO 80303

(303) 497-6521 FTS 320-6521

Dr. William E. Roper

Assistant Director

Research and Development (Civil Works)

US Army Corps of Engineers

20 Massachusetts Avenue, N.W.

Washington, DC 20314-1000

(202) 272-0257

Dr. Erwin L. Schaffer

Assistant Director, Wood Products Research

Forest Products Laboratory

Forest Service

US Department of Agriculture

One Gifford Pinchot Drive

Madison, WI 53705-2398

(608) 264-5672 FTS 364 5672

Mr. Charles E. Smith

Research Program Manager

Technology Assessment and Research Branch

Minerals Management Service

Department of the Interior

Mail Stop 647 National Center

Res ton, VA 22091

(703) 658-7752

Mr. Robert J. Smith

Chief, Structures Branch

Directorate of Engineering

and Construction

Office, Chief of Engineers

US Army Corps of Engineers

HQUSACE (CEEC-ED)

Washington, DC 20314-1000

(202) 272-0220

xvi
Alternates Members

Dr. Clifford J. Astill

Program Director

Division of Fundamental Research

for Critical Engineering Systems

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

(202) 357-9500

Dr. Mehmet K. Celebi

Research Civil Engineer

Branch of Engineering Seismology

and Geology

US Geological Survey

345 Middlefield Road, M/S 977

Menlo Park, CA 94025

Mr. Michael J. Changery

Chief, Applied Climatology Branch

National Climatic Data Center

Federal Building

Asheville, NC 28801

(704) 259-0765 FTS 672-0765

Ms. Janina Z. Mirski

Chief, Southern/Western Structural

Division

Veterans Affairs

811 Vermont Avenue, N.W.

Washington, DC 20420

(202) 233-2394

Mr. Ronald J. Morony

Director, Innovative Technology

and Special Projects Division

Office of Policy Development and Research

Room 8232

Department of Housing and Urban Development

Washington, DC 20410-6000

(202) 755-0640

Dr. John B. Scalzi

Program Director, Division of Fundamental

Research in Critical Engineering Systems

National Science Foundation

1800 G Street, N.W.

Washington, DC 20550

(202) 357-7710

Dr. James F. Costello

Senior Structural Engineer

Division of Engineering

Office of Nuclear Regulatory Research

US Nuclear Regulatory Commission

Mail Stop 007NL

Washington, DC 20550

(301) 443-7709

Mr. Lucian G. Guthrie

Structural Engineer

Engineering and Construction Directorate

Office, Chief of Engineers

US Army Corps of Engineers

HQUSACE (CEEC-ED)

Washington, DC 20314-1000

(202) 272-8673

Dr. James R. Houston

Chief, Coastal Engineering Research Center

Waterways Experiment Station

US Army Corps of Engineers

P.O. Box 631

Vicksburg, MS 39180-0631

(601) 634-2000

Mr. Robert R. Ledzian

Senior Staff Assistant for Research

Office of Liaison-Engineering Research

Bureau of Reclamation

Department of the Interior

18th & C Streets, N.W.

Washington, DC 20240

(202) 343-6703 _v:f
JAPAN PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

1989

Dr. Nobayuki Narita

Chairman

Director-General

Public Works Research Institute

Ministry of Construction

1 , Asahi , Tsukuba-shi

,

Ibaraki-ken 305

Tel. C298-64-2211

Dr. Toshio Iwasaki

Deputy Director-General

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Yoshio Adachi

Secretary-General

Director, Planning and Research

Administration Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Tokunosuke Fujitani

Chief, The Second Research

Laboratory

Applied Meteorology Research

Division

Meteorological Research Institute

Japan Meteorological Agency

1-1, Nagamine, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-51-7111

Dr. Kazuhiko Kawashima

Head, Earthquake Engineering

Division

Earthquake Disaster Prevention

Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Yoshikazu Kitagawa

Head, Civil Engineering Division

International Institute of

Seismology and Earthquake

Engineering

Building Research Institute

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Mr. Minoru Fujiwara

Head, Bridge Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Yasuyuki Koga

Head, Soil Dynamics Division

Construction Method and Equipment

Department

Public Works Research Institute

Ministry of Construction

1 , Asahi , Tsukuba-shi

,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Masaya Hirosawa

Director

International Institute of

Seismology and Earthquake

Engineering Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Mr. Kiyoshi Kurashige

Head, Typhoon Research Division

Meteorological Research Institute

Japan Meteorological Agency

1-1, Nagamine, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-51-7111

xviii
Dr. Norihisa Matsumoto

Head, Fill Type Dam Division

Dam Department

Public Works Research Institute

Ministry of Construction

1 , Asahi , Tsukuba-shi

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Michio Okahara

Head, Foundation Engineering

Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Tatsuo Murota

Director, Structural Engineering

Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Shin Okamoto

Director, Research Planning

and Information Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Shinsuke Nakata

Head, Structural Division

International Institute of

Seismology and Earthquake

Engineering

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Norio Oyagi

Director, Third Research Division

National Research Center for

Disaster Prevention

Science and Technology Agency

3-1, Tennodai, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-51-1611

Dr. Setsuo Noda

Director, Structures Division

Port and Harbour Research

Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka,

Kanagawa-ken 239

Tel. 0468-41-5410

Mr. Shoichi Saeki

Dii-ector, Bridge and Structure

Department

Public Works Research Institute

Ministry of Construction

1 , Asahi , Tsukuba-shi

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Keiichi Ohtani

Head, Earthquake Engineering

Laboratory

National Research Center for

Disaster Prevention

Science and Technology Agency

3-1, Tennodai, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-51-1611

* Dr. Yasushi Sasaki

Director, Earthquake Disaster

Prevention Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

xix
Dr. Masaaki Seino

Head, Seismology and Volcanology

Research Division

Meteorological Research Institute

Japan Meteorological Agency

1-1, Nagamine , Tsukuba-shi

Ibaraki-ken 305

Tel. 0298-51-7111

Dr. Hiroyuki Yamanouchi

Head, Structural Dynamic Division

Structural Engineering Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Dr. Katsutoshi Tanimoto

Director, Hydraulic Engineering

Division

Port and Harbour Research

Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka,

Kanagawa-ken 239

Tel. 0468-41-5410

Dr. Yutaka Yamazaki

Head, Large-scale Structure

Testing Division

Production Department

Building Research Institute

Ministry of Construction

1, Tatehara, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2151

Mr. Ken-ichi Tokida

Head, Ground Vibration Division

Earthquake Disaster Prevention

Department

Public Works Research Institute

Ministry of Construction

1, Asahi , Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Mr. Satosu Yanase

Director, Geographic Department

Geographical Survey Institute

Ministry of Construction

1, Kitazato, Tsukuba-shi

Ibaraki-ken 305

Tel. 0298-64-1111

Dr. Hajime Tsuchida

Director-General

Port and Harbour Research

Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi

Xanagawa-ken 239

Tel. 0468-41-5410

* Mr. Koichi Yokoyama

Head, Structure Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

Dr. Takaaki Uda

Head, Coastal Engineering

Di vision

River Department

Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi,

Ibaraki-ken 305

Tel. 0298-64-2211

* Secretary

XX
Task Committee US Side

Strong-Motion Instrumentation *A.G. Brady

Arrays and Data M.K. Celebi

A.F. Franklin

TaT d W. D. juyner

F.G. McLean

H. Meyers

Large-scale Testing Programs *H.S. Lew

A.J. Eggenberger

C.E. Smith

R.J. Smith

C. Repair and Retrofit of Existing

Structures

*A.J. Eggenberger

P.E. Gurvin

H.S. Lew

R.D. McConnell

R.J. Morony

J.B. Scalzi

D. Evaluation of Structural

Performance

*G.R.

M.K.

R.D.

R.D.

J.B.

R.J.

Fuller

Celebi

Marshall

McConnell

Scalzi

Smith

E. Natural Hazard Assessment and

Mitigation Through Land Use

Programs

*S.T. Algermissen

C.S. Barrientos

A.G. Brady

G.R. Fuller

R.D. Marshall

Japanese Side

*H. Tsuchida

S. Hattori

M. Ichikawa

Y. Ishiyama

K. Ghtani

Y. Sasaki

H. Yamanouchi

*K. Chtani

Y. Koga

S. Noda

S. Okamoto

K. Yokoyama

H. Yamanouchi

*M. Hirosawa

T. Iwasaki

S. Okamoto

H. Shinohara

*S. Okamoto

T. Hanafusa

K. Masamura

T, Murota

N. Narita

M. Okahara

H. Shinohara

H. Yamanouchi

K. Yokoyama

*K. Kawashima

S. Hattori

T. Iwasaki

M. Nagaoka

Y. Sugimura

xxi
Task Ctemmittee-continued US Side Japanese Side

F. Disaster Prevention Methods *S.C. Liu *0. Ueda

for Lifeline Systems M.K. Celebi K. Kawashima

J.D. Cooper T. Murota

Tj. • T.iii 1 1 1 LA c Noda

J.B. Scalzi K. Ohtani

J.S. Spencer Y. Sasaki

S. Wu

H. Ooil Behavior and Stability *A.G. Franklin *Y. Sasaki

During Earthquakes F.G. McLean Y. Koga

R.J. Smith fc MMU 1 VAX N. Matsumoto

M. Okahara

Y. Sugimura

H. Tsuchida

I. Storm Surge and Tsunamis *H. Meyers *T. Uda

POL J- 1C1 1 lA^O M Tr*h JL iUVUnU "i V^wt^

E.N. Bernard K. Masamura

L. Butler H. Takahashi

K. Tanimoto

J. Wind and Earthquake Engineering *J.D. Cooper *S. Saeki

for Transportation Systems A.G. Franklin M. Fujiwara

J.H. Gates T. Iwasaki

J.B. Scalzi K. Kawashima

N. Narita

M. Okahara

Y. Sasaki

K. Yokoyama

*Chairman

xxii
RESOLUTIONS OF THE TWENTY-FIRST JOINT MEETING

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS (UJNR)

Public Works Research Institute

Ministry of Construction

1 , Asahi , Tsukuba-shi , Ibaraki 305

May 16-19, 1989

The following resolutions are hereby adopted:

1. The Twenty-First Joint Panel Meeting provided an opportunity to

exchange valuable technical information which was beneficial to both

countries. In view of the importance of cooperative programs on the

subject of wind and seismic effects, the continuation of Joint Panel

Meetings is considered essential.

2. The following activities have been conducted since the Twentieth Joint

Meeting:

a. Guest researchers from both countries performed joint research that

advanced the state of wind and earthquake engineering.

b. Technical documents, research reports and proceedings of workshops

were exchanged.

c. The Japanese MOC document, "Manual for Repair Methods of Civil

Engineering Structures Damaged by Earthquakes" was translated into

English, published, and soon will be disseminated to appropriate US

engineering community members.

d. The Japanese MOC reports on Base Isolation Systems for Buildings

were discussed with NIST in March 1989, for translation into

English. They will be published and disseminated to appropriate

U.S. engineering community members.

e. Workshops and planning meetings were held:

a) Fourth Joint Technical Coordinating Committee on Masonry, Task

Committee (B), San Diego, CA, October, 1988.

b) International Workshop on Sensor Technology Applied to Large

Engineering Systems, Task Committee (D), NIST, Gaithersburg MD in

September, 1988.

c) Fifth Bridge Workshop, Task Committee (J), Tsukuba, May 9-13,

1989.

d) Workshop on Disaster Prevention for Lifeline Systems, Task

Committee (F), Tsukuba, May 11 - 13, 1989.

e) Planning Meeting on Cooperative Research for Remedial Treatment

of Liquefiable Soils, Task Committee (H), Tsukuba, May 13 - 14,

1989.

xxiii
f. Held a special Panel Theme on "Japanese and U.S. Code Development

Process and Code Enforcement Regulations" to provide a better

understanding of the history and evolution of both countries'

building regulations.

3. Experiences from post-earthquake fact finding team visits to Armenia in

December 1988 were reported by both sides. The Panel recognizes the

importance of using these findings and experiences and encourages

similar exchanges of information from future post-disaster

investigations. Information transfer will include reports, seminars,

and special Panel sessions as used at the Twenty-First Joint Meeting.

4. The Panel will continue seeking methods to contribute to the

International Decade of Natural Disaster Reduction (IDNDR) such as

exchanging Proceedings of Task Committees and Panel Meetings with their

respective National Committees' of IDNDR.

5. The Panel approves each Task Committee's Mission Statement reports

developed during the Twenty-First Joint Meeting. These reports present

Task Committee objectives, current and proposed scope of work, and past

and expected accomplishments.

6. The Panel recognizes the importance of the following items:

a. Both sides are performing work contributing to the Panel's Program

on Large-scale Testing of Precast Seismic Structural Systems

(PRESSS). Information about the program will continue to be

exchanged during the coming year.

b. Continue to exchange data and information on application of active

and passive response control systems for buildings and other

structures and encourage the appropriate Task Committees, by joint

work, to advance these technologies.

c. Collect strong motion data on the performance of buried pipeline

systems for 3-D downhole information.

d. Obtain experimental verification of the effectiveness of

retrofitting and strengthening methods for structures and soils.

e. Disseminate information to all Panel members about opportunities

for exchanging researchers to each country such as using the

Japanese STA Fellowship Program and the related NSF Fellowship

Programs

7. Joint planning will begin on cooperative work in a new research area on

verifying soil-pile interaction models for marine structures subjected

to earthquake loads. This may be planned through correspondence by

both chairmen of T/C (H) who will inform both Panel Side Secretaries

about progress.

8. The Panel endorses the following proposed Task Committee workshops and

planning meetings

xxiv
a. The Fifth Meeting of JTCCMAR, T/C (B) will be scheduled in Tsukuba,

Japan October 1989, and a US - Japan Precast Seismic Structural

Systems (PRESSS) Program Coordinating Meeting, T/C (B) will be

planned for Japan in October, 1989.

b. Task Committees (C) and (D) are planning a joint workshop on

Evaluation, Retrofit, and Strengthening Structures Damaged by High

Winds and Earthquakes in May 1990.

c. Workshop on Earthquake Hazard and Risk Assessment, T/C (E) will be

held in Japan in 1990.

d. Workshop on Disaster Prevention for Lifeline Systems, T/C (F) will

be held in the U.S. in 1990.

e. Workshop on Design and Treatment Method for Mitigating Liquefaction

Induced Damage, Task Committee (H) will be held in Japan in October

1990.

f. Workshop on Storm Surge and Tsunamis, T/C (I) will be held in

Honolulu in October or November 1990.

g. Sixth Bridge Workshop, T/C (J) will be held in the U.S. just prior

to the 22nd UJNR Joint Meeting.

Scheduling for the workshops and planning meeting shall be performed by

the US and Japan chairmen of the respective Task Committee with

concurrence of the Joint Panel chairmen. Results of each activity

shall be presented at the next Joint Panel Meeting.

The Panel recognized the importance of continued exchange of personnel

,

technical information, research results, and recorded data that lead to

mitigating losses from earthquakes and strong winds. The Panel also

recognizes the importance of using available large-scale testing

facilities in both countries. Thus, these activities should continue

to be strengthened and expanded and, as appropriate, share Task

Committee activities work at other meetings that have technical

interest in the Task Committee activities. To facilitate these

exchanges, the Panel will provide official endorsement.

The Twenty-Second Joint Meeting of the UJNR Panel on Wind and Seismic

Effects will be held at NIST, Gaithersburg , Maryland, U.S., May 1990.

Specific dates, program, and itinerary will be proposed by the U.S.

Panel with concurrence of the Japan Panel.

XXV

Theme I

Wind Engineering

A Summary of the Federal Highway Administration's

Wind Engineering Research Program

by

Harold R. Bosch 1

ABSTRACT

Since the dramatic collapse of the Tacoma Narrows

Bridge (Washington) in 1940, the Federal HighwayAdministration has actively pursued a program of

research aimed at advancing our understanding of

the effects of wind forces on structures and

ensuring the aerodynamic stability of long-span

bridges. This research has involved scale model

testing in wind tunnels, analytical

investigations, numerical modeling,

instrumentation development, full scale

measurements at bridge sites, software

development, coordination of national research,

and technology exchange. This paper will attempt

to highlight some of the research activities

which have been undertaken during the past 50

years with the aim of providing an overview of

the Administration's wind engineering research

program.

KEYWORDS: Bridge; cable-stayed; dynamics;

stability; suspension; wind.

1. INTRODUCTION

Almost a half century ago, the dramatic collapse

of the Tacoma Narrows Bridge sparked a major

investigation into the effects of wind on suspension bridges. To coordinate the many

activities which were to be undertaken, the

Advisory Board on the Investigation of Suspension

Bridges was formed. The Board was broadly

representative of engineers responsible for

specific suspension bridges, research engineers

having competence in aerodynamics and suspension

bridge theory, and representatives of industry

with demontrated ability and leadership in the

fabrication and erection of suspension bridges

[1].

The first action of this Board was to recommend

that the Bureau of Public Roads (BPR), now the

Federal Highway Administration (FHWA), cooperate

actively in a broad attack on the problems of the

aerodynamic behavior of suspension bridges. The

Bureau quickly responded by initiating agreements

with: the Washington Toll Bridge Authority to

conduct model studies in a special wind tunnel at

the University of Washington [2]; the Golden Gate

Bridge and Highway District for development of

suitable bridge instruments, installation, and

operation on the Golden Gate Bridge [3]; the

American Institute of Steel Construction for

analytical studies by the late Dr. Friedrich

Bleich [4]; and the Oregon State Highway

Department for theoretical analysis and static

load testing on a suspension bridge model.

This cooperative research served to identify the

mechanism of the wind action which caused the

failure of the original Tacoma Narrows Bridge and

firmly established the stability of the proposed

design for the new bridge. In the process, it

demonstrated that properly scaled model studies

conducted in suitable wind tunnels could be used

effectively to evaluate structural designs and

would reproduce quite faithfully the observed

bridge behavior.

The research also provided a number of

significant observations concerning the response

of suspension bridges to wind action [5]:

o Girder-stiffened suspension bridges are prone

to restricted vertical or torsional

oscillations. The strength and amplitude of

these oscillations can be predicted from wind

tunnel study,

o Girder-stiffened suspension bridges are also

prone to catastrophic vertical or torsional

oscillations depending on their depth to width

ratio. Generally, narrow and deep structures

have a tendency for vertical motion while wide

and shallow structures have a tendency for

torsional motion,

o Truss-stiffened suspension bridges are also

subject to minor restricted oscillations;

however, their main danger is from flutter

resulting from a coupling of modes. The

motion is usually dominated by the torsional

component.

o In addressing the question of coupling, Dr.

Bleich applied the classical theory of flutter

to the behavior of suspension bridges. His

analysis indicated that an effective means for

combating the catastrophic behavior of trussstiffened bridges would be to increase the

torsional frequency through use of top and

bottom lateral systems.

o Although the use of deck vents or slots proved

to be effective on the new Tacoma Narrows

Bridge, they are not generally effective on girder-stiffened suspension bridges.

Although this early research provided

considerable insight into the problem of bridge

aerodynamic stability, it revealed some inconsistencies and left a number of questions

unanswered. Those involved in the work

recognized the need for continued research

leading to definite and consistent design

criteria which would ensure the aerodynamic

stability of suspension bridges and for

establishment of a more permanent wind tunnel

facility which would enable investigation of

1 Research Structural Engineer

Federal Highway Administration

McLean, Virginia 22101-2296 U.S.A.

3
specific designs until suitable criteria are developed. Many felt that the Bureau of Public

Roads should take on this obligation and

^coordinate the research efforts.

2. WIND RESEARCH PROGRAM

BPR, and later FHWA, gladly accepted this

responsibility and embarked upon a broad research

program which has involved coordinating national

and international wind investigations, sponsoring

contract research, conducting laboratory and

field studies in-house, and providing technical

guidance through committees, panels, or research

councils. For more than 40 years, the Bureau has

continued to play an important role in ensuring

the aerodynamic stability of long-span bridges.

Some of the major activities (Table 1) may be

grouped into the following broad categories:

o Building a wind tunnel facility. Few wind

tunnels in the U.S. are suitable for

investigating the effects of wind forces on bridges and none are totally dedicated to that

purpose.

o Developing laboratory equipment. Wind tunnel

laboratories require specialized test

equipment and custom support hardware which

often is not available commercially.

o Conducting scale model studies. Designs were proposed for many new suspension and cablestayed bridges which required testing.

Alterations were proposed for a number of

existing bridges which also required testing.

o Developing instrumentation systems. Durable,

automated instrumentation systems were required to perform long term measurements at

remote bridge sites. Few, if any, suitable

systems were available commercially.

o Performing full scale measurements. Detailed

wind and bridge response measurements are

needed at bridge sites to establish the exact

nature of local winds and to characterize

bridge dynamic behavior.

A brief overview of specific projects within each

of these categories will be presented in the

sections which follow.

3. AERODYNAMICS LABORATORY

Following the reconstruction of the Tacoma

Narrows Bridge, the temporary bridge testing

facilities at the University of Washington were dismantled to make way for other new buildings.

In response, the Bureau made a careful survey of

existing wind tunnel facilities in the United

States to determine if one might be adapted for

the purpose of studying the effects of wind

forces on suspension bridges. BPR determined

that it would be more cost effective and

expeditious to build its own specialized

facility. During the 1950's, the aerodynamics

laboratory and wind tunnel (Figure 1) were

designed, constructed, and placed into service at

BPR's Fairbank Highway Research Station in

McLean, Virginia. The wind tunnel is a low

velocity, open-circuit, laminar flow tunnel

designed primarily for study of wind loads on long-span bridges [6,7]. During normal

operation, air enters through the double inlet

section of the tunnel, passes through a diffusion

section containing a series of screens, and exits

through a 6 by 6-ft (1.8 by 1.8-m) nozzle into

the test area. Wind speeds up to 50 fps (15.2

mps) can be generated and the vertical angle of

the wind can be changed by rotating the nozzle up

and down.

4. LABORATORY INSTRUMENTATION

4.1 Turbulence Generator

During the period 1977-1983, FHWA sponsored

research at Colorado State University to develop

a new technique for laboratory simulation of

large scale turbulence suitable for use in bridge

section model studies [8]. This research led to

design and implementation of a unique turbulence

simulation system in the FHWA Aerodynamics

Laboratory. The concept of this system is shown

in Figure 2. The main components of the

turbulence generator are two arrays of

symmetrical airfoils and an electrohydraulic

driving system. The airfoil assemblies are located just downstream of the wind tunnel

nozzle. The airfoils at location A oscillate 180

degrees out of phase to produce horizontal

velocity fluctuations while those at location B

oscillate in phase to produce vertical

fluctuations. Both sets of airfoils are driven

by hydraulilc actuators and each can be operated

independently through a master control system

(Figure 3).

4.2 Force-Balance

During the period 1985-1986, FHWA designed and

implemented a high frequency force-balance for

measuring lift, drag, and pitching moment forces

on bridge section models (Figure 4). The forcebalance system actually consists of two separate

balance mechanisms, one located at each end of

the model. This rather unique feature enables

detection of unbalanced loads on the bridge. The

balance system is mobile and can be rolled into

and out of the test section as necessary.

Section models mounted in the force-balance can

be rotated through a range of 180 degrees to vary

the angle of attack.

5. SCALE MODEL STUDIES

5. 1 Suspension Bridge Studies

During the period 1960-1968, wind tunnel studies

(Table 2) were conducted to evaluate the

aerodynamic stability of seven suspension bridges

[9-16]: San Pedro-Terminal Island (California),

Golden Gate (California), Tagus River (Portugal),

Lion's Gate (Canada), Narragansett Bay (Rhode

Island), New Quebec (Canada), and Bosporus

(Turkey). Most studies were aimed at assessing

proposed designs for new bridges, while some,

like Golden Gate, were to investigate proposed

alterations. All of these experiments were conducted under laminar flow conditions with the

4
exception of Lien's Gate, where a parallel bridge

was placed upwind of the model for a portion of

the testing. Rigid section models with geometric

length scales ranging from 1:^0 to 1:75 were

designed and fabricated for each test. The

models were mounted on a spring suspension system

which was carefully adjusted to provide the

correct vertical and torsional frequencies. Wind

speeds were varied from 0 to 50 fps (15.2 mps)

while angle of attack was varied from -5 to +10

degrees. Model response was observed, recorded

on strip chart, and later evaluated to obtain

data on logarithmic increment, logarithmic

decrement, amplitude, frequency, and critical

velocity.

5.2 Generic Flutter Studies

During the period 1969-1971, wind tunnel studies

were performed on a dozen generic sections in

conjunction with analytical work being conducted

by Sabzevari, Scanlan, and Tomko [17]. The

models were chosen to be representative of truss,

box, and girder type bridge sections. The

research centered on developing an analytical

model for bridge flutter, in which the

aerodynamic information was to be obtained from

experiment. Behavior of the rigid section models

was observed in the wind tunnel and aerodynamic

force coefficients were inferred from the

response of the models. Experimental procedures

were developed for efficient and reliable

extraction of the necessary aerodynamic

information.

5.3 Cable-Stayed Bridge Studies

During the period 1972-1980, FHWA conducted a series of model studies to investigate the

aerodynamic behavior of cable-stayed bridge

designs. Wind tunnel investigations were

performed on the design for Sitka Harbor Bridge

(Alaska), 14 design alternates for LulingDestrehan Bridge (Louisiana), and the steel box

design alternate for East Huntington Bridge (West

Virginia). Like earlier experiments, these tests

employed rigid section models, were conducted in

laminar flow, and vertical wind angle was varied.

In contrast with the earlier work, flutter

coefficients were extracted, mean forces (lift,

drag, and moment) were measured, and vortexinduced bridge response was carefully documented

[18,19].

5.4 Bridge Rehabilitation Studies

During the period 1981-1983, Phase I of a comprehensive investigation into the aerodynamic

behavior of the Deer Isle-Sedgwick Bridge (Maine)

was accomplished. Section model tests were

performed on a 1:25 scale section model of the

existing bridge as well as three proposed

modifications [20]. Laminar flow conditions were

maintained throughout this phase of testing. A series of separate tests were conducted to

identify general response characteristics, to

extract flutter coefficients, to measure mean

forces, and to study vortex-induced response.

5.5 Turbulence Simulation Studies

During the period 1984-1986, experimental

procedures were developed to conduct section

model studies in sumulated, large-scale

turbulence [21]. Wind tunnel experiments

incorporating simulated turbulence were performed

on Deer Isle (Phase II) and Golden Gate section

models using FHWA's unique turbulence generator

[22]. The bridge models were subjected to

various levels of turbulence and the resulting

response compared to results from previous

studies in laminar flow.

5.6 Generic Section Model Studies

During the period 1987-1988, special model

studies were conducted on generic section models

representing "Box" and "H" structural shapes, seven of each, with depth to width ratios ranging

from 0.10 to 1.00. Initially, all the shapes

were tested in both laminar and simulated

turbulent flow conditions without varying the

wind angle [23]. Flutter coefficients were

obtained for each test condition. Similar tests

are being conducted at the Chinese Aerodynamic

Research and Development Center and the results

from both laboratories are being compared.

Following these experiments, another series of

tests were performed in conjunction with research

on analytical modeling of vortex-induced response

[24]. For this study, three shapes exhibiting

strong response to vortex shedding were selected

from the original set of fourteen. Testing was performed only in laminar flow and attention was focused on regions of resonant response. The

variation of response amplitudes within the

resonant region (s) was carefully recorded as was

the nature of vortex shedding in the wake and the

coherence of shedding along the span of the

model

.

6. FIELD INSTRUMENTATION

6.1 Melpar System

In 1969, a data acquisition system was designed

to monitor and record wind velocities and

direction at remote bridge sites and to document

their effects on suspension bridges [25]. The

system was designed to operate unattended, being

triggered into operation by wind velocities or bridge displacements which exceed predetermined

threshold levels. To characterize the turbulent

nature of the wind, 3-component propeller type

anemometers were selected. These were to be

spaced at intervals along the bridge span to sense wind components normal, parallel, and

vertical relative to the roadbed. To

characterize the bridge response, servo type

accelerometers were selected. The accelerometers

were to be installed in pairs, one near each edge

of the roadbed, at key points along the bridge

span. Accelerometers were also to be placed in a bridge tower to identify bending, twist, and sway

of the tower. Provisions were also made for use

of an aerovane type wind sensor which provides

basic wind speed and direction data similar to

5
that obtained by the National Weather Service.

The data acquisition equipment included circuits

or modules for signal conditioning, amplifying,

filtering, multiplexing, and digitizing the

incoming analog signals. A dual buffer was provided for intermediate storage of the

digitized data. As one buffer fills, the system

automatically switches to the other buffer and

the data is dumped to one of two 9-track, 800 BPI

digital tape drives. The system was designed to

automatically sample up to 70 input channels at

8-bit resolution using a system clock rate of 500

samples per second and have the capability to

store 10 hours of continuous wind and bridge

data.

6.2 Kinemetrics System

By 1978, FHWA was expanding its field measurement

program and developed a need for a second data

acquisition system. A new system, similar to the

prototype, was designed [26]. This new instrumentation system incorporated at number of

technological advances and enhancements:

o 12-bit analog-to-digital conversion,

o Variable number of channels (61 max),

o Variable sampling rate (90,000 max),

o Variable filter frequencies and gains,

o Dual 1600 BPI 9-track tape drives,

o Special offset amplifiers,

o Uninterruptable power supply.

7. FULL SCALE MEASUREMENTS

Full scale measurements at bridge sites (Table 3) are an essential part of the wind research

program. The wind data obtained provides

detailed information regarding prevailing

conditions (speeds, directions, angle of attack)

and turbulence properties (intensity, spectrum,

scale) which is often not available from other

sources. The structural response data provides

information on the structure's dynamic properties

(frequencies, mode shapes, damping) which can be

compared with computed values. Wind and

structural measurements are often used in the

design of wind tunnel experiments and to evaluate

the effectiveness of aerodynamic retrofits. They

can also provide a means for determining the

adequacy of experimental and analytical

techniques.

7. 1 Ambient Vibration Surveys

In 1969, BPR awarded a contract to Teledyne

Geotronics to investigate the feasibility of

using the small amplitude vibrations normally

present in suspension bridges to measure natural

frequencies, mode shapes, and damping factors.

Teledyne had previously been quite successful in

using such an approach to determine the dynamic

characteristics of large buildings. The survey

technique was demonstrated on two moderate sized

suspension bridges, the Newport Bridge (Rhode

Island) and the Chesapeake Bay Bridge (Maryland).

The test results were quite encouraging. On the

Newport Bridge, the contractor identified 20 mode

shapes, including five symmetric vertical, four

anti -symmetric vertical, the first symmetric and

first anti-symmetric torsional, six lateral, and

three tower modes [27]. Mode shapes were in good

agreement with calculated values although

measured frequencies were about 20% higher.

Damping estimates ranged from 3% for the first

lateral mode to 0.k% for the fourth antisymmetric vertical mode. Similar information was obtained for the Chesapeake Bay Bridge [28].

7.2 Wind and Bridge Response Studies

7.2.1 Newport Bridge

In 1971, FHWA installed its new wind and bridge

response instrumentation system on the Newport

Bridge (Rhode Island). Completed in 1969, the

Newport Bridge is a truss-stiffened suspension

bridge with a 1600-ft (488-m) main span, 688-ft

(210-m) side spans, and 48-ft (14.6-m) roadway.

Although this instrumentation was not fully

deployed due to conflicts with painting crews on the bridge, it was used to record wind data from

a portion of Hurricane Doria in August 1971. A total of five 3-component anemometers spaced at

350-ft (107-m) intervals along the span were in

place at the time. Mean wind speeds of 60 mph

(97 km/h) with peak gusts to 75 mph (121 km/h) were observed at the site. Wind data was recorded continuously for 10 hours. Wind angle

ranged from 20 degrees for mild winds to 3 degrees for strong winds [29,30].

7.2.2 Sitka Harbor Bridge

In 1972, the instrumentation system was installed

on the Sitka Harbor Bridge (Alaska). The Sitka

Bridge is the first cable-stayed bridge in the

United States to carry vehicular traffic. It is

a five-span structure with a total length of 1053

ft (321 m) and a main span of 150 ft (137 m).

The deck is 38 ft (11.6 m) wide and is supported

by two steel box girders 30 by 72 in (76 by 183

cm). The two towers consist of free-standing

pylon legs with no portal framing. Five 3-

component anemometers and 10 accelerometers were

mounted along the bridge span, 3 accelerometers

were mounted at the top of one tower, and a strong motion accelerograph was installed at the

base of one tower (Figure 5). Much of the wind

observed at this site was parallel to the

structure and did little to excite the span.

Vortex-induced motion of the towers perpendicular

to the plane of the stay cables was frequently

observed, however. The tower legs were typically

excited in the fundamental mode at wind speeds of

2k to 26 mph (39 to 42 km/h) and a frequency of

1.16 cps [31,32].

7.2.3 Perrine Bridge

In 1977, FHWA installed the instrumentation on the Perrine Bridge (Idaho). The Perrine Bridge

is a 1500 ft (457 m) long structure over the 500

ft (152 m) deep Snake River canyon. The main

span of the bridge is a steel deck arch with a span of 993 ft (303 m) and a rise of 210 ft (64

m). The deck, stringers, and floor beams rest

6
atop box-shaped columns which span to the top

chord of the arch truss. These columns are not

braced and range in length from 7 to 1 60 ft (2.1

to 48.8 m). Pairs of accelerometers were mounted

at mid height of three of the westerly columns

(one short, one long, and one medium length) to

monitor vibrations normal and parallel to the

structure. To measure stresses in the columns,

strain gages were mounted in the middle of each

face near the base of the column. Three 3-

component anemometers were installed on booms

extended from the top chord facing west (one near

the crown and one near each quarter point).

Wind, traffic, and column vibration measurements

were recorded on digital tape and strip chart for

later analysis [33]«

7.2.4 Pasco-Kennewick Bridge

In 1979, the instrumentation system was installed

on the Pasco-Kennewick Bridge (Washington). The

Pasco-Kennewick Bridge is the first segmental

concrete cable-stayed bridge constructed in the

U.S. The bridge is 2503 ft (763 m) long with a main span of 981 ft (299 m). Six 3-component

anemometers were mounted along the deck with

spacings ranging from 33 to 525 ft (10 to 160 m).

Accelerometers were attached to each edge of the

deck section at six stations along the main and

side spans and three accelerometers were

installed in the top of one tower. During the 24

month monitoring period, 49 hours of wind and

bridge response data, representing approximately

100 events, were recorded [34]. Unfortunately,

the highest wind speeds encountered were only 27

mph (43 mps). As expected, the bridge exhibited

very little response to the given conditions.

Measured frequencies and mode shapes compared

quite well with computed values. Measured wind

spectra was also quite similar to selected

empirical models.

7.2.5 Deer Isle-Sedgwick Bridge

In 1981, FHWA refurbished the instrumentation

system and installed it on the Deer Isle-Sedgwick

Bridge (Maine). The Deer Isle Bridge is a girder-stiffened suspension bridge built in 1939

and similar in cross section to the original

Tacoma Bridge. Immediately following the

collapse of the Tacoma Bridge, this bridge was stiffened through addition of longitudinal and

diagonal stays as well as cable clamps. Despite

these modifications, the bridge has continued to

oscillate in certain storm conditions. The Deer

Isle Bridge has a main span of 1080 ft (329 m)

with side spans of 484 ft (148 m) and approach

spans of 130 ft (40 m). The deck is 23.5 ft (7.2

m) wide and the girders are 6.5 ft (2.0 m) deep.

Six 3-component anemometers were mounted on booms

facing seaward. Two aerovane type instruments

were installed on the north bridge tower, one at

the top and one near the bottom about 33 ft (10

m) above the water. Temperature sensors were

located at the top and bottom of the same tower

and outside as well as inside the instrument

house located just north of the bridge. Twelve

accelerometers were installed on the bottom

flange of the stiffening girders at six stations

along the main span and north side span. Three

accelerometers were installed at the top of the

north tower to measure tower bending and sway.

More than 600 hours of wind and bridge response

data have been recorded and data collection is

continuing. Although the prevailing wind

direction is from the northwest, wind events from

most directions have been sampled. Wind speeds

up to 80 mph (129 km/h), including a portion of

Hurricane Gloria, have been captured. Bridge

motion has been primarily in the vertical degreeof -freedom and frequencies as well as mode shapes

compare reasonably well with calculations

[35,36].

7.2.6 Luling-Destrehan Bridge

In 1984, a new instrumentation system was installed on the Luling-Destrehan Bridge

(Louisiana). The Luling Bridge is the first

major U.S. cable-stayed bridge to be built in the

hurricane region of the Gulf of Mexico. The

bridge consists of a five-span superstructure

made entirely of weathering steel. The main span

is 1235 ft (376 m) with side spans of 500 and 260

ft (152 and 79 m). The roadbed is 82 ft (25 m)

wide and supported by twin trapezoidal box

girders 14 ft (4.3 m) deep. Twenty-four groups

of stay cables radiate from the two A-frame

towers which are 400 ft (122 m) tall. The stay

groups consist of either two or four cables each.

Five anemometers have been installed atop special

luminaire poles on the median barrier, three in

the main span and one in each back span [37].

Thirty accelerometers were mounted on the bridge,

18 on the cables, 10 in the box girders, 2 at the

top of the south tower. Storm activity at the

site in 1984-85 was infrequent and maximum wind

speeds of only 30 mph (48 km/h) were observed.

As expected, the bridge remained quite stable

[38]. Since that time, portions of Hurricanes

Juan and Florence have been recorded, but wind

speeds failed to exceed 35 mph (56 km/h). Data

has been analyzed to determine bridge deck and

cable frequencies as well as motion amplitudes.

The results of measurements compare well with

calculated values.

8. CONCLUSIONS

In this paper, the author has attempted to

provide a brief summary of the Federal Highway

Administration's wind engineering research

program which has spanned half a century. The

detailed tasks and results of each project

highlighted could not be presented here and not

all of the investigations performed could be

mentioned. Some of the activities not covered

involve application of computational fluid

dynamics, development of computer software,

implementation of design aids, cataloging of

research results, dynamic calibration of wind

sensors, and automation of wind tunnel testing.

This research program, and others, have

contributed significantly toward advancing our understanding of the effects of wind forces on structures and have ensured the aerodynamic

7
stability of modern long-span bridges.

9. REFERENCES

1. Vincent, G.S., "Research as Related to

Highway Bridges," Proceedings, American

Institute of Steel Construction, National

Engineering Conference, April 1962.

2. Farquharson, F.B., Smith, F.C. and Vincent,

G.S., "Aerodynamic Stability of Suspension

Bridges with Special Reference to the Tacoma

Narrows Bridge," University of Washington,

Bulletin 116, Parts I-V, June 1 9^9-June

1954.

3o Vincent, G.S., "Golden Gate Bridge Vibration

Studies," ASCE, Journal of the Structural

Division, Volume 84, ST6, October 1958.

4. Bleich, F. , et.al., "The Mathematical Theory

of Vibration in Suspension Bridges," BPR,

Government Printing Office, 1950.

5. Vincent, G.S., "A Summary of Laboratory and

Field Studies in the United States on Wind

Effects on Suspension Bridges," Proceedings,

Symposium on Wind Effects on Buildings and

Structures, Teddington, England, June 1963.

6<. Vincent, G.S., "Wind Tunnel for Aerodynamic

Testing of Section Models of Suspension

Bridges," Public Roads Magazine, Volume 30,

No. 2, June 1958.

7. Bosch, H.R., "Aerodynamics Laboratory," FHWA

Laboratory Flyer, Report No. FHWA/RD-88/070,

August 1988.

8o Cermak, J.E., Bienkiewicz, B., and Peterka,

J. A., "Active Modeling of Turbulence for

Wind-Tunnel Studies of Bridge Models," FHWA,

Report No. FHWA/RD-82/148, February 1983.

9. Wiles, E.G., "Aerodynamic Studies on Proposed San Pedro-Terminal Island

Suspension Bridge," BPR Report, 1960.

10. Vincent, G.S., and Wiles, E.G.,

"Investigation of Aerodynamic Effects of

Proposed Alterations of the Golden Gate

Bridge," BPR Report, November 1960.

11. Vincent, G.S., "Aerodynamic Studies on Proposed Tagus River Suspension Bridge," BPR

Report, 1961.

12. Vincent, G.S., "Aerodynamic Studies on Proposed Tagus River Suspension Bridge," BPR

Supplemental Report, 1961.

13- Cayes, L.R., "Aerodynamic Studies on Lion's

Gate Suspension Bridge," BPR Report,

February 1964.

14. Cayes, L.R., and Michael, D.L., "Aerodynamic

Stability of Proposed Narragansett Bay

Suspension Bridge," BPR Report, December

1965.

15. Cayes, L.R., "Aerodynamic Stability of

Proposed New Quebec Suspension Bridge," BPR

Report, October 1966.

16. Cayes, L.R., "Aerodynamic Stability of

Proposed Bosporus Suspension Bridge," BPR

Report, June 1968.

17. Scanlan, R.H., and Tomko, J.J., "Airfoil and

Bridge Deck Flutter Derivatives," ASCE,

Journal of the Engineering Mechanics

Division, December 1971.

18. Bosch, H.R., "Aerodynamic Investigations of

the Luling Louisiana Cable-Stayed Bridge,"

FHWA, Report No. FHWA-RD-77-161 , July 1978.

19. Cayes, L.R., "Aerodynamic Stability of

Proposed Ohio River Cable-Stayed Bridge,"

FHWA, Report No. FHWA-RD-77-1 57 , December

1977.

20. Bosch, H.R., "A Wind Tunnel Investigation of

the Deer Isle-Sedgwick Bridge (Phase I),"

FHWA, Report No. FHWA/RD-87/027 , August

1987.

21. Huston, D.R., and Bosch, H.R., "Active

Turbulence Generation for Section Model

Studies," Proceedings, Sixteenth U.S. A.-

Japan Conference on Wind and Seismic

Effects, May 1984.

22. Huston, D.R., Bosch, H.R., and Tanaka, H.

,

"The Effects of Large-Scale Upstream Gusting

on the Aeroelastic Behavior of Section

Models," Proceedings, Eighteenth U.S. A.-

Japan Conference on Wind and Seismic

Effects, May 1986.

23. Huston, D.R., "Flutter Derivatives From 14

Generic Deck Sections," Proceedings, ASCE

Structures Congress, Orlando, 1 987

24. Ehsan, F., and Bosch, H.R., "Modeling the

Effect of Spanwise Coherence of Aerodynamic

Forces in Full-Bridge Response,"

Proceedings, Twentieth U.S. A. -Japan

Conference on Wind and Seismic Effects, May

1988.

25. American Standard, Inc., "An Instrumentation

System to Monitor and Record Wind

Velocities, Directions, and Effects at

Suspension Bridge Sites," FHWA, March 1971.

26. Bourgeois, B.S., et.al., "Wind

Instrumentation of the Luling Bridge,

Luling, Louisiana (Volume I)," Louisiana

Transportation Research, Report No. FHWA/LA79-1 ST, March 1982.

27. McLamore, V.R., "Ambient Vibration Survey of

the Newport Bridge," Teledyne Geotronics,

Report No. 0370-2150, March 1970.

28. McLamore, V.R., "Ambient Vibration Survey of

the Wm. Preston Lane Memorial Bridge,"

Teledyne Geotronics, Report No. 0370-2152,

March 1970.

29. Gade, R.H., and Scanlan, R.H., "Aspects of

Hurricane Winds as Recorded at an Instrumented Bridge," Proceedings, Eighth

U.S. A. -Japan Conference on Wind and Seismic

Effects, May 1976.

30. Gerhardt, C.L., "Monitoring and Recording

Wind Velocities, Directions, and Effects at

Long-Span Bridge Sites," FHWA, Report No.

FHWA-RD-76-10, August 1975.

31. Gade, R.H., "Status of Investigation of

Aerodynamic Behavior of the Sitka Harbor

Cable-Stayed Bridge," Proceedings, Symposium

on Full Scale Measurements of Wind Effects

on Tall Buildings and Other Structures,

London, Ontario, June 1974.

32. Nelson, D.D., Greene, D.M., and Keay, N.S.,

"Wind and Bridge Motion Data Recorded at the

Cable-Stayed Bridge, Sitka, Alaska," FHWA,

Report No. FHWA-RD-78-57, May 1977.

33 • Cheng, D.H., "Aerodynamic Stability of Long

Columns with Elastic End Constraints," FHWA,

Report No. FHWA/RD-81/133, February 1982.

8
3h. Bampton, M.C.C., et.al., "Pasco-Kennewick

Cable-Stayed Bridge Wind and Motion Data,"

FHWA, Report No. FHWA/RD-82/067 , February

1983.

35. Bampton, M.C.C., et.al., "Deer IsleSedgwick Suspension Bridge Wind and Motion

Analysis," FHWA, Report No. FHWA/RD-86/ 183

,

March 1986.

36. Bosch, H.R., "Full Scale Studies on the Deer

Isle-Sedgwick Suspension Bridge,"

Proceedings, Fifth U.S. National Conference

on Wind Engineering, November 1985.

37. Thompson, H.A., et.al., "Proposed Plans for

Measurement of the Aerodynamic Behavior of

the Luling, Louisiana Cable-Stayed Bridge,"

Public Roads Magazine, Volume 45, No. 3,

December 1981.

38. Bruce, R.N., et.al., "Dynamic Response to

Wind and Wind Climate of the LulingDestrehan Bridge at Luling, Louisiana,"

Louisiana Transportation Research, Report

No. FHWA/LA-82-1ST, July 1986.

Table 1. Research Time^-Phase Chart

RESEARCH YEARS

ACTIVITY 1950's 1960 's 1970's 1980's

WIND TUNNEL

CONSTRUCTION

TURBULENCE

GENERATOR

FORCE-BALANCE

SUSPENSION

BRIDGE STUDY

GENERIC

FLUTTER STUDY

CABLE- STAYED

BRIDGE STUDY

BRIDGE REHAB.

STUDY

TURBULENCE

STUDY

GENERIC MODEL

STUDY

AMBIENT VIBR.

SURVEY

WIND 5 BRIDGE

MOTION STUDY

9
Table 2. Wind Tunnel Model Studies.

BRIDGE OR MODEL LOCATION DATE OF TEST TYPE OF FLOW

SAN PEDRO

TERMINAL ISLAND

CALIFORNIA

U.S.A.

1960 LAMINAR

GOLDEN GATE

(Alterations)

CALIFORNIA

U.S.A.

1960 LAMINAR

TAGUS RIVER PORTUGAL 1961 LAMINAR

LION'S GATE CANADA 1964 LAMINAR

TURBULENT

NARRAGANSETT

BAY

RHODE ISLAND

U.S.A.

1965 LAMINAR

NEW QUEBEC CANADA 1966 LAMINAR

BOSPORUS TURKEY 1968 LAMINAR

GENERIC DECK

SECTIONS

1969-1971 LAMINAR

SITKA HARBOR ALASKA

U.S.A.

1970 LAMINAR

LULINGDESTREHAN

LOUISIANA

U.S.A.

1972-1976 LAMINAR

EAST HUNTINGTON

(Steel Box Alt)

WEST VIRGINIA

U.S.A.

1977 LAMINAR

DEER ISLESEDGWICK

MAINE

U.S.A.

1981-1983 LAMINAR

GOLDEN GATE CALIFORNIA

U.S.A.

1984-1985 LAMINAR

TURBULENT

DEER ISLESEDGWICK

MAINE

U.S.A.

1985-1986 LAMINAR

TURBULENT

GENERIC "BOX" § "H" SECTIONS

1987-1988 LAMINAR

TURBULENT

Table 3. Full Scale Field Measurements

BRIDGE LOCATION DATE OF MEASUREMENT

NEWPORT RHODE ISLAND 1969

CHESAPEAKE

BAY MARYLAND 1969

NEWPORT RHODE ISLAND 1971

SITKA

HARBOR

ALASKA 1972-1976

PERRINE IDAHO 1977-1978

PASCO

KENNEWICK

WASHINGTON 1979-1980

DEER ISLESEDGWICK

MAINE 1981 -Present

LULINGDESTREHAN

LOUISIANA 1984 -Present

10
Figure 1 . FHWA Wind Tunnel

A = Horizontal Gust Control B = Vertical Gust Control

C = Linkage

D = Hydraulic Actuators

E = Bridge Model

Figure 2. Concept of Active Turbulence Generator.

^-g; —Turbulence Generator

Figure 3. Control -Driving System

11
TUNNEL- MODEL

' DUAL BALANCE

Figure 4. Dual Balance on Mobile Support,

A11-A13 (SEE INSET A)

WEST \. 52'ACLRWC KEAST

JAPONSKI U- 150" _ 4S0' . — 150' SITKA

1 ft = .30 m

•A2 —o

A4

aA6•A8

Al A3 AS A7

I

:

Wl —a

1"

W2 IW3

'JW4 —

1

PREVAILING

INSET A

WIND

PARALLEL TO ROADWAY

' / /

A12/

Al

'normal to ROADWAY

PARALLEL TO

ROADWAY

SANORTH

A12 A10

A9

A13 W5

INST. HOUSE

• AG

• T

(SEE INSET B)

INSET B

W NORTH VERTICAL

' 1

\ /

/

t EAST

' /

V ALONG

U ACROSS

POSITIVE OUTPUT WITH WIND

FROM INDICATED DIRECTIONS

INSTRUMENTATION

A1-A13 ACCELEROMETERS

(VERTICAL MOTION) W1-W5 3-D ADEMOMETERS

(INSET A) AG ACCELEROGRAPH

T TEMPERATURE Figure 5. Instrumentation Configuration - Sitka.

12
Wind Resistant Design Manual for

Highway Bridges in Japan

by

Nobuyuki Narita1

, Koichi Yokoyama2 and Hiroshi Sato3

ABSTRACT

The new wind resistant design manual for highway

bridges has been required recently in Japan. The working

group was organized in the Japan Road Association in

1984 to prepare the manual. The working group was

reformed into the wind resistant design committee in

1986, and the first draft of The Proposed Wind Resistant

Design Manual For Highway Bridges was completed.

The contents of the Manual are as follows.

1. GENERAL REMARKS

2. GUIDELINE OF WIND RESISTANT DESIGN

3. WIND PROPERTIES USED IN DESIGN

(Basic Wind Speed, Design Wind Speed, Intensity of

Turbulence, etc.)

4. WIND-INDUCED VIBRATIONS OF BRIDGE DECK

(Prediction, Evaluation, Method of Alleviation, Flutter, Galloping, Vortex-Induced Vibrations, Gust

Responses)

5. WIND-INDUCED VIBRATIONS OF TOWERS AND

BRIDGE MEMBERS

6. WIND RESISTANT DESIGN AT ERECTION STAGES

7. WIND TUNNEL TESTING METHODS

KEY WORDS: Wind resistant design, Highway

bridge, Design manual, Wind-induced vibration.

1. INTRODUCTION

The design of short-and medium-span highway

bridges in Japan has been made according to The Specification For Highway Bridges (hereinafter referred to as

'the Specification'). With regard to wind resistant

design, the Specification prescribes design wind load of

bridges and dimensions of pipe structure, however, there

is not a sufficient provision for wind-induced vibrations

in the Specification.

The wind resistant design of long-span bridges in Japan has been mostly based on The Wind Resistant

Design Criteria For Honsyu-Shikoku Bridges (hereinafter

referred to as 'the Criteria'). The Criteria was originally

formulated in 1967 and revised in 1976 by the ad hoc

committee of the Japan Society of Civil Engineers. The

Criteria prescribes the design wind load including the

effect of gust and the evaluation method of windinduced vibrations based on wind tunnel testing.

Since the revision of the Criteria, wind engineering

has made a remarkable progress and a considerable

amount of aerodynamic data of bridges have been ac- quired. The wind resistant design for limited kinds of

bridges can be made without wind tunnel testing if

reliable formulae for the estimation are provided, The

prediction of wind-induced vibration will become more

reasonable and reliable if the effect of turbulence is incorporated appropriately in the design.

From these reasons, the new wind resistant design

manual for highway bridges has been required recently

in Japan. The working group was organized in the Japan

Road Association in 1984 to prepare the manual. The

working group was reformed into the wind resistant

design committee in 1986, and the first draft of The

Proposed Wind Resistant Design Manual For Highway

Bridges (hereinafter referred to as 'the Manual') was

completed.

In this paper described are some of the features of

the Manual. It should be noted that the final version of

the Manual, which will be published in 1989, will not be

necessarily the same as the first draft described in the

paper.

2. PROCEDURE OF WIND RESISTANT DESIGN

USING THE PROPOSED WIND RESISTANT

DESIGN MANUAL

The procedure of wind resistant design using the

Manual is shown in Fig. 1. The wind properties (basic

wind speed, design wind speed, turbulence intensity etc.) are decided first. Then the design wind load is calculated according to the provisions of the Specification,

and the static design of the bridge is made. This procedure is the same as the existing wind resistant design

method.

At this stage, where major dimensions of the bridge

are determined, wind-induced vibrations to be studied

further are chosen considering the design wind speed,

the deck width and the span length of the bridge (Table

1). For bridges with short span length, no study on

wind-induced vibrations is required.

For other bridges, wind-induced vibrations specified

above should be predicted using the formulae provided

in the Manual. Then the wind-induced vibrations are evaluated. The formulae were established considering

wind tunnel test results. Since only a few parameters

are included in the formulae, the prediction error of the

formulae is not negligibly small. Therefore, the formulae

are established so that their prediction may become

1. Dr. Eng., Director General, Public Works Research Institute,

Ministry of Construction, 1, Asahi, Tukuba-Shi, Ibaraki-Ken,

305, Japan

2. Head , Structure Div., PWRI, ditto

3. Senior Research Engineer, Structure Div., PWRI, ditto

13
safer. It means that evaluation 'good' based on the

formulae is almost always 'good', however, evaluation

'no good' based on the formulae is not always 'no good'

because of the safety margin of the formulae. In case

that the evaluation based on the formulae turns out

to be 'no good', the bridge engineer can modify the

desing or he can predict wind-induced vibrations by

means of wind tunnel testing. The evaluation based on

wind tunnel testing has priority over that based on the

formulae.

("start )

i wind properties 131

|

DESIGN" WIND LOAD

PREDICTION OF WIND-INDUCED

VIBRATIONS BY FORMULAE (4)

WIND TUNNEL

STUDY ( 4. 7 )

MINOR

Number in the Parenthesis shows corresponding chapter

in the Mannal

( END )

DESIGN AT THE

ERECTION STAGE 16)

Fig. 1 Procedure of wind resistant design

using the Manual

3. WIND PROPERTIES USED IN DESIGN

Wind properties are modeled fundamentally according to Davenport [1978, 1982], namely, lower mean

wind speed but higher turbulence in rough terrain and

low altitude, and higher mean wind speed but lower

turbulence in smooth terrain and high altitude. The

terrains are classified into four, namely, rough sea (Terrain I), open farmland (Terrain II), suburbs (Terrain

III) and city centres (Terrain IV).

3.1 Basic wind speed U10

The basic wind speed U10 is defined as the mean

wind speed over farmland (Terrain II) at an elevation of

10m, averaged over a period of 10 minutes. The probability that the annual maximum mean wind speed

exceeds the basic wind speed is 10% in a 50-year period.

Using the meteorological data at weather stations in

Japan, extreme wind speeds were estimated. The basic

wind speeds were classified into 4 categories, namely

35m/s, 40m/s, 45m/s, 50m/s. It is shown in Fig. 2.

3.2 Design wind speed Ud

The design wind speed can be obtained from the fol- lowing formula.

Ud = U10Eul 0)

where, Eul: correction factor for altitude and terrains

The design wind speeds and the correction factors are shown in Fig. 3.

14
Table 1 Type of wind-induced vibrations

to be studied further

Bridge Deck Criteria Wind — induced Vibration

to be studied

Truss Deck for Cable -Stayed or Suspension Bridge

if L>385B/U,j then Flutter

Open Solid Deck for Cable — Stayed or Suspension Bridge

L >200B/Uti

L>346B/U,i

and B< 5D

L>385B/U,i

Vortex — induced

Vibration

Galloping

Flutter

Closed Solid Deck for Cable-Stayed or Suspension BridgeL>200B/U( |

L>346B/U,i

and B< 5D

L>577B/U<i

Vortex - induced

Vibration

Galloping

Flutter

Girder Bridge L>200B/U(1

L>346B/U«]

and B< 5 D

Vortex — induced

Vibration

Galloping

where, L : main span length B ; bride deck width Ud : design wind speed D : bridge deck depth

Basic Wind Speed

45

50 m/s

0

Fig. 2 Basic wind speed

15
3.3 Turbulence properties

Typical values for turbulence properties such as turbulence intensities and power spectral density functions are provided in the Manual. Turbulence intensity Iu is shown in Fig. 4.
4. PREDICTION OF WIND-INDUCED VIBRATIONS

BY THE FORMULAE

4.1 Flutter

Reduced critical wind speeds for flutter of solid decks

are shown in Fig. 5. It seems that the reduced critical wind speed increases with an increase in deck width to depth ratio (B/D). The lowest reduced critical wind

speed is about 2.5. As for a truss deck, the critical wind

speed varies much with minor difference of configuration of truss and floor deck. But the reduced critical wind speed is higher than 2.5.

The formulae for the prediction of the critical wind

speed of flutter is as follows:

Ucf = Ucfo Ef

Ucfo = 2.5 f0B

(2)

(3)

where,

Ucf:

Ef: Ucfo:

f»:

corrected critical wind speed for flutter,

correction factor for the effect of turbulence,

critical wind speed for flutter in smooth flow,

natural frequency of the 1st torsional mode.

Sinusoidal wave form observed in smooth flow

changes into a random form in turbulent flow. The

effect of turbulence on flutter, which is not so clear at the moment, will be incorporated into Ef.

4.2 Galloping

Galloping may take place for a solid deck whose deck

width to depth ratio (B/D) is relatively small. Fig. 6

suggests that the critical wind speed for galloping is very

high when B/D is over 5. For a solid deck whose B/D is

less than 5, the lowest reduced critical wind speed is

about 4.5. The formula for the prediction of the critical wind

speed of galloping is as follows:

Ucg = Ucgo Eg

Ucgo =4.5 frjB

(4)

(5)

where,

Ucg:

Eg:

Ucgo:

ft?:

corrected critical wind speed for galloping

correction factor for the effect of turbulence

critical wind speed for galloping in smooth

flow

natual frequency of the 1st bending mode.

Sinusoidal wave form observed in smooth flow

changes into a random form in turbulent flow. It was

found that onset velocity for negative damping, which

causes galloping, increases remarkably in turbulent flow

(Narita et al. [1987]). The effect of turbulence will be

incorporated, for example, into Eg.

4.3 Vortex-induced vibrations

Vortex-induced vibrations may take place for a solid

deck. The amplitude decreases with an increase in mass

or mass moment of inertia and structural damping. The

amplitude is very sensitive to the geometrical configuration, however, the amplitude decreases with an increase

in deck width to depth ratio (B/D).

On the other hand, the variation of reduced wind

speed for the occurrence of vortex-induced vibrations is smaller than that of the amplitude.

In the turbulent flow, the amplitude becomes smaller

than in smooth flow. This effect becomes more signifi- cant with higher turbulence intensity and cross sections

of large B/D. The ratios of the amplitude of vortexcfo tgB

O critical wind speed

of flutter

<^ no flutter was

observed up to the wind speed

h

( -3°< angle of attack < 3"

) O O o 6

h —e °t-r~

1 L 1 1 1 1

i i i i i 10 11 13 U 15

Fig. 5 Reduced critical wind speed for flutter of solid deck

17
U ego

Angle of Attack

i +

3" O :0-

;-3"

Arrow means no galloping was

j observed up to the wind speed

Ucgo

tr/Q D = 5

= 4.5

7.5B/D

Fig. 6 Reduced critical wind speed for galloping

of solid deck

Ev2 i.o

Iu

Fig. 7 Correction factor for effect of turbulence on

amplitude of vortex-induced vibrations

induced vibration in turbulent flow to that in smooth

flow suggested by several wind tunnel tests are shown in

Fig. 7. The formula for the prediction of the critical wind

speed and amplitude of vortex-induced vibrations is as follows:

UcV = r j3t? • ftj B bending (6)

L 00 • f0 B torsion

A = Ao Evl Ev2 (7)

Ao = fOT?(D/B)2 /(mr5T7) bending (t?/B) (8)

"U0(D/B)3 /(Ir 50) torsion (deg.)

where,

Ucv:

A:

Ao:

Evl:

wind speed for the maximum amplitude of

vortex-induced vibration

reduced wind speed for vortex-induced vibration (Provision of Hanshin Expressway

Public Corporation [1984] seems to be ap- propriate.)

corrected maximum amplitude of vortex- induced vibration

maximum amplitude of vortex-induced vibration for rigid model in smooth flow

correction factor for vibrational mode

18
(about 4/ff)

Ev2: correction factor for the effect of turbulence (Fig. 7)

oct\, ad : effect of sectional shape. (Provision of

Bridge Aerodynamics [1981] seems to be

appropriate.) mr, Ir: reduced mass or reduced mass moment of

inertia

5tj, 50: logarithmic decrement

5. PREDICTION OF WIND INDUCED VIBRATIONS BY WIND TUNNEL TESTING

Wind tunnel testing increases the reliability of wind

resistant design of bridges. Since the prediction of windinduced vibrations based on the formulae may often

provide safer value, there is a fair chance that a wind

tunnel study provides an economical design.

As is mentioned above, turbulence has significant

effects on wind-induced vibrations. Therefore, standard

wind tunnel testing methods in turbulent flow (full aeroelastic model test, taut-strip model test) are described in

the Manual as well as conventional testing methods

(spring-mounted rigid model test, measurement of

steady aerodynamic forces).

6. EVALUATION OF WIND-INDUCED

VIBRATIONS

6.1 Flutter and galloping

The following inequalities should be satisfied.

Ucf>Ur Flutter (9)

Ucg>Ur Galloping (10)

Ur= Ud Erl Er2 (11)

8. CONCLUSION

The new wind resistant design manual is now being

prepared in Japan. The main features of the proposed

wind resistant design manual for highway bridges are as follows.

a. The basic wind speeds of the whole country are provided.

b. The design wind speed and the properties of wind

turbulence can be estimated considering the altitude

of structure and the terrain around the structure.

c. The critical wind speed for flutter and galloping can be obtained by simple formulae as well as by wind

tunnel testing.

d. The amplitude of vortex-induced vibrations can be

estimated by simple formulae as well as by wind

tunnel testing. The effect of turbulence on vortex- induced vibrations are incorporated.

e. The method of verification of flutter, galloping and

vortex-induced vibrations are provided.

f. . The guideline of wind resistant design for bridge

members and for bridges at their erection stages are described.

g. The standard wind tunnel testing methods are described.

ACKNOWLEDGEMENT

The authors would like to express their gratitude to

all the members of wind resistant design committee in Japan Road Association. The discussion at the committee was essential for preparing the paper as well as the Manual.

REFERENCES

where,

Ur: reference wind speed for flutter and galloping

Erl: correction factor for the effect of gust and

rapid growth of flutter or galloping

Er2: safety factor

6.2 Vortex-induced vibrations

The following inequality should be satisfied.

Ucv>Ud (12)

In case that Ucv ^Ud, the amplitude A should be ex- amined in view point of safety of the bridge, fatigue and

serviceability.

7. OTHERS

Besides the wind resist?.nt design method of bridge

deck mentioned above, thr Manual describes the design

methods for structural nembers such as towers and

cables, and the design methods for bridges at their erection stages.

Davenport, A.G. 1978, "Wind structure and wind

climate",

Safety of Structures under Dynamic Loading, Trondheim, TAPIR, V.l, pp.209-256.

Davenport, A.G. .1982, "The interaction of wind and

structures', Engineering Meteorology, Amsterdam,

Elsevier, pp.527-572.

Hanshin Expressway Public Corporation 1984, Proposed

Design Rule on Dynamic Effects of Wind on Bridges,

(in Japanese).

ICE 1981, Bridge Aerodynamics, Proposed British

Design Rules, TTL.

Narita, N., Yokoyama, K., Sato, H. and Nakagami, Y.

1987, "Aerodynamic characteristics of continuous

box girder bridges relevant to their vibrations in

wind", Seventh International Conference on Wind

Engineering, Aachen, V.4, pp. 283-292.

19
Field Measurements of Wind Pressures

by

Kishor C. Mehtai

ABSTRACT

This paper describes a field project sponsored by the

National Science Foundation to obtain a reliable data

base of wind induced pressures on building surfaces. A

test building with the dimensions of 30 x 45 x 1 3 ft (9.1 x 13.7 x 4.0 m) and a 160 ft (48.8 m) tall meteorological

tower are constructed in an open field. Wind speed,

wind direction, and building surface pressures are measured simultaneously for a duration of 15 minutes

when wind speeds exceed 20 mph (8.9 mps). The

computerized data acquisition system records data at the rate of 10Hz for each of 40 channels. The

meteorological data and pressure data instruments are

installed and calibrated. Preliminary terrain

characteristics and pressure results are obtained.

Specific goal-oriented data collections are planned for the future.

KEY WORDS: Wind; Pressures; Field Data; Building

1. INTRODUCTION

As evidenced by wind-induced building damages around

the world, many low-rise structures are susceptible to severe damage in windstorms. A greater understanding

of what pressures the wind exerts on a building will assist in mitigating this damage. Toward that end, a

unique facility has been constructed at Texas Tech

University which permits research in wind effects on a

low-rise building in the field.

The facility has the capability to measure pressures on

the building surface at as many as 40 locations

simultaneously. The 30 x 45 x 13 ft (9.1 x 13.7 x 4.0 m)

test building can be rotated, allowing positive control over wind angle-of-attack. Meteorological conditions

are monitored by instruments mounted on a 160 ft (48.8

m) tower located at the site. The test building and the

tower are shown in Figure 1

The research, sponsored by the National Science

Foundation (NSF), concentrates on acquiring a reliable

database of external and internal pressures on the test building. Other preliminary objectives include

characterization of the terrain and verification of the

pressure measuring system. This paper describes the

facility, instrumentation, data acquisition system and

preliminary results.

2. FIELD FACILITY

The field facility has two main components: the building

and the meteorological tower. As shown In Figure 2, the outer test building is mounted on a track for rotation and

the inner data acquisition building is fixed on a concrete

slab. Pressure taps are mounted on surfaces of the

test building while the computerized data acquisition

system is housed in the fixed building.

The facility is located on a flat, open site, about two

miles from the university campus. The city of Lubbock

is situated on the high plains of Texas, elevation 3300 ft (1000 m) above sea level. The terrain is very flat, typically changing elevation at a rate of 5 In 1000. There

are no significant hills or valleys within 20 miles (32 km)

of the city. Most of the land is cultivated, the chief

crops being cotton and sorghum. The flat and open

nature of the countryside minimizes possible anomalies

in wind characteristics caused by terrain. Additional

descriptions of the surrounding terrain can be found in papers by Levitan, et al. [1] and Lakas, et al. [2].

The test building is a prefabricated metal building. It has

smooth skin with no architectural features, making the

building shape as simple as possible. The building has one door and one window. The entire test building is constructed on a rigid frame undercarriage. This

carriage has a wheel in each corner which rests on a

circular steel track embedded in the concrete slab. The

undercarriage rests on the slab except during the

rotation operation of the building. Hydraulic jacks at each corner lift the building (total weight of

approximately 16,000 lb). Motors at two opposite

corners rotate the building at the rate of 5 fps (1 .5 mps).

The building can be rotated a full 360°, providing positive

control of the wind angle-of-attack. Anchor bolts are placed at 1 5° intervals around the circle.

The data acquisition building is a fixed structure of

reinforced CMU construction, 12x12x8 ft (3. 7x3.7 x

2.4 m). The meteorological instruments are mounted on

a 160 ft (48.8 m) high three-legged guyed tow«?r. The

tower is guyed at two levels. 70 ft (21.3 m) and 130 ft (39.6 m). The base of the tower is 150 ft (45.7 m) west

1 Professor of Civil Engineering, Director, Wind

Engineering Research Center, Texas Tech University,

Lubbock, Texas 79409.

20
of the test building. Retractable booms extending 6 ft

(1 .8 m) from the tower are provided for wind speed

anemometers at heights of 13, 33, and 70 ft (4, 10 and

21 .3 m). A reference pressure box is located 50 ft (15.2

m) west of the test building. The box, of dimensions 5 x 4 x 3 ft (1 .5 x 1 .2 x 0.9 m), is placed below the ground

surface. An 8 in. (20 cm) diameter PVC pipe extends

underground from the reference pressure box to the

data acquisition building.

3. INSTRUMENTATION

The meteorological instrumentation consists of four

wind speed anemometers, two wind direction vanes, two

temperature sensors, one relative humidity sensor, and

one barometric pressure sensor. The wind speed

anemometers are the 3-cup type and are located on booms at heights of 1 3, 33, and 70 ft (4, 10, and 21 .3 m)

and at the top of the tower, at the height of 160 ft (48.8

m). Wind direction vanes are placed at heights of 33

and 160 ft (10 and 48.8 m). Temperature, relative

humidity, and barometric pressure sensors are mounted

at the 1 3 ft (4 m) level on the tower.

Differential pressure transducers are used to measure

external and internal pressures on the building. Two

types of differential pressure transducers are used,

thirteen Validyne model DP103-22N-7-S-4-H and twentyeight Omega model PX 163-005 BD 5v. The transducers

have a range of ±0.2 psi (±1.38 kPa). Validyne

transducers have removable stainless steel

diaphragms, and are of the variable reluctance type.

Omega transducers have a fixed silicon diaphragm.

Validyne transducers are more versatile and rugged.

The data acquisition system consists of an IBM PC XT

microcomputer with an internal 20-megabyte hard drive.

Analog voltages from the instruments are converted to

digital form by a MetraByte DAS-8 high-speed A/D

converter. The DAS-8 has a 12-bit resolution and an

input range of ±5 volts. The system can record data at rates of up to 1200 samples per second (total for all channels). The data is streamed directly to a 20-

megabyte Bernoulli removable cartridge drive, thus

providing an easy way to transport the data from the

field to the campus.

The pressures on building surfaces are measured

through 3/8 in. (10 mm) diameter pressure taps mounted

flush with the surface of the building. A 4 in. (102 mm)

long 3/8 in. (10 mm) inside diameter flexible plastic

tubing connects the pressure tap to a three-way

solenoid valve. Ten inches (254 mm) of 3/16 in. (5 mm)

inside diameter flexible plastic tubing connects the

solenoid valve to the differential pressure transducer.

The tubes are long enough to assemble the system in the field and are short enough to provide good response

without distortion. A general setup of the pressure

transmission system is shown in Figure 3. Approximately 35 ft (10.7 m) of 3/16 in. (5 mm) inside

diameter flexible plastic tubing is used to connect the

transducer to the reference pressure pipe in the data

acquisition room. Dynamic response of the system is shown in Figure 4. The mean response values are good

for frequencies up to 100 Hz; however the standard

deviation values are not unity beyond 20 Hz.

Recognizing that distorted responses will give ratios

other than unity, undistorted measurements can be

obtained with this system up to the frequency of 20 Hz.

Details for calibrating the pressure measuring system

are given in a paper by Ng and Mehta [3].

4. CURRENT PROGRESS

A wind tunnel study to determine the influence of the

surrounding buildings on the external pressures has

been carried out. Results of this study are outlined in the paper by Lakas, et al. [2].

Meteorological data have been collected during strong

winds since early 1988. Of the first 63 records

collected, 31 are of stationary winds. Analysis of the 31

records, each of 15-minute duration with mean wind

speeds between 21 and 33 mph (9.4 and 14.8 mps), at the height of 33 ft (10 m), provides values for the power

law coefficient, roughness length, and turbulence

intensity. These values are shown in Table 1 . Detailed

analysis of these data are given in a master's thesis

completed at Texas Tech University [4]. The azimuth zones for terrain surrounding the site are indicated in Figure 5.

Typical recordings of wind speed, wind direction, and

pressure fluctuations, recorded simultaneously, are shown in Figure 6. The pressure values are positive, as

expected, since this pressure tap is in the windward wall. The recordings indicate large fluctuations in wind speed

and direction encountered in the field. Analysis of these

recordings are in progress.

5. CONCLUSIONS

The field facility is constructed and operational. It is capable of collecting data that will assist in understanding wind loads on low-rise buildings. The

facility has the potential for many other types of

research. Topics such as measurement of the total wind force exerted on the building, load sharing and load

path research, and effects of varied building geometry on pressures are possibilities. In addition, research in building ventilation is under consideration.

6. ACKNOWLEDGEMENTS

The research presented here is supported by NSF Grant

No. CES8611601. Eleanora Sabadell and Michael P. Gaus, both of NSF, provide a continued encouragement

to pursue field research. John D. Holmes, Senior

Research Scientist from CSIRO, Australia, is a Fulbright

scholar at Texas Tech during March-June, 1989, to

21
conduct research on this project. His significant

contributions in calibrating reference pressure and

analyzing recorded data are acknowledged.

7. REFERENCES

[1] Levitan, M.L., Mehta, K.C., Chok, C.V. and

Millsaps, D.L., "An Overview of Texas Tech's

Wind Engineering Field Research Laboratory,"

Proc. . Sixth U.S. Conference on Wind

Engineering, University of Houston, Houston,

TX, March 1989.

[2] Lakas, B.D., Mehta, K.C., Kiesling, E.W., and

Dunn, J.R., "Wind Tunnel Simulation of the Texas

Tech Research Building: Effects of Terrain

Features and Building Orientation," Proc. Sixth

U.S. Conference on Wind Engineering,

University of Houston, Houston, TX, March

1989.

[3] Ng, H.H.T. and Mehta, K.C., "Pressure

Measuring System for Wind-Induced Pressures on Building Surfaces," Proc . Sixth U.S.

Conference on Wind Engineering, University of

Houston, Houston, TX, March 1 989.

[4] Chok, C.V., "Wind Parameters of the Texas Tech

University Field Site," a master's thesis

submitted to the Graduate Faculty of Texas Tech

University, Lubbock, TX 1988,90 pp.

Table 1 . Preliminary Terrain Parameters at the Test Site

Azimuth

Zone

(see Fig. 5)

No.

of Records

Power-law

Exponent a

Surface

Roughness

Length

Z0 , ft (cm)

Turbulence

Intensity

at 33 ft (10 m)

270° -70°

70° -160°

160° -210°

210° -270°

24

3

4

0.14 (0.14-0.17)'

0.15 (0.14-0.16)

0.11 (0.10-0.13)

0.045 (1.14)

0.059 (1.50)

0.007 (0.18)

0.17

0.18

0.18

* Range of values

22
Fig. 2. Cutaway view of the test building

Fig. 3. Pressure transmission system

23
(

ST/

i

kNDAR

(

D (

i

EV—IA" nc

I

4

—

1

1

1

H

MEA

i

N

1 V— -i I 1 >—

1—• —

12 5 10 20 50 100

FREQUENCY, Hz

Fig. 4. Frequency response of ths pressure transmission system

SCALE - MILE

Fig. 5. Map of the site

24
LU

rx

C/)

0)

LJJ cc

Q.

e,

o

Hi

LU

£L

(/) a

z

CD

<D z

o

O

LU

CC

Q

O

110 10

TIME (min)

Fig. 6. Wall pressure, wind speed, and wind direction records

25
Transfer of Research Results in Wind Engineering

by

J. Eleonora Sabadell1

ABSTRACT

Brief comments on three types of transfer of

research results will be presented here first, on

the transfer of knowledge in general; second, on

the transfer in the area of wind engineering; and

third, when these exchanges occur between

countries at different levels of development and

resources.

1. TRANSFER OF KNOWLEDGE

Scientific and technical knowledge, in order to be

put to use, has to follow certain processes: it must be generated (research); it has to be

communicated and analyzed (information and

intelligence); then assessed for adoption

(evaluation, scenarios, alternatives); and lastly it has to be implemented into action (utilization,

diffusion) when the innovation is proven

appropriate (Vlachos, 1978).

The generation of new knowledge and the issues

involved in identifying, organizing, managing and

supporting needed research in general, and in

wind engineering in particular, will not be

discussed in this presentation.

Many of the problems that can arise during the

process of communicating knowledge can be

captured into two questions: where does the

information come from, and where does it go.

The different characteristics of the source of

information as well as the material itself, e.g., disciplinary or multidisciplinary, private or public,

basic or applied, to name only a few, are great in

number and complex in nature. The same can be

said about the receptor of this knowledge. What

has been found is that sender and recipient often

work at quite different "wavelengths".

The capabilities for analyzing and determining the

quality and appropriateness of any exchanged

piece of information vary widely among cultures,

disciplines, localities, sectors, institutions, and

levels of government. This can be a substantive

problem, especially for the recipient, when its

capacity is lower than the source. Then transfer

may become dumping.

Next, and when information has been transformed

into intelligence, the adoption process starts. For

this purpose further evaluation is necessary, e.g., by building scenarios for different alternative use

of the new idea, by ascertaining that it will be

useful and timely in solving the specific problem

at hand, by evaluating the costs and benefits of

the different choices. During this part of the

process the capabilities of the organizations and

professionals involved in these evaluations are of

critical importance. Unfortunately, this kind of

expertise is not always available.

The last stage is the adoption of the new idea

and its incorporation into a technique, measure,

regulation or policy addressing a particular

problem. The institutional arrangements

supporting, carrying out and disseminating the

innovation in its various forms, is the key factor

for a successful transfer. The existing variety of

national, regional, and local organizations is large

enough as to make coordinated efforts in the

adoption of innovations not an easy task.

An extra step that is often neglected, is the

formal evaluation of the gains/losses made by

adopting and implementing the innovation.

Feedback on successes and failures is not easy to

obtain and this is probably due in great measure

to the amount of the free flow of information

allowed or possible within and between countries.

2. TRANSFER OF NEW KNOWLEDGE IN

WIND ENGINEERING

Wind engineering is a relatively young and

interdisciplinary area which addresses a variety of

issues, e.g., buildings aerodynamics, structural

systems and materials and wind-load criteria, wind

energy, dispersion of pollution, transport of

1 National Science Foundation, Washington, D.C.,

20550, USA.

26
particulates, airflow over complex terrain, to

name only few. Other emerging concerns are in

the area of social and economic damages

assessment and mitigation, legal and insurance

considerations, risk-benefit analysis (Meroney,

1987).

Wind engineering is also a very good example of

the complexity of a given research field, and of

the difficulties and opportunities investigators face

when transferring the generated new knowledge

to the practitioner.

To begin with, many disciplines are involved in

wind engineering: fluid mechanics, aerodynamics,

meteorology, structural mechanics, architecture,

mathematics, computer science, forestry,

agriculture, economics, psychology, and other

social sciences. The difficulty of communicating

between disciplines is a recognized problem, as it

is to relate researchers within a single discipline

but in different topical areas, e.g., wind

characteristics and wind transport of pollutants.

In the United States the acknowledgment by the

construction and insurance industries, and by

regulatory agencies that results of wind

engineering research should be integrated, or at

least tested, in the design and construction of, and

standards and codes for buildings and structures,

has proven to be laborious, lengthy and often

ignored especially for single-family homes. This

fact is distressing because it has been estimated

that losses from windstorms exceeds three billion

dollars annually (Petak and Atkisson, 1982) and

this figure continues to grow.

On the other hand, death and injuries caused by

wind-related natural hazards are low because

warning systems have improved markedly in the

last two decades especially for hurricanes. In the

case of tornadoes, warning is far from perfect, but

engineered buildings and structures designed and

built following adopted codes and better

techniques can successfully protect their occupants

and suffer limited structural damage. This is not

the case for low-rise and mobile houses and

commercial buildings where most loss of life and

injuries still take place. These are some examples

of how new knowledge has been transferred in

certain areas better than in others.

The wind engineering community is composed of

scientists and engineers whose work is much

better integrated than in most other areas of

inquiry. This fact is probably due to the youth of

the field and to the early recognition of the

complex nature of this very real problem. Even

so, two main groups of researchers have been

operating almost from the start: the wind

tunnel/physical simulation group, and the full-scale

structural behavior investigators. To these two

groups all the other subgroups are attached.

Communication between these researchers has

been quite good but they continue to operate

mainly on their own areas of interest. It is evident

that there is room for improvement in

cooperation and communication, as it is between

the producers and the users of innovations. It has

been found, time and time again, that the results

of research where the user participates in a

project from the start, have a much higher

probability to be adopted into practice. Many

wind engineering specialists have arrived to

another important conclusion, and this is: that

knowledge transferred is constrained by economic,

legal and political factors far more than by the

availability of technical and scientific knowledge.

3. TRANSFER OF RESEARCH RESULTS

BETWEEN COUNTRIES

During the past four decades there have been

many activities in the transfer of knowledge from

countries with advanced state of experience to

countries with less expertise, capabilities and

resources. These transfers have been supported

by governments, international organizations in all areas of aid, as well as financial institutions. The

results have been mixed, with successes where

cultural, economic, educational and other

differences were well understood by the donor or conveyer country, and with failures where the

transfer of knowledge or technology was done "as

is" without any adjustment to the new

environment.

Similarly, any wind engineering technique,

measure, policy or any other innovation generated

by a developed country, located mainly in a temperate zone, with stable population, rich in

resources, with strong human capabilities and an

excellent communication system, can be

successfully transferred to another country that

has limited resources and is often located in a completely different climatic zone of the world,

only if those differences are well recognized from

the start of any exchange, and if the

consequences of any change are carefully

assessed.

Another important point to be made is that

transfer of knowledge does not occurred in only

27
one direction. To the contrary, countries at any

stage of their development have much to learn

from each other.

4. REFERENCES

1. Vlachos, Evan, A Conceptual Model of the

Knowledge Transfer Process, "Water

Knowledge Transfer", by Neil Grigg, Ed.,

Water Resources Publications, Fort Collins,

Colorado, 1978.

2. Meroney, Robert N., Proceedings of

Seminar/Workshop on Wind Engineering,

"The Past to the Future", Colorado State

University, 1987.

3. Petak, W. J. and Atkisson, A. A., "Natural

Hazard Risk Assessment and Public Policy:

Anticipating the Unexpected", SpringerVerlag, New York, 1982.

28
Wind Resistance in Manufactured Housing in

Hurricane Zones

N

G.Robert Fuller, P.E.

1

ABSTRACT

Each year in the United States, significant damage to manufactured (mobile) housing is caused by high winds,

hurricanes and tornadoes. In 1987, 115 deaths

occurred in manufactured housing from wind-related

causes. This paper is a review of recent research and

investigations into the anchorage and installations of

manufactured housing in hurricane zones.

KEYWORDS : Anchorages ; foundations ; hurricanes

,

manufactured housing; tie-downs, wind resistance.

1. WIND DAMAGE IN THE U.S.

According to a study funded by the National Science

Foundation (Wiggins, 1977) wind damage to buildings in the United States had exceeded $2 billion per year by

1976. Estimated annual wind damage in 1988 was $4 to $5 billion. This doubling of cost from 1976 to 1988 was

not only due to inflation, but also due to increased

housing construction, particularly in coastal ands

affected by hurricanes. The cost due to wind damage is

higher than that due to any other natural hazard,

including earthquakes, volcanoes, expansive soils and

floods. Storms and weather kill more people in the U.S.

than any other natural hazard averaging approximately

350 people per year (Gaus, 1985). Wind is one of the

worst natural hazards in the United States [1 ].

Wind caused damage to buildings is most widespread for non- engineered and marginally engineered woodframe

structures, such as housing, motels, stores, shopping

centers, schools and churches. It was because of

extensive damage to these types of structures, that the

American Society of Civil Engineers (ASCE) in 1985

established a "Task Committee on Mitigation of Damages

Due to High Winds." Included in its purpose was that the

committee should "apply recent advancements in building aerodynamics to study this problem." [2]

The Tasks of the Committee were as follows

:

To apply new knowledge in building aerodynamics and

wind engineering to make buildings, especially nonengineered buildings,more resistant to high winds.

. To review wind load related provisions and national

and local building codes to determine their adequacy

.

. To study the enforcement of provisions by building

code officials, and compliance by contractors and

builders.

. To recommend improvements in codes, standards,

and design.

. To prepare reports for publication in ASCE Journals.

Initial results of this Task Committee were reported at an ASCE Structures Congress in New Orleans,

September 1986, at a session on "Mitigation of

Hurricane Wind Damage". Another conference was then

held in Kansas City in November 1987, sponsored by the

U.S. Wind Engineering Research Council and major

Building Code organizations.

Because of the activities of the ASCE Task Committee

and reports submitted to the two conferences, the

National Science Foundation in 1 988 awarded a grant to the University of Missouri - Columbia to study the

response of wood framed houses to high winds and to develop a realistic and practical analytical model that

could be used to predict the response. [3]

As reported in one of the original papers under this NSF

Grant; "Great progress has been made in recent years in understanding the nature of wind damage to buildings and

other structures, and how to mitigate such damage.

It is now known that wind damage caused by hurricanes,

tornadoes, and straightline winds can be greatly reduced

by improving building construction practices, and by

having better tie-down systems for mobile homes and

parked aircraft. The costs associated with such

1 Chief, Manufactured Housing Compliance Branch,

U.S. Department of Housing and Urban Development,

Washington, D.C. 20410-8000

29
improvements are rather modest the potential benefits are high." [1]

Recently, an issue of "Disaster Preparedness Report"

by the National Weather Service, contained a summary

of weather-related hazard fatalities in the United

States. This report reveals that 341 lives were lost in 1987 and 349 in 1986; caused by weather (lightning,

floods, tornadoes, high winds, heat, winter storms,

hurricanes and tropical storms). [4]

Wind-related deaths totalled 1 15 in 1987 : high winds -

33, thunderstorm winds - 23, tornadoes - 59. Of

these, 51 deaths were in Texas; although 22 other

states reported wind-related deaths. May 22 was the

single deadliest day in 1 987 when a tornado killed 30

people in Saragosa, Texas.

Analysis of locations where wind-related deaths

occurred revealed that 30 deaths were in manufactured

(mobile) homes: 24 from tornadoes, 1 from high wind,

and 5 from thunderstorm strongwinds. Manufactured

housing, which accounts for 655 of all U.S. housing, had

4193 of all tornado deaths reported in 1 987 [4].

Each year there are significant losses in terms of

injuries, deaths, and property damage associated with

manufactured homes exposed to high winds. This is

critical when almost a third of all new single-family

homes in America are manufactured housing.

2. WIND EFFECTS ON HOMES [5]

Manufactured home construction is by nature

susceptible to damage by extreme winds. The units are

lightweight and have large surface areas exposed to wind pressures. Single-wide units, 12'0" to 16'0"

(3.7m to 4.9m) wide by 50"0" to 74*0" (15.3m to 22.6m) long, offer little resistance to overturning and

sliding, except that provided by diagonal steel straps

attached to screw-in ground anchors (See Figure 1).

Tornadoes, hurricanes, and "straight" winds can each

create unique problems for manufactured housing. The

sudden high intensity winds from tornadoes account'for

many deaths and injuries each year, and manufactured

homes are frequently destroyed by tornado winds as low as 1 10 mph (176 km/hr). Long duration hurricane

winds produce cyclic loads that cause fatigue in roofing

and wall siding, and can weaken nailed connections and

ground anchors. Straight winds can reach speeds of

65-80 mph (104-128 km/hr) but usually cause only

moderate damage.

External pressures pull walls, roof, and floor apart,

except for windward walls. Drag forces tend to roll or

slide the unit if it is not adequately anchored. Local

pressures also cause material and connection failures at corners anft roof eaves. Complete roof failures may

occur because of wind uplift forces.

However, for the purpose of this paper, concentration

will be on tiedown, anchorage and foundation failures of

manufactured homes. Most manufactured housing is constructed on longitudinal steel chassis beams

supported by piers of unreinforced concrete masonry

units (emu), spaced from 8'0" to 10'0" (2.4m to 3.1m)

on center. Diagonal steel straps run from the steel

chassis beams to steel rod anchors screwed into the

ground. Some units, mostly double- wide homes, have

perimeter unreinforced emu walls with anchor bolts to the exterior wood frame.

Tiedown /anchorage failures range from the homes

shifting off the foundation piers to complete pullout of

ground anchors, allowing the unit to roll or vault in the

wind. A shift off the piers can damage plumbing under

the floor, or the piers punch through the floor. A more

destructive failure occurs when the frame ties or over-the-roof straps break. This failure usually results

in total destruction of the unit, because it is free to roll

or tumble, A similar situation occurs if the ties stay

intact, but if the ground anchors pull out of the soil.

Setting the proper size and type of anchorage is not always a simple operation. Relatively stiff soils are

capable of resisting large pullout loads, but are highly

resistant to penetration by screw-in ground anchors to the required depth. Loose soils may permit easy

installation of ground anchors, but offer little resistance

to pullout. [5]

3. INSTALLATION INVESTIGATION [6]

As a result of recent information collected on the

number of deaths and injuries and amount of property

damage associated with manufactured housing exposed

to high wind forces, the U.S. Department of Housing and

Urban Development (HUD) launched an investigation in

1 988 into the installation of manufactured housing in hurricane (coastal) zones. This study was conducted by

HUD's Monitoring Contractor, The National Conference

of States on Building Codes and Standards (NCSBCS),

with coordination by HUD engineers in the Manufactured

Housing Compliance Branch. A major part of the study

was the on-site inspection of 66 homes in ten coastal zone states (Alabama, Delaware, Florida, Georgia,

Louisiana, Maryland, North Carolina, New Jersey

,

Damage to manufactured homes results from

aerodynamic wind pressures that develop as air flows over and around the unit.

30
Rhode Island and Texas), from April to June in 1988. 4.4 Marriage Vail Connections (Double-vide Units)

:

State and local inspectors were asked to take NCSBCS

and HUD engineers to "better installed" homes to evaluate installation practices considered to be

satisfactory by the installers and local government

inspectors. The purpose of the inspections was to determine not only the typical methods of installation

used, but to understand the interaction between

homeowners, dealers, installers, home manufacturers,

local inspectors, and state regulators. The study

included inspections conducted at manufactured housing

parks, planned subdivision communities, and individual

homeowner sites.

NCSBCS and HUD engineers identified specific

deficiencies observed during on-site inspections,

obtained information from installers, enforcement and

building code officials, and reviewed installation

manuals, ground anchor test data, and manufacturer's

foundation and anchor designs. All this technical data

was then analyzed to evaluate the on-site conditions

observed to determine acceptable practices. A list of

major deficiencies was then developed to be used in future training sessions for building code and State

enforcement personnel. Finally, a "Manufactured Home

Installation Inspection Guidebook" was drafted to be used

as a training document, and also to be used by

manufacturers and designers to prevent deficiencies

from recurring.

4. I NCORRECT INSTALLATIONS

Several deficiencies in the installation process were

discovered, as shown in Figures 2 and 3, and in Photographs *1 through *18. The primary ones are summarized in the following seven categories

:

4.1 Site Preparation

:

Improper water drainage, topsoil not removed,

incorrect placement of footings, lack of vapor barrier,

poor soil bearing conditions.

4.2 Foundations

:

Footings not below frost level; omitted under large

opening column supports, perimeter walls, or ridgebeam

posts; improper construction of block piers; caps or shims under steel beams; unreinforced block piers over 30" (0.8m) high.

4.3 Tiedowns

:

Ground anchors at wrong angle, not in soil to correct

depth, or extending out of ground an excessive distance,

improper details or anchor test data, straps improperly

installed or not tightened properly

.

Inadequate connections at roof,floor or end walls;

improper insulation or weatherproofing to prevent air or moisture infiltration; differential settlement or

unlevel units.

4.5 Mechanical

:

Defects or improperly installed plumbing, drain lines,

heat ducts, or fixtures.

4.6 Electrical Work

:

Improper field installation of fixtures and wiring

connections; transportation damage to fixtures,

appliances, or electrical system.

4.7 Special Site Conditions

:

Inadequate design of foundation for flood plain,

expansive clay soils, or earthquake forces; inadequate

ventilation under homes.

5. CONCLUSIONS:

It becomes obvious from the data collected and

investigations reported herein that there are several courses of action required to improve the resistance of

manufactured housing to high wind forces. All corrective actions will involve specific organizations

and levels of government. These are, in the opinion of

the author, as follows

:

. Research organizations (Universities and consultants);

. Designers (Manufacturers, anchor manufacturers, and

consulting engineers);

. Local government and building officials (Building code

and standards and inspectors);

. Installers and dealers

;

. Federal and State Governments;

5 . 1 Research Organizations

:

Funding by the Federal Government and the

manufactured housing industry needs to continue, to develop analytical techniques for design of manufactured

housing subjected to tornado, hurricane, or other high

wind forces. Risk and hazard mapping development

needs to continue, for incorporation into model building

codes and standards.

31
5.2 Designers:

Manufacturers need to provide complete designs of each

model for resistance to high wind forces, and provide

specific details to show how wind forces are

transmitted through the superstructure (shear vails and

roof and floor diaphragms), to the frameand to the

foundation and anchors. Test data for design needs to be supplied by anchor manufacturers and designers need

to incorporate all information in adequate installation

instructions and requirements.

5.3 Local Government and Building Officials

Uniform inspection and enforcement programs need to be

established by all states and municipalities for the

installation and foundations of manufactured housing.

The American National Standards Institute (ANSI),

American Society for Testing and Materials (ASTM),

and National Conference of States on Building

Codes and Standards (NCSBCS) need to continue to promulgat standards and guidelines for adoption by local

governments.

5.4 Installers and Dealers

These industry representatives need to participate

actively in trade organizations to obtain guidelines,

standards, training, and local or state regulatory

requirements. They also need to provide follow-up

inspection to Assure that homes continue to be level and

are tied down properly . They also are required by the

Federal Regulations to notify manufacturers of

non-compliances with the Federal Standards and defects

in construction.

5.5 Federal and State Governments

State Governments need to develop uniform regulations,

standards and enforcement programs, including onsite

inspections of foundations and installations. The Federal

Government, under the Manufactured Housing Program

administered by HUD, should continue to foster

research, develop guidelines for local governments,

participate in and coordinate work of ANSI, ASTM, and

NCSBCS, enforce the regulatory program for the design

and construction of manufactured housing, and adopt

updated wind design standards, as appropriate.

6. REFERENCES

[Tfliu, H. and Turner, E.J. ; 1 989 : "Wind Damage

Mitigation Strategies for the United States". University

of Missouri, Columbia, MO 6521 1 , Proceedings of Sixth

U.S. National Conference on Wind Engineering, March

1989, Houston, TX.

[2 ] Fuller , G.R., 1 988 : " Anchorages and Foundations

for Wind Resistance of Manufactured Homes," U.S. Department of Housing and Urban Development,

Proceedings of 20th Joint Meeting of US-Japan

Cooperative Program in Natural Resources (UJNR),

National Institute of Standards and Technology , S P 760,

January 1989, Washington, DC

[3] Liu, H.; Gopalaratnam, V. and Nateghi, F; 1 989

"Improving Wind Resistance of Wood-Frame Houses,"

University of Missouri, Columbia, MO Proceedings of

Sixth National Conference on Wind Engineering, March

1989, Houston, TX.

[4] Peters, B.; 1988. "Disaster Preparedness Report,

Natural Hazard Deaths in 1987 in the United States,"

National Weather Service, Southern Region, Ft. Worth,

TX 76102.

[5] McDonald, J.R. and Mehnert, J.F.; 1989: "A Review

of Standards of Practice for Wind Resistant

Manufactured Housing," Texas Tech University

,

Lubbock, TX 79409, Proceedings of Sixth U.S. National

Conference on Wind Engineering, March 1989, Houston,

TX.

[6] U.S. Department of Housing an Urban Development;

1989: "Preliminary Study of the Installation of

Manufactured Homes and Installation Monitoring by

States and Local Governments," National Conference of

States on Building Codes and Standards, Inc. January

1989,Herndon,VA.

32
33
Manufactured

Home

Note: Strap not shown

gure 2 Improper Blocking and Shimming

Manufactured Home

Steel

Beam

Over Extension

of metal stand

Strap not placed around beam,

per manufacturer's instructions,

-Footing of insufficient

size

-placed

on soil containing

organic material or

disturbed uncompacted

soil

-Footings above

frost penetration

Ground anchors not

suitable for the soil

type, with insufficient

depth

Ground anchor not

parallel to strap.

INCORRECT METHOD OF INSTALLATION

gure 3 Incorrect Installation and Foundation

34
* Perimeter Footings

Not Below

Frost Line

* Improper Location

Of Piers

Photograph # 1

* Improper Cell

Alignment

* No Footing Under

Pier

Photograph # 2

* Improper Cell

Alignment

* Absence Of Pier

Cap

Photograph # 3

35
Noncompacted Soil

Underfooting

Inadequate Protection

Of Bearing Soil

Photograph # 5

* Organic Material

Under Footing

* Deteriorating Metal

Stand

Photograph # 6

36
* Missing Pier

Cap

Photograph # 8

Improper Shimming

(without a pair)

Improper Frame

Tie Buckle

Photograph # 9

37
Photograph #11

* Wrong Angle Of

Anchor For Tie

Photograph #12,

38
* lmproper Buckle

Fastening

* Shimming Without

A Pair

Photograph # 13

* Improper Buckle

Fastening

* l-Beam Not Bearing

On Pier

Photograph # J4

*

* Unacceptable Wrapping

Of Frame Tie

* Poor Site Preparation

Photograph # \S

39
Excessive Strap

Angle

Photograph # IG

* Improper Pier

Alignment To

Marriage Wall

* Dryer Vented

Beneath Home

Photograph # 17

40
Theme II

Earthquake Engineering

Behavior of the Nagara Dam During the

Earthquake of December 17, 1987

, by

Yasushi SASAKI 1 ' Tetsuro KUWABARA2 ) Masao NISHIMURA3 ' Ikuji SEKO4 ) Jun KASHIWAGI 5 ' Yoshiki AOKI 6 >

SYNOPSIS

During the Chiba-ken Toho Oki Earthquake of

December 17, 1987/ 27 components of strong

motion records were obtained at the Nagara

Dam of the Water Resources Development

Public Corporation located at a distant of

about 29 km from the epicenter. Hie

behavior of the dam during this earthquake

was analyzed from these records and the

dynamic properties of the dam were

examined. Moreover, observed behavior of

the dam was compared to the result of the

dynamic finite element analysis.

KEY WORDS

Strong motion observation, Record of

strong-motion acceleration, Dynamic behavior

of earth dam, and Dynamic finite element

analysis.

1. OUTLINE OF THE EARTHQUAKE AND THE

STRONG-MOTION OBSERVATION FACILITIES

AT THE NAGARA DAM

1.1 Outline of the Earthquake

An earthquake of JMA Magnitude 6.7 occurred

near the offshore of eastern Chiba

Prefecture at 11:08 AM on December 17, 1987.

According to the Japan Meteorological Agency

(JMA) , the epicenter of this earthquake was located at 140°29' east longitude and 35°21'

north latitude with the hypocenter depth of

58 km. 1 ) The seismic intensity of JMA scale

was V (very strong) at Katsuura, Chiba and

Choushi, and was IV (strong) at Tateyama,

Tokyo, Yokohama, Mi to, Ajiro, Kumagaya and

Kawaguchiko. The range where the seismic

intensity of V felt was almost within 50 km

of the epicentral distance within Chiba

Prefecture.

W 0' 1* IQOkfp/

l

t 0^20 km

< Nagara DamT"" 7 )

\ vr

-A

'

Fig. 1 Plan of the Nagara Dam

Dr., Eng., Head, Ground vibration Division,

Earthquake Disaster Prevention Department,

PWRI, M0C

Research Engineer, Ground Vibration

Division, ditto

3) Visiting Research Engineer, Ground

Vibration Division, ditto

4) Chief, the First Research Department,

Research Institute, WRDPC

5) Head, Fill Dam Division, ditto

6) Fill Dam Division, ditto

43
Damages caused by this earthquake were

reported; 2 persons dead, 161 persons

wounded, 72,698 houses damaged, and the

total amount of damages to the public

facilities was about 6.88 billion yen. The

damage features of this earthquake are reported in more detail in a separate

paper . 2 '

1.2 Outline of the Nagara Dam

The Nagara Dam is a zoned earthfil dam

constructed at the border of Nagara Town and

Ichihara City, which creates the reservoir

of the capacity of 10 million cubic meters.

This reservoir is one of the water

conveyance facilities for supplying water to

the Boso coastal industrial zone and its

vicinities in Chiba Prefecture as well as residents in Chiba City and Kujukuri coastal

region. Plan of the dam and its maximum

cross section are respectively shown in

Fig. 1 and Fig. 2. The 52 m high dam is

constructed on the diluvium sandy layers.

The cross section of the dam consists of

three zones and counterweight fills.

Construction materials for the dam are Kanto

loam (Zone I, impervious zone), sandy soils

(Zone II) and silty mudstone (Zone III).

Wing dams for impervious purposes are located at both the right and left bank

sides, and drains are provided inside the

main Dam.

The dam site is located at a distance of

29 km from the epicenter in the

west-northwest direction. An acceleration

of about 260 gal was observed at the base of

the dam. As a result of this earthquake

motion, maximum settlement of 30 mm and a horizontal displacement of about 20 mm were

observed in addition to hair cracks (maximum

depth of 2 m) at several places in the

pavement on the dam crest. However, this

anomaly was not so severe and would not

affect the safety of the dam. 3 ) Settlement

distribution and location of cracks are

shown in a separate paper. 3 ) Since it was

the stage of test filling to the reservoir

during the earthquake, the water level

during the earthquake was lower (EL.: 57 m)

than the surface of counterweight fill (EL.

60 m) .

1.3 Strong-Motion Observation Facilities at

the Nagara Dam

Three-component strong motion accelerographs

were installed at 10 places as shown in

Figs. 1 and 2 within the dam. 4 ' G-l and G-7 are electromagnetic type accelerometors.

And G-8 to G-10 are popularizing type

accelerometers which were installed on the

concrete base on the dam crest and inside

the water-collecting tunnel. A ten-second

delay circuit was employed for records of

G-l, G-2, G-4 and G-6. All records of each

strong-motion accelerograph are stored on a magnetic tape in analogue form.

Specifications of accelerograph are shown in

Table 1.

Table 1 Specifications for Accelerograph

Transducer

Place Code G-l —C-7 C-8-— G-10

Tr «ntducec Type Electromagnet lc Movable Coll Type Mezoclectr ic Shear Type

Natural Frequency 3.0 fix Approx. 1 kHz

Damping Constant More than 11

Max . Haeeur ing Acceleration

1000 gal 1000 gal

Meaeur ing Frequency

Range

0.3 — 30 Hz 0.1-*-O0 Hz

2. BEHAVIOR OF THE DAM DURING THE EARTHQUAKE

Strong-motion records were obtained during

the event at 9 places out of ten other than

G-9 accelerograph. Since the record is of

analog type, it was converted into the

digital quantity at the interval of

1/100 sec. and was processed by a band-pass

filter shown in Fig. 3. Maximum

accelerations are shown in Table 2. Since

the records obtained by G-8 and G-10 are not

synchronized with others, time was adjusted

for G-8 and G-10 in such a manner that the occurrence time of maximum value of vertical

motion becomes equal to that observed by

other accelerographs. The maximum value at

the base (G-l) was 256 gal in the

u ps tr eam-downs t r earn direction, 135 gal in

the dam axis direction, and 86 gal in the

vertical direction and at the crest (G-6)

,

it was 371 gal in the upstream-downstream

Design High Water Level in. ti unn Crest with Shrinkage Allowance c !. 6 R^

ad Design Crest

Design Full Water Level i' .ii h.tvio x

Water Level when Earthquake occurr

Ground Line in Field

Izumiya Mudstone Layer

Lower Sand Layer

Fig. 2 Cross Section of the Nagara Dam

Upper Sand Layer^ Downstream Rock Impervious

C-1 Sheet

Relief Well'

44
direction, 353 gal in the dam axis

direction, and 326 gal in the vertical

direction. The ratio between maximum

acceleration of G-l and G-6 was respectively

1.4 times (upstr earn-downstream direction),

2.6 times (dam axis direction) and 3.8 times

(vertical direction). Time history of

acceleration in the upsui earn-downstream

direction and the displacement waveforms

determined by numerical integration in

frequency domain at G-l, G-4 and G-6 are

shown in Fig. 4. Maximum acceleration at

G-4, inside the dam, is smaller than that at

the base (G-l) . At the dam crest,

relatively large acceleration values

continue after the main shock compared to

the base . By examining the waveform of the

main shock portion in Fig. 4, it can be

known that a phase lag occurs in the record

of each point and waves propagate from base

to the crest of dam. Distribution of

maximum acceleration in the

upstr earn-downstream direction at 6 points is

shown in Fig. 5(a). Acceleration with the

level same as the maximum acceleration at

the crest occurred at the upstream and

downstream slopes, and the values became

almost half at the roidheight of the

enbankment. Also, the distribution of the

maximum displacement derived from

integration of acceleration is shown in

Fig. 5(b), and displacement orbit of each

0.25 0.333 30 31 Frequency f (Hz)

Fig. 3 Correcting Function of Band-Pass Filter

Table 2 Maximum Acceleration Observed

at the Nagara Dam

Place OOdO

Meaaur ing Direction

Cod.

Ha«laua

Acceleration

(9*11

1 Plan

Code

Heaaur ing Direction

Code

Kaxlaue Acceleration

19*1)

C-l G-l

C-l

X

y

z

2 St US

86

G-4 0-8

G-4

X

Y

1

171 3S3

32C

6-2 C-l

c-:

X

Y

I

2C3

204 123

C-7

C-7 C-7

XY

X

2S7

288 179

0-3 G-3

C-3

XY

t

347 288 122G-S G-S

C-8

X

*

X

497 385

329

G-4

G-4 G-4

X

Y

I

174 ISO

137

C-10

C-10 C-10

XY

t

274

147 109

C-5

G-S

C-S

X

Y

X

380

301 182

° pleasuring Direction Codec

Xi Upatreae-Downetreea Direction

Yi Daa Axis Direction

X: vertical Direction

Meiclaua acceleration is a corrected value after band-pass filter.

.100 -»o -MM

G-e a„

371 eat

0 » 19 IS H

«''»"' h''«

, 'W"m'''h''4«"'«''Vo'''> »!«.!

G-4 ft,

174 (el

o s to is n n so ss «o

,, «s"M''»

, 'M

,, »s

,, Vo''n'

, #o"aV'w"

I 6.0 G-6d,

G-4 d„

wo

JOO G-l a.

2S6 (.l

0 i 10 If To 7* M IS 4b 4S SO SS 40 SS 10 IS SO ts *0 SS 100 I0S (a) Acceleration

4>

G-i d.

0 S 10 is :o ?S 30 JS »0 <s so ss to SS TO IS 40 SS SO 9S 100 10S

(b) Displacement

Fig. 4 Time History of Observed Acceleration and Calculated Displacement

(Upstream-Downstream Direction)

(a) Acceleration

Fig. 5

(b) Displacement

Distribution of Maximum Acceleration and Displacement

(Upstream-Downstream Direction)

45
Fig. 6 Displacement Orbits at Observation Points

record is shown in Fig. 6. By looking at

the displacement orbit, a large displacement

directed from the downstream (north) to the

direction 30° to 45° biased on the left bank

northwest) can be recognized. This

displacement occurred at about 8 sec. as shown in the displacement time history in

Fig. 4(b). This portion of the waveform

gives the maximum horizontal displacement

for all observation points.

3. WAVE PROPAGATION IN THE DAM AND THE

ESTIMATION OF DYNAMIC PROPERTIES OF

EMBANKMENT MATERIALS 5 *

3.1 Natural Frequency of the Dam

Since the record of G-6 is the result of an earthquake motion at G-l which was amplified

within the embankment, the natural frequency

of the embankment can be derived from the

transfer function between both points. The

transfer function is calculated as the

square root of the ratio of power spectrums

of acceleration records between two

observation points. As an example, the

transfer function at the central section of

the dam is shown in Fig. 7 to Fig. 9. In

the horizontal components of the

upstream-downstream direction and the dam

axis direction, no dominant frequency

showing significant amplification is

recognized, and the response magnification

is low as a whole. On the other hand, it can be known that the vertical component has

a high amplification at around 2.7 to

3.0 Hz. Frequencies having large transfer

magnifications recognized between base (G-l)

and crest (G-6) are shown below in the order

starting from the largest magnification.

Upstream-downstream direction (refer to

Fig. 7): 5.4, 1.4 to 1.7, 4.2, 9.0 to 9.6,

10.7 to 10.9 Hz Amplified in a wide

range of frequency.

Dam axis direction (refer to Fig. 8): 1.6,

2.9 to 3.2, 4.6 to 4.7, 6.3 to 6.6, 8, 8.7

to 9.3 Hz ... Amplified in a wide range of

frequency.

Vertical direction (refer to Fig. 9): 2.7

to 3.0, 5.4, 7.7 to 7.9, 13.7 to 14.0 ... There are significant dominant frequencies.

Primary natural frequency of the maximum

cross section was 1.6 Hz, which is derived

from Matsumura's shear vibration solution^*

of a wedge-shaped embankment. From this

solution and the above mentioned dominant

frequency, the natural frequency of the dam

is considered to be 1.6 Hz for horizontal

shear vibration in both upstream-downstream

and dam axis directions and 2.7 to 3.0 Hz

for vertical vibration.

3 .2 Wave Propagation Velocity within the

Embankment

The upward propagation of an earthquake

motion within the dam is seen in Fig. 4.

Then, the propagating time of an earthquake

motion from point (G-l) to the other upper

observing points was determined by the two

methods explained below. In the first

method, cross correlation between time

history of recorded waveforms at two

observation points was taken as shown in

Fig. 10, and then phase lag time which gives

the peak correlation was used as a propagating time (Method 1) , which means the mean propagating time for the overall

waveforms. In the second method, attention

is paid to the pulse showing the maximum

amplitude which is considered to be the same phase at different observation points as shown in Fig. 11, and the time difference

between two observation points of the rising

zero cross time of the pulse is used as the

propagating time (Method 2). The time

derived by the Method 2 is the wave propagating time for a particular phase of

the maximum amplitude. Values of wave propagating time within the embankment

derived from acceleration components a x and

ay in the upstream-downstream and the dam

axis directions in these two methods are

shown in Fig. 12(a). It can be known that

the values of propagating time derived from

both methods almost coincide with each

other. Also, from the phase spectrum of

transfer function between two observation

points, the time difference determined from

the gradient of the phase angle

^/frequency f also coincides with the

propagation time as shown below (refer to

Fig. 7 to Fig. 9)

r =0/(2^/) (l)

Where, propagating time between two

observation points is , their elevation

difference is H, and the unit mass of the

embankment between two points is , then the

shear wave velocity is given by

V, = H/ r (2)

Shear modulus is obtained by

g = p yj ( 3 )

46
G-1 —G-6

5 WI i to it • IV r«outtci f ihii

7* i • i

S 41* B t-

-) - ito —>M /I* / h-1

:in

1 «

-J •

: r: '

Fig. 7 Transfer Function of Acceleration Records

(Upstream-Downstream Direction)

§ i.i

i •

• 01

f.UI

• .•01

. MB

X I'M a

G-1 — G-4

r«ouCiti f inti rntouChCi r i nil "«0l(<i f inl>

iiii

rncaucxi r mn

Fig. 8 Transfer Function of Acceleration Records

(Dam Axis Direction)

FM0UOCT f IHfl

a. •»

G-4— G-6

fXOKMCI f IK*>

rMOuCKf f mil

tr1—

a

I ; S

*.•>

• 01

G-1— G-6

rwouttft r imi

fAh-- vtyifc

Fig. 9 Transfer Function of Acceleration Records

(Vertical Direction)

47
c

o

4J

(0

l-l*u ec 0O O

w «-< w «-< O 4> u O

u o

upper Row

Method 1 Lower Rav; Method 2

G-2

0.21 0.19

G-3

0.17 0.14

U«c )

Opper ftowi Method 1 Msthod 5

Time Lag (sec.)

(a) Wave Propagation Time r

(sec)

Fig. 10 Cross Correlation Coefficient of

Acceleration Records (Upstr earn-Downstream

Direction, Method 1) a

300

200

100

0 -100

-200 -300

(GflU

:G-6 J ax

""ill I III nil iiiImiiiiiiiImi iim ill in i ill iilim mull lilt mill I err I 7.2 7.3 7.4 7.5 7.6 7.7 7.8 7.9 8.0 8.1 8.2

Upper *».- Hethod 1 Lower luwi Method 2

G-6

121 137 G-2Q G-30 G-4

129 ( 159 ( 450

142 S 193 { 386

G-

(b) Shear Wave Velocity Vs

(GfiL)

»i"""'" inililniiiitlntminlnumtilinnimlit milium iiiliiiiiinil

)

Upper »o»i _ , Sx Method 1

fa 6

Lower Rawt ... Method 2 261 335 G-4

3620

2660

G-20 G-3

275 C 356

333 \ 525

G-30

(

.s." L

I iiiilm iIiiiiiiiiiIiii iiniilii iiiiiil.ii.inm mil I . err

)

7.2 7.3 7.4 7.5 7.6 7.7 7.B 7.9 6.0 8.1 8.2

Fig. 11 Wave Propagation

(Upstream-Downstream Direction, Method 2)

Vs and G which are derived from the above

are shown in Fig. 12(b) and (c)

respectively. As shown by this Fig., the

average shear wave velocity for the whole

embankment determined from the records in

the upstream-downstreare direction at G-l and

G-6 is about 210 m/sec. On the other hand,

the average shear wave velocioty for the

upper portion of the dam determined from the

records in the upstreara-downstream direction

at G-4 and G-6 is about 120 to 140 m/sec.

This difference is thought to arise from the

difference of the strain level in the

embankment and confining stress level as stated later.

3.3 Dynamic Properties of the Dam Materials

Seismic prospecting which was made after the

earthquake gives a formula which enables to

estimate the shear wave velocity and the

shear modulus as a function of the

overburden pressure. ^ These are shown as vertical distribution curves in Figs. 13 and

14. These figures also indicate the results

obtained by the methods 1 and 2 from the

records of acceleration in the

upstream-downstream direction. The results

determined from the records of earthquake

motion have the values 20 to 30% lower for

V, and 40 to 50% lower for G compared to Vso

Opper »ow! 'y Method 1 Lover Ravi Method 2

G-6(

522

J 522

i 1010 912 G-2

226

226

O G-30 G-4 0j G-SO

( 284 ( 2660 \ { 904 >

( 257 ( 1610 )S 1300

)

(c) Shear Modulus

G-1 («9l/cm»)

G

Fig. 12 Results of Acceleration Records

Analysis between Observation Points

and Go at a very small strain level which was obtained from seismic prospecting.

Especially at the upper half portion (G-4 to

G-6), V s decreased to x/3 to 1/4 and G

decreased to 1/10 approximately compared to

the lower half portion (G-l to G-4). Both

V, and G are close to the results of seismic

prospecting near the upstream slope (G-l to

G-3) and the downstream slope (G-l to G-5)

Shear strain that occurred within the

embankment and the counterweight during this

earthquake will be determined as explained

below from the multiple reflection theory

assuming that the dam is as a homogeneous

semi-infinite layer (uniform shear wave propagation velocity V,

)

. If the velocity

of the earthquake motion observed on the

ground surface is 0 o it) , then shear strain

7ft, z) at a given depth z can be obtained

by

r (t,z) =

l

2V,

Uoit+ j-)- «o(t- ~)|

.(4)

Time history waveforms of shear strains

generated within the dam and counterweight

were respectively calculated by using the

velocity waveforms of G-6 and G-2, and the

maximum value at the depth z was determined. The results are shown in

48
a x Vs (m/s)

0 200 400 by Seismic Prospecting

z(m) z(m) 2(m)

Fig. 13 Distribution of Shear Wave Velocity by

Acceleration Records Analysis

(Upstream-Downstream Direction)

( kgf/cm*)

by Seismic Prospecting

I

II

G0 = 119 z

0 4 G0

= 457 z 04

G0

= 378 z 04

z(m) z(m) z(m)

Fig. 14 Distribution of Shear Modulus by Acceleration Records Analysis

(Upstream-Downstream Direction)

\

: Maximum Shear Strain Distribution

by Dynamic Analysis

I Maximum Shear Strain Distribution

by Multiple Reflection Theory

: Average Shear Strain by the

Maximum Relative Displacement

Tzx {%)

0.1 0.2 0

Tzx (%)

0.1 0.2 0.3 T

Tzx(%)

0 0.1

z (m) z (m) z (m)

Fig. 15 Distribution of Maximum Shear Strain

(Upstream-Downstream Direction)

z (m)

49
Fig. 15. Also in the same figure, the

maximum value of average shear strain is

shown as reference, which was obtained from

the division of the relative displacements

at G-4 and G-6 to G-l by the relative height

of these two points to the base. From the

time history waveform of strain shown in

Fig. 16, it can be known that the strain of

higher than 10-4 occurred repeatedly in the

dam and counterweight fill and the maximum

strain was in the order of 10~3

. By the occurrence of shear strain of this level,

the shear modulus of the dam is considered

to be lowered to about one-half in average

of the initial shear modulus Go estimated

from the results of seismic prospecting.

Especially, a decrease of one order occurred

in the upper half portion of the dam

compared to the lower half portion.

The displacement amplification between the

crest and the base for resonance can be

determined if the embankment material is

assumed to be viscoelastic. By utilizing

this theoretical solution of shear vibration

in a stationary stated , the coefficient of

viscosity can be calculated and, from the

viscosity constant of the material can be

calculated.

Displacement amplification factor becomes

2.25 (=6.00/2.67) from the maximum

displacements at crest G-6 and base G-l

(refer to Fig. 5(b)), and the coefficient of

viscosity is determined from the following

equation:

2.4048 ^

u 7 (5

7 zx of the Counterweight

Strain Time History by Velociy Record on

the Surface (G-l ^ G-2) V t = 130 m/s

o. jo CIS

0. 10 0.01

0.00 •COS

-0. 10 -0. 13

• 0. JC -C.11 1 .21 m

V 'id' IS 20 »' 30 IS «i so ss to os io n to is to »s 100 ios

7 zx of the Embankment

... Strain Time History by Velocity Record on

the Surface (G-4 % G-6) V, = 120 m/s

0 I i c IS 70 JS 10 IS «0 «S SO SS SO OS 10 ss 10 OS SO SS 100 I OS

7 zx of the Embankment

... Strain Time History by Velocity Record on

the Surface (G-l % G-6) Vs = 210 m/s

o.?s see

-:.:s

lO.ise*

lljjlAOl •.a .

where, height H* = 52 m by considering a standard triangular section for the dam,

unit mass of the material is given by p = 1.75xl03 kg/m3

, its shear modulus G = 80MPa

(800 kgf/cm2

), coefficient 7 = 0.12 which is

determined from the displacement

amplification factor of 2.25, and then 7] = 5.6xl0^Pa*s. The coefficient of viscosity

and damping constant h has the following

relation:

TjlO

2G

eh

GT (6)

As the period of the pulse, which can give

the maximum amplitude in the displacement

waveform of base G-l, is given by T =

(7. 56-7.37)x4 = 0.76 sec, and if this value

is used as T then h becomes h = 0.29. On

the other hand, from the pulse at crest G-6,

T - (7.80-7. 53)x4 = 1.08 sec. so that h = 0.20 is obtained. In the damping constant

calculated as stated above, it should be

noted that the recorded displacement

waveform is not the one at resonance. From

this theoretical relation between the

deformation of the embankment during resonance and its viscosity coefficient, the

coefficient of viscosity of V = 5.6 MPa's

and the damping constant of h = 0.2 to 0.3 are obtained for the embankment material.

20 2} SO SS 40 <S SO SS 10 SS SO SS 10 OS SO SS 100 I OS

Fig. 16 Time History of Shear Strain by

Velocity Records on the Surface

(Upstream-Downstream Direction)

BEHAVIOR OF THE DAM ESTIMATED BY

FINITE ELEMENT ANALYSIS

4.1 Input Data

7)

f s

1! l\ 1-4- II !

I

1 •

1 i

ii i ,

i

Fig. 17 Finite Element Model

In order to estimate the behavior, stress

and deformation of the dam during the event,

two-dimensional finite element analysis

using QUAD-4 8 ) was performed for the maximum

cross section shown in Fig. 2.

The model for analysis is shown in Fig. 17.

Counterweight portion was limited to 55 m from G-2 point in the upstream side, the

number of nodes is 359 and the number of

elements is 351. The observed record for 30

seconds at G-l point; (horizontal and

vertical directions, refer to Fig. 4 for

50
Table 3 Dynamic Properties used in Dynamic Analysis

Properties

Zone

Shear Modulus

Shear Modulus in Small Strain G_

StrainDependent of

Shear Modulus

Poisson '

s

Ratio Damping Constant

, (7.32-e) 0.6 G„ = 100-—: '—am

0 1+e

4.0x10

o 4.0xio" 3 +r

0.40

h - 0.15 + 0.25 4.0x10 + 7

II (2.17-e) 0.4 G = 1100- : —om 0 1+e

1.3x10 0.35

0 1.3xl0~3+7

h » 0.25 + 0.25

1.3x10 +7

III G0 = 330i^-.m °- 1.0x10 0.35

0 1.0xl0~ 3 +7

h - 0.18 + 0.25 1.0x10 +7

Counterweight

'o ^S2^L. om

o.e 4.0x10

-3

0.40

0 4.0xl0~ 3 +7

h - 0.15

4.0x10 +7

+ 0.25

horizontal motion) was input uniformly to

the bottom boundary of the analysis model.

Since there were no test values for the

dynamic properties of the embankment

materials, they were determined based on the

published results of studies performed on a similar or the same kind of material.

However, the seismic prospecting was

conducted near the surface of the dam after

the earthquake as stated before and the

velocity of an elastic wave was measured

down to the depth of several meters in each

zone, so that the results of this

measurement were also taken into account.

Dynamic properties used in the analysis are shown in Table 3.

As shown in Table 3, dynamic shear modulus

Go is given by function of confining

stress (mean principal stress a m ) and

density (void ratio e) . Therefore,

incremental embankment analysis was made

before the dynamic analysis for determining

static stress and deformation. And Go calculated from the results of the above

analysis was given to each element. The

properties of the material used in the

incremental embankment analysis were determined from the quality control test

results during embankment work and the

strength test results on undisturbed samples

of materials after completion of

embankment. The dissipation damping between

the dam and foundation was assumed to be

0.25, which is somewhat large because the

foundation was a sand layer, the difference

in physical properties between the sand

layer and the dam is considered to be

relatively small, and frequency

characteristics of earthquake motion are relatively similar between them.

4.2 Results of the Analysis

Distribution of the maximum response

acceleration are shown in Fig. 18. The

maximum response values at each portion

correspond to the maximum values of input

earthquake motion. The response

acceleration decreases inside the dam and

increases at the crest and the slopes of the

dam, which agrees fairly well with the

distribution tendency of measured values.

However, the vertical maximum acceleration

at the crest of the dam is considerably

small compared to the observed value, so that the large amplification from the middle

elevation to the crest of the dam is not

well simulated. Fig. 19 shows the response

acceleration time history and spectrum at

the crest of the dam (G-6 point) . The

waveforms significantly indicate a response

to long period components of input values,

and indicated large responses with a spike-shaped short period which are seen in

observed records do not appear. Similar

results are obtained also at other

observation points. Transfer function of

horizontal acceleration at the crest of the

dam is shown in Fig. 20. In this case,

response is hardly made to the frequency

components higher than 3 Hz, which indicates

the predominance of low-order vibration mode

with a simple deformation shape. Tne

response frequency of the primary mode is

1.6 Hz and almost coincides with the primary

natural frequency derived from the records

of the earthquake motion observed and

Matsumura's solution explained in the

previous chapter. Fig. 21 shows the

displacement time history for the crest of

the dam (G-6) and the inside of the dam

(G-4) observed and analyzed, where both

observed values and analyzed values are the

relative displacement to the base (G-l)

.

The analyzed values are smaller than

observed values at the crest of the dam and

it is to the contrary inside the dam. This seems to occur from some difference between

the properties of the materials of the dam

used in the analysis and those of actual

materials and also from the effect of errors

in the calculating method. However, the

waveforms are similar to each other, and the

analyzed values agreed relatively well with

the observed records inside the dam. From

this, it is considered that analyzed values

which are very close to actual values may be

obtained even for strains or stresses which

occur inside the dam. Fig. 22 shows the

51
G-6 UNIT: 9al

(371.3SS)

(a) Distribution of Maximum Horizontal Acceleration

300

100

0 -100

•200

-300 .400

l i 0

355 gal

I I I 5

i i ISl i i i i l i '

' 3d

5

4 00

300 200

100

0 -100

-J00

-300

.400 -

I i i i i i 0 5

m 88 gal

i > i 10

I I I < IS

I I I I I I I I I I If

200

5 10 FRE0UEIICT F I HZ I

o.s

s 10 FREOUEHCT F IHZ)

(b) Vertical Direction a. (a) Horizontal Direction a

x Fig. 19 Analyzed Response Acceleration of the Dam Crest (G-6)

52
Vibration

Mode

1.000

1.100

1. 000

I.JO0

1.000

0.500

1st

2nd

3rd

4 th

1.00 10.

X

fMOuCK:

Fig. 20 Transfer Function of the Dam Crest (g-6)

Observed

1.5 em

i i i i i i i i i i i 0 5 10

i i i IS

Analyzedeu .• 2.0 cm

I I I I I I I I' I I I t I I I I I I I I I I 0 S 10 IS 20

' ' ' 3<i S

"'A.'jseci

• 4.8 em

Observed

r

1 1 1 1 1 0 s

ICMIAnalyzed

3.1 cm

VV-MM

I 1 0 5

i 1 1 20 » 3'd 5EC '

(a) Inside of the Dam (G-4) (b) Crest of the Dam (G-6)

Fig. 21 Comparison of Relative Displacement in Horizontal Direction

UNIT: kgf/cm 2

(b) Maximum Shear Strain

Fig. 22 Distribution of Maximum Shear Stress and Strain

53
distribution of maximum shear stress and

maximum shear strain obtained by the

analysis. Both values are dynamic values

only due to earthquake motion without

containing static stress and strain.

Maximum shear strain occurs almost in the

horizontal (vertical) direction. From the

results shown in Fig. 22, the vertical

distribution of maximum strain inside the

dam and counterweight is determined and

indicated in Fig. 15. According to this,

the maximum strain is almost of the same

order of the value easily determined by

equation (4) from the records. According to

Figs. 15 and 16, maximum value of shear

strain inside the dam is large and exceeds

10~3. Shear modulus decrease made by this

large strain, reduces the generated shear

stress. Also, in the distribution of shear

stresses, they simply increase from the

surface of the dam to the depth direction

and are slightly concentrated in low

elevation portion of Zone I.

The behavior of the dam estimated by the

finite element analysis stated above has

almost agreed with the behavior clarified

from the records of actual observation.

5. CONCLUSIONS

1) Earthquake motion occurred during this

event can be characterized by a large

pulse of displacement toward the

downstream direction.

2) Primary natural period of the Nagara Dam

is about 1.6 Hz and is almost equal to

the theoretical solution.

3) Shear wave velocity and shear modulus

determined from the observation records

are 20 to 30% lower for shear wave velocity and 40 to 50% lower for shear

modulus than those obtained from

experimental formulas based on seismic

prospecting and laboratory tests.

4) Maximum shear strain generated at the

main dam and counterweight was about

10-3.

5) Damping constant was about 0.2 to 0.3.

6) The behavior of the dam clarified by the

records of strong earthquake motion has

been almost substantiated by the

behavior of the dam estimated from the

dynamic finite element analysis.

ACKNOWLEDGEMENT

The authors would like to express the

deepest gratitude to Mr. Kimoto, Chief of

the Boso Channel Construction Office, Water

Resources Development Public Corporation,

and other persons who offered the recorded

data to the authors and are making efforts

for maintaining strong-motion accelerographs

for observing the precious records of the

strong-motion earthquake.

i

REFERENCES

1) Japan Meteorological Agency: Outline of

the Earthquake Occurred in December 1987

(in Japanese) (Jan. 1988).

2) Sasaki, Y. : Damage by the Chiba-ken

Toho-Oki Earthquake, 21th Joint Meeting,

Wind and Seismic Effect, UJNR (May,

1989).

3) Kimoto, E. , Yamazaki, A., Fukushima, R.

,

Seko, I., and Tanaka, H.: Situation of

the Nagara Dam for Boso Channel in the

Chiba-ken Toho-Oki Earthquake (in

Japanese) , Tsuchi-Tc—Mizu, No. 73, pp.

89-95 (Jan. 1988)

4) Seko, I., Kashiwagi, J., Saito, A.,

Aoki, Y. , and Kogo, K.: Observation

Records at the Nagara Dam for the

Chiba-ken Toho-Oki Earthquake of

December 17, 1987 (in Japanese), Report

of Fill Dam Division, No. 8704, the

First Research Department, Research

Institute of Water Resources Development

Public Corporation (Jan. 1988).

5) Sasaki, Y., Nishiroura, M. , and Kuwabara,

T. : Dynamic Properties of the

Embankment Materials of the Nagara Dam

estimated from the Strong Motion Records

during the Chiba-ken Toho-Oki Earthquake

of December 17, 1987 (in Japanese),

Technical Memorandum of Public Works

Research Institute, No. 2702 (Jan. 1989).

6) Matsumura, M. : Deformation of Earth

Embankment during Earthquake (in

Japanese), Journal of Research No. 28,

Public Works Research Institute (Oct.

1934)

7) Kashiwagi, J., Aoki, Y., Tsukamoto, K.

,

and Kogo, K. : Dynamic Behavior of the

Nagara Dam during the Chiba-ken Toho-Oki

Earthquake (in Japanese) , Symposium on

the "Behavior of the Ground and Earth

Structures during the Chiba-ken Toho-Oki

Earthquake", pp. 205-212, JSSMFE (Jan.

1989)

8) Idriss, I.M., Lysmer, J., Hwang, R.

,

Seed, H. Boiton: QUAD-4:A Computer

Program for Evaluating the Seismic

Response of Soil Structures by Variable

Damping Finite Element Procedures,

Report No. EERC73-16, University of

California, Berkeley (1973).

54
Analysis of Structural Response Records

During the Whittier Narrows,

California, Earthquake, October 1, 1987

A. Gerald Brady 1

ABSTRACT

The extensively instrumented building at 12440

Imperial Highway in Norwalk, California, was excited by horizontal ground accelerations

peaking at 0.2 to 0.3 g during the M5 . 9 Whittier

Narrows, California, earthquake, October 1,

1987. Specific frequencies corresponding to

lower frequency modes are chosen from the

accelerograms and their spectra. Refiltering to remove all frequencies lower than each chosen

frequency, integrating for displacements, and

separating out translations, torsions, and inplane bending, permit graphic evidence of the

amounts of each of these motions that are present at each modal frequency. The

instrumentation was intended to provide just

this type of engineering information.

KEYWORDS: Analysis; Building instrumentation;

Combined modes; Structural response.

1 . INTRODUCTION

The Whittier Narrows earthquake, October 1,

1987, magnitude 5.9, triggered 52 stations in

the National Strong Motion Instrumentation

Network operated by the U.S. Geological Survey

for both the USGS and other federal and local

agencies (Figure 1). These 52 stations

represent 362 data channels (Etheredge and

Porcella, 1987).

2 . NORWALK STATION

The Norwalk building, directly south of the

epicenter and 15 km distant, was extensively

instrumented prior to the current USGS project

for instrumenting special structures. It is a 7-story moment-resisting steel-frame building,

instrumented jointly with Bechtel Power

Corporation, the owner. Dimensions are 141m x 41m x 35m high. There are 27 channels from

accelerometers within the building, and 6

channels at two ground level sites 30m and 13lm

from the building.

Horizontal peak ground accelerations were 0.2-0.3g at these free-field sites, 0.16-0.2g in

the basement, and .41g at roof level.

3 . PURPOSE OF THIS EXTENSIVE INSTRUMENTATION

When extensive instrumentation is installed, it

is usually designed to investigate a few

important modes at lower frequencies, and

occasionally, special dynamic behavior.

Accelerograms from significant events provide

this data. Resonances in fundamental modes are often clear, after the strong input ground

motion is over (Figure 2). During the strong

motion, transients in higher modes are also

present, sometimes concealing the contribution

from lower modes.

Both lower mode resonances and higher mode

transients may be the important, damaging, or destructive motions. The instrumentation here

should show both, since it includes the

basement, 1st and 2nd floors for damaging

transients, as well as the 5th and 8th levels.

The accelerations during this event were not

sufficient to cause structural damage.

This paper shows how the use of various options

in the USGS processing programs (Converse, 1984) can give a visual representation of the

structural displacements at frequencies close to

natural resonant frequencies. A few of the more significant results are presented.

4. PROCEDURE

We investigate the total structural response

displacements, as represented by the motion at

the accelerometer locations, at specific

frequencies, which have been previously obtained

from Fourier spectra.

4.1 Filter out all content at lower frequencies

than the frequency of interest, and

integrate twice for displacement, which puts

a frequency-squared attenuation onto the

higher frequencies. This leaves

displacements at the chosen frequency as the

only prominent motion.

4.2 Separate the translations from the torsion

and horizontal in-plane bending of the floor

slabs so that each can be looked at alone, even though the low-frequency modes might

contain combinations of these motions

(Figures 3 and 4).

5. AMPLITUDES OF SEPARATED COMPONENTS

Figure 5 shows displacements at four levels, in

a graphic display of the north-south translation

component, of a mode whose frequency was 0.76 Hz

(Table 1). The corner frequency for the highpass filter was 0.66 Hz. Figure 6 shows the

only motion recorded in the east-west direction,

a translation, in which a modal frequency of

2.4 Hz is sought. Both plots use consistent

scales for each time history.

Table 1 lists the modal frequencies for N-S and

USGS, Menlo Park, CA 94025

55
E-W translations, and torsion, measured during

the resonating part of these responses. The

frequencies of these listed modes are closely

proportional to the expected sequence 1, 3, 5.

Transient peaks at the 8th level indicate the

response during strong motion input. Transient

peaks at the basement, downhole caisson and

freefield levels indicate the extent to which

the building amplifies or attenuates the

freefield motion.

Table 2 lists frequencies and transient peak

values for rocking and in-plane bending.

Freefield rocking was not measured. In-plane

bending is not associated with any specific

input patterns. No measure of amplification or attenuation by the building for rocking or inplane bending is therefore available.

Table 1

Resonant Frequencies

and Transient Modal Peaks

Mode: 12 3 N-S Translation

Resonant Freq. (Hz) 0 .76 2.3 3.6

Transient peaks 8th 6 0.4 0.03

(cm) B/C 2 0.25 0.04

GL 2 0.25 0.07

' Translation

Resonant Freq. (Hz) 0 .84 2.4 4.0

Transient peaks 8th 3 .2 0.35 0.06

(cm) B/C 1 .0 0.12 0.05

GL 0 .85 0.18 0.07

The frequencies listed in Tables 1 and 2 are identified in Figure 7. From the transient

peaks in Table 1 the ratio of basement peak to

ground level peak is calculated and plotted

against the corresponding frequency. As can be seen from this figure, there is a trend that

low-frequency modal components be amplified, and

high-frequency components attenuated.

6. CONCLUSIONS

Options available in the AGRAM programs permit

plots to be drawn that show clearly the amount

of each modal component (translations, torsion,

in-plane bending) at each modal frequency.

Torsion

Resonant Freq.

Transient peaks

(y rad)

Note: 8th:

B:

C: GL:

(Hz) 1.08 1.4 3.9

8th 57 65 4

B/C 18 35 4

GL 26 5

8th level or roof.

Basement level

Caisson bottom level

Ground level, 30 m and 131 m from

the building

Table 2

Among the modes studied, there is a tendency for

low-frequency modal components (less than 2 Hz)

to be amplified (by 1 to 12 times), while higher

frequency modal components (greater than 2 Hz)

to be attenuated (by 0.6 to 1 times).

7. REFERENCES

Etheredge, E. , and R. L. Porcella (1987).

Strong Motion Data from the October 1, 1987

Whittier Narrows earthquake. U.S.G.S. Open-File

Report 87-616.

Rocking and In-Plane Bending

Mode: 1 Rocking

Resonant Frequency (Hz) 1.1

Transient peak (y rad) 10

-4

4

In-Plane Bending

Resonant frequency (Hz) 0.76 2.4

Transient peak offset

at center (cm) 8th 1.8 0.8

B 0.3 0.15

5.5

Converse, April (1984). AGRAM, a series of Note: 8th: 8th level or roof

computer programs for processing digitized B: Basement level

strong-motion accelerograms, version 2.0.

U.S.G.S. Open-File Report 84-525.

56
Whittier Narrows Earthquake

Peak Horizontal Ground Accelerations (g)

118° 15" 118P OCT 117* 45'

Figure 1. Epicenter of October 1, 1987, M5.9 Whittier Narrows earthquake.

Peak ground accelerations at Norwalk, 15 km south of epicenter, are 0.24 and

0.29g.

North

1st Mode 1st Mode 2nd Mode

Strona Motion 10-12s later Strona Motion

Freq.: 0.76 Hz Freq.: 0.76 Hz Freq: 1.6Hz

Modal Deflection* - cm.

Figure 2. Estimates of 1st and 2nd mode N-S translation mode with combined inplane bending of the floors, scaled from the original accelerograms prior to

digi tizing

.

57
I

TX4W>L/tTtC\/

Tcrr*L 2>/SPc ftC&Aet/T^

AyB.C

Bl

I

A2 r- —

Al - 81 -ci

A2 - -C2

C3 JVsrts. 2>o£~

IN- ££>V2v*Vt?

AS -C3 - -283

TORS/O*/ : A - C

W-PL/wr 8£*&'*J$: 3~i-(4 + c)

Figure 3. Grouping of sets of recorded channels

for investigations of 4 different motions: N-S

translation, in-plane bending, E-W translation,

and torsion. Transducers are placed at the

basement level, the 1st, 2nd, 5th, and 8th (or

roof) floor levels, and in a foundation caisson,

9.1 m below basement level.

Figure 4. Plan view of an instrumented floor

with the N-S motions (A, B, C), split into

separate translation, torsion, and in-plane

bending

.

58
tyOfTtf-SOUTH T#4AJSC/}7-/Ors/ >0-66#*

oct87 14 42 00 040 SECONDS

Figure 5. N-S translation displacements after filtering the relevant

expressions and integrating twice. Displacements at 8, 2, 1, and Basement

levels are shown.

-0-35 i i i i i i i i 1 1 i i i i i i i i i i i 2 4 6 8 10 12 14 18 18 20

Sri.

0

I I I I I I I I I I I I I I I I I I I I 2 4 6 8 10 12 14 16 18 20

2nd

Figure 6. E-W translation displacements, filtered at a frequency designed to

show 2nd mode, at 8, 5, 2, 1, and Basement levels.

59
2

1

yoA/

0

0

A/-S B&*2>.

Rock.

3

J

4- 5 6

(«*)

Figure 7. Building vibrational modes whose basement and downhole motion is

amplified or attenuated with respect to the freefield motion. Modes are identified by their frequencies. There is a tendency for the low frequency

modes to be amplified, and the high frequency modes to be attenuated.

60
Behaviours of Dams

During Chiba off Earthquake

by

Norihisa MATSUMOTO*1

, Nario YASUDA*

2

[SUMMARY]

The paper describes the behaviours of dams during

December 17, 1987 Chiba off earthquake. The magnitude of the event was 6.7 on the Richter scale. Houses

and public works in Chiba Prefecture suffered damages

due to this earthquake. We made investigations on the

eleven dams which were located within about 30 kilometers from the epicenter. Surface cracking on the

dam crest was found in two earthfill dams. However,

cracks were limited only in the shallow depth. There

were relations between the foundation geology and

behaviours of dams.

[KEYWORDS] Dam, Earthquake damage, Foundation

geology, Chiba off earthquake

1. INTRODUCTION

The earthquake of a magnitude of 6.7 on the Richter

scale occurred on December 17, 1987, with the hypocentral depth of 58 km, at 35° 21' North latitude and

140° 29' East longitude, according to the Japanese

Meteorological Agency. The paper describes the behaviours of dams during this earthquake.

2. INSPECTION OF DAMS IMMEDIATELY AFTER

THE EARTHQUAKE

Inspection of dams were made at forty four dams

immediately after the earthquake. These dams were

all higher than fifteen meters and located in the areas where the Japanese Seismic Intensity Scale was equal to

or greater than 4, and/or recorded peak acceleration was

larger than 25 gals.

As a result of inspections, unusual behaviours were

reported at five dams as shown in Table- 1.

3. FIELD INVESTIGATIONS

We conducted field investigations on eleven dams as

listed in Table-2. These dams include not only dams

in Table- 1, but also those dams located in the areas within thirty kilometers from the epicenter where the

damage of earth structures were mostly observed.

Yamakura Dam listed in Table- 1, was excluded from our

field investigations, since the unusual behaviour of

Yamakura Dam was cracking of the abutment of the

bridge in the reservoir not but the damage of the dam.

The dams within the thirty kilometer radius from the

epicenter were all fill dams. Therefore, Kameyama Dam,

a concrete gravity type dam was added to the investigation, although its epicentral distance was greater than

thirty kilometers. Kori Dam being farther than thirty

kilometers from the epicenter were also added to the

investigation, since the acceleration records were obtained at this dam. Locations of the dams were illust- rated in Fig. 1 , which also shows the outline of the area geology. Of eleven dams investigated, three—Kori,

Nagara, and Konakagawa—were built on foundations of

the Quaternary, and the other eight dams were on the

late Tertiary.

We also made investigations on damaged river em- bankments of Tone river which are also earth structures

similar to fill dams.

3.1 Investigation of dams

3.1.1 Kori dam

Maximum cross section of the dam is shown in Fig. 2. The foundation consists of Kasamori formation and

Jizodo formation of the Pleistocene of the Quaternary

period, and the impervious core of the dam was founded on the mudstone of Kasamori formation. The core was

built with weathered mudstone, and the sands were used

for random materials. The downstream slope were protected with grass, and furnace slugs were used for

upstream riprap.

Caretakers of the dam were cleaning the boat in the

reservoir at the time of earthquake, and they told us

that they had not felt much shaking during the earthquake. Accelerometers were installed at three locations,

the dam crest, middle elevation and the base. Although

the accelerations in the dam axis direction at the crest and the base were not recorded, the up-down stream

direction and vertical component were recorded. The

peak value in the horizontal and vertical direction at the base was 21 and 13 gals, respectively, and the one at the crest was 62, and 177 gals, respectively.

Since the instrument at the foundation malfunctioned

*i Dr. of Eng., Head, Filltype Dam Division, Dam Department

Public Works Research Institute, Ministry of Construction,

Japan

*i Research Engineer, Filltype Dam Division, Dam Department Public Works Research Institute, Japan

61
during the main mothion, the peak values at the foundation were less than the real ones. Leakage were measured by using the weir at the

downstream toe of the dam. Leakage was the same

value of 152 litre per minute before and after the earthquake and no change was observed at the dam.

3.1.2 Kameyama dam

The section and elevation of Kameyama Dam are shown in Fig-3. The geology of damsite consists of a

series of mudstone and sandstone of Tertiary age. The

sandstone outcrops on the higher elevation of the left alutment. The staff was in the dam management office

when the earthquake occurred and they felt big shaking.

But, nothing on the desk overturned in their office. Uplifts and leakage are measured two times a month at Kameyam Dam. Fig. 4 shows the leakage and uplift

before and after the earthquake. Uplift No.8 and No. 14

increased shortly after the earthquake, deceasing after- wards. No damage was observed at Kameyama Dam.

3.1.3 Katsuura Dam

Fig. 5 shows the section of Katsuura Dam. Sndstone

known as Kiyozumi formation of Miocene, Tertiary age

comprises the foundation rock. Kiyozumi sandstone

which underlies the Kurotaki uncontormity is a hardened sedimentary rock. Weathered mudstone and loam

were used for core materials and mudstone for random

materials as shown in Fig. 5. Although four sccelerometers were installed at the downstream slope, no record was obtained due to malfunctioning of the instruments.

Wet portions were found at the downstream berm.

Since these wet areas gathered moss, they seemed to have no relations with the earthquake. No damage was

observed at the dam.

3.1.4 Onjuku dam

Typical cross section of the dam is shown in Fig. 6. Bedrock of the site is comprised of mudstone of the

Pliocene, Tertiary age. Weathered mudstone and loam

are used for zone I, i.e. impervious material, and mudstone is used for zone II, i.e. semi-pervious material. The

water level of the reservoir was 30 to 40 cm below the

full reservoir level, when the event occurred. The leakage measured at the weir of the downstream increased

just after the event, but returned to the previous value in a few days. The normal leakage amounts 40 litre per

minute. No damage was observed at dam.

3.1.5 Nakuma dam

Fig. 7 shows the elevation and section of Nakuma

Dam. The foundation consists of alternation of thin

mudstone and sandstone of the Pliocene of Tertiary age. The core material is built with weathered mudstone and

loam, and randum zone with mudstone. The upstream

slope was protected with flexibly connected concrete

blocks. Leakage is not measured at the dam. No damage

was found at the dam.

3.1.6 AzumaNo.ldam

The section of the dam is shown in Fig. 8. The

foundation rock consists of mudstone of the Pliocene of

Tertiary age. Mudstone is used for random zone and

weathered mudstone and loam for core materials.

Neither change of leakage nor damage was found due to the earthquake.

3.1.7 AzumaNo.2dam

The plan is shown in Fig. 9. Azuma No.2 Dam is very

close to No.l Dam and the geology of the foundation is very similar to No.l dam. Neither damage nor change of

leakage was observed.

3.1.8 Arakine dam

The section of the dam is shown in Fig. 10. The

geology of damsite is almost the same with that of

Azuma No.l Dam or No.2 Dam. Impervious core was

built with loam and westhered mudstone, and random

zone with mudstone. Although accelerometers were

installed, no record was obtained due to disorder of the

instruments. The water level was 20 to 25 cm below

the normal pool level which is equal to the crest elevation of non-gated overflow spillway. According to the

talks of the operationnal personnel, the wave of 1 meter

height occurred during the earthquake, and some amount

of water overtopped the spillway crest. The leakage

measured at the downstream drain varied from 62.4 liter/ min before the earthquake to 1 14.0 liter/min just after

the event and to 108.0 liter/min on February 10th,

1988, as shown in Fig. 11. No damage was observed at the dam.

3.1.9 Uryukodam

Section and elevation of the dam are shown in Fig. 12. No damage was found at the dam.

3.1.19 Nagaradam

Fig. 13 shows the typical cross section of Nagara

Dam. The dam was just under the test filling of the

reservoir after completion when the dam is subjected to the earthquake. The settlement was observed at the

dam. Cracking of the asphaltic concrete of the dam

crest pavement and opening of the concrete block of the

spillway were observed at Nagara Dam.

Fig. 14 illustrates the geologic formation of damsite.

There is Kazusa group below El. ±0m, and is Shimousa

group above it. Both groups belong to middle Pleisto- cene age. Kazusa group is comprised of mudstone

(El.-45.0m below) and of sand (El.-45.0m above).

Shimousa group which exists higher than El.±0m consists of medium to fine grained sand (El ±0m to 20m)

and of 2 to 3m thick mudstone at about El. 23m. Above

62
it, there are sands of Jizodo and Narita formation.

Number of blow of SPT of these sands exceeds 50, if they are not weathered. The minimum number of blow

of SPT was 30 in the foundation sands of Naga Dam.

Loam is used for impervious Zone I, and silty mudstone is used for outer shell Zone III, and sands are used

for Zone II. The water level of the reservoir was very low when

the event occurred and it was about El. 57.0m which is approximately equal to the elevation of the surface of

upstream counterweight. The leakage from the drain

increased from 44 liter/min before the earthquake to 106 liter/min 30 minutes after the earthquake, and re- duced to 45 liter/min 24 hours after the earthquake.

The increase of pore water pressures was 0.1 kgf/cm2

at the maximum.

Cracking of asphaltic concrete of the dam crest was

observed at four locations, all of which coincide with the

location of piezometer hole. This implies that cracking

is related with the stress concentration around the hole

of asphaltic concrete pavement. The Water Resources

Development Public Corporation (referred to WRDPC

hereafter), the owner of Nagara Dam, excavated the

exploratory pit to investigate the nature of cracking.

Opening of the crack was so small that nobody could

recognize it without very carefull observation. The

direction of cracking was almost perpendicular to the

dam axis. According to this investigation, cracking

proved to reach two meter depth at maximum. Later,

cracked portion was removed and backfilled with- the same impervious material.

The maximum settlement in the dam body was 30 mm at the locations of middle elevation of upstream

slope and left abutment as shown in Fig. 15. 700 mm

camber had been provided for Nagara dam. The amount

of settlement due to the earthquake is much less than

the amount of total embankment consolidation ex- pected for the dam.

The embankment settled several centimeters due to

the earthquake at the contact with the spillway concrete

block, while the settlement of 1 to 2 cm had already

occurred before the earthquake. The opening of construction joint of the spillway block amounted 95 mm

at the top just after the earthquake, although the joint

had already been open to some amount before the earthquake. Ten accelerometers with three component sensors was placed at Nagara Dam. Peak acceleration at the

dam base was 262 gals, and at the dam crest was 369 gals.

Accelerograms recorded at Nagara Dam contain long

period component of 1.0 Hz as well as short period component. Assuming that the natural period T(sec) of

shear wave vibration is given by T=4H/VS , where H is

thickness of the layer in meters and Vs is shear wave

velocity in meter per sec, T=4H/VS = 1.0 sec, if H=125m

(dam height plus the depth of the sand layer) and Vs =

500m/sec.

3.1.11 Konakagawa dam

Section of the dam is shown in Fig. 16. Several

cracks were found at the dam crest paved with asphaltic concrete. As shown in Fig. 17, there were three cracks

in the up-down stream direction near the left abutment,

and three cracks in the dam axis direction, which are 2

to 3 mm wide. The concrete block for the upstream

protection was displaced 2 to 3 cm at the contact

of concrete block and the crest pavement. The slope of

the concrete block protection work changes abruptly at

El. 36.94m, and concrete blocks are separated 15 to 16 cm at this elevation in the central part of the dam.

Furtheremore swelling of the protection work was

observed in the lower part of this area.

3.2 Investigation of river embankments

River embankments were suffered damages such as cracking, slide and settlements due to Chiba off earthquake in and around Sahara City. Most of damaged

embankments were located where the foundation is comprised of the old river bed deposit, and SPT value

of the foundation is 10 from surface to 5 m, and ranges

from 10 to 20 to the further depth of 10m. There

were liquefaction craters in the nearly paddy field. Hitachi Tone river embankment was subjected to 50 cm

subsidence. This embankment was also located in the

old river bed. STP value was 1 to 2 up to 5 meter depth,

and 30 at maximum from 5 to 9.5 meter in depth. The

portions of the embankment also consisted of a ex- tremely loose sand.

4. CONCLUSIONS

1) There were no major damages of dams due to 1987 Chiba off Earthquake.

2) Unusual behaviours were observed at Nagara and

Konakagawa Dam.

At Nagara Dam, the foundation motion seemed

to be larger than other sites, because firstly

there are thick sand deposits at the site foundation, and secondly the site is closer to the main

earthquake area. Hair craking occurred at the

creast in the direction perpendicular to the dam

axis. The depth of cracks ranged 1 to 2 meters

and the crack did not pass through from upstream to downstream surface.

Cracking and swelling occurred at Konakagawa

Dam presumably because the upstream slope

had an abritmt change. The same type of damage

had also occurred in the similar dams due to 1983 Mid Japan Sea Earthquake.

3) Damage of river embankment in Tone River were

caused by liquefaction of loose sand deposit.

The difference of damage between river embankments and embankment dams is mainly due

to the difference of foundation condition.

4) Acknowledgements

63
We wish to thank River Bureau of Ministry of

Construction, River Division of Chiba Prefectures

and Water Resorces Development Public Corporation for their help to our study.

REFERENCES

1) N. Matsumoto, N. Yasuda, "Behaviours of Dams during 1987

Chiba Off Earthquake", Dam Engineering, No.25, 1988.

2) M. Ohmori, Y. Mizuyama, M. Horiguchi, "Kanto Area",

Kyoritsu Press, 1986 (Geologic map was recompiled by

Wakisaka based on this book).

3) Water Resources Development Public Corporation, "Investi- gations on Nagara Dam Related to Chiba Off Earthquake",

1988.

4) J. Kashiwagj. M. Aoyama, K. Tsukamoto, K. Kogo, "Analysis

of Nagara Dam during Chiba Off Earthquake", Symposium

on Behaviours of Foundation and Earth Structures during

Earthquake", Japanese Geotechnical Sosiety, 1989.

5) Tone River Project Office, MOC, "Damage of River Structures", 1988

64
Table- 1 List of dams reported unusual behaviours

Name of Dams Type of Dam Height (m) Unusual Behaviours

Arakine Earth Fill 33.5 Increase of leakage

Azuma No. 1 EaithFill 24.5 Small failure of road slope near dam

Azuma No. 2 Earth Fill 21.0 Small failure of road slope near dam

Nagara Earth Fill 52.0 Settlement, and cracking of asphaltic concrete pavement of dam crest

Yamakura Earth Fill 22.5 Cracking and displacement of abutments

of bridge near dam

Table- 2 Investigated Dams

Name of Dams Type of Dam Height (m)

Year of

Compretion

Owner

Epicentral

Distance

Kori Earth Fill 38.2 1972 Chiba Prefecture 53

Kameyama Concrete Gravity 34.5 1981 Chiba Prefecture 38

Katsuura Earth Fill 29.0 1971 Chiba Prefecture 32

Onjuku Earth Fill 23.5 1978 Onjuku Town 23

Nakuma Earth Fill 18.0 1976 Irrigation District 20

Azuma No.l Earth Fill 24.5 1977 Ohara Twon 21

Azuma No.2 Earth Fill 21.0 1985 Ohara Twon 20

Arakine Earth Fill 33.5 1978 Chiba Prefecture 21

Uryuko Earth Fill 19.0 1951 Irrigation District 10

Nagara Earth Fill 52.0 1985 WRDPC 28

Konakagawa Earth Fill 19.4 1946 Irrigation District 24

Table-3 Peak acceleration at Nagara Dam

Location G-l G-2 G-3 G-4 G-5 G-6 G-8

X 262 208 353 180 382 369 497

Direction* Y 138 270 288 151 298 354 386

Z 86 124 124 139 180 324 328

Accelerometer PTK130 PTK130 PTK130 PTK130 PTK130 PTK130 SMI OA

•Direction X: Up-down stream direction

Y: Dam axis direction

Z: Vertical direction

65
1 I I I

I 1 50 40 30 20 10 0

Distance in Kilometers

Fig. 1 Location of Dams and Local Geology

H Accelerometer

Fig. 2 Section of Kori Dam

66
Drain holes

Fig. 3 Section and Elevation of Kameyama Dam

4.0

.sE

V

00 I 3.0

J3

—°o

o

Nov./87 Dec/87 Jan./88 Feb./88 Mar./88

Earthquake

^- No.6

/-No.8

^-No.14

Nov./87 Dec/87 Jan./88 Feb./88 Mar./88

Earthquake

Fig. 4 Leakage and Uplift in Kameyama Dam

67
8.0 m

Crushed Gravel

Curtain Grouting

Fig. 5 Section of Katsuura Dam

7.0 m

\ EL. 66.25 m

Curtain Grouting

Fig. 6 Section of Onjuku Dam

Fig. 7 Profile and Section of Nakuma Dam

68
EL. 75.5 m

8.0 m

Curtain grouting

Fig. 8 Section of Azuma No. 1 Dam

Fig. 9 Plan of Azuma No. 2 Dam

EL. 78.9 m

Fig. 10 Section of Arakine Dam

69
120

100

>—

3

C

"s| 80

u

<L> ooAM«j a 60

40

9

5. 9

1

>

DO OQ

O ( > 0

Nov./87 Dec/87 Jan./88 Feb./88 Mar./88

Fig. 1 1 Leakage in Arakine Dam

112.0 m

Fig. 12 Section and Profile of Uryuko Dam

G-6, C-8 (79.5, 80.2)

G: Accelerometer (Elevation)

Fig. 1 3 Section of Nagara Dam

70
El. 20m

El.Om

Narita Formation

(Sand)

Jizodo Formation

(Sand)

Mudstone

Shimousa Group

Kongoji Formation

Kasamori Formation

(Sand)

Kasamori Formation

(Mudstone)

> Kazusa Group

1-14 m?^mm<D®nmf&

15 &my^it&!dz>&T

71
! !

Fig. 16 Section of Konakagawa Dam

Inclined Outlet Works Dislocation and Swelling

Fig. 17 Location of Cracks in Konakagawa Dam

(

72
Survey for Estimation of Seismic Damage in

South Kanto Area

BY

Hideaki Oda*l and Shigeru Tamura*2

SUMMARY

"Outline of Emergency Countermeasure Activities

against Earthquake Disasters in South Kanto Area

" was agreed and made in public by the Central

Disaster Prevention Council on December 6, 1988,

along with "the Survey for Estimation of Seismic

Damage in South Kanto Area" which was conducted

for the use of reference in making up the

Outl ine. This report aims to present the result of the

Survey such as the surface acceleration, fire

damages, damages to buildings and personal

injury with the degree of the imminency of a great earthquake occurrenm in South Kanto Area

and the present earthquake disaster countermeasures in Japan for background information.

KEY WORD: Survey for Estimation of Seismic

Damage, Earthquake, Earthquake Disaster

Countermeasures, Outline of Emergency Countermeasure Activities against Earthquake Disasters, South Kanto Area, Fire Damage'

1. BACKGROUND OF THE SURVEY FOR ESTIMATION OF

SEISMIC DAMAGE

1. 1 Earthquake Occurrence in Japan

Japan is located in the cacum-Pacif ic seismic

zone where the crustal movement is the most active and said to be one of the most famous seismic countries in the world.

While the Japan archipelago and the surrounding

continental shelves amount to only 0.1% of the

total area of the world, the energy of the

earthquakes emitted from there holds about as much as 10% of what the earth generates in total

In the vicinity of the Japan Archipelago the

four plates of the Pacific, the Philippine Sea,

the Eurasian, and the North American exist bordering one another, and on these interplates

a great earthquake is likely to break out as a result of the particular concentration of their

strain energy.

There are many types of earthquakes such as the

Great Kanto Earthquake and the Tokai Earthquake

along the Pacific coast which are though to break out periodically.

Along the Pacific coast, some types of great

earthquakes has occurred periodically such as

the type of the Great Kanto earthquake and that

of the Tokai Earthquake.

The former broke out in 1605, 1707 and 1854. and

the latter in 1703 and 1923 (Fig. 1). In the South Kanto Area (consisted of Tokyo

metropolitan, Kanagawa, Chiba and Saitama prefecture ;hereinafter the same) three types of

great earthquakes which would cause serious

damages are expected to break out, the type of

the Great Kanto Earthquake, referred to above,

that of the Earthquake occurring right beneath

the South Kanto Area and that of the Earthquake

occurring off the Boso Peninsula with the mutual

contact among the Continental Plate, Philippine

Sea Plate and the Pacific Plate (Fig. 2).

Since this region is the political and economic

center of Japan and population as well as various functions have been greatly and densely

accumlated, the resultant damages are likely to

be tremendous and in a wide range taking

secondary disasters such as fires into account.

1. 2 The Present Earthquake Di saster Countermeasures in Japan

Earthquake disaster countermeasures in Japan

have been promoted on the basis of the Disaster

Countermeasures Basic Act (1962) and the Basic

Plan for Disaster, a guiding principle adopted

by the Central Disaster Prevention Council.

In addition, lessons learned from the San Fernando of Earthquake of February 1971, as to re- quired countermeasures against earthquake disaster on a larger city of modern building structures, the Central Disaster Prevention Council

has adopted the "Essentials of Earthquake Countermeasures for Larger Cities" in May 1971. The same council has also decided in August

1975, and in May 1983, on "Contemporai y Promotion of Urgent Disaster Prevention Countermeasures". The present disaster countermeasures are being

promoted in accordance with those laws and plans

With respect to the emergency countermeasures

against earthquake disaster, upon the breaking

*1

Director, Earthquake Disaster Countermeasures

Division, Disaster Prevention Bureau, National

Land Agency

*2

riiief, Earthquake Disaster Countermeasures

Division, Disaster Prevention Bureau, National

Land Agency

73
out of a disaster, or when such is anticipated

the national administrative organs, local

governments and public cooperations will proceed

to take various emergency countermeasures to prevent or control such a disaster.

Should an eartquake occur, the muncipality will

at first establish a Municipal Headquarters for

Disaster Countermeasures to execute emergency

operations.

If necessary, the prefectural government would

establish its Headquarters for Disaster Countermeasures.

In the meantime, National Land Agency will make

the assessment of it<; gravity, and if deemed

necessary will call up the Meeting of Ministries

and Agencies related to disaster countermeasures

for exchanging pertinent information.

Should the need be recognized, the national government would establish the Headquarters for

Major/Extraordinary Disaster Countermeasures,

in accordance with the Disaster Countermeasures

Basic Act, and proceed to excute a comprehensive

disaster emergency countermeasures.

Concerning the promotion of earthquake prediction, since the proposal, in 1964, of an earthquake prediction plan by the Geodesy Council (an

advisory organ for the Minister of Education),

a systematic research on the earthquake prediction has started.

In 1969 the Coordination Committee for earthquake Prediction was established in the Geographical Survey Institute, the Ministry of Construction, for facilitating exchange of information among various research organizations and

for the composite assessment and judgement on such information.

This committee has designated 2 areas of intensified observation, and 8 areas of specified

observation where the earthqudKe prediction is being explored.

A special emphasis is being placed on the improvement or the enchancement of earthquake prediction at the Tokai Area.

In 1976 the Headquarters for Earthquake Prediction Promotion(of which secretarist ib established in the Science & Technology Agency) was established within the Cabinet for further promoting a comprehensive and systematic execution

of related policies.

These systems for promoting earthquake prediction are as shown in Fig. 5.

1. 3 Outline of Emergency Count e rmeasure Activities against Earthquake Disasters in South

Kanto Area

For the earthquake countermeasures in the South

Kanto Area, improvement and development of the

system have been made according to the Disaster

Countermeasures Basic Act, Basic Plan for Disaster Prevention prepared by the Central Disaster

Prevention Council, operational plans for disaster prevention prepared by the designated

administrative organs, local plans for disaster

prevention prepared by the prefectures and

Essentials of Earthquake Countermeasures for

Larger Cities determined by the Central Disaster

Prevention Counci 1. However, in the event of occurrence of an earthquake causing a disaster of extreme severity

extensively in the South Kanto Area, close cooperation is required among the ministries and

agencies responsible for the individual activity

of emergency measures in order to perform them more efficiently and smoothly.

The Outline of Emergency Countermeasure Activities against Earthquake Disasters in South Kanto

Area thus was formulated according to the purport of the "Contemporary Promotion of Urgent

Disaster Prevention Countermeasures" decided by

the Central Disaster Prevention Council to provide the outline for four activities of information, transport, medical service and first-aid

which are specified as to be examined urgently

among other activities for extensive and comprehensive disaster emergency countermeasures to be

followed, by the Headquarters for Extraordinary

Disaster Countermeasures and other related

organs over the prefectural boundaries.

This outline of activities is based on the

survey for estimation of seismic damage.

Besides, the contents of outline of activities

is as shown in Table 1, and flow chart of

government activities in Fig. 6.

2. SURVEY FOR ESTIMATION OF SEISMIC DAMAGE IN SOUTH KANTO AREA

2. 1 The Beginning

The Survey for Estimation of Seismic Daiwe in South Kanto Area was conducted in order to know,

should there occur an earthquake of an intensity

similar to that of the Great Kanto Earthquake

which caused the largest damage in the past, although the possibility of occurrence is not

imminent, what phenomena would likely to occur

and what extent of damage be caused in South

Kanto Area which it is considered would be greatly affected in such event and thus to serve

for furtherance of earthquake disaster counter- measures including preparation of the outline of

emergency countermeasure activities and improvement of the accuracy of plans for earthquake

disaster prevention.

Such estimation of damage is of course very

difficult as there are many aspects involved

which are not yet academically clarified such as behaviors of ground in an earthquake, effects on structures and mode of extension of fire.

74
But, upon the presently available knowledge, the

damage estimation was made by a unified method

over the South Kanto Area. This damage estimation is in no way to predict

what seismic damage is to occur and, as a matter

of fact, its result has but to have some allowance according to the difference in the preconditions as well as the estimation method.

But, it is expected that it will serve for

raising disaster prevention awareness of the

inhabitants concerned, and examination and promotion of the earthquake disaster countermeasures by the administrarive organizations

concerned.

Damage estimation was made with 7 panels established through cooperation of the government,

ministries and agencies and other related organizations and discussion at the "South Kanto Area

Seismic Damage Estimation Survey Committee" in National Land Agency, and the survey period was

fiscal 1981 to fiscal 1987.

2. 2 The Premise

The hypocenter of the assumed earthquake was Suruga Bay (Dislocation model by Kanamori and

Ando, 1973), and scale was magnitude, 7.9, and

conditions at the time of occurrence were 3 cases in Table 2.

2. 3. The Outline of the Result

2.3.1 Surface acceleration

Based on the conditions of damage in the Great

Kanto Earthquake, the values of bedrock acceleration were set up by the distance from the hypocenter, and the value of surface acceleration

was calculated for each mesh (1km by 1km) by

multiplying the amplification factors of 18 ground types.

The values of surface acceleration thus assumed

are as shown in Fig. 7, the area giving a surface acceleration equivalent to a seismic intensity of "6" or higher accounts for about 36

percent of the whole area.

2. 3. 2 Liquefaction

Obtaining the PL valuo of each of the meshes of

68 ground types having sandy soil distributed

within a depth of 20m, possibility of liquefaction was estimated for each mesh according to the result of the study by Iwasaki et al. The result of estimation of the possibility of

liquefaction is as shown in Fig. 8, with the

land along the rivers, reclaimed land along the

coast and dune area giving higher possibility of

1 iquefaction.

2. 3. 3 Damage to buildings

(a) Wooden buildings

Taking the cases of earthquake disaster in the

past for reference and classifying th r buildings

into 4 types according to the time of construction and the number of stories, damage due to the shock was esimated from the deformation obtained for each m^sh by input of the surface

acceleration and that due to liquefaction was estimated with reference to the damage in the

Ni igata Earthquake.

Fig. 9 shows the result of estimation of the

damage to wooden buildings.

"Heavy damage" accounts for 4.7 percent, and

"half damage" for 5.1 percent, but those buildings of "heavy damage" which go to collaspe will

not be much.

(b) Non-wooden buildings

Damage due to earthquake shock was estimated by

comparing the prevailing earthquake-resisting

performance of the existing buildings with the

required earthquake-resisting performance

obtainable from the seismic movement and

building property.

Also, in reference to the damage in the Ni igata

Earthquake, damage due to liquefaction was

estimated. . The result of estimation of the damage to non- wooden buildings is shown in Fig. 10. "Heavy damage" accounts for 3.5 percent, and

"half damage" for 3.7 percent, but those buildings of "heavy damage" which go to collaspe are

very few.

2. 3. 4 Fire damage

With the cases of earthquake disaster in the

past taken for reference and corrections made

for the time and season, the number of fire cases was set up for each mesh, and in consideration of the fire fighting capacity, the number

of spreading fires leading to fire of urban area was estimated.

Then, the spread areas were assumed, with the

spreadabi 1 ity (incombustible zone rate) and stop

of fire from spreading by road, etc. taken into

consideration for each mesh.

The result of estimation of the fire damage is as shown in Fig. 11 through 13. The burnt area is 2-6 percent, the number of the

households in the burnt area is 1,220 to 3,770

thousands, and that of the buildings destroyed

is 980 to 2,600 thousands.

2.3.5 Lifeline facilities damage

Based on the cases of earthquake disaster in the

past and in consideration of earthquake ground

motion, liquefaction and fire spreading, damage

was estimated.

The result of damage estimation is as shown in Table 3, as a whole, electricity has the function obstructed for 21 43 percent, water supply

for 32 percent, sewerage for 15 percent and

damage to the subscribers' telephone accounts

for 14-37 percent.

2. 3. 6 Personal injury

Based upon the data of the earthquake occurring

75
after the Great Kanto Earthquake with involving

death of more than 300 persons, the relation between the number of buildings destroyed and

burnt down and the personal injury (killed and

wound) was set up for estimation of the personal

injury (Table 4). As a whole, 83 to 152 thousand persons are estimated to be killed, and 120 to 205 thousand

persons to be injured,

3. CONCLUSIONS

According to the result of this survey, if a large scale earthquake such as the Great Kanto

Earthquake should occur in the South Kanto Area,

the resultant damages are likely to be

tremendous.

Since the prediction of a earthquake in the

South kanto Area is quite difficult for the time

being, it is especially helpful and useful to promote and sophisticate emergency countermeasures activities systems after the occurrence of

disasters in order to minimize damages, and so the manual about emergency countermeasures

activities was formulated this time.

Of course, it is very important not only to take

emergency countermeasures after the occurrence

of disaster but also to take every possible

earthquake disaster countermeasure beforehand.

Then it is and will be required to implement

such countermeasures effectively with emphasis

place on three points of (1) promotion of disaster prevention programs in urban areas, (2) strengthening disaster prevention systems and

raising disaster prevention awareness and (3) promotion of earthquake prediction .
• Urte EtrtkajMiet •carre* it t<w II cuttry tlosj tic Pacific Cout

' Urtt lirtMaake* vxatH it tkc 19 c«lw» tlM| Ike fscif ic Coast

! Urte (irtMtaket secured it Ike 20 cutm aloet lie Pacific Cout

M: dart f«f eartaejuie

Fig. 1 Hypocenters of Large Scale Earthquakes Distribution Nap in Recent Years

77
w n m m m »3 « k © s s

MttUMg Cxm mot* Dnmm* ftr*nt*nt

as a * o is s

9 a s !i

•5BJS • BBSrffiBtS-BB

Cpp«*«S « Wflftqu*** IHWW***

MaUng fit* fin* miiM rav *

• Q»KSSIta ISBWJWStflCDBII

Snnni up oow«ton*) pim «nd local oUm tar d

5wm| up dhawar pnp

-•«««!*»«

fx:

•bbbmbbb

£B?93?£®Bfl

-•WWWHCBB

"-•sjsaaacDM

Fig. 3

Framework of Earthquake Disaster Countermeasures

Municipal Oiimi ftmniion Council Pnrfcctural Diuun Putmiun Council

<P * IB WBB

Ow*J DrUMtr rNiwBTWioft Council

M«tdqmntn for Mtiaafil*aoidtMry Onatm Coumrrmcuum IMrmutr of

Nl^/PliriiB Mtrtnwrl

man

crown)

Municipal Haadajuawarl (or DrtMMf

Countanaaaaurai (Hud of tha municipafityl

c i n i

(BBIffBSOB)

Pitfcaynjl Haadquanan for CounatwnaaaMW» iPiiwji in y

*!ESOTf5H«n

SUB JJMSB!

Haadquanm tor Paaaajt Counojbmm> in of Dauanaod AdnManoWOrun

and Oni(Md local Ad>Mtirarh«

Onjara

?B«coea

BS BHfflOS

*3ABB

us • Bases

SIB SEW**»J

hauanca and diaaawiinaaion of pnrdicooni and wamtnji.Advt*

rna/Onjtnnf evacuation. Spacrfy.

ing warrwif mm, fancrotncy measures Uia and atoropnation

of land and burldinp. Cipaiabon

ntnmm Fir* Fighting

* 85 O groups ^L**TM OBwtnJMd

LoaJ Public

EmBffBjiCB «

load pn»<fWion, «id.

Imcfftncv mcuum, JEVlSrtS

Enwfcfvcy me«Mim.

OpeiJiion ordkm, Stoute of tuppJin 0*d»

«f. Ek-propriaiian, «c.

KSftBfl

ImtfBvncy m*a»unB oil

BBBCDfiB

6W-KHB

Saoraaa of iiipplia n. «c. Ordan. avproprtauona, ate

Fig. 4

Main Emergency Measures to be Taken upon a Major Disaster

78
C»od*t* Cue*

i ill" ' - " \

nsa5£]ita©J*g x a *

8S»*OTM«5©< S 51 . 1 0StS)

l«l II

r itm

1 . ffl»*S)©8Wt

aacomwtw

I. »l III 11. a**em*

SWI7

BS»?S«8S(S44.4|ftB}

mm Bissau

(S54.8KB)

s a ?

x aj » *K«am

SSM Nil

» ft n

* b as

a±«saiK

smews*

ssaasaas saaaa* sst*h»

X-tP-'JMGaa**

3ia», SSSAWi

nis-iisiBAaax »« a in i i«

ffl«S«D«Jni2>9-

a«3-ft7D-flsafta« laa^WlMl

Owl U»N—I ad hWfcw Dw» Cl

wrasses, ssis-sis iRaa^ssfifSTDfls

•MM O* OM«l KfeMfr. Ml

Fig. 5

Systems for Promoting Earthquake Prediction

79
Table 1

Outline of Eaericncr Countcrasaiure Activities

ereln.t

tsrthquak* Dlsasttrs In South Kanto District

CONTEXTS

Preface

Section 1 Background of Foraulatlon

Section 2 Purpoaea and Cberactere

Section 3 Application of the Outline

Section 4 Anendaent of the Outline

Section S Recognition of Large-scale Earthquake* aa

the Preniaa of the Outline

Chapter 1 GENERAL RULES

Section 1 Governaont-s Activity Syetea

Section 2 Wide-Range Biapateh and Aaaiatlng Activities

Section S Operation of the Dlaaater Relief Law

Chapter 2 INFORMATION AND PR ACTIVITIES

Section 1 Baaic Pollclea for Information and PR Activitiea

Section 2 Jointly Owned Inforaation

Section 9 Inforaation Llaiaon Sratea

Section 4 Publio Announeeaent of Inforaation

Section S PR Activitiea

Chapter J ACQUISITION Or TRAFFIC FOR EHERCENCY TRANSPORTATION

AND EMERGENCY TRANSPORTATION ACTIVITIES

Section ) Baaic Poliolaa for the Acquisition of Traffic and

Easrgsoey Transportation Activitiea

Section 2 Measures for Securing Traffic

Section 3 Eaergency Transportation Activitiea

Chapter 4 RESCUE, FIRST-AID AND MEDICAL ACTIVITIES

Section i Basic Policies for Rescue, First-Ald and Medical

Activities

Section 2 Rescue and First-Aid Activities

Section 3 Medical Activities la Diaaater Area

Seetiea 4 Wide-Range Medical Activities la Rear Base Section 5 Activities of Preventing Ipidoaies

Chapter ( PSCCuTO«NT AND SUPPLY ACTIVITIES FOR FOODS, DRINKING WATER AND DAILY NECESSARIES

Section 1 Baaic Polioiee for . Procuring and Supplying

Activities for Foods, Drinking Water and Daily

Neceaaarlea

Section 2 Basle Rels Allotaent in Coaaodity Procureaent and

Supply Activitiea

Section 3 Coaaodity Procureaent Syatea and so forth

Chapter « EMERGENCY ACCOMMODATING ACTIVITIES

Section 1 Baaic Policiee for Eaergency Accoaaodating

Activitiea

Section 2 Baaio Role Allotaent for Eaergency Accoaaodatlon

Activitiea

Section 3 Aeooaaodatioa In Prefecture or Metropolis in

Disaster Area

Section 4 Accoaaodatlon outsids the Disaster Area

Chapter 7 ACTIVITIES OF EMERGENCY MEASURES FOR LIFE LINE

FACILITIES

Section 1 Basic Policies for the Activities of Eaergency

Heaaures for Life Line Facilities

Section 2 Baaic Rela Allotaent of the Activitiea for

Eaergency Measures for Life Line Facilities

Section 3 Activitiea for Eaergency Meaaurea for Life

Llaa Facilities

Section 4 Precautions for Eaergency Recovery Activitiea

Appendix Table: Jointly Owned inforaation

CO

G

80
Collection

of Information

. Call of .Meetlno, .of galatad Ministries t Aaancljs Establishment .of Headquarters for Disaster Courtax- —aattrae ;bf "Designated' Administrative Orcans. «te.

Conference (or tstiblXshaont of Headquarters So; Extraordinary

' Oliaster Ccante'raeasurss

Cabinet procedures for DoeTTrTcISa , pf Emerqencx Situation fof Disasters

and Establishment o: headquarters

for extraordinary Dmr.i; Countar»easure»

Declaration of .Eaerceney Situation

for Oliaitexa ajid.- Proclamation of Establishment of Headquarter! for Extraordinary Disaster Counterneaauraa

Dispatch of eedquartars Personnel

Establishment of Headquarters for Extraordinary Disaster Counts measures

(So. S lulldlne or Tachlkawa)

equal t»

pWhen Specifically^

L xaeulred J

Local Public Corporatloni

Requests

lapleaentatlon of tha Operations

Pertaining .to Eaaroency Measures of Headquarters for Extraordinary Olsastai Coun terse a sure

s

Headquarters

Centerer.ce Headquarters

Secretariat.

j Determination of I lasle Policies !

• for Eaereeaey

• Measures

I Information

! Collection,

J Arrangement

{and Dissemination

! Comprehensive

! Coordination of

jlsporten.t natter!

! Concerning .

• Emergency

! Countarmaesures

roraulatlon of

"! Transport Plan,

• etc.

j Ceaprehenslv*

• Coordination

! Concerning

• ErnVroeney

! Countermaesoros

Information Communication

Information Communication

Request for Emergency

Transport

Adjustment

Implementation of tha Operation!

Pertaining to Emergency Keasursi

of Designated

Xdalnlatrative

Ortans. etc.

Designated

Administrative

Organs, etc.

I Implementation of j Disaster Emergency! Ceuntermeaeures

Abolition of Declaration

Abolition. of

.

Haadouartsrs for Extraordinary Dlsasta: Countermeasures •••

Abolition of Headquarters

Fic e Flow Chart of Government Activities

81
Table 2 Conditions at time of occurrence of earthquake

Case 1 Case 2 Case 3

1. Season Winter Saae as left Autuan

2. Day of week Weekday Saae as left Saturday

3. Tiee At about 5 P. . At about 2 a. a. At about noon

4. Wind velocity 4a/sec Saae as left lOa/sec

5. Wind direction Saitaaa and Chiba: NW Saae as left S

Tokyo: NNW

Kanag*va:N

6. We* t her Fine Saae as left Saae as left

T.Buaidity 50X Saae as left SOX

8. Tide level AP 1.4a Saae as left AP 1.0a

Fig. 7 Surface acceleration distribution map

82
Fig. 8 Liquefaction possibility distribution map

"otai "navy da«aoa" - iulldlnga partly or wholly tilting

with pillars and b.ajaa haavlly daoaoad and not

oaabla diractly.

"Half da»ag«- - amldlivja having plllara and baana

partly daaagad and walla (ailing off conaidarabla

but yat oaabla.

Fig. 9 Wooden building damage distribution map

83
Fig. 11 Burnt area map (Case 1)

84
Fig. 12 Burnt area map (Case 2)

Lagand

Burnt rata loot a Burnt rata 50%

Fig. 13 Burnt area map (Case 3)

85
Table 3 Damage to lifeline facilities

Ilea Case 1

(evening in winter)

Case 2

(aidni(ht in winter)

Case 3

(noon in autuan)

Electricity 4 2% 2 \% 4 2%

Water supply 3 2%

Sewerage 1 5%

Telephone Z7% 1 i% 3 1%

Table 4 Personal injury

killed (1.000 persons) injured (1.000 persons)

Case 1 Case 2 Case 3 Case 1 Case 2 Case 3

South Kanto Area 1 5 0 8 3 1 5 2 2 0 3 1 2 0 2 0 5

References

(1) the Research Group for Faults (1980) :Active Faults Sheets in Japan, the Publisher of Tokyo University, P13

(2) Kazuo Hamada (1986) Earthquake Prevention, the Publisher of Morikita, P47

(3) Earthquake Disaster Counterraeasures Division, Disaster Prevention Bureau, National Land Agency (1988)

Earthquake Disaster Counterraeasures in Japan

(4) the Central Disaster Prevention Council (1988) :0utline of Emergency Countermeasure Activities against

Earthquake Disasters in South Kanto Area

(5) National Land Agency (1988) rResult of Survey for Estimation of seismic Damage in South Kanto Area

86
Seismic Design Considerations for Offshore

Oil and Gas Structures

by

Charles E. Smith*

ABSTRACT

The ability to determine the vibrational response of

offshore platforms subjected to earthquake ground motions

and to design for the resulting loads has received

considerable attention since the quest for oil and gas . resources extended to seismically active zones worldwide.

It is the intent of this paper to review the major design

considerations for offshore structures that are to be

located in areas of potential seismic risk. Subjects of

major importance relate to area seismicity, soil response,

energy transfer between soil and platform foundation,

platform response, and the consequence to operations

resulting from that response. It is concluded that for

the successful development of oil and gas resources and

other mineral deposits in offshore environments, careful

consideration must be given to the potential earthquake

hazard, and appropriate design criteria must be evaluated

and used to ensure safe and pollution- free operations.

KEYWORDS: Earthquake hazard, offshore platforms,

offshore structures, oil and gas operations, seismic

design criteria.

T. " INTRODUCTION

During the past decade there has been a marked

increase in the exploration and production of

hydrocarbons in seismically active regions of the

world. In the United States much of the effort has

focused on the areas offshore southern California,

and to a lesser extent, offshore Alaska. Both steel

jacket and gravity-based platforms have been proposed

to meet the operational requirements for developing

resources in these areas-*-. The components of a typical

steel- jacket platform are shown in Figure 1, and a general schematic of a typical arctic gravity platform

is shown in Figure 2 . In addition to the traditional

environmental forces of wind, wave, current, and ice

(for arctic areas), a significant consideration in

the design of offshore facilities for seismically

active areas is their structural response to earthquake

ground motions.

However, the comprehensive inclusion of this con- sideration in a design process is a formidable task as shown in Figure 3. Major elements of this task include

the characterization of area seismicity '

, sitespecific response of the soil^' , transfer of

seismic energy from soil to platform '

1

, platform

response to absorbed energy ', and the con- sequences associated with platform response to the

earthquake induced motions^- 1 • ^ . An offshore

facility should have sufficient strength and ductility

to satisfy its intended purpose (drilling and

production) during any anticipated loading condition

without undue expense or risk. Design criteria should

provide an engineer with the necessary procedures and

parameters to enable him to develop a platform system

(superstructure, structure, and foundation) that will

have acceptable performance characteristics during a seismic event.

This paper reviews the major features which should be

considered during the process of developing sitespecific seismic design criteria for an offshore

installation. In addition, ongoing research programs

are presented to illustrate possible future

technological developments . Some of the requirements

are rigidly stated in design standards, such as the

American Petroleum Institute (API) Recommended

Practice 2A(RP2A) 13

, whereas others are based on the

application of experience and probabilistic methods.

It is noted that offshore structures have several key

structural and environmental differences as compared

with typical onshore buildings, for which most of the

experience on earthquake response has been obtained.

Major differences include the following: (1) geology

and response of subsea soils (lack of offshore seismic

data and remoteness of most offshore sites) , (2) a water column that exerts pressure on the soil deposit

supporting the facility and provides a source of

loading and dampening to the structure (effects of

soil-fluid-structure interactions), (3) the special

characteristics of the structure (deep pile

foundation, large mass, large number of redundant

members , large overturning moments , and the fact that

the mass is concentrated at the top as opposed to being distributed over its height as in most building

structures) , and (4) the unusual properties of many

offshore soils. Thus, the dynamic behavior of an offshore platform during an earthquake may be

significantly different from its onshore counterpart,

requiring a different set of design guidelines.

2. AREA SEISMICITY

For offshore structures located in areas subjected to earthquakes, the first step required in developing

appropriate design criteria is a characterization of

the seismicity of the proposed site. The engineer

needs a qualitative description of the seismic threat

which gives both the severity and likelihood of future occurrences of ground motions. As with land-based

structures the time of occurrence, location of faults,

magnitude of future earthquakes, likelihood of ground

failure, (such as liquefaction and sliding) and the

attenuation of ground motions with distance from the source represent major uncertainties in an assessment

of the offshore seismic hazard^.

The difficulties that occur in evaluating the

seismicity of an offshore site differ from those

normally encountered ashore. These difficulties stem

from the fact that the seismic history of offshore

areas is generally less complete than for neighboring

onshore regions. The ability to delineate active

faults or to accurately locate offshore earthquakes

is generally less exact than for onshore events because

of the lack of offshore regional seismograph

networks''--'. Thus, the engineer must rely heavily on

the predictions and extrapolations of seismologists

and geologists in formulating an earthquake source model for a particular offshore site. The procedure

taken in the development of the source model may

range from the extremes of a fully deterministic

approach to a fully probablistic method.

^Technology Assessment and Research Branch,

Minerals Management Service, Reston, Virginia 22091

87
Approaches such as these have inherent risks and are possible sources of unwarranted conservatism. However,

until further knowledge is obtained on the regional

tectonic environment, regional and local geology, and

the compilation of historical records of earthquake

occurrences , these approaches are the only prudent

actions that can be taken.

3. SOIL RESPONSE

Ground motion characteristics can be modified by the

soil properties existing at a particular site as shown in Figure 4, and it is easy to perceive that

offshore ground motions might differ from onshore

motions in several aspects. In fact, the soil response

considerations can be one of the most important factors

in the seismic design of an offshore structure and

its foundation.

The differences between onshore responses and offshore

responses can be attributed to several factors^.

The attenuation factors may be different in saturated

seafloor soils, especially if they are soft and contain

entrapped gas. In addition, there are other variables

which could cause differences in the responses : the

reflection of seismic waves at the water column-soil

interface could interact with site -specific wave forms

to alter their characteristics, the weight of the water

column, acting as a confining pressure on soil mass, could change the vertical component of motion, or local

soil profiles could change the response if the sediments

are wedge-shaped as they depart the shore (the

entrapment or expansion of the seismic waves may lead to focusing or defocusing effects at a given site) . All or part of these factors could contribute to the different

response characteristics of offshore soils

There are two general approaches for formulating the

response of a particular site to a given earthquake.

The soil response could be based on data obtained from

an accumulation of strong motion records or it could be

derived from an analytical method based on a soil

response model, the latter incorporating the soil

properties as determined by field and laboratory tests.

In principle, the accumulation of ground response

records from a number of earthquakes could be used to characterize the response at a particular site. These

records could be scaled to provide response data for

earthquakes of varying intensities. However, because

few records are available on the response of seafloor

sediments , the current accepted procedure is to extrapolate onshore data for use at offshore sites.

Research is being performed, as will be explained in a later section of this paper, to collect and analyze

seismic data taken from actual seafloor sites.

Analytical methods are available to produce response

data at a particular site based on the assessment of

the base (bed rock) excitation. A "shear beam" or "soil column" can be modelled by using a number of

layers which take into account soil properties and

actual soil layer thicknesses.

Selection of appropriate inputs for the base excitation

would come from a seismicity source model as previously

discussed. The results of such an analysis are shown in

Figure 5. It should be noted, however, that this method

of analysis can only provide meaningful results if the

appropriate distribution and nonlinear stress-strain

properties of the soils are used in the calculations.

When selecting ground motion records for use in an

analysis or design, the particular type of information

specified must be guided by engineering judgment. Only

those ground motion parameters which directly influence

the structural response of the structure need to be

considered. Peak acceleration, for example, has little

to do with the response of a structure that has a three

or five -second fundamental period. Spectural velocity

or peak ground velocity, however, appears to be

particularly useful in the analysis of structures with

fundamental periods in the range of typical offshore

platforms 1 '

. 4. SOIL- STRUCTURE INTERACTION

With the soil response defined for a particular site,

the question remains: how is the soil energy transmitted

to the structure. The central focus of this energy

transfer is the coupled behavior between the foundation

elements of the structure and the supporting soil

components. Specifically, the reference here is to the local interactive forces and displacements between

the soil mass and the structural elements.

For offshore platforms, there are two distinct aspects

of the soil-structural interaction process which are inherent to the structure's physical size and mass. One

aspect is the relative significant deformation between

the so-called free-field soil system and the structural

elements of the foundation system. This is especially

true of pile- supported structures where the piles are subjected to large cyclic lateral loads. The other

aspect is the possibility that the dynamic response of a structure -foundation system might result in the

transmission of a significant amount of energy back

into the soil mass. Thus, the existence of the

structure could modify the motion of the soil system.

For instance, in the case of an offshore gravity

structure the calculated dynamic response resulting

from earthquake- induced ground motions can be

drastically reduced by including soil-structure

interaction effects^.

The main objective for developing improved soilstructure interaction models is to improve the

analytical capabilities of the engineer to determine

the vibrational response of structures. The features

of such a model are shown in Figures 6 and 7 . Having

this model , the engineer can evaluate the potential

for structural damage or failure and modify the design

to mitigate earthquake effects. The soil-structure

model should be capable of reflecting the three

dimensional, inelastic, nonlinear, hysteretic, damping,

and cyclic degrading characteristics of the various

components of the soil - structure system. Several

models have been developed for this purpose, but due

to the complex nature of the problem, additional

research is required and is being pursued to further

define the technique^ • ^

5. STRUCTURAL RESPONSE:

The problems associated with the analysis and design

of offshore structures subjected to earthquake forces

are numerous and complex due to their many interactive

components. Consideration must be given to the fact

that these structures are either partially or totally

immersed in water and will, therefore, exhibit

different dynamic characteristics than if they were

in air. Also, due to their large mass, there is a strong interaction during ground shaking with the

soil below and adjacent to their foundation elements.

Figure 8 shows some of the considerations that must

be given to the components in the analysis of a soilpile-structure system. Much information has been

published on fluid-structure interaction and soil

foundation interaction inherent in steel -jacket and

gravity-based offshore structures. Figure 9 illustrates the frequency response of four different

height platforms with the fluid-structure interaction

not included and Figure 10 shows the response of the same platforms considering the interaction.

88
Numerical procedures for carrying out dynamic analyses

of a complete offshore structure foundation system are similar to those for determining the dynamic response

of other types of structures. However, the performance

characteristics of an offshore structure, since it

incorporates a number of unique features not found in

onshore building construction and because of the

magnitude of the environmental load which it must

resist, rely greatly on the inelastic behavior of its

components. Information on the inelastic, nonlinear

structural behavior is essential for a realistic

assessment of strengths and risks associated with an offshore system ' .

The specific algorithm used to describe performance

characteristics of platform elements should be

carefully considered. The inelastic behavior of braces,

legs, and joint subassemblages is very complex and can deteriorate under the severe cyclic load reversals

typically experienced during earthquake ground

shaking Many analytical procedures have

been proposed for use in finite element computer

programs which mimic the postbuckling loadings and

deformations of structural members^"*.

The fluid- structure interaction or hydrodynamic effect

manifests itself in two primary ways during the

earthquake excitation of an offshore structure. This

effect acts as an apparent additional mass due to the

acceleration of the structure and also serve as a source of energy dissipation. It should be noted - that if the structural members are flooded, the

entrapped water mass must be included when calculating

the effective dynamic mass of the structure. The so- called "added mass," is taken in the analysis as being equivalent to the mass of water displaced by

the member projected in the direction of its motion.

Changes in the mass and its assumed distribution can produce important effects on the frequency and mode

shape responses of the structure .

The deck weight of a steel- jacket platform may comprise

only a small portion of the total system weight;

however, due to its position at the top of the

structure, it can have a very important effect on the

vibration properties of the structure. Therefore,

the engineer must consider a reasonable range of deck

weights when investigating the response of a platform.

This is also true for other analytic assumptions, such

as foundation stiffness and effects of marine growth.

Research has shown that the interaction between soil and

foundation elements is a significant factor in absorbing

energy and limiting loads transmitted to the jacket^.

Therefore, because many uncertainties exist in a seismic

analysis, parameter studies should be undertaken to determine their influence on the dynamic response of a particular platform to a series of ground motions. This

may be the only way in which the engineer can estimate

with some assurance, that the structure will perform

adequately during a seismic event^'^.

Because of the large number of interactive systems,

for purpose of analysis the structure is reduced to a complex mathematical model which can best be solved on

a computer. However, analytical methodologies have been

and are currently being verified in the field by

comparing the measured earthquake ground motion

responses of actual platforms to the responses derived

9 ft 0 7 from such analytical models" '. One should,

however, not totally accept analytical results without

understanding the significance of the approximation

incorporated in them. Early studies on the use of the

time history method of analysis illustrate this point.

If a structural analytical model is subjected to several

earthquakes of nominally similar intensity, there is a wide variation in the computed responses. These results

suggest that not only the peak intensity, but also the

duration and phasing of ground motion pulses are significant to the platform's response. For the

designer, an implication such as this means that he

should review his analytical results with caution. His

prediction of the loads and stresses in the structure

based on past earthquake responses, might not match the

reality of responses to future earthquakes^.

6. DECK RESPONSE

A seismic analysis of a typical offshore structure

has two major objectives. First, it must provide

information that can be used to ensure that the

structure as a unit will remain intact during an

event, and second, it must provide information that can

be used to ensure that the equipment and facilities on

the deck will also retain their integrity during the

event. Basically, the structure acts as a filter,

altering the frequency content of the energy being

absorbed at the base of the platform. Figure 11

illustrates the magnification in the vertical frequency

response at deck level of four different height

platforms subjected to the same vertical base

excitation.

Ground motion components having frequencies close to the

natural frequency of the platform are amplified while

those away from the natural periods of the structure are phased out. Thus, the local vertical and horizontal

accelerations to which the facilities and equipment on

the deck are subjected may be significantly higher than

the peak ground accelerations. It is important that

this fact be recognized so that appropriate design

criteria can be formulated to prevent or minimize

damage to facilities and equipment on the deck due to an earthquake that the structure might survive with

little or no damage. The usual procedure is to develop

representative "deck response spectra" based on the

response of the structure to several different

earthquake records. With this spectra, and depending

upon the natural period of the equipment or the

equipment/deck support, either a static, a pseudostatic, or full-modal analysis can be used to determine

required design forces. In essence, the natural

period of the equipment or the equipment/deck

supports determine its response to the seismic event.

If the equipment natural period is in the region of high

energy input from the deck response, there could be a substantial amplification in the vibrational response.

In a like manner, if the natural period of the system is

such that no amplification is caused by the deck

response, the system only undergoes a rigid body

translation, and its motion is assumed to be equal to the deck's motions. Although most equipment has a high natural frequency and would respond statically

to earthquake loading, the designer should be aware

of situations in which the flexibility of deck support

members might increase the systems dynamic response.

For example, a very heavy container supported on a series of flexible beams which is subjected to vertical

excitation or a very tall vessel on flexible supports

which is subjected to rocking motion from the lateral

excitation

.

Because it is desirable to prevent deck equipment

failures from happening before failures occur in the

supporting structure, deck equipment is usually designed

for allowable material stresses proportionally lower

than those used within the structures. It is usually

assured that equipment tie-downs tend to be the weakest

link in the system and, as such, should be designed

for even lower allowable stresses. In general practice,

a one-third increase in the allowable material stress

is permitted when incorporating load conditions,

including those for earthquake- induced motion, in the

design of deck members and equipment with no increase

89
allowed for equipment tie-downs or brackets.

There are, however, two subsystems relating to decksupported facilities, that should require a more

elaborate design methodology. Both cranes and drilling

rigs (due to their long, slender structure and

correspondingly long natural periods) can exhibit

independent natural responses to a seismic event with

corresponding increases in forces and moments within the

structures. Two recent references, one relating to cranes11 and the other to drilling rigs , show the

advantages of using a more refined dynamic analysis of

the system to ensure that members are designed for

realistic load conditions.

A few comments are made concerning safety-ralated

equipment such as fire water lines, well control

equipment, and personnel escape systems. The general consensus is that these items should be configured by

assuming that, even with prudent attention to design

variables such as prescribed loadings and material

strengths, a failure will occur. That is, a failuremode type of analysis should be conducted, with

increased controls and redundancies designed into the

systems to safely handle certain failure situations

and mitigate the resulting hazard to personnel, the

structure, or the environment.

7 . GUIDELINES AND STANDARDS

The American Petroleum Institute (API) , through several

advisory committees, has established requirements for

the seismic design of offshore platforms and these

guidelines are stated in its Recommended Practice

1

-i (RP)2A . The API guidelines employ a dual - obj ective

design philosophy in that a structure must provide

adequate elastic strength and stiffness to withstand

an earthquake of an intensity which has a reasonable

"likelihood" of not being exceeded during the life of

the structure. The guidelines also state that the

platform should have adequate inelastic ductility so that it is able to withstand rare -intense earthquakes

without collapsing, although major structural damage

might occur.

Until recently, the API guidelines gave explicit

recommendations for meeting the strength and the

ductility objectives. The ductility calculation

requirements, involving the calculation of inelastic

deformations and strain energy, have been replaced by

statements recommending good seismic design practices

such as brace configuration and joint details. Several

recent experimental and analytical studies have shown

that approximately the same ductility goals can be

accomplished by following these recommendations.

allowable limits. As a second level of verification,

some platform designs are required to demonstrate

sufficient reserve capacity to prevent collapse, though

not local failures , of the platform under a rare intense

earthquake

In lieu of a lengthy site-specific determination of the

intensity and the characteristics of seismic ground

motion, the MMS permits the use of "defensible"

standardized spectra applicable to the region of the

installation site. The spectra must reflect those sitespecific conditions affecting frequency content and

energy distribution.

When using the "time history" method for structural

analysis, the MMS requires that at least three sets of

ground motion time histories be used. These may consist

of recorded or constructed (synthetic) earthquake time

histories. The manner in which they are used must

account for the potential sensitivity of the structure's

response to variations in the phasing of the ground

motion records. Ground motion descriptions are required

to consist of three components corresponding to the two

orthogonal horizontal directions and to the vertical

direction.

8. SEAQUAKES

Another area of concern, especially for floating

production systems where seismic effect are normally not

considered or for other floating facilities subjected to long-term exposure periods, is a phenomenon called

seaquake. During a seismic event, earthquake energy

propagates within the earth's crust in the form of shear

or compression waves. However, due to the inability of

water to transmit shear, this energy is propagated

through the water column in the form of compression

waves. The literature contains relatively little

information on the propagation of such waves or their

effects on ships or offshore structures. Several

instances have been reported of ships being severely

shaken by seaquakes, with resulting damage indicating

large deck accelerations"^.

Although there is little data available on this

phenomoneon, the possibility exists that earthquakes

may impose just as severe loadings on floating

structures as for structures located on land. Since

the mechanisms behind the characteristics of seaquakes

are not well established, further investigations are necessary to better explain their probability of occurrence and the effects that they could have on permanent floating installations.

9. RESEARCH

The API is currently developing a load and resistance

factor design (LRFD) format for fixed offshore

platforms. As part of this effort, LRFD factors for

fixed offshore platform sites in seismically active

areas will be specified. The LRFD format should provide

a more consistent quantification of the bias and

uncertainities contained in the seismic design process .

The Minerals Management Service (MMS) requirements are aligned with the API guidelines and are published in 30

CFR Part 250 of the U.S. Code of Federal •5 1 Regulations Ji

. The MMS, in verifying the design of

platforms that may experience seismically- induced

loadings, requires that dynamic analyses of the

structures be conducted using earthquake - ground motions

of the intensity appropriate to the site in question.

The motions can be decided by applicable ground motion

records or by response spectra consistent with the recurrence period appropriate to the design life of the

platform. For design purposes, joint and member stresses

from a "strength level" analyses must remain within

The MMS has initiated two research projects

to better understand the effects of earthquake loadings

on the dynamic response of offshore platforms. The

Seafloor Seismic Data Study project is being conducted

by the Sandia National Laboratories to obtain and

analyze seafloor earthquake motion data for seismically

active areas off southern California. The program is

focusing its efforts on the use of the Seafloor

Earthquake Measurement System (SEMS) to collect and

store seafloor seismic events. Figure 12 llustrates

how seismic information obtained in this manner can

be used to evaluate soil and structural response

models where actual field measurements are being

compared to calculated data.

The SEMS concept consists of two subsystems; a seafloor

package and a shipboard unit as illustrated in Figure

13. The seafloor package captures and records seismic

activity, and transfers the data cn-cormand to the

shipboard subsystem via an underwater telemetry system.

The shipboard unit records the telemetered digital data

90
on magnetic media. Situated on the seafloor is the

seismic instrumentation package. The seafloor

electronics package, referred to as the Data Gathering

System (DAGS) , which is connected to a seismic probe

embedded in the seafloor sediments. The probe contains

three accelerometers , configured triaxially to record

the seismic motions, and a two-axis magnetometer to determine the absolute orientation of the accelerometers. The DAGS consists of a microcomputer controller, a data acquisition interface, volatile Random

Access Memory (RAM) , permanent nonvolatile bubble

memory, an acoustic telemetry system, and a battery

power pack. The major attributes of the DAGS include

low power consumption for extended system life; broad

seismic dynamic range - sensing capability; efficient

use of limited digital data storage capacity; and a remote interrogation of the system from the surface.

Low power consumption is achieved by use of CMOS

technology and other low power components wherever

possible. A nonvolatile magnetic bubble memory is used

for long-term data storage (5 years) and is only activated when significant earthquake events are detected.

This bubble memory device has the advantage of being

de-energized for long periods of time without loss

or degradation of stored data.

The SEMS uses low power accelerometers which are capable of measuring acceleration levels from 0.05

milli-G's to 10 G's over frequencies ranging from 0.2

Hz to 20 Hz. Each accelerometer signal is digitized

at a 100 Hz rate by a high-precision data acquisition

subsystem. The digital accelerometer data is placed

into volatile RAM and analyzed by the microprocessor.

The DAGS incorporates earthquake event detection

through the use of pretrigger RAM buffering. The RAM

is used to hold data while a determination is made

whether these data should be saved or discarded.

Short-term and long-term running averages are calculated and used to define the strength of seismic

activity. If the system declares that the current

data correspond to an earthquake event , the data

are stored in the bubble memory. Allocation of the

bubble memory is based on a hierarchal scheme which

retains only the most significant seismic events.

Communication with the seafloor package is accomplished

with the use of an acoustic telemetry data link. The

DAGS can be controlled from a surface computer, thus

allowing data retrieval, diagnostic analysis, and

variability in operating parameters. The surface

instrumentation consists of acoustic transmitting and

receiving circuitry, referred to as the Buoy Repeater

System (BRES), and a controlling IBM PC compatible

computer. The PC serves as a user-friendly interface

and allows direct communication with the seafloor

instrumentation package. Additionally, data retrieved

from DAGS is stored on floppy disks, and the earthquake data can be analyzed and graphically viewed on- board the ship during the interrogation^.

By the end of the year (1989) it is envisioned that two

SEMS units will be deployed offshore southern

California. The two instruments, approximately 150

miles apart will form a seismic array and will provide

an invaluable source for seafloor measurements of

earthquake -induced seismic motions.

As an adjunct study, a project entitled "Seismic

Response Analysis of Offshore Pile Supported Structures"

is being conducted by the University of California, San

Diego, to assess the reliability of current

state-of-the-art computational pile-soil interaction

models used to predict the of energy transmitted to a platform. Seismic accelerograms will be obtained from

the SEMS units and from onboard adjacent instrumented

platforms. Data that Tokai University recently

obtained from an instrumented structure offshore

Japan will also be used in the study . It is

anticipated that these studies will attest to using

present day standards to accurately predict platform

responses and, if necessary, to propose changes to design standards so that they are more reliable.

10. CONCLUSION

This paper was written to provide the reader with an understanding of the key considerations that need to be reviewed in determining the response and performance

of offshore facilities to earthquake induced ground

motion. The simplistic manner in which these con- siderations are presented should not be inferred as indicative of the procedure. Determining the response

of these systems to intense seismic events represents a very complex, and at this time, not fully understood

problem. The references cited should be consulted to acquire a more complete understanding and appreciation

of these aspects and factors.

What is known is that offshore platforms have exhibited

an inherent capability to perform well during

earthquakes. In large measure, this capability is a consequence of the attention to details put forth in

designing structures for the large wind, wave, and

current forces that the platforms must resist. Because

of the extensive research being conducted by both

Government and industry, it is felt that the required

technologies have been and will continue to be

developed to provide safe and reliable operations in

seismically active areas.

11 . ACKNOWLEDGEMENTS

The author would like to thank Dr. Peter W. Marshall,

Shell Oil Company, and Professor Robert G. Bea,

University of California at Berkeley, for their time

and effort in providing technical information and

references used in preparing this paper. Also, to

Dr. Nabih Bedewi , Advanced Technology and Research

Corporation, the author's appreciation is extended

for providing plots and assistance in representing

the example platform excitations. Special thanks are due Richard Giangerelli, Minerals Management Service,

for his timely review and helpful comments. Finally,

I must record my gratitude to Mr. John Gregory,

Minerals Management Service, for providing much moral

support and for being so understanding about unfinished

work during the preparation of this paper.

12. REFERENCES

1. Brian Watt Associates, Inc., "A Study on the

Feasibility of Production, Storage and Loading

Systems in the North Aleutian Basin

(OCS Lease Sale 92)," 284/BWA, Houston, Texas,

September 1985.

2. Shimazaki, K. , and Nakata, T. , "Time Predictable

Recurrence Model for Large Earthquakes," Geophysics

Research Letters, 7, 1980, p. 279-282.

3. Wiggins, R.A., Sweet, J. and Frazier, G.A., "Seismic

Risk for Offshore Structures," OTC 3113,

Proceedings , Offshore Technology Conference,

1978, Vol. I, p. 529-533.

4. Idriss, I.M., Dobry , R. and Power, M.S., "Soil

Response Considerations in Seismic Design of

Offshore Platforms," OTC 2355 Proceedings . Offshore

Technology Conference, 1975, Vol. Ill, p. 191-205.

91
5. Bea, R.G., "Moderator's Report," Session on Offshore

Earthquake Geotechnology , International Conference

on Recent Advances in Geotechnical Earthquake

Engineering and Soil Dynamics, April 26 -May 2, 1981,

St. Louis, Missouri.

6. Bea, R.G., Audibert, J.M.E. andAkky, M.R.

"Earthquake Response of Offshore Platforms" Journal

of the Structural Division, Proceedings . American

Society of Civil Engineers, Vol. 105, No. ST2

February, 1979, p. 377-400.

7. Bea, R.G., Litton, R.W. , Nour-Omio, S. and Chang,

J.Y., "A Specialized Design and Research Tool for

the eling of Near-Field Pile-Soil Interactions,"

OTw 4806, Proceedings . Offshore Technology

Conference, 1984, p. 249-254.

8. Liou, D.N. and Penzien, J., "Seismic Analysis of an Offshore Structure Supported on Pile Foundations,"

Report No. UCB/EERC-77/25 , Earthquake Engineering

Research Center, University of California, November

1977.

9. Marshall, P.W., "Design Considerations for Offshore

Structures Having Non-Linear Response to Earthquakes," Preprint 3302, American Society of

Civil Engineers Convention, October 1978, Chicago,

Illinois

10. Popov, E.P., Mahin, S.A. and Clough, R.W.

"Inelastic Response of Tubular Steel Offshore

Towers," Journal of Structural Engineering,

Vol. Ill, No. 10, American Society of Civil

Engineers, October 1985.

11. Cardenas, J., "Drilling Rigs Designed for

Earthquakes," Petroleum Engineer International,

May 1988, p. 68-71.

12. Long, R.J., Watmough, D. and Effenberger, M.

"Seismic Evaluation of Platform Cranes," OTC

6171, Offshore Technology Conference, May, 1989.

13. "Recommended Practice for Planning, Designing,

and Constructing Fixed Offshore Platforms," API

RP2A, 17th Edition, American Petroleum Institute,

Washington, D.C., April 1987.

14. Arnold P., "Seismic Risk Analysis with Combined

Randon and NonRandon Earthquake Occurrence , " OTC

3111, Offshore Technology Conference, May, 1978.

15. Page, R.A. , "Evaluation of Seismicity and

Earthquake Shaking at Offshore Sites," OTC 2354,

Offshore Technology Conference, May, 1975.

16. Reece, E.W., Ryerson, D.E. and McNeill, R.L.,

"Long-Term Measurements of Ground Motions Offshore,"

PROCEEDINGS . VOL. 1, International Conference on Recent Advances in Geotechnical Earthquake

Engineering and Soil Dynamics, Rolla, Missouri,

April 26-May 3, 1981.

17. Bea, R.G., Arnold, P., Idriss, I.M., Beebe, K.E.

and Marshall, P.W. , "A Study of Soil-Pile-Structure

Systems in Severe Earthquakes," OTC 2749, Offshore

Technology Conference, May 1977.

18. Pawsey, S.F., Knudson, W.C., Vahdani, Si and

Bach, G.F., "Seismic Response of Gravity Structures

with Soil Interaction," PROCEEDINGS OF BOSS '85 . Delft, Netherlands, Jul 1985.

19. Nogami , T., "Soil/Pile Interaction Due to Seismic

Response of Subsea Soils," Technology Assessment

and Research Program for Offshore Minerals

Operations- -1988 Report, OCS Study MMS 88-0057,

Minerals Management Service, Reston, Virginia.

20. Finn, W.D.L., Martin, G.R. and Lew, M.K.W.

"Application of Effective Stress Methods for

Offshore Seismic Design in Cohesionless Seafloor

Soils," OTC 3112, Offshore Technology Conference,

May 1978.

21. Marshall, P.W. , "An Overview of Recent Work on Cyclic Inelastic Behavior and System Reliability,"

PROCEEDINGS , Structural Stability Research

Council, Lehigh University, Bethlehem, Pennsylvania

1982.

22. Mahin, S.A., Povov, E.P. and Zayas , V.A.

"Seismic Behavior of Tubular Steel Offshore

Platforms," OTC 3821, Offshore Technology

Conference, May 1980.

23. Marshall, P.W., Gates, W.E., and Anagnostodoulos

S., "Inelastic Dynamic Analysis of Tubular Offshore

Structure," OTC 2908, Offshore Technology Conference, May 1977.

24. Bea, R.G., "Earthquake and Wave Design Criteria for

Offshore Platforms," Journal of the Structural

Division, Vol. 105, NoST2 , American Society

of Civil Engineers, February 1977, pp. 401-419.

25. Mariow, R.S., "Effects of External Hydrostatic

Pressure on Tubular Beam- Columns , " Final Report,

SWRI Project Number 06-1184, Southwest Research

Institute, San Antonio, Texas, February 1988.

26. Chen, J., Ullmann, R.R. and Mason, A.B.,

"Measurement of Earthquake Ground Acceleration

and Structural Response of a Fixed Offshore

Platform," OTC 6172, Offshore Technology Conference,

May 1989.

27. Mason, A.B., Chen, J., Beck, J.L. and Ullmann, P.R.,

"Model Parameter Identification of an Offshore

Platform from Earthquake Response Reloads,"

ASCE Structural Congress, American Society of

Civil Engineers, San Francisco, California,

May 1989.

28. Wilson, B. , "Earthquake Analysis Studies for

Offshore Platforms," Annual Meeting, American

Petroleum Institute, San Antonio, Texas, April 1982.

29. PMB Systems Engineering, Inc., "Seismic Load Effects

Study," Prepared for the American Petroleum

Institute, API PRAC Project 22 Technical Advisory

Committee, 1985.

30. Moses, F. , "Development of Preliminary Load and

Resistance Design Document for Fixed Offshore

Platforms," API PRAC Project 85-22, Case Institute

of Technology, Cleveland, Ohio, January 1986.

31. U.S. Code of Formal Regulations, 30 CFR,

Part 250- -Oil and Gas and Sulphur Opertions in the

Outer Continental Shelf, Subpart I - Platforms

and Structures.

32. Seines, P.B., Hove, K. and Bungum, H. , "Seaquakes:

A Potential Threat to Offshore Structures," Third

International Conference on the Behavior of Offshore

Structures, Cambridge, Massachusetts, August 1982.

33. Sleefe, G.E., "Seafloor Seismic Data Study,"

Technology Assessment and Research Program for

Offshore Minerals Operations- -1988 Report,

OCS Study MMS 88-0057, Minerals Management Service,

Reston, Virginia, 1988.

92
SOFTWARE /

TOOL '

V) z

..—-STRUCTURAL RESPONSE/OAMAGE

ANALYSIS

SEISM ICITY

soil- pile

Interaction

SITE EFFECTS

ft'

<1

TRANSMISSION / ATTENUATION

Figure 3. Areas of consideration for offshore

earthquake engineering technology.

(Taken from Reference 17)

Figure 1. Typical components of a steel -jacket

offshore platform.

figure 2. Typical arctic gravity platform. 93
Mil)

0.80

« 0.40

SZ 0.40 E

*BP<r>

.01 .03 .] .3 1 Nfiod (S*condt)

o.ao

£ o.»o

s5 0.40 |

|

0.20

0.00

.01 .03 .1 .3 1 Ptfiad (Socondt)

Figure 5. Example results of ground response analysis

for an offshore platform site. (Taken from

Reference 4)

FREE-FIELD

ANALYSIS

FID

INERTIAL MOMENTS

AND SHEAR FORCES FROM STRUCTURE

RELATIVE DISPLACEMENT

DUE TO INTERACTION

BENDINC MOMENTS I DOMINATED BY LATERAL

INTERACTION LOADS

LE DISPLACEMENT

AT TIME t TRANSITION

t

FREE FIELO DISPLACEMENT

TIME HISTORIES

BENOINC MOMENTS

DOMINATED BY EARTHQUAKE

DISPLACEMENTS

FREE FIELO DISPLACEMENT

AT TIME 1

.MIC

Figure 7. Qualitative description of soil-pile

interaction under seismic loads. (Taken

from Reference 20)

LINEAR STRUT, BEAU ft D I A P m RA tj M

ELEMENTS

mtt rtELO SOIL ELEMENT

Figure 8. Typical finite element model required to

evaluate the soil-pile-structure system.

(Taken from Reference 17)

1.00

i

io

I

g

I

o

U-zI2

400 FT MODEL

600 FT MODEL

800 FT MODEL

1000 FT MODEL

PERIOD (sec)

Figure 6. Offshore soil-pile-structure interaction Figure 9. Frequency response at deck level

model for earthquake response. (Taken from horizontal base excitation fluid- structure

Reference 20) interaction not included.

94
1.00

o

I

g 0.40

0.00

- 400 FT MODEL

600 FT MODEL

800 FT MODEL

-1000 FT MODEL

4 6

PERIOD (sec)

1 -i

0.8 -

a!

^ 0.6

o

s

5

0.04

!

0.12

400FT MODEL

FT MODEL

800 FT MODEL

1O0OFT MODEL

0.16

PERIOD (sec)

Figure 10. Frequency response at deck level due to horizontal base excitation;

fluid- structure interaction included.

Figure 11. Frequency response magnification at deck

level due to vertical base excitation.

OFFSHORE PLATFORM RESPONSE TO EARTHQUAKES ——MEASURED AND CALCULATED

Figure 12 . Evaluation methodology for offshore platform

analysis and design.

THE SEMS CONCEPT

Figure 13. Illustration of the Seafloor Earthquake

Measurement System (SEMS) for incorporating long-term data recordings,

95
Centrifuge Dynamic Model Tests in Pnri

By

Takamasa Inatomi^, Motoki Kazama^, Setsuo Noda-^ and Hajime Tsuchida'*

SUMMARY

A centrifuge model test is one of the testing

method on geotechnical engineering. This

method is known to satisfy the similitude for

model tests of geotechnical materials. The

autors have developed earthquake simulators

which can be accomodated in the centrifuge of

the Port and Harbour Research Institute

(PHRI), and carried out dynamic model tests

in centrifugal field. The earthquake

simulator is of electrohydraulic type and can generate scaled large earthquake motions. The

paper introduces and illustrates the outline

of the earthquake simulators and some

experimental results using this apparatus.

KEY WORDS ; Centrifuge, Dynamic Model Test,

Earthquake, Test Equipment

1. INTRODUCTION

Soil structures and foundations in Japan have

often suffered from great earthquakes.

Studies in the dynamic geotechnical field are very important in Japan.

A centrifuge model test is one of the testing

method on geotechnical engineering. Using

this method we can make the same stress con- dition in a model ground as that in a real

structure. Because the physical properties of

geotechnical material, no matter which are

static one or dynamic one, depend on its

confining pressure, the use of a centrifuge

is effective for both static and dynamic

model tests. So far, only static model tests

have been carried out in the centrifuge of

PHRI since the centrifuge was constructed in

1981 (Ref.1). The use of the PHRI centrifuge

for dynamic tests had to wait until recently

when the authors decided to start developing

the earthquake simulator which can be accom- modated in the existing PHRI centrifuge

(Ref.2). Obviously the centrifuge model test

satisfies the similitude for models made of

geotechnical materials and is, at present,

the most promising method for testing model

grounds and structures. However, many difficulties had to be overcome for developing the

earthquake simulator and for applying to a dynamic model test in the centrifuge. The

purpose of this paper is to introduce the

outline of the earthquake simulator for PHRI

and to show some results of application to

the dynamic problem.

2. DESIGN OF THE EARTHQUAKE SIMULATOR

2.1 Scaling Law

The general scaling relationship for a dynamic model test is shown in Table 1 (Ref.

3). The scaling law for a centrifuge dynamic

model test is, also given in Table. 1,

obtained by substituting n$ , n6 , rip, n for

1/n, 1, 1, n in the general scaling T_aw,

respectively. A small-scale model and a fullscale structure are related through this

relationship. According to this scaling law,

if a model is constructed at a 1/50 scale, it

is subjected to a 50G centrifugal acceleration to simulate the prototype structure. In

the dynamic test, the frequency of the input

motion applied to the model is 50 times

higher, and the acceleration is 50 times

larger. For example, in order to simulate a real earthquake motion, which has a peak

acceleration of 0.2G, a duration of 60s and a frequency content up to 6Hz in the scaled

model, the model is subjected to an excitation which has a peak acceleration of 10G, a duration of 1.2s and a frequency content up

to 300Hz. If the mass of the model container

and specimen is about 100kg, the capacity of

the shaking force must be about 10kN.

In this scaling law it is found that the

stress condition at homogeneous points is the

same in the model and the full scale

structure using the same material. That is a major merit of the centrifuge model test

compared with the model test in ordinary 1G

field. Using the centrifuge we can take

account for the fact that the ground itself

change its properties under the action of its

own weight.

1 Dr. Eng., Chief of Structural Dynamics

Laboratory, Structures Division, Port and

Harbour Research Institute, Ministry of Transport, 3—1 —1 » Nagase, Yokosuka, Japan.

2 Senior Research Enginner, Structures Division, ditto

3 Dr. Eng., Director, Structures Division,

ditto

4. Dr. Eng., Director General, ditto

96
2.2 Shaking System Selection

In order to design a centrifuge earthquake

simulator, we had to consider advantages and

disadvantages of various shaking systems.

The authors examined various approaches such

as the 'Bumpy Road' of Cambridge University,

electromagnetic devices, electro hydraulic

equipments, explosive firing and piezoelectric actuator(Ref .4.-8). With the Bumpy Road,

we will have to make a great deal of effort

to change the input motion and to modify the

PHRI centrifuge. The electromagnetic devices

are too heavy and large to mount on the swing

platform, and explosive firing seemed difficult to control the input motion. Moreover

the authors could not find a large piezoelectric actuator in Japan. Consequently, it was decided to attempt to design an electrohydraulic type of shaker. Though this system is

a little complex, we need no improvement on the PHRI centrifuge. This system satisfies

the required capacities for shaking and the

required reliability. Fortunately, the

authors could find an accumulator, a servo

valve and shut off valves which can be

mounted on the swing platform. The reliability of the servo valve and of the shut off

valves and the safety of the accumulator in

a 60G centrifugal field were confirmed by the

authors before making up the whole system.

3. OUTLINE OF THE EARTHQUAKE SIMULATORS

We made two earthquake simulators till today, one is a trial production system for feasibility study and the other is a version up

system for practical experiment. But the

fundamental mechanism of both system are the

same. The earthquake simulator consists of an

oil pressure supply unit, an actuator unit, a control unit and an oil tank. Fig.1 illustrates the arrangement with all the units of

the version up system on the swing platform.

Fig. 2 shows the layout of the oil flow

circuit of the simulator. The oil supply unit

consists of an accumulator to supply oil

pressure to the servo valve and three shut

off valves. The A-shut off valve reduces oil

leakage from the servo valve. The B-shut off

valve is used in case of emergency. The Cshut off valve with flow control valve is

used to avoid an impulse liquid flow to the

servo valve. Once we accumulate oil pressure,

we can experiment ten and several times

during one flight.

3.1 Trial Production system

The model container box of the trial production system is fixed on the cylinder of the

actuator. The dimensions of the shaking box

are ^OOmm(L) X 180mm(D) X 270mm(H). It is

supported by eight arms from four beams. The

container is made of titanium to save the

total shaking mass. The major specifications

of the system are shown in Fig.3.

The system controller is composed of two

control boxes. One is mounted on the swing

platform and the other is located in the

control room. They are connected by eight

lines through slip rings. The controller has

a sequence circuit to control the experiment

from start to end. That is to say, if an

operator presets the experimental conditions

such as input wave, duration of shaking,

amplitude and frequency, the only thing the

operator needs to do during the flight is to

press one button. The servo valve receives

electrical signals from a function generator

or Read Only Memory (ROM) and the actuator

generates the input motion. Typical

earthquake motions are memorized in the four

ROMs. Each ROM has a length of 8192 words and

we can select the time interval of the input

motion from 0.001ms to 10ms.

3.2 Version up System

We improved on the trial production system to overcome the following drawbacks.

(1) We can not see the phenomena in visual

because the control box hinders the soil

container.

(2) We can select only one input motion

during one flight because we have to set the

input motion conditions before the centrifuge

operation.

(3) It is inconvenient for making a model and

for conducting several model tests

continuously because the model container is

fixed on the cylinder.

(4.) The shaking force is small for exciting a satisfactory large model.

The major actuator specification of the

version up system is also shown in Fig.3.

The shaking force of the version up system is

four times larger than that of the former

one. And various model containers including a stacked ring apparatus are available because

of using a shaking table. The dimensions of

the shaking table are 700mm X 350mm. We can

control this system and another instruments

such as high speed camera from control room

by a computer through RS232C interface and

can set various test conditions. Input

digital data are written in ROM whose

capacity is 32K words, we can select input

motion used in the test by setting the ROM

address.

3.3 Data Acquisition System

There are two current way to collect data as follows. One is ordinary way with the

transducer installed in the model, the other

is optical way using cameras. In the former

method we use a dynamic amplifier on the

swing platform to save the slip ring channels

and to improve the SN ratio. In the case of

centrifuge dynamic model test, as an object

is a very fast phenomena, we operate the

analogue data recorder from a few seconds

97
before the experiment starts to avoid failure

in collecting the data. The data collected byanalogue data recorder are converted into

digital data after experiment. We can use both a high speed camera and an ordinarycamera, which are effective for the phenomena

with large deformation. We can trace movement

of the target set up in the model by

analyzing photographs.

4. APPLICATION OF THE SYSTEM TO THE DYNAMIC

GEOTECHNICAL PROBLEM

4.1 Residual deformation of the gravity type

caisson during earthquake

The authors discussed about sliding deformation of the gravity type caisson during ear- thquake using the model as shown in Fig. 4.

The model ground was constructed by pouring

the dry Toyoura sand naturally from 20cm

above the water surface into water. The model

ground had wet density of 1.89 g/cm3 and

relative density of 38'& We used water as a pore fluid material to avoid liquefaction,

because the sliding behavior of the caisson

on the ground without liquefaction is different from that with liquefaction. The input

motion was the integrated displacement of the

acceleration observed at Hachinohe Port in

the 1968 Tokachi-oki earthquake, M=7.9. Four

stage tests ware conducted as shown in

Table. 2. Fig. 5 shows the time histories

measured in the stage-4 test. It is found

that the caisson slides in a moment when

large acceleration takes place. The residual

deformation at the top of the caisson and the

maximum response acceleration of the caisson

are shown in Fig. 6 and Fig. 7 with the

maximum input acceleration in prototype

scale. The following trends about the sliding

behavior are found in these figure.

(1) The residual deformation increases with

increasing the input acceleration. The

caisson subjected to input acceleration of

200Gals slides 10cm in prototype scale.

(2) The pore water pressure did not rise

remarkably, because the model fluid

material was water. This result is predicted

by the scaling law about the permeability of

the fluid material.

(3) The pore water pressure decreases when

the caisson is sliding.

(4) The maximum response acceleration of the

caisson did not so much increase with

increasing the input acceleration because of

caisson sliding.

4.2 Confining pressure effects on the shear

wave velocity of the sandy layers

It is well known that a shear modulus of the

sand is in proportion to square root of

effective confining pressure aV . From

elastic wave theory, it is also known that

the shear modulus is equal to pVs . Thus,

the shear wave velocity is in proportion to

(Q~v )°'

. This fact is important for

evaluating the dynamic properties of the

centrifuge model ground constructed. The

authors studied on this point with the saturated sandy layers. Fig. 8 shows the cross

section of the model and the ground depth in

prototype scale based on the scaling law. The

sand and the pore fluid material used in the

model and the model construction method are the same as those of the caisson sliding

experiment described above. We conduct experiment with two different relative density

model ground. Since it was difficult to detect the shear wave velocity directly in time

domain, we estimate it by the response characteristics of the model ground in frequency

domain. The input motion is a white noise,

that frequency content of the acceleration is

constant with frequency, which is generated

by the authors in considering the response

characteristics of the servo valve. Table.

3

lists the test condition in prototype scale.

In all cases we calculated the transfer

function between the input acceleration and

the acceleration response of the sand layer

surface and search for the first natural

frequency of the ground(fl). Typical transfer

function is shown in Fig.9.

4.2.1 Average shear wave velocity of sandy

layer

If it can be assumed that the model ground

has an average shear wave velocity Vs, Vs can be identified by multiplying f1 by 4H, in

which H represents layer thickness. Fig. 10

shows the relation between the average shear

wave velocity and the effective confining

pressure at the middle depth of the ground.

In this case the average shear wave velocity

did not strongly depend on strain level

because the model ground had high stiffness.

The solid line described in this figure are regression line which is revolved using

equation Vs = a(oV ) b

, in which a and b

represents constant number.

4-2.2 Distribution of shear wave velocity

with depth

We estimated distribution of the shear wave velocity with depth by the following

procedure.

(1) First we divide the ground into five

layers in case-2 shown in Fig. 8. In case-1

the ground is divided into three layers. It

is assumed that the divided each layer has a constant shear wave velocity.

(2) We regard the shear wave velocity of the

ground from surface to -2.4m /depth as the

average shear wave velocity estimated by the

method described above at 1 0G centrifugal

field.

(3) Next we assume the shear wave velocity of

the ground from -2.4m to -4.8m depth as Vs2.

Using Vs1 determined on the former stage and

Vs2 assumed here, we can calculate the

98
response of the ground based on multiple

reflection theory. If the dominant frequency

calculated is agree with that obtained by the

experiment at 20G centrifugal field, we adopt

Vs2 as the shear wave velocity of the ground

from -2.4m to -4.8m. If not so, we assume new

Vs2 and calculate again. We calculate several

times till obtaining good agreement.

(4) We can estimate the shear wave velocity

of the ground under -4« 8m depth similarly.

Fig. 11 shows the distribution of the shear

wave velocity obtained by this procedure. The

line described in this figure inclines 0.25

power of the depth through the center point

of the middle layer. These results indicate

that the confining pressure affects on the

shear wave velocity of the sandy ground even

if its depth is less than 12m. And close

agreement between the experiment and the

element test conducted in the 1G field was

obtained (Ref.9). The method described here

is one of effective way for evaluating dynamic properties of soils in centrifugal field.

4.3 Modelling liquefaction

The research about liquefaction is important

object using centrifuge because liquefaction

is directly related to the effective

confining pressure generated by its own

weight. During recent several years some

experiment about liquefaction using

centrifuge have been conducted (Ref. 10-13).

It is necessary for modelling liquefaction in

nG centrifugal field to reduce permeability

of the fluid material n times smaller than

that of prototype. Thus we studied relation

between the coefficient of permeability and

the viscosity of silicon oil as a fluid

material. The permeability tests were

conducted under 20 degrees centigrade using

Toyoura sand of relative density 50% and 10%.

Fig.12 shows the test results with viscosity.

The results at the point of 1 centistokes

represents that using a water as a fluid

material. Fig.13 shows permeability reduction

ratio of the silicon oil compared with the

water. In this figure the coefficient of

permeability of a 50 centistokes silicon oil

is about 30 times smaller than that of the

water. It is also found that the variation of

the permeability is almost independent of the

relative density.

Fig. 14 shows the cross section of the model

constructed in a stacked rings. The experiment was conducted in a 30G centrifugal field

using 50 centistokes silicon oil. The model

ground was constructed by removing the

saturated sand into silicon oil without

mixing the sand with the air, after getting

the air out from the sand steeped in the

silicon oil under vacuum. The ground

constructed has relative density of 86& The

input motion used here is the same as that of

the caisson sliding test described above.

Table 4 lists the maximum response acceleration of the ground. The maximum response

acceleration increases from bottom to

surface. Fig. 15 shows the time histories

converted into prototype scale. The rise of

excess pore pressure was remarkable at the

point of the large acceleration took place.

Fig.16 shows the distribution of the maximum

excess pore pressure with depth. The brake

line described in this figure represents

initial effective stress. The excess pore

pressure generates with increasing acceleration level and cause perfect liquefaction at

the upper part of the layer in stage-3 test.

According to further investigation, there was few settlement of the ground during vibration.

5. CONCLUSIONS

We introduced the earthquake simulator in

PHRI centrifuge and explained some results of

application to dynamic geotechnical problem.

The earthquake simulator developed by the

authors shows satisfactory performances for

testing ground and structure models in the

centrifuge. Obviously, the earthquake

simulator in the centrifuge has a wide

applicability to geotechnical problems such

as soil-structure interaction. But there are many problems had to be overcome for

applying to the dynamic geotechnical testing

as follows.

( 1 ) Technical problem

* Development of more convenient equipment

* Model construction method

* Effects of side friction and rigid end wall

of model container

* Choice of the fluid material

* Evaluation method of constructed model

properties in the centrifugal field

(2) Internal problem of centrifuge model test

* Coriolis effect

* Influence of difference centrifugal acce- leration with model depth

* Relation between the dynamic properties

and frequency, that is to say, an effect

of strain rate with cyclic loading.

* Impossibility of modelling structures made

of large material such as gravel.

The technical progress of the dynamic

centrifuge testing and the so-called

'modeling of models' analysis will solve

these problems, and fruitful results are

expected from the model tests in the coming

few years.

REFERENCES

1) Terashi,M., "Development of PHRI

Geotechnical Centrifuge and Its

Application, " Report of the PHRI Vol.24,

No. 3, 73-122, (1985).

2) Kazama, M. , Inatomi, T. and 0htsuka,K.,

"Development of an Earthquake Simulator

99
for the PHRI Centrifuge," Technical note

of the PHRI, No. 607, 35p.

(in Japanese 1988).

3) Inatomi,T, Kazama,M. and Ohtsuka, K.,

"Development of Earthquake Simulator in

PHRI Centrifuge and Its Application",

Proc.of 9WCEE, Vol.VIII, 831-836, (1988).

4.) Schofield, A. N., "Dynamic and Earthquake

Geotechnical Centrifuge Modelling," Proc.

of Int. Conf. on Recent Advance in

Geotech., 1081-1100, (1981).

5) Scott, R. F., "Centrifuge Model Testing

at Caltech," Proc. of Soil Dynamics and

Earthquake Engineering, Vol.2, No. 4, 188-

198, (1983).

6) Zelikson, A., Devaure, B., and Badel, D.,

"Scale Modelling of Soil Structure

Interaction during Earthquake Using a Programmed Series of Explosions during

Centrifugati on," Proc. of Int. Conf. on Recent Advances in Geotech., 361-366,

(1981).

7) Arulanandan, K., Canclini, J. and

Anandarajah, A., "Simulation of Earthquake Motions in the Centrifuge,"

Proc. of ASCE, GT5, 730-742, (1982).

8) Whitman, R.V., "Experiments with Earthquake

Ground Motion Simulation," Proc. of Symp. on the application of centrifuge modelling to geotechnical design, 281 -299 f (1984).

9) Iwasaki, T., Tatsuoka.F. and Takagi,Y.:

"Shear Moduli of Sands under Cyclic' Torsional Shear Loading", Soils and Foundations, Vol.18, No.1, 39-56, (1978).

10) Schofield, A.N. and Venter, K.: " Earthquake Induced Pore Pressure in the

Foundation of Sea Dyke", CUSD/DS0ILS/TR150, Dept. of Engineering,

Cambridge University, 17p. (1984).

11) Whitman, R.V., Lambe,P.C. and Kutter,

B.L.:"Initial Results from a Stacked Ring

Apparatus for Simulation of a Soil Profile", Proc. of Int. conf. on Recent

Advance in Geotech., 1105-1110, (1981).

12) Lee, F.H. and Schofield, A.N.:"

Centrifuge Modelling of Sand Embankment

and Islands in Earthquakes", Geotechnique

38, No.1, 45-58, (1988).

13) Hushmand,B., Scott, R.F. and Crouse, C.B.

" Centrifuge Liquefaction Tests in a Laminar Box", Geotechnique 38, No.2, 253-

262, (1988).

14) Koga, Y., Taniguchi, E., Koseki, J. and

Morishita, T.:" Sand Liquefaction Tests

Using a Geotechnical Dynamic Centrifuge",

Proc. of the 20th Joint Meeting of

US. /JAPAN Panel on Wind and Seismic

Effects, US/JAPAN Natural Resources

Conference, 8p, (1988).

Table. 1 Scaling relationship for a dynamic model test.

Quantity General Scaling law in the

(Model/Prototype) scaling law centrifugal field

Length 1/n

Mass density Tip 1 Strain 1 Acceleration n Time (n e £f/n ) 1 {2

(n£ n fi /n|)- 1 / 2 n £ ni

1/n

Frequency n Displacement 1/n

Stress nP n

g nJ! np ngna /

£

1 Stiffness matrix 1 Pore water pressure

U[ nf/n

g

) 1/2 /np

n

p n

g n

i 5/n £

1 Coefficient of 1/n

permeability matrix

Bending stiffness of pile 1/n^

Table 2 Test condition

Stage Centrifugal Model Prototype

No. Acc. at 3.62m Amax ( G

Amax ( Gals

1 50.3 2.4 47

2 50.2 5.6 1 09

3 50.2 10.1 197

4 49.9 12.8 252

Amax: The max. acceleration of the input motion

100
Table 3 The model test condition

Target centrifugal Maximum acceleration of the input motion

acceleration in prototype scale, unit: Gal

Flight No. 1 2 3 4

50G 60,61

Case-1 30G 85,84

10G 111,110

50G 29 53 79

40G 34 61 80 89

Case-2 30G 43 82 126 \ 15

20G 66 1 24 209 276

10G 107

Table. 4 The maximum acceleration response o i the ground

The maximum response acceleration

Measure point in prototype scale : Gal

Stage-1 Stage-2 Stage-3

A 5 U5 165 231

A 4 115 1 50 213

A 3 101 1 30 195

A 2 76 110 1 89

A 1 75 135 31 5

1600

Fig.1 Arrangement of all the units on the swing platform.

101
external

coupler

on swing platform

flow

shut off valve control valve —k3 —

• v» ( accumu-^^

v

1a *° r

oil

tank

„shut off valve HiHn?"] shut off

X valve

model container —TM i . ;

•

'

' ; - actuato

servo valve

n

Fig. 2 Oil flow circuit

Trial Production System

Shaking Box

Version up System

Bucket Stacked Rings

/

High Speed Camera

Quantity Trial production system Version up system

Shaking force about + 12kN about + 50kN

Maximum displacement + 3mm + 6mm

Maximum velocity + 30cm/s + 4-5cm/s

Volume of the accumulator 4. liter 10 liter

Frequency content (nominal) up to 300Hz up to 250Hz

Payload 92kg 200kg

Maximum usable oil pressure 30MPa (ordinary use 21MPa)

The servo valve is controlled by LVDT displacement feed back system.

Fig. 3 Specification of the earthquake simulator for PHRI centrifuge.

102
D1

23cm — =U=I„

.'. • PI \

• 10cm >A3

6cm

• * • . .

.

•

' :

• •'• n AO- •

'

* 40cm »

O Accel erometer

• Pore pressure

meter

DO

Actuator

Fig. 4 Cross section of the gravity type caisson model.

A2: Backfill Surface Acceleration (G)

0,

•15 J

^ DO: Actuator Displacement (mm)

0 ^ AfWWV^ v *^\, o^-

-3 J

-15^ AO: Input Motion Acceleration (G)

-15

.j A3: in the front of the Caisson Acceleration (G)

0 - . - , » . M>H

15 J

3 D1 : at the top of the Caisson Displacement (mm)

0 r

-3 J

P1 : at -5cm from backfill surface Pore Pressure(kPa)

0

-30

,j, ... ff

. ....»! ., I

1 1 1 1

I 1 1 1 1 1

1 0.5s 1.0s 1.5s

Fig. 5 The time histories measured in the No. U experiment.

103
4-> c

Q)E

0)u

<o . r— c

Q. O

t/> (/) •i- to T3 !-

(O >— O

10

=3 <U -O -C

•i- +->

I/) QJ <+. Q£ O

10 -

5-

(cm)

in prototype scale

0 100 200 (Gals)

Maximam acceleration of

the input motion (AO)

Fig. 6 The residual displacement

at the top of the caisson.

co

f—p

«rs i.(V

0)o

u

<c

QJ ,

l/l c c

o o

CL l/l w (/> 0) •r- i. (0u

E

3 0) E -C

•f- -p K

<0 4—

O

200'

1n prototype scale

100 200 (Gals)

Maximum acceleration of

the input motion (AO)

Fig. 7 The maximum response

acceleration of the caisson.

Accel erometer

Pore pressure

meter

equivalent

prototype depth

Vsl

Vs2

Vs3

Vs4

Vs5

Fig. 8 Cross section of the model and the equivalent prototype depth.

104
uj O U. l—i go t— z o

< 2

15

10

5

180

90

at a 306 centrifugal

acceleration

(flight No.l)

0. 100. 200. 300. 400,

natural frequency

fl

0. 100. 200. 300.

Frequency-(Hz)

Fig. 9 Typical transfer function

at a 30G centrifugal field.

500.

o

400. 500.

E—

>>

+Ju

o

0)>

>

s_

IT3 <D -C

oo

<u

s.><

450-i

400-

350-

300

250 J

A CASE— 1 (D r =48%)

O CASE- 2 (D r =50%)

a v'

V s =219.4 Uv' ) 0 34 1 V s =280.3 (a v' )

0 226

1

i3 (tf/nf)

Effective confining pressure

at the middle depth of the ground

Fig.10 Relation between the average shear

wave velocity and effective confining

pressure at the middle depth of the

ground

.

Shear wave velocity(m/s)

0 500

-4

4J Q

•12

\

\

\

T

4

i

\

\

V —\—

\

\

\»-

t

I

t

1

Shear wave velocity(m/s)

0 500

E

-C

4-> Cl

CU Q

12

' V

\

\

\-»

\

\

\

t

I

1

CASE-1 CASE-2

Fig. 11 Distribution of the shear wave velocity with depth.

105
5X10

60 80 100

Viscosity (cs)

Fig. 12 Relation between the viscosity and the coefficient of permeability.

O) -p E <o

<u ' Q- <—

<+• o o

c

+-> o

c u

(U -r- O !-

•f- 00

1/80

1/60

1/40

1/20

o

<_>

O Dr=50%

Dr=70% Dr: Relative density

I—i viscosity range

presented by maker

Viscosity (cs)

Fig. 13 Permeability reduction ratio of the silicon oil compared with the water.

106
unit: cm

Toyoura sand saturated

with 50cs silicon oil

A o Accel erometer

Dl P • Pore pressure meter

X

f- • P3 • -vo"

. • Ot

• A1 0

• P4

• P2 •

.PI

:>A5 -r

3

»A4

•A3 —

n

?A2

30

Fig. H Cross section of the liquefaction model using stacked rings.

P4-

P3-

P2--

P 1 Jm

O— Stage-1

A—A Stage-2

O—O Stage-3

Initial effective

stress.

4-

50 100 kpa

Excess pore pressure.

Fig. 16 Distribution of the maximum excess pore pressure

107
100.

0.

100.

70.

-70.

60.

-GO.

20.

-20,

400,

400.

200,

200.

200.

-200.

300.

0.

300.

300.

0.

300.

P1 : -36.2cm under the layer surface Pore Pressure (kPa)

P2 : -26.7cm under the layer surface Pore Pressure (kPa)

P3 : -16. 6cm under the layer surface Pore Pressure (kPa)

P4- s - 6.4cm under the layer surface Pore Pressure (kPa)

A1 : Shaking Table Acceleration (Gal)

A2 : Stacked ring at +9. 5cm above base Acceleration (Gal) —— "W^fftyVifl|^^ *

A3 : Stacked ring at +19. 5cm above base Acceleration (Gal)

-

kU : Stacked ring at +29. 5cm above base Acceleration (Gal)

A5 : Stacked ring st +36. 5cm above base Acceleration (Gal)

"i r

0. 20,

i i i i i i i i i 40. 60. 80. 1 00. 1 20,

TIME-SECONDS'

Fig. 15 Time histories of stage-3 in prototype scale.

108
Modeling of Earthquake Induced Pore Pressure Behavior

by

Richard H. Ledbetter

ABSTRACT

Centrifuge model testing with simulated

earthquake excitation can provide an extensive

data base for the detailed study of earthquake

induced prototype behavior, including pore fluid

pressure response. In centrifuged models,

prototype stresses can be produced at points in

the model corresponding to those in the full

scale structure by creating an artificial

gravity field in the model. Each soil element

in a centrifuged model may thus be expected to

undergo the same response history as corresponding elements in a prototype for a given earthquake excitation. Centrifuge model

tests for the study of earthquake -induced

behavior have been conducted on the large

geotechnical centrifuge at Cambridge University,

England. The study of actual field-measured

pore pressure behavior and response under

earthquake loading compared to the centrifuge

model tests shows that the pore pressure

behavior in the centrifuge tests is consistent

with the field measured behavior.

KEYWORDS: Centrifuge; earthquake; pore pressure.

1. INTRODUCTION

Study is indicating the same consistent behavior

of actual pore fluid pressure under earthquake

loading and physical tests yielding equivalent

prototype response to earthquake excitation.

Direct instrument measurements of excess pore

water pressure induced by earthquakes have been

made; (1) Ikehara (1970) reported measurements

below a railroad embankment, (2) Ikuta et al.

(1979) reported measurements near the basement

walls of a building, (3) Ishihara (1981)

measured pore water pressure rise at two depths

within the artificial island of Ohgishima which

felt three earthquakes, (4) Ishihara et al.

(1981) measured excess pore water pressure at

two depths within the artificial island of Owi,

(5) Harp et al. (1984) reported measurements in

lake shore sediments, and (6) Holzer et al.

(1988) reported measurements from the Wildlife

liquefaction array.

Centrifuge model testing with simulated

earthquake excitation provides an extensive data

base for the detailed study of earthquakeinduced prototype behavior, including pore fluid

pressure response. In centrifuged models,

equivalent prototype stresses can be produced at

points in the model corresponding to those in

the full scale structure by creating an artificial gravity field in the model

(Schofield, 1981). The ability to induce

prototype stresses in a model is important

because soil properties and responses are dependent on effective stresses. Each soil

element in a centrifuged model may be expected

to undergo the same response history as corresponding elements in a prototype for a given earthquake excitation.

In a centrifuged model, N is the model scale

factor and the model is subjected to inertial

accelerations of Ng, where g is the acceleration

due to earth's gravity. For a saturated soil

model subjected to an earthquake excitation, the

behavior will depend on both the interaction

between inertial effects and the local

1. U.S. Army Engineer Waterways Experiment

Station Vicksburg, Mississippi 39180-0631

109
generation and dissipation of pore fluid

pressures. The time scales for inertial

2 effects, N, and for diffusion, N , can be

brought together by slowing the diffusion

process by using a more viscous pore fluid

(Schofield, 1988). The time scales will be

identical if a fluid of viscosity N times

greater than water is used for modelling at Ng

inertial acceleration.

The United States Nuclear Regulatory Commission

and United States Army Corps of Engineers

sponsored a series of centrifuge model tests to

provide data for the study of earthquake induced

settlements/deformations and liquefaction. The

tests were conducted on the large geotechnical

centrifuge at Cambridge University, England.

Details of the Cambridge centrifuge and

associated procedures for simulated earthquake

testing have been described by Schofield (1981)

.

This paper presents typical centrifuge model

pore fluid pressure response to earthquake

excitation which is consistent with field

measurements and indicates that pore fluid

pressures respond to the peak motions of

excitation. The test data and field

measurements indicate that pore pressures

dissipate rapidly and low permeability confining

layers and/or low permeability liquefied soil

may be necessary to sustain pore pressures at an

elevated state in a liquefied soil mass.

2. FIELD PORE PRESSURE BEHAVIOR

Figure 1 (National Research Council, 1985) shows

the soil profile at Owi Island, Japan, where the

pore pressures in Figure 2 (National Research

Council, 1985) were measured by Ishihara (1981)

at the 6 and 14 meter depths during the

Mid-Chiba earthquake in 1980. The measurements

were in silty fine sand. An accelerograph

measured surface motion. Note that: (1) the

pore pressures responded immediately to the peak

motion, they did not accumulate from the start

of the earthquake, and they started dissipating

immediately after the peak motion (they did not

hold or continue to increase with the earthquake

continued motion), (2) the pore pressure at

the 14 meter depth did not dissipate as rapidly

as that at the 6 meter depth, and (3) the soil

profile shows clay layers above and below the 14

meter depth. The less rapid dissipation at the

14 meter depth illustrates the effect of the

clay layers bounding the zone and that low

permeability confining layers may be necessary

to sustain pore pressures.

Figure 3 (Harp et al. 1984) shows pore pressure

measured at a 1.3 meter depth in a saturated

gravelly sand at the shore of Convict Lake,

California, during the Mammoth Lakes earthquakes

of 1980. The recorded surface acceleration is

superimposed on the pore pressure record

illustrating that the pore pressure followed the

earthquake motions; peaking with the peak motion

and not accumulating with continued motion. Low

permeability layers would have been necessary to

confine the gravelly sand in order for pore

pressures to have been sustained. The site

liquefied during the main shock and underwent

lateral spreading; the pore pressure transducer

was installed after the main shock and recorded

\ the aftershock sequence.

3. LABORATORY PORE PRESSURE BEHAVIOR

3.1 Simple Shear Test

Figures 4 thru 6 (Ishihara et al. 1988) show

pore pressure behavior of loose, medium, and

dense Fuji river sand in undrained cyclic simple

shear laboratory tests using irregular mutually

perpendicular motions from three earthquakes

recorded in Japan. Ishihara and Nagase (1988)

were studying the effects of irregular and

multi- directional loading on soil during

earthquakes. Shown in each figure are the

surface recorded earthquake components of motion

that were used as test input and the shear

110
strains and pore pressures for two stress

ratios. Of note in these figures is that the

pore pressures respond to the peak motions and

do not significantly accumulate. This pore

pressure behavior is consistent with field

measured responses. The pore pressures in these

tests are held at the levels induced by the peak

motions be cause the tests are undrained, thereby

locking in the pore pressures and not allowing

them to dissipate during or after the earthquake

loading, as might occur under field conditions

unless there were confining low permeability

layers

.

3.2 Centrifuge Test

Figure 7 shows the schematics of a typical

centrifuge physical model test illustrating an embedded structure, displacement transducers

(LVDT) , accelerometers (ACC) , and pore pressure

transducers (PPT). Earthquake simulation motion

is input at the model base and transmitted

through the model, primarily by shear stresses

but also by normal stresses due to bending

modes. De-aired blended silicone oil with a viscosity of N times the prototype viscosity was used as the pore fluid to model the drainage

conditions in the prototype during the

earthquake. Tests by Eyton (1982) have shown

that the stress -strain behavior of fine sand is

not changed when silicone oil is substituted for

water as a pore fluid.

Figures 8 and 9 show the typical pore pressure

response character in centrifuge tests with

homogeneous models of Leighton Buzzard and

Banding sand, ranging in relative density from

45 to 70 percent. The pore pressure records

have been filtered and smoothed so that the main

changes of the mean response can be compared to

the earthquake acceleration input records.

Major changes in the pore pressures were in

response to the peak motion stages of the

records; they stay at about the same level or

start dissipating unless a larger motion peak

occurs, do not accumulate with cyclic loading,

and dissipate after the earthquake. Figure 10

shows the dissipation behavior of pore

pressures. The pore pressure character in the

centrifuge is consistent with field behavior.

These sands were very fine, with nominal grain

sizes of 0.1 and 0.19 mm for the Leighton

Buzzard and Banding sands respectively, which is

why they tended to hold the pore pressures

during earthquake loading.

4. CONCLUSIONS

The study of actual field measured pore pressure

behavior and response under earthquake loading

compared to the laboratory test results of

special multi-directional loading simple shear

and centrifuge model tests is indicating that

pore pressure behavior in centrifuge tests is

consistent with field measured behavior.

Centrifuge earthquake simulation testing can provide prototype truth data concerning

earthquake- induced response in soils and can provide an extensive data base for detailed

study, verification of analysis techniques, and

to fill the voids in the knowledge of pore

pressure behavior under earthquake excitation.

5. ACKNOWLEDGMENTS

Financial support for the centrifuge

investigations by the U.S. Nuclear Regulatory

Commission, the U.S. Army Corps of Engineers,

and cooperative research from Cambridge

University, England, is gratefully acknowledged.

Appreciation is expressed to NRC, WES and Office

Chief of Engineers for granting permission to

publish this paper. Gratefully acknowledged are

the contributors to this work, in alphabetical

order

:

L. L. Beratan, U.S. Nuclear Regulatory

Commission

W. D. Liam Finn, University of British

Columbia, Vancouver, Canada

A. N. Schofield, Cambridge University,

England
R. S. Steedman, Cambridge University, England

6. REFERENCES

1. Eyton, D.G.P. 1982. "Triaxial Tests on

Sand With Viscous Pore Fluid," Part 2,

Project Report, Engineering Department,

Cambridge University, Cambridge, England.

2. Harp, E.L., Sarmiento, J., and Cranswick, E.

1984. "Seismic-Induced Pore-Water Pressure

Records From The Mammoth Lakes, California,

Earthquake Sequence Of 25 To 27 May 1980,"

Bulletin of the Seismological Society of

America, Vol 74, No. 4.

3. Holzer, T.L. , Youd, T.L. , and Bennett, M.J.

1988. "In Situ Measurement of Pore Pressure

Build-Up During Liquefaction," Wind and

Seismic Effects, Proceedings Of The 20th

Joint Meeting Of The U.S. -Japan Cooperative

Program In Natural Resources Panel On Wind

And Seismic Effects, U. S. Department of

Commerce, National Institute of Standards

and Technology.

4. Ikehara, T. 1970. "Damage to Railway

Embankments Due to the Tokachioki

Earthquake," Soils and Foundations, Japanese

Society of Soil Mechanics and Foundation

Engineering, Vol. 10.

5. Ikuta, Y. , Maruoka, M. , Mitoma, T. , and

Nagano, M. 1979. "Record of Lateral

Pressure Taken During Earthquakes," Soils

and Foundations, Japanese Society of Soil

Mechanics and Foundation Engineering, Vol.

6. Ishihara, K. 1981. "Measurements of Insitu

Pore Water Pressures During Earthquakes,"

Proceedings of The Int. Conf . on Recent

Advances of Geotechnical Engineering and

Soil Dynamics, St. Louis Missouri, Vol. I.

7. Ishihara, K. , Shimizu, K. , Yamada, Y. 1981.

"Pore Water Pressures Measured In Sand

Deposits During An Earthquake," Soils and

Foundations, Japanese Society of Soil

Mechanics and Foundation Engineering, Vol.

21.

8. Ishihara, K. , and Nagase, H. 1988.

"Multi-Directional Irregular Loading Tests

On Sand , " Soil Dynamics and Earthquake

Engineering, Vol. 7.

9. National Research Council. 1985.

"Liquefaction of Soils During Earthquakes."

10. Schofield, A. N. 1981. "Dynamic and

Earthquake Geotechnical Centrifuge

Modelling," Proceedings of the Int. Conf. on Recent Advances in Geotechnical Engineering

and Soil Dynamics, St. Louis, Missouri, Vol.

III.

11. Schofield, A.N. 1988. "Geotechnical

Centrifuge Modelling In 1988,"

CUED/D - SOILS/TR .213, Cambridge University

Engineering Department, Cambridge, England.

112
Figure 1. Owi Island soil profile for the Mid-Chiba earthquake,

September 25, 1980 (from National Research Council,

NRC,1985)

113
10 20

Time in second

40,60

Figure 2. Surface acceleration components and pore water

pressure responses at Owi Island, Mid-Chiba

earthquake (from NRC, 1985)

114
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DAY ISO TIME 17/0/45

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and horizontal acceleration, May 1980 Mammoth Lakes

earthquake sequence (from Harp et al. 1984)

115
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116
embedded

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direction of shaking

TESTS RSS.90 and RSS.91

ISOMETRIC SKETCH OF SOIL MODEL WITH AN EMBEDDED STRUCTURE

FIG. NO.

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TABLE OF EVENTS

FLIGHT e-Lcvn CARTHOUAKC

MAX. AMPLITUDE IK)

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TEST RSS.91

MODEL SAT EMBD

DESCRIPTION

TEST RSS.91 . Embedded structure, saturated soil DESCRIPTION AND TABLE OF EVENTS

FIG NO.

ure 7. Schematics of a typical centrifuge physical model test

117
PORE PRESSURE RESPONSE CHARACTER IN CENTRIFUGE TESTS

Ol

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Peak value- 5.197

RSS91 EQ1 PORE PRESSURE RESPONSE

Peak value- 7.288

\

Time (sees)

10 15

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centrifuge tests with homogenous sand models

118
PORE PRESSURE RESPONSE CHARACTER IN CENTRIFUGE TESTS

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Peak value- 11.123

—r— 15

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Time (sees)

—i

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Figure 9. Typical pore fluid pressure response character in

centrifuge tests with homogenous sand models

119
Figure 10. Typical pore fluid pressure dissipation behavior in

centrifuge tests with homogenous sand models

120
Seismic Design Method of Earth Embankment With

Seismic Coefficient

BY

Yasuyuki KOGA1

* and Osamu MATSUO2 *

ABSTRACT

The current practices in the

seismic design method of earth

embankments both in Japan and the U.S.

are briefly described, and future

research needs are identified. Next,

some results of experimental and

analytical study which has been

conducted to improve the seismic

coefficient method as a seismic design

method for earth embankments are

presented

.

KEYWORDS : Seismic stability,

Embankment, Seismic coefficient,

Liquefaction

1 . INTRODUCTION

During the past few decades, there

have been major advances in our

knowledge of the seismic behavior of

earthen embankments, such as

embankment dams, river levees, and

road embankments, through the

continuous efforts of strong

earthquake motion observation,

experimental and analytical studies.

As a result of such efforts, the

knowledge of liquefaction of subsuface

ground during earthquakes and its

associated damages to earth structures

has remarkably progressed. Some of the

results have already been introduced

in design codes in Japan. However, the

seismic design method for embankment

structures does not seem to have been

established but much yet remains to be

improved

.

This paper first outlines the

current seismic design methods for

embankment structures in Japan,

compares with those in the U.S., and

identifies future research needs

concerning the seismic design methods.

In the latter part of this paper, the

study on the seismic coefficient

method as a seismic design method for

embankment structures is presented.

2. OUTLINE OF SEISMIC DESIGN FOR

EMBANKMENTS

2.1 Practices in the Seismic Design

Methods in Japan

In this paper, "embankment

structures" generally indicate

embankment dams, road embankments, and

river levees. In Japan, "Dam Design

Standard" published in 1957 was the

first design code which specifies

seismic design for embankments.1 ' This

design code adopted the pseudostatic

analysis method combining the sliding

surface method and the seismic

coefficient method. Although the code

has since been revised several times,

there is no substantial change in the

seismic design method. The current

edition revised in 1978, however,

recommends to conduct, in addition to

the pseudostatic analysis, dynamic

analysis and/or model vibration tests

to ensure the seismic integrity of

dams V

For road embankments and river

levees, the seismic design has not

generally been applied. But some codes

require to maintain seismic stability

exceptionally for high embankments

which are founded on soft ground.

Accordingly, there recently appear

some cases that seismic design or

investigation is individually applied

to embankments in the metropolitan

area

.

Table 1 . summarizes current design

codes in Japan which specify the

seismic design method for earth

structures. This table includes codes

for various types of earth structures

other than those for embankment

structures. It is shown in the table

that:

(1) All of the codes adopt

pseudostatic analysis method as

seismic design procedure.

(2) The analysis is based upon the

sliding circle method and the

Modified Fellenius method is

adopted for the slice calculation.

(3) Design seismic coefficient kh is

basically given by Eq.(1), which is

commonly used for other engineering

structures such as bridges and

retaining walls.

kH = Ai A2 A, k0 (1)

/ Head, Soil Dynamics Division,

Construction Equipment and Method

Department, Public Works Research

Institute, Ministry of Construction,

Japan

,

2) Senior Researcher, ditto.

121
where

k0 : standard horizontal seismic

coefficient

,

A^ seismic zone factor,

A2 : ground condition factor,

A3 : correction factor depending on

type, use, material property of

the structure.

The value of the standard

horizontal seismic coefficient k0

is

empirically given as 0.2 in most codes

and 0.15 in some codes. The seismic

zone factor takes one of the values

of 1.0, 0.85, or 0.7, depending on the

seismic activity of the area. The

ground condition factor a, is given as

0.9. 1.0, 1.1, or 1.2, depending on

the stiffness or fundamental period of

the ground. The factor A3 takes 1.0 as

a general principle, but code-1 in

Table 1 gives A3 = 0.5-0.7 for

embankment structures. This reduced

value is considered to be due to the

material characteristics of earthen

embankment that soil is more ductile

than steel or concrete. In other

words, soil does not necessarily fail

at the time of peak stress application

but usually fails due to the cyclic

loading during earthquakes. The value

of A3 =0.5-0.7, however, is at most the

empirical one.

Since the 1964 Niigata Earthquake,

it has been widely recognized that

liquefaction of the ground

significantly affects the seismic

stability of embankments. Hence, most

of the codes shown in Table 1 require

to assess liquefaction potential of

the ground at sites where

susceptibility of liquefaction is

suspected

Codes-6, 7, 9 and 11 adopt safety

factor calculation formula in which

effect of liquefaction is directly

considered. In the formula, the

shearing strength of saturated soil

during earthquakes is given by Eq.(2).

According to the equation, shearing

strength is reduced due to the build

up of porewater pressure, which is

usually estimated from the

liquefaction resistance factor Fl.

Ti = c* + (<Xo' - Au) tan<*>' (2) where

7-, : shearing strength of saturated

soi 1

,

c', <t> : shearing strength parameter in

terms of effective stress,

Oo : initial effective confining

stress

,

Au : excess porewater pressure due

to earthquake loading.

In seismic stability analyses of

embankments by most Japanese design

codes which consider liquefaction

potential of saturated sandy deposits,

liquefaction strength of the deposit

is given by Eq.(2) using excess

porewater pressure. The only exception

is code-1 1, which allows the usage of

dynamic strength based on generated

strain without using excess porewater

pressure in addition to the above

mentioned soil strength.-*

As described above, it is rather

rare at present to perform a seismic

design or investigation for road

embankments and river levees on the

basis of design codes. Meanwhile,

there are some cases that a detailed

investigation which applies design

code method and seismic response

analysis method is performed for the

particular newly planned and/or

existing embankments. For embankment

dams, static shear strength parameters

are generally used.

Finally, with regard to embankment

dams, which are defined as those with

a height of 15m or more, the design

procedure is based on the seismic

coefficient method, not necessarily

considering liquefaction-induced

failure. This is guaranteed by the

fact that the liquefaction

susceptibility is precluded by removal

of loose sandy strata and wellcompaction of the embankment during

construction. Further, it is verified

that the above design procedure with

careful construction would be a conservative one because the dams

constructed that way have never been

damaged during past earthquakes.

2.2 Practices in the Seismic Design

Methods in the United States

In this section*, current practices

in the seismic design of embankment

dams in U.S. are briefly summarized

based on materials available to the

authors

The report , "Safety of Dams Flood

and Earthquake Criteria '," which was

made up by the Committee on Safety

Criteria for Dams established in the

National Research Council, provides

valuable information on the current

practices in U.S. According to the

report, it seems that many of the

federal agencies responsible for dam

construction have been adopting the

dynamic analyses, liquefaction

potential analyses and permanent

deformation analyses instead of

pseudostatic analyses based on the

seismic coefficient method. In

particular, it is noteworthy that the

U.S. Army Corps of Engineers rules out

the seismic coefficient method. As for

the state governmental agencies, not

all the states furnish the earthquake

122
design provision. Especially, many of

the states located in low seismic

activity zones do not prepare the

provision or use the conventional

seismic coefficient method. In

contrast, some states located in high

seismic activity zones require dynamic

analyses and liquefaction potential

assessment in addition to the

pseudostatic analyses. In California,

the state where the seismic activity

is the most active in U.S., there are

cases that the post-earthquake

stability assessment of an existing

dam against flow failure was performed.

2.3 Summary of Seismic Design

Practices in Japan and U.S

The current practices in the

seismic design of embankment

structures both in Japan and U.S. may

be summarized as follows:

In Japan

(1) Seismic design provisions are now

under preparation even for relatively

small-size embankment structures such

as river levees, road embankments and

the like.

(2) Many codes require liquefaction

potential analyses.

(3) Pseudostatic method with circular

sliding method and seismic coefficient

is mainly used.

(4) In addition to the conventional

pseudostatic method, the partly

improved method, which takes loss of

shear strength due to the build up of

porewater pressure during earthquakes

into account, is adopted in many

design codes.

In the United States

(1) Seismic design procedure used is

shifting from the conventional

pseudostatic method to dynamic

analysis, liquefaction potential

analysis, and permanent deformation

analysis

.

(2) Some agencies are adopting such

up-to-date seismic design procedures

as the post-earthquake stability

analysis against flow failure.

2.4 During- and Post-Earthquake

Stability Analysis

It has been recognized that there

are two types of eartquake-induced

failure in terms of time when the

failure occurs; i.e., during- and

post-earthquake failure. Because these

two types of failure have different

mechanisms, their failure potential

should be evaluated with different

procedures. Dur ing-earthquake

stability analysis includes the

pseudo-static analysis and permanent

deformation analysis, and is concerned

with finite damages. On the other

hand, post -ear thquake stability

analysis includes the pseudo-static

analysis using steady-state strength ' and is concerned with flow failure.

The latter may be required for such

embankment structure that the

embankment itself or the sandy strata

in the underlying ground is of a fairly low density.

In Japan, however, there have been

very few cases of flow failure in

spite of frequent earthquake

occurrence. This probably is caused by

the following reasons: that the sandy

layer existing in a high seismicity

area may have been densified by

frequently experiencing earthquakes

since its formation; and that most of

the embankment dams which have been

constructed are fairly well-compacted.

On the other hand, in Japan,

earthquake damages to road embankments

or river levees are considered to

have much social impact to the

economic activities or potential loss

of life, even if the damages are

within the moderate range. Thus, the

procedure which can predict the extent

of damage is needed.

2.5 Future Research Needs

Based on the discussions above,

the future research subjects which are

needed to establish the reliable

seismic design procedure for

embankment structures may be as

follows:

(1) On the during-earthquake stability

evaluation

(1-1) Improvement of pseudostatic

analysis method using seismic

coefficient

-- Methodology on how to

determine the dynamic shear

strength of soil and the

magnitude of seismic

coefficient is particularly

* important.

(1-2) Improvement of permanent

deformation analysis techniques

-- Representation of soil stressstrain relation should be

essential both in the

conventional technique and in

the rigorous technique based

on the elasto-plastic approach.

(2) On the post-eathquake stability

evaluation

123
-- Development of stability

analysis technique using steadystate-strength of soil

-- Development of assessment

technique to identify if steadystate is reached is particularly

required

(3) Development of reliable

measurement techniques of soil

density and dynamic shear strength

3. P ON THE PSEUDOSTATIC ANALYSIS

METHOD WITH SEISMIC COEFFICIENT

In this chapter is presented a part of the study that has been

carried out in the PWRI to improve the

conventional pseudostatic analysis

method. The purposes of this study

are

(1) To clarify how to determine the

dynamic shear strength of soil and

seismic coefficient to be used in

the analysis, and

(2) To correlate the safety factor

with the damage extent of the

embankment

Case studies of river levees

damaged by an earthquake and model

shaking table tests as objective case

histories are described in the

following sections.

3.1 Shear Strength of Soils for

Seismic Stability Analysis

3.1.1 Types of Shear Strength of Soils

Table 2 is a summary of shear

strengths for seismic stability

analysis of embankments. Among them,

following three strengths frequently

used are investigated.

St : Ti = (Tn'o tan<f>' (3) S2 : T( = ((Tn'o-u8) tan<*>' = (Jn'o(1 -ue/an 'o) tan* .... (4) S3: n = On 'o tan</>D = <7n 'o R(N. Ti) (5)

where

(7n'0

: initial effective confining

stress

,

<t>': shearing resistance angle,

0d: dynamic resistance angle,

R: dynamic strength ratio,

u9 : excess porewater pressure due

to earthquake loading.

Strength s3 is called as a dynamic

strength, which is determined as a magnitude of dynamic stress to cause a reference strain 7i under a certain

number of loading cycles. Strength S2 is based on the concept that shearing

strength is reduced through buildup of

excess porewater pressure due to

earthquake loading, whose detail has

been described in Section 2.1.

3.1.2 Damage Examples

Analyzed are three sections of

river levee which was damaged during

the 1983 Nihonkai-chubu Earthquake,

liquefaction of the foundation being

the cause. These three sections were

100m apart with each other. While the

configurations of the embankment

sections were the same, the thickness

and the liquefaction potential of the

sandy layers were different, as shown

in Fig . 1

3.1.3 Calculation Method

Stability analyses were conducted

using circular sliding method of the

modified Fellenius method (Eq.(6)).

F RETfJ

s E(WRsina + kWy)

_

RE{c'£ + (W-u b)cosa tan*'} (6) E(WRsin« + kWy)

where

Tt : shearing strength of soil,

c', shearing strength parameters in

terms of effective stress,

W: weight of slice,

k: seismic coefficient,

1: length of sliding arc,

u: porewater pressure,

b: width of slice,

R: radius of sliding circular arc,

a; gradient of sliding arc to

horizontal,

y: vertical distance between

gravity center of slice and

center of sliding circle.

(refer to Fig. 2)

3.1.4 Stability Analysis Results

In Fig. 3 are the stability

analysis results which show the

relationship between seismic

coefficient k and safety factor Fs for

three sections. In the figure, S2

indicates results of calculation

without seismic coefficient, and S2*

vice versa. Calculation results

indicate that the safety factor Fs

depends much on the strength used. In

both cases of S

2 and S2

1 , Fs decreases

remarkably when k exceeds a certain

value, which reflects that the excess

porewater pressure is very sensitive

to the seismic coefficient k.

Moreover, estimated maximum

response accelerations from seismic

response analyses are shown as a max in

the figure. Following formula was used

to evaluate the seismic coefficient k.

124
k-C(Nc)xJfflM- (7)

where

Cr(Nc): equivalence coefficient,

i.e, coefficient to

convert response

acceleration history into

seismic coefficient,

g: acceleration of gravity.

Equivalence coefficient Cr was

evaluated as in the figure by means of

the cumulative damage concept. Seismic

coefficients calculated in this manner

are plotted in the same figure. The

validity of this method will be

discussed in the next section.

As a result, the safety factor

calculated for strength was fairly

large, which cannot explain the fact

that respective embankments were

damaged during the earthquake. Fig.

4

shows a relationship of safety factor

Fs for strengths S?, S2 1 , S3 and

observed settlements of the

embankments. All the relations appear

reasonable in that the settlement is

large when Fs is small. However, Fs

is too small in the cases where S2 and

St 1 were used, considering the fact

that some embankments in the vicinity

of these three sections were hardly

damaged. Therefore, it is considered

that Fs=1.0 cannot be a threshold of

stability when using S2 and 82'.

On the other hand, it seems

reasonable to use S3, as the

settlement of embankment of 6 m high

was approximately 50 cm corresponding

to Fs=1 . 0

.

3.2 Investigation of Seismic

Coefficient

3.2.1 Equivalent Seismic Coefficient

Consider first an embankment in

Fig. 5. The response acceleration a of

sliding block B and shear stress t at

reference point P on a sliding surface

during a random seismic motion will

show such time histories as (a-1 ) and

(b-1 ) in Fig. 6. If the embankment is

shaken as a rigid body, both time

histories a-t, T-t are similar. Next,

consider a damage to this embankment

against this seismic motion. Another

seismic motion with a cyclic time

history can cause a damage of the same

amount to the embankment. The time

histories of acceleration and shear

stress are shown in (a-2) and (b-2).

As the damage to an embankment is

essentially caused by the deformation

of soil elements of the structure, the

stress time history in (b-1) is

equivalent to that with amplitude Teq

and number of cycles Neq in (b-2). The

relation is governed by the fatigue

characteristics of a soil. The

amplitude ratio Cr between (b-1 ) and

(b-2) are obtained as below.

Cr = ^_ (8) Tmax On the other hand, according to

the assumption of rigid body motion,

the ratio of acceleration to shearing

stress remains constant as below.

SUax _ 3ea (9) Tmax Teq

Therefore, we get

-^a- = -^a_ = Cr (10)

3max Tmax That is, random response acceleration

time history (a-1 ) has been converted

to cyclic response acceleration (a-2)

by use of equivalence coefficient Cr

which is determined from fatigue

characteristics of soils and

equivalent number of cycles.

Moreover cyclic response can be

easily converted to static problem

then shearing stress Teq is generated

at point P on sliding surface by

applying seismic coefficient kQQ =aeQ /g

to the sliding block B in Fig. 5.

Consequently equivalent seismic

coefficient is given by the following.

keq = -^9- = Cr (11)

3.2.2 Model Shaking Table Test

A series of shaking table tests

were conducted for embankment model

shown in Fig. 7 by use of 6 types of

acceleration wave form as given in

Table 3. Cases 1, 2, 3 were excited

with sinusoidal waves of constant

loading cycles, 20, and different

frequency. Case 4, 5, 6 were excited

with random waves, which are shown in

Fig. 8.

As a consequence of shaking tests,

the relationship between the maximum

input acceleration and cumulative

crest settlement were obtained as in

Fig. 9. This figure shows that the

magnitude of the input acceleration to

cause a certain settlement was larger

for random waves than cyclic waves.

Meanwhile stability analyses for

the model in Fig. 7 using dynamic

strength specified with respect to

number of loading cycles, Nc=20, and

failure strain, 7i=5%, resulted in

Fig. 10. This shows that the critical

seismic coefficient k to give safety

factor Fs=1.0 equals to 0.19.

Combining the coefficient and

input acceleration shown in Fig. 8,

equivalence coefficient Cr is obtained

125
by use of Eq.(10). Fig. 11 shows the

relationship between the crest

settlement and Cr thus obtained. It

is not decisive how large the critical

settlement corresponding to Fs=1.0

should be adopted. If we take that

is 5-10 mm, then the equivalence

coefficient Cr is approximately 1.0

for cyclic loading and 0.3-0.6 for

random loading. In the figure are

plotted the equivalence coefficients

Cr, which were obtained for Nc=20 by

applying the cumulative damage

concept to the initial liquefaction as

a failure criterion. Both coefficients

from the concept and shaking table

tests coincide when critical

settlement Scrit =1 -2mm is adopted. As

the shear strain of the soil is small

at the time of initial liquefaction,

the crest settlement for this stage is

supposed also to be small. It can be

said that equivalence coefficients

agreed fairly well in that sense.

3.3 Conclusions

Applicability of sliding surface

method with seismic coefficient as

aseismic calculation method of

embankment was reviewed on the basis

of damage examples and model shaking

table tests of embankments.

Following results were obtained:

(1) In a sliding surface method with

seismic coefficient, it is adequate

to use dynamic strength in terms of

total stress based on reference

strain as the shear strength of

soils

(2) Seismic coefficient should be

obtained by multiplying equivalence

coefficient to the maximum

acceleration as shown in Eq.(11).

Equivalence coefficient can be

roughly estimated by use of

cumulative damage theory.

(3) The safety factor obtained using

above mentioned shear strength of

soil and seismic coefficient can be

related to the settlement of an

embankment

4. SUMMARY

The current practices in the

seismic design method of earth

embankments in Japan and the U.S. were

reviewed on the basis of the design

codes available to the authors. Then

the future research needs were

identified.

Among various seismic stability

analysis methods, the seismic

coefficient method, i.e., pseudostatic

analysis method, was investigated on

the basis of case histories. The

basis on how to determine the values

of dynamic soil strength and seismic

coefficient in the pseudostatic method

was obtained from the study.

REFERENCES

1 ) Japanese Committee on Large Dam : Dam Design Standard, 1957.

2) : Dam Design Standard, Second

revised edition, 1978.

3) Koga,Y. and Shimazu,A. and

Taniguchi,E. : Consideration of

Earthquakes in Highway Earthworks,

Proc. 18th UJNR Joint Meeting, 1986.

4) Committee on Safety Criteria for

Dams, National Research Council : Safety of Dams--Flood and Earthquake

Criteria, National Academy Press, 1985.

5) Poulos,S.J., Castro, G. and France,

J.W. : Liquefaction Evaluation

Procedure, Journal of Geotechnical

Engineering, Vol.111, No. 6 , pp. 772-792

June,1 985.

6) Koga,Y. and Matsuo,0. : Stability

Analysis of Seismically Damaged

Embankments, Proc. 3rd International

Conference on Case Histories in

Geotechnical Engineering, 1988.

7) Koga,Y. and Matsuo,0. : Seismic

Coefficient in Seismic Stability

Analysis of Embankment, Proc. of 9th

WCEE, Vol.3, pp. 103-108, 1988.

126
Table 1 Codes for earthquake-resistant design of earth

structures in Japan

No. Agency

Title of code

or standard

Obiective

earth structures

Shape of

sliding surface

Slice calculation method

Magnitude

of seismic

coefficient

Excess pore water pressure Minimum

safety factor Remarks

1 Public Works Research

Institute, Ministry of Construction

New earthquake

resistant design method

(tentative! (1977)

Embankment dam,

River dyke, Road embankment

Circular Fallen. us method

(composite circle, modified

Fellenius method. Wedge method)

A, A2 ko>« (0.5 -0.7)

|ko-0.2)

1.2 and over for damDynamic analysis is desirable for a large scale dam

A,: seismic zone factor Ai: ground condition

factor

2 Japanese National Committee on Large Dams

Dam design standard (2nd revised

edition. 19781

Empankment dam

of 1 5 meters high and over

Circular (curve and

straight line)

Fellenius method

(composite circle, wedge method and

modified Fellenius method)

Htgh seismicity zone: 0.15-0.25 0.12-0.20 Low seismicity zone: 0.12-0.20

0.10-0.15

- 1 .2 and over Ditto

3 Japan Housing

Corporation,

Industrial Relocation Areas Promotion

Corporation, Japan River Association

Revised disaster prevention flood control reservoir technical standards

(tentative), (1980)

Embankment dam

below 1 5 meters

high

Circular Fellenius method High seismicity zone: 0.15 Low seismicity zone: 0.12

4 The Japan Port & Harbor

Association (under the auspices of Ports and

Harbors Bureau, Ministry

of Transport)

Technical stan- dards for port and harbor

facilities (1979)

Slope of cohesive

soil

Circular Modified Fellenius method(Seismic coefficient by area) x (Subsoil

coefficient) x (Importance

coefficient)

- (Normally 1.3) Application of design

seismic coefficient gives excessive safety. Liquefaction to be studied

separately.

Slope of sand and

gravel

Straight line Straight line sliding surface methodDitto 1.0 and over

5 Agricultural Structure Improvement Bureau,

Ministry of Agriculture, Forestry and Fisheries

Design standards

for land improve* ment project plan (dam design)

(1981)

Embankment dam

of 15 meters and

over

Circular Fellenius method

(Wedge method,

seismic coefficient

circle method)

High seismicity zone:

0.15. 0.12 Low seismicity zone:

0.12, 0.10

- 1.2 and over Liquefaction to be studied

separately.

6 Industrial Location and

Environmental Protection Bureau,

Ministry of International Trade and Industry- Japan Mining Industry Association

Construction standards for tailings dam

(1982)

Tailings dam Circular Modified Fellenius method

High seismicity

zone:

0.15 Low seismicity zone:

0.12

considered (1.2 and over) Checking method is subject to handling of excess

pore water pressure. Safety factor will also be

subiect to change

accordingly.

7 Minamitama New Town

Development Office, Tokyo Metropolitan Government

Design standards

for housing site construction works

(1984)

Embankment with

slope height

7 meters and over. Cutting with slope

height 10 meters and over

Circular Modified Fellenius method

0.15 considered Embankment

1.1 Cutting- 1.2

8 The Japan Gas

Association

Guideline for buried LNG Tank

(1981)

Embankment

surrounding buried tank

Circular

(straight line, composite

sliding surface)

Modified Fellenius methodkh = aia?a-ko

<ko =0.15»

1.1 a) in design seismic

coefficient is a correction

factor in underground

portion. Liquefaction to be studied

separately.

9 Public Corporation for Housing and Urban Development

Guideline for Seismic design of housing site (tentative) (1984)

Land for housing (embankment, cut

slope and natural

slope)

Circular CompositeModified Fellenius methodkh = A,-A z .A,-k0 lko=0.2, A,=0.5)

considered 1.0

10 Japanese National Railway

Guideline for seismic design of railway embank- ment (tentative) (1983)Railway embank- mentCircular (Not specified) 0.2 1.2

11 Japanese Road

Association Manual for Road Earthworks

(1986)

Road embankment

on soft ground or

of high importance

Circular, composite

Modified Fellenius method

k»,= f t-fi-kho

(k ho = 0.15l

considered 1.0 v \ = A,

127
Table 2 Classification of soil strengths

Type Strength criteria Drainage

condition

**)

Loading

pattern

* * * 1 Stress path &

strength

Time of strength

mobilized

Corresponding

loading condition Reference

Static strength

(yi'lUIIU Mnnnfnnnnc KKJl IUU3

lo ading)

Mobilized maximum IIIU1I1 cfroeG allcba nnriAr ullUci

a monotonous

loading

Drained a A during earthquake

(high permeability)

Dead weight +

seismic force JNCOLD (1978)

Undrained a B,B' during earthquake

(impulsive earth- quake)

do do

Dynamic

strength

(Cyclic loading)

Maximum stress of stress time

history to cause

a reference strain during a cyclic

loading

Undrained b C during earthquake do Seed (1966)

Ishihara (1980)

Maximum static strength after a cyclic loading

Tnromnlptol v

drained*)

(Pore pressure during a cyclic loading

remains

constant)

n UUlUlg cdJT U ILJUdrvtr

&

soon after earthquake

tin

Dead weight

IMP / -t Q"70\

PWRI (1975)

HUDC (1984)

JMA (1980)

I Inrfrainprf*

)

E UUI I1IK CCU UlUUtfnC

&

soon after earthquake

Seismic force

Dead weight

Seed (1979)

Tokyo Metro. Govnt. (1983)

Seed (1979)

Drained*'

(Pore pressure returns to

initial pressure)

c F long after earthquake

do

N.B. * > This drainage condition refers to that for a static loading after a cyclic loading.

An undrained condition is assumed during a cyclic loading. **» Refer to Fig.Al ***> Refer to Fig.A2

Fig.A2 Stress paths Fig.Al Loading patterns

and shear strengths to obtain shear strength

128
(a) Shape of the embankment

(b) Soil profiles at three locations

Site A

e e

CO

I

2

3

4-

5

6

7

8

9

10

II

12

IVM

IS

IS

17

18

19

20

Fill

Fine sand

S P T

N - volue

0 10 20 30 40 50

Silly clo>

Site B

E E

Medium

\ iHDsl_

, Medium

\ sand

N - volue

0 10 20 30 40 50

•Fill

Fine sand

Coarse,

iond Medium

sond

Silly cloy

S P T

Silly cloy

Site C

E

o

CO

N - volue

0 10 20 30 40 50 FillMedium

sond

Medium

sond

Medium

sond

Cootse

sond

Medium

sond

Medium

sand

Silly cloy

S P T

/

Fig.1 Shape of the embankment and soil profiles of the

ground at three locations

129
Fig. 2 Key sketch for sliding surface method

o 3

<_> O~ 2 o

CO

1 1 1 1 I

1 1 1 1

1 r \ Site A

N. 5 >\

- \

2 \ /Asmax = X \ \Cr(20)

NC i

\ \

\ \.

(III

l60golV

= 0.678^

\ V

\ \\

. ... 1 .... 1 I...

0 0.05 U.IU U.ID U U.U3 U.IU UI3 (J U.UD U.IU U.ID

Seismic coefficient Kh ( =O.G78xKs) Seismic coefficient Kh ( = 0.689xKs) Seismic coefficient Kh ( = 0.689xKs)

T I I I I I I II I III i i I . i i I i i

I I i l r I I I I I I I l I I I I I I I I l I • • I I i lii I i 0 0.05 0.10 0.15 0.20 0 0.05 0.10 0.15 0.20 0 0.05 0.10 0.15 0.20

seismic coefficient K s

I : Asmox/g)

0.10 0.15

0.20 0 0.05 0.10 0.15 0.20

Surface seismic coefficient Ks ( =Asmax/g) surface seismic coefficient Ks ( : Asmax/g) Surface

Fig. 3 Seismic coefficient vs. safety factor

150

OO

E

i—I—i

1—

r

(Site C)

Strength S3

Strength S^' (Without seismic food

Strength S2

(With seismic load)

(Site B)

a, (Site A)

Ol—1—1—1 1 I 1

1 0 0.5 1.0

Safely (actor Fs

Fig. 4 Safety factor vs. crest settlement

130
Fig. 5 Key sketch for indicating the sliding block B and the

point P on arbitrary sliding surface

r Random t_ response *

(a-1)

Equivalization

r Cyclic i

« response *

(a-2)

( Qeq , Neq)

2

^Statical problem]

(a-3)

r. _ aeq *eq —W

(b-3)

(c-3)

Fig. 6 Transformation of irregular response problem to

pseudostatic problem

Sand ground 3

250

(Unit :cm)

Fig. 7 Model configuration

131
Table 3 Input acceleration wave forms

Case Wave Form Frequency f *> Duration td f xtd No. of stops

Max. Acc. Range

1 Regular (N=20) f= 2.5 Hz 8 sec 20 4 117 -409

2 do ( do ) f=10Hz 2 sec 20 10 172 - 542

3 do ( do ) f= 5 Hz 4 sec 20 6 90 - 358

4 Irregular (Shock Type) fo= 5 Hz 5 sec 25 14 94 - 991

5

. (Vibration d0 Type)

f0 = 5 Hz 5 sec 25 8 112 - 832

6 .Vibration . do ( Type' fo= 2.5 Hz 10 sec 25 7 95 - 774

*) f0 = Predominant frequency for irregular wave

gCase4 Hachinohe NS(1968,Tokachioki EQ, Main shock)—.

0 Case 5 Aomori EW (1968, Tokachioki EQ, Aftershock) I

.Case 6 Aomori EW (1968, Tokachioki EQ, Aftershock) !

—i 1 i_ » 10 Time (sec)

Fig. 8 Input acceleration wave form

Excitation Acceleration ain (gal)

0 200 400- 600 800 1000

1

E

9

CO

O

CO>

Eo

Fig. 9 Input acc. vs. cumulative crest settlement

132
0 0.1 0.2

Seismic coefficient : k,„

Fig. 10 Safety factor vs. equivalent seismic coefficient

.2o

«

uc

J2

CD>

'5

CT LU

1.0

0.5

Case Method

b c Remarks

1 O Regular

f „ 2_ 5Hz

wave

2 A do f=10Hz

3 do f=»5H*

4 € • ,rregu,ar Shoek Type wave

5 do Vibration Type

6 D do do

Cr-1.0

Case1 Case1.2.3(Method a) Case 2

Case 3

Case 6 (Method a) CaseS (Method a)

Cr- 0.46 — Case 4 (Method a)

Case 6

Case S

J_

5 10

Reference Crest Settlement 3 (mm)

Fig. 11 Failure criteria vs. equivalence coefficient

133
Seismic Monitoring of Buildings: Analyses of Seismic Data

by

M. Celebi*

ABSTRACT

The main objective in seismic instrumentation of structures is to facilitate response studies that lead to improved understanding of the dynamic behavior and the

potential for damage to structures under seismic loading. The purpose of this paper is: (1) to review the

status of the programs for strong-motion instrumentation of structures in the United States and discuss var- ious procedures and instrumentation schemes designed

to best acquire response data from buildings and (2) to

discuss preliminary results derived from recorded re- sponse data obtained from well-instrumented structures

during the recent earthquakes.

KEYWORDS: earthquakes, records, spectra, instrumentation, diaphragm, interaction

1. INTRODUCTION

The purpose of this paper is to provide some insight

into the objectives of programs for seismic monitoring

of structures and review the status-quo with examples

of records obtained from structures.

The main objective of any seismic instrumentation program for structural systems is to improve the understanding of the behavior, and potential for damage, of

structures under seismic loading. The acquisition of

structural response data during earthquakes is essential

to develop and confirm methodologies used for analysis

and design of earthquake-resistant structural systems.

This objective can best be realized by selectively instrumenting structural systems to acquire data on strong

ground-motion, and the response of structural systems

(buildings, components, lifeline structures, etc.). As a

long-term result one may expect design and construction practices to be modified to minimize future earthquake damage.

As part of its earthquake hazard reduction program,

the United States Geological Survey (USGS) has pursued a structural instrumentation program to further

structural response studies in selected targeted seismic

regions shown in the map in Figure 1. A general re- view of the USGS program objectives and procedures is provided by Celebi and others [1].

It is expected that a well-instrumented structure for

which a complete set of recordings has been obtained,

would provide useful information to:

check the appropriateness of the design dynamic model (both lumped mass and finite element) in the elastic

range;

determine the importance of non-linear behavior on

the overall and local response of the structure;

follow spreading of the non-linear behavior throughout the structure as the response increases, and investigate the effect of the non-linear behavior on fre- quency and damping;

correlate the damage with inelastic behavior;

determine ground motion parameters that correlate

well with building response and damage; and

make recommendations to improve seismic codes.

The primary factors affecting selection of structures for

instrumentation within the general USGS program are

related to structural systems and construction materials that are likely to be repeated and those that have

sufficient long-term engineering interest as well as siterelated factors. These factors and parameters are weighted in the decision-making process of ^election of structures for strong-motion instrumentation and can be cat- egorized as follows:

(a) structural parameters: materials of construction,

structural systems, geometry, discontinuities, importance, age, and interest.

(b) site parameters: probability of occurrence of a

large earthquake, proximity to faults, site conditions,

and expected damage and loss.

In the State of California, the most extensive program

for instrumentation of structures is being conducted by

the California Division of Mines and Geology (CDMG).

To date, approximately 100 structures in California have

been instrumented by the CDMG program [2]. The

CDMG program aims to monitor typical buildings and

structural systems according to a pre-established matrix reflecting structural types and heights. Therefore,

in California, the USGS program concentrates on instrumentation of non-typical structures of special engineering interest.

*U.S. Geological Survey, 345 Middlefield Road,

Menlo Park, CA 94025

134
SEATTLE

SAN

FRANCISCO

BAY

LOS ANGELES,

ORANGE COUNTY

BOSTON

Figure 1. Targeted seismic regions for the USGS instrumentation of structures

program.

Outside the United States, a significant number of buildings and other structures have been instrumented in

Japan by various governmental and private organizations [3]. Italy has embarked upon an impressive structural monitoring program that also encompasses historical monuments [4J. It is known that there are instrumented structures in New Zealand and the Soviet

Union.

2. CODE-TYPE INSTRUMENTATION

In the United States, independent of the USGS and

CDMG programs described above, various codes in ef- fect, whether nationwide or local, recommend different

quantities and schemes of instrumentation. The Uniform Building Code (UBC) [5] recommends that for

the two top seismic zones (Seismic Zones 3 and 4) a minimum of three tri-axial accelerographs be placed in

every building over six stories in height, with an aggregate floor area of 60,000 square feet or more, and

in every building over 10 stories in height, regardless

of floor area. According to this recommendation, the

tri-axial accelerographs are typically located in a building as shown in Figure 2a. The City of Los Angeles

adopted the UBC recommendation in 1966—thus en- abling numerous sensors in many buildings to record

the motions during the 1971 San Fernando earthquake.

However, in 1983, the City of Los Angeles changed the

requirement of three accelerographs to only one—to be

placed at the top of those buildings, meeting the UBC

criteria described above.

Experience from past earthquakes, as well as the 1971

San Fernando earthquake, shows that the instrumentation guidelines given by the UBC code, for example,

although providing necessary data for the limited analyses projected at the time, do not provide sufficient data

to perform the model verifications and structural analyses now demanded by the profession. For example,

torsional or rocking motions of a structure cannot be

identified unless the structure is properly instrumented

in key locations to obtain the necessary data. Typically, an extensively-instrumented building should contain sensors in key locations to record translational,

torsional as well as rocking, motions of the structural

system as seen in Figure 2b. Whenever physically possible, an extensive instrumentation scheme should include a free-field station. Extensive instrumentation for

special cases such as buildings with flexible diaphragms

and base-isolators requires variations in the schemes of

instrumentation and are schematically shown in Fig- ures 2c and 2d.

A recent example of records obtained during the 1 October 1987 Whittier-Narrows earthquake from UBCtype recommended instrumentation at the 900 South

Fremont Building in Alhambra, California is shown in

Figure 3. From these records, we can clearly identify

the fundamental periods of the twelve-story building in

its translational mode in the two horizontal axes at appoximately 2 seconds. This is demonstrated in Figure 4

showing the same recorded tri-axial accelerations (vertical, north-south and east-west directions) at the basement, 6th floor and 12th floor and the Fourier spectrum

corresponding to each component. However, it is impossible to identify non-translational contributions (e.^.,

torsional) and/or interaction effects {e.g., rocking), if any, to the behaviour of the building—particularly when,

as in this case, the tri-axial accelerographs were located near the center of rigidity of the building.

135
UBC RECOMMENDATION

IDEAL EXTENSIVE INSTRUMENTATION

TWO PARALLEL SENSORS

AT A DISTANCE APART TO

EVALUATE TORSION (ALL TO

EVALUATE TRANSLATION)

VERTICAL SENSORS IN THE BASEMENT TO

EVALUATE ROCKING

EXTENSIVE INSTRUMENTATION

EXTENSIVE INSTRUMENTATION SPECIAL STRUCTURES

FLEXIBLE DIAPHRAGM BASE-ISOLATED BUILDING

Figure 2. (a) Instrumentation according to Uniform Building Code recommendations, (b) An ideal extensive instrumentation for a regular building, (c)

Instrumentation to record response of a flexible diaphragm, (d) Instrumentation to record response of a base-isolated building.

ACCELERATION RECORDS FROM UBC TYPE RECOMMENDED INSTRUMENTATION

AT

900 SOUTH FREMONT BUILDING (ALHAMBRA)

(WHITTIER-NARROWS EARTHQUAKE—OCTOBER 1, 1987)

VERTICAL HORIZONTAL (380 OEQREES) HORIZONTAL (SO DEGREES)

Figure 3. Acceleration records obtained at the 900 South Fremont Building (Alhambra) during the Whittier-Narrows earthquake of October 1, 1987

(Code-Type Instrumentation).

136
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ure 4. Acceleration records obtained at the 900 South Fremont Building

(Alhambra) during the Whittier-Narrows earthquake of October 1, 1987

and the corresponding Fourier spectra.

137
While UBC recommendations are deemed not to be sufficient, nonetheless, records obtained during the 1971

San Fernando earthquake from buildings in downtown

Los Angeles instrumented in accordance with UBC rec- ommendations were used to derive relationships for determining fundamental periods of structures incorporating non-linear behavior [6]. These relationships have

been incorporated into ATC-03-06 code provisions [7] and the 1988 edition of UBC [5].

3. EXTENSIVELY INSTRUMENTED STRUCTURES

Imperial County Services Building is perhaps the source

of the most widely studied building-response data. During the 1979 Imperial Valley earthquake (M3 = 6.5), the

building sustained substantial damage, primarily to its soft first story, and as a result was later demolished.

The building was extensively instrumented with thirteen channels of accelerometers, a central recorder, and

a three- component free-field accelerograph. The locations of sensors throughout the building are seen in Fig- ure 5. The following conclusions were derived from the

study of the complete set of records (Figures 6 and 7) [8]:

the motion of the ground floor was affected by soilstructure interaction, as shown by the agreement of

the data obtained with mathematical models that incorporate soil-structure interaction,

the peak acceleration was approximately 50% higher

and the nature of motions considerably different at the ground-floor level in comparison to the free-field

motions,

during the earthquake, the fundamental period of the

building increased significantly (in the order of 150%),

compared with the pre-earthquake periods determined

from ambient vibration tests. Such drastic changes in

the fundamental dynamic characteristics can only be

attributed to the inelastic behavior of the structure

during the earthquake,

the initiation of the inelastic behavior and structural

damage as apparent on the records and subsequent

analyses reconfirmed that soft first stories exemplified

in the Imperial County Services Building design are vulnerable during earthquakes unless specific design

considerations are implemented.

Significant results have also been deduced from the study

of the records obtained, during the 1979 Imperial Valley

earthquake, at the Meloland Road-Interstate Highway 8

crossing [9]:

peak accelerations recorded on embankment sites adjacent to each abutment were up to 30% higher than

those recorded at the central column, indicating that

the embankment fill material influenced the motion

of the abutment,

relative motion between the abutments and the sur- rounding fill material was evident,

the recorded deck motions showed rotational as well as translational components.

During the M3 = 6.1 Morgan Hill earthquake of 24 April

1984, a complete set of acceleration records depicting

the behavior of a flexible roof diaphragm were obtained

from the West Valley College gymnasium in Saratoga,

California. The gymnasium was instrumented by the

California Division of Mines and Geology (CDMG) [10,

The roof diaphragm of the 34.14 m x 43.90 m rectangular, symmetric gymnasium consists of 0.95 cm plywood

over tongue-and-groove sheathing attached to steel truss- es supported by reinforced concrete columns and walls.

A three-dimensional schematic of the gymnasium and

the locations of the eleven channels of force-balance ac- celerometers (connected to a central recording unit) and

associated directions are provided in Figure 8. Only the

main lateral load-resisting elements (shear walls) are depicted in the figure; however, the finite element model

(also in the figure) developed for dynamic analyses took

into account the columns as well. The recorded acceleration responses are shown in Figure 9. There were no

free-field instruments at the site. Three sensors placed

in the direction of each of the axes of the diaphragm facilitate the evaluation of in-plane deformation of the diaphragm. Other sensors placed at ground level measure

vertical and horizontal motion of the building floor, and

consequently allow the calculation of the relative motion

of the diaphragm with respect to the ground level.

The following conclusions were derived based on studying the peak accelerations, relative displacements and

Fourier spectra of the records obtained at the roof diaphragm level [12, 13]:

the recorded peak accelerations in the N-S and E-W

directions of the ground floor as well as the diaphragm

differ considerably. The recorded peak ground-floor

acceleration in the N-S direction (perpendicular to

the narrower 34.14 m dimension) is 0.10 g while in

the E-W direction (perpendicular to the 43.90 m dimension), it is 0.04 g. Consequently, the displacements in the stronger (N-S) direction are larger than

those in the weaker (E-W) direction. Such differ- ences in ground level accelerations are not generally

accounted for in design considerations. The wall motion at the roof level is 1.5 times the amplitude of the

floor level, while the horizontal motion of the center

of the roof is 3 times the motion at the edges during

the resonances in the two horizontal axes of the roof

diaphragm evaluated 3.8 Hz and 3.9 Hz, respectively.

The fundamental frequencies of the diaphragm are determined from the recorded accelerations to be 3.8

and 3.9 Hz in the N-S and E-W directions, respectively.

at a frequency of about 4 Hz, the center of the roof

diaphragm attains a peak acceleration of 0.42 g and

0.20 g in the N-S and E-W directions, respectively.

The corresponding roof-edge peak accelerations are 0.14 g

(average) and 0.065 g

(average) in the N-S and

E-W directions, respectively. Therefore, there is an

amplification of motions at the center of the roof as compared with the edges by a factor of approximately

3 for both directions. From ground floor level to the

edges, the same factor is approximately 1.5 in both

directions.

138
E-W SECTION

ACCELEROMETEfK

4 th FLOOR

ROOF

6 th

5th

4 th

3rd

2nd

GROUND

FLOOR

SHEAR-WALL

t© ®^ ©t

1 FREE-F ACCELEROGRAPH

2nd FLOOR GROUND FLOOR

Figure 5. Floor plans and typical vertical section of the Imperial County Services

Building. Locations of the sensors are identified.

IMPERIAL VALLEY EARTHQUAKE OCTOBER 15. 1979

(M L =6.6, M s

=6.5)

MOTIONS RECORDED IN THE STRUCTURE

A « g. a, ROOF (NROOF (N-S)

ROOF (N-S)

ROOF (E-W)

S)

fi\>:J\jv;A/VX~'

GROUND FLOOR (E-W)

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4TH FLOOR (E-W)

2ND FLOOR (E-W)

2ND FLOOR (N-S)

-2ND FLOOR (N-S)

2ND FLOOR (N-S)

GROUND FLOOR (N-S)

GROUND FLOOR (N-S)

-GROUND FLOOR (UP)

a K> TIME. IN •ECOND8

14

Figure 6. Acceleration records obtained at the Imperial County Services Building

during the Imperial Valley earthquake of October 15, 1979.

139
IMPERIAL VALLEY EARTHQUAKE OCTOBER 15, 1979

(M L= 6.6, M s

= 6.5)

COMPARISON OF FREE-FIELD AND GROUND FLOOR MOTIONS

AT THE IMPERIAL COUNTY SERVICES BUILDING

NORTH

Free field .24g

..29 8 Ground floor

UP

Free field

Ground floor

.190

EAST

Free field

Ground floor .24 8

^—- 32g

8 10 12 14

TIME, IN SECONDS

16 18 20

Figure 7. Comparison of free-field and ground floor motions recorded at the Imperial County Services Building during the Imperial Valley earthquake

of October 15, 1979.

140
furthermore, a very simple finite-element model consisting of beam elements (columns and shear walls)

and plane-stress elements (diaphragm) [Figure 8], is used to match the fundamental frequency of the roof

diaphragm. Within acceptable tolerances, the calculated displacement of the center of the diaphragm

with those from recorded motions are matched for

5.0% damping as seen in Fl0 -ire 10 which also shows

the Fourier spectra derived from calculated responses.

the calculated or the recorded displacements do not

compare well with those from either the ATC-7 rec- ommended formulas [14] or by calculations based on

UBC seismic forces [5]. In addition, the recorded or calculated first mode of the structure is that of the diaphragm only and the frequencies, recorded or calculated, do not match with the frequencies determined

from standard period formulas for structures. Therefore, in the design-analysis process, the standard code

formulas for building periods should not be used to

predict the periods of buildings with large-span diaphragms.

4. TORSIONAL RESPONSE OF A UNIQUE

BUILDING

The 483 ft. (147 m) tall, 32-story 1100 Wilshire Finance

Building in Los Angeles, California was instrumented in

1986 during its construction. The unique building plan

abrubtly changes at the 12th level from a nominally

rectangular shape to a triangle. General dimensions and

the 21-channel instrumentation scheme of the building

are provided in Figure 11. The building sits on spread

footing founded on shale. Scaled horizontal acceleration

time-histories at different levels of the building obtained

during the October 1, 1987 Whittier-Narrows earthquake are provided in Figure 12. The records show that

the motions at each level in the 208 direction at the NW

corner of the building differ considerably from those in

the SE corner. These differences indicate that torsional

motions contribute significantly to the translational motions (e.g., the peak accelerations of parallel motions are

0.19 g and 0.10 g on the 32nd floor and 0.26 g and 0.12 g

on the 13th floor). The difference between the NW208

and SE208 accelerations at each instrumented level represents the rotational acceleration at that level (after

division by the perpendicular distance between the two

sensors). The recorded accelerations in the NW208 direction, the representative rotational acceleration (the

difference between NW208 and SE208 accelerations),

and corresponding Fourier spectra (all plotted to the

same scale) for each of the top three levels are shown

in Figure 13 and in Figure 14 for the ground level and

the basement. The Fourier spectra of the total translational and the representative rotational motions at these

top three levels show that the superstructure responded

with a closely coupled torsional mode at a predominant frequency of approximately 0.8 Hz and another

at 0.7 Hz. Furthermore the Fourier spectra amplitudes

and the comparison of the amplitudes of the translational and torsional contributory accelerations show the

significant level of torsional energy at the top three instrumented levels; however, the level of energy reduces

considerably at the ground level and is not significant

at the basement.

While further study of the records from this building

are currently underway, it is important to note that the

frequencies derived from the records presented herein

do not in any way compare with the frequencies used

in the design-analysis process. For example, the fundamental frequency of the building has been calculated

(using fixed-base assumption) as 4.9 s (or approximately

0.2 Hz). We draw the following conclusions from the

study of the motions recorded:

higher modes dominate the response of the building

to this earthquake.

the dominant behavior of the building is a coupled

torsional- translational mode (the torsional contribution to the translational motions recorded is very significant).

5. CONCLUSIONS

The objectives of seismic instrumentation of structures

and various programs are reviewed in light of lessons

learned from studying response data obtained from structures during past earthquakes. UBC-recommended instrumentation is discussed and compared with extensive

instrumentation that permits complete response study

of structures. It is shown that code-type instrumentation does not allow detection of torsional or rocking

motions of a structure.

It is important to invest in seismic monitoring of structures. We have learned from study of the records obtained from instrumented structures. Well-instrumented

structures can provide information on the real-life laboratory behavior of a structure that might not be retreviable using only the code- type recommended instrumentation, nor could it always be predicted during the design/analyses processes which requires certain idealizations. Examples presented show (a) that soft first stories

require careful consideration, (b) that closely coupled

higher torsional modes dominate the response of a vertically irregular structure, and (c) that the behavior of a flexible diaphragm can be identified by appropriately

recording its response. These results may be significant

in future analyses and modeling of such structures.

6. REFERENCES

1. C/elebi, M., §afak, E., Brady, G., Maley, R., and Sotoudeh, V., 1985, Integrated instrumentation plan

for assessing the seismic response of structures—

A

review of the current USGS program, USGS Circular 947.

2. Huang, M., Ventura, C, and Shakal, A., 1989,

Strong-motion records from buildings, advance copy

of paper submitted to ASCE Structures Congress,

San Fransisco, CA, June 1989.

3. Watabe, M., and Nagata, M., 1989, Seismic instrumentation and its implications for high-rise buildings in Japan, advance copy of paper submitted

to ASCE Structures Congress, San Fransisco, CA,

June 1989.

141
4. Basin, M., and Bongiovanm, G., 1989, Seismic instrumentation of buildings in Italy, advance copy

of paper submitted to ASCE Structures Congress,

San Fransisco, CA, June 1989.

5. Uniform Building Code, International Conference

of Building Officials, Whittier, CA, 1970, 1973, 1976,

1979, 1982, 1985, 1988 editions.

6. Mulhern, M. R., and Maley, R., 1973, Building period measurements before, during and after the San

Fernando earthquake, in San Fernando, California

earthquake of February 9, 1971, U.S. Dept. of Commerce, NOAA, vol. I, Part B, pp. 725-733.

7. Applied Technology Council, 1978, Tentative provisions for the development of seismic regulations

for buildings, Applied Technology Council—ATC

3-06.

8. Rojahn, C. and Mork, P. N., 1982, An analysis

of strong motion data from a severely damaged

structure—The Imperial County Services Building,

El Centro, California, in The Imperial Valley, California, earthquake of October 15, 1979: U.S. Geol.

Surv. Prof. Pap. 1254.

9. Rojahn, O, Ragsdale, J. T., Raggett, J. D., and

Gates, J. H., 1982, Main-shock strong-motion records

from the Meloland Road-Interstate Highway 8 overcrossing, in The Imperial Valley, California, earthquake of October 15, 1979: U.S. Geol. Surv. Prof.

Pap. 1254.

10. Huang, M. J., Shakal, A. F., Parke, D. L., Sherburne, R. W., and Nutt, R., 1985, Processed data

from the strong motion records of the Morgan Hill

earthquake of 24 April 1984—Part II: Structural

response records, Office of Strong Motion Studies,

Report. No. OSMS 85-04, CDMG-SMIP, Sacramento, CA.

11. Shakal, A. F., Sherburne, R. W., and Parke, D. L.,

1984, CDMG strong motion records from the Morgan Hill, California, earthquake of 24 April 1984,

Office of Strong Motion Studies Report No. OSMS

84-7, CDMG, Sacramento, CA, 101 p. 12. Celebi, M., Brady, G., §afak, E., and Converse, A.,

1986, Performance of an earthquake-excited roof

diaphragm, Proc. ASCE Conference on Dynamic

Response of Structures, U.C., Los Angeles, March

31-April 2, 1986, pp. 863-870.

13. Celebi, M., Bongiovanni, G., §afak, E., and Brady,

G., 1989, Seismic Response of A Large-Span Roof

Diaphragm, in print Earthquake Spectra (May 1989

issue).

14. ATC-7, 1981, Guidelines for the design of horizontal wood diaphragms, Applied Technology Council,

Palo Alto, CA.

142
Figure 8. Three-dimensional schematic of the West Valley College Gymnasium

(Saratoga, California) and the locations and directions of eleven channals of force-balance accelerameters. Only the main lateral load resisting

shear walls are shown (top figure). The finite element model (quarter of

the gymnasium-bottom figures) shows the columns as well as the shear

walls all of which are modeled as beam elements.

143
Structural Response - Area I

144
MORGAN HILL EARTHQUAKE

WEST VALLEY COLLEGE GYMNASIUM

CENTER OF ROOF DIAPHRAGM

MORGAN H ILL EBRIHOuBbE : SBRB I 0GB CII NNHSi-l U« CnBN 7 IMOOC 21 M5l 0 1 SPlBCCMCNI FROM RCCORDCO BCCCl

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Figure 10. Recorded and calculated displacements at the center of the roof

diaphragm in the N-S and E-W directions respectively. The calculated

displacements are from dynamic analyses using the finite element model

and ground level motions as input. The Fourier spectra corresponds to calculated displacements.

145
ORIENTATION

OF SENSORS

Figure 11. General dimensions and locations as well as orientations of the strongmotion instrumentation scheme of 1100 Wilshire Finance Building (Los

Angeles, California).

146
147
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13. On the left column are the NW208 component accelerations and their

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12th levels respectively. On the right are the representative rotational

accelerations (differences between NW208 and SE208 recordings) and

their Fourier spectra corresponding to the same instrumented levels.

148
14 On the left column are the NW208 component accelerations and their

Fourier spectra (all plotted to the same scale) for the gound level and

the basement respectively. On the right are the representative rotational

accelerations (differences between NW208 and SE208 recordings) and

their Fourier spectra corresponding to the same instrumented levels.

149
Development of Advanced Reinforced Concrete Buildings

Using High-Strength Concrete and Reinforcement

by

Tatsuo Murota* and M

ABSTRACT

This paper describes the outline of a MOC's

new research project on reinforced concrete

buildings using high-strength and highquality concrete and reinforcement. The

project is carried out by BRI from 1988 to

1992. Concrete of 300 to 1200 kgf/cm 2

strength and reinforcement of 4000 to 12000

kgf/cm2 yield strength are treated in the

project. Guidelines for structural design

and construction of reinforced concrete

buildings using concrete of less than 600

kgf/cm2 strength and reinforcement of less

than 6000 kgf/cm2 yield strength are expected

as fruit of the project.

KEYWORDS : high-strength concrete, highstrength reinforcement.hlgh-rise reinforced

concrete building.

1. INTRODUCTION

Recently, research and development on

reinforced concrete buildings using highstrength concrete has been actively promoted

in Japan. Based on these activities, highrise RC buildings which have images

remarkably different from those of common RC

buildings have come to appear in Japan.

Technology development on such a new field of

RC building construction seems to be more

actively promoted in the future and it is sure that the Ministry of Construction will

be requested to have technical standards for

building administration for such buildings,

sooner or later.

Based on the background, the Ministry of

Construction (MOC) decided to manage a fiveyear research project titled as "Development

of Advanced Reinforced Concrete Buildings

Using High-Strength Concrete and

Reinforcement" (hereinafter 'New RC). New

RC Project started in 1988 fiscal year with

major objectives as follows:

(1) To survey and analyze results of research

and development conducted by private

companies using a proper common measure to

establish a technology system for advanced

reinforced concrete buildings (New RC

buildings). If necessary, to execute

basic researches for the systematlzation.

(2) To understand the mechanical behavior of

buildings using high-strength concrete and

steel bars and to develop design systems

suitable for New RC buildings.

Results of the project will be utilized as technical standards for building

Hlrosawa

administration and will encourage further

research and development on advanced

reinforced concrete buildings.

2. RANGE OF MATERIAL STRENGTH

Strength of concrete and reinforcement

treated in the project ranges from 300 to

1200 kgf/cm2 (4300 to 17400 psi) for concrete

and from 4000 to 12000 kgf/cm2 (57 ksi to 174

ksl) for reinforcement. In Fig.l major

research fields in the project are shown on a plane where vertical axis is yield strength

of reinforcement and horizontal axis is

concrete strength.

In the figure, small zones A and B surrounded

by dotted lines correspond to areas for

current common RC buildings and for high-rise

RC buildings of about twenty through thirty

stories constructed in the Ministei—ofConstruction's agreements specified in

Article 38 of the Building Standard Law.

respectively.

Zones I. II- 1, II-2 and III which are zones

treated in the project, cover very large area compared to two zones stated above.

Therefore It is obviously unreasonable to

think that we could understand mechanical

behavior of New RC buildings simply by

extrapolating ' the knowledge on current

reinforced concrete structures.

So. in the project, theoretical examination

of experimental data on current knowledge on high-strength concrete structures will be

emphasized to construct new technology

systems. This means that current technical

knowledge on reinforced concrete buildings

will also be re-examined in the project.

Because of the wide range of material

strength, the character of researches in the

project will be basic and the project may not

yield much practical results In some zones

far apart from the current reinforced

concrete technology. Practical results are expected in Zones I and II— 1. because these

zones are comparatively close to the boundary

of current technology.

Director. Structural Engineering Department,

Building Research Institute

Director. International Institute of

Seismology & Earthquake Engineering,

Building Research Institute

150
3. OBJECTIVES OF RESEARCH AND DEVELOPMENT

AND FINAL RESULTS EXPECTED

Table 1 shows five objectives of research and

development and the corresponding final

results expected in the project. Some of the

results will be applicable to refine current

reinforced concrete technology. Guidelines

for structural design or construction in the

table will not give full details of

technology but will give basic principles

that are not enough for practical use'. It Is because the guidelines which details

technical specifications for structural

design or construction often have a tendency

to Impede the development of relevant

technology, and also future technical

development will possibly change the highstrength concrete technology so much as we

can not predict.

4. RESEARCH ORGANIZATION

Building Research Institute is in charge of

conducting the project. Research committees

are set up in Japan Institute for

Construction Engineering (JICE) as shown In Fig. 2 to organize universities, private

companies, etc. Three committees are assigned

to determine research schemes, to coordinate

research works, to promote the project and to

integrate research results. The chairman of

these three committees is Prof. H. Aoyama.

University of Tokyo.

Technical Committees are in charge of making

research programs in detail. Implementing

research works and integrating research

results in five particular fields. The

chairmen of the technical committees are

Prof. F. Tomosawa, University of Tokyo for

Concrete' Committee. Prof. S. Morita. Kyoto

Unlvers.ity for reinforcement Committee,

Prof. 'S. Otani. University of Tokyo for

Structural Element Committee and Prof. T. Okada. University of Tokyo for Structural

Design Committee. Construction and

Manufacturing Committee will start in 1991.

These committees have cooperative relations

with various groups as shown in Fig.2.

5. ITEMS FOR RESEARCH AND DEVELOPMENT

Research items under consideration in each

technical committee are as follows:

Concrete Committee

(1) Development of materials necessary for

making high- and ultra high-strength

concrete.

Quality standards of the materials.

(2) Physical properties of high- and ultra

high-strength concrete.

(3) Mix proportion design, casting and curing

works and quality control.

Reinforcement Committee

(1) Development of high- and ultra highstrength reinforcement. Mechanical

properties of reinforcements.

(2) Mechanical properties of confined

concrete.

(3) Constitutive equation of RC elements and

application of the finite element method.

(4) Bond between concrete and reinforcement.

Anchorage and arrangement of

reinforcement.

Structural Element Committee

(1) Mechanical properties of beams and

columns.

,(2) Mechanical properties of shear walls.

(3) Effect of shear force on beams, columns

and shear walls.

(4) Mechanical properties of beam-column

connections and frames.

(5) Mechanical properties of foundations.

Structural Design Committee

(1) Methods for modeling and analysis of

behavior of structural frames

in each zone.

(2) Practicable types of structural frames.

(3) Design loads and requirements for

structural performance.

(4) Design methodology.

6.. CONCLUDING REMARKS

The scope of the research and development In

New RC Project includes research fields far

beyond frontiers of current knowledge on

reinforced concrete structures, where it is sure for us to encounter much difficulty.

In spite of that,, the- project dares to

challenge the difficulty In order to know the

future scope of reinforced concrete

technology. This project has aroused great

interest . in private companies, universities,

etc. and the cooperation and assistance

proposed by those organizations will produce

much fruit in this project.

One of the most important technology elements

in high-strength reinforced concrete building

engineering is related to quality control of

concrete. In Japanese practice in the

structural design of reinforced concrete

buildings, the accuracy of quality control of

concrete has never been reflected positively

on structural design. This project will pay

attention to the effects of the accuracy of

quality control of concrete on structural

design of high-strength reinforced concrete

buildings.

151
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154
Guidelines for Identification and Mitigation of Seismically

Hazardous Existing Federal Buildings

by

H. S. Lew1

ABSTRACT

This report, Guidelines for

Identification and Mitigation of

Seismically Hazardous Existing Federal

Buildings, was prepared by the

Interagency Committee on Seismic Safety

in Construction in support of the

National Earthquake Hazards Reduction

Program, the President's plan to

implement the Earthquake Hazards

Reduction Act of 1977 (Public Law 95-

124) . The Guidelines are intended for

consideration and use, as appropriate,

by Federal agencies in their plans for

mitigation of seismic hazards in

existing buildings. Some Federal

agencies have their mitigation plan in

operation. It is not the intent of

these Guidelines to supercede the

existing plans.

KEYWORDS: buildings; earthquakes;

earthquake hazard;

evaluation; existing

buildings; federal

agencies; guidelines;

mitigation; seismic safety;

strengthening

1. INTRODUCTION

1.1 Background

Existing buildings that were not

designed to be earthquake-resistant

constitute one of the major potential

hazards to life and property. A

building may be vulnerable due to ground

shaking caused by the earthquake , soil

liquefaction, landslides, surface

rupture, and also by tsunami. The huge

inventory of existing buildings makes

the reduction of the earthquake hazard a

complicated and costly issue that must

be worked out carefully with full

consideration of the economic

impact.

There are several factors which

make it technically and

economically difficult to identify

and upgrade many buildings

which could be hazardous in terms

of life safety or post-earthquake

operational capability:

(a) while most areas of the

United States are susceptible to some

level of seismic hazard,

many areas have not

required buildings to be

designed to resist

seismic forces;

(b) seismic design criteria

have been upgraded over

the years and it may not

be feasible to bring all

existing buildings

,

including those which

were designed according

to old codes but may not

meet current codes, into

conformance with the

provisions of current

codes; and

lNational Institute of Standards

and Technology, Gaithersburg , MD

20899

155
(c) historically, some buildings rot designed to

current codes have

performed adequately during

earthquakes, while other

buildings designed to

specific codes have not

performed satisfactorily.

Thi report provides guidelines to

r^deral agencies for use in their

programs to identify, evaluate, and

strengthen existing buildings to reduce

the threat to life safety and to reduce

major damage to critical and essential

buildings

.

1.2 Scope

The basic goal for stxerigthening of

existing buildings is to provide life

safety and, where necessary,

post-earthquake operational capability.

Buildings vulnerable to structural

damage pose the most serious hazard to

life safety. First priority must be

given to identifying structures that

might be susceptible to structural

damage and/or collapse. Nonstructural

components that could be a threat to

life safety should be identified next.

Buildings exempted from investigation

under these guidelines (Sect. 3.2.1)

should be investigated for nonstructural

hazards, but these investigations could

be incorporated into regular maintenance

inspections or even be a special

investigation for buildings housing

toxic and explosive substances.

and scheduling the seismic

evaluation are presented. The

final decision is left to the

responsible agency based on budget

constraints and overall agency

programs

.

Section 3 describes a procedure

for identification of hazardous

buildings and Section 4 describes

qualitative evaluation procedures.

Section 5 makes reconmendations

for requirements for levels of

acceptable performance for

existing buildings. Section 6

suggests mitigation techniques.

It is recognized that Federal

agencies must have flexibility in

dealing with hazardous buildings

which can have significantly

different characteristics.

Additional constraints are imposed

when a building is leased or built

under a grant program instead of

being owned by the agency.

1.3 Definitions

Critical Building - a building

which, in case of "failure" may

cause secondary effects such as release of toxic substances, fire or explosion.

Essential Building - a building

which must be safe and useable for

essential or emergency functions

during and after a major

earthquake.

This report presents a systematic

methodology for identifying hazardous

conditions and supplying decision makers

with information on the extent of the

hazard and the feasibility of mitigation. A viable strategy for mitigation

of hazards can be developed and targets

for implementation established.

Section 2 provides reconmendations

pertinent to implementing a mitigation

program for hazardous buildings.

Guidelines for prioritizing buildings

Federal Building - a building

owned, leased, assisted, or

regulated by Federal agencies

.

2. program for seismic eva1imt0n

Of existing buiidings

Implementing a program for seismic

evaluation of existing buildings

will depend upon availability of

funds and the priority of

mitigation action relative to

other requirements . It would be

156
generally infeaslble , in terms of time

and funding, for an agency to carry out

detailed engineering analyses on every

building in its inventory. Therefore,

simplified methods will be required for

identifying vulnerable buildings and

needed retrofit measures. Priorities can be set only when there is knowledge

of the degree of hazard, potential

effect on current occupants or agency's

mission, and the feasibility of

mitigation.

Following is a reconraended schedule

which is reasonable and feasible:

a. All buildings located in NEHRP*

seismic map areas 6 and 7 or UBC* seismic zones 3 and 4 and

with cccupancy of 100 or more

people should be evaluated

within five years.

b. All critical and essential

buildings located in NEHRP

seismic map areas 6 and 7 or in

UBC seismic zones 3 and 4 should

be evaluated within five years.

c. All other buildings in NEHRP

seismic map areas 6 and 7 or in

UBC seismic zones 3 and 4 should

be evaluated within eight years

.

d. All critical and essential

buildings located in NEHRP

seismic map areas 3, 4 and 5 or in UBC seismic zones 1 , 2A and

2B should be evaluated within

eight years.

e. Buildings in NEHRP seismic map

areas 3 , 4 and 5 , or in UBC

seismic zones 1, 2A and 2B

should be evaluated, on a case- by-case basis, within twelve

years.

f. All nonstructural components,

which may be a threat to life

safety, of buildings in NEHRP

seismic map areas 3 through 7 or UBC seismic zones 1 through 4

should be evaluated

within five years.

3. IDENnFICATION OF HAZARDOUS

BUILDINGS

3.1 General

The potentially large cost of an

engineering investigation of all

existing buildings in areas of

seismic risk makes it necessary to

approach the identification of

hazardous buildings in a carefully

planned manner. No single

approach would meet the need of

all agencies, because:

a. The number of buildings

in agency inventories

varies greatly.

b. The diversity of

buildings; i.e., use,

occupancy, size,

location, type and age,

complicates establishing

a single approach.

c. Availability,

completeness, and

accuracy of information

on each building vary

widely.

d. The best strategy for

prioritizing the

investigation and

mitigation efforts of

each agency would not be

uniform.

In general, buildings located in

high seismic hazard areas should

be evaluated first. The following

screening factors may be

considered in establishing an

effective evaluation program.

See references

157
3.2 Screening Factors 3.2.3 Critical Buildings

The first step in dealing with a large

inventory of existing buildings is to

apply a screening process that

eliminates unnecessary evaluation and

identifies buildings requiring further

evaluation.

3.2.1 Primary Screening Factors

Buildings, except critical and essential

buildings, which fall in any of the

following categories need not be

evaluated for earthquake vulnerability:

a. Those designs that meet or exceed the provisions of: the

NEHRP Recommended Provisions, or

the 1976 edition (or later) of

the uniform Building Code.

b. Those located in NEHRP map areas

1 and 2 or in UBC seismic zone

0.

c. One story wood-frame and one story pre-engineered metal

buildings.

d. Buildings, except essential

buildings, occupied by fewer

than 6 people.

e. One- and two-family houses which

are two stories or less.

3.2.2 Secondary Screening Factors

After the primary screening, agencies

should set priorities for qualitative

evaluation and retrofitting. As a

minimum, the priorities should consider:

Seismicity

Site conditions

Structural types

Occupancy

Building use

Some methods for prioritizing structures

are suggested in Appendix A.

a. All critical buildings

located in NEHRP map

areas 3 through 7 or in

UBC seismic zones 1 through 4 should be

evaluated. The type of

evaluation will depend

upon the nature of the

hazardous substance,

such as toxic chemicals

and explosives,

contained in the

buildings.

b. Critical buildings

located in NEHRP map

areas 1 and 2 and in UBC

seismic zone 0 need not

be evaluated.

3.2.4 Essential Buildings

a. All essential buildings

located in NEHRP map

areas 6 and 7 or in UBC

seismic zones 3 and 4

should be evaluated in

the initial phase of

program.

b. Essential buildings

located in NEHRP map

areas 3, 4, and 5 and in

UBC seismic zones 1, 2A

and 2B may be deferred

from the initial phase

of program.

In some instances, it may be more

economical and reasonable to

evaluate, at the same time, a

complete complex consisting of a

number of buildings in lieu of

individual priorities.

4. EVAIUATTON OF EXISTING

BUILDINGS

Qualitative Evaluation based on

examination of available design

documentation and field inspection

158
shall be carried out for the structural

system and for each exterior or interior

nonstructural system or component which

may pose a seismic hazard. It should be

a two step approach: preliminary

evaluation and detailed evaluation.

4.1 Preliminary Evaluation

The preliminary evaluation is intended

primarily to reduce further the large

inventories and avoid unnecessary

investigation costs. It should

determine, at the least practical cost,

whether the structure provides an

acceptable degree of safety or a more

extensive evaluation is required.

For preliminary evaluation, as well as determining the acceptable degree of

safety, ATC-21 (FEMA-154) Handbook on

"Rapid Visual Screening of Buildings for

Potential Seismic Hazard" (ATC-1988)*

may be used. The Handbook provides a

procedure for Determining if a building

needs further analysis without

performing a detailed structural

analysis

.

4.2 Detailed Evaluation

A detailed evaluation of buildings not

eliminated in the preliminary evaluation

will be required to determine the

earthquake resistant capacities of

critical elements of the building and

the extent of any deficiencies. ATC-14

"Evaluating the Seismic Resistance of

Existing Buildings" (ATC-1987)* or AIC22 "Seismic Evaluation of Existing

E^ldings" (ATC-1989)* may be used.

The ATC-14 methodology includes a stateof-the-art review of existing documents

and has incorporated information from

earlier methodologies. It is based on

assumptions that one or more of the

following events pose danger to human

lives:

o the entire building collapses o portions of the building

collapse

o components of the

building fail

and fall

o exit and entry routes

are blocked, preventing

the evacuation and

rescue of the occupants

The fundamental approach in ATC-14

is to ascertain whether there is a

complete lateral resisting system

with a coherent load path and

whether appendages and veneer are properly attached. The adequacy

of seismic performance of the

structural system and components

and exterior and interior

nonstructural systems is expressed

in terms of the Earthquake

Capacity Ratio. The methodology

is applicable to all parts of the

U.S.

ATC-22, which is based on ATC-14,

provides a step-by-step procedure

for evaluating existing buildings

.

The methodology is to identify

structural weaknesses that have

been observed in past earthquakes

to lead to failure and falling of

components or to partial or total

collapse, with an attendant loss

of life.

4.3 Earthquake Capacity Ratio

The earthquake capacity ratio can be expressed in terms of the ratio

of seismic capacity to seismic

demand of critical structural

members. The capacity to demand

ratios can be computed using the

procedure described in ATC-14,

ATC-22 or other appropriate

procedures.

The earthquake capacity ratios for

the critical elements of the

lateral force resisting system are

indices of the structural

See references

159
resistance of the existing

building. The lower the values the

higher the potential risks. If the

lowest value of the earthquake capacity

ratio is less than unity, the building

should be considered a life safety

hazard and so reported to the building

owner.

5. REQUIREMENTS FUR MITIGATION OF

HAZARDOUS BUILDINGS

Determine the level of acceptable

performance for an existing building

exposed to earthquake forces based on

the predominant performance requirement

imposed on the building, i.e. , basic

life safety of the occupants or

post-earthquake operational capability.

5.1 Basic Life Safety Requirements

The basic life safety requirements for an existing building are that during a

major earthquake it: 1) does not

collapse, partially or totally and 2)

performs adequately to provide

unobstructed ingress and egress. This

level is the minimum performance

requirement for an existing building

normally occupied by personnel and

located in NEHRP map areas 3 through 7 or in UBC seismic zones 2A, 2B, 3, and

4. This level should provide life

safety for the occupants, containment of

hazardous or lethal contents, and a safe means of egress after a major

earthquake. However, the damage

incurred by the building may not be

repairable. Recctimended actions for the

abatement of the structural deficiencies

in buildings with basic life safety

performance requirements are:

a. Where the building's

earthquake capacity ratio

as defined in Section 4.3,

is 0.80 or greater, no

action is required.

b. Where the building's

earthquake capacity ratio

is less than 0.80 but

greater than 0.50,

mitigation is

reccnmended within

10 years.

c. Where the

building's

earthquake capacity

ratio is 0.50 or

less, mitigation is

recommended within

5 years.

5.2. Post-Earthquake Operational

Capability Requirements

Post-earthquake operational

capability requirements are intended to provide continued

operation or function of the

building during and immediately

after a major earthquake. This

level is the maximum performance

requirement for an existing

building. Recctimended actions for

the abatement of the structural

deficiencies in buildings with

post-earthquake operational

requirements are:

a. Where the

building's

earthquake capacity

ratio is 0.90 or

greater, no action

is required.

b. Where the

building's

earthquake capacity

ratio is less than

0.90 but greater

than 0.50,

mitigation is

reccnmended within

10 years.

c. Where the

building's

earthquake capacity

ratio is 0.50 or

less, mitigation is

reccnmended within

5 years.

160
5.3 Cost Impact Study

Perform a cost impact study to determine

a reasonably accurate estimate of the

total costs to strengthen the building.

If the total cost exceeds projected

budgetary constraints, the continued use of the building should be limited to

storage; the building should be phased

out; or other strategies, such as lowering seismic hazard exposure by

changing occupancy requirement should be

assessed carefully.

6. MITIGATION TECHNIQUES

6.1 Strengthening Methods

Strengthening the structural and

nonstructural systems of an existing

building may be a viable strategy for

mitigating earthquake hazards. Various

strengthening methods may be applied.

The basic principles of earthquakeresistant design should be followed. It

is recotmended that the guidelines given

in the Technical Manual of the Army,

Navy, and Air Force "Seismic Design

Guidelines for upgrading Existing

Building" or "Techniques for Seismically

Rehabilitating Existing Buildings" by

URS/John A. Blume & Associates* be used.

These documents provide guidelines for

upgrading both structural and

nonstructural members. They provide the

conceptual development for seismic

upgrading and detailed techniques for

strengthening structural and

nonstructural elements with illustrated

examples. They further provide

guidelines for the cost effectiveness of

upgrading existing buildings based on

data obtained from the preliminary

evaluation and the detailed structural

analysis

.

6.2 Quality Assurance Requirements

The requirements for inspection

and material testing for new work

also apply to modification of

existing structural components or

systems. However, special

procedures are necessary to assure

the quality of alterations

involving those techniques which

are no longer used in new

construction. Therefore, the

overall adequacy of a repair or

strengthening program cannot be

guaranteed by conformance of work

to code and testing requirements

for new construction alone.

APPENDIX A

METHODS FOR PRIORITIZING

INVESTIGATIONS OF EXISTING

STRUCTURES

A number of methods for

establishing priorities have been

developed. Each agency should

determine the factors suitable to

its purposes . The following brief

descriptions of some methods are offered for information only.

a. The Naval Facilities

Engineering Conniand has

developed a computer program

to search automated data

files on structures and

prioritize them by applying

factors selected according to

size, age, replacement cost

and usage. This program can be easily expanded to include

other factors if desired.

b. "Handbook on Establishing

Priorities for Seismic

Retrofitting of Buildings"

by Federal Emergency

Management Agency.

Whenever repair and strengthening

procedures are implemented, quality

assurance should be as rigorous as that

required for new construction. * See references

161
7. REFERENCES

ACI 318, "Building Code Requirements for

Reinforced Concrete, ACI 318-83",

American Concrete Institute, Detroit,

Michigan 1983.

Army, Navy, Air Force, 1988, Seismic

Design Guidelines for Upgrading Existing

Buildings . Departments of the Army (TM809-10-2), Navy, (NAVFAC P355.2) , and

the Air Force (AFM 88-3, Chap. 13, Sect.

B), Washington, D.C.

ATC, 1987, Evaluating The Seismic

Strength of Existing Buildings . Applied

Technology Council Report ATC- 14,

Redwood City, CA.

ATC, 1988, Rapid Visual Screening of

Buildings for Potential Seismic Hazards:

A Handbook. Applied Technology Council

Report ATC-21 (FEMA-154), Applied

Technology Council, 3 Twin Dolphin

Drive, Redwood City, CA 94065.

ATC, 1989, Draft Seismic Evaluation of

Existing Buildings . Applied Technology

Council Report ATC-22, Redwood City, CA

94065.

NEHRP, 1988, NEHRP Recomrended

Provisions for the Development of

Seismic Regulations for New Buildings . Building Seismic Safety Council,

Washington, D.C.

UBC, 1988, Uniform Building Code

(Revised Every Three Years') . International Conference of Building

Officials, Whittier, CA.

URS/John A. Blume & Associates, Draft

Techniques for Seismically

Rehabilitating Existing Buildings . Federal Emergency Management Agency,

Washington, D.C. 1989.

162
i

U.S. Army Corps of Engineers Seismic Research Program

by

William E. Roper

1

ABSTRACT

The Corps Research Program is focused on addressing the seismic problems of the Civil

Works Water Resource Program. It includes

seismic challenges related to new construction

as well as operating existing structures and

repair, maintenance and rehabilitation of

existing structures. The research program is

developed cooperatively with field users, the

laboratories, and headquarters program managers.

The research program has four primary areas:

(1) liquefaction potential of fine-graded soils,

(2) seismic response of concrete dams and

abutment structures, (3) earthquake hazard

evaluations for engineering sites, and (4)

improvement of foundation soils susceptibility

to liquefaction.

KEYWORDS: Liquefaction; Seismic; Earthquake.

1. INTRODUCTION

The Corps Seismic Research Program is very

applied in character and carefully structured to meet the highest priority needs of the Civil

Works program. To ensure proper prioritization

of the research program, close coordination

between the major user and developer

organization is required. In the Corps of

Engineers, these organizations are the

headquarters Research and these organizations

are the headquarters Research and Development

Directorate, the field District offices, the

laboratories, and the headquarters Technical

Monitors (representing the functional

headquarters directorates other than R&D).

Changes this past year to effect better

coordination among the four principal

organizational elements have resulted in the

establishment of field review groups for all

research program areas. The field review groups

are composed of project engineers and managers

who are specialists in the seismic and

earthquake engineering areas from representative

Corps districts across the United States. They

are individually recommended by the headquarters

R&D Directorate and appointed by the field

Division Engineers. They participate in an annual program review of the Seismic Research

Program with the laboratory specialists and

headquarters Technical Monitors. This annual

meeting reviews the content of the proposed

program and establishes the priority of

individual work units within the overall

program. An illustration of the principal

elements involved in the program development and

the interrelationships is shown in Figure 1. This approach is proving very effective in

ensuring that field input is incorporated into

the research program. The field review group

members also act as technology transfer agents

to assure that the new technology tools are effectively utilized in the field. This has

resulted in a much closer working relationship

between the field, laboratory and headquarters

Technical Monitors.

The current Seismic Research Program has four

major components that will be summarized

separately in the remainder of the paper.

2. EARTHQUAKE HAZARD EVALUATION

This effort is a part of the Rock Research

Program that is conducted through the

Geotechnical Laboratory at the, U.S. Army

Waterways Experiment Station. The objective of

this program is to provide earthquake, geologic

and seismic logic evaluation procedures for

estimating site specific earthquake ground

motions for engineering evaluations at major

Corps projects. It is primarily focused on evaluating the durability of dams where failures

would be catastrophic. The Corps need reliable

methods to assess the sources and the maximum

potential for earthquakes. We must also develop

methods in order to judge the susceptibility of

sites to the effect of earthquakes, assign

site-specific earthquake ground motions, and be

able to generate accelerograms and response

spectra for the engineering analyses.

The current work expands the Waterways

Experiment Station collection of strong-motion

earthquake records and develops techniques for

field evaluations for estimating susceptibility

of sites. This work will also generate

parameters for earthquakes, ground motions and

develop characterizations of ground motions for

engineering analysis. These characterizations

will be used to improve computer program models

for generating synthetic seismograms.

Some specific accomplishments from this research

include seismic zoning diagrams for the eastern

United States as shown in Figures 2 and 3. The zone diagrams provide guidance to our field

organizations for seismic design considerations.

Also, charts for earthquake ground motion

related to damage and for ground motions related

to earthquake magnitude have been devel6p'ed \ using available data. A methodology for

assessing susceptibility of a foundation soil to' liquefaction by earthquake shaking has also been

developed

.

Milestones during FY 1989 include publication of

a catalog of recommended accelerograms and

development of accelerograms for hard rock

formations. This effort is scheduled to be

completed in 1991.

Assistant Director, Research and Development

U.S. Army Corps of Engineers, HQUSACE

(CERD-C), Washington, DC 20311-1000

2

E. L. Krinitzsky, Principal Investigator

Geotech Lab., Waterway Experiment Station

Vicksburg, MS 39180-0631

163
3. SEISMIC RESPONSE OF CONCRETE DAMS AND

ABUTMENT STRUCTURES

This effort is part of the Structural

Engineering Research Program at the Waterways

Experiment Station. The research is addressing

the coupled response of the structure-reservoirfoundation system to earthquake ground shaking.

The results will be used to revise the Corps

criteria for seismic analysis of concrete dams

and abutment structures. The current Corps

criteria assumes linear elastic behavior of the

structure and loads as determined by the seismic

coefficient method. Recent research indicates

that these criteria do not adequately consider

the locally expected earthquake motion, the

dynamic behavior of the structure, the

hydrodynamic response to the reservoir, the

dynamic effect of the foundation, and monolithic

linear behavior of the structures concrete.

The analysis procedure that has been developed

is applicable to seismic analysis of nonoverflow

monoliths for concrete gravity dams. The

analysis procedure includes the dynamic behavior

of the structure, hydrodynamic response of the

reservoir, dynamic behavior of the foundation,

and absorption of the reservoir bottom. A rigorous time history finite element analysis

program was developed to include behavior of the

foundation and reservoir interactions. This

analysis procedure has demonstrated that tensile

stresses significantly change form slight

variations of the input variables. Therefore,

it is important to understand the importance of

each part of the problem, such as foundation

interaction, and appropriate values to correctly

model them nominally.

A simplified procedure taking into account the

multiple parameters has been developed into a computer program titled "SDAM." SDAM

(Simplified Dam Analysis) is an interactive

computer program which performs a seismic

analysis of nonoverflow monolith of a concrete

gravity dam. The hydrodynamic loads are developed by using a simplified procedure

developed by Dr. A. K. Chopra, University of

California at Berkeley. The hydrodynamic and

gravity loads are calculated based on an input

response spectra or a given spectra acceleration .

SDAM generates a finite element grid internally

from the six dimensions shown in Figure 4. This

grid is never seen by the user so the user does

not have to be concerned with node and element

number. SDAM uses the simplified loads and

applies them to the finite element grid and then

calculates stress values at the centroid of each

element. The centroidal stresses are then

extrapolated to produce stresses on the upstream

and downstream faces of the dam as shown in

Figures 5 and 6. Since the maximum seismic

stresses occur on the surface of gravity dams,

these plots are valid for determining levels of

stress occurring in the structure.

Specific milestones for this year include

documentation of a 3-D analysis for spillway

monoliths, and publication of a technical report

on Chopra's simplified procedures for spillway

analysis

.

4. LIQUEFACTION POTENTIAL OF FINE GRAIN SOILS

The liquefaction effect is part of the soils

programSjjConducted at the Waterways Experiment

Station. The objective of this research is to

develop criteria and methods for evaluating the

liquefaction susceptibility of fine grain soils.

This will address a problem relevant to a number

of Corps dams which are located in seismicly

active areas and are underlaid by fine grain or plastic alluvial soils. Presently, the cyclic

strengths have been judged against standards

developed for clean sands. Recently, available

data on liquefaction occurrences show that

fine-grained soils may also be susceptible to

liquefaction. A better understanding of the

limits of applicability of present analysis

methods to such soils and what modifications of

these methods is needed.

The research approach will include: (1)

evaluation of field occurrences of fine-grain

soil liquefaction, (2) development of a laboratory test program to assess dynamic

strength and liquefaction susceptibility of

fine-grain soils on the basis of plasticity,

grain size distribution, and density, and (3)

analysis of test results and developed

procedures for determining in-situ liquefaction

potential in soils containing various amounts

and types of fines. The effort will include

evaluation of field occurrences of clayey soil

liquefaction worldwide as to ground motion

characteristics, in-situ test data, and

engineering properties. This will include a compilation and analysis of data on dynamic

laboratory test for materials close to the "A"

line on the Casagrande plasticity chart. The

study will also include the examination of data

bases at various geological earthquake engineering research institutes in the People's Republic

of China.

During 1989, 576 isotropically-consolidated

cyclic triaxial tests, on controlled-mixture

specimens will be conducted to evaluate

functional relationships between liquefaction

resistances and fines type and content. Hollow

Robert L. Hall, Principal Investigator

Structure Lab., Waterways Experiment Station

Vicksburg, MS 39180-0631

4J. P. Koester, Principal Investigator

Geotech Lab., Waterways Experiment Station

Vicksburg, MS 39180-0631

164
cylinder cyclic torsional stress tests will also

be conducted to assess an appropriate factor to

convert "conventional" cyclic triaxial strengths

to cyclic simple sheer strengths. This work is

being accomplished under contract with the

University of Colorado at Denver.

5. IMPROVEMENT OF FOUNDATION SOILS

SUSCEPTIBILITY TO LIQUEFACTION

This research is being conducted as part of the

Repair, Evaluation, Maintenance and Rehabilitation (REMR) research program at the Waterways

Experiment Station. This study is to determine

the feasibility and effectiveness of various

techniques for improving liquefiable foundation

conditions under existing structures and to ensure safety against failure due to earthquake

excitation. The research results will be of

direct benefit to the Corps dams where current

evaluations reveal inadequate margins of safety

against earthquake-induced liquefaction. The

research may offer cost effective alternatives

for remedial action to improve earthquake-safe

dam foundations.

Guidelines for remedial action have been

developed as a part this research effect and

published as a technical note GT-SR-1.2 in the

REMR notebook.

6. CONCLUSION

The requirements for practical, applied

engineering tools for seismic design will

continue to grow, particularly in the area of

remedial action. In developing these

engineering tools, a partnership between the

field user and laboratory researcher is

important for providing review and feedback to

the research program. This partnership also

assists in transferring the new technology into

field application.

R. H. Ledbetter, Principal Investigator

Geotech Lab., Waterways Experiment Station

Vicksburg, MS 39180-0631

165
Figure 1. Research Program Development Process

166
80°00' 75°00' 70°G0'

NORTHEASTERN STATES

SEISMIC SOURCE ZONES

LEGEND

BOUNDARY BETWEEN SEISMIC ZONES

3DE MODIFIED MERCALU INTENSITY

SCALE

100 150 Mi

80 00' 75-W

50 0 50 100 150 KM

70 00

Figure 2: Seismic Source Zones for Northeastern United States

167
85°00' 80°00' 75°0O'

35°00

30=00

SCALES

50 100 150 Ml

25°00'

SOUTHEASTERN STATES

SEISMIC SOURCE ZONES

LEGEND \BOUNDARY BETWEEN SEISMIC ZONES

MODIFIED MERCALLI INTENSITY .**

I 85°00'

50 0 SO 100 150 KM

e7-o<r tt'oa woo-

' SCALES

SO —1_

-1

100 Ml

50 0 SO IOOM

I I I I

PUERTO RICO

I I 97'0ff WOO- 85*00

80°(xr

Figure 3: Seismic Source Zones for Southeastern United States

168
Figure 4: SDAM Finite Element Grid System

169
FOLSUM, EQ2-HV

SIGMA- 1 UPSTREAM

© 10130088

<S>.0 1.0 2.0 3.0 4.0 5.0 6.0 7.0 8.0 9.0 10.0

STRESS - PSI CX102

)

Figure 5: Stresses on Upstream Face of Folsum Dam Using SDAM

170
FOLSUM, EQ2-HV

SIGMA- 1 DOWNSTREAM

© 10130088

8

oo

•

.0

27

zrCM

•

•

.0

18

IT) f

• CD

y

ID

8

OO

8

<&.0 1.0 2.0 3.0 4.0 5.0 6.0 7.0 8.0 9.0 10.0

STRESS - PSI CX102)

Figure 6: Stresses on Downstream Face of Folsum Dam using SDAM

171
The History and Status of Strong-Motion Earthquake in Japan

By

Keiichi OHTANI 1

. Hiroshi TAKAHASHI 2

SUMMARY

The strong-motion earthquake observation

in Japan has the progress history over

35 years. Japan has about 1,000

observation sits and more than 2,000

strong-motion seismographs. However,

that whole situation is not always

understood to the full . Strong-Motion

Earthquake Observation Council was

conducted the Distribution plan for the

target to push on with installation

after 10-15 years by the grasping of

that whole situation as possible as they

can. This report describes the general

view of the history and present status

on strong-motion earthquake observation

and the outline of the distribution plan

in the future

.

Keyword

Strong-motion earthquake observation,

Strong-motion Earthquake Observation

Council, Distribution plan, Observation

network

.

1. INTRODUCTION

The strong-motion earthquake observation

in Japan, which try to catch the seismic

motion during strong earthquake without

scale-over, has passed away over 35

years. At the beginning age, the

concerned organizations of strong-motion

earthquake observation were scarcely and

the whole situation was grasped fully.

By the increasing of observation sites,

manifold of observation equipment,

complexity of the concerned

organizations nowadays, it was very

difficult to understand the present

status of the network of strong-motion

earthquake observation at nowadays

.

This report describe the outline of

present status of strong-motion

earthquake observation in Japan and the

distribution plan for the target to

10-15 years in the future placed under

the making clear the state of

observation through the reconsideration

works of "distribution Plan of the

Strong-Motion Earthquake Observation" by

conducting the Strong-Motion Earthquake

Observation Council ( SMEOC ) . And also,

we discuss that progress history by

considering with the lacking of detailed

discussion of that history and adjusted

literatures .

2. THE PROGRESS HISTORY

The persons concerned in trial

construction of a seismograph obtaining

such strong-motion records held

meetings, the "Committee for the

Standard Strong Motion Accelerograph

was organized, and the actual work of

this committee was started after the

grant-in-aid for Developmental

Scientific Research from the Ministry of

Education in the fiscal year 1951 had

been obtained. This instance was the

starting point for the strong-motion

earthquake observation in Japan. Thus,

in March 1953 a prototype of a seismograph for measuring the

acceleration of strong-motion

earthquakes was completed, and it was

named "SMAC type" after the initial

letters in the name of the committee.

This accelerograph No . 1 was installed in

an underground vault of the Earthquake

Research Institute of the University of

Tokyo, and the observation was started

on June 29, 1953. On the other hand, by

the Grant-in-aid for A.rchi tectural

Technique Research from the Ministry of

Construction in the fiscal year 1954 , another type of strong-motion

seismograph was designed by the same

committee, according to specifications a little simpler than those of the SMAC

type, and this new type was accomplished

in May 1955 and named "DC type".

On October 9, 1953 the work of making

the project for observation of

strong-motion earthquakes was commenced

by the Seismological Subcommitte,

Disaster Prevention Committee, Resources

Council, Prime Minister's Office, and

was completed as "Recommendation

concerning the Project of Strong-motion

Earthquake Observation" on January 31,

19 55, which was submitted to the Prime

Minister. This recommendation demand

the realization of the following points.

1) To complete the facilities for

observation of strong-motion

earthquake in three years.

2) To establish an organization for

liaison between observers and to

entrust this organization with

agreements about observational

operations, exchange of

observational informations and data,

*~I Head , Earthquake Engineering

Laboratory, National Research Center for

Disaster Prevention, Science Technology

Agency

*2 Former Di r ec t or -Ge ner a 1 , NRCD? , STA

172
and with carrying out of various

sorts of study.

3) To take appropriate measures for

obtaining the budget which is

required for measurement of

strong-motion earthquakes and for

administration of the said liaison

organization

.

With this recommendation as a momentum,

into the budget under the jurisdiction

of the Ministry of Construction there

were included the expenses necessary for

examination the antiseismic resistance

of buildings, and 25 strong-motion

accelerographs were additionally

installed in the fiscal year 1956 and

1957. Further, as a result of

encouragements concerning establishment

of strong-motion accelerographs, given

by the Seismological Society of Japan,

the Architectural Institute of Japan and

the Japan Society of Civil Engineers,

the number of accelerographs set up in

buildings by civil expenses was

gradually increased. In 1958, an

accelerograph of SMAC type was set up in

a dam for the first time.

With the increase of the number of -' accelerographs as is mentioned above,

the records of measurements began to be

accumulated, and the necessity of

materialization of the liaison

organization mentioned above in 2) and

3) was more and more increased. For

some time after the said recommendation

the Seismological Subcommittee of the

Resources Council played the role of the

liaison organization, but with the

increases in the number of

accelerographs and records, a stronger

organization became indispensable. For

this reason, apart from the Resources

Council, the Strong-Motion Earthquake

Observation Committee was organized, and

commenced its activity from December 21,

1956. However, this committee had no

definite estimate at all, so that the

maintenance of accelerographs and

measurements were tolerably continued by

favourable cooperation from manufactures

of accelerographs and civilians and by a part of Grant-in-aids for Developmental

Scientific Research allocated to the

Earthquake Institute of the University

of Tokyo. And then, in November 1958

the Resources Council was presented "the

Report on Strong-Motion Earthquake

Observation" to the Minister of State

for Science and Technology by the

collection of state of the strong-motion

earthquake observation under the

considering the re-examine stage of the

central mechanism for this observation,

and demand the promotion of the work of

stronti-motion earthquake observation.

By the efforts of the persons concerned

in this committee, "Strong-Motion

Earthquake Records in Japan, Volume 1«" was published in March 1960, and

thereafter the style of this publication

has t been regarded as the standard of

record publication in Japan.

It happened that on June 16 , 1964 , the

Niigata Earthquake occurred and a considerable number of buildings and

bridges were damaged from destruction of

the soft ground of their foundations and

became unusable. From this fact,

demands were made for the establishment

of earthquake-proof construction

technique on the soft ground. In

addition, a strong-motion

accelerograph , which had been installed

by the chance in a building,

accomplished its function and furnished

valuable scientific data which are

considered to clarify the dynamic

properties of loose soil and the

changing process of buildings in

Niigata, and this fact reaffirmed the

necessity for increasing of

accelerographs. Motivated by these

circumstances, the recommendation

entitled "On Intensification and

expansion of Antiseismic Engineering

Study" was submitted to the Prime

Minister from the Science Council of

Japan on November 17, 19 64. This

recommendation demanded the following

measures

:

1) Increase in the number of

strong-motion seismographs for

engineering use and practical

utilization of the data.

2) Special measures , for establishment

of antiseismic engineering technique

on the soft ground.

3) Intensification and expansion of

education and study of earthquake

engineering and soil engineering at

the universities and institutes.

After this recommendation was submitted,

in the fiscal year 1965 the Strong

Earthquake Observation Center was

established in the Earthquake Research

Institute, University of Tokyo, and a part of the expenses necessary for the

administration of Strong-Motion

Earthquake Observation Committee was the

first time recognized to be defrayed at

national expense. Motivated by the

Niigata Earthquake, however, while on

the one hand the establishment of

accelerographs became again active, on the other hand an increase in the budget

for the Strong Earthquake Observation

Center was not to be expected, and

therefore the necessity of

intensification of the said liaison

organization was afresh keenly felt.

Thus, on June 21, 1967, the

173
Strong-Motion Earthquake Observation

Council was established, the expenses

for its administration being recognized

as as estimate allocated to the National

Research Center for Disaster Prevention,

Science and Technology Agency, and this

council was now able to promote the

project of strong- motion earthquake

observation in cooperation with hitherto

existing Strong-Motion Earthquake

Observation Committee. This council as

a liaison organization has been

preparing a plan of distribution of

strong-motion accelerogr aphs all over

the country and making an effort to

secure the budget corresponding to the

increase in the number of accel erogr aphs

and records.

The Strong-Motion Earthquake Observation

Council, in compliance with the demands

from every field of life, has collected

the data concerning all of observation

points where strong-motion

accelerographs are installed all over

the country and registered them in the

"Resister Book". This book contains the

data as of March 1969, and the

additional ones have been

supplemented. In 1972, the Strong-Motion

Earthquake Observation Council presented

a future plan of distribution of

strong-motion accelerographs with the

consideration of the importance in

distributing them in the vacant area of

the country so as to record the

strong-motion earthquakes which will

occur at any part of Japan, and also in

measuring the ground vibration of

various types of soil conditions during

strong-motion earthquakes . The

standards of distribution recommended in

this plan are as follows:

1) Each of the prefecture should

have 20 stations as an average

expect Tokyo, Osaka and Hokkaido,

and the stations in each prefecture

should be distributed as uniformly

as possible

.

2) In a city having more than 200,000

population, the number of

accelerographs should be increased

according to the population, because

of the importance of the prevention

of earthquake disasters in such a congested area . 3) In principle, two or more

accelerographs should be installed

at each stations

.

According to these standards the total

number of accelerographs will be 2,100.

The plan gives the priority of the

installation to the ground having

especially soft soil and to the region

where a high probability of the

occurrence of an earthquake disaster is

expected

.

At the almost same time for the study of

distribution plan as above mentioned,

the Science Council of Japan presented

the Demand entitled "On Research

Promotion for the Prevention of Seismic

Hazards of the Constructions" to the

Prime Minister on October 1971. On the

occasion of presentation of this demand,

the Society of Construction Workers, the

Architectural Institute of Japan, the

Japan Society of Mechanical Engineers,

the Japan Society of Agricultural Civil

Engineering, the Japan Society of Civil

Engineers, the Seismological Society of

Japan, UNESCO and the other societies

were presented the demand of almost same

contents to the authorities concerned.

It was strongly requested the following

aims in that demand which will be

practised the appropriate measures

without delay at the government office

concerned

:

1) Propulsion of installation of

strong-motion accelerographs and

intensity of management structures.

2) Special measures for the promotion

of antiseismic engineering research

for the general structures include

the ground.

3) Promotion of the research for the

establishment of the new design

criteria of the buildings, such as a piped type structure, off-shore

structure, plant structure, which

were not experienced the strong

earthquakes .

With the stimulation of the active

cooperation between the organizations

concerned with the strong-motion

earthquake observations, the

recommending with the installation of

accelerographs for the high rise

buildings based on the administrative

advice of the application to the Article

38 of the "Building Standard Code", the

number of the installations increased

remarkably by the private organizations.

The observation points were remarkably

concentrated to the big cities ,such as Tokyo, Osaka. And, the new problems,

which the maintenance and keeping of

accelerographs could not followed up by

the rapid increasing of the number of

accelerographs , were arise. On August

1974, the Administrative Management

Agency was done the administrative

inspection to the aseismic measures at

the big cities. Based upon this

inspection, the following recommendation

was presented to the Ministers of

related organizations.

"The actual data for the propagation

of seismic waves and the dynamic

behaviour of structures are not so enough, because the dynamic

anti-seismic design method is nev/ly

developed. To improve the

174
the following urgent measures,

1) The adjustment of the

strong-motion earthquake observation

network which can be caught the

large earthquakes occurred at any

regions

.

2) The strengthening of the

organization for the collection and

analyses of the observed data.

3) The appropriate maintenance of

accelerogr aphs and the advice of

accelerograph installation for the

public or important structures.

By the considering with the progress of

earthquake engineering and engineering

seismology, the Resources Council was

looked over again the implementing

condition for the recommendation

presented on 1955, and presented the new

recommendation entitled "On the

strong-motion earthquake observation for

the necessity to estimate the seismic

risk", which was putted stress on the

promotion of observation at the ground,

to the Minister of State for Science and

Technology. This recommendation was

requested the following measures;

1) To adjust and expand the network

at ground for the understanding of

characteristics of seismic motion

during large earthquakes at the

different soil conditions of

different areas.

2) To attempt the improvement the

recording accuracy for the grasping

of totally properties of strong

motion

.

3) To attempt the adjustment of the

collection and management systems

for the observed records

.

4) To attempt the adjustment of the

regulation and propulsion systems

intentionally and synthetic to the

totally researches and developments

for the ground motion during large

earthquakes

.

The International Workshop on

Strong-Motion Earthquake Instrument

Arrays was held under the sponsorship by

the National Science Foundation and

UNESCO-, and the auspices of the

International Association for Earthquake

Engineering in Hawaii in May 1978. The

goal of this workshop was to develop a a workable plan for the possible future

deployment of dense strong-motion

earthquake instrument arrays with

primary emphasis on ground motion

studies. The following resolutions,

should be implemented, was approved;

1) The International Association for

Earthquake Engineering in

collaboration with the International

Association of Seismology and

Physics of the Earth's Interior form

an International Strong Motion

The Ministry of Construction was

recognized the expansiveness of the

current accelerographs , the complication

of maintenance and replaying of the

records, was set about the development

of the popular edition accelerographs of

the having the definite accuracy and low

cost, and also having the easy treatment

by using the new technology focused to

the electronics on June 1979. This

development was conducted based on the

regime of the technical estimation of

construction, 5 manufactures were

subscribed to this development. As the

result of the estimation, these 5 accelerographs from 5 manufactures were

approved to reach the developmental

targets. These targets were instituted

to the following;

1 ) To be able to record

simultaneously 2 horizontal

components and 1 vertical component.

2 ) To be able to record the

acceleration in the range between 30

to 1,000 gals in the frequency range

0.3 - 15 Hz, and to prepare the

replay or digitized systems with the

interval less than 1/100 sec.

3) The accuracy of the replay or digitized acceleration is not

exceeding with (recorded

acceleration X 10 % + 10 gals).

4 ) To be able to record several

times, and maintain the operating

time of more than 2 minutes by the

one start

.

5) The starting is done by the

vertical acceleration, and it is

able to set up the starting

acceleration with the range 5-20

gals . 5) The cost should be sett

low-priced

.

By the reason of the necessity to

increase the dense-array strong-motion

earthquake observations, the

Strong-Motion Earthquake Observation

Council examined the new type of

seismograph. In October 1983, after two

years discussion, the council published

the report relating the new type of

strong-motion seismograph. The main

specifications of this seismograph has

the following;

1) To be able to record the seismic

motion of acceleration with the

range of 0.1 -2,000gals.

2) To be able to record faithfully

within the frequency domain of 0.05

- 25 Hz . 3) To be able to record the absolute

time data with the accuracy of

1/1000 sec simultaneously with the

seismic motion.

4) To be able to record from the

initial movement part by using 5 sec

delay equipment

.

175
5) To be able to record the same

timing sampling and same absolute

time by the connection with plural

seismographs

.

6) To easy use the analysis of record

from using the digital cassette

tape .

Based on the present analysis of the

some conditions of development of

strong-motion earthquake observation,

the Strong-Motion Earthquake Observation

Council was established the working

committee to study the new distribution

plan for the index of the project of

strong-motion earthquake observation for

the target of 10 - 15 years in the

future in 1983. In September 1989, this

council was published the report of

distribution plan of strong-motion

seismographs". The following sections

of this report is explained the present

status and the future plan of

installation of seismographs based on that report

.

As is mentioned above, the SMAC type

accelerograph was developed in 1953.

Based on the experiences of the use of

SMAC type in the subsequent period,

improvements upon this were put one

after another. The SMAC type has now 12

variations: A, A-2, B, B-2, C and C-2,

which are waxed paper type, and D, E,

E-2 and Q, which are scratch record film

type, and M and T, which are analogue

cassette tape type. To correspond with

the digital measurement, the Ministry of

Construction was developed new SMAC-MD

in March 1989. The main properties of

this digital instrument are as follows;

1) To be able to measure 9 channels

for the maximum.

2) To be able to increase the

capacity of the recording memory, to

be able to max. 4 Mbyte by using 1 Mbyte IC memory card.

3) To be able to read the maximum

acceleration value in a moment.

4) To be able to transmit the seismic

information

.

5) To have the compensation of 3

hours stoppage of electric current.

3 . PRESENT STATUS

As is mentioned above, the history of

strong-motion earthquake observation is

nearly concerned with the project by

using the SMAC type accelerographs

.

Nowadays, the whole condition of the

strong-motion earthquake observation is

limited to the condition of using SMAC

type accelerographs. It is the present

status that the whole situation off

using the other type seismographs, such

as moving coil type, servo magnetic type

and so on, is not arranged. Therefore,

the Strong-Motion Observation Council

was investigated the almost 70 % of the

whole condition in cooperation with the

manufactures .

Fig. 1 shows the distribution of SMAC

type accelerographs (include DC type, and so forth) registered to the

Strong-Motion Earthquake Observation

Council as of March 1989. Fig. 2 shows

the increasing trend of installed

accelerographs , which include the

non-registerd accelerographs. the total

number of accelerographs is reached to

over 1,500 instruments. Fig. 3 shows

the number of recorded earthquakes

.

More than 1,500 earthquakes were

recorded in the over 35 years till the

fiscal year 1989.

To investigate the distribution of

seismographs all over the country, the

map which was devised to the mesh from

the origin located Nihonbashi, Tokyo,

was made. We made the 2 kind of mesh

map, 25 km X 25 km mesh (25 km mesh map)

and 50 km X 50 km mesh (50 km mesh map).

We did not considered the part of small

island area. Fig. 4 and Fig. 5 show the

25 km and' 50 km mesh lines superposed on

the map of Japan, respectively. Figs. 6

and 7 show the outline of Japan by the

mesh map, respectively. 25 km mesh map

has 644 meshes, and 50 km map has 177

ones .

The installed area of SMAC type

accelerographs illustrate to the 25 km

mesh map in Fig. 8. The mesh , which has

more than one SMAC at least, is counted

to 198, the ratio is 30.6 %. Fig. 9 shows the same figure to the 50 km mesh

map. The installed meshes are 112, the

ratio is 63.3 %.

Now, we use many kind of seismographs

for the observation at the same to the

use of SMACs. Fig. 10 shows the

installed area of all kind of

seismographs. The installed meshes of

25 km map are 292, the ratio is 40.5 %.

The vacant meshes are 485. Fig. 11

shows the same manner map of 50 km mesh.

The installed meshes 146, the vacant

meshes are 31

.

4. DISTRIBUTION PLAN OF SEISMOGRAPH S

The Strcng-Motion Earthquake Observation

Council was published the distribution

plan on February 1972, as mentioned

above. This plan is recommended the

2,100 accelecographs installation all

over the country. During 25 years from

the publishing this plan, the

installation and collection of records

were smoothly underwent. Nevertheless,

based on the consideration of the more

176
increase trend of concentration of the

seismographs at the big city, the

developments of the research for the

seismic risk and the earthquake

prediction, it was important thing to

reconsider the distribution by the base

of efficiently and economically

observation. The Strong-Motion

Earthquake Observation Council was

established the "sub-committee for the

examination of new distribution plan" in

November 1983. As the result of

examination, the council was adjusted

the report on the distribution plan of

seismographs all over the country on

September 1988.

At the examination, it is aim to the

efficiently collection of data by the

correction of unbalanced installation

and the consideration of local seismic

risk. And also, that works were

enforced by the esteem of the way of

thinking of former plan and the changing

of social and economical states in

present. This new plan was made by the

visual point to the observation at

ground, which has very high capability

of utilization of records. The

observation at the structures will be

made the appropriate plan by the each

organization

.

The large or huge earthquake will, occur

at the off-shore of Pacific ocean, the

moderate or large earthquake will occur

at the Japan sea are and the moderate

and shallow earthquake will occur at the

inland area. The seismic activities

based on the historical data show the

local properties, but generally

speaking, any place in Japan has the

possibility of attacking of the

strong-motion earthquakes by the

distribution of the active fault all

over the country. For the catching of

the record during earthquake at the one

area at rare intervals, the fundamental

distribution is necessary to install the

equal density all over the country. And

then, it is reasonable to change the

distribution density at locally by the

consideration of the research result of

the seismic risk and the prediction of

earthquake occurrence . The aim of

distribution was set up to the two kind

of areas, the one is the mesh of

fundamental observation (fundamental

mesh) for the covering to similar

density to the whole Japan, the other is

the mesh of intensified observation

(intensified mesh).

The distribution distances were

determined to the following objects by

the considering with the target of

observation

.

1) To be able to record the vibration

of Seismic Intensity 5 (JMA scale)

at the earthquake of greater than

M6.5.

2) To be able to record the vibration

of Seismic Intensity 4 at the

earthquake of greater than M6.0.

The empirical equations of the relation

•between the Magnitude and the area (km ) of the. greater than some intensity were

presented by Prof. Muramatsu and Dr.

Katsumata. Prof. Muramatsu was

presented the equation between M and

area (Sv), which is the area of greater

then intensity 5, as shown Eq . 1. And,

Dr. Katsumata was presented the equation

between M and Siv (the area of greater

than intensity 4) as shown Eq . 2.

log Sv = M - 3 . 2 ( 1

)

log Siv = 0.82M - 1.0 (2)

It is assumed that the area of

generation of greater than intensity 5 or 4 vibration is the square of L (km),

L = 45 km for the M6 . 5 by the Eq . 1, and

L = 91 km for M6 . 0 by the Eq . 2 . By the

reference of this value, the size of

fundamental mesh is defined to the 50 km

X 50 km for the equality distribution of

seismographs based upon the above items

1 ) and 2 )

.

The size of intensified mesh i's defined

to the condition to be able to record of

Seismic Intensity 5 at the earthquake

greater than M6.0. From Eq . 1, L = 25

km for M6 . 0 . Then, the size of

intensified mesh is determined to 25 km

X 25 km. The area of intensified mesh

was determined the following three

items

.

(1) The area of high seismic activity

It is a nearly impossible to predict

the future earthquake occurrence

except the greater than M8.0, like

the Tokai Earthquake, because the

prediction for the smaller earthquake

is the face of research. And, some

researchers were presented the method

to estimate of the strength of

seismic motion for future earthquake

by statistic manner. But, they have

the great difference by the

difference of analytical method and

the using data. Therefore, The

relatively high seismic activity area

was determined by the proposed method

by the "Development of New

Anti-seismic Design Method"by the

Ministry of Construction (1972

1976). Fig. 12 shows the section of

seismic activities. In this figure,

A is the area of high seismic

activity, B is the moderate activity

and C is low activity. The

intensified area was determined by

this high seismic activity zone.

(2) Area of specific and intensified

177
observation, designated by the

Coordinating Committee for Earthquake

Prediction

.

The designation of area of specific

observation is given to areas of

historical disastrous earthquakes,

active faults, high seismicity, and

to areas important from social and

economic aspects, like densely

populated area. When anomalous

phenomena which could be considered

precoursors to an earthquake, such as land uplift or subsidence, are

observed by either the nationwide

network of geodetic surveys, seismic

observation and tide-gauge

observation or by other observations

in the area of specific observation,

such an area is designated as an area

of intensified observation. Fig. 13

shows the area of specific and

intensified observation , which were

revised on 1978. These areas are

appropriated to designate the

intensified mesh of the strong-motion

earthquake observation. Some parts

of these areas are duplicated to the

high activity zone. Then, the areas

of western Akita Pref . and

northwestern Yamagata Pref.,

southwestern Niigata Pref. and

Northern Nagano Pref . , Iyo-nada and

Hyuga-nada region are determined to

the intensified observation mesh of

this distribution plan.

(3) The densely populated areas

In the densely populated areas, the

number and sort of constructed

structures are so many, it is

predicted to the social effect- is

very severe at the earthquake hazard.

And, the disaster of large and

important structures has a possibility of secondary hazard

unless the hazard of themselves. By

consideration of this point, the

densely populated areas and the

construction sites of large and

important structures are designate to

the intensified area of this plan.

Based on the determination of

fundamental mesh and intensified mesh,

the mesh which should be installed

seismographs in the future and that

distribution plan were examined. The

examined ideas are as follows;

1) Dissolution of vacant mesh of

seismograph

All of vacant fundamental and

intensified mesh will be set 3 observational sites. At one site, 2 seismographs are installed at the

ground and structure. It is

necessary that more seismographs on large and long structures will be

installed. Table 1 shows the object

meshes for examination by considering the duplicated meshes

between the fundamental and

intensified observation. The map of

the object meshes for examination is

shown in Fig. 14. This figure

explain the final distribution' of

the fundamental and intensified

meshes. There are 92 fundamental

meshes, 151 intensified meshes foe

the area of high seismic activity

zone, 93 intensified meshes foe the

area of specific observation and 50

intensified meshes foe the area of

intensified observation, designated

by the Coordinating Committee for

Earthquake Prediction. And, the

total number of meshes is 386. The

distribution of installed

seismographs is shown in Fig. 15,

which was drawn on the Fig. 14. By

the Table 1, the vacant meshes are

20 (22 %) in the fundamental

observation area, and 140 (48 %) in

the intensified observation area. To

dissolute these vacant meshes

according with the above mentioned

target of installation, it is

necessary to set up 480 (= 160 X 3)

observation sites and 9 60 (= 4 80 X

2) seismographs.

2) Densely installation for the

scarcely number of seismographs mesh

It is no little number of meshes

v/hich were set up the plural

observation sites already. But, It

is important to attempt the densely

installations, according with the

above target (1) The meshes which

has not 3 observation site, are 55 % of . total fundamental observation

meshes, and 60 % of intensified

observation meshes. The installed

condition is not same with each

mesh, which has one seismograph at

least. If it is assumed that 2 observation sites, and 2 seismorraphs for each site will be

need to attemDt in the future, 80

( 72 X 0.55 X 2") sites, 160 (80 X 2)

seismographs for the fundamental

meshes, and 190 ( 1 54 X 0.6 X 2 ) sites, 380 (190 X2) seismorraphs for

the intensified meshes will be need

to install in the future.

For the achievement to the targets of

(1) and (2) together, it is necessary to

install 750 sites, 1500 seismographs in

total in the future. They are the

minimum numbers for the achievement to

the targets. Therefore, it is necessary

that more seismographs will be installed

on the structure. In addition with the

increasing of installation within the

fundamental and intensified observation

meshes in this distribution plan, the

178
following items are requested.

1) To propel the installation on the

ground at the new observation site.

2) To request the plural installation

on the ground at the different soil

conditions, if possible. And, to

strongly request the installation on

the rock layer.

3) To attempt the dissolution of the

vacant mesh at the densely populated

area, with dependent of the above

mentioned items

.

According with the publication of this

report of new distribution plan, the

actualization of this plan will be

desired by the specific cooperation of

every organizations with reflection of

this plan's point of view to the

enforcement installation planning of

the seismographs.

5 . ACKNOWLEDGMENT

The Report of Distribution Plan of

Strong-Motion Seismographs was completed

by the specific efforts of the member of

the Strong-Motion Earthquake Observation

Council, the Secretary Meeting of this

council and the Working Committee for

studying the New distribution Plan.

This report was arranged in order by the

authors, quote from that Report. The

authors would like to thank every

concerned gentlemen. And, for the

history of the strong-motion earthquake

observation, we feel uneasy to escape

some items our memory. We shall be

happy if you will point out.

6 . REFERENCES

1) "The Report of Distribution Plan of

Strong-Motion Seismographs" edited by

the Strong-Motion Earthquake

Observation Council, published by

National Research Center for Disaster

Prevention, 1988

2) I. MURAMATSU "Relation between the

Distribution of Seismic Intensity and

the Magnitude of Earthquake"

Education and Research Report of

University of Gifu, 4(3)] 1969,

pp.168 - 176

3) M. KATSUMATA, K. TOKUNAG

A

"Corresponding with the Area of

Seismic Intensity 4 and the Magnitude

of Earthquake, and the Seismic

Intensity and the Acceleration"

Shinken-jiho 36 1971, pp.89 - 96

4) S. HATTORI, Y . KITAGAWA, M. OTUKA

"Regional Coefficients of Earthquake

for Aseismic Designs" BRI Research

Paper, No. 7b, 1977

179
Table 1. The Object Meshes for Examination

Total No.

of

Meshes

Mesh No.

of

Installed Area

Mesh No.

of

Uninstalled Area

Meshes for Fundamental.

Observation (50kmX50km) * 9 2 7 2 2 0

Meshes

for

Intensified

Observation

(25kmX

25km)

I : High Seismic

Zone 1 5 1 6 0 9 1

II : Area of Specific Obs. 9 3 2 9 4 5 1 1 5 4 4 2 1 4 0

HI : Area of Intend fied Obs. 5 0 4 3 7

Total 3 8 6 2 2 6 1 6 0

* Excepted the Duplicated Meshes for Intencified Observation

O 200 km

<f

'

Fig. 1. Distribution

Number Number

Districts of of

Places

* SMACS

Hokkaido (HK) 48 82

Tohoku (TH) 62 115

Tokyo (TK) 182 488

Kanto (KT) 94 163

Aichi (AC) 29 52

Chubu (CB) 10S 183

Osaka (OS) 38 93

Kinki (KK) 47 72

Chugoku (CO 43 65

Shikoku (SK) 30 53

Kyushu (KS) 35 59

Total 712 142S

SMAC type + DC type

Strong-Motion Accelerogr aphs

(as of March 1989)

180
150

o

<

in 100

LL

o

cr

LU

CD

3 50 -

~ 1500

1000

- 500

r> o

<

CO

LL

o

crLU

CD

Q

LU

h-

<

O

(J

<

Fig. 2. The Increasing Trend of. Installed Accelerographs

150 -

CO

LU

<

ZD

O

IE

Hcr<

LU

LU 100

Q

cro

o

LU

cr

Tokochiokl Eq . M 7.9

LL

O

cr

LU

CD

50

55

Niigata Eq . M 7.5

/

Chiboken-Toho-oki

Eq . M 6.7

Nihonkoi-Chubu

Eq . M 7.7

Miyagiken-oki

Eq . M 7.4

1500

1000

500

0

60 65 70 75 80 85

YEAR

CO

LU

LL < O 3

a

cr x

LU im cr 2 < 3 LU

Q Q LU

LU Q

i- cr < o

_i cj 3 LU 2 cr 3

O

(_) <

Fig. 3. The Number of Recorded Earthquakes

181

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187
Summary of Lifeline Earthquake Engineering

James D ABSTRACT

This paper reviews the current status of lifeline

earthquake engineering design practice and major

research efforts underway which ultimately will lead to the establishment of more formal seismic design

criteria. Earthquake engineering studies have been

conducted for many years. However, it was not until

the 1971 Ss.i Fernando earthquake that major focus was placed jn the damage to structures and facilities other

than buildings. At that time, C. Martin Duke defined a new subarea of earthquake engineering he termed

"lifeline earthquake engineering." His general

definition of lifelines encompassed those facilities

which are required to support and maintain emergency

operations and to facilitate longer term reconstruction

following an earthquake. He noted that little

attention had been given to the seismic design of

transportation, gas and liquid fuel, water and sewer

and electric power and communications facilities.

Since 1971, major research has been conducted which has

lead to the identification of lifeline design

guidelines and establishment of a major lifeline

earthquake program in the United States, which is

summarized herein.

KEYWORDS; Earthquake; lifelines; research; design.

1. INTRODUCTION

Lifelines are defined as those utilities, facilities,

and structures that are required to function following

an earthquake to facilitate search and rescue, provide

emergency services, allow for the movement of goods and

materials and form a network which is required for post

event reconstruction. Electric power and

communications, gas and liquid fuel, transportation and

water sewer systems form todays modern lifeline

network. Each lifeline is comprised of numerous

components, some of which are critically vulnerable to earthquake induced damage and whose loss of function

renders the lifeline useless. For example, a break in

a major gas, oil or water transmission pipeline can cause system failure. However, a break in a gas or water service line may eliminate service to a very

local area only. In either case, failure may induce

secondary problems such as fire caused by escaping gas

from a broken pipe or the inability to control fires

becasue of a broken water pipe.

Attention was focused on the lifeline problem following

the 1971 San Fernando earthquake. Since then, numerous

post earthquake investigations have been made and

damage assessment reports prepared documenting the

performance of lifelines. Basic research is providing

engineers an understanding of how and why components of

lifelines perform the way they do. Yet little has been

published in the open literature to provide detailed

guidance for the design and construction of lifelines

to resist strong ground shaking.

The design and construction of integrated lifelines

involves the application of multidiscipl inary topics

and experience gained from previous earthquakes where

weaknesses in design, construction, system

architecture, and management have been highlighted.

The performance and reliability of lifelines can affect

a broad geopolitical area requiring the involvement of

community leaders, public officials and the private

sector to mitigate damaging effects. Pre-earthquake

by

. Cooper1

considerations include the identification of expected

variations in earthquake intensity, engineering

factors, and policies influencing risk and reliability.

The planning process for each lifeline system will

significantly influence the expected outcome of the

effects of a seismic event. The process includes an understanding of the geologic factors that produce the

seismic intensity levels that cause structural damage

and ground failure. This knowledge supports systematic

evaluation of hazards and risks required for planning

reliability needed to mitigate earthquake damage.

The easiest, most cost effective way in which to mitigate seismically induced damage is to upgrade

seismic design and construction procedures for new construction. This approach will take decades to

enhance the seismic resistance of lifelines, but will

ultimately reduce the potential for catastrophic impact

in the event of a great earthquake. Limited seismic

design procedures and details for lifelines have been

developed over the past 20 years. However, some newer structures incorporating these procedures have been

exposed to relatively strong earthquakes and have

performed quite satisfactorily. Design and

construction costs associated with enhanced seismic

resistance is still being evaluated. A range of from

one to 20 percent increased project costs for upgraded

seismic resistance has been suggested and is highly

dependent on the level of seismic detailing

incorporated into the design. Nonetheless, these costs

distributed over time become an inexpensive investment

to significantly reduce the sudden loss potential from

a great earthquake.

A second method to mitigate seismically induced damage

is to retrofit existing lifelines. Retrofit is

typically much more costl

v

than planning in advance for

the seismic design of new construction. However,

retrofit can be cost effective even in regions outside

of the traditionally thought of seismically active

zones. For example, if potentially vulnerable

transportation routes are identified, a large number of

bridges along these critical routes could be

"restrained" at low cost. Since this retrofit

technique typically costs between 10 and 20 thousand

dollars per bridge, 150 bridges could be retrofit for

the approximate cost of one new structure! For other

types of lifelines, retrofit may not be cost effective

except in unusual circumstances. Retrofit details and

designs are becoming more common in traditional

seismically active areas, although retrofit is not yet

a generally accepted policy in these areas.

Following is a summary of recent and ongoing design and

research activity that will have important impact on the seismic design and construction of lifelines.

2. LIFELINE DESIGN METHODS

The American Society of Civil Engineers established the

Technical Council on Lifeline Earthquake Engineering

(TCLEE) in 1974 to establish the means by which the

civil engineering profession could undertake a comprehensive role in elevating the state of the art of

lifeline earthquake engineering. In 1974, no major

1 Federal Highway Administration, McLean, Virginia

22101-2296

188
organization or agency was committed to the problem

area. The Society's goals and structure provided a unique opportunity to focus attention on this area. Technical committees corresponding to each of the

lifelines were established to study the problems of

planning, design, construction and operationof each

lifeline to mitigate the effects of earthquakes and

develop procedures with which sound design could be

achieved. Additionally, a committee on seismic risk

was established to develop procedures for seismic risk

analysis, including risk from geological hazards.

One of the major goals of TCLEE was to develop and

assemble advisory notes for each respective lifeline,

the ASCE publication, "Advisory Notes on Lifeline

Earthquake Engineering" (1984) presents the most

comprehensive overview of the current (1989) state of

the art of seismic design practice for each topical area

and provides some insight into how lifeline facilities

have performed in past earthquakes. The publication

provides information and approaches in developing

improved designs for new lifeline facilities.

Significant research has been conducted or funded by

the National Science Foundation (NSF), the Federal

Emergency Management Agency (FEMA) and the National

Institute of Standards and Technology (NIST) which

warrants an update of this publication.

Since 1984 some utilities, privately owned gas and oil

companies and individual agencies have developed their

own criteria for seismic design. A general approach

for the design of lifelines has evolved and is outlined

herein. Perform or conduct site specific studies which

reflect the hazard exposure to the lifeline.

Typically, geological and seismological investigations

are conducted which identify local and regional geology

and historical earthquake activity. In some cases,

field exploration studies are conducted. Criteria are developed which are based on risk considerations and

potential ground motions including accelerations,

velocities and displacements at the site, distance from

causativefault and duration of shaking. Structural

design parameters are established which include the

development of design spectra. Special structural

considerations including estimates of damping and

ductility are factored into the criteria. Allowable

deformation and force criteria are then developed.

Special detailing particularly for foundations,

anchorages and connections are then identified. These

general steps have been applied to the design of

electric power and communications facilities, gas and

oil transmission systems and transportation structures

including bridges.

3. FUTURE FEDERAL LIFELINE DESIGN GUIDES

In 1977, the Congress of the United States enacted the

Earthquake Hazard Reduction Act of 1977 to reduce risks

of life and property from future earthquakes. This act

forms the basis for the formation of the National

Earthquake Hazards Reduction Program (NEHRP). Several

specific objectives cited in the NEHRP include the

charge to develop seismic design and construction

standards for Federal use; develop guides or facilities

that are Federally owned, constructed or financed to ensure serviceability following an earthquake and

coordinate the development of guides for the

consideration of seismic risk in the development of

Federal lands.

In 1978, the Interagency Committee on Seismic Safety in

Construction (ICSSC) was established as part of the

NEHRP to assist Federal departments and agencies

involved in construction to develop earthquake hazard

reduction measures for incorporation in their ongoing

programs. The measures will be based on existing

standards when feasible.

In meeting its responsibilities, the ICSSC cooperates

with State and local governments and private

organizations in developing nationally applicable

earthquake hazard reduction measures. Five

subcommittees of the ICSSC are responsible for

responding to the charge contained in the NEHRP:

1. Standards for New and Existing Buildings

2. Lifelines

3. Evaluation of Site Hazards

4. Federal Domestic Assistance, Leasing and Regulatory

Programs

5. Post Earthquake Response Investigations

The mission of the Lifelines Subcommittee is to identify existing guidelines or standards for seismic

design, construction and retrofit of energy,

transportation, water, and telecommunication systems;

to recommend Federal adoption of such standards when

found adequate; and to encourage development of new standards where there are significant omissions. The

Subcommittee will also study techniques for evaluating

the seismic vulnerability of existing lifelines and for

improving their resistance to seismic effects and ease

of repair.

The Lifelines Subcommittee is also considering

strategies which will permit identification of those

lifeline facilities important in the emergency,

immediate recovery, and long-term economic recovery

periods, and provide guidance for appropriate levels of

seismic protection for each type.

The subcommittee is currently in the process of

reviewing technical literature pertaining to seismic

resistant design methodology of lifelines prior to

developing a formal recommendation to the ICSSC or specific guidelines for Federal agencies use.

4. LIFELINE RESEARCH

Several Federal agencies and organizations are actively

pursuing lifeline earthquake engineering research.

They include the National Science Foundation (NSF), The

National Center for Earthquake Engineering Research

(NCEER) at the State University of New York in Buffalo,

and the Federal Emergency Management Agency (FEMA).

The NSF is funding several research studies relating to siting and geotechnical systems research which focuses

on the fundamental engineering issues related to ground

shaking and the effects it has on geological -structure

interactions. Specific engineering investigations are being conducted to better understand the process of

subsurface ground damage to lifeline systems including

utility lines and energy transmission systems. Studies

include a University of Colorado investigation on the

effect of geological layered strata in modifying the

strong ground motion experienced by tunnels and buried

pipelines, including amplification of ground motion at the surface.

The 1987 Whittier Narrows, California earthquake caused

extensive damage to water distribution pipelines.

While theoretical modeling of earthquake induced damage

has been actively pursued, theoretical models are based

upon damage mechanisms that are not well understood,

and there is very little data related to the direct

observation of earthquake induced damage. Such data is

needed to validate these models so that they can be

used with confidence in deriving seismic design

criteria. At present, there is no code for buried

pipel ines.

189
An Old Dominion University study involves a thorough

field investigation and analysis of the damage to water

lifeline systems around Whittier, California. In addition, the predictions of the existing models are being compared with the observed behavior, and based on this comparison the models are being modified to correlate with the observations.

A field experiment is being performed by Weidlinger

Associates in which underground pipelines are placed

within the Parkfield segment of the San Andreas Fault

Zone. The experiment will help examine the validity of

current assumptions regarding the behavior of

underground continuously welded steel pipe and jointed

ductile iron pipe subjected to near surface fault

offset. Both active and passive instrumentation is

included to permit evaluation of important assumptions

made in current prediction and design methods.

The Whittier Narrows earthquake also offers an opportunity to assess the impact that research

conducted since the 1971 San Fernando earthquake has

had on the earthquake design and performance of

lifelines. Dames and Moore, Inc., is conducting

research including the collection, review, and

documentation of lifeline system performance data from

the October 1, 1987 Whittier Narrows earthquake; the

development of preliminary response and recovery models

for different lifeline systems using data from the

Whittier Narrows earthquake and the 1971 San Fernando

earthquake; and assessment of the impact that research

in the lifelines area conducted after the 1971

San Fernando earthquake has had on the improvement of

seismic design and emergency response and preparedness

procedures for lifelines.

The NSF also sponsors the NCEER at Buffalo, New York in

cooperation with the State of New York, and public and

private corporations. NCEER is conducting research to improve basic knowledge about earthquakes, engineering

practice and the implementation of seismic hazard

mitigation procedures to minimize loss of lives and

property. A major focused effort is underway 10

conduct research in the area of lifeline system

failure. Here, effort is directed at two lifeline

systems; namely crude oil transmission systems and

water delivery systems.

Research on water delivery systems addresses issues

relating to ground motion studies; system performance,

vulnerability and serviceability; and risk assessment

and societal impact. Research on crude oil

transmission systems involves the development of

seismic risk assessment for a crude oil pipeline system

which traverses large spatial areas. Factors will

include studies of peak ground accelerations and

permanent ground displacements. In addition to ground

motion studies which include space-time correlation and

geological, topographical and ground motion data

analyses, numerous detailed studies are being conducted

to better understand the hazards associated with

liquefaction and large ground deformation, soilstructure interaction, wave propagation and fault

crossings. Detailed studies on system vulnerability,

response and serviceability include evaluating damage

and repair of piping systems, establishing system

reliability and enhancing analysis capability.

Completion of these studies will result in a greater

fundamental understanding of how these two types of

systems respond and interact when subjected to earthquake activity. These results will provide

important information needed for the development of

detailed seismic design guides for piping systems.

In recent years, FEMA has become a major sponsor of

lifeline earthquake engineering studies. Recently, the

Building Seismic Safety Council completed the

development of a seismic hazard abatement action plan

for lifelines for FEMA to provide a basis for the

planning of a long-range lifelines research program.

The plan was developed through a workshop consensus

process utilizing the expertise of numerous professionals and organizations that deal with the

diverse areas of design, construction, operation, and

maintenance of lifeline systems. The Action Plan

recommended 67 priority projects related to lifeline

systems. Details are contained in the FEMA publication

entitled "Abatement of Seismic Hazards to Lifelines:

An Action Plan" and the six volume FEMA report entitled

"Abatement of Seismic Hazards to Lifelines:

Proceedings of a Workshop on Development of an Action

Plan."

FEMA sponsored a follow-up study to develop expert

recommendations and consensus for prioritorizing and

implementing the action. An ad hoc committee on lifelines was established for performing this task.

The committee observed that earthquakes pose a profound

threat to the reliability and continued survivability

of all lifeline systems. The committee made

recommendations to implement a nationally coordinated

and comprehensive program for mitigating the risk to

lifelines from seismic and other natural hazards and

focus project activities in the following four highpriority project areas: Improve the awareness and

education on lifeline hazards mitigation to lifeline

service providers, users, and regulators; Develop

information on vulnerability of lifeline systems to natural hazards and identify procedures for minimizing

lifelines vulnerability; and Develop and recommend

regulatory actions for adopting hazard mitigation

standards and criteria for lifeline systems.

Finally, FEMA initiated a study with EQE, Inc., to assess the potential impact of damaging earthquakes on lifeline systems in general and determine the

vulnerability of a yet to be identified single lifeline

in one region as a case model. This initial effort is

expected to contribute a considerable amount of needed

information to promote awareness on the importance of

mitigating hazards to lifeline systems.

This summarizes the major research efforts being

conducted which will add to the basic understanding of

the performance of lifeline systems. Additional

research is being conducted by private organizations

and institutes but is generally not available to the

publ ic.

5. CONCLUSION

Although significant attention has been focused on the

effects earthquakes have on lifelines, relatively

little guidance is available for the seismic resistant

design of these facilities. The most definitive

seismic design criteria available in the United States

for a specific set of structures is the American

Association of State Highway and Transportation

Officials Bridge Specifications. More general

seismic design guidance for lifelines is documented in

the ASCE publication "Advisory Notes on Lifeline

Earthquake Engineering."

Important research has and is being conducted that will

provide information on which to update seismic design

guidance for lifelines. Minimal fiscal resources could

be invested now to update design guidance for selected

lifeline structures. For example, much has been

190
learned about the dynamic effects of fluids in tanks,

piping behavior, welding and detailing and anchorage

systems to update design guidance for new tanks and

piping systems. Of course, many unanswered questions

remain which are identified in detail in the FEMA

publication "Abatement of Seismic Hazards to Lifelines:

An Action Plan." However, many new lifeline structures

will be constructed before research studies which are identified in the Action Plan can be completed. As

engineers, we must apply today's available knowledge

from lessons learned in past earthquakes to the design

of tomorrows lifeline structures.

6. REFERENCES

1. Abatement of Seismic Hazards to Lifelines: An

Action Plan, Federal Emergency Management Agency

Earthquake Hazards Reduction Series 32, FEMA,

Washington, DC, August 1987.

2. Abatement of Seismic Hazards to Lifelines:

Proceedings of a Workshop on Development of an Action Plan, Volumes 1-6, Federal Emergency

Management Agency Earthquake Hazards Reduction

Series 26-31, FEMA, Washington, DC, July 1987.

3. Advisory Notes on Lifeline Earthquake Engineering,

American Society of Civil Engineers, New York,

New York, 1983.

4. Evaluation of Site Hazards, U.S. Geological Survey

Open File Report 88-362 prepared for the

Interagency Committee on Seismic Safety in

Construction (ICSSC TR-10), Reston, Virginia, 1987.

5. National Earthquake Hazards Reduction Program:

Fiscal Year 1988 Activities, Federal Emergency

Management Agency, Washington, DC, 1988.

6. Reducing Earthquake Hazards: Lessons Learned from

Earthquakes, Publication 86-02, Earthquake

Engineering Research Institute, Berkeley,

California, 1986.

7. Second-Year Program in Research, Education and

Technology Transfer, Technical Memorndum NCEER-87-

0028, National Center for Earthquake Engineering

Research, Suny, Buffalo, New York, March 1988.

191
A Strong Motion Data Catalog for

Personal Computers

by

Lee Wesley Row, III

1 and Herbert Meyers2

ABSTRACT

Technical advancements in the field of

earthquake engineering can be impeded by

lack of essential strong motion data.

Catalogs which document these records

are valuable tools for engineers and

researchers needing to locate data

applicable to specific studies. The

personal computer-based strong motion

catalog, called SMCAT, combines the

efficiency of automated data management

systems with the convenience of personal

computers to provide users with an

alternative information source that can

be accessed interactively. The 12,700

record database which documents the

largest public domain holding of strong

motion data includes information on

triggering events, recording station

characteristics and peak ground motion

values. This paper describes the

development and features of this

catalog.

1. INTRODUCTION

Many studies regarding the effects and

potential impacts of strong ground

motions rely on the analysis of historic

accelerograms. The need for convenient

access to existing strong motion data

has been widely documented. The U.S.

National Workshop on Strong Motion

Instrumentation notes that research

advancements in the field of earthquake

engineering can, and may often, be

impeded by insufficient data (ref . 4)

.

With the thousands of processed

accelerograms already in existence, it

can be a major undertaking for engineers

and researchers to locate all of the

data applicable to a particular study.

Strong motion catalogs which document

this vital data resource have proven to

be extremely useful for investigators

needing to locate data pertinent to

their application. Although these

catalogs have appeared in a variety of

formats, such as printed listings,

on-line information banks, and graphic

work stations (see refs. 1, 2 and 3)

,

they all tend to have their drawbacks as

far as either being inconvenient,

limited in scope or not up-to-date. An

alternative solution to documenting this

resource is a personal computer-based

catalog system known as SMCAT. SMCAT,

an abbreviation for Strong Motion

Catalog, utilizes the many advantages of

personal computers to provide a

convenient database which contains ample

information.

The catalog documents the National

Geophysical Data Center (NGDC) and World

Data Center-A (WDC-A) Strong-Motion Data

Archive — the largest public domain

holding of digitized strong motion data.

The catalog is accessed via a

menu-driven, front-end program which

allows for interactive searches and

retrievals from the database. Special

characteristics of the catalog provide

for a multitude of features such as

simple updates and linkages with

existing databases. A combination of

the efficiency of automated database

management systems with the convenience

of personal computers attributes to the

practicality and effectiveness of this

catalog.

The SMCAT catalog is a dBase III+

formatted database which documents the

contents of the NGDC/WDC-A Strong Motion

Archive. The initial release of the

catalog contains over 12,700 entries or

records. Each record in the database is

comprised of 29 fields of information

which describe the triggering event

parameters, recording station

characteristics, and peak recorded

ground motion values for every processed

accelerogram in the WDC-A archive (table

1) . Most of the information fields used

in the catalog are those deemed most

useful for engineers and researchers by

the U.S. National Workshop on Strong

Motion Instrumentation (ref. 4).

Additional useful data such as recording

Cooperative Institute for Research in

the Environmental Sciences (CIRES)

,

University of Colorado/NOAA, Boulder,

Colorado 80309

2 National Geophysical Data Center, NOAA,

Boulder, Colorado 80303

192
site geology, fault plane parameters and

structural characteristics were left out

of the present catalog because of the

difficulty involved in obtaining much of

this information.

The scope of the catalog encompasses all

of the public domain strong motion data

dating from 1933 (the first U.S.

accelerograph recording) to the present

that were contributed to WDC-A from

numerous government agencies,

educational institutions, and industries

located around the world. Although

almost half of the records originated

from accelerographs installed in the

United States, there are also a

significant amount that originated from

fifteen other countries (fig. 1)

•

Records in the archive are

representative of a broad range of

recording environments from structural

to free field (fig. 2). Structural

records include those that were obtained

from a diverse selection of

dams, bridges, and utilities, as well as

various types of buildings. Recording

site geologies cover almost the entire

spectrum from hard rock to alluvium and

fill.

2 . DATA COLLECTION

Information used in the catalog was

collected from three main sources. Most

of the data pertaining to the triggering

event and station characteristics are

extracted from reports and technical

papers published by agencies responsible

for the collection and processing of

accelerograms. Unfortunately, most of

these reports are published rather

promptly and, hence, do not contain

comprehensive information.

Seismological catalogs were used as a

supplement to fill in any information

regarding the triggering event.

Finally, any information that still

remained unknown, such as peak ground

motion values, was obtained from the

archived accelerogram data files.

Most of the collected information was

entered into the catalog database

manually. However, automated procedures

can be used whenever large sets of data,

such as the CALTECH series or the Taiwan

SMART- 1 array data, are available in

redundant file formats. In this case, a short FORTRAN or C program was guickly

produced to extract as much information

as possible from the file headings and

data points.

In theory, the process of incorporating

information into the SMCAT database

seems rather straightforward; in

reality, however, the process is

typically complex. The difficulty in

developing this type of catalog can be

attributed to the hodgepodge of data and

documentation formats. Ideally, all of

the information should be contained in

the heading of the accelerogram data

file; this would not only aid in the

development of catalogs, but would

undoubtedly benefit the users of these

data. Compilation of this catalog would

become a trivial task if all of the

necessary information were written into

the files in an internationally

standardized format.

3 . INFORMATION RETRIEVAL

The catalog is disseminated to users in

a dBase III+ format on 5 1/4 inch floppy

diskettes. A front-end program is

included for interactive searches and

retrievals. This menu-driven program

acts as an interface between the user

and the catalog enabling the user to

access the database without dBase III+

software. A search is initiated by

defining the search parameters and

specifying the range of each parameter.

For instance, a typical search may

include locating all of the records

obtained from free field sites located

at a maximum distance of 20 kilometers

from the epicenter and that recorded a

peak acceleration greater than 20% g.

Up to five fields can be selected to

specify a search condition (table 2)

.

The data output can be directed to a printer, monitor, or stored in an output

file. Stored output can be re-searched

with a new set of parameters if

necessary. Three types of output file

formats are available: dBase III+,

ASCII, and Lotus 1-2-3. This option

greatly increases the flexibility of the

product by not restricting the user to a

dBase III+ format.

4 . ADDITIONAL FEATURES

The use of dBase III+ as a database

management system technology greatly

increases the capabilities of this

catalog. Several features inherent in

dBase and other features that have been

designed into the catalog program

provide an adequate foundation for users

to expand upon. New fields can be added

to the catalog without significantly

affecting the performance of the search

program. Suitable candidates for

additional fields could pertain to site

geology, structural characteristics, or

fault plane parameters. Any additional

fields that are not part of the original

database structure cannot be accessed by

the program. For users that prefer

193
management software other than dBase,

the SMCAT program will export the entire

database, or a selected portion of the

database, to either an ASCII or Lotus

1-2-3 formatted file.

If modification of the existing database

is not desirable, then the catalog can

be linked to an external database

through a related field such as event

date. Use of the program can be avoided

altogether by exporting the database to

the desired file format and using the

appropriate software to access the

records. Editing or altering of the

catalog can easily be accomplished by

using one or more of the editing

utilities available in dBase.

5. CONCLUSIONS

The SMCAT strong motion catalog is

designed to assist engineers and

researchers in locating accelerograms

suitable for particular applications.

With the diverse and extensive selection

of historic recordings that are

documented in this catalog, a user can

find records applicable to. most studies.

Special features contribute to the

flexibility of the catalog which enable

users to tailor or apply the system to

suit specific needs.

6. REFERENCES

1. Coffman, J.L. and Godeaux, S. 1985.

"Catalog of Strong-Motion

Accelerograph Records," World Data

Center-A for Solid Earth

Geophysics Report SE-38, 79 p.

2. Polhemus, N.W. and Jones E. 1984.

"Interactive Database for

Graphic Retrieval of Strong-Motion

Accelerograms," in Proceedings

of the Eighth World Conference on

Earthquake Engineering, July

1984, San Francisco, California, v.

2, p. 87-91.

3. Converse, A., 1978. "Strong-Motion

Information Retrieval System

User's Manual," U.S. Geological

Survey Open File Report 79-289,

51 p.

4. Iwan, W.D. 1981. "U.S. Strong-Motion

Earthquake Instrumentation,

Proceedings of the U.S. National

Workshop on Strong-Motion

Earthquake Instrumentation," April

1981, Santa Barbara,

California, 55 p.
1

i Fields that Comprise SMCAT

Type of Character

Field Name Description Field Width

EyDAl JCi LVent Date "KTi i yn /~\ "v* "i /*

JN UlUcilC D

EQTIME Event Time Numeric oo

T?T)TT EF1 ATI JLA1 Epicentrai Latiuuae iMumeric cD

E_NS Latitude Nortn/soutn r lag tnaracter 1

EF1JjUNCj bpicentrai Eongituae Numeric "7 E EW JjOngiLUQe iLaSt/r/est flay 1

EQNAME Event Name tnaracter J U

DEFlri rocai ueptn rjumeric

INT Maximum Intensity (MM) Numeric 2

TUT *^ mm ^ ^» i ^ / TUT \ Magnituae \^t./ Numeric A4

LUUH otation toae tnaracter -7/ ^UiTlF'JJNrjIM 1 T>^ 4" •v*! i m ^ aymv*\ /-\ >^ £N 4» instruinenu ^omponcnu 1 O T A rp T/*MvT b 1 Al 1UJM station .Location Lnaracter J U

Oil A rpi? btate (recording site; L.naracter Z

COUNTRY Country (recording site) Numeric 5

STRUCTURE Structure Type Character 3

EP1DIST Epicentral Distance Numeric 3

MAXACCEL Maximum Acceleration Numeric 10

MAXVEL Maximum Velocity Numeric 8

MAXDISP Maximum Displacement Numeric 7

INSTRUMENT Instrument Type Character 7

FLOOR Floor Number (recording Site) Numeric 2

STALAT Station Latitude Numeric 6

S NS Latitude North/South Flag Character 1

STALONG Station Longitude Numeric 7

S EW Longitude East/West Flag Character 1

TAPE NGDC Data Set Name Character 7

TYPE Data or Record Type Character 1

FILE Data Set File Number Numeric 4

Total Characters: , 190

i

Table 1. Table lists the 29 fields that make up the SMCAT database.

195
Searchable Fields in the SMCAT Database

Field Units Range

Event Date

Epicentral Coordinates

Event Name

Focal Depth

Maximum Intensity

Magnitude

Station Name

Country (recording site)

State (recording site)

Structure Type (recording site)

Building Floor Number (rec. site)

Epicentral Distance

Maximum Acceleration

Maximum Velocity

Maximum Displacement

Station Coordinates

NGDC Data Set Name

UTC

degrees

kilometers

MM

MT

floor #

kilometers

gals

cm/sec

centimeters

degrees

1933 - present

global

0 to 9999

1 to 12

1 to 10

-9 to 99

0 to 9999

0 to 99999

0 to 9999

0 to 999

global

Table 2. Table lists the fields that can be searched by the SMCAT program.

Up to five fields can be searched at one time with the program.

196
QUANTITY OF PROCESSED STRONG NOTION RECORDS ORIGINATING FROM VARIOUS COUNTRIES

IND RUN AUS F IJ CAN N IC SAL PER NZD CH I SSR PAP HEX JAP I TA PRC TA I USA

Country

HDC-A/NGDC DATA HOLDINGS

Figure 1. Chart shows the quantity of processed strong motion records

originating from various countries represented in the SMCAT catalog. One

record is defined here as a single component record with only one phase of

processing. A record with all three phases of processing (raw, filtered,

and response spectra) is considered three separate records. Country codes

are as follows: IND: India, RUM: Rumania, AUS: Australia, FIJ: Fiji, CAN:

Canada, NIC: Nicaragua, SAL: El Salvador, PER: Peru, NZD: New Zealand,

CHI: Chile, SSR: Union of Soviet Socialist Republics, PAP: Papua New Guinea,

MEX: Mexico, JAP: Japan, ITA: Italy, PRC: Peoples Republic of China, TAI:

Taiwan, USA: United States. Less than ten records originated from each of

the countries of India, Rumania, and Australia.

197
QUANTITY OF PROCESSED STRONG MOTION RECORDS RETRIEVED FROM VARIOUS STRUCTURES

(6.0)! NUMBER

v.v.v.;.;!;/.;!

X'Xv'.vXv;

BRIDGES BUILDINGS DAMS FREE FIELD UTILITIES OTHER

Structure

UDC-A/NGDC DATA HOLDINGS

Figure 2. Chart shows the quantity of processed strong motion records retrieved

from the various structures represented in the SMCAT catalog. One record is

defined here as a single component record with only one phase of processing. A

record with all three phases of processing (raw, filtered, and response spectra)

is considered as three separate records.

198
Nonlinear Seismic Analysis of

Concrete Gravity Dams

by

Luci an G. Guthri e

ABSTRACT

This paper describes an investigation of the

behavior of concrete gravity dams subjected to

earthquake motions of sufficient strength to

induce cracking. Nonlinear, dynamic finite

element analyses of three dams subjected to the

Parkfield earthquake motion are conducted with

the ADINA 84 code. In the analyses the

essential characteristics of static preloading,

bidirectional seismic motion, dynamic concrete

cracking, and hydrodynamic interaction are modeled. The results show that cracked zones

can propagate through the cross sections at

various elevations. The U.S. Army Corps of

Engineers' guidance for seismic evaluation of

gravity dams compares conservatively to these

results but incorrectly locates the elevation of

cracking through the section in some cases. A procedure to estimate the relative permanent

displacement across such cracked sections is

developed using the sliding block analysis.

KEYWORDS: Concrete gravity dams; earthquake

analysis; finite element analysis.

1. INTRODUCTION

U.S. Army Corps of Engineers publication ETL

1110-2-303 (U.S. Army CE, 1985) provides design

guidance for determining the earthquake-induced

loading, performing the dynamic stress analysis,

and evaluating the results for concrete gravity

dams. That guidance requires an analysis to

estimate the extent of cracking, and a sliding

stability analysis on planes where the greatest

cracking occurs. Cracking is indicated when a linear elastic analysis calculates combined

maximum tensile stresses greater than 0.1 times

the unconfined compressive strength of the

concrete using damping ratios of 7 or 10

percent. The sequence of analysis is shown in

Figure 1.

Where cracking is indicated an estimate of the

location and extent of cracking and an evaluation of the severity of this cracking are required. To accomplish this a dynamic

nonlinear analysis of the dam is required to

determine the location and extent of cracking.

This is followed by a sliding stability analysis

of any section above a plane determined to be

cracked

.

To confirm or appropriately modify the guidance

given in ETL 1110-2-303, as described in Figure

1, an analytical investigation was conducted on

the behavior of concrete gravity dams subjected

to earthquake motion of sufficient strength to

induce cracking (Mlakar, 1987). This paper

describes that investigation and reports on the

results

.

2. DYNAMIC NONLINEAR ANALYSIS INVESTIGATION

2. 1 Tasks of Investigation

The investigation encompassed the following

tasks

:

1) Performed nonlinear analyses on three given

nonoverflow gravity dam cross sections of 185,

300, and 638 feet (56, 91, and 194 meters) in

height using earthquake-induced ground motion of

sufficient strength to induce cracking. The

results of the analyses were used to estimate

the extent of cracking possible and any

permanent displacements likely to occur along

planes of maximum cracking.

2) Evaluated the interim procedure contained in

ETL 1110-2-303 for analyzing the response of a gravity dam to a maximum credible earthquake.

3) Developed a simplified method of analysis to

estimate upper bounds for permanent

displacements due to vibratory ground motion

.without resorting to multiple acceleration

time-history analyses.

2.2 Particulars of the Nonlinear Analyses

2.2.1 Computer Code and Discretization

The nonlinear analyses were performed using the

general purpose finite element code ADINA (ADINA

Engineering, 1984). The cross section of each

structure was modeled using 9-node,

isoparametric, quadrilateral, plane stress

elements of unit thickness. The discretization

of the finite element meshes is shown in Figure

2. This degree of discretization was found to

adequately reproduce the stress distributions

within gravity dams (Cole and Cheek, 1986).

U.S. Army Corps of Engineers, HQUSACE

(CEEC-ED), Washington, DC 20314-1000

199
2.2.2 Constitutive Modeling

The constitutive behavior of the concrete of the

dams was described with the ADINA concrete

material model (Bathe and Ramaswamy, 1979). For

the static and the seismic loadings imposed,

this description provided an essentially linear

behavior in compression and a linear behavior in

tension up to the stress level at which cracking

occurred. When this level of stress was reached, spatial zones of cracked material were

defined. In these zones the tensile stiffness

across the cracked surface was reduced to zero

and the shear stiffness was reduced to half the

uncracked value. The effects of strain rate on material behavior were approximated as described

in Mlakar, 1987.

2.2.3 Loadings

In the nonlinear analysis, both static and

seismic loadings were considered. The

self-weight of the cross section was applied as

a mass proportional body force. The hydrostatic

loading of the reservoir was accomplished

through a set of pressure loadings along the

upstream surface of the structure. Both of

these loadings were linearly increased from zero initially to their full static value in a rise

time of 2 seconds. This rise time was selected

to achieve equilibrium under the static loading

prior to the arrival of the strong earthquake

motion. The horizontal and the vertical

earthquake loading were applied as uniformly

prescribed displacements along the base of the

dams. Finally, the hydrodynamic loading of the

reservoir was approximated by adding

concentrated nodal masses on the upstream face

corresponding to the distribution used in the

linear elastic analysis as described in Cole and

Cheek, 1986.

The earthquake record (Figure 3) used in the

analyses was the N65W horizontal and vertical

components of the 1966 Parkfield, California

earthquake recorded at Temblor No. 2 Station,

CIT File Nos. B037-1 and B037-3. This loading

is representative of strong earthquake motions

likely to be encountered at rock sites, upon

which all Corps of Engineers concrete dams are founded

.

2.2.4 Discussion of Results

When excited by an earthquake motion strong

enough to initiate cracking, the dams sections

analyzed cracked completely through their cross

sections. This is consistent with limited scale

model tests (Niwa and Clough, 1980). In some cases the cracked zones grew through stable

stages. In other cases, the cracking was virtually instantaneous once it began. The

practical implication of this may be that dams

which crack during an earthquake do crack

completely through their cross section.

However, further analytical and experimental

study is required to support this conclusion in

general

.

The three dams analyzed cracked completely at

various elevations. The shortest structure

cracked through its base while the taller ones cracked through the upper elevations. The

elevation of complete cracking may be sensitive

to the flexibility of the foundation which was

not considered in this nonlinear analysis.

However, a state-of-the-art linear analysis

which included this feature (Fenves and Chopra,

1984) indicates that the maximum total stress

occurs at the upstream edge of the base. This

is the same location at which cracking initiated

in each of the nonlinear analyses conducted in

this investigation.

The Procedure of ETL 1110-2-303 conservatively

evaluated the safety of each of the four

dam-earthquake combinations considered in this

work. But in two cases, the simple linear

procedure was misleading in that the elevation

of cracking through the cross section differed

from that observed in the nonlinear analysis.

It is important that this elevation be correctly

predicted so that a subsequent estimate of

permanent displacement along the cracked plane

is meaningful. This prediction requires that

the inelastic behavior following initial

cracking be incorporated in the evaluation

procedure. This might be simply achieved with a static nonlinear resistance function and an inelastic response spectrum.

3. STABILITY OF CRACKED SECTIONS

It has been shown that extremely strong ground

shaking can cause cracking through the cross section of a concrete gravity dam. In fact, this

may have occurred as shown in Figure 4, at the

Koyna Dam during the December 1967 earthquake

(Chopra and Chakrabarti, 1971). The Corps'

analytical procedure for the maximum credible

earthquake (Figure 1) would require an estimate

of permanent displacement in such a case. A

simple procedure for this estimation is

developed and illustrated in^Mlakar, 1987.

3. 1 Description of Sliding Analysis

In the proposed procedure the cracked surface is

assumed to be planar and either horizontal or inclined (Figure 5). The portion of the

monolith above this crack is assumed to be a rigid body subject to sliding along the cracked

plane as the portion of the monolith below the

crack moves horizontally due to the ground

motion. This idealization resembles the sliding

block analysis widely employed for earth

embankments (Newmark, 1965), (Franklin and

Chang, 1977), (Hynes-Griffin and Franklin, 1984)

and should be no less applicable to gravity

dams

.

The limiting acceleration is calculated from

equilibrium of the monolith above the crack and

the amplification of ground motion is computed

from an approximation consistent with the

simplified stress analysis of gravity dams as described in Cole and Cheek, 1986.

200
3.2 Discussion of Results 5. REFERENCES

The procedure leads to a conservative result in

each of the two examples illustrated in Mlakar,

1987. The method may be improved through an experimental examination of the frictional

coefficient employed to compute the limiting'

acceleration. An analytical study of the

amplitude and frequency modulation of ground

motion within the cracked structure may also

lead to a better procedure.

3.3 Overturning Stability

Corps of Engineers criteria (U.S. Army CE, 1985)

states that an overturnig mode of failure of a gravity dam due to an earthquake-induced

excitation is not likely because of the large

base width to height ratios of gravity dam cross sections and the nature of the excitation—peak

oscillatory motion of very short duration.

4. CONCLUSIONS

The funding limitation for the Mlakar, 1987

study permitted the investigation of a rather

limited number of cases. In addition,

foundation interaction effects were not taken

into account, only one earthquake record was

used, and the hydrodynamic loading of the

reservoir was approximated by adding

concentrated nodal masses on the upstream face

corresponding to the distribution used for the

linear elastic analysis. So, broadly applicable

conclusions are not supported by this study

alone. However, the following observations can

be made from the results of the study:

1 ) When excited by an earthquake motion of

sufficient strength to initiate cracking, the

three dam cross sections analyzed cracked

completely through. In some cases, this

cracking progressed through stable stages and in

others, it was virtually instantaneous.

2) The cracking through the dam cross sections

occurred at the base, at the elevation of

downstream slope change, and at a lower

elevation. The shortest structure cracked

through its base while the taller ones cracked

through the upper elevations.

3) The procedure in ETL 1110-2-303

conservatively evaluated the safety of the four cases considered. However, in two cases the

simplified dynamic analysis procedure

incorrectly located the elevation at which

cracking occurs through the cross section.

4) The sliding block analysis is a rational

bases to estimate the permanent relative

displacements along cracked planes through a dam

cross section.

ADINA Engineering, 1984, "Automatic Dynamic

Incremental Nonlinear Analysis Users Manual,"

Report AE 84-1, ADINA Engineering, Watertown,

MA.

Bathe, K. J., and Ramaswamy, S., 1979, "On

Three-Dimensional Nonlinear Analysis of Concrete

Structures," Journal of Nuclear Engineering and

Design , Vol 52, pp 385-409.

Chopra, Anil K. , and Chakrabarti, P., 1971, "The

Koyna Earthquake of December 11, 1967 and the

Performance of Koyna Dam," Report UCB/EERC 71-1,

Earthquake Engineering Research Center,

University of California, Berkeley, CA.

Cole, R. A., and Cheek, J. B., 1986, "Seismic

Analysis of Gravity Dams," Technical Report

SL-86-44, US Army Engineer Waterways Experiment

Station, CE, Vicksburg, MS.

Franklin A. G., and Chang, F. K. , 1977,

"Earthquake Resistance of Earth and Rock-Fill

Dams: Permanent Displacement of Earth

Embankments by Newmark Sliding Block Analysis,"

Miscellaneous Paper S-77-17, Report 5, US Army

Engineer Waterways Experiment Station, CE,

Vicksburg, MS.

Hynes-Griffin, M. E., and Franklin, A. G., 1984,

"Rationalizing the Seismic Coefficient Method,"

Miscellaneous Paper GL-84-13, US Army Engineer

Waterways Experiment Station, CE, Vicksburg, MS.

Mlakar, Paul F. , 1987, "Nonlinear Response of

Concrete Gravity Dams to Strong

Earthquake-Induced Ground Motion," Technical

Report SL 87-7, US Army Engineer Waterways

Experiment Station, CE, Vicksburg, MS.

Newmark, N. M., 1965, "Effect of Earthquakes on Dams and Embankments," Geotechnique , Vol 15, No.

2, pp 139-160.

Niwa, A., and Clough, R. W., 1980, "Shaking

Table Research on Concrete Dam Models," Report

UCB/EERC 80-05, Earthquake Engineering Research

Center, University of California, Berkeley, CA.

U.S. Army Corps of Engineers, 1985, "Earthquake

analysis and Design of Concrete Gravity Dams,"

Engineer Technical Letter 1110-2-303,

Washington, DC.

201
No

No Stop.

Design Is Satisfactory.

Yes

Repeat Analysis

with 7% Damping.

Repeat Analysis

with 10% Damping.

Stop.

Design Is Satisfactory.

Estimate Extent of

Cracking and Perform a

Static Sliding Stability Analysis

on Planes Where Greatest

Cracking Is Indicated.

Figure la

Figure 1 . Sequence of analysis for the maximum credible

earthquake (DA 1985) (Sheet 1 of 2)

202
Compute Permanent

Displacement During the

Earthquake.

Stop.

Design Is Satisfactory.

Figure lb

Figure 1. (Sheet 2 of 2)

203
2500 r2000 -

1500

1000 -

500

500 1000

Y

1500 2000

Figure 2. Finite element grid of Russell Dam.

204
00

a

aH

4Jau

41

l-t41y

3

a. Horizontal

oo

a

o

(0h

41

41U

U

<3

12 16 20

Time sec

b. Vertical

24 28 32

Figure 3. Parkfield earthquake motions used in analyses

205
206
w

b. IMPENDING SLIDING UPSTREAM

Figure 5. Free body diagram of cracked dam.

207
Seismic Microzonation

By

Walter W. Hays

U.S. Geological Survey

Reston, Virginia 22092

ABSTRACT

This paper reviews the basic principles of

seismic microzonation—the multidisciplinary

process that leads to the division of a region into smaller areas expected to

experience the same relative severity of an earthquakes physical effect such ground

shaking. Since the original pioneering

research by Japanese scientists and engineers

in the early 1900 's, many significant

advances in understanding have been made.

The time seems to be right for renewed

efforts in seismic microzonation to be made

worldwide.

1. WHAT IS TO BE DONE?

1 . 1 Design Goal

The goal is to devise a set of standard

procedures that can be used throughout the

world to produce seismic microzonation

products. Products produced in a seismic

microzoning study are applied in land-use,

building codes, design and construction

practices, repair and strengthening of

existing buildings, and response and recovery

planning. These applications can save lives

and economic resources.

Although the pre -earthquake environment is more optimal, experience shows that seismic

microzonation studies are accepted more readily as a strategy for reducing losses

from earthquake hazards in the postearthquake environment. Political

considerations are usually less of an impediment in the post-earthquake

environment.

1 .2 Background

Seismic microzonation, the division of a region into smaller areas expected to

experience the same relative severity of an earthquake hazard (for example, ground

shaking, surface fault rupture, earthquakeinduced ground failure, tectonic deformation,

or tsunami runup) is an important part of the

process of evaluating earthquake hazards and

assessing the risk in an urban area. Seismic

microzonation is the part of the process of

evaluating earthquake hazards that provides

the prospective user of an area with the

design criteria that will permit him to

select the most suitable part of the area for

the proposed use.

2. HOW WILL IT BE DONE?

2.1 Compilation of Seismic Microzonation

Experience

Seismic microzonation has been performed in

many countries throughout the world.

However, there is no standard procedure for

seismic microzonation, and the results have

varied widely from country to country (see

proceedings of the three international

conferences on seismic microzonation held in

the United States and the proceedings of the

microzonation conference held in Algeria).

2.2 Technical Procedure

The key to the creation of a standard

procedure in seismic microzonation is to

obtain explicit answers to the following

questions:

Where are the earthquakes occurring now?

Where did they occur in the past?

Why are they occurring?

How often do earthquakes of a certain

size (magnitude) occur?

How big (severe) have the physical

effects been in the past?

How big can they be in the future (e.g.,

next 50 years)?

How do the physical effects vary

spatially and temporally?

How have these physical effects impacted

various types of buildings and lifeline

systems?

Although these questions appear to be simple,

the answers typically require detailed

research and technical studies that integrate

geologic, seismological , and engineering data

on two scales:

1 ) Evaluation of seismic hazards on a regional scale: (a map scale of about

1:100,000 to 1:1,250,000). This part of

a microzoning study establishes the

physical parameters of the region needed

to evaluate the earthquake hazards of

ground shaking, surface fault rupture,

tectonic deformation, and tsunami

runup. Technical tasks such as the

following are required:

a. Compilation of a catalog and map of

the prehistorical , historical, and

current seismicity.

208
b. Performance of neotectonio studies

(mapping, age dating, and trenching)

to acquire information on recurrence

times in the past several thousand

years not provided by historical

seismicity

.

c. Preparation of a seismotectonic map

showing the location of active

faults and their correlation with

seismicity

.

d. Preparation of a map showing

seismogenic zones and giving the

magnitude of a maximum earthquake

and the frequency of occurrence for

each zone.

e. Specification of regional seismic

wave attenuation laws and their

uncertainty.

f. Preparation of probabilistic groundshaking hazard maps in terms of peak

bedrock acceleration, peak bedrock

velocity, exposure times, and

probabilities of nonexceedance

.

Evaluation of seismic hazards on an urban scale: (a map scale of about

1:5,000 to 1:25,000). This part of a microzonation study integrates the

seismotectonic and other physical data

acquired in the region of the study

(Part 1 above) with site-specific data

acquired in the urban area to produce

seismic microzonation maps. Technical

tasks such as the following are required

:

a. Acquisition, synthesis, and

integration of existing and new

geologic, geophysical, and

geotechnical data to characterize

the soil/rock columns in terms of

their physical properties and their

response to various levels of ground

shaking.

b. Preparation of ground -shaking hazard

maps showing the dynamic

amplification factors for soil/rock

columns in terms of amplitude and

frequency composition of ground

shaking and the level of dynamic

shear strain for a range of seismic

loads.

c. Preparation of a map showing the

potential for surface fault rupture

and tectonic deformation.

d. Preparation of a map showing the

potential for liquefaction.

e. Preparation of a map showing the

potential for landslides.

f. Analysis of the vulnerability of

various types of buildings and

lifeline systems under a range of

seismic loads.

3. WHAT LEADS US TO BELIEVE IT CAN BE DONE?

3. 1 Basic Data

The basic data required for seismic

microzonation' are available in many countries

throughout the world; what is missing is a standard procedure. Microzonation on the

regional and urban scales requires the best

available information on: 1) seismicity, 2)

the nature of the earthquake source zone, 3)

seismic wave attenuation, 4) local ground

response, and 5) building and lifeline

response.

3 . 2 Scientific and Engineering Problems

A number of technical issues (i.e., questions

for which expert judgment is divided between

"yes" and "no") have been identified for the

problem of microzoning the ground-shaking

hazard. They are summarized below to provide

examples of their range and complexity.

Considerations of structural response and

potential vulnerability will not be discussed

here. Similar technical problems can be

stated for the ground-failure hazard.

Seismicity - The record of historical

seismicity in both the United States and

other countries varies considerably in length

and completeness. Lack of completeness can introduce biases in statistical analyses

unless careful judgments are made.

Incorporating geologic evidence ^6f recent

faulting as well as geodetic data improves

the likelihood of establishing the best

possible recurrence rates for earthquakes.

If geologic and geophysical data are not

available, it may be extremely difficult to

estimate the maximum magnitude in an area,

and indeed, it is possible that a number of

geographic areas may not have experienced

their maximum magnitude earthquake. Use of

the record of historical seismicity alone may cause underestimation of the maximum

magnitude.

The issues include the following:

a. Will the uncertainty involved in using

catalogs of instrumentally recorded and

felt earthquakes representing a short

time interval and a broad regional area permit a precise specification of the

frequency of recurrence of major

earthquakes on a local scale?

209
b. Can the seismic cycle of individual

fault systems be determined accurately

and, if so, can the point in the cycle

be specified?

c. Can the location and magnitude of the

largest earthquake that is physically

possible on an individual fault system

or in a seismotectonic province be

specified accurately? Can the frequency

of this event be specified?

d. Can seismic gaps be identified and their

earthquake potential evaluated

accurately?

e. Can discrepancies between the geologic

evidence for the occurrence of major

tectonic movements in the geologic past

and the evidence provided by current and

historical patterns of seismicity in a geographic region be reconciled?

Seismogenic Zones - No standard method has

been adopted for delineating seismogenic

zones. Usually, each cluster of earthquake

foci on active faults is considered as a source zone; however, scientific judgment is

involved in drawing the boundaries of source

zones. For example, one danger is that two or more regions having different

seismotectonic characteristics will be

incorrectly combined and the resultant

analysis will suggest some average but

nonexistent physical condition. In defining

seismogenic zones, all available information

is used to establish the physical

correlations between earthquake occurrences

and geologic processes and tectonic

structures, including: 1) location of the

boundaries of crustal blocks which are undergoing contrasting displacements, 2)

history of vertical and horizontal regional

tectonic movements, 3) the seismic cycle and

history of active faults, and 4) tectonic

stress. Each seismic source zone is chosen

so that it encloses an area of seismic

activity and, to the extent possible, an area

of related tectonic elements. Although timedependent models are now available,

earthquakes are commonly assumed to have

equal probability of occurrence anywhere in a source zone, to have an average rate of occurrence that is constant in time, and to

follow a Poisson distribution of recurrences.

The technical issues include the following:

a. Can seismic source zones be defined

accurately on the basis of the record of

historical seismicity? On the basis of

geology and tectonics? On the basis of

the record of historical seismicity

generalized by geologic and tectonic

data? Which approach is most accurate?

b. In assessing the earthquake groundshaking hazard for a region, can a magnitude be assigned accurately to the

largest earthquake expected to occur in

a given period of time on a particular

fault system or in a particular seismic

source zone?

c. Can the physical effects of earthquake

source parameters such as stress drop

and seismic moment be quantified and

incorporated in zoning maps?

Seismic Wave Attenuation - Characterization

of the ground motion close to an active fault

is one of the most important yet most

difficult parts of the problem of

constructing a ground-shaking hazard map.

The empirical strong ground motion data are currently too limited to resolve all of the

technical issues concerning the attenuation

characteristics of both near- and far-field

ground motion, even though unique groundmotion data have been acquired in the near

field in the 1979 Imperial Valley,

California, earthquake and in other locations

worldwide. These data have reinforced

current thinking in some areas and revised it

in others, but have not resolved all of the

controversial issues concerning seismic wave attenuation. Frequency-dependent effects of

the transmission path on earthquake ground

motion have not been quantified fully because

of limited data. Observational and

instrumental data indicate that the regional

seismic attenuation rates depend on the

physical properties (i.e., Q structure) of

the Earth's crust and upper mantle in a region, that the attenuation rates can vary

considerably from region to region, and that

Q is frequency dependent.

Attenuation curves are required to specify

how values of peak ground motion (or spectral

velocity ordinants) decrease as distance from

the causative fault increases. Such curves

are essential when constructing a zoning map

of the peak-acceleration ground-shaking

hazards. The problem is that many peakamplitude attenuation curves having

substantial differences exist in the

literature. The question of magnitude

dependence of attenuation is important in

probabilistic ground-shaking hazard

estimation because it sharply influences the

estimated level of maximum ground motion in

two cases: 1) areas having a high rate of

seismicity, and 2) when long periods of time

are considered.

The technical issues include:

a. Can the complex details of the

earthquake fault rupture (e.g., rupture

dimensions, fault type, fault offset,

210
fault slip velocity) be modeled

accurately enough to give precise

estimates of the amplitude and frequency

characteristics of ground motion close

to the fault? Far from the fault?

b. Do values of peak ground-motion

parameters or spectral velocity

ordinants saturate at large magnitude?

Local Ground Response - Since the early

1900 's, literature of earthquake engineering

and engineering seismology has recognized and

documented that structures founded upon

unconsolidated material (soil) are damaged

more frequently and usually more severely in

earthquakes than structures founded on rock. The damage distribution on many

occasions (for example, in the 1967 Caracas,

Venezuela, and the 1985 Mexico earthquakes)

has been recognized as being related to the

specific properties of the site geology.

Many past studies have used empirical groundmotion data and analytical models to define

the frequency-dependent effects that have

been and still are controversial; only

acquisition of ground-motion data recorded at

sites underlain by rock and a variety of soil

columns close to the fault from large- to

great-magnitude earthquakes will resolve

these arguments.

The technical issues include the following:

a. For various soil types, is there a discrete range of peak ground-motion

values and levels of dynamic shear

strain where the ground response is

repeatable and essentially linear? Is

there a range where nonlinear effects

dominate?

b. Can the physical effects of selected

physical properties of the soil and rock

column (e.g., thickness, lithology,

geometry, water content, shear-wave

velocity, and density) be modeled

accurately? Which of these physical

properties control the spatial

variation, duration, and amplitude and

response characteristics of ground

motions in a geographic region for the

fault-site geometries?

c. Can the variation of ground motion with

depth below the surface be modeled

accurately in order to estimate the

ground-shaking effects on underground

lifeline systems?

3 • 3 Seismotectonic Analogs

The optimal approach is to select countries

that have analogous seismotectonic

settings. Most of the zones of seismic

activity in the United States have

counterparts and analogs in other areas of

the world. Much more can be learned about

the tectonic setting, earthquake mechanics,

earthquake hazards, and risk for parts of the

United States by studying tectonic analogs in

other countries. Certain aspects of source zones in the U.S. that are not clearly

understood (i.e., pronounced overburden

masking basement feature, lack of long

historic record of seismicity, etc.) become

impediments or road-blocks to understanding,

particularly if efforts are concentrated

solely on specific features in the U.S.

Critical keys to the understanding of major

strike-slip faults like the San Andreas fault

may come from research and field mapping of

similar features in say Turkey, Guatemala,

New Zealand, Venezuela, the Philippines,

Japan, Alaska, Iran, Pakistan, western China,

and the U.S.S.R. The similar statement can

be made for normal and thrust faults. A

comparison of some of these fault systems

indicates that parts of them are in various

stages of their earthquake cycle. By

studying analogous critical earthquakegenerating features in other countries, it

may be possible to "catch" forerunning or precursory features of large shocks and to

apply that experience prior to the occurrences of the next large earthquake on,

say, the San Andreas fault or other fault

systems in the U.S.

Intraplate earthquakes and the state of

stress in Australia, Canada, northern Europe,

the U.K., parts of Africa, and peninsular

India show many similarities to those

associated with the Central and Eastern

U.S. Many shocks in those areas seem to occur along old fault systems that have moved

many times throughout geologic history in

response to various plate-tectonics events.

The configuration of major tectonic elements

and the reactivation of fault systems in the

Southeastern U.S., the site of the Charleston

earthquake in 1886, are similar to those of

west Africa near Accura, Ghana, and the Benue

trough of Nigeria. The tectonic setting of

the New Madrid seismic zone in the Central

U.S. is similar to that of the seismic zone

that extends into the interior of Australia

near Adelaide. Several zones of intraplate

shocks are similar to those of the eastern

and Central U.S. in that they are characterized by very large areas for a given

level of energy release.

A great deal can be learned from studies in

other countries about the repeat time of

large earthquakes along given segments of

strike-slip and convergent plate

boundaries. The average repeat time is a function of the long-term rate of plate

movement and the geometry of the rupture

zone. Variations in repeat time at a given

place appear to be associated with the length

211
of the rupture zone and the amount of seismic

slip associated with the last large shock in

that zone.

Since the historic record of earthquakes

along the Alaska-Aleutian arc is so short,

information from convergent zones near Japan,

New Zealand, India, Pakistan, the U.S.S.R.,

the Lesser Antilles, Mexico, Central and

South America, and other similar areas can be

applied to earthquake-related problems for

Alaska and the Aleutians. Similarly, the

unusual style of plate motion to the north of

Puerto Rico and the Virgin Islands —thrust

faulting on nearly horizontal planes with the

slip vector nearly parallel to the plate

boundary—is similar to that in the western

most Aleutians and in the Andaman Islands.

4. WHAT IS THE BENEFIT?

4.1 Potential Applications of Seismic

Microzonation Products

Applications of seismic microzonation

products (maps, data, analyses) can be made

in terms of land-use, building codes,

construction practices, repair and

strengthening of existing buildings, and

response and recovery planning. The benefits

in any given country where seismic

microzonation has been performed include:

a. Improving the current building code,

identifying options for modifications

that incorporate the scientific and

engineering lessons learned from past

destructive earthquakes.

b. Improving of regional and urban land-use

practices, identifying options for

alternatives to current practices that

might reduce potential losses.

c. Improving design and construction

practices for new buildings, specifying

options for alternatives to current

practices that might be more effective

in ensuring high quality.

d. Improving the current practices to

repair and strengthen existing

buildings, suggesting options for

alternatives to current practices that

might be more effective.

e. Evaluation of plans for emergency

response and disaster recovery.

4.2 Implementation

Three groups of professionals will implement

the new knowledge provided by seismic

microzonation: the design professional, the

urban and regional land use planner, and the

emergency manager.

5. REFERENCES

1. University of Washington, Proceedings of

the First International Conference on Microzonation for Safer ConstructionResearch and Application, Seattle,

Washington, October 30 - November 3,

1972, 2 volumes.

2. University of Washington, Proceedings of

the Second International Conference on Microzonation for Safer ConstructionResearch and Application, San Francisco,

California, November 26 - December 1,

1978, 3 volumes.

3. University of Washington, Proceedings of

the Third International Earthquake

Microzonation Conference, Seattle,

Washington, June 28 - July 1, 1982, 3 volumes.

4. Ministry of Urban Development and

Construction, Proceedings of the

International Conference on Seismic

Microzonation, Ech Cheliff, Algeria,

October 11-12, 1984, 2 volumes.

5. Thiel, Charles C. , 1988, Seismic

Microzonation: An Approach to Seismic

Land-Use Planning in_ Hays, W. W.

,

(Editor), Proceedings of Conference 41

on "A Review of Earthquake Research

Applications in the National Earthquake

Hazards Reduction Program: 1977-1987,

U.S. Geological Survey Open-File Report

88-13-A, pp. 320-342.

6. Woodward Clyde Consultants, 1984,

Seismic Microzonation of Ech Cheliff

Region, Algeria, Report prepared for

Controle Technique de la Construction,

Algiers, Algeria; 6 volumes.

7. Hays, W. W. , 1985, The Realization of a Seismic Microzonation Study in the Chief

Region of Algeria Following the 1980 El

Asnam Earthquake, _in_ Kusler, Jon A. (Editor), Proceedings of International

Symposium on Housing and Urban

Development After Natural Disasters, pp.

51-57.

8. Panel on Seismic Macro- and Microzonation,

1981, in State-of-the-art in Earthquake

Engineering: Proceedings of the Seventh

World Conference on Earthquake Engineering,

Special Volume, pp. 47-68.

212
Full Scale Test on a Three-Storied

Wood-Framed Building

by

Tatsuo MUROTA* and Motoi YASUMURA**

Abstrac t

A full-scale wood-framed building

of three stories was subjected to the

lateral loads, and the structural performance against the seismic and wind

loads were analyzed. The experiments

were carried out on the specimens

whose floor and walls perpendicular to

the loading direction were not

sheathed with sheet materials as well

as on the complete structure and the

influence of the shear stiffness of a diaphragm and that of a bearing wall

perpendicular to the loading direction

on the structural performance was investigated. The experimental

results were compared 'with the .calculation from the finite element,

model. frame model and also Tuomi's

model based on the load-slip relations

of a nail joint, and the applicability

of these calculation methods to the

design procedure was investigated.

KEY WORDS : S ta t i c loading, Finite element analysis, Horizontal deformation,

Ultimate load. Shear wall, Diaphragms.

1 ntroduct ion

In relation to the revision of

Japanese Building Standard Law [1] in

1987, the full scale tests on a Woodframed House of three stories were

carried out to investigate the structural safety against the seismic and

wind loads.

Building Standard Law, Enforcement

Order requires that the structural

safety of wooden houses having more

than two stories and/or exceeding

500s' in total floor area should be

confirmed by means of structural

calculation.

A purpose of this study is to

analyze the performance of the shear

walls and diaphragms in an actual

structure, and to present the useful

data for the structural design of the

three storied Wood-framed Houses.

Description of specimens

The specimen was a full-scale Woodframed Construction of three stories

whose plans and elevations are shown

in Figs.l and 2. Each story had 66

square meters floor area of 9. 1 meters

long in the longitudinal direction

(loading direction) and 7.28 meters

wide in the transversal direction.

Each story had three continuous walls

in the longitudinal direction at the

both sides and in the center of the

plan. ..Exterior walls were sheathed

with 9.5 millimeters thick Douglus-fir

plywood outside and 12 millimeters

thick gypsumboard inside. Interior

walls were sheathed with 12 millimeters thick gypsumboard on both

sides. Nail spacings were 10 centimeters around the perimeters of a sheathing sheet and 20 centimeters

along the intermediate support. Subfloor was 12 millimeters thick lauan

plywood and roof sheathing was 9.5

millimeters thick Douglus-fir plywood.

No elastmeric adhesives were applied

to the joints between subfloor and

floor joists except for the joint just

under the wall.

Specimens were divided into following three types( See Fig. 3).

Specimen A: Specimen whose floor framings were not sheathed with subfloor

and walls in the transversal direction

* Director, Structural Engineering

Department, Building Research

Institute, Ministry of Construction.

**Dr.Agr., Senior Research Engineer.

Aerodynamics Division, ditto.

213
were sheathed with no sheathing

materials.

Specimen B: Specimen whose floor framings were sheathed with subfloor but

whose wall framings in the transversal

direction were not sheathed with

sheathing materials.

Specimen C: Specimen whose floor framings were sheathed with subfloor and

wall framings were sheathed with

sheathing materials.

Test methods

The lateral static load was applied

at the top of the specimen with three

hydraulic jacks as shown in Fig. 5. Horizontal and vertical displacements

of each story were measured with the

electric transducers and tension force

of a hold-down bolt connecting each

story was measured as shown in Fig. 6. In the Specimen A, the monotonously

increasing loads were applied horizontally at the top of three continuous

walls so that each displacement of the

loading point was equal. In the

Specimens B and C, the reversed cyclic

loads as shown in Fig. 7 were applied

so that the horizontal load applied to

the central wall was one and a half

times as much as those applied to the

south and north walls.

Forced vibration tests were also

carried out before and after the

static loading test of the Specimens B and C -by employing a rotating-mass

generator situated on the ceiling of

the third story, and the response of

each story was measured with the

electromagnetic pickups.

Experimental results

Hor i zon ta 1 resistance

The relation between the horizontal

loads and horizontal displacements at

the top of the south. central and

north walls of the Specimen A is shown

in Fig. 8, and the relations between

the total applied loads and average

displacement of the south, central and

north walls of the Specimens B and C are presented in Figs. 9 and 10

respectively. Table 1 summarizes the

lateral load for the prescribed displacement ratio and the maximum load.

The prescribed displacement ratio is

defined by the ratio of horizontal

displacement to the height of loading

point in the Specimen A and the ratio

of average displacement ofthree walls

to the height of the loading point in

the Specimens B and C. The lateral load for the prescribed

displacement ratio and the maximum

load of the south and central walls in

the Specimen A was respectively 36 to

48% and 55 to 69% smaller than those

of the north wall. Horizontal displacement rajio for the maximum, load

of the south and north walls of the

Specimen A was both 0.032 and that of

the central wall was approximately 20%

smaller than those of the south and

north walls. . The total lateral loads

of three walls in the Specimens B and

C were respectively 20 to 23% and 10

to 15% smaller than those of the

Specimen A. This may be caused by the

difference of the shear stiffness and

strength between the south and north

walls. The lateral load of the

Specimen C was approximately 10%

greater than that of the Specimen B. It is considered that the load applied

to the south wall was not transferred

to the north wall in the Specimen B because of the torsional deformation

of the specimen, while the torsional

moment was supported by the walls in

the transversal direction in the

Specimen C. Horizontal displacement

for the maximum load in the Specimen B (average of three walls) was approximately two-thirds of that of the

Specimen C. This summarizes that the

Specimen B had less deformabi 1 i ty than

the Specimen C because of the lack of

the shear stiffness of the walls in

the transversal direction.

214
Horizontal displacements of each

story are presented in Figs. 11 to 13.

They showed that the horizontal deformations were the most remarkable on

the second story in all the specimens.

Description of failure

In all the specimens the failure

concentrated on the second story especially on the south and central walls.

Fig. 14 shows the failure of the south

wall in the. Specimen C. Exterior

plywood buckled at the compression

side and were torn at the tension side

around an opening, and the nail joints

connecting plywood and studs failed

along the vertical fasteners of

plywood. Similar failures were observed on the interior walls sheathed

with the gypsumboard.

Specimens B and C. The distribution of

forces to the walls in the Specimens B and C was obtained from the relation

between the shear deformation and the

lateral load in each wall of the

Specimen A. The parenthesized values

in thfi figures are the ratio of the

amount of the bearing walls. The

amount of the bearing walls is

defined by the products of the wall

coefficient and the sum of the length

of the bearing walls without openings.

The ratio of shear forces of each wall

was almost constant regardless of the

horizontal deformation, and close to

the ratio of the amount of the bearing

walls. This proves that the lateral

load can be distributed according to

the amount of the bearing walls in

case of designing the lateral stiffness and strength of the structure.

Effects of floor stiffness and perpendicular wal Is

The horizontal deformations of the

floor in the Specimens B and C are presented in Figs. 15 and 16

respectively. These figures show that

the floor of each story in the

Specimen B turned remarkably according

to the increase of the horizontal

deformation, while the torsional

deformation of the floor in the

Specimen C was very small in comparison with the Specimen B. This

proves that the torsional moment

caused by the difference of the shear

stiffness of the walls in the longitudinal direction were supported by

the walls in the transversal direction

and it reduced the torsional deformation of the floor. Figs. 15 and 16 show

also the shear deformations of the

floor of the Specimens B and C were

very small.

Distribution of shear forces

Figs. 17 and 18 show the ratio of

the lateral forces supported by the

continuous walls of each story in the

Uplift of bear i ng walls

Fig. 19 shows vertical displacement

of the bearing walls on the first

story of the south wall, and Figs. 20

to 22 show drawing forces of the bearing walls at the end of each story in

the Specimen C. Drawing force of the

bearing wall on the third story was very small, and it became larger on

the lower story than the higher story.

Drawing force of the north wall was

the largest on the first story, and

that of the center was the smallest

and approximately a half of that of

the north wall.

Dynamic characteristics ,

Forced vibration tests using a rota t i ng-mass vibration generator were

carried out in order to evaluate the

dynamic characteristics of the

Specimens B and C. Figs. 23 and 24 show typical

frequency response curves for EWdirection (the longitudinal direction)

in forced vibration tests, and Table 2 summarizes the natural frequencies

evaluated from the response curves.

215
Microtremor exited vibrations of the

completed house 'with the live load

were also measured after finishing the

exterior and interior! The natural

frequencies of the completed house

evaluated from the fourier spectra of

observed records are summarized in

Table 2. The natural frequencies of torsional mode of the Specimens B and C

were compared to evaluate the-effects

of the walls in the transversal direction on the torsional stiffness. The

results were summarized that the

natural frequency was risen from 4.8

to 8. 8Hz by sheathing the walls in the

transversal direction. This suggests

that the floor of the specimen was sufficiently rigid so that the perpendicular walls increased the torsional

stiffness.

.In order to investigate t'he effects

of, damages of the bearing walls on dynamic characteristics, the natural

frequencies of translational mode

before and after the static loading

tests of the Specimen C were compared.

The results were summarized that the

damage decreased the natural frequency

from 5.8 to 3.1Hz. The natural

frequency of the completed house

decreased from 5.8 to 4. 8Hz comparing

with the unfinished structure. It seems the natural frequency decreased

because of the increase of weight of

finish and live load.

Comparison of the calculated values

with the experimental results

Horizontal displacements of each

story in the Specimen C were compared

with the calculated values computed

from the finite element model, frame

model and Tuomi's model.

In FEM model, walls were modeled

with the triangular plates as shown in

Fig. 25. The modulus of elasticity and

the shear modulus of elasticity of

wall elements were assumed

3 6, 400kgf/cm2 and 1 . 28 5 kg f/cm 2 for ex- terior walls and 30, 000kg f/cm2 and 750

kgf/cm2 for interior walls respectively from the racking "test of wall

panels.

In Frame model, shear walls, lintel

walls and end joists were modeled with

the beam element as shown in Fig. 26.

The modulus of elasticity of .a column

element of 12-by 10cm in section corresponding to the shear wall of

ninety-one centimeters in length was assumed 1, 254, 300kgf/cm 2 for exterior

walls and 812, 800kgf/cm 2 for interior

walls from the racking test of wall

panels. The equivalent section of lintel walls was assumed in consideration

of the composite action between lumber

and plywood. Rigid zone was also as- sumed at the joint of beams and

col umns. Tuomi's model[2] based on the slip

of nail joint between lumber and

plywood sheet in shear wall(See

Fig. 27) was applied to predict the

shear deformation and ultimate

load[3][4]. The slip modulus and

strength of a nail joint was assumed

691kgf/cm and lOOkgf for plywoodlumber joints and 238kgf/cm and 38 kg

f

for gypsumboard-1 umber joints from the

racking test of wall panels.

Figs. 28 to 30 show the comparison

of the experimental results in the

Specimen C with the computed results

from each model. For the computation

of the shear resistance of each story

in the Specimen C, the shear resistance of each wall was simply summed

up neglecting the torsional

deformation. The calculated results

from each model had a tendency to un- derestimate the horizontal stiffness

when the horizontal displacement was very small (when 7 is approximately

less than 0.03) but it seems that the

calculation from these models are applicable to the practical use. Table 3 shows the ultimate load of

each story computed from Tuomi's

model. The computed value of the

second story was the smallest which

coincides with the fact that the

second story was most damaged in the

experiment, and the computed value of

216
the second story showed relatively

good agreement with the maximum load

in the experience. This proves that

Tuomi's model is applicable to predict

the ultimate properties of Wood-framed

building.

Concl us ions

To summarize the results of this

study, the following conclusions were

1 ead.

1. The shear resistance of the north

wall of the specimen was approximately

one and a half times as large as that

of the south wall. However, the difference of shear resistance between

the south and north walls affected

very little on the tors iona 1 deformation of the structure in case it had

sufficient lateral stiffness in the

transversal direction.

2. The shear deformation of the diaphragm was small enough to assume that

the diaphragm was rigid in case of

calculating the lateral resistance of

the construction.

3. The lateral forces distributed on

the walls were approximately proportional to the amount of the bearing

walls.

4. Drawing forces at the end of the

wall were the largest on the first

story and the smallest on the third

story. This suggests that the drawing

forces of the walls on the upper story

should be considered to design those

of the lower story.

5. FEM model, Frame model and Tuomi's

model can be applied to the practical

use to predict the lateral resistance

of Wood-framed building, and Tuomi's

model is proper to predict the ultimate properties of Wood-framed

build i ng.

References

1. Building Standard Law, Notification

No. 56, Ministry of Construction, 1982.

2. TUOMI.R. L. and McCUTCHON, W. J. "Predicting Racking Strength of Lightframe Walls", Preprint for ASCE

Meeting. October 1977.

3. YASUMURA.M. "Racking Resistance of

Wooden Frame Walls with Various

Openings", CIB W18/19-1 5-3. September

1 986.

i. YASUMURA, M. , MUROTA. T. et al.

"Experiments on A Three-Storied Wooden

Frame Building", Proceedings of the

1988 International Conference on Timber Engineering, Vol.2, September 1 988.

217
Table 1. HORIZONTAL RESISTANCE, MAXIMUM LOAD AND

DISPLACEMENT RATIO FOR MAXIMUM LOAD

• (Unitrkfff)

DISPLACEMENT* 1/600 1/300 1/200 1/150 1/120 1/60 Pmax Rmax»*

SPECIMEN

A

SOUTH: 2,050 3,190 3,890 4,500 5,090 6.450 8,700 0.032

CENTER: 1,750 2,350 3.240 3,450 3,550 3,680 4.160 0^025

NORTH

:

3 , 870 5 , 610 6 , 800 7 , 660 8 , 550 11 , 310 13 , 590 0 . 032

SPECIMEN B 6,240 9,080 10,650 12,690 13,980 16 ,660 16.700 0.016

SPECIMEN C 6,970 9,900 11,520 13,570 14,890 17,970 18,670 0.026

* Ratio of horizontal displacement to the height at the loading point

** Displacement ratio for the maximum load

Table 2. NATURAL FREQUENCY OF SPECIMENS

SPECIMEN FIRST MODE SECOND MODE

Hz Hz

SPECIMEN B BEFORE 4.8 6.5

LOADING (TORSION)

BEFORE 5.8 8.8

LOADING (TORSION)

SPECIMEN' C

AFTER 3.1 7.0

LOADING (TORSION)

FINISHED 4.8 7.2

HOUSE (TORSION)

Table 3. Ultimate Strength of each wall

computed from the Tuomi ' s model (Unit : kg-f

)

Wall South Center North Total

IF 6 .110 5,104 10.837 22,051

2F 6 , 902 4,246 9 , 807 20 , 955

(18 . 670)*

3F 8 , 170 4 , 248 10,577 22 , 995

* The parenthesized values represent the ultimate

load of the experiment in the Specimen C.

218
o

CO

c\

1 9 100

o

CO

CM

rFirst Floor

9 100 1 Second Floor

9 100

Third Floor

Flc.l Plans of a specimen.

u u

nonn

South . East

u u u u

D 0 D

a

0 g

•

North West

Klc.2 Elevations of a speclraen.

Specimen A Specimen B Specimen C

Kltf.U KInils of specimens.

219
Fie. 7 History of loading.

KIc.C HcA*urc»cnt of a icnvlon force.

50 100 150 200

Horizontal Oisplccement (mm)

Fie. 8 Load -d l sp Uccmt n t curves In the Specimen A.

-20

Klc.9 l.oad-cl I sp Laccmcn t curves In the Specimen II.

220
3F

3 00

Horizontal Displacement (mm)

L-2 0

Kir. 10 [.oad-dlsplaccmcnt curves In the Specimen C.

2F

1F

"

2F

1F

0 5 0 1 00 %

Horizontal Displacement

Vl20

Kit:. 11 Horizontal tie forma t Ion of cuch story In the Specimen A.

3F

2F

0 50 1 00 *

Horizontal Displacement

1 / 120

Flc.12 Horizontal deformation of each story In the Specimen D.

1F

0

XV—

V

r

50 100%

Horizontal Displacement

•z&bbbhmi

1 A20

i

. \ m mum i ~ n m — r* ir^iurii'

I

Klg.13 llor J zonln 1 deformation of cuch story In the Specimen C. MEM

ill

... J^M;^

Flir.M Fnllurc of llic south wull In the Specimen C.

221
0.38 0.21

(0.39) (0.22)

(0.24)

l

4-

0.32 0.27

(0.28)

i

0.40

(0.39)

-1

0.34 0.24 0.41

(0.27) (0.25) (0.48)

I

1

1 0.41

(0.48)

1 0.36

(0.39)

I- 0.33

(0.27)

I—

0.33

(0.24)

0.22

(0.22)

I 0.21

(0.25)

1 0.23

(0.28)

I

0.42

(0.39)

-1

0.46

(0.48)

-1

0.45

(0.48)

I

I 0.32

(0.39)

0.32

(0.27)

1-

0.31

(0.24)

I

7

0.20

(0.22)

(0.25)

0.21

(0.28)

I

I 0.49

(0.39)

0.49

(0.48)

H

0.48

(0.48)

_J

N N N

1/300 1/120 1/60

r'ltf.17 Dlslrlhiitloii of llic horizontal lo.nl In the Specimen II.

222
f l 1 0.33 0.21 0.46

(0.39)

1—

(0.22)

-4—

(0.39) —\

0.31 0.22 0.47

(0,27)

1—

(0.25)

-4—

(0.48) —

1

0.25 0.26 0.48

(0.24) (0.28) (0.48)

1

1 1

S C N

1 0.33

(0.39)

f—

1 0.22

(0.22) —f—

0.45

(0.39)

0.31

(0.27)

1—

0.21

(0.25) H—

0.48

(0.48) —

1

0.28

(0.24)

1

0.22

(0.28)

1

0.51

(0.48)

_J

S C N

1 I 1 0.35 0.21 0.44

(0.39)

1-

(0.22)

-4-

(0.39)

0.32 0.18 —

1

0.50

(0.27)

h-

(0.25)

-+-

(0.48)

0.29 0.21

H

0.51

(0.24) (0.28) (0.48)

1 I }

sen

1/300 1/120 1/60

FlC 18 Distribution of the horizontal load In the Specimen C.

1/300

1/120

1/60

003 181. 097 214a 124 088 1 62 003

083 433• 245. 243 202 341

4 63

0 34

0_79 7 2 478 773

3 68 2 94 4 56 0 57

(Un!t:mm)

Fir. 19 Vertical displacements of the south wall on the first story.

223
1

X South

a Center

• North

05 10 15 20 *10"

Horizontal Displacement (D/H)

Flff.20 Urawlnt force of the walls on the third story.

4*

0 5

X South

a Center

• North

1 0 1 5 2 0 *10"

Horizontal Displacement (D/H)

Flc.il Drowlne force of the walls on the second story.

C 2

o

/

./

^

X South

a Center

• North

0 5 1 0 1 5 2 0 »10"

Horl2ontol Displacement (0/H)

I' Ik''-'- Dr.-i»lnK force of the walls on tlie first slor

224
6 4 5

Frequency (Hz)

Flu. l'rcn,ucncy response curves for EW- d 1 rcc 1 1 on In the Specimen D.

01 | £2 I 01

I 162

I 91

I ie?

I 91

Fit. 27 RacVlnc aodel of « shear «lo<nC In TuooL's model.

225
Tuomi ' s Model

Frame Hod el

Finite Element Model

Experiment

o 5 TO 15 X10'3 .Horizontal Displacement Ratio(D/H)

Flr.2« Cose*rl<on of the «x.;«rlm«ntil results with th« calculated

results on the first story of the Spec laen C.

Tuom i

' s Hode

1

nt Model

m e n t

O 5 10 15 XlO3 Horizontal Displacement Ratio(D/H)

Fir. 29 Coeparlson of the experiaentJ.1 results with the calculated

result* on the second acory of the Specimen C.

Tuom i

' s Model

Frame Model

Finite Element Model

Experiment

0 5 10 15 xicr3 Horizontal Displacement Ratio('D/H)

Coaparl jon of the e xp« r la* n ta 1 reaulca with th«j calculated

results on the tMrd storj of the Speclsen C.

226
Damage by the Chiba-ken Toho-Oki Earthquake

by

Yasushi SASAKI

1

ABSTRACT

An earthquake of magnitude 6.7 occurred off

the eastern shore of Chiba Prefecture at

approximately 11:8 AM on December 17, 1987.

This earthquake took two human lives, when

they were crushed to death under fallen

concrete block walls and a stone lantern,

and caused damage to various civil

engineering structures and buildings.

Reports were already published on this

earthquake, and studies are still being

continued. In this report, the author

summarizes the damage of this earthquake

based on the two papersl)»2) published

previously.

KEY WORDS : Earthquake, Disaster, Chiba-ken

Toho-Oki Earthquake.

1. OUTLINE OF EARTHQUAKE

According to the Meteorological Agency 3 ), the epicenter of the earthquake has located

at 35°21' north latitude and 140°29' east

longitude with the hypocenter depth of

58 km, and the magnitude of the earthquake

was 6.7.

Seismic intensity of V on the JMA scale was recorded at Katsuura, Chiba, and Choshi, IV

137. 00 I.'19.110

37.00

35.00

at Tateyama, Tokyo, Yokohama, Mito, Ajiro,

Kirmagaya and Kawaguchiko, and III at Ohshima, Mishima, Utsunomiya, Chichibu,

Niijima, Maebashi, Nikko, Koufu, Onahama,

Snizuoka, Shirakawa, Karuizawa, Iida and

Hachi jou j ima

.

Fig. 1 shows the seismic intensity at

various places. It also shows approximate

boundaries between the seismic intensities V

and IV, between IV and III and between III and II. The range where the seismic

intensity of V was observed was the area at

the distance of less than 50 km from the

epicenter within Chiba Prefecture. It was reported that the intensity of earthquake

motion varied even within the range of

seismic intensity of V in Chiba Prefecture

shown in Fig. 1 according to the results of

detailed survey. :

Choush i

Main Shock

JMA aclaaic

Intanalty ical« Acceleration ot

« j( thqu-ka [94!

)

NO f«el lng allqht

I II HMk

Mo f«eLlncj 1.5 •Light

D.I

fltfOM

IV Strong

25

v v«ry •tcoog

SO

VI Dle#«tr cu»

VII Vary

400

6 7 3 9 L0 11 12

137.00 139.00 Ml. 00 Fig

.

Distribution of Kawasumi's

Intensity near epicenter

(after Harukawa 3 ') Fig. 1 Seismic intensity (JMA scale)

at various places Head, Ground Vibration Division,

Public Works Research Institute,

Ministry of Construction

227
Harukawa 3 ' conducted a questionnaire

survey by asking questions of the residents

in 47 cities, towns and villages in Chiba

Prefecture and determined a seismic

intensity distribution in Chiba Prefecture

based on about 2,900 answers obtained. The

results of Harukawa's survey are shown in

Fig. 2. As "Kawasumi's seismic intensity"

is used as indexes of seismic intensity in

his survey, Kawasumi's intensity is compared

to the JMA scale in Fig. 2.

According to the results of the survey, it can be known that there were areas near Tougane City and Chounan Town where the

earthquake motion was locally stronger.

From the study of earthquake mechanism,

Mizone 5) reported that this earthquake was

caused by the strike slip fault in the

northeast-southwest direction on P-axis

(principal axis of maximum stress) which was created inside the plate in the Philippines

Sea. One of nodal planes (potential shear

plane) is almost directed toward north-south

which is the direction of distribution of

aftershocks.

By this earthquake, many strong-motion

seismographs for engineering purposes

installed in Tokyo and Chiba Prefecture were

trigered. Recorded maximum values were summarized in the prompt report of the

earthquake.^' Among these records,

numerically processed and corrected for

instruments by PWRI are shown in Fig. 3 which were obtained on ground and

underground near ground surface. However,

part of the records were not digitized

since the maximum acceleration values were

not large. Fig. 3 also shows the maximum

acceleration^' (maximum value of the

resultant in two horizontal directions)

observed by the Tokyo Gas Co.

Fig. 4 shows the relation between maximum

acceleration and epicentral distance.

Fig. 4 also shows the values for Class 1 ground of the following regression

equation**) which indicates the distance

attenuation of maximum horizontal

acceleration obtained from previous records

of strong earthquakes:

A max = 987.4 x io 0 - 216 M

232.5 x 10 0 - 313 M 403.8 x io 0 - 265 M

x ( A + 30) -1 - 218 .

(Class 1 ground)

(Class 2 ground)

(Class 3 ground)

(1)

Misato interchange

G.L. -5

In premises of PWRI G.L. -2 Suigou Bridge

G.L. -15

LC. TR, b'-D

37 32 32

N-S E-W U-D

58 61 28

N-S E-W U-D

95 113

On the ground at Shin-Katauihika

Bridge (Smac-b 2

)

X-S 1 E-W L'-D

53 70 46

Three-span continuous

bridge on Metropolitan

Expressway G.L. -4

N'-S E-W U-D

47 43 21

On the ground at

X-S E-W L'-D

66 77 27

On the ground at Kannonzaki (SMAC-B 2 )

I X-S E-W U-D

118 166 39

35.00

On the ground at Onigawa Bridge (SMAC-Ej)

N-S E-W U-D

204 146 51 On the ground at the estuary weir of the Tone River (SM-10A) N-S E-W U-D

130 143 44

On the Nagara

basement of Dam

LG. TR. U-D

138 262 86

On the ground at Araatsu Overbridge {SMAC-Q)

N-S E-W U-D

1 44 58 40

2 104 101 42

35.00

O

Observed value (resultant in 2 directions) by Tokyo Gas Co. Observed values on the ground {respective values in two directions before numerical processing instrument correction) by

strong-motion seismographs installed by the Ministry of Construction and others; unit in gal; value shown outaide fig. 3 is the maximum acceleration after numerical processing instrument

corrections

.

Fig- 3 Maximum acceleration at various places

141.00

228
Range of data collected by Txakyo Gas Co.

Epicentral distance (km)

* Values of max. horizontal acceleration and max. vertical

acceleration after numerical processing instrument corrections are shown black. Readings from analog records are shown in white.

• Data recorded by Tokyo Gas Co. are resultant max. acceleration for two horizontal directions.

Fig. 4 Epicentral distance and maximum

acceleration

where, M: Magnitude, A : epicentral

distance

.

Tendency of the records obtained this time

was almost similar to those of previous

earthquakes, the size of acceleration was also similar to those of previous

earthquakes, and the attenuation due to distance was greater compared to the

regression equation based on many previous

earthquake records.

As understandable from Figs. 3 and 4, large

maximum acceleration values were observed in

part of the region. This occurred probably

because of local variation in ground

conditions.

If the acceleration in two horizontal

orthogonal directions and the acceleration

in a vertical direction are observed at a certain observation point and are vector -synthesized, then it is possible to obtain the locus of acceleration vector at

that point. This kind of locus is called

the orbit of acceleration record. The orbit

has a complicated shape but is normally

distorted in a certain direction and the

acceleration in that direction is large.

This direction is called the principal axis

of the acceleration of earthquake motion or the dominant direction of vibration. ^)

The principal axis can be determined also

relative to velocity or displacement.

Principal axis of acceleration and the

principal axis of displacement of the

records obtained by the present earthquake

are shown in Fig. 5. Earthquake motion is

dominant in the direction of the principal

axis, and some dominant directions of

earthquake motion at observation points

shown in Fig. 5 are orthogonal to the

epicentral direction but others are not,

thereby indicating the irregularity. This seems to occur because the earthquake motion

is strongly affected by local topography or ground conditions near the observation

points.

(a) Acceleration records

(b) Displacement records

Fig. 5 Principal axis of earthquake motion

(projection of dominant principal axis

to horizontal plane)

2. OUTLINE OF THE DAMAGE

The outline of the damage caused by the

earthquake is shown in Tables 1 and 2.

This earthquake took two human lives, when

they were crushed to death under fallen

concrete wall blocks (Ichihara City, Chiba

Prefecture) and a stone lantern. (Mobara

229
Table 1 General Aspect of damage (As of

March 22, 1988, surveyed by the

National Land Agency)

Class Unit Amount of damage

Per sons Dead Persons 2 persons Wounded Per sons 161 persons

Houses Totally

collapsed

Half collapsed

Par tially

damaged

Houses

Houses

Houses

16

102

72,580

Table 2 Damage to facilities (as of March

22, 1988, partially as of February

29, surveyed by Chiba Prefecture)

Facilities Number of places

Amount of damage

(million

yen)

Civil Facilities related to 153 2,781

engineer lng Ministry of Construcfacilities tion Facilities related to 13 197 Ministry of Agriculture,

Forestry I Fishery

(fishing ports) Facilities related to .10 5 Ministry of Transport

(ports & harbors)

Subtotal 176 2,983

Other Farmland, agricultural 390 2,688

fac il i ties, facilities

etc Agricultural common use facilities Agricultural crops Non-common use

fac il it ies

6 59

4

3

Devastation of forest 100 864 land National school 1 4 facilities

Public school facilities 55 235 Cultural properties 2 3 Children's welfare 3 5 facili ties The disabled rehabilita- 4 26 tion S welfare facilities

Welfare facilities foe 2 10 the aged

Subtotal 563 3,901

Total 739 6,884

City, Chiba Prefecture), and 161 persons were wounded. Among the damage to houses,

broken ridge roofing tiles were remarkable.

Ridge roofing tiles of many houses were broken at many places mainly in the eastern

part of Chiba Prefecture close to the

hypocenter. Among approximately 72,700

damaged houses, most houses were reported to

be partially damaged, and the contents of

damage were mostly broken ridge roofing

- lies .

Distribution of the damaged wooden houses is

shown in Fig. 6. Statistics of the total

number of houses were not available, so that

the total number of households in cities,

towns and villages at the end of February

1988 was used instead. And ratio of the

Tougane City

Kujukuri Town

Chousei Village

Damage ratio (%)

icen ter

L to less than 10

L0 to less than

20 to less than 40

| 40 to less than 60

60 and over

Fig. 6 Distribution of damage ratios for

houses by city, town and village

number of damaged houses to the total number

of households was determined as damage ratio

for each city, town and village and is shown

in Fig. 6. Damage ratios of houses in

Tougane City and Chousei Village exceed 60%,

and districts having large damage ratios are distributed mainly around both these

districts. This means that slightly larger

earthquake motions acted mainly to both the

districts during the earthquake.

Cracks which might cause slope failure were noticed in slopes at 10 places (2 places in

Narutou Town, 3 places in Matsuo Town, 1 place in Isumi Town, 1 place in Chounan

Town, 2 places in Tougane City, and 1 place

in Ichihara City in Chiba Prefecture), so

167 persons from 47 households in two cities

and 4 towns evacuated to safe places.

Damage to public and other facilities

occurred at 739 places in both Ibaraki and

Chiba Prefectures, and the total amount of

damage was reported to be 6.88 billion yen.

Among the damage to civil engineering

facilities, the damaged facilities under the

jurisdiction of the Ministry of Construction

are shown by kind of facility in Table 3.

Facilities which were relatively damaged,

will be explained later mainly with respect

to the facilities under the jurisdiction of

the Ministry of Construction. Damage given

to river facilities were the settlement and

cracking of levees and revetments. Damage

to road facilities under the jurisdiction of

the Ministry of Construction were the

settlement of fills and depression of road

surfaces which occurred on National Highway

No. 126 at Daikata in Tougane City.

230
Table 3

( Damage to the facilities under jurisdiction of the Ministry of Construction (as of April 4, 1988,

surveyed by the Disaster Prevention Section, Ministry

of Construction)

Class

Projects executed

by central government

Projects executed

by Local (Prefecture!) Government

Total

Number of place*

Amount

Number of places AmountNumber of places Amount

Riven

Steep

slope*

Roads

Br idges Sewerage

11

1

(illion

yen)

9(2

5«

45

1

90

5

S

(million

yen)

1,082

2

629

52 248

56

1

91

5

5

(million

yen)

2,044

2

68}

52 248

Total 12 1,016 US 2.01] 158 3,029

Facilities under the jurisdiction of the

local government were damaged at 45 places

in rivers, 5 bridges, 90 places on roads and

5 places of sewerage. Damage to sewer facilities occurred in both Tougane and

Mobara Cities; manholes and pipe ducts were

damaged (at 883 places in Tougane City & 61

places in Mobara City) , and the Choho pump

station as well as water supply pipes and a sludge disposal yard of the Kawanakajima

terminal treatment plant were damaged in

Mobara City. Places of damage to rivers,

bridges and road facilities related to subsidized projects in Chiba Prefecture are

shown in Fig. 7. It can be known in this

Figure that more damage occurred also in the

eastern part of Chiba Prefecture which is

close to the epicenter.

O Damage to roads Damage to rivers A Damage to bridges

Fig. 7 Places of damage to rivers, bridges

and roads (related to subsidized

projects in Chiba Prefecture)

In addition to the damage to facilities, the

phenomena of sand boils were observed at

various places. The places where sand boils were surveyed in detail by Chiba

Prefectures^-' and Yasuda et al.H) One

of the characteristics seen at the places of

sand boils was that they occurred not only

in reclaimed land, at beaches and lakes, but

also in reclaimed land in mountainous

regions. By comparing the epicentral

distance of the places where liquefaction

occurred to the limit epicentral distance

explained by Kuribayashi and Tatsuoka ,^2)

it can be known that the liquefaction

occurred during the earthquake at the places

having longer epicentral distances compared

to previous experience. An example of the occurrence of liquefaction at a place

farther than Kur ibayashi-Tatsuoka 's limit

epicentral distance was also reported with

respect to the Lake Hachirougata 13 ' , and

the liquefactio seems to occur at the places

exceeding the limit epicentral distance

depending upon ground conditions or earthquake motion characteristics.

As seen in the cases of Chounan Junior High

School and the levee of the Tone River, some structures are considered to be damaged

directly by the liquefaction. But

structures located near the places of sand

boils were not always damaged during this

earthquake. Probably, the reason of this

feature may be related to the depth and

thickness of liquefied layer. Liquefaction

in Urayasu City located along the coast of

the Tokyo Bay will be explained later.

3. DAMAGE TO RIVER FACILITIES

In the main stream of the Tone River, the

phenomena of sand boils were observed not

only in protected low land's and flood

channels near the^Tone River Lower Reach

Construction Office, which manages the lower

reach, but also at other places.

The places, where the phenomena of sand

boils was observed, were the areas which were considered to be old river courses or the crossing portions of old rivers in all cases; the ground seems to be easily

liquefied in these areas. In a wide place

such as flood channel, the sand boiling was observed along continuous crack-shaped holes

in the probable direction of the bank of the

old river. In an area along the levee, the

sand boiling continued in the direction

parallel to the alignment of the levee.

Photo 1 shows the sand boil in the flood

channel near 39 km of the right bank.

Near the right bank of 47 km of the Tone

River shown in Photo 2, the sand boil was observed in a paddy field at the protected

low land side, at the toe of the slope at

the protected land side, and at the berm on

the water side. In the berm at the

protected low land side, several stripes of

cracks 20 cm wide and approximately 160 cm deep were created, and the levee crown was

231
Photo 1 Sand boil in flood channel near

39 km (Sawara City) at the

right bank of the Tone River

Photo 2 Cracks in inner berm near 47 km at

the right bank of the Tone River

settled. Though the amount of settlement

was not accurately found, it looked about

30 cm maximum during observation in field

surveys.

Transverse cracks were observed near the

places where the maximum settlement occurred

during field surveys, but these cracks were

not found immediately after the earthquake.

Therefore, in some cases, cracks in levees seem to appear on the surface with a time

delay after the dispersion of pore water

Photo 3 Depressed levee at the right bank

upstream of Yokotone Gate

Photo 4 Depressed levee at the right bank

downstream of the Yokotone Gate

pressure inside the liquefied ground. This

should be noted during the survey after an earthquake

.

Photos 3 and 4 show the situation of damage

near the Yokotone River Gate located at the

confluence of the Yokotone River near 40 km

at the left bank.

Photos 5 and 6 and Fig. 8 show the situation

of damage near 1.25 km at the left bank of

the Hitachi Tone River.

232
4.3m

—I / 1 v 1— Private house

i- 7n: ,/ 1 ;o.r m oc £a , ml3nd Hitachi Tone River / > L-o

2. I {No trace of sand Doll)

Revetment - Not damaged

Photo 5 Overall view of the damaged levee

near 1.25 km at the left bank of

the Hitachi Tone River

<<

N

<'

\

*> 3

\ :/

J

1

*.l ' \\

"j i

h 0 X o w

\

Crack 10 to

20 cm wide U-shaped gutter - Not damage

Crack 20 cm wide

,

30 to 100 cm deep

Fig . 8

Settlement of crown

of 50 to 80 cm

Damage to the levee near 1.25 km at

the left bank of the Hitachi

Tone River

Photo 6 Cracks in the levee near 1.25 km at

the left bank of the Hitachi

Tone River

The levee at this place had been constructed

in a tidal marsh previously, and the

phenomenon of sand boiling was not observed

near the damaged place. The levee was relatively low with a height of about 2 m, and the material of the levee was sand of

almost uniform grain size. Automatic Ram

sounding and Dutch Cone sounding were performed here immediately after the

earthquake, and qc value of the levee was 7 to 10 kg/cm so that the material was loose.

Also, part of the levee was so loose that it

settled by itself.

Immediately below the levee, loose sand of

almost uniform grain size, probably having a similar grain size distribution to that of

the levee, was located to a thickness of 1.5

to 2.0 m and, below this sand, there was a sand layer which was relatively dense.

At the protected low land side of the

damaged place, private land filled to 1.0 to

1.5 m above the original ground was located

adjacent to the damaged place. There were

U-shaped concrete gutters at the boundary

between the private land and the river

reservation. But no damage to the private

land and the concrete gutters was noticed.

One of the causes is considered to be the

liquefaction in the levee below the

groundwater level and in a very thin portion

of the loose sand layer immediately below

the levee. And one of the remote causes

will be the raised groundwater level inside

the levee by the fill on the private land at

the protected low land side as explained

above

.

233
4 . DAMAGE TO ROAD FACILITIES

On the National Highway No. 126 at Daikata

in Tougane City in Chiba Prefecture, a difference in level of about 7 cm was created on the fill at the connection with

an overbridge above JR Tougane Railway Line

by the earthquake, also a depression of

about 20 cm deep, 4 ra wide and 100 m long was created on the road surface of the

National Highway No. 126 at Daikata.

Depression 4 m wide, LOO m long and about 20 cm deep in the form of a hollow

.

view foe Photo 10

To Choushi

Approx. 4 m

A depression about 20 cm deep

to the form of a hollow

Street gutter

Cement stabilization

treatment

Fig

.

9 Depression of road surface on the

National Highway No. 126 at

Daikata, Tougane City (No. 10 point

shown in Table 3 and Fig. 6)

Photo 7 Difference in level created at the

connection with an overbridge on

the National Highway No. 126 at

Daikata

Photo 8 A depression created on the road

surface on the National Highway

No. 126 at Daikata in Tougane City

Photo 9 Damage to road surface on the

Iioka Ichinomiya Line of

Prefectural Road at Ichinomiya

Photo 7 shows the situation of emergency

repair of the fill at the connection to the

Daikata Overbridge and the difference in

level at the curb, and Photo 8 and Fig. 9 shows a partially depressed road surface.

The portion where a depression was created

on the road surface was on a low earth fill

which probably had been widened before the

earthquake, and only the newly filled

portion was settled resulting in a depression with the maximum level difference

of 20 cm on the road surface.

Photo 9 shows the situation of damage to roads on I ioka-Ich inomiya Line of the

prefectural road at Ichinomiya. In this

section, the road runs through pine forests

on flat land, and the road surface meandered

over a length of 400 m. On December 18,

1987 when the field survey was conducted, a level difference of 4 to 5 cm created on

234
part of the road surface had been already

repaired by asphalt paving by emergency

measures

.

The meandering in this section was created

by the earthquake in such a manner that the

road surface was waved upward, downward and

laterally so that the paved surface was separated from the raised or lowered from

the roadside gutters. At the furthest north

end portion of the damaged section, upward

and downward wave on the road surface had

the wavelength of 15 to 16 m, the wave

height of the swell was about 8 cm, and the

swell with 5 to 6 waves continued. The

difference in level between the road surface

and the top of roadside gutters was the

largest at the central part of the damaged

section, and the maximum difference in level

was approximately 18 cm. Gaps of about 5 cm

were observed between both the ends of

gutters and the road and private land,

thereby indicating the trace of horizontal

movement of paved road relative to the

adjacent natural ground.

Photo 10 shows the situation of the collapse

of cut-slope at a road side on the National

Highway No. 128 at Katsuura.

At the time of the occurrence of the

earthquake, the site for the highway was in

the stage of spraying works with the top

soil and plants excavated from the face of

slope as shown in Fig. 10. And the surface

layer of weathered sandstone slope collapsed

from the earthquake. According to the

workers working on the site at the time of

the earthquake, the whole of the face of the

slope collapsed in a moment leaving a cloud

of dust. No spring water from the slope was recognized. The first report of the damage

indicated the collapsed volume of about

1,500 . Part of collapsed earth was scattered to the National Highway No. 128

and the Katsuura toll road, but most of the

collapsed earth was stopped by the

protection fences installed for the slope

works, causing almost no disturbance to road

traffic.

In addition to these damage, relatively

minor damage, such as cracks in road

surfaces and differences in level, were

caused also on the roads in Ibaraki

Prefecture. Fig. 11 shows the situation of

damage which occurred on the National

Highway No. 125 near the boundary between

Sakuragawa Village and Higashi Village.

This section is a fill section 2 to 3 m high

crossing exit roads, and transverse cracks

near the boundary between fill and cut and

roadside cracks in fill portion were caused.

Portion where

che spray work

was completed

Elevated portion

of Katsuura toll \ road

Kamogawa Katsuura

(a) Front view of the damaged portion

Portion where spray work was completed

Elevated Katsuura

I Toll Road

//A, / Protection

///A 1 '1 fence

National Highway No. 128

Toll Road

Karoo^awa Katsuura

(51 Side view of the damaged portion (c) Plan view of the damaged portion

Fig. 10 Damage to the National Highway No. 128 (Katsuura City)

(No. 11 point in Table 3 and Fig. 6)

235
Fig. 11 Damage to National Highway No. 125 (at Sakuragawa)

5. SAND BOIL IN URAYASU DISTRICT

Phenomena of sand boil were recognized after

the earthquake also in Urayasu City and

Takasu in Chiba City located along the Tokyo

Bay in Chiba Prefecture.

Among them, an example of sand boil occurred

near Mihama in Urayasu City as shown in

Photo 11. The sand boil occurred in a strip-like area at the south side of the

Urayasu City Hall. Fig. 12(a) shows the

range of sand boil summarized by the city

office. Fig. 12(b) shows the former coastal

line read from an old topographic map

(surveyed in 1945) near the relevant area. Ground in this area was formed by filling-up

recently, and the whole neighboring area including the strip-like area where the sand

boil occurred this time has newly filled

ground.

Nevertheless, the phenomenon of sand boil

occurred only in a part of the area along

the Sakai River, and this seems to be

Photo 11 Sand boiling in Urayasu City

(a) Range where sand boiled

(b) Range of sand boil and the location of

former coastal line based on partially

corrected map in 1945

Fig. 12 Places of sand boiling in Urayasu

City (No. 25 point in Table 3 and

Fig. 6)

236
strange. The reason may not be found unless

a detailed survey is made. However, the

former coastal line coincides with the

lagoon-shaped portion projecting out in the

form of a sand bar/ so that there may be a thin earth layer containing relatively fine

grained soil below and near the surface

layer. Thus, it is possible that the

groundwater level in the thin earth layer was locally higher than that in the

surrounding portion.

Fortunately, damage to structures in the

sand boiling area as not reported except for

cracks in road surfaces, houses were not

deformed even one week later. Although occurrence of liquefaction in the ground was apparent during this earthquake, it did not cause damage to structures at some placed.

The reason for this should be reviewed in

detail based on the information concerning

the magnitude of ground motion, ground

conditions, groundwater levels, etc.

6. DAMAGE TO BURIED STRUCTURES

The earthquake caused damage not only to houses and fill structures such as embankments but also to sewer pipelines,

manholes, water supply pipes and gas pipes

which were buried underground as stated

previously . ^)

According to the case studies of damage to buried structures by the earthquakes which

occurred in the past, the direct causes of

the damage are the liquefaction of the

ground and the accompanied flow failure of

the ground in many cases. Liquefaction of

the ground and the flow of the ground often

accompany the sand boil or cracks in the

ground surfaces, so that these abnormal

states on the ground are greatly helpful for

detecting the location of damage to underground structures which otherwise

cannot be found directly by visual

observation. However, damage to underground

structures will not cause any abnormal

states on the nearby ground surface in some

cases. This occurs because the mechanism of

causing damage to the buried structures is

very complicated.

Therefore, it will be checked here whether

the location of damage to buried structures

by the present earthquake can be correlated

to the changes in the nearby ground

surface. This checking will be made with

respect to the damage to water supply pipes.

According to the data reported by the Chiba

Prefectural Government, 652 cases of damage

to the waterworks in the whole of Chiba

Prefecture were reported. Breakdown of

these cases is: 122 cases in Keiyou

district, 11 cases in Kimitsu district, 16 cases in Katori district, 3 cases in

Youkaichiba district, 222 cases in Sanbu

d-istrict, 257 cases in Chousei district, and

21 cases in Isumi district. The number of cases of damage occurred in both Chousei and

Sanbu amounts to 73% of the whole.

Table 4 Damage to water supply pipes and

kinds of pipes

(Unit: Number of places)

Form of damage

Kind of DIP© Broken Detached Broken

Total

joint joint pipe

Asbestos 14 6 59 79 pipe

Vinyl pipe 43 10 28 81 Steel pipe 14 6 10 30 Ductile cast 1 9 0 10 iron pipe

Total 72 31 97 200

Table 5 Damage to water supply pipes and

pipe diameter

(Unit: Number of places)

Form of damage

Pipe diameter

Broken Detached Broken Total

joint joint pipe

Less than 52 16 71 139 100 mm

100 to 16 10 20 46 200 mm

200 to 2 4 6 12 300 mm More than y- 2 1 0 3 300 mm

Total 72 31 97 200

>\s Damage to water works {Giousei 87, Sanbu 113)

•s Damage to roads (Chousei 37, Sanbu 24) : Hatched portion shows mountainous area (hilly land).

o: Areas where the damage is concentrated.

Fig. L3 Distribution of the damaged

portions of water supply pipelines

in Sanbu and Chousei districts

237
villages are concentrated in these areas and

thus the density of water supply pipes there

is relatively high. On the other hand, the

damage to roads occurred more in the

districts along valleys in mountainous areas and in the districts near the coast. Most

damage to these roads as made on earth fills.

From the situation of this distribution, it

is difficult to find certain characteristics

of the relation between the damage to roads

and damage to water supply pipes. However,

according to the damage distribution in a small range, damage at more than 6 places was concentrated in 13 circular ranges with

the radius of 1 km. Topographic features of

these 13 ranges (districts) are as shown in

Table 6.

Topographic features of the districts, where

the damage as concentrated, were the change

in the ground conditions.

As shown in Table 6, some districts have

damage to roads while other districts have

damage to water supply pipes but no damage

to roads among the 13 districts. Thus, the

location of damage to underground water

supply pipes cannot be detected based on the

Table 6 Characteristics of the districts where the damage to water supply pipes as concentrated

Region District name

Density of damage

to waterworks:

Number of cases

shown in ( )

Density of damage

to roads: Number of cases

shown in ( )

Characteristics of topographic outline at damaged place

Chouse

i

(1) Downtown area in Mobara City4.1 (13) 0.3 (1) Fill, and boundary between fill and other topography (such as dune)

(2) Kokufuzeki area in Mobara City

2.9 ( 9) 0.6 (2) Flattened land, plain at the bottom of valley,

or boundary between low position of dune and slope (steep slope)

(3) Central part in Chounan Town

3.5 (11) 2.2 (7) (Same as (2) above] + high fill

(4) North area in Ichinomiya Town2.9 ( 9) 0 (0) Natural levee, sand bar, and boundary between

these and other topography (reclaimed land,

plain on coast)

(5) Near Shin-Ichinomiya

Bridge in Ichinomiya Town

0.6 ( 2) 0.6 (2) Natural levee

(6) Toroni area in Ichinomiya Town

1.6 ( 5) 1.0 (3) [Same as (2) above) + plain on coast

Sanbu (7) Built-up area in Tougane City

4.1 (13) 0 (0) Fill (Most Downtown area was developed after

1975)

(8) Yamaguchi area in Tougane City

2.5 ( 8) 0 (0) Boundary to low position of terrace in low hinterland [Locally the same as (2) above]

(9) Daizudani area in Tougane City2.2 ( 7) 0 (0) [Same as (2) above) fill, low hinterland

(10) Matsunokyo area in Tougane City

1.9 ( 6) 0.3 (1) (Same as (2) above] + fill

(11) Central part of Narutou Town6.0 (19) 0.6 (2) Fill, flattened land (reclaimed land),

boundary between natural levee and plain on coast

(12) West area of Ohami Shirasato Town

3.2 (10) 0 (0) [Same as (2) above) + fill

(13) East coast area of Ohami Shirasato Town

0.3 ( 1) 1.6 (5) Sand bar

Note: Density of damage Number of damage cases/km

Among the places of damage to waterworks in both the districts, the restoration works

were performed at 87 places in Chousei

district and at 113 places in Sanbu

district. Relations between the damage and

the kind and diameter of pipes in 200 places

in total are shown in Tables 4 and 5. And

the places of damage are shown in Fig. 13.

Fig. 13 also shows the places of damage in

both Chousei and Sanbu districts, among the

damage to roads indicated in Fig. 7.

The kinds of pipes which were more damaged were asbestos pipes and vinyl pipes;

contents of damage were broken pipe body in

the case of asbestos pipes and broken joints

in the case of vinyl pipes. Pipes with

smaller diameters were more damaged, and

most damage as broken joints and broken pipe

bodies of small pipes with diameters smaller

than 100 mm.

From Fig. 13 showing the distribution of the

places of damage to water supply pipes, it can be known that the damage to water supply

pipes is distributed in the portions close

to mountainous regions and mountainous

portions along valleys at the western end of

Kujukuri low land. This occurs because

238
abnormal states on ground surfaces alone,

such as by damaged roads. However, if there

are damaged roads and water supply pipes are located nearby, then the water supply pipes

are also damaged in many cases.

Damage to structures buried relatively close

to the ground surface are caused by the

deformation of the ground such as settlement

or cracks near the buried structures or by

large strains in the ground created during

the earthquake. Therefore, for detecting

the location of damaged underground

structures, it will be helpful to know the

abnormal states (residual deformation)

occurring on the ground and also to find the

traces of dynamic strains created during the

earthquake. However, at present, it is

difficult to clarify the proper traces

capable of suggesting the places where large

strains actually occurred in the ground

during earthquakes.

7. BEHAVIOR OF THE NAGARA DAM

The Nagara Dam was under construction in

Nagara Town, Chousei County, Chiba

Prefecture by the Water Resources

Development Public Corporation. And hair

cracks were observed at several places on

the crown of this dam body and a slight

settlement was measured after the

earthquake. Accurate records of anomaly and

valuable acceleration records by

strong-motion seismographs were obtained

during the earthquake since the dam

reservoir was in the stage of test

filling. 16 '

The Nagara Dam is a zone type earth fill

dam (embankment dam) 52 m high.

Construction work for the dam was started in

January 1973, the embankment was built from

October 1980 to May 1985, and test filling

of water to the reservoir was started in

April 1986. Water level during the

earthquake was near El = 57.0 m.

Fig. 14 shows the plan of the Nagara Dam,

and Fig. 15 shows its section. According to

the visual observation immediately after the

earthquake, no anomaly was recognized on the

riprap work for the slope of the

embankment. But it was reported that hair

cracks were recognized at several places on

the top asphalt and on the pavement surface

of a shifted road near the crown. Fig. 16

shows the location of the hair cracks

noticed

.

Also, traces of several cm of settlement of

embankment were found at the transitional

part between the spillway and dam body.

Also, slightly increased gaps in concrete

block joints of spillway were recognized at the top.

On January 1988, a follow-up survey was made

of the cracks previously recognized at the crown of the embankment. The crown asphalt

was peeled off and lime water was poured

into the hair cracks. The maximum depth

where the traces of lime water were recognized was about 1 m. When the

discontinuous surface was scraped with a planting trowel, the surface was peeled off

to the depth of about 2 m maximum. The hair

cracks were not continued from upstream side

to downstream side, and it was verified that

the permeability coefficient of the

embankment directly below the hair cracks

satisfied the standard value of construction

management, thereby assuring impermeability.

Measuring points for surveying were

installed on the crown of the dam and on the

upstream and downstream slopes, and the

measurements were made on December 14 before

the earthquake. The results were compared

to the results of surveying performed on

Fig. 14 Plan of the Nagara Dam

(including settlement)

Counter-weight H.W.L.

C -2nS

(59.0) /zone mfi

/ Silty-Y" ^mudstone^ Zone I ,„ ft > Loam

^

{j-l\d6.U)

G-6, G-8(70.5, 80.2;

G-5(59.0)

G -4(59.0) zone n

'

Sandy soil

Upper sand layer

B ! Strong-motion seismograph

Basement Lower sand laver

Note: Elevation (m) of strong-motion seismograph installed is shown

in ( ) .

Fig. 15 Cross section of the Nagara Dam (including the location of

strong-motion seismographs)

239
(-)Y (downstream side) —*" © A" (+)Y (upstream side)

Displacement value before

and after the earthquake

4 lcmJ

o is shown in ( )

t oad

Hair cracks in crown asphalt

Fig. 16 Location of verified hair cracks and the distribution of

horizontal displacements of dam body

December 21 after the earthquake, and the

settlement and horizontal displacement of

the embankment were arranged as shown in

Figs. 14 and 16.

Settlement of the crown of the embankment

was about 20 mm, and also a maximum

settlement of 30 mm was recognized at the

toe of the upstream slope. The horizontal

displacement occurred toward the downstream

side at both the upstream and downstream

slopes, and the maximum horizontal

displacement at the downstream slope was

about 23 mm.

The strong-motion seismographs were installed at 10 places within the dam and on natural ground on the Nagara Dam site.

Location of the strong-motion seismographs

installed is shown in Fig. 15. Among the 10

strong-motion seismographs, 7 of them were

of electromagnetic type, and the remainder

were popularizing type. Nine out of ten

seismographs were trigered during the

earthquake and the records of 27 components

were obtained.

Maximum acceleration values recorded are

shown in Table 7. At the foundation surface

(G-l) directly below the dam body, 262 gal was recorded in upstream-downstream

direction and 138 gal in dam axis

direction. At the crown (G-6), 369 gal was recorded in upstream-downstream direction,

and 354 gal in dam axis direction.

Horizontal acceleration in

upstream-downstream direction was amplified

by 1.4 times approximately.

In addition, pore water pressure gauges were installed at the Nagara Dam; bore hole type

at 15 places and placing during construction

stage type at 41 places. Earth pressure

cells were also installed at 6 places.

According to the pore water pressure gauge

installed near EL-46 m, a maximum rise of

about 0.05 kgf/cm^ in pore water pressure

was recognized in Zone III.

At the upstream side of the embankment

adjacent to Zone III, a counter -weight

comprising loam and Narita sand was constructed up to EL = 59 m. This

counter -weight was not fully compacted, and

thus a possibility of a rise in pore water

pressure by the action of earthquake motion

acted to the inside of counter -weight was considered. This possibility will be

roughly examined hereinafter.

The counter-weight about 20 m thick can be

presumed to have almost homogeneous soils.

Therefore, the shearing strain y (z, t) due

to the propagation of SH wave at the point

with the depth Z in the soil layer can be

expressed as shown below by using the

velocity v (t) at the ground surface (Z = 0) .

Y (z, t) = u (t + z/Vs)

- u (t - z/Vs) /(2-Vs) (2)

where, Vs is the shear wave velocity in the

counter -weight.

Since an accurate value of Vs is unknown, it

240
Table 7 Maximum acceleration recorded on the strong-motion

seismographs installed at the Nagara Dam

Instrument

Code

G-l

G-l

G-l

G-2

G-2

G-2

G-3

G-3

G-3

G-4

G-4

G-4

G-5

G-5

G-5

Measur ing

direction

XY

Z

XY

Z

XY

Z

Maximum

acceleration

(gal)

262

138

86

270

208

124

353

288' 124

180

151

139

382

298

180

Instrument

Code

G-6

G-6

G-6

G-7

G-7

G-7

G-8

G-8

G-8

G-10

G-10

G-10

Measur ing

direction

X

Y

Z

X

Y

Z

X

Y

Z

XY

Z

Maximum

accelera tion

(gal)

369

354

324

262

294

177

497

386

328

281

148

111

is assumed to be Vs = 20 0 m/s based on tfte wet bulk density (Yt = 1.62 tf/m3 ) . As

the value of v(t), a time history of

velocity derived by integrating the

horizontal acceleration at G-2 was used for

calculating the shearing strain inside the

counter -weight from equation (2), and the

distribution of its maximum values in the

depth direction is indicated in Fig. 17(a).

U/ av'o 0.0 0.05 0.10

(a) Presumed distribution o£ shearing (b) Distribution of pore strains in depth direction water pressure ratios

in depth direction

: Where the records in upstream-downstream direction were used,

o: Where the records in embankment axis direction were used.

Fig. 17 Estimated maximum shearing strain

and estimated pore water pressure

inside the counter-weight fill

* Measuring direction codes:

X: Upstream-downstream

direction

Y: Dam axis direction

Z: Vertical direction

t

Here, the pore water pressure rising

characteristics in the saturated soil layer

due to the repeated action of shearing

strain is needed, however the soil testing

data is not available. Then, the following

relation^) which was derived by the

author et al for Sengenyama sand is used.

0.05)

1.076 I (y. 0.05) + 0.633 (3)

°vo " " " a where, Ya * s tne amplitude (%) of shearing

strain

.

An example of the time history of the

shearing strain which would have occurred

inside the counter-weight fill is shown in

Fig. 18. Pore water pressure ratio occurred

inside the counter-weight was determined by

using equations (2) and (3) and is shown in

Fig. 17 (b)

.

C0 > = STRAIN

0.05

-o.nn

-n. o:.

-0. 10

0.0 Z = M»m

820

0.0502

f w

i

I) .". 10

Fig. 18 An example of time history of

estimated shearing strains

(Z = -15 m)

241
From these results, the pore water pressure

ratio at the depth of about 15 m inside the

counter-weight becomes almost equal to 0.03

to 0.05. From this, it can be known that

the pore water pressure of about

0.05 kgf/cm2 can be generated.

8. CONCLUSION

The earthquake reported above gave a feeling

of relatively large earthquake motions to people in the metropolitan region. But the

magnitude of the earthquake was 6.7 and the

depth of the hypocenter was deep, at 58 km.

Therefore, the damage to structures occurred

only in local, limited areas where the

ground conditions were poor, and the life of

citizens was not greatly disturbed. Road

traffic was restricted in some sections

immediately after the earthquake but, in

general, this did not cause traffic

disturbance. However, there was a danger of

the occurrence of slope failures in certain

areas, and the residents had to evacuate to safe places. Also, roofing tiles of many

houses damaged at the same time, and a long

period of time was necessary for repairing

the roofs due to the shortage of labor and

materials

.

Places of boiiing of sand and water were distributed in a wider range compared to the

previous experience. This disclosed the

possibly of dangerous places very

susceptible to the earthquake on the new ground located near the metropolitan zone. This should be taken as a . lesson for the

future

.

Acknowledgement

In the stage of the present survey, the

author received data and materials from many

organizations and persons including the

Flood Control Section, Disaster Prevention

Section, and Disaster Measures and Surveys

Office, River Bureau, Ministry of

Construction; Road Disaster Prevention and

Measures Office, Road Bureau, Ministry of

Construction; Tone River Lower Reach

Construction Office, Kasumigaura

Construction Office, and Chiba National

Highway Construction Office, Kanto Regional

Construction Bureau; Ryugasaki Civil Works

Office, Ibaraki Prefecture; Public

Waterworks Section, Chousei and Sambu Civil

Works Office, Waterworks Bureau, Chiba

Prefecture; City of Mobara and City of

Tougane; and waterworks construction firms

in Chousei and Sanbu. Tokyo Gas Co. gave

the author the values of maximum

acceleration immediately after the

earthquake. And the Water Resources

Development Public Corporation gave the

author the records taken at the Nagara Dam.

The author would like to express the

greatest gratitude here to all the persons

concerned

.

REFERENCES

1) Sasaki, Y. : Prompt Report on the

Chiba-ken Toho-Oki Earthquake of

December 17, 1987, Civil Engineering

Journal 30-1, pp. 47-55, 1988.

2) Sasaki, Y . , Matsumoto, H . , Kuwabara, T. and Seko, I.: Damage by the Chiba-ken

Toho-Oki Earthquake, Tsuch i-to-kiso

,

Vol. 136, No. 12, pp. 5-11, 1988.

3) Earthquake and Volcano Division,

Meteorological Agency: Summary of

Earthquakes and Volcanoes in December

1987.

4) Harukawa, M. : Seismic Intensity

Distribution in Kujukuri Plain and the

Vicinities during the Chiba-ken

Toho-Oki Earthquake and the Problems

Suggested (2nd report), Proceedings of

Symposium by Kanto Branch of the Japan

Geological Society, pp. 18-24, 1988.

5) Mizoue, M. : Outline of the Chiba-ken

Toho-Oki Earthquake, Proceeding of

Symposium by Kanto Branch of Japan

Geological Society, pp. 1-8, 1988.

6) Conference for Promoting the Strong

Earthquake Observation Project: Prompt

Report on Strong-Motion Accelerograms

No. 37, published by the National

Research Center for Disaster

Prevention, Science and Technology

Agency, 1988.

7) Osawa, R. : Special Issue, Earthquake

Measures by Enterprises, Part 1,

Earthquake Measures by Tokyo Gas Co.,

News from Earthquake Engineering

Promotion Society, No. 93, 1987.

8) Kawashima, K . , Aizawa, K. and

Takahashi, K. : Method of Estimating

Maximum Earthquake Motion and

Earthquake Response Spjectrum (Part 4),

Re-examination of the distance

attenuation formulas for maximum

acceleration, velocity, displacement

and response spectrum, PWRI Technical

Memorandum No. 199 3, 1983.

9) Yoshimi, S., Sasaki, Y. and Kuwabara,

T. : Riview of the Principal Axis of

Three-dimensional Earthquake Motions in

Strong Earthquake Records, Proceedings,

43rd Annual Meeting of the JSCE, Part

1, pp. 936-937, 1988.

10) Kusuda, T. et al . : Liquefaction

Phenomena Observed in the Chiba-ken

Toho-Oki Earthquake, Proceedings,

Symposium by Kanto Branch, Japan

Geological Society, pp. 31-34, 1988.

11) Yasuda, S., Touno, I. and Morimoto,

I.: Survey Report of the Chiba-ken

Toho-Oki Earthquake, Chishitsu & Chousa, 1988.

12) Kuribayashi, E. , Tatsuoka, F. and

Yoshida, S.: Ground Liquefaction

History after the Meiji Era in Japan,

PWRI Technical Note No. 30, 1974.

13) Sasaki, Y.: Lessons from the Damages

Caused by the Nihonkai-chubu Earthquake

(1983) - Lessons from the damage to the

242
levees of the Hachirougata Lagoon - Kisokou, pp. 77-84, May 1987.

14) Kanto Central Committee, Japan Gas

Association: the Chiba-ken Toho-Oki

Earthquake and City Gas, 1988.

15) Geographical Survey Institute, Public

Works Research Institute and Building

Research Institute, Ministry of

Construction, National Land Development

Technology Center: Report on the

Development of Disaster Information

System, pp. 176-190, 1988.

16) Kimoto, E. , Yamazaki, A., Fukushima,

R. , Seko, I. and Tanaka, H.: Situation

of the Nagara Dam of Bosou Water

Conveyance Channel during the Chibaken

Toho-Oki Earthquake, Tsuchi-to-Mizu,

pp. 89-95, 1988.

17) Sasaki, Y. and Masui, H.: Calculations

of the Generated Amount of Pore Water

Pressure based on Cumulative Shearing

Strains, Proc. 43rd Annual Meeting,

JSCE, Part 3, pp. 448-449, 1988

243

Theme III

Storm Surge and Tsunami

Observations of Tsunamis on the West Coast of the

United States

by

James F. Lander1

ABSTRACT

Although the recorded history extends back 150

to 200 years for the United States West Coast

the effect of tsunamis there has been studied

little. Major harbors of Seattle and San

Franciso are well protected by narrow entrances

and the southern coast of California is

protected by a broad continental slope and

offshore islands. Fourteen tsunamis with

amplitudes of 1 meter or more or which were

destructive have been reported. However, until

1960 none caused more than a few thousand

dollars in damage. The effects of the 1946

Aleutian tsunami were stronger near Santa Cruz,

California while the Chilean tsunami of 1960

caused about $1,000,000 in damage to southern

California. The 1964 Gulf of Alaska tsunami was

by far the most severe causing $13,000,000 in

damage and 16 fatalities, mostly in Crescent

City, California. These different areas of

principal effects probably relate to the

different orientations of the source of the

wave. The 1812 Santa Barbara, California

tsunami, 1960 Chilean tsunami and 1964 Alaskan

tsunami are considered as the design tsunamis.

Future tsunamis from the Shumagin gap, in

Alaska, and Peru-Chile gap are potentially

damaging. Population growth, coastal and harbor

development, increased recreational use of

beaches, and increased number of small fishing

and recreational craft are increasing the risk

of damage from future tsunamis.

Keywords: Tsunami; U.S. West Coast; Tsunami

Hazard

1. INTRODUCTION

Permanent settlement of the West Coast by

Europeans began in 1769 with the establishment

of the Spanish mission at San Diego, and by

American and British traders in Washington and

Oregon territories in the early 19th century.

Thus, a period of 150 to 200 years or less in some areas is available for the historical study

of tsunamis. Recording tide gages were put into

permanent operation in San Francisco, and San

Diego, California and Astoria, Oregon, in 1854.

These gages were probably the earliest in

operation in the Pacific. In 1854 these gages

recorded unusual oscillations which were

identified as sea waves from the Tokaido and

Nankaido, Japan, earthquakes. These may have

been the first recorded tsunamis identified with

earthquakes, and allowed the scientists of the

time to calculate the average depth of the ocean between Japan and California. Despite this

relatively long history, the study of West Coast

tsunamis has received little attention.

Tsunami warnings for the West Coast are provided

by the regional facility at Palmer, Alaska. Due

to the response time they cannot provide

warnings for locally generated tsunamis on the

West Coast. Warnings for remote sourced

tsunamis are sent to each state government and

military commands and relayed to county

officials. Response to the 1964 warnings was mixed. Some officials succeeded in making

evacuations and others did little. In fact, in

response to the warning, people crowded the

wharfs in San Francisco and Los Angeles to see

the 1964 tsunami arrive.

2. TECTONIC SETTING

The West Coast owes most of its good fortune

with respect to the tsunami hazard to geography

and tectonics. The major population centers of

Seattle and the San Francisco Bay are well

protected by large bays with restricted

entrances. The coasts of northern California,

Washington and Oregon are .rugged and exposed to

the sea. The communities there are small,

seldom more than a few thousand people.

Southern California is protected by a broader

continental slope and a chain of offshore

islands.

Unlike most of the Pacific coasts, the West

Coast of the United States does not have

typical subduction zone tectonics. The

principal earthquake faults including the famous

San Andreas are strike-slip without notable

vertical offset and occur mainly on land. The

1906 San Francisco earthquake did not produce a tsunami even though portions of the break were under water. Offshore areas in southern

California do have normal and reverse faulting

University of Colorado, CIRES, Boulder,

Colorado, 80309.

247
and have caused small local tsunamis four or

five times in historical times. Figure 1 shows

the main tectonic features relative to tsunamis

on the West Coast.

3. LARGE AND DAMAGING TSUNAMIS OBSERVED ON WEST

COAST

Fourteen tsunamis have been observed on the West

Coast with amplitudes of one meter or more.

Nine have caused some damage including one

tsunami with an amplitude of less than one meter (see Table 1). Remote sourced tsunamis

from Japan, Kamchatka, the Aleutian Islands,

Gulf of Alaska, Hawaii, and Chile have all

caused at least minor damage; Local tsunamis

have been reported observed four or five times

but some of the earlier reports are

questionable. The 1812 tsunami was once

believed to have had a 15 m run up but now is

believed to have had only a 3.5 m wave. The

1927 California tsunami was the best documented

local tsunami. Prior to 1960 no tsunami caused

more than a few thousand dollars of damage. The

most destructive tsunamis were the 1960 tsunami

from Chile and, the 1964 tsunami from the Prince

William Sound, Alaska which was the most

important. Even these two events which caused

about $1,000,000 and $13,000,000 in damage

respectively on the West Coast were overshadowed

by the much more significant effects elsewhere

which received most of the scientific attention.

Due to the orientation of the generating zones these are thought to be the most hazardous

sources possible for tsunamis and are the design

tsunamis for remote sourced tsunamis for the

West Coast.

4. COMPARISON OF THE 1946. 1960 AND 1964

TSUNAMIS

Figure 2 shows the amplitudes recorded or observed on the West Coast of the United States

for the 1946, 1960 and 1964 tsunamis. The 1946

Aleutian tsunami was devastating to Hawaii and

resulted in the establishment of the Pacific

Tsunami Warning Center. On the West Coast it was observed widely but damage was limited to

the area north and south of the San Francisco

Bay. A maximum amplitude of 2.7 meters was

observed at Noyo, California, where 100 fishing

vessels were washed ashore. At Santa Cruz,

California, one man was drowned and houses and

sheds flooded.

The 1960 Chile tsunami's greatest effects were

in the Long Beach/Los Angeles, California, area. There was one fatality, and $500,000 to

$1,000,000 damage there, mostly to small craft

and floating boat slips. It caused minor damage

elsewhere including $30,000 to two boats at

Crescent City, California.

The 1964 tsunami most strongly affected the

Washington, Oregon, and northern California

coasts with most of the damage (about

$8,000,000) at Crescent City. Another

$1,000,000 damage was doie to small craft at San

Rafael and Sausalito, California, marinas well

inside the San Francisco Bay. Abundant loose

logs from the lumbering industry contributed to

the damage on the coasts of Washington, and

Oregon. The waves tended to follow stream and

rivers concentrating damage there. Figure 3

shows the effect at Seaside, Oregon. The

effects were notably less in southern

California. A total of 16 fatalities and about

$13,000,000 in damage occurred. One fatality

occurred the following day at Klamath River,

California, due to continued high wave activity

in resonance with the continental shelf.

5. RISK FACTORS

The local tsunamis are small and limited in

effect. The length of expected fault rupture is

believed to limit the maximum wave heights to 4

to 6 m, and the range of coastline affected to

about 100 km. Also, there are no large rivers

to create submarine deltas and deposits or steep

sided trenches to give rise to a tsunami due to

a submarine landslide. The history of local

tsunamis may be too short to be fully confident

that the worst has been observed. This is

particularly true for the area off the coasts of

Washington and Oregon where there is evidence of

subduction of the Juan de Fuca Plate but no history of earthquakes or tsunamis from the area

of subduction.

Seismic gaps in the Shumagin, Alaska, area and

in the Peru-Chile area are potential source

areas for remote sourced tsunamis damaging to

the West Coast in the future and may have

effects similar to the 1960 and 1964 tsunamis.

A particular hazard to the northern part of

coast is damage caused by logs which litter the

beaches. These act as battering rams and also

end up as substantial debris on land. In the

southern area much of the damage is caused to

small craft insecurely berthed in floating

slips. Currents set up by the tsunamis can dislodge the boats which became additional

floating projectiles.

The tsunami hazard is growing due to changing

demographics including: population growth, the

increased use of beaches for recreation,

increased numbers of small fishing and pleasure

craft, and increased development of harbor

areas.

6. CONCLUSIONS

1. The West Coast probably has experienced its

design tsunamis in the 1812, 1960, and 1964

events with the 1954 event being the most

destructive.

2. The West Coast is relatively protected from

tsunamis due to protected coasts and harbors.

There is apparently only a minor local tsunami

risk.

248
3. A significant current hazard probably exists

from remote-sourced tsunamis from seismic gap

areas in Shumagin, Alaska, and South America.

4. The tsunami hazard is growing due to

changing demographics with increased coastal

development, recreation use of beaches,

population growth, and numbers of small craft.

5. Logs on the beaches of the Northwest and

small craft in floating slips add to the hazard.

6. Several localities such as Crescent City,

Half Moon Bay, Santa Cruz, and Avila, California

report larger amplitude waves than other

communities which may relate to the shape of

their harbors or to their exposure to the open

ocean. Tsunamis follow rivers inland affecting

communities which also cluster near the river

mouths.

7. Variations in the tsunami height reported at

locations along the coast and for tsunamis are probably related to the orientations of the

tsunami source.

7. REFERENCES

1. Agnew, Duncan Carr, "Tsunami History of

San Diego," Earthquakes and Other Perils,

San Diego Region, San Diego Association of

Geologists, 1979, p. 117-138.

2. Bache, A.D., Report of the Superintendent

of the United States Coast Survey for

1855, Washington, D.C., 1856, p. 96, and

appendix, 50 p. 342-346.

3. Bancroft, Hubert H., History of

California, Vol. 4, 1886, p. 78.

4. Berkman, S.C. and J.M. Symons, The Tsunami

of May 22, 1960 as Recorded at Tide

Stations, United States Department of

Commerce, Coast and Geodetic Survey,

Washington, D.C., 1964, 79 p.

5. Byerly, Perry, "The California Earthquake

of November 4, 1927," Bulletin of the

Seismological Society of America, Vol. 20,

No. 2, June 1930, p. 53-66.

6. Green, C.K. "Seismic Sea Wave of April 1,

1946, as Recorded on Tide Gages,"

Transactions of the American Geophysical

Union, Vol. 27, No. 4, August 1946, p.

490-500.

7. Grouginis, V.J., J.W. Joy, and R.R. Putz,

The Reported California Tsunami of

December, 1812, unpublished manuscript,

120 p.

8. Heaton, Thomas H. and Stephen H. Hartzell,

"Earthquake Hazards on the Cascadia

Subduction Zone," Science, American

Association for the Advancement of

Science, Washington, D.C., 1987, Vol. 236,

p. 162-168.

9. Heaton, Thomas H. and Parke D. Snavely,

Jr., "Possible Tsunami Along the

Northwestern Coast of the United States

Inferred from Indian Traditions," Bulletin

of the Seismological Society of America,

Vol. 75, No. 5, 1985, p. 1455-1460.

10. Hogan, O.W., W. W. Whippe, and C. Lundy,

Tsunamis of 27 and 28 March, 1964, State

of Washington Coastline, United States

Army Corps of Engineers (Seattle,

Washington, unpublished file report, 1964

.29 p.)

11. Joy, J.W., "Tsunamis and their Occurrence

along the San Diego County Coast",

Westinghouse Ocean Research Laboratory

Report, No. 68-567- 0CEAN-RL, July 15,

1968, San Diego, California.

12. Larsen, L.H., "Tsunamis Hazard Along

Washington's Coast," Natural Hazards in

Washington's Coastal Zone, Washington

State Department of Emergency Services,

Olympia, Washington, 1979, p. 35-38.

13. Magoon, O.T., Structural Damage by

Tsunamis, American Society of Civil

Engineers Specialty Conference on Coastal

Engineering, 1965 Proceedings, Santa

Barbara, California, 1965, p. 35-38.

14. Marine Advisers, Inc., Examination of

Tsunami Potential at the San Onofre

Nuclear Generating Station, Report A-163,

Los Angeles, California, September 1965,

59+ p.

15. McCulloch, D.S., "Evaluating Tsunami

Potential," in Evaluating Earthquake

Hazards in the Los Angeles Region, United

States Geological Survey Professional

Paper 1360, United States Geological

Survey, Reston, Virginia, 1985, p.

375-413.

16. National Academy of Sciences Committee on the Alaska Earthquake, The Great Alaska

Earthquake of 1964, Oceanography and

Coastal Engineering, National Academy of

Sciences, Washington, D.C., 1972, 556 p. 17. O'Brien, M. P., Preliminary Report on Seismic Sea Waves from Aleutian Earthquake

of April 1, 1946, Technical Report Wave

Project Fluid Mechanics Laboratory,

University of California at Berkeley:

April 25, 1946.

18. Roberts, James A. and Chen-Wu Chien, "The

Effects of Bottom Topography on the

Refraction of the Tsunami of 27-28 March

1964: The Crescent City Case," Ocean

Science and Ocean Engineering 1965--Joint

Conference Marine Technology Society and

American Society Limnology and

Oceanography, Washington, D.C., Trans.,

Marine Technology Society, Washington,

D.C., Vol. 2, 1964, p. 707-716.

19. Salsman, Garrett S., The Tsunami of March

9, 1957 as Recorded at Tide Stations,

Technical Bulletin No. 6, United States

Coast and Geodetic Survey, July 1959, 18

P- 20. Soloviev, S.L. and Ch.N. Go, A Catalogue

of Tsunamis of the Eastern Shore of the

Pacific Ocean, Academy of Sciences of the

USSR, "Nauka" Publishing House, Moscow,

1975, 204 p. (translation by Canada

Institute for Scientific and Technical

Information, National Research Council,

Ottawa, Ontario, Canada K1A 0S2, 285 p.)

249
21. Spaeth, M.G. and S.C. Berkman, Tsunami

of March 28, 1964, as Recorded at Tide

Stations and the Seismic Sea Wave Warning

System, The Great Alaska Earthquake of

1964, Oceanography and Coastal

Engineering, National Academy of Sciences,

1972, p. 38-110.

22. Symons, J.M. and B.D. Zetler, The Tsunami

ii of May 22, 1960 as Recorded at Tide

Stations, Preliminary Report, United

States Department of' Commerce, Coast and

Geodetic Survey, Washington D.C., 1960, 29

P- 23. Wiegel, R.L., "Tsunamis," Earthquake

Engineering, Prentice-Hall, Englewood,

Cliffs, New Jersey, 1970, p. 253-306.

24. Wilson, Basil W. and Alf Trum, "Runup

Heights of the Major Tsunami on North

American Coasts" in The Great Alaskan

Earthquake of 1964; Oceanography and

Coastal Engineering, National Academy of

Sciences, Washington, D.C., 1972, p.

138-180.

25. Zerbe, W.B., Tsunami of November 4, 1952, as Recorded at Tide Stations, United

States Department of Commerce, Washington,

D.C., 1953, 62 p.

250
Fig. 1. Schematic of major tectonic features

relating to tsunamis on the west coast

251
Table 1.

Date

Tsunamis Affecting the West Coast of the U. S.

Damaging or at Least 1 Meter Amplitude

Amplitude (M) Locality Most Affected Source

1812

1859

1862

1868 (Aug.)

1868 (Oct.)

1877

1878

1896

1927

1946

1952

1957

1960

1964

1975

3.5

4.6

1.2

1.8

4-5

3.6

2.0

1.5

1.8

2.7

1.4

1.0

1.7

4.3

0.4

Gaviota (S. Calif.

)

Negligible Damage

Half Moon Bay (C. Calif.)

Schooner Damaged

San Diego (S. Calif.

)

San Pedro (S. Calif.

San Francisco (C. Calif.)

Gaviota (S. Calif.)

Wilmington (S. Calif.)

Unspecified Damage

Santa Cruz (C. Calif.)

Dike Destroyed, Ship Damaged

Surf and Avila (S. Calif.)

Santa Cruz (S. Calif.)

Some Damage, 1 Killed

Crescent City & Avila

(N. & S. Calif.)

San Diego (S. Calif.

)

Minor Damage

Crescent City & Los Angeles

(N. & S. Calif.)

$1,000,000 in Damage, 1 Killed

Crescent City (N. Calif.)

$12,000,000 in Damage, 16 Killed

Catalina Island (S. Calif.)

$2,000 in Damage

S. Calif.

Unknown

S. Calif.

Chile

N. Calif.

Peru-Chile

Chile?

Japan

S. Calif.

Aleutian Is.

Kamchatka

Aleutians

Chile

Gulf of Alaska

Hawaii

252
CO

LUm2

UJQDH

-J

CL<

o>

Q. <Owcw©

.c

(0o

"3

55o&

a

—

o

§1

O)00 a -5 8 1

llo <o

a>.c icCD —

(0 . — o

cS<S

a> Mill . j&i — to E T3 is

•I

TO© IS

253
Figure 3. Location of major tsunami damage in Seaside, Oregon (from Spaeth and Berkman, 1972).

254
Present State of Tsunami Numerical Simulation

by

Chiaki Goto' 1 and Katsutoshi Tanimoto" 2

SUMMARY

Present state of tsunami numerical simulation

which is generally carried out to predict

tsunami hazard in Japan is briefly reviewed.

As recent knowledge, simulations for

transoceanic propagation and soliton fission

of tsunamis and behavior of spread materials

due to tsunami are also introduced. Finally,

through a knowledge of numerical error,

remarks for computation are explaned.

KEY WORDS : Tsunami, Distant Tsunami. Soliton

Fission, Behavior of Spread Materials.

Numerical Computation, Numerical Error.

L INTRODUCTION

Located in northwestern part of the Pacific

Ocean seismic belt, Japan has been facing to tsunami hazard and is widely known as the

place where tsnamis often attack. Especially,

the Sanriku Coast in northeastern coast of

main island of Japan experienced crushing

calamities caused by the Great Sanriku

Tsunamis occured in 1896 and in 1933. Due to the two tsunamis, more than 30.000 people

were killed and about 10.000 houses were

lost. Thus, the Japanese term "tsunami" had

a chance to become the international word.

For the past twenty years, many tsunami walls

and breakwaters have been constructed along

Japanese coast line as the tsunami protection

works. However most of them Were constructed

for the purpose tof the pretection against

middle class of tsunamis and a few structures

were designed against huge tsunamis. For coun termeasures against huge tsunamis, the

total protection plan with the construction

of defence structures, the management of

hazardous area and tsunami warning system

should be considered. From the reason and

the limited data avilable on tsunamis, many

research attempts for the prediction of

tsunami hazard have been made by numerical

simulat ion.

2. NUMERICAL COMPUTATION OF TSUNAMI

2.1 Computation in deep sea region

The finite difference and the finite element

methods have been developed for reproducing

the tsunami propagation based on the long

wave theories. Here, a numerical simulation

of tsunamis by using the finite difference

method which is usually carried out to predict tsunami hazard is introduced. The

computation of tsunamis is generally carried

out in two separate parts: deep sea and

shallow sea area, because of a limit of

computer memory capacity and computation

time. In deep sea. the computation area is covered with coarse grid and a coast line is

approximated as a vertical wall. However the

grid length might be varied dependent on the

length of tsunami source, general grid length

is set so as to be about 5 km long and become

gradually smaller approaching a shallow sea. Figure 1 shows a computation area on Sanriku

Coast located in northern Japan 1 '. Computation areas are divided four regions,

region A to region D, the grid lengths are

5.4 km. 1.8 km. 0.6 km and 0. 2 km

respectively. In this figure, tsunami sources reported by Hatori 2 ' are also presented. The

numerals inidicate the occarence year of

t sunam i s.

Generally, the Governing equations in deep

_l 1 I 1 1—I 37 Fig. 1 Computation region

Senior research engineer. Port and Harbour

Research Institute. Ministory of Transport

" 2 Head of Hydraulic Engineering Division.

Port and Harbour Research Institute,

Ministory of Transport

255
sea are the linear long wave theory as

f ol lows,

t = 0 rain. HOKKAIDO

ataj

at

a_N at

+ gh

a_M a_N ax

+ ay

In

dxaa

= o

+ gh^ = 0

ay

= 0

(1)

(2)

(3)

where

(x.y)

t

g

hn

(M. N)

spatial coordinates,

time.

gravity acceleration,

still water depth,

tsunami height.

tsunami discharge flux in the

(x, y) di rect ion.

There are some cases which takes terns of

the Coriolis and the bottom friction into

consideration. The effects of the terns,

however, are usually small because tsunamis

attacking Japan islands travel only short

distance in deep sea.

The initial tsunami profile is assumed to be

the same as crustal movement calculated on a fault model. The seabed deformation can be

approximated as instantaneous change and the

initial water surface can be also asuumed to have the same profile as the defornated sea bottom by earthquake.

The computation for the deep sea. mentioned

above, are primarily performed with the

purpose of calculating the input condition

for the computation of shallow sea areas.

In addition, the reliability of the

computation results should be considered

only for the sea of deeper than 200 n.

Figure 2 shows a result of computation in

deep sea for Great Meiji Sanriku Tsunami

occured in 1896. 11 The vertical scales of

the figures shown are exaggerated compared to the horizontal ones. The maximum tsunami

height at t = 0 s is 4m.

2.2 Computation in shallow sea region

The computation for a shallow sea is carried

out for the purpose of predicting tsunami

behavior in the vicinity of the coast and on

the land. Generally the grid size is less

than 50 m. The governing equations used for

the computation is the shallow water theory,

which takes sea bottom friction into

consideration, as follows.

aM3T

+ $ = 0 ( (4)

TOHOKU

t = 1 2 min.

Fig. 2 Computation result for Great Meiji

Sanriku Tsunami ( in deep sea).

+ §t£? MvTT-F- = 0 (5)

+ NvHP-TTF- = o (6)

where

d : total water depth, d = h + n.

n : Manning's roughness coefficient 31

.

As the offshore boundary condition in shallow

sea region, the results of the computation in deep regions is given. Homma' s weir

formula 41,51 is usually applied to the

estimation of overtopping rate or over flow

256
(a) Computation result with both

effects of run-ups on land and

coastal structures.

(b) With effct of run-up on land < c > Without both effects,

and without effect of coastal

structures.

Fig. 3 Computation results for the 1896 Great Meiji Sanriku Tsunami (in shallow sea area).

MEASURED RECORDS

ASAUCHI

CALCULATED RESULTS

FISHING PORT

YONESHIRO RIVER

Fig. 4 Computation results for

rate on sea walls and breakwaters. As the

boundary condition at the tsunami front on

land, there are two methods. One is a method

that assumed a weir formula 6 ' and the other

is a method that a continuous topography is

approximated by series of discontinuous

horizontal steps 7 '. Because the reliability

of both methods are not clear from a physical

point view, we have to use the fine grid near

shore 1 i ne.

An example of computational results is shown

in Fig. 3

.

1

' The computation is carried out

for the Great Meiji Sanriku Tsunami at Miyako

bay. The computatioal results at 40 minutes

after tsunami occurence are presented for the

different conditions: i.e. with and without

tsunami run-up and effects of structures.

the 1983 Nihonkai-Chubu Tsunami.

The result which is taken both run-up on land

and structures into account is apparently

different with tsunami velocity distribution

from those without consideration.

Fugure 4 8 ' shows a result of computation of

the 1983 tsunami occured at Japan Sea. In this figure, tsunami run-up heights and

inundation area are compared with records.

Since the grid size of the computation is 30 m long and is considered to be fine

enough, it seems that the results is quite

reliable. Aida 9 ' ,lB) proposed parameters

denoted K and k to evaluate the accuracy of

numerical result as follows;

log K = (1/N) log K . (7)

and

257
120-E UO'i )60*E IBO'E I60*W 140*V 120'W 100'W BO'W

(a) Water height distribution at 15 hours after tsunami occurence

(b) Tsunami travel-tine curve

Fig. 5 Computation results of the 1960

Chilean Tsunami.

K = J (log K;)VH - dog KK (8)

where

N : numbers of data.

Ki : ratio of record to numerical result.

The values of K and K is 0.99 and 1.28

respectively in this computation.

3. RECENT DEVELOPMENTS IN TSUNAMI COMPUTATION

3. 1 Distant tsunami

Numerical computation of the transoceanic

propagation of tsunamis generated on the

Chilean coast to Japan can be carrired out by use of a super computer.

Figures 5 and 6 11,,12> show numerical results

of the 1960 Chilean Tsunami. The governing

MAXIMUM HEIGHT () RUN-UP HEIGHT (m)

0 10 20

—— SHALLOW WATER THEORY

LINEAR LONG WAVE THEORY

1 OBSERVED RECORDS

Fig. 6 Computation results of the 1960

Chilean Tsunami ( tsunami height

along Sanriku Coast).

equations of the computation are the linear

Boussinesq equations with Coriolis terms

described in the latitude-longitude

coordinate system as follows :

J?

+ Kcktih^v + % i = 0 (9)

$ +

= -« + l§x^3p

l < 10 > $ ireSx ft ' « + lei^ < ll >

where

< X. 4>> : latitude and longitude

coord ina tes.

f : Colioris factor.

(M.N) : tsunami discharge flux in the

(X. i|)) d i rec t i on.

258
In the computation, grid length is set for

1/6 degree, which corresponds to 20 km along

the equator.

Figure 5 shows distribution of tsunami height

at 15 hours after tsunami occurence and

travel-time curves drawn for every 1 hour.

The difference between tsunami heights

simulated with the linear long wave theory

and those simulated with shallow water

theory are exhabited along the Sanriku coast

in Fig. 6. Since the linear long wave theory

does not included sea bottom friction terms,

the resulting tsunami heights are large

compared with observed data. On the other

hand, the result with the shallow water

theory including wave nonlinearity and bottom

friction shows good agreement with observed

data. The CPU time to reproduce 30 real

hours is 30 minutes with the aid of super

computer NEC SX-1.

3.2 Soliton Fission of Tsunami

The Nihonkai-Chubu earthquake which occured

in the middle part of Japan Sea in 1983

generated a great tsunami. Many accurate

data of run-up heights, inundation areas and

tide gauge records were obtained and

collected. In addition, soliton fissions as

a result of the interaction between the

nonlinear and the dispersion effects during

the tsunami propagation was observed and

recorded at the front of tsunami.

A tsunami formed in a solitary wave train is

quite dangerous to protection works because

of its rapid growth in height. Detail of the

soliton fission of tsunamis should be

analized by numerical computation.

Figure 7 13 ' shows an example of numerical

computation for evolution of solitons. The

computation is carried out by the nonlinear

dispersive long wave theory with bottom

friction terms 14 '. The result has small

numerical oscillations becase coarse grid

size is adopted for the limitation of

computer memory. Therefore: the simulation

does not seem to be carried out relevantly.

However we can recognize in the figure that

run-up front of the first and the second wave

have growing to solitary wave train. As described above, the numerical computation

for soliton fission of tsunamis has been

developed recently, but there are so many

problems related to soliton behavior left unso 1 ved.

3.3 Spread Materials Due to Tsunamis

As for the behavior of materials caused

by a tsunami, only a few studies have been

3. 40ka

(ml

Fig. 7 Computation result for tsunamis

with sol iton f issions.

made. They are limited to the numerical

simulation for the spread of timber and oil.

As for the timber spread, it is reported that

about 4.000 timbers were floated away from

Miyako port on North Sanriku coast when the

Tokachi-oki tsunami hit in 1968. Once

floated and washed away by tsunamis, timbers

often become destructive forces to houses and

port facilities.

The simulation model of the timber spread is

developed by Goto 15 '. The motion of timber

consists of two parts, a avereged motion and

diffusion. The averaged motion is calculated

for the mean current induced by tsunami, on combining the drag and added mass effects of

timbers. The diffusion is estimated by a assumption that spread of timbers around

their center of gravity follows a random

process.

Figure 8 shows a result of practical

application for numerical simulation of

timber spread. It is assumed that 6.000

timbers are being stored and are washed away

by tsunami. In the figure, black points

indicate location of timbers at each time.

For oil spread, it is reported that oil tank

in Niigata city faced on the Japan Sea was damaged by the 1964 earthquake and oil was discharged and spread over the water.

259
Hei River

Ryugas«k<

Breakwater

t'i/Omin. t'lOOmin. t'UOmin. t«120min.

fig. 9 Computation results for oil spreading.

Unfortunately, five houres after the

earthquake the oil caught on fire and fire

continued 15 days.

The spread of oil by tsunami was first

computed by Goto 16). The method without the

capillarity can be applied to the gravityinertia and gravity-viscous regimes. In a large oil spill, the surface tension becomes

important after several days, while a tsunami

may continue for several hours. Therefore,

the method is sufficient to predict the oil

transported by tsunami.

Figure 9 shows an example of the results.

There are two tanks at points A and B. Each

of tanks contains 1,000 m

3 of oil and is surrounded by dikes of 40 cm high. It is

assumed that the tanks are ruptured by an

earthquake and leaked oil is stored inside

the dikes. When the tsunami water level

exceeds the surrounded dikes, oil begins to be floated and transported.

4. NUMERICAL ERROR AND REMARKS FOR

COMPUTATION

4.1 Classification of Numerical Error

Once the initial tsunami profile is determined exactly, the accuracy of tsunami

numerical simulation is closely related with

the governing equations and the grid length

used. Especially, aliasing and truncation

errors related with grid length is very

important because the errors appear as the

artificial dispersion or dissi pat ion effects

and causes the phase shift or the damping of

260
tsunami height 17 '- 181

.

4.2 Governing Equations

Usually, the long wave theory is used as a model representing the propagation of a tsunami. The theory is derived by the

assumption which vertical acceleration of

water particle is small as compared with the

acceleration of gravity and the relative

depth is small. The equation is described

the motion of long waves as follows :

(local term ) + (pressure term)

+(nonlinear term) + (dispersion term) = 0.

If the dominant terms of the equation

consists in the first two terms, it is called

the linear long wave theory, if three terms,

the shallow water theory, and if all terms,

the nonlinear dispersive long wave theory.

For a tsunami occurred in the sea near Japan,

the dominant terms of the equation can be

considered as only a local and a pressure

terns. And the nonlinear term is kept small

while the tsunami is traveling in deep seas

and the dispersion term is known as small in

both areas 19 '. For on the reason, the linear

long wave theory is used in deep sea, whereas

the shallow water theory is used in shallow

seas.

In conventional simulations, there are some cases used by the linear long wave theory as governing equations even if in shallow seas. However the accurate waye celerity is faster

than ones by the linear long wave theory and

is described by the equation as follows :

C = f~gb[ 1+ (3/2) (H/h) - <2jih/L) a /6 + •••]

(12)

Where

C : tsunami celerity.

H : tsunami height.

L : tsunami length.

Therefore, we have to pay attention to the

difference of celerities which causes the

phase sift and gives serious influences to maximum tsunami height and velocity

distribut ion.

4. 3 Grid Length

Aliasing and truncater errors of the

numerical computation are very important,

which causes calculated tsunami height to be

smaller. In order to keep error small, grid

length Ax should be proposed to satisfy the

following equation.

CbT/Ax > 30 (13)

where

Cb : wave celerity of the liner long wave

theory.

T : period of tsunami.

Wave celerity is proportional to the square

root of sea depth, therefore, the computed

grid length should be made gradually smaller

as the tsunami approachs shallower.

To obtain more accurate results, we can

use of the knoeledge of the artificial

dispersion effects introduced by numerical

scheme because the effects have the similar

to the physical correct dispersion. The

condition, that long wave theory without

dispersion terms can give the same wave profile computed with the theory with

dispersion effects, requires the relation as following :

(Ax/hW l-UAt/Ax)' = 0.5 (U)

Therefore, the grid length of numerical

simulation should be determied so that the

relation Eqs. (13) and (14) are satisfied.

Even if there are no breaking waves, the

boundary condition at the wave front on land

has an important influence on the accuracy of

the numerical results. Goto and Shuto 19 ' proposed a condition related with grid size

which are successfully applied to the

problem, as following

Ax/sgT < lQ

-4

(15)

where

s : bottom slope from the shore line to the point of 100m deep.

5. CONCLUSION

In the paper, numerical simulation method for

tsunami propagation and run-ups is briefly

reviewed. Recently, tsunami numerical

simulation for run-up on land and transoceanic propagation been developed and

appried to practical problems. But. even

now. the accurate velocity and wave force

distributions necessary to calculate

satabilities of coastal structures are quite

difficult and there are so many problems

which remain unsolved.

REFFERENCES

1) Shuto. N. . C. GOTO and F. Imamura : Use

of numerical simulation as a means of

tsunami warning, Proc. of JSCE. Vol.393,

pp. 18 1-189. 1 98 5 ( i n Japanese).

2) Hatori. T. : Tsunami sources off

northeastern coast of the Pacific Ocean

in Japan. Earthquake. Vol. 27, 1 9 74 (in

261
J apanese)

.

3) Houston. J.R. and H. L. Butler : A numerical model for tsunami inundation,

U.S. Army Eng. W. E. S. . Tech. Rept. Hl-79-

2. 5 4 p. 197 9.

4) Goto. C. and N. Shuto : Effects of large

obctacles on tsunami inundation, Terra

Scientific Publishing Company, pp. 551-

525. 1983.

5) Aida. I.. H. Tsubokawa and M. kawaguchi : Numerical experiment on behavior of

tsunamis exceeding the design height of a sea wall - Case studies for Matsuzaki and

Taro -, Earthquake. Vol.41, pp. 343-350.

1988(in Japanese).

6) Aida, I : Numerical experiments of

inundation of tsunamis. Bull. Earthq.

Res. Inst.. Vol.53, pp. 4 41-446. 1 9 7 7 < i n

J apanese)

.

7) Iwasaki, T and A. Mano : Two-dimensional

numerical computation of tsunami run-ups

in the Eulerian description, Proc. of

26th Conf. on Coastal Eng.. JSCE. pp. 70-

74. 1 97 9 (in Japanese).

8) Uasegawa, K. and K. Inagaki : Numerical

simulation of the Nihonkai-Chubu Tsunami.

Unic Cop., 1984(in Japanese).

9) Aida. I. : Tsunami source models of the

1983 Nihonkai-Chubu Tsunami, Bull.

Earthq. Res. Inst.. Vol.59, pp. 93-104,

1984(in Japanese).

10) Aida, I. : Numerical simulation of old

tsunam is originated off the Sanriku

coast. Bull. Earthq. Res. Inst.. Vol.56,

pp. 7 1-10 1. 1 9 7 7 ( i n Japanese).

11) Imamura, F. . 0. Nagano. C. Goto and N.

'

Shuto : Numerical simulation of the

transoceanic propagation of the 1960

Chilean Tsunami, Proc. of 34th Conf. on Coastal Eng.. JSCE. pp. 1 72-17 6. 1987 ( in

Japanese).

12) Goto, C. . F. Imamura and N. Shuto : Study

on numerical simulation of the transoceanic propagation of tsunami Part 1: Governing equation and mesh length.

Earthquake. Vol.41, pp. 5 15-52 6. 1 9 88 ( in Japanese).

13) Hasegawa, K. and K. Inagaki : Numerical

simulation of the Nihonkai-Chubu Tsunami

(Part 2). Unic Cop., 1 98 7 ( i n Japanese).

14) Goto. C. : Equations of nonlinear

dispersive long wave for a large Ursell

number. Proc. of JSCE. Vol.351, pp. 193-

202, 1984(in Japanese).

15) Goto. C. : A simulation model of timber

spread due to tsunamis, Proc. of 30th

Conf. on Coastal Eng., JSCE. pp. 594-597.

1 98 4 ( in Japanese).

16) Goto. C. : A simulation model of oil

spread due to tsunamis. Proc. of JSCE.

Vol. 3 57, pp. 2 17-224. 198 5 ( in Japanese).

17) Imamura, F. and C. Goto : Truncation

error in numerical tsunami simulation by

the finite difference method. Coastal

Eng. in Japan. Vol.31, No. 2. pp. 245-263.

1988.

18) Sayama, J., F. Imamura and C. Goto and N. Shuto : Study on an accurate numerical

simulation method for tsunamis in deep

sea, Proc. of 34th Conf. on Coastal Eng..

JSCE, pp. 17 7- 18 1 . 1 98 7 ( i n Japanese).

19) Goto. C. and N. Shuto : Numerical

simulation of tsunami propagations and

run-ups. Terra Scientific Publishing

Company, pp. 439-451, 1983.

262
Theme IV

U.S.-Japan Cooperative

Research Program

U.S.-Japan Coordinated Earthquake Research Program on

Masonry Buildings— Seismic Test of Five Story Full Scale

Reinforced Masonry Building, Part 2

By

Shin Okamoto*. Yutaka Yamazaki^

Takashi Xaminosono"' and Masaomi Teshigavara

SUMMARY

In order to investigate the validity of the

design method and to know the total behavior of a RM building, static cyclic loading and pseudo

dynamic tests of a five story full scale RM

building was performed from mid-November, 1987 to

the end of January, 1988. The test building in the

extended service load range was very stiff with

only minor structural distress. It was found that

the characteristics of the test building in

elastic range are very sensitive to the condition

of foundation. The maximum lateral force capacity

reached 968 ton, at an overall building drift of

4/800 ( 0.5% ) in the yield load phase test.

Response of the pseudo dynamic test using the

Taft EW 1952 ( 300gal ) was within the hysteresis

loop obtained from the yield load phase test. The

ultimate deformation limit state was marked by

rapid strength degradation at an overall building

drift of 7/800 ( 0.875%).

The maximum shear carrying capacity and loaddeformation relationship of the test structure was

simulated well by a non-linear frame analysis.

KEY WORDS: Reinforced Masonry, Full Scale Test,

Static Cyclic Load, Pseudo Dynamic Test, NonLinear Frame Analysis

1. INTRODUCTION

Under the auspices of the UJNR on Wind and

-Seismic Effects, both the U.S. and Japan are working since 1984 on a coordinated earthquake

research program on masonry structures. The target

of the Japanese program is the development of

comprehensive design guidelines for medium rise RM

structures, in particular, the five story

apartment building, to meet the countrie's need of

high density residential construction. Based on a detailed understanding of the structural behavior

of RM building derived from analytical and

experimental research programs on materials,

components, sub-assemblages and planar frame

structures, the design guidelines (draft) was

developed

.

The five story full scale prototype test on a reinforced concrete masonry building was planned

and carried out in order to experimentally verify

analytical models which are assumed and dealt with

in the draft of the design guidelines. The main

portion of the full scale test consisted of three

static cyclic load phases, those were service load

phase, yield load phase, and ultimate load phase ( see Table 1). The lateral load distribution was

derived from the Japanese Building Code. Overall

test plan, design and construction, and brief test

results were presented at the last (20th) UJNR

meeting held in Washington D.C. May 1988 [Ref.l].

In this paper, adding to the review of tht

test structure and loading method, results of each

static cyclic load phase test and pseudo dynamic

test carried out just before the ultimate load

phase test are discussed. Pseudo dynamic test was performed in order to simulate the response of the

damaged building due to the secondary shock after

the main shock of an earthquake. Finally, large

amplitude of deformation was applied to the test

structure in order to obtain the deformation

capacity of the test structure and the deformation

characteristics of the structural elements in the

ultimate state.

This paper also discusses on the reason why

there is much difference between ultimate capacity

calculated by the design guidelines and the one

obtained from the test, as well as on the elastic

behavior by mean of frame model analysis. A frame

model analysis was applied because it is very

simple and very convenient for current design

method

.

2. GEOMETRY AND LOADING OF TEST STRUCTURE

Even though detailed reference on the

geometry and loading of the five story test

building should be made to [Ref.l], a brief

iteration of the most important geometrical data,

load application and reference notation is

presented in Fig.l to allow quick cross

references.

The five story full scale test specimen

represented a module of a prototype apartment

building with four parallel load bearing frames

with openings (Frames Yl through Y4, Fig. 1) in

the critical loading direction under

investigation. The floor area is 13.79m (loading

direction) x 15.19m (transverse direction,

including 2.4m wide balcony) and the total

building height is 14m from the top of foundation

to the roof slab (each story is 2.8m in height).

The standard and corner concrete units used in the

test building are shown in Fig. 2. Running bond was used in the test building. Both vertical and

horizontal rebars were arranged in 20cm spacing

module basically. Wall length ratios are 15.20cm/m in the loading direction, and

23.30cm/m in the transverse direction,

Dr. of Eng., Director, Research Planning & Information Department, Building Research

Institute, Ministry of Construction, Japan.

Dr. of Eng., Mead, Production Department,

Building Research Institute, Japan.

Senior Research Engineer, Production Department,

Building Research Institute, Japan.

Dr. of Eng., Research Engineer, Production

Department, Building Research Institute, Japan.

265
respectively. If the test building consists of 8

dwelling units per floor plan (4 of the test

building modules side by side), those values are 14.89cm/m and 18.15cm/m in each direction,

respectively.

Loads were applied to the center line (X2) of

the test structure by means of 11 servo-controlled

hydraulic actuators depicted schematically in Fig.

1(a). For the cyclic load testing, floor level

loads were applied in forced control as a fraction

of the roof level force as stipulated by the

Japanese Building Code. Roof level loading was

under displacement control to eliminate the

torsional mode. For the pseudo dynamic seismic

simulation test the five-story displacements were applied to one of the actuators at each floor

level while the other servo-controlled actuator

was force slaved to the first. Only at the roof

level, where three actuators applied the lateral

loading, see Fig. 1(a), the two outside actuators

were driven in displacement control again to

eliminate the fundamental torsional mode.

Wall and beam/girder components were

denominated by W or G, respectively, followed by a number and letter sequence indicating the plan

location. Preceding digits 1 to 5 denote the story

level under consideration. Thus, e.g., jWj^

denotes the first story wall labeled 1A at the

lower left hand corner in Fig. 1(b) at the

intersection of reference lines Yl and XI, etc.

3. SERVICE LOAD PHASE RESPONSE [Ref.2]

Crack pattern development in selected frames

in loading direction is shown in Figs. 3 and 4. Base shear versus displacement relationship

obtained through static cyclic loading tests is

shown in Fig. 5 and corresponding roof level

displacement envelope is shown in Fig. 6 together

with indications of the major events detected

during loading.

The service load phase covers the loading up

to a nominal base shear stress level of 8.0kg/cm

which is approximately twice the nominal value

corresponding to the 0.2G service load level

required by the Japanese Building Code.

Up to the loading of 4. Oka/cm ^ nominal base

shear stress level, the test structure showed very

little distress and a very stiff overall response

(The first story stiffness was 3,237ton/cm). Very

small flexural cracks in the lintel beams and a slight opening of horizontal bed joint cracks in

first and second story wall elements were observed

at this loading stage. Also, cracking at slab

corner started from this load level. The response

of the test building up to this load level can be

classified as virtually linear elastic.

During the loading between A.Okg/cm^ and

8.0kg / cm nominal base shear stress level,

flexural cracks at the horizontal bed joints of

walls and the lintel beams in all stories

developed, as shown in Fig. 3. Very small

diagonal cracks starting from a corner of small

rectangular openings located at the bottom of

first and second story wall elements ^Vl^^ and -jW^

started to develop. Cracks at floor slabs,

perpendicular to the loading direction, started to open due to flexural deformation of beams

concerned. Almost no cracks were observed in the

orthogonal walls Xj to X3.

The first yield in any instrumented

reinforcement was recorded in the flexural

reinforcements of long wall elements ^W^ and iWg,

and in the building corner reinforcement in i*gat

all at S.Okg/cm^ nominal base shear stress level, as shown in Fig. 7.

The first story secant stiffness at this

stress level decreased in 2,417ton/cm from initial

elastic stiffness, 3,237ton/cm.

4. YIELD LOAD PHASE RESPONSE [Ref.2]

Crack Development and Reinforcement Yield

Development

During the loading up to an overall building

drift angle of 1/1,200 (0.083%), wall element jW^g

and a joint element located above the first story

wall i^2B ex hibited diagonal cracking. At the

loading of 9.0kg/cm nominal base shear stress

level, horizontal bed joint cracks of transverse

wall elements were detected in the first story.

Figure 7 shows flexural reinforcement yield

development up to a building drift angle of 2/800

(0.25%). According to this figure, the

reinforcements in all first story wall elements

along the base of the building yielded up to a building drift angle of 2/800 (0.25%). The

flexural reinforcement arranged at lower side of

most beam elements also yielded until this

deformation level. The flexural reinforcements

arranged at upper side of all beam elements except

G

7g and o^6B not viel d due to existence of RC

tloor slat).

At 2/800 (0.25%) drift angle load level, the

first diagonal cracks occurred in most of all wall

elements in the first story and also in the second

story long wall element 2^3. The stiffness of the

building significantly dropped at this deformation

level due to this diagonal cracking as seen in

Fig. 6. The first diagonal cracking in a beam

element also was detected in the second story

short beam 2^5B" Remarkable diagonal cracking

from a corner of a small rectangular opening

located at the bottom of the first story wall

element occurred, and radiated cracking in

floor slabs adjacent to wall edges (including

straight cracking along with a beam face) occurred

as well at this deformation level.

At a deformation level of 3/800 (0.375%),

diagonal cracks in first story wall elements

increased in number. Face shell spalling in the

second story short beam 2^5B as we H as toe

crushing of the long wall elements jW^ and i^q, which indicate the onset of major structural

deterioration, started at this deformation level.

In the first story orthogonal walls, vertical

cracking was detected at the place distant 1.0-1.2

meters from the walls arranged in the loading

direction, which is considered to be produced by

excessive deflection of the walls in the loading

266
direction. Lateral load still increased up to a deformation level of 4/800 (0.5%) drift angle at

which the maximum lateral load (base shear)

968tons was recorded. This almost corresponds to

the total building weight, 996tons.

Change in Stiffness and Deflection Mode

The two cycles service load level loading was held after the service load phase and the yield

load phase responses in addition to the first

service load level loading as shown in Table 1. Table 2 shows story stiffnesses of the test

building under various loading level. It is

clearly understood from this table that all the

story stiffnesses changed at almost same rate of

stiffness degradation. This also indicates that

the building was considerably damaged under the

yield load phase response up to a deformation

level of 4/800 (0.5%) drift angle.

A plot of story drift percentage for each

story, shown in Fig. 8, indicates that the

deflection mode did not significantly change

through cyclic loading up to the maximum lateral

load level. One can just notice that the first

story drift percentage is slightly increasing with

increasing building drift.

Story rotation of the long wall element is

depicted in Fig. 9. It is clearly noticed that

flexural deflection is definitely caused by the

rotation at the foot--orf~ the first story wall

element jW^ and that, among three stories above

the 3rd story, the sign of the story rotation is

just opposite to the corresponding one in the

first story, even the magnitude of story rotation

at upper three stories is small. The fact

mentioned in the latter suggests existence of

considerable flexural reaction by beam elements in

these stories. The shear and flexural deflection

percentages in the total deflection of the first

story wall element jWr are 40% and 60%

respectively through the building deformation

level of the drift angles of 1/800 (0.125%)- 6/800

(0.75%).

The shear deflection versus story deflection

relationship in the first story wall elements

1 wIB • I W

2B and 1 W 3' depicted in Fig. 10, also

clarifies that the shear deflection is, roughly

speaking, 50% or a little less in the total story

deflection up to the building deformation level of

4/800 (0.5%) drift angle.

Yield Hinge Development

The consecutive yield hinge development is

depicted in Fig. 11. This almost coincides with

reinforcement yield development as shown in Fig.

7. The difference bet ween them is due to two

kinds of difinition of yielding, that is,

reinforcement yielding and structural element

flexural yielding, as schematically shown in Fig.

12. The point 1 in Fig. 12 just corresponds to a reinforcement steel yielding point (2,000 micro

strain in this case), whereas the point 2 was

determined as a point where sudden change in slope

on stress-strain relationship was observed, at

which the structural element concerned is

considered to have yielded.

Based upon Fig. 11, hinges were made at feet

of most wall elements (65% of total number of

hinges which were made finally) in the first

story, and at second and third story short beams,

2G^g and ^Gcg, until a building drift angle

reached 1/800 (0.125%). Among those hinges, the

ones at feet of two first story long wall

elements, iWe and i^q> were first formed at

1/1,200 (0.083%) building drift angle. Up to

2/800 (0.25%) building drift angle, all the first

story wall elements and some second story ones

yielded at their feet. Beam elements, almost 60%

of all, also yielded at their lower side

throughout the entire building.

Up to 3/800 (0.375%) building drift angle, 64

- 66% of total number of hinges were made in wall

and beam elements arranged in frames Yl and Y4,

whereas 80% in those arranged in frames Y2 and Y3

in which long wall elements were arranged.

During the loading up to 4/800 (0.5%)

building drift angle, another yield hinges were newly made mostly at those elements arranged in

frames Yl and Y4 so that hinges more than 80% were formed in these frames until this drift

angle. Most of all beam elements (83%) yielded at their lower side by then. Upper side of beam

elements did not yield throughout the testing

except second and fourth story beam elements in

fra me Yl and two second story beam elements in

frame Y4. This is considered to be caused by

existence of RC floor slab reinforcements.

Strain Distribution in Reinforcing Bars

Many strain gages were attached to surface of

reinforcing bars vertically and horizontally

arranged in walls, beams and RC floor slabs to measure effectiveness of those bars especially

arranged in the orthogonal elements. All the

vertical reinforcements in the frame X^ yielded at

upper surface of building f oundation(line h\),

when a building deformation level reached 1/1,200

(0.083%) drift angle, and also yielded at line H

2

, 90 centimeters above the line Hj, under 1/800

(0.125%) drift angle loading. The first story

vertical reinforcements in the interior frame X?

yielded under relatively large deformation as compared with those in the exterior frame Xj.

All those reinforcements in the frames Xj and

X? yielded at floor slab surface seimaltaneously

when flexural reinforcements arranged in concerned

wall elements in the loading direction yielded.

The yielding of vertical reinforcements arranged

in the walls which are located perpendicular to

the loading direction, thus clearly showed

effectiveness of all those reinforcements against

lateral loading.

Similar tendency was observed on floor slab

reinforcement associated with beam flexural

reinforcement. However, strain level through all

the measurements of slab reinforcements were very

low as compared with the one in vertical

reinforcements arranged in orthogonal walls, such

as the frame X,.

267
5. PSEUDO DYNAMIC TEST [Ref.3]

The objective of the subsequent f ive-degree0 f_freedom pseudo dynamic test was twofold, namely

to investigate the response characteristics, i.e.,

natural frequency, mode of response and lateral

force distribution, of the test structure after

damage against LOG level lateral load

accumulated, and to obtain the dynamic response of

the damaged building to an after-shock or a subsequent seismic event with a 0.3G maximum

acceleration.

The earthquake record of the Taft EW 1952

with a time window from 2.68 to 8.98 seconds was

selected as the driving seismic motion for the

pseudo dynamic test, since the first mode response was dominant with virtually equal amplitudes in

both the positive and negative directions under

preliminary computer analysis using the record.

The pseudo dynamic floor level response of

the damaged test structure is depicted in Fig. 13

for the roof floor level displacement and base

shear force time histories. After 6.3 seconds

passed in pseudo dynamic response, the free

vibration response produced a natural period of

vibration of 0.54 second, slightly longer than the

measured forced vibration response (0.41 sec.)

after the yield load phase test.

Visual inspection of the test structure,

i.e., trace of crack propagation and damage area assessment, did not indicate any additional

structural damage or stiffness deterioration

during the 300gals secondary seismic event. The

obtained hysteretic response (Fig. 14) to the

secondary seismic event stayed well within the

hysteresis envelope obtained under the yield load

phase test. Figure 14 again indicates that the test

structure can still perform in a stable manner

without increased loss of structural integrity.

Finally, Fig. 15 depicts the variation of

peak response of the test structure along the

height. The dominance of the first mode response

is clearly visible as well as the limiting

response envelope obtained at the ultimate lateral

load limit state during the cyclic testing of the

yield load phase. Of particular interest is the

shape of the lateral load distribution over the

building height, Fig. 15a, which is almost linear

(inverse triangular) as opposed to the shape of

the imposed lateral force distribution for the

cyclic load testing which was based on the

Japanese Building Code. The straight line inverse

triangular response is indicative of the formation

of flexural mechanisms at the base of the first

story wall elements. This tendency is emphasized

by almost straight line lateral displacement

distributions along the building height, see Fig.

15b, compared to the displacement envelopes

obtained during the cyclic yield load phase test.

6. ULTIMATE LOAD PHASE RESPONSE [Ref.3]

Subsequent to the maximum lateral load limit

test and the pseudo dynamic seismic test, the test

structure was subjected to increased cyclic

lateral deformation levels, beyond the 4/800

(0.5%) building drift at maximum load, to

investigate the strength degradation

characteristics and to evaluate the ultimate

failure mechanism. Overall building drift levels

of 5/800 (0.625%), 6/800 (0.75%) and 7/800

(0.875%) were imposed in single load cycle with

full reversal. At a total building drift of 7/800

(0.875%) the ultimate deformation load test was

terminated since a lateral load carrying capacity

of less than 50% of the maximum lateral load level was obtained on the reverse deformation cycle.

The base shear versus drift relationship for

the overall building (roof level) and the first

story level are depicted in Fig. 5. In the

initial deformation cycle load levels of 98%, 90%

and 69% of the maximum lateral load were achieved

at overall building drift levels of 5/800

(0.625%), 6/800 (0.75%) and the 7/800 (0.875%),

respectively. A comparison between the overall

building drift and the first story drift shows

first story drift levels of 1/102 (0.98%), 1/70

(1.43%) and 1/42 (2.38%) for the corresponding

overall building drift levels of 5/800 (0.625%),

6/800 (0.75%) and 7/800 (0.875%). The cyclic

ultimate deformation load test showed that 80% of

the maximum lateral load capacity was still

maintained at overall building drift of 1/120

(0.83%) and first story drift of 1/60 (1.67%).

The lateral deformation characteristics of the

test building over the building height is

depicted in Figs. 16 and 17. Individual story

displacements and story drifts are plotted over

the height of the test structure for various

overall building drift levels in Figs. 16a and

16b, respectively. A significant increase in first

story deformation and drift for overall building

drift levels of 6/800 (0.75%) and 7/800 (0.875%) can be seen, with very little additional

deformations in the upper stories two to five.

The rapid increase in story drift without an increased rotational component is a clear sign

for shear deformations and the development of a shear mechanism in the first story walls. This

tendency is also depicted by the overall

deformation modes for lateral load bearing frame

Y2 in Fig. 17, where deformation states at 3/800

(0.375%) and 7/800 (0.875%) radians are compared.

The additional upper story deformations are negligible compared with the first story

displacements

.

The above findings are confirmed by large

diagonal cracks in all first story walls as depicted by the ultimate crack pattern of all four

load bearing frames in Fig. 18. The opening of

these diagonal cracks in the first story wall

components occurred at deformation levels of 5/800

(0.625%) and 6/800 (0.75%) building drift, which

corresponds to the increased lateral deformations

in the first story at these load levels. Few of

the cracks in the upper stories as well as the

cracks in the floor slabs and in the transverse

walls extended or widened during the ultimate

deformation load phase test. Thus, the overall

deformation mechanism encountered in the test

structure clearly shifted from an overall flexural

behavior up to and at the ultimate load limit

state (5/800 (0.625%) building drift) to a shear

268
dominated mechanism in the first story during the

ultimate deformation load phase test.

7. FRAME ANALYSIS [Refs. 4 and 5]

Model

The test building is idealized to be frames

composed of beam members representing the wall and

beam components. In order to represent the

nonlinearlity of members, walls and beams, onecomponent model consisting of flexural spring at

the both edges of a member and a shear spring at

the center of a member is considered. Beam model

has rigid zones at both ends. Flexural spring

characteristics representing flexural behavior and

shear spring characteristics representing shear

behavior of a member are modeled to be a trilinear one with cracking and yielding points.

For flexural spring, yield rotation angle(

T ) is assumed to be 1/800 for lm long wall and

earn, 1/1,500 for 2m long wall and 1/3,000 for 4m

long wall. Maximum flexural strength is

calculated by the equation in Ref. 7. Those

values are listed in Table 3. For shear spring,

nominal shear cracking stress is assumed to be

15kg/cm^. Therefore shear cracking deformation

angle( Y c

) becomes almost 1/4,000.

The value of Young's Modulus (E) is defined

from the prism compressive test results. In this

analysis mean values of E and Fm are 196t/cm and

196kg/cm , respectively. Shear Modulus (G) is

gotten assuming that poisson's ratio( V ) is 0.2

(G=86t/cm2

).

The joint part of wall and beam is treated as rigid zone in this analysis. In the elastic

analysis this rigid zone is assumed as illustrated

in Fig. 19 and in the inelastic analysis full part

of beam-wall joint is assumed to be rigid zone. The effective width of transverse members for

elastic stiffness are determined by Ref. 7.

Typical component and subassemblage test can

be traced with this beam model and the validity of an inelastic beam (member) model used in this

paper is verified [Refs. 4 and 5].

Foundation

Test building is fixed to testing floor

(depth = 2.00m). Test building is too rigid to

consider the foundation be fixed. In this

analysis foundation beam (1=121. 67xl0^cm,

A=10,575cnO and testing floor (I=600x 10 5cir, , A=18, 000cm ) and the rotational stiffness (see

Table 3c) due to prestressing bars are

considered

.

Results of Elastic Stage

Four cases listed in Table U are analized.

Among Model 1 through Model 3, different effective

width of transverse members to inertia moment is

considered. In Model 1, rectangular section

without taking account of transverse members is

considered. In Model 2, the full width of

transverse members are taken into each member's

inertia moment. The inertia moment of Model 3 is

defined according to the design guidelines

[Ref. 7], The design guidelines recommend that

incremental factor of inertia moment due to

transverse members is up to 2.0, and inertia

moment of almost all members with transverse

members are limited to be 2.0 times as that of

rectangular section (I 0

). These three cases,

Model 1 through Model 3, are assumed to be under

fixed end condition. In Model 4 the effect of

the foundation is considered, other conditions

besides the foundation are the same as

corresponding ones in the Model 3.

The vibration periods are tabulated in Table

5. Experimental results are also listed in this

table. Experimental results are obtained from the

flexural matrix gotten by the each floor unit load

test. Model 4 gives the closest results to the

test ones among these Models. The limitation of'

effective inertia moment up to 2*I Q

, is

reasonable. The recommendation for effective width

of transverse members in the design guidelines is

also reasonable. Vibration modes between Model 3

and Model 4, are different each other. It clearly

shows the necessity to consider the effect of

foundation condition in elastic stage.

Overall Behavior in Inelastic Stage

Monotonic loading analysis is carried out.

Inertia moment and foundation condition are the same as the Model 4. In this case, however, full

part of the panel zone surrounded with walls and

beams is taken into account as a rigid zone. This

is based on the fact that the purpose of this

analysis is to predict the lateral load capacity

of the test building from the component strength

at their critical section, and also that the

stiffness degradation of each member is so dominant to the total stiffness of the test

building that rigid zone of wall-beam panel is

assumed not to affect on the total stiffness of

the test structure. Maximum moment capacity of

each member is calculated with considering the

effect of all re-bars in transverse members.

Figure 20 shows the each story shear force

vs. story drift angle relationship. Good

agreements between test results and this analysis are observed. This means that the effect of rebars in transverse member on flexural strength and

characteristics of flexural and shear springs are evaluated properly in this analysis. These spring

characteristics are defined based upon the

component test results conducted in past years.

Development of Yield Hinges

Test result shows that the yielding of wall

occurred at first in the nominal shear stress

of 8kg/cm loading step. But it means that the

extreme edge re-bar reached its yield strain at

this loading level. The re-bars' strains in

middle part of wall Wc increased rapidly from

building drift angles of 1/800-2/800 (0.125%-

0.25%). This fact matches well to the analytical

results.

269
8. CONCLUSIONS Ultimate Load Phase Test

Conclusions obtained from the seismic test on the five story full scale reinforced masonry

building are listed as follows:

Service and Yield Load Phase Tests

1. The test building showed very little distress

and a very stiff overall response against the

loading of 4.0kg/cm^ nominal base shear stress

level which corresponds to the 0.2G design load

level. The first story drift angle of the

test building against the 0.2G design load

level was 1/4,530. Almost no crack was

observed under this design load level loading.

The test building certainly exceeded design

requirements of carrying lateral load capacity

corresponding to 0.5G base shear which is

specified in a draft of a seismic design

guidelines for medium rise RM buildings.

2. The actual lateral load capacity obtained

during the test at an overall building drift

angle of 4/800 (0.5%) was 968tons which is

almost equivalent to the weight of the test

building (dead load + live load) or a base

shear coefficient of LOG.

3. The overall development of failure mechanisms was governed by formation of flexural yield

hinges at the base of the first story walls,

especially walls Wg, W^ and Wo, and also at the

beam ends throughout the entire building. All

first story walls clearly yielded in flexure

long before the development of major diagonal

cracks which ultimately dominated the first

story wall limit behavior.

4. Significant portions of the transverse walls

contributed to the overall load transfer in the

form of wide flanges to the four load bearing

' frames. Activation of all the transverse wall

reinforcement certainly contributed to the high

maximum lateral load capacity. Large

contribution of floor slab to beam strength

can be considered from the observed

transverse crack patterns in the reinforced

concrete floors and also from strain

distribution in the floor slab reinforcements

associated with flexural reinforcements in the

beams

.

Pseudo Dynamic Test

5. In spite of the comparatively large magnitude

of the input acceleration, 300 gals, response

of the specimen was within the hysteresis loop

obtained from the yield load phase test, and

there was no progress in damage of the

specimen. The lateral force distribution was more similar to inverted triangular

distribution than that imposed in the static

loading test, and also the deflection mode

was nearly linear compared with the one in the

yield load phase test. The natural period of

0.54sec. was obtained from the free vibration

by means of the pseudo dynamic test method.

6. Over 5/800 (0.625%) building drift angle,

lateral load capacity decreased gradually and

deflection started to concentrate to the first

story. Flexural deflection mode was changed to shear failure mode which was dominated by only

the first story displacement. After 6/800

(0.75%) building drift angle, shear cracks at

the first story walls expanded and lateral

load capacity of the test building decreased

rapidly with concentration of deflection to the

first story. The overall building drift angle was approximately 1/120 (0.83%) when the

strength decreased to 80% of the maximum

strength.

Frame Analysis

7. Total behavior of the test building was

simulated by a non-linear frame analysis. In

this analysis, wall and beam characteristics

are assumed based on a component test and

simple beam theory. The elastic analysis

showed that rigid zone recommended in the RM

design guidelines is reasonable. And the

ultimate analysis also showed that the lateral

load capacity of the test building can be

estimated by the flexural strength at critical

section of each component. It is also

important for discussing the elastic behavior

of such a rigid structure to consider the

condition of foundation. And in inelastic

stage, all the reinforcing bars in the

transverse members are effective on the

ultimate shear carrying capacity.

ACKNOWLEDGMENTS

The five story full scale masonry test was carried out as a part of the U.S. -Japan

coordinated earthquake research program under the

auspices of the UJNR on Wind and Seismic Effects.

In Japan side, the program was promoted under the

Promotion Committee for Masonry Research (PROCMAR,

chaired by Prof. H. Umemura), the Technical

Coordinating Committee for Masonry Research

(TECCMAR, chaired by Prof. T. Okada) and the

Building Construction Committee for Masonry

Research (BLDCMAR, chaired by Prof. K. Kamimura).

The writers would like to express their gratitude

to these committees. The contributions of the

U.S.-TCCMAR members, especially Prof. F. Seible,

UCSD, in numerous discussions and joint meetings

also are greatly appreciated.

The writers would like to record the generous

cooperation of Dr. H. Mizuno, Messrs H. Kato,

H. Isoishi, T. Nishi, and M. Matsuno for

performing the full-scale RM building test.

REFERENCES

[1] S. Okamoto, Y. Yamazaki, T. Kaminosono, and M. Teshigawara, "Seismic Test of the Five Story

Full-Scale Reinforced Masonry Building", 20th

Panel Meeting on Wind and Seismic Effects,

UJNR Cooperative Research Program, Washington

D.C., May 17-20, 1988.

[2] Y. Yamazaki, et al., "Seismic Test of Five

270
story Full Scale Reinforced Masonry BuildingOutline of Test and, Service and Yield Phase

Response", Proceedings of the Fourth Meeting

of the U.S.-Japan Joint Technical Coordinating

Committee on Masonry Research, San Diego,

October, 1988.

[3] T. Kaminosono, et al., "Seismic Test of the

Five-Story Full-Scale Reinforced Masonry

Building-Pseudo Dynamic Test and Ultimate

Phase Response-, "Proceedings of the Fourth

Meeting of the U.S. -Japan Joint Technical

Coordinating Committee on Masonry Research,

San Diego, October 1988.

[4] M. Teshigawara, "Introduction of Dynamic

Inelastic Frame Analysis on Reinforced Masonry

Structures," Proceedings of the Second U.S.-

Japan Joint Technical Coordinating Committee

Meeting on Masonry Research, Keystone, USA,

1986.

[5] M. Teshigawara, T. Kaminosono, Y. Yamazaki and

T. Nishi," Seismic Test of Five Story Full

Scale Reinforced Masonry Building- Analyses on Service and Yield Phase Response-",

Proceedings of the Fourth Meeting of the U.S.- Japan Joint Technical Coordinating Committee

on Masonry Research, San Diego, October, 1988.

[6] "Design Standard for Reinforced Concrete

Building," Architectural Institute of Japan.

[7] T. Kaminosono, et al., "Draft of Design

Guidelines for Reinforced Masonry Buildings,"

Proceedings of the Third Meeting of the U.S.-

Japan Joint Technical Coordinating Committee

on Masonry Research, Tomamu, Hokkaido, Japan,

October 1987.

271
Table 1 Test and Loading Sequence

Test

Phase

Response

Level

cracking

V

ultimate

[displacement

1. Static (cyclic)

Service

Load

Level

[kg/cm2

]

Load

Tests

[kg/cm2

]

Displ.

Tests

[rad]

I 3 l lcycle

| A llcycle

| 4 [lcycle

| 4 |2cycles

3cycles

3cycles

9 2cycles

| 4 |2cycles

1/800 2/800

3/800

4/800

2.

FORCED

VIBRATION

ESS

t/i2Qo| 2cycles

2cycles

3cycles

3cycles

3cycles

3/300

6/800

7/800

lcycle

lcycle

lcycle

PSEUDO

DYNAMIC

UNIT

LOADING

ma

1777A

Table 2 Story Stiffness of 5 Story Full Scale Test Building

(ton/cm]

story

4kg/cra2 Stress Level 8kg/cm2

Stress

Level

1/800 building

Drift Angle

Virginal Level

Loading

After Service

Load Phase

After Yield

Load Phase

5 1975

(100)

1660

(84.2)

230

(H.6)

1641

(83.1)

844

(42.7)

4 2520

(100)

1906

(75.6)

340

(13.5)

1978

(78.5)

1240

(A9.2)

3 2763

(100)

2158

(78.1)

363

(13.1)

2180

(78.9)

1329

(48.1)

2 3237

(100)

2401

(74.2)

394

(12.2)

2464

(76.1)

1525

(47.1)

1 3237

(100)

2392

(73.9)

291

(9.0)

2417

(74.7)

1554

(48.0)

272
Table 3 a) Characteristics of Member-BeamMyi My?

889 9,150

1,228 9,589

1,355 9,940

1,566 9,940

My ^ : Ultimate moment when lower side is in tention [t.cm]

My2^ Ultimate moment when upper side is in tention [t.cm],

including the re-bars in slab (28D10 and 8D13)

Mc : Flexural cracking moment = My/3

I = 15.6*105cm

2

I < 2*1 I =bD3 /12 (7.8*105

cm

4

) Iy=l/800rad.

A : Area for shear stiffness = 1,501cm #1.2, 1.2:effect of RC slab

Qc= Strength of shear crackajig = 15kg/cm #A = 27.0ton

Foundation Beam. I=721.67*105cm , A=28, 575cm 2

Member

Rp3

5^3

• 4G

3,2U

3

Table 3 b) Characteristics of Member-WallW4A

W

5

r My I,Qc, y,A

5 10,544

4 8,178 1=249. 55*105

3 8,781 Qc=68.0 ton

2 6,784 Iy=l/1,500

1 4,192 A=3,781 cm

z

5 31,301

4 34,997 I=1164.0*105

3 43,395 Qc=143.3

2 47,091 £y=l/3,000

1 50,787 A=7961

5 7,559

4 11,555 1=249. 55* 105

3 16,200 Qc=68.0

2 20,510 £y=l/l,500

1 21,058 A=3,781

cm

4

W4B

My : Ultimate moment [t.cm]

Ts = 15kg/cm

Qc = Ts*A*1.2. 1.2 : Effect of transverse wall

A : Area for shear stiffness and strength

273
Table 3 c) Characteristics of Rocking

Location Krn(t.cm)

i Qr,ot-h wall £?n*lf)^ 2m length wall 420*10

4m length wall 1,700*104

KT

=2* 32*Es*D2 /l=105D2

Es=2,100t/cmo 32=8cm2

1: length of PC-bar ( 320cm)

Dtwall length

Table 4. Analytical Condition of Model 1-4

Model 1 Model 2 Model 3 Model 4

Transverse Not considered Full width According to Same as the

member considered the RM D.G[7] Model 3

Foundation Not considered Not considered Not considered Considered

(Fix) (Fix) (Fix)

Shear area Rectangular only

E = 196 ton/cnT G = 86 ton/cm

In the elastic analysis for Model-1 to Model-4, rigid zone of wall-beam

joint is assumed as shown in Fig. 19 (according to A.I.J, standard

for RC structure [Ref.6]).

Table 5 Natural Period (sec.)

MODE

Model

Ml M 2 M 3 M 4 M 5 TEST

1

2

3

4

5

0.184 0.12 0.159

0.055 0.039 0.048

0.029 0.022 0.026

0.020 0.016 0.018

0.016 0.013 0.015

0.169 0.166

0.052 0.051

0.028 0.027

0.019 0.019

0.015 0.016

0.157

0.049

0.026

0.019

0.013

FIX END CONDITION

* note: It is very difficult to evaluate the proper rotational stiffness

of foundation because of its pre-stressed PC bars. Model A ignores the

effect of this pre-stressed. In supplementary analysis, Model 5,

rotational spring due to pre-stressed PC bars was changed to be 1,000

times rigid as compared with Model 4. Natural period does not changed so much(Tl=0.169 sec in Model A, Tl=0.166 sec. in Model 5). And the first

vibration mode of Model 5 closes to the test results than that of Model

4. It is supposed that rotational stiffness assumed in Model A would be

enough rigid.

274
XI X2 X3

a) Test Setup and Load Distribution

® © ©

13.6m (44 - -8") 'II ! 5.4m (17--9') 2.8m (9 -2") 5.4m(l7 -9-J

b) Plan Geometry and Reference Notation

Fig.l Geometry and Loading of Test Structure

275
390

f

"s

290

o

1 190 1

Standard Corner

(1=290)

Corner

(1=190)

Fig. 2 Concrete Block Units [scale in mm]

•p- sac

•

r

i

4— r

(

t J 1

I pn T T ~

l"

i i

| rn

-r r

¥ 1 1 1 i

Frame Y1 Frame Y3

Fig. 3 Crack Pattern of Frames -Yl and -Y3 at T=8kg/cm^

276
Frame Y1

Fig. A Crack Pattern of Frames

Frame Y3

-Yl and -Y3 at R

t =4/800 rad.

c

o

or<

IDX

toUJ

to<

CO

1200

BOO

400

0

-400

-800

1F DISP. (mm)

-56 -28 0 28 56

RF DISP. (mm)

1 F IT'

Q

i

- i

• I

22.9

15.2

7.6

0

-7.6

-15.2

-22.9

E

o\cn

(0

(O

UJ

a.

r-tO<

UJX

to<2

;uu -16 -8 0 8 16

IF DRIFT ANGLE(xl/800rad.)

b) 1st Story Drill

1200 -140 -70

c 800

o

+-

or 400

<

UJX 0

coUJ

10 -400

<

CD

-800

-1200

— 70 140 1 1— 1 —1—

(

.1.1.1. -I.I.I.

22.9

15.2

7.6

0

-7.6

-15.2

E

u\C71

"» 1

to

toUJ

cc

to<

UJX

to<

2

-8-4 O 4 8

TOTAL DRIFT ANGLE

(xl/800rad.)

a) Building Drift

-22.9 2

Overall Building Drill Levels (4) R = 4/800rad. or 0.500V.

(7) R= 1/800rad. or 0.125% (5) R = 5/800rad. or 0.625%

(2) R = 2/000rad. or 0.250% (?) R = 6/800rad. or 0.750%

(3) R = 3/800rad. or 0.375% (7) R = 7/0OOrad. or 0.875%

Fig. 5 Base Shear vs. Displacement Relationship

Obtained through Static Cyclic Loading Tests

277
c

o

S 800 - JZ

V)

cu

COo

m

x lOkg/cm* «. 9kg/cm'

Slcg/cm* *

6kg/cm'

4kg/cmf

3kg/cm*

o

o

00

*r

n

o

o

00

in

n

o

o

00\

H

OO

CO\

r>- n

« 500

^Diagonal cracking at wall jW2B occurred

j

I •To* crashing started at

1st story long walls.

•Host of all sialn flexural

. rebars in the beams yielded.

*Fac« .shells of beam jCjjj L started to fall down.

Diagonal cracking at 1st story walls occurred.

Cracking from lat story small openings started.

Diagonal cracking of beam 2**5B occurred.

Shear crack at wall-beam Joint element between

wall jWjb and beams ^Gjg and jGjjj occurred.

Host of all main flexural rebars in the

L 1st story walls yielded.

Shear crack at wall-beam joint element between

wall |V2Q and beams 2<»1B snd 2^25 occurred. Beam flexural rebars yielding started.

•Cracks along horizontal bond Joints of lat

atory wall in Frames XI and XI occurred.

1*5 main flexural rebar .yielded.

10 20 30 40 50 60 70 80 90 100 110 120

Top Displacement (mm)

Fig. 6 Base Shear vs. Top Displacement Envelop

Curve for Static Cyclic Loading Tests

278
4 5

J

4

RF

5F

4F

3F

5

2F

WlB

3—

5|I_5JI 115

j5E

! Wa

m

Gab

5 5

W7B

RF

5F

4F

3F

5

2F

Wee

Ws

" W4B

|5 I

Y

RF

5F

4F

3F

2F

45

Wt1B

00

4! !5

a

5

^ I 5 G6B

G7B

WlOB B

RF

5F

4F

3F

5

2F

W9B

1 4 4 4 3 4 4 1

Building Drift 2 -1/1670(0.060%) A -1/800(0.125%)

1 -1/2000(0.050%) 3 -1/1200(0.083%) 5 -2/800(0.250%)

Fig. 7 Consecutive Reinforcement Yield Development

279
~ 40r

30

20

10

1/800 2/800 3/800 4/800

BUILDING DRIFT ANGLE

Fig. 8 Story Drift Percentage during

Yield Load Phase Pesponse

a i 'tea * mm o ftn a -itM x -l't« » -vtsf

-6 -4 -2 0

STORY ROTATION (xlO" 3

)

Fig. 9 Story Rotation of Wall W5

Fig. 10 Shear Deflection vs. Story Deflection

on First Story Walls

280
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Numerals : indicate a building drift angle scaled

in x 1/800

• : strain gage locations

X : dead strain gage locations

Fig. 11 Yield Hinge Development
Free Vtbr»t1on - 600. co

-600.

Story Shetr Force

1 F

A7

Free Vlbrttlon

6 8

Tine (sec.)

a) Time History of Horizontal b) Time History of Lateral Force

Displacement

Fig. 13 Pseudo Dynamic Test Floor Level Response to Taft EW 1952 (0.3g)

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HYSTERESIS LOOP

of the YIELD

LOADING PHASE

- " 5 10 15 -200 2F DISP. (mm)

• -400

--600

Fig. 14 Relationship Between Base Shear and the First

Story Displacement During Pseudo Dynamic Test

Aft* - -

FLOOR ^LOAD DISTRIBUTION

.* In the STATIC TEST

-200 -100 0 100

LATERAL FORCE (ton)

a) Lateral Force Distribution

at Peaks of the Response

200

- • • -

• DEFLECTION MODE

of the YIELD

LOADING PHASE

-60 -40 -20 0 20 40 60

DISPLACEMENT (mm)

b) Deflection Mode at Peaks

of the Response

Fig. 15 Peak Response to Taft EW 1952 (0.3g)

282
LOADING LEVEL (x1/800rad.)

Rf

_ +1 +2+3 +4 +5 +6 +7

40 80 120

FLOOR DISP. (mm)

LOADING LEVEL ^

(xl/800rad.)

+1 +2+3+4+5 +6 +7

j i_ j i 10 20 30

STORY DRIFT ANGLE

(xlO'Vad.)

a) Building Deflection Mode b) Loading Level vs. Story Drift Angle

Fig. 16 Progressive Deformation History

W4A W5 W4B

FRAME-Y2 (+3/800rad.)

W4A VV5 W4B

FRAME-Y2 (+7/800rad.)

Fig. 17 Overall Deflection Patterns of Frame Y2

283
VV7A

c) Frame Y3

I^i l 'l — — J

1

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:1 1

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-0^ ••"-•\?yai^±>/r":

§5 K &

Wsa Woa Wha Wia

d) Frame Y4

V/oe W9B

Fig. 18 Ultimate Crack Patterns in Lateral Load Bearing Frames

284
rigid zone

Db/4 IV /A DC/

4

1 J L Db

n

Dc/4-1

Dc

t _ 1 Db/4

frane line

Beam

wal 1

Fig. 19 Assumption of rigid zone

T.A

4 F

, YSjjS 5 F

:—~"T EST

-

r

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— 200 TON — 1/800 R — 1/200 R r— 200 TON

'Testj— 1/800. R

L- 1/200 R|

~ii r i i i r- STORY DRIFT

DRIFT ANGLE ( RAD X E-03 )

Fig. 20 Story Shear vs. Story Drift Relationship

285
Influence of Horizontal Reinforcement on the

Shear Resistance of Masonry Walls

by

Charles W. C. Yancey^

ABSTRACT

Ten concrete masonry wall specimens were tested

to determine the effect of varying the amount

and distribution of horizontal reinforcement on

in-plane shear resistance. The walls were

constructed with hollow concrete block and

varying amounts of bed joint reinforcement and

reinforcing bars located in grouted bond beams

.

They were subjected to reversed cyclic, inplane lateral loads and essentially a constant

axial compressive load. Results of the tests

indicate that relatively small amounts of

horizontal reinforcing are effective in

increassing post-cracking strength. In-plane

shear strength did not increase proportionally

with increasing amounts of reinforcing.

KEYWORDS: axial load; bond beams; concrete

masonry; cracking load; cyclic

loading; energy absorption;

horizontal reinforcement; shear

failure; shear walls; ultimate

strength

1. INTRODUCTION

1.1 Background

This series of tests initiated the second phase

of a masonry research program being conducted

at the National Institute of Standards and

Technology (formerly the National Bureau of

Standards). The primary objective of the

research program is to gather data necessary in

defining in-plane shear capacity and

deformation behavior of shear-dominated masonry

walls. During the first phase of the program,

unreinforced masonry walls were tested to

obtain a benchmark for comparison of results

from tests on reinforced walls. Variables

investigated during the Phase I testing

include: magnitude of axial compressive stress,

aspect ratio, masonry type, mortar type, and

grout strength. Phase II testing is divided

into three series:

horizontally-reinforced walls, verticallyreinforced walls, and walls with combined

vertical and horizontal reinforcement. This

paper reports the results from tests

conducted on ten horizontally-reinforced,

masonry walls.

2. TEST SPECIMENS

2.1 Materials

The masonry units were 8x8x16, hollow, twocore, concrete blocks with a compressive

strength of approximately 1800 psi, based on

gross cross-sectional area. The mortar was

proportioned in accordance with ASTM

designation Type S. The water/cement ratio

used for the grout was based on trial

batches to produce a compressive strength

approximately equal to that of the block

units. Bed joint reinforcement consisted of

9-ga, ladder-type, galvanized wire. Bond

beam reinforcement consisted of combinations

of #3, #4 and #5 deformed bars.

2.2 Wall Details

The walls were fabricated in the laboratory,

using running bond construction. They were

48 in. long and 7 5/8 in. thick, and had a

net area of the horizontal cross section of

196.5 sq. in. The height of the walls was

56 in. , resulting in an aspect ratio

1 National Institute of Standards and

Technology, Odthersburg, MD 20899

286
(height/length) of 1.17. An initial axial

force of 39,300 Ibf was applied to all walls.

This force caused an axial stress of 200 psi on

the net area. Based on results a series of

test run during Ihase I, this axial stress

level was sufficient to promote a shear mode of

failure. As shown in figure 1, there were two

different bond beam configurations: l)a single

bond beam at mid-height, and 2) two bond beams

at the second and sixth courses. Table 1 summarizes the details of the ten specimens.

3. TEST PROTOCOL

3.1 Test Setup

The walls were tested in a structural test rig

consisting of servo-controlled hydraulic

actuators that are capable of applying any

combination of forces or displacements in the

six-degrees-of freedom at the top of a wall

specimen. The six degrees of freedom are three

translations along and three rotations about

the three orthogonal axes. Vertical loads were

applied through the use of four actuators which

were attached to the ends of the upper and

lower crossheads . In-plane lateral load was

applied by an actuator which was aligned with

the central plane of the wall. Upper and lower

loading shoes were attached to the ends of the

walls to simulate a fixed-fixed boundary

condition.

3.2 Instrumentation

The horizontal and vertical actuators were

equipped with load and displacement transducers

to measure the overall forces and moments

applied to the upper crosshead and

displacements of the actuator pistons as they

moved the crosshead. The in-plane horizontal

displacement of the walls was measured by

linear variable differential transformers

(LVDTs) which were vertically spaced along the

loaded edge as show in figure 2. vertical and

diagonal displacements were measured using four

LVDTs which were mounted on one face of the

specimens. A Timber of specially designed

strain gage devices, called leaf spring

transducers (LSTs) , were attached to one face

to measure localized deformations in the block

units. The LSTs were positioned such that they

did not cross mortar joints. Figure 3 gives

the locations of the LSTs deployed on wall

specimen Rl.

3.3. Test Procedure

The test procedure was standardized for all

specimens except wall R8. The reversed,

cyclic loading history illustrated

schematically in figure 4 was used as the

guide. First, the specified vertical

compressive load, 39,300 lbf, was applied.

While maintaining this load at in-plane

lateral displacements were applied at the

top edge of the walls. The first

displacement excursion was 0.02 in. , or

about 25% of the estimated displacement at

first cracking. A first cracking

displacement of 0.08 in. was predicted based

on test results from Phase I. Three

reversed cycles of lateral displacement were

applied at each increment of 0.02 in. up to

the maximum excursion of 0.06 in.

Thereafter, the displacement was increased

until a major crack formed (i.e. First Major

Event or IME). The direction of the

crosshead was then reversed to a

displacement equal to that recorded at EME.

The degradation and stabilization cycles

indicated in figure 4 were followed as the

walls were subsequently loaded to failure.

As the lateral displacement magnitude

increased, the axial load gradually

increased. Wall R8 was tested under load

control in the vertical direction.

4. TEST RESULTS

All ten of the wall specimens exhibited

shear modes of failure as characterized by

cracks propagating along each major

diagonal. Table 2 presents a summary of the

results for two limit states: 1) first

cracking and 2) ultimate load. The steel

content percentages given in column 4 were

computed by dividing the area of the

reinforcement (i.e. bars or galvanized wire)

by the product of the wall's thickness and

height. The total horizontal load at first

cracking and the corresponding lateral

displacement at the top of wall, the

magnitude of the ultimate load and the

lateral displacement at ultimate load are listed in columns 5 through 8.

287
The displacement values, referred to as "global in-place displacements," are those

measured by the displacement transducer located

in the horizontal load actuator.

Figure 5 presents load-displacement curves for

four walls with varying amounts and

distributions of reinforcement: 1) wall Rl was

unreinforced and provides a benchmark for

comparing response characteristics; 2) wall R4

had no bond beam but contained joint

reinforcing at every horizontal mortar joint;

3) wall R5 contained a mid-height bond

beam,with 2 - #4 bars, but had no joint

reinforcing; and 4) wall RIO had a mid-height

bond beam, with 2 - #4 and 1 - #5 bars, and

joint reinforcing at every course. The load

and deflection values at EME and ultimate are noted on the curves. The bracketed nuafoers

indicate the nunber of loading cycles applied

to the specimens at the time of occurrence of

the limit state.

Figure 6 provides a qualitative comparison of

behavior for the same four walls in the form of

schematic drawings of the crack pattern at the

end of testing.

The energy absorbed during each cycle of

loading was computed using the loaddisplacement data. Bar charts quantifying the

energy absorption magnitude per cycle are presented in figure 7 for walls Rl, R4, R5, and

RIO. The vertical scale is the same for all

four bar charts to more clearly illustrate the

relative magnitudes of energy absorption.

5. DISCUSSION OF RESULTS

The qualitative and quantitative results from

wall Rl provide a baseline of comparison for

the other wall specimens. The majority of the

cracks that existed in wall Rl at the end of

testing were formed in the first few cycles of

inelastic deformation and progressed as testing

continued. Strength and stiffness degradation

occurred as the result of grinding of mortar

joints and crushing of block units along cracks

in the bed and head joints. The FME and

ultimate load, occurring in the direction of

initial deflection, were of the same magnitude.

The number of cycles of loading before reaching

ultimate was 17. Referring to figure 7, it is

observed that the maximum energy level in any

cycle was about 3 kip-in. Comparing the

results from walls R2 and R4 with those of

wall Rl, it was observed that cracks formed

in the two specimens with bed joint

reinforcement at about a 20% higher load

than in the unreinforced wall. The cracks

tended to remain closed, and there was less

grinding and crushing along cracks in walls

R2 and R4 than in wall Rl. As borne out by

figure 6, cracking was more uniformly

distributed over the faces of R2 and R4 than

for Rl. Ultimate load was increased by the

presence of bed joint reinforcement placed

either at every course or every other

course. The displacement and number of

cycles at ultimate load for walls R2 and R4

were approximately double the corresponding

value for wall Rl. (Comparison of results

from R2 with those from R4 indicate no

significant difference in wall response as a

function of the spacing of the bed joint

reinforcing.

Referring to table 2, the comparative

entries for wall Rl and walls R5 or R6

indicate a significant increase in the

ultimate load levels as a result of using a

reinforced bond beam. The presence of the

mid-height bond beam confined crack

propagation to the lower half of the

specimens for the first 20 cycles of

loading. The final crack pattern for wall

R5 is presented in figure 6. There was

considerable grinding of mortar joints and

crushing and griridLng of block units as the

blocks moved relative to each other. When

the test results from walls R5 and R6 are taken as a bond beam set and compared to the

joint reinforcing set formed by walls R2 and

R4, it is observed that the individual

contribution of each form of reinforcement

to an increase in ultimate load is about the

same. Moreover, comparable values were

obtained for deflection at ultimate load,

number of cycles at ultimate load and first

cracking load from each set.

The displacement at ultimate load was between 3/16 and 1/4 in. for all reinforced

walls except RIO (mid-height bond beam and

joint reinforcing every course). The

displacement value for wall RIO was about

1/2in. Its first cracking load level was

significantly higher than any other in this

288
series. Moreover, wall RIO had the second

highest ultimate shear stress in the series.

When comparing the ultimate shear resistance of

this wall with that of walls R6 and R8, which

had ahout the same steel content, the

significantly higher values obtained by RIO may

be attributable to the presence of the bed

joint reinforcing.

The following observations summarize the

preliminary results:

1. Small amounts of horizontal

reinforcing are effective in

increasing post-cracking strength.

2. In-plane shear strength does not

increase proportionally with

increasing amounts of reinforcing.

3. Bed joint reinforcing placed in

alternating courses is as effective

in increasing in-plane shear

strength as when placed in every

course.
TABLE 1 - TEST SPECIMEN DESCRIPTION

SPECIMEN

VT A XJTT7 NAME

ASPECT

RA110

NO. OF BOND

BhAMb

BOND BEAM

LOCA1 10JN

BOND BEAM

T) T7 TXTt? RElNr . JOINTT) 1? T KTT7 PATTERN

JOINT

Tl T7 T XT REINF.

AMOUNT

Rl 1.17 0 N/A N/A UNREINF. N/A

R2 1 . 17 0

XT V A N/A XT / A N/A EVERY

OTHER

COURSE

9 ga (5 16"

on center

R3

R4 1.17 -0- N/A N/A EVERY

COURSE

9 ga @ 8"

on center

R5 1 1

"7 1.17 1 MID-HEIGHT I - #4 T TXTT1 T? T XT'P UNREINF

.

XT / A N/A

R6 1.17 1 MID -HEIGHT 3 -#5 UNREINF. N/A

R7 1 . 17 2 2nd & 6th

courses

1 - #5 T TX7T"> T XT UNREINF. XT / A N/A

R8 1 . 17 41 MID-HEIGHT 3 - #5 T TXTH T XT UNREINF

.

XT / A N/A

R9 1.17 1 MID-HEIGHT 1 - #3 EVERY

OTHER

COURSE

9 ga @ 16"

on center

RIO 1.17 1 MID-HEIGHT 2 - #4

1 - #5

EVERY

COURSE

9 ga @ 8"

on center

Rll 1.17 2 2nd & 6th

courses

1 - #5 UNREINF. N/A

Specimen broken before testing

290
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291
T

\ 1, % % * * * * *

X X X X * * * * *

X X X X * * *f * . *

\ \ X X *******

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X X X X X

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56

« BOND BEAM

***********. XXXXXXXXXXXXXXX . X * * **** *********** xxxxxxxxxxxxxxxxx-. . *y ' A*- *

xAX

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V X **************** X X X XXXXXXXXXXXXX ******** - XXX XX XXXXXXXXXSXXXX XX X X X X X X X X X X X X X X XX '*\xxx\xxxxxxx\x.xxxxxxxxxxx < ************** * * * * * *********** x x x x x x x x x x x

MID-HEIGHT BOND BEAM

7r

56

* BOND BEAM

* BOND BEAM

BOND BEAMS AT 2ND AND 6TH COURSES

Figure 1 - Location of Bond Beams

292
44'

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Figure 3 - LST Locations

293
200(800)

150(400)

125(200)

2O0(-80O)

Figure 4 - Cyclic Loading History Used For Masonry Walls

294
30.

18. -

6. -

-S. -

18. -

) I coonOS lincit ho

30.-0.2S

FltM-O.Ot.-H.lH m.T(-o.c».-i>.t)lio)

-0.15 -0.05 0.05 0.1S

DISPLACEMENT (IN.)

35.

0.2S

21. -

'

7. - in

a

-21. -

-35.

( ):COORDS

| liCYCI.1 HO

-0.40 -0.24 -0.08 0.08 0.24

DISPLACEMENT (IN.)

0.40

UNREINFORCED WALL (Rl) NO BOND BEAM, JOINT REINF. EVERY COURSE (RA)

SO

( JiCOWtDS

I 1 1 oral no

mi(-o.ii,-M.i)iioi

-0.3S -0.21 -0.0? 0.07 0.21

DISPLACEMENT (IN.)

MID-HEIGHT BOND BEAM

NO JOINT REINFORCING (R5)

0.35 -0.65

nu(-o.i«,-M.t)lit|

-0.39 -0.13 0.13 0.39

DISPLACEMENT (IN.)

0.6S

MID-HEIGHT BOND BEAM

JOINT REINF. EVERY COURSE (RIO)

Figure 5 - Load Displacement Curves

295
Figure 6 - Crack Patterns at End of Testing

296
SPECIMEN R1

UNREINFORCED WALL

S 10 15 20 25 Cyck

SPECIMEN R4

so

20

10

NO BOND BEAM

JOINT REINF. EVERY COURSE

hIl |nl«

t inl». ill L

10 15 20 25 30 3S

Cyd«

SPECIMEN RS SPECIMEN R10

90'

MID-HEIGHT BOND BEAM

NO JOINT REINFORCING

20 .

10-

6 10 15 20 25 30 35 40

Cycto

SO

20

10

MID-HEIGHT BOND BEAM

JOINT REINFORCING EVERY COURSE

0 6 10 15 20 25 SO 35 40 4S

Cycle

Figure 7 - Energy Absorption Per Cycle

297
Inelastic Behavior of Reinforced Concrete Bridge Pier— Effect of

Bilateral Loading and Circular Hollow Section

by

Toshio Iwaskai 1

, Osamu Ueda 2

, Kazuhiko Kawashima 3

, Kinji Hasegawa

1* and Takeshi Yoshida 5

ABSTRACT

Dynamic loading tests using large specimens

of reinforced concrete bridge piers have been

conducted to study an effect of bilateral

loading and circular hollow section on the

nonlinear hysteresis behavior of reinforced

concrete bridge piers.

It was found from the tests that

deterioration of load bearing capability is

larger in the specimens subjected to the

bilateral loading as compared with the

specimen subjected to unilateral loading. It was also found that the specimen with hollow

section subjected to a series of step-wise

load reversals shows much significant failure

than the specimen with solid section due to

spalling-off of the concrete into the hollow

space as well as the spalling-off of concrete

outside the specimen.

KEYWORDS: Seismic design, reinforced concrete

( RC ) bridge piers, highway bridges, dynamic

loading tests, ductility

1 INTRODUCTION

in two horizontal directions.

On the other hand, for high RC piers, hollow

sections are often adopted for aiming to

reduce the weight of pier.

In the RC piers with hollow section, core concrete is not fully confined by the tie

bars surrounding the outside of concrete

since there is a hollow space inside the

pier, so that it is anticipated that the

hollow section has less amount of ductility

as compared with the solid section. Because

few experiments have been made for the hollow

sections, study for the nonlinear behavior of

RC piers with hollow sections seems to be

quite important for adopting the RC hollow

cross sections for high bridge piers.

From the viewpoint described above, the

bilateral loading in two horizontal

directions was conducted by using two loading

actuators. Loading tests were also conducted

for RC pier models with circular hollow

section. Effect of the bilateral loading and

hollow section on the nonlinear hysteresis

behavior of RC piers was studied.

A number of highway bridges has been damaged

in the past earthquakes. Particularly as represented by the damages caused by the

Miyagi-ken-oki Earthquake in 1978,

significant damages of the RC bridge piers

were likely to be developed by strong ground

shaking compared to the superstructure or foundation. Seismic design for highway

bridges has been made in accordance with the

seismic coefficient method in which

structural members are designed so that they

would stay within elastic range against

lateral force equivalent to the horizontal

acceleration of 0.2 G to 0.3 G. It is

important, however, to properly evaluate the

dynamic characteristics of RC piers including

energy dissipating capability.

Many studies have been made on nonlinear

hysteretic behavior of RC piers under various

loading conditions such as static, dynamic

and earthquake -like loadings. However, in

such loading tests conducted in the past, the

loading was mostly made only in one direction. The seismic force acting to RC

piers during an earthquake comprises twodirectional components. It is required to

evaluate the effect of simultaneous loading

2. TEST SPECIMENS

2.1 Specimens for Bilateral Loading

Three specimens were used for the bilateral

loading. As shown in Fig. 1 , the specimens

have the rectangular section of 80 cm x 4-0 cm, and the height of 2.U m from the base to

the loading point. Thus, the shear span

ratio defined as a ratio of the pier height

to the effective width is 6.8 and 3-2 for

weak axis and strong axis, respectively. The

main reinforcement is of eleven SD30 with a diameter of 13 mm ( longitudinal

1 Dr. of Engr. , Deputy Director-General,

Public Works Research Institute, Ministry

of Construction, Japan

2 Former Director, ' Earthquake Disaster

Prevention Department, Public Works

Research Institute, Ministry of

Construction, Japan

3 Dr. of Engr., Head, Earthquake Engineering

Division, Earthquake Disaster Prevention

Department, Public Works Research

Institute, Ministry of Construction, Japan

A Research Engneer, ditto

5 Former Assistant Research Engineer, ditto

298
reinforcement ratio of 1.0 % ). It was placed only along the long side wall with

covering thickness of 5 cm. All the axial

reinforcements are arranged continuously from

the footing to the top of the pier. Lateral

ties are of SR24 with a diameter of 9 mm and

placed at 20 cm interval, which is equal to

the half of the thickness of 40 cm in shortside direction.

High-early-strength portland cement with the

maximum grain size of aggregate of 10 mm is

adopted. Unconfined compressive strength on

the day of loading test was 311 to 4-02

kgf/cm 2

.

2.2 Specimens for Circular Hollow Section

Two specimens as shown in Fig. 2 were used to

study the effect of circular hollow section.

The specimens have the circular hollow

section with the outside diameter of 800 mm

and inside diameter of 518 mm, and the height

h is 2.475 m from the base of the pier to the

loading point. For the main reinforcement,

fifty SD30 bars were placed for both

specimens P-4.6 and P-4.7. Diameter of the

main reinforcement was varied as a parameter

to be investigated, i.e., <j> 10 mm for P-4.6

and 4> 13 mm for P-4-7. Thus, the main

reinforcement ratio is 1.2 % and 2.1 % for P4.6 and P-47 respectively. Main reinforcement

was placed continuously from the footing to

the top of pier. The lateral ties of SR24.

with <t> 9 mm were placed at the intervals of

20 cm.

High-early-strength portland cement was used

for concrete with the maximum grain size of

aggregates of 10 mm.

3. LOADING PROCEDURE

In the experiments for studying the bilateral

loading effect, the specimen was anchored to

the reaction floor with bolts as shown in

Fig. 4. Two dynamic actuators, which were installed to the reaction wall in such a manner that two loading directions became

orthogonal to each other, were fixed to the

head of the specimen. Although axial force

of about 5 to 20 kgf/cm2 acts to bridge pier

due to dead load of superstructure, it was disregarded in this test because of the

restrictions of the experimental equipment

and its limited effect.

Two loading directions were set equal to the

two principal axes of the section as shown in

Fig. 5, in which the first and second loading

directions are defined here as the direction

normal and parallel, respectively, to the

long side of the cross section. For

comparison with the bilateral loading

condition, unilateral loading was also made,

in which loading was made in the first

loading direction.

The yield displacement corresponding to one ductility factor was defined ; 6i for the

first loading direction and 62 for the second

loading direction. It was defined as the

displacement at loading point at which

reinforcing bars at the extreme tension fiber

firstly reached yield strain. When subjected

to the loading in the relevant direction, the

yield displacements 61 and 62 are 12 mm and 5 mm, respectively.

The specimens were subjected to a series of

step-wise increasing symmetric displacement

cycles as shown in Fig. 5.

In the first loading direction, the loading

amplitude was increased with the step-size of

one ductility factor such that 1 6 1

, 26 1, 36 1,

In the second loading direction, the

step-size for the loading amplitude was assumed one ductility factor ( 62, 262, 36 2 ,

• • • ) for P-4-9 and two times of ductility

factor 6 2 for P-50 ( 26 2 , 4.6 2 , 66 2 , ••)•

Phase between the first and second loading

amplitudes was shifted by 90 degrees as shown

in Fig. 6, so that hysteresis between the two

loading amplitudes becomes elliptic. Thus,

the number of loading is 10 cycles in the

first loading direction, but it is 10.5

cycles in the second loading direction. No

loading was made in the first and last 0.25

cycle in second loading directions . The

loading rate was determined in such a manner

that the velocity amplitude in the first

loading direction became constant with 10

cm/ sec.

In the experiments for studying the effect of

hollow section, the lateral force was applied

with an actuator as shown in Fig. 7.

The specimens were subjected to a series of

step-wise symmetric displacement cycles. At

each step, 10 cycles of " loading with the same displacement amplitude was carried out. The

step-size was determined by increasing the

displacement in each step by a displacement

ductility factor of one. It should be noted

here that due to malfunction of measurement

system the yield displacement 6y used in this

test was determined from the strain of main

reinforcement at 10 cm above the base

.

Therefore, the yield displacement 6y used in

this test is about 4- % larger than the

correct value.

The loading velocity was assumed as 25

cm/sec. The axial force equivalent to the

dead load of the superstructure was also

disregarded in the test.

299
4. EFFECT OF BILATERAL LOADING

4..1 Failure Mode

Failure mode after step-wise loading is shown

in Figs. 8, 9 and 10, in which cracks,

spalling-off of concrete and buckling of main

reinforcement normal to the first loading

direction are presented. When flexural

failure occurs at the base of the pier

subjected to unilateral loading, the failure

normally progresses in the order of

a) horizontal cracks in concrete at the base,

b) spalling-off of concrete near the base,

c) outward buckling of main reinforcement, and

d) rupture of main reinforcement.

Even in the bilateral loading the same progress of the failure basically occurs. In

the unilateral loading, the spalling-off of

concrete occurs at a time on the whole plane

orthogonal to the first loading direction.

In the bilateral loading the spalling-off

begins from a corner and then spreads to the

whole plane orthogonal to the first loading

direction as the loading progresses.

Table 1 shows the progress of damages in

terms of loading displacement at which

spalling-off of concrete occurs and the

number of main reinforcement ruptured. The

spalling-off of concrete begins from 46

1

loading in the unilateral loading. In the

bilateral loading, the spalling-off begins

from 4-6 1 and 36 i loading in specimen P-4-9 and

P-50, respectively.

As to the rupture of main reinforcement, the

rupture begins from 66 i loading in the

unilateral loading. Four reinforcements are ruptured at this loading step. On the other

hand in P-4-9, the rupture begins at 66

1

loading, in which 6 reinforcements are ruptured. In specimen P-50, the rupture

begins from the 4.6 1 loading in which one bar

is ruptured, and 7 bars are ruptured at the

next loading step.

From such evidence, it can be said that the

damages generally progress more promptly in

bilateral loading compared to unilateral

loading. Number of main reinforcement

ruptured significantly depends on loading

pattern. This is certainly developed because

the bars, . specially at the corner, are subjected to alternation of larger tension

and compression in the bilateral loading than

in the unilateral loading.

4.. 2 Strength and Ductility

Fig. 11 shows hysteresis loops between

loading force and lateral displacement in the

first loading direction. It is seen in Fig.

11 that the hysteresis curves of P-49 and P50 have considerably round shapes compared to

the unilateral loading.

Fig. 12 shows the comparison of the envelope

of the hysteresis loops in the first loading

direction.

It can be seen in Fig. 12 that envelope of'

the hysteresis loop of P-49 is almost

identical with that of P-37, while the

envelope of the hysteresis loop of P-50 is

much smaller than that of P-37. It seems that as long as the ratio between the loading

displacement in the first and second

directions is 6 2 /6i, the effect of bilateral

loading is less significant. On the other

hand, when the ratio is doubled to 262/61,

the bilateral loading effect becomes

considerable.

Fig. 13 compares the hysteresis loops in the

second loading direction. In both specimens,

a projected loop can be observed at the

beginning. This corresponds to the loading

at the first 0.25 cycle in which loading was

made only in the second loading direction.

Table 2 summarizes the yield load Py, the

maximum load Pu, yield displacement 6y and

ultimate displacement 6u. Only averaged

values developed between in positive and in

negative loading directions are presented in

Table 2.

Although the ductility factor is 17 % smaller

in P-4-9 than that in P-37, the yield load Py

and the maximum load Pu are almost the same between P-49 and P-37. However, the

ductility factor and the maximum load Pu of

P-50 are 25 % and 12 % smaller than those of

P-37. Thus, the bilateral loading effect is

significant for P-50.

4.-3 Energy Dissipating Capability

The hysteresis loop generally has a curve presented in Fig. 14., and the area surrounded

by the hysteresis loop, represents the energy

absorbed by the pier during one loading

reversal. Fig. 15 compares the amounts of

energy absorbed during load reversals due to

deformation in the first loading direction.

Energy absorption increases up to 361 loading

during which difference between the three

specimens is less significant. Over 361

loading, the energy absorption of the

specimen P-50 significantly decreases due to

rupture of the main reinforcements, while the

energy absorption of the specimen P-49 takes

the maximum at 4.61 loading and then decreases

thereafter.

Energy absorption capability was also studied

in terms of an equivalent hysteretic damping

ratio he which is defined as

300
, i aw ;_ /

1 \

in which, W and AW represents an elastic

energy and an energy dissipated dixring one load reversal, respectively, as defined in

Fig. U.

It is interesting to note that the hysteretic

damping ratio is larger in P-50 than P-49» even for the loading displacement up to 36 1.

This is due to the hysteresis loop with round

corner.

5. EFFECT OF HOLLOW SECTION

5.1 Failure Mode

The failure mode on the plane parallel to the

loading direction after each load reversal is

shown in Figs. 18 and 19. Flexural failure

occurred in the three specimens at their

bases. Because the main reinforcement ratio

is 0.9 % higher in P-47 than in P-46 by the

tie reinforcement ratio being the same, the

failure at the base is much significant in P47 than in P-4.6, i.e., after load reversal

with 5cy> concrete at the base is completely

crushed and distortion similar with the shear

failure occurred. Three and six main

reinforcements were ruptured during 66y and

76y load reversals, respectively, in P-46>

while no rupture of the main reinforcement

was developed in P-47 up to 66y loading.

This clearly shows that the deformation

developed at the base of P-47 is close to the

shear failure rather than the flexural

failure

.

Spalling-off of concrete into the hollow

space occurred from 76y loading in P-47 and

from 5<5y loading in P-48. This is greatly

different from the failure mode developed in

P-29 with solid cross section.

5.2 Strength and Ductility

Fig. 20 shows the hysteresis loops between

loading force and displacement of pier. Fig.

21 compares the envelopes of the hysteresis

loops presented in Fig. 20.

In accordance with the difference of the

failure mode described in the preceding

section, there is a significant difference in

the envelope of the, hysteresis loop. In the

specimen P-29, the capability for supporting

the lateral force decreases gradually as the

number of main reinforcement ruptured

increases. Even in the specimen P-46,

deteiioration of the load in accordance with

the increase of loading displacement is not

necessarily significant. However, in the

specimen P-47, the load significantly

decreases after reaching the maximum

strength. In the hollow circular section,

the concrete is confined at the outside by

the main reinforcement as well as lateral

ties but not confined at the inside. This is

considered to be related to the sudden

decrease in the strength due to the spallingoff of concrete into the hollow space.

5.3 Energy Dissipating Capability

Fig. 22 shows the equivalent hysteretic

damping ratio defined by Eq . ( 1 ) . It should

be noted here that the yield displacement is

not identical between the hollow and solid

sections. It is seen in Fig. 22 that

decreasing rate of the damping ratio after

reaching the maximum value is much larger in

hollow section than in solid section.

6. CONCLUSIONS

In order to study the effect of the bilateral

loading and of hollow section on the

nonlinear hysteretic behavior of RC piers, a series of dynamic loading tests was conducted. From the results presented

herein, the following conclusions may b^

deduced:

(1) Effect of bilateral loading

i) When the specimens are subjected to a series of step-wise increasing symmetrical

bilateral displacement cycles with the step

size of one ductility factor for the relevant

directions, progress of the failure and

deterioration of load bearing capability is

approximately identical with the specimen

subjected to the unilateral load reversals.

However, the step size in the second

direction is doubled ( 25 2 /6i ), the progress

of failure and deterioration of load bearing

capability is substantially decreased as compared with the specimen subjected to the

unilateral load reversals.

ii) Amount of the energy dissipation

during load reversals is smaller in the

specimen subjected to the bilateral loading,

in particular, ratio between the loading

displacement in the first and second

directions becomes 262/61.

(2) Effect of hollow section

i) In the circular hollow section, the

concrete is confined at the outer side by

main reinforcement and lateral ties.

However, because there is an empty space

inside, spalling-off of concrete into the

inner space develops.

ii) In consequence, especially in a specimen having a high main reinforcement

301
ratio, the load bearing capability suddenly

decreases after the spalling-off of the inner

concrete. The final failure occurs without

the rupture of the main reinforcement.

iii) Decreasing rate of the equivalent

damping ratio after reaching the maximum

value is much larger in hollow section than

in solid section.

REFERENCES

1) Iwasaki, T. , Kawashima, K. , Hagiwara, R.

,

Hasegawa, K. , Koyama, T. and Yoshida, T. : Experimental 'Investigation on Hysteretic

Behavior of Reinforced Concrete Bridge Pier

Columns, 17th Joint Meeting, Wind and Seismic

Effects, UJNR, 1985.

2) Iwasaki, T. , Kawashima, K. , Hasegawa, K.

,

Koyama, T. and Yoshida, T. : Effect of Number

of Loading Cycles and Loading Velocity of

Reinforced Concrete Bridge Piers, 19th Joint

Meeting, Wind and Seismic Effects, UJNR,

1987.

3) Iwasaki, T. , Ueda, 0., Kawashima, K.

,

Hasegawa, K. and Yoshida, T. : Dynamic

Strength and Ductility of Reinforced Concrete

Bridge Piers, 20th Joint Meeting, Wind and

Seismic Effects, UJNR, 1988.

4.) Kawashima, K. and Koyama, T.: Effect of

Number of Loading Cycles on Dynamic

Characteristics of Reinforced Concrete

Bridge Pier Columns, Structural

Eng. /Earthquake Engr., Proc. JSCE, Vol. 5,

No. 1, pp. 183 - 191, 1988, /,. 5) Kawashima, K. and Koyama, T. : Effect of

Cyclic Loading Hysteresis on Dynamic

Behavior of Reinforced Concrete Bridge

Pier, Structural Eng. /Earthquake Eng., Proc.

JSCE, Vol. 5, No. 2, pp. 343 - 350, 1988,

10.

302
Table 1 Comparison of Failure Modes

Loading Type Unilateral Loading Bilateral Loading

Spec i men P - 3 7 P - 4 9 P - 5 0

Initiation of Spa 1 ling of

Cover Concrete

Step 4 Step 4 Step 3

Number of Main Reinforcement

Cut during Loading

Stepl~ 3 0 0 0

Step 4 0 0 1

Step 5 0 0 7

Step 6 4 6

Table 2 Comparison of Load and Displacement Performance

Loading Type Un i lateral Load Lag Bilateral Loading

Spec imen P - 3 7 P - 4 9 P - 5 0

Load

Yield Load P y (tf) 7 . 1 7 . 2 5. 9

Maximum Load Pu (tf) 8 . 4 8 . 1 7 . 4

Pu/Py 1 . 18 1 . 13 1 . 2 5

Displacement

Yield Displacement

<5 n (cm)

0. 9 1 1 . 0 4 0 . 8 8

Ultimate Displacement

6 v (cm)

5 . 7 5 . 4 4 . 1

{ »/J « 6 . 3 5 . 2 4 . 7

303
Fig.1 Test Specimens for Bilateral Loading

Fig. 2 Test Specimens with Circular Hollow Section

304
Second

Loading

Direction

-nS^" nSi

First Looding

Direction

to) Unilateral Loading in P~37

Second

Looding

Direction

nSz

-nSi

-n8z 3>nSi

First Loading

Direction

(b) Bilateral Loading in P-49

nS, -

-nS,

nSi -

-nSi

nS? -

-nS? -

)0 Cycles

First Loading

Direction

(a) Unilateral Loading in P-37

10 Cycles

First Loading

Direction

Second

Looding

Direction

10.5 Cycles WWVWvW

(b) Bilateral Loading in P-49

Second

Loading

Direction

2nS?

-n8A

r

-2nS2

First Loading

Direction

Ic ) Bilateral Loading in P-50

nSi

-n8t

2nSj

-2n8 2

(0 Cycles

10 .5 Cycles

First Loading

Direction

Second

Looding

Direction

(c) Bilateral Loading in P-50

Fig. 5 Loading Locus in Unilateral

and Bilateral Loading

Fig. 6 Loading Hysteresis in First

and Second Direction

Dynamic ^ Actuator ,

Direction of Excitation

Loading Jig/' o

3£

o

Footing

Reaction Floor

Fig. 7 Experimental Set-up in Circular Hollow Section

305
---

- -

-- —

-

—

mwm

—

—

SBm

(a) step 1 (b) step 2 (c) step 3 (d) step 4. (e) step 5 (f) step 6

Fig. 8 Failure Mode for Specimen with Unilateral Loading (P-37)

Di—

p

— - i

or-

~ =» *

/ I fig

(a) step 1 (b) step 2 (c) step 3 (d) step 4 ( e ) step 5 (f) step 6

Fig. 9 Failure Mode for Specimen with Bilateral Loading (P-4.9)

SB magi

(a) step 1 (b) step 2 (c) step 3 (d) step i (e) step 5

Fig. 10 failure Mode for Specimen with Bilateral Loading (P-50)

306
6 _ 4 w 0 a

S

"«

- 4

- 6 1 r*r *r'^«--^—

-KX) -80 -60 -40 -20 0 20 40 60 DISPLACEMENT (MM)

-60 -40 -20 0 20 40 60 80 100 DISPLACEMENT (MM)

-60 -40 -20 0 20 40 60 80 100 DISPLACEMENT (MM)

(a) Unilateral Loading (P-37) (b) Bilateral Loading (P-49) (c) Bilateral Loading (P-50)

Fig. 11 Comparison of Hysteresis Loops in First Direction

_ 4

B

2 — o a

< - 2 o

4

- 6

-10

1 \ u\ :

. L

: x f—i ! LEGEND —UreUrtiral Lontng 1P-37I

.

—BiloHfol Loodinj IP-491 —e*at«rol Looding (P-50)

'

-CO -80 -60 -40 -20 0 20 40 60 80 100 DISPLACEMENT (MM)

-36 -24 -12 0 12 24 36 48 60 DISPLACEMENT (MM)

-60 -48 - 36 -24 -12 0 12 24 36 48 60 DISPLACEMENT (MM)

(a) Bilateral Loading (P-4-9) (b) Bilateral Loading (P-50)

Fig. 12 Comparison of Envelope

of Hysteresis Loops

Fig. 13 Comparison of Hysteresis Loops

in Second Direction

FORCE P

DISPLACEMENT

8

Fig.H Definition of Equivalent

Hysteretic Damping Ratio

-i 1 r- LEGEND

Unilateral Loading (P- 37)

Bilateral Looding (P- 49)

Bilateral Loading (P- 50)

2 4 6 8 DISPLACEMENT fx 80

Fig. 15 Comparison of Energy

Dissipation

j i i i_

LEGEND

Unilateral Looding (P-37)

Bilateral Loading (P-49)

Bilateral Loading (P-50I

2 4 6 DISPLACEMENT (x8i)

Fig.16 Comparison of Equivalent

Hysteretic Damping Ratio

307
I

I

(a) step 1 (b) step 2 ( c ) step 3 (d) step U (e) step 5 (f) step 6 (g) step 7 (h) step 8

Fig. 17 Failure Mode for Specimen with Circular Section (P-29)

(a) step 1 (b) step 2 (c) step 3 (d) step A (e) step 5 (f) step 6 (g) step 7 (h) step 8

Fig. 18 Failure Mode for Specimen with Circular Hollow Section (P-4-6)

(a) step 1 (b) step 2 (c) step 3 (d) step U (e) step 5 (f ) step 6

Fig. 19 Failure Mode for Specimen with Circular Hollow Section (P-4-7)

308
•100 -80 -60 -40 -20 0 20 40 60 80 100 DISPLACEMENT [MM)

(a) Circular Section

(P-29)

Fig. 20

-100 -80 -60 -40 -20 0 20 40 60 80 100 DISPLACEMENT IMM)

(b) Circular Hollow Section

(P-46)

-100 -80 -60 -40 -20 0 20 40 60 80 100 DISPLACEMENT (MM)

(c) Circular Hollow Section

(P-47)

Comparison of Hysteresis Loops

I P-29)

ion (P-46)

ion (P-47)

DISPLACEMENT [lit] DISPLACEMENT C«8|]

Fig. 21 Comparison of Envelope Fig. 22 Comparison of Equivalent

of Hysteresis Loops Hysteretic Damping Ratio

309
U.S. Coordinated Program for Masonry Building

Research— A Fourth Year Status

by

James L. Noland1

ABSTRACT

The U.S. Coordinated Program for

Masonry Building Research is

comprised of 28 specific research

tasks. The primary purpose is to

support development of a limit state

design standard for masonry

buildings. The. U.S. program has, in

turn, been coordinated with a parallel program in Japan. Overall,

the U.S. program is approximately 50%

complete. Several individual U.S.

research Tasks are complete and

others well underway. Many technical

reports have been published and

significant technical accomplishments

made .

KEYWORDS: Reinforced masonry;

TCCMAR; Limit state design; status.

1 . INTRODUCTION

1.1 Overview

The U.S. Coordinated Program for

Masonry Building Research is an

integrated set of specific research

tasks being executed by the U.S.

Technical Coordinating Committee for

Masonry Research (TCCMAR/U. S . ) . The

U.S. Program, in turn, is coordinated

with a parallel program conducted in

Japan to exchange information and

concepts for the mutual benefit of

both countries.

The U.S. program addresses structural

issues and is intended to provide

design & criteria recommendations

necessary to support limit state

design techniques for reinforced

masonry. The program consists of

both analytical and experimental

research tasks and is organized on a project basis as shown in Figure 1.

The individual research tasks are

coordinated and time-phased, as shown

in Figures 2 and 3, to avoid

duplication of effort and to transfer

information between the task

Atkinson-Noland & Associates

Boulder, CO 80302, U.S.A.

researchers on a timely basis.

Research tasks which comprise the

U.S. Coordinated Program are listed

in Table 1 with a brief statement of

purpose for each. This paper is an

update of the previous status report

(1) and is based upon a detailed

status report of November 1988 (2)

.

1 . 2 Current Status

By May 1989 the U.S. program will

have been under way for approximately

four years. This paper briefly

reviews the activities, progress, and

accomplishments of the program during

that period.

2. ADMINISTRATIVE CHANGES

Dr. H.S. Lew of the National

Institute of Standards and Technology

(USA) has replaced Dr. Scribner as

TCCMAR/U. S.-UJNR Co-Liaison. Refer

to Figure 1

.

Gregg Borchelt, Director of

Engineering, Brick Institute of

America has replaced G. Dalrymple on

the Industry Participation Panel and

Ro Whitlock on the Industry Observers

Panel. Refer to Figure 1.

3 . PROGRESS

3 . 1 Individual Task Status

The status of individual research

Tasks in terms of percent completion

is listed in Table 1. A more

complete description of the progress

and status of each task may be found

in reference (2) . The completion

percentages are based upon the total

time period of each task as opposed

to the time period of Fall 1985 to

December 31, 1987 used in the

previous report (1).

3 . 2 Reports

The number of technical reports

published by TCCMAR/U . S . researchers

continues to increase. Arrangements

have been made so that the reports

are available to anyone from the

Earthquake Engineering Library in

310
Berkeley. The reports available are

listed in Table 2.

A large number of technical papers

have been presented and/or published

in various meetings and conferences

including JTCCMAR meetings. These

are listed in reference (2).

3.3 Design Standards Development

A critical issue in the U.S. is, and

has been the transfer of technology

developed by research into practice.

In the U.S. an important factor in

implementing research into practice

is the incorporation or provision for

new technology by the design

standards

.

In the U.S., design standards are

developed by non-governmental

professional groups using various

forms of consensus processes. Hence,

the NSF-supported U.S. Coordinated

Program for Masonry Building Research

cannot develop standards, but will

develop recommendations (Task 10).

Discussions and negotiations over the

last few months have served to

establish mechanisms and processes

for transfer of technology (i.e.

research data and recommendations) to

a consensus design standards

development committee sponsored by

U.S. professional societies. A oneyear pre-standardization effort to

collect and document background

information pertinent to limit state

structural design and to develop a draft limit state design standard for

masonry has been organized. The prestandardization work along with the

technology developed by the

coordinated masonry research will be

utilized by the national masonry

standards committee to produce a limit state design standard for

masonry. The basics of the process

are illustrated by Figure 4.

4. TECHNICAL ACCOMPLISHMENTS

The U.S. program is proceeding

generally according to expectations.

At this point in time, which is about

the half-way point, technical

accomplishments are beginning to

become apparent. It must be

recognized, however, that the major

technical results are yet to come.

Because the primary objective of the

U.S. program is to provide a basis

for limit state design and an

associated generic modeling

capability, the following seem

significant among the results

achieved to date:

a. A finite element model (FEM) has

been formulated and implemented

capable of post-peak strength

response of reinforced shear

walls. The material models have

the capability to represent the

reduction in peak compressive

strength due to tensile cracking

and allow reorientation of

initial cracks to a final crack

state. The FEM is formulated

such that Gauss points are

essentially independent thereby

allowing use of a coarser mesh in

modeling. Correlation with

experimental results has been

very good. Work will continue in

this area to verify correlation

with experimental results.

b. Structural component models (SCM)

can now predict yield and maximum

force levels for in-plane

flexural reinforced masonry

walls. Correlations with

experimental results are very

good. Studies have led to the

development of capacity reduction

factors for in-plane flexural

masonry walls. Work will

continue to develop a family of

SCM's for system analyses.

c. A non-linear lumped parameter

model (LPM) has been developed to

do time-history analyses of

masonry structures. A new nonlinear "spring" element has been

introduced to represent masonry

shear walls in building system

models. Studies have been done

to understand the dynamic

behavior of masonry structures

under ground motions with

different v/a ratios. The LPM

has been extended and used to

provide understanding of the

dynamic behavior of masonry

structures allowed to uplift.

311
For future design applications

and studies, U.S. ground motions

have been categorized with

respect to v/a rates.

d. A database has been developed for

the cyclic behavior of singlestory reinforced masonry shear

walls with different in-plane

compressive loads and

reinforcement rates. Limit

states have been identified for

application in the design

recommendations to be made.

e. Tests of two-story coupled

masonry shear walls have revealed

the need for high accuracy in the

modeling of the coupling beams.

Limit states for this type of

system were identified.

f. Force-deflection characteristics

of prestressed precast concrete

plank floor diaphragms have been

established experimentally both

parallel and perpendicular to

plank direction. Limit states

have been identified. Data has

been accumulated for other floor

systems as well. This

information will be the basis of

spring elements for use in the

lumped parameter models.

g. Reinforced concrete and clay

masonry walls responded

elastically when subjected to

out-of-plane dynamic loads

typical of design ground motion

in soil type 1 in the highest

U.S. seismic zones. The walls

had reinforcement ratios of 0.07

balanced to 0.5 balanced,

vertical loads of 0 to 80 psi

(36.3kg) and slenderness ratios

of 40 to 50. Under design

motions, maximum lateral

displacements were about 6 inches

(2.4 cm). For higher levels of

ground motion, the walls

sustained vertical loads, but

some permanent lateral deflection

was noted.

h. Tests of grouted hollow clay and

concrete unit prisms in

compression indicated that

masonry is a generic material,

i.e., that no unusual behavioral

traits are present to separate

grouted clay and concrete

masonry. The compressive tests

established stress-strain curves

including the post-peak

descending branch and limit state

values. Tests of prisms loaded

in eccentric compression

established strain gradient

effects and Kl , K2 , K3 stress

block values for strength design.

5 . RE S OLUTIONS OF THE FOURTH JTCCMAR

MEETING

1. The fourth meeting of JTCCMAR

was held in San Diego,

California, U.S.A. on October

17, 18, and 19, 1988. The

meeting was attended by 18

researchers from the United

States, 16 researchers from

Japan, 7 observers from the

United States and 10 observers

from Japan.

2. There was a successful exchange

of information and ideas between

the researchers from both

countries on the subjects of

masonry material behavior,

behavior of assemblies, behavior

of systems, testing procedures,

analytical/modeling methods,

results of the Japanese tests on

a 5~story building, seismic

design procedures and future

work

.

3. It was recognized that, while

progress has been made in

developing an understanding of

the behavior of masonry, there

is still an urgent need for

additional information. The

participants therefore recommend

that sponsoring agencies take

appropriate measures to continue

support for masonry research.

4. The participants recognize that

both countries have many common

research needs regarding masonry

construction as well as

mechanical properties of masonry

units, components, and systems.

5. It was recognized that

continuous and in-depth effort

is required to establish the

312
seismic performance

characteristics of masonry

structures and to develop

improved construction and design

procedures. However, the

participants realize that the

research programs of both

countries primarily address

central issues. They recognize

that, while the knowledge

developed will provide a basic

foundation for improved masonry

technology, many other important

issues exist which should be

addressed

.

6. Significant progress has been

made by both sides in testing

and modeling masonry materials,

components, subassemblages , and

prototypes which will lead to a better understanding of masonry

structural behavior and to the

development of adequate seismic

design procedures for masonry

buildings

.

7 . The participants recognized that

the full-scale test carried out

in japan has and will make a great contribution to the

understanding of the overall

behavior and performance of

masonry structural systems.

They recommend that another

full-scale test on a masonry

structural system be carried out

in the United States. It is

also recommended that

information obtained from both

full-scale tests be exchanged.

8. The technical communication

which has been carried out by

sub-groups and individuals of

TCCMAR/U.S. and TCCMAR/ Japan has

been valuable and should be

continued. Researchers should

be exchanged as required to

enable a meaningful exchange of

concepts and information between

individuals and groups with

similar research interest.

9. The participants from both

countries agree that proper

seismic performance of masonry

structures requires

consideration of damage control

and minimization of life safety

threats

.

10. It was recognized that the joint

project will result in an

improved technological base for

the design of RM buildings.

11. It was recognized that the

results of this coordinated

research program could also be

beneficial to other countries

which have a seismic hazard,

especially developing countries.

12. Considering that the U.S.

program is planned to continue

until 1993 and considering the

benefits of the mutual exchange

between the U.S. and Japanese

programs, it is hoped that the

joint relationship be continued

so the mutual benefits may be

maximized

.

13. The participants realized that a reinforced masonry building

association should be

established in Japan as soon as

possible to promote, by

continuous effort, the

development and promulgation of

a more rational design procedure

for RM buildings, including the

low-rise category, so that the

joint U.S. -Japan relationship

may be continued.

14. It was recognized, by both

sides, that new concepts in

masonry technology should be

investigated to enhance economy

and reliability.

15. The fifth meeting of JTCCMAR

will be held in Japan in the

fall of 1989.

6 . ACKNOWLEDGEMENTS

Primary financial support for the

U.S. side of the Joint program has

been provided by the National Science

Foundation. Dr. A.J. Eggenberger is

the cognizant program manager. The

support of the U.S. masonry industry

is gratefully acknowledged.

The benefits of the U.S. -Japan

cooperation are recognized. The

313
generous sharing of data and

experience by TECCMAR/ Japan has been

most welcome

.

7 . REFERENCES

1 . Proceedings : 20th Joint Meeting

of the U.S. -Japan Cooperative

Program in Natural Resources - Panel on Wind and Seismic

Effects, N.J. Raufaste, ed .

,

NIST, January 1989.

2. Status Report: U.S. Coordinated

Program for Masonry Building

Research' TCCMAR/U . S . , November

1988.
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318
TABLE 1 - RESEARCH TASKS COMPRISING THE U.S. COORDINATED PROGRAM

FOR MASONRY BUILDING RESEARCH

Category Task % Title - Purpose

(Researcher) Complete

1.0 1.1

(Atkinson)

100 Preliminary Material Studies — To

compare the behavior of clay and

concrete unit masonry. To provide a basis for selection of the type or types

of masonry to be used in subsequent

tasks. To establish standardized

materials test procedures for all the

experimental tasks.

1.0 1.2

(Hamid)

( Brown)

95 Material Models -- To measure the

parameters required for development of

the flexural compression stress-block.

To determine uniaxial and biaxial

material properties for analytical

models (Tasks 2.1 and 2.2) including

post-peak behavior. To evaluate nonisotropic behavior.

1.0 1.3

(Atkinson!

-0- Material Tests -- To critically review

and assess existing tests of masonry

materials and assemblages to determine

the usefulness of data produced with

respect to the needs of analytical

models and design methodology developed

in the program. To revise existing

tests as required and/or suggest new

tests. The work will be done in

coordination with Category 2 and 10

Tasks to establish accuracy

requirements

.

2.0

2 . 0

2.1

(Englekirk)

2.2

(Ewing)

60 Force-Displacement Models for Masonry

Components -- To develop forcedisplacement mathematical models which

accurately characterize reinforced

masonry components under cyclic loading

to permit pretest predictions of

experimental results. To develop models

suitable for parameter studies and

models suitable for design engineering.

65 Strain Analysis Model for Masonry

Components -- To develop a strain model

for reinforced masonry components in

conjunction with Task 2.1. To identify

regions of large strain thus assisting

in experimental instrumentation

planning. To develop a simplified model

to be used to provide data for strength

design rules and in-plane shear design

procedures

.

319
TABLE 1

Category Task % Title-Purpose

(Researcher) Complete

2.0 2.3 65

(Kariotis

)

2.0 2.4(a) -0-

(Porter

)

2.0 2.4(b) 15

(Mayes

)

( Sveinsson

)

3.0 3.1(a) 90

( Shing)

3.0 3.1(b) 10

(Hegemier

)

(Seible)

Dynamic Response of Masonry Buildings -- To develop a generalized dynamic

response model to predict interstory

displacements using specified time

histories. To correlate forcedisplacement models and to investigate

force-displacement characteristics of

structural components in the nearelastic and inelastic displacement

range. To provide data for building

test planning.

Dynamic Response of Diaphragms — To

develop an analytical non-linear model

of load displacement history of

horizontal diaphragms. To provide

associated displacements and stiffnesses

for an integrated dynamic spring model.

This Task will provide a computer model

extension using a lumped-parameter mass

parameter spring of the experimental

data collected from Tasks 5.1 and 5.2.

This work will provide input for Task

2.3.

Response of Reinforced Masonry StoryHeight Walls to Fully Reversed In-Plane

Lateral Loads -- To experimentally

establish the behavior of story-height

walls subjected to small and large

amplitude axial force, and bending

moments considering various

reinforcement ratios and patterns.

Development of a Sequential Displacement

Analytical and Experimental Methodology

for the Response of Multi-Story Walls to

In-Plane Loads — To develop a reliable

test methodology for investigating

structural response, through integrated

analytical and experimental studies of

three-story reinforced hollow unit

masonry walls. The methodology will be

the basis of studying the response of a full-scale masonry research building in

Task 9.4. To develop analytic models in

conjunction with Tasks 2.1, 2.2, and

2.3.

Dynamic Out-of-Plane Response of

Reinforced Masonry Walls — To develop

analytical models based upon the results

of Task 3.2(b) which can be used to

predict out-of-plane response of masonry

walls of various shapes, sizes, and

internal construction. To conduct

response studies based on independent

variation of parameters. The models

will interface with the models of Tasks

2 . 3 and 2.4(a)

.

320
TABLE 1

Category Task % Title-Purpose

(Researcher) Complete

3.0 3.1(c)

( Klingner

)

40

3.0 3 . 2

(Hamid)

(Mayes

)

(Harris

)

95

3.0 3.2

( Adham)

(Mayes

)

100

4 . 0 4 .

1

( Priestley)

20

Response of Reinforced Masonry Two-Story

Walls to Fully Reversed In-Plane Lateral

Loads -- To establish the behavior of

two-story walls subjected to small and

large amplitude reversals of in-plane

lateral deflections, axial force and

bending moments considering the effect

of openings, floor-wall joint details,

reinforcement ratios and coupling

between shear walls. To develop

analytical models in conjunction with

Tasks 2.1, 2.2, and 2.3.

Response of Reinforced Masonry Walls to

Out-of-Plane Static Loads -- To verify

the behavior of flexural models

developed using material models, to

evaluate the influence of unit

properties, bond type and reinforcement

ratios upon wall behavior. To provide

stiffness data for correlation with

dynamic wall test results [ (Task

3.2 (b) ]

.

Response of Reinforced Masonry Walls to

Out-of-Plane Dynamic Excitation -- To

experimentally determine effects of

slenderness, reinforcement amounts and

ratios, vertical load and grouting on

dynamic response as needed for

mathematical response models and the

development design coefficients for

equivalent static load methods.

Response of Flanged Masonry Shear Walls

to Dynamic Excitation -- To

experimentally investigate the dynamic

behavior of flanged shear walls, in

particular, the behavior of T-section

walls and the significance of dynamic,

as opposed to static or quasi-static

testing, for in-plane loading. To

develop analytical models to investigate

the flange-web shear lag phenomena, and

to identify the interaction between

flange width, height, reinforcement

content, and ductility level. (In

conjunction with Tasks 2.1, 2.2, and

2.3) .

4 . 0 4.2

( Hegemier

)

80 Floor-to-Wall Intersections of Masonry

Buildings — To determine the

effectiveness of intersection details to

connect masonry wall components. To

construct a nonphenomenological

analytical model of intersection

behavior for use in building system

models .

321
TABLE 1

Category Task % Title-Purpose

(Researcher) Complete

5.0 5.1

(Porter)

85 Concrete Plank Diaphragm Characteristics

-- To investigate experimentally

concrete plank diaphragm floor

diaphragms with stiff supports to

determine modes of failure and stiffness

characteristics including yielding

capacity in terms of distortion as needed for masonry building models.

5.0 5 . 2

(Johnson)

(Porter)

85 Assembly of Existing Diaphragm Data -- To assemble extensive existing

experimental data on various types of

floor diaphragms, to reduce to a form

required for static and dynamic analysis

models

.

5.0 5.3

( Porter)

-0- Concrete Plank Diaphragm Characteristics

Continuation -- To investigate

experimentally the behavior of concrete

plank floor diaphragms with flexible

supports to determine modes of failure

and stiffness characteristics including

yielding capacity in terms of distortion

as needed for masonry building models.

6.0

6.0

7.0

8 . 0

6 .

1

(TBD)

6 . 2

(Tulin)

7 .

1

( Abrams

)

8 .

1

(Hart)

-0- Grouting procedures for Hollow Unit

Masonry -- To identify methods of

grouting hollow unit masonry such that

the cavity is solidly filled and

reinforcement is completely bonded.

100 Reinforcement Bond and Splices in

Grouted Hollow Unit Masonry — To

develop data and behavioral models on

the bond strength and slip

characteristics of deformed bars and lap

splices in grouted hollow unit masonry,

as needed for building modeling.

Small Scale Models — To provide

experimental test data on the dynamic

behavior of three-story reinforced

concrete masonry buildings built with

1/4 scale hollow concrete units. To

demonstrate the viability of

constructing and dynamically testing

reduced scale building system models for

basic behavior studies.

95 Limit State Design Methodology for

Reinforced Masonry -- To select an

appropriate limit state design

methodology for masonry. To select and

document a procedure to compute

numerical values for strength reduction

factors. To review program experimental

research tasks to assure that

statistical benefits are maximized and

proper limit states are investigated.

322
TABLE 1

Category Task % Title-Purpose

(Researcher) Complete

8 . 0 8 . 2

(Hart)

15 Numerical Reliability Indices -- To

develop numerical values of

statistically-based strength reduction

factors using program experimentally

developed data, other applicable data,

and judgment. To complete development

of the methodology.

9.0 9 .

1

( Kariotis

)

100 Design of Reinforced Masonry Research

Building - Phase 1 -- To develop the

preliminary designs of the potential

research buildings which reflect a significant portion of modern U.S.

masonry construction. to select a single configuration in consultation

with TCCMAR which will be used as a basis for defining equipment and other

laboratory facilities using methods

developed in Category 2 tasks and the

associated load magnitudes and

distributions.

Phase 2 - To prepare final drawings and

specifications for construction of the

five-story test specimen.

9.0 9.2

( Hegemier

)

(Seible)

75 Facility Preparation -- Define, acquire,

install and check-out equipment required

for experiments on a full-scale fivestory reinforced masonry research

building

.

9.0 9.3

(Hegemier

)

(Seible)

(Priestley)

-0- Full Scale Masonry Research Building

Test Plan -- To develop a detailed and

comprehensive plan for conducting static

load-reversal tests on a full-scale

five-story reinforced masonry research

building

.

9.0 9.4

(Hegemier)

(Seible)

( Priestley)

Full Scale Test -- To conduct

experiments on a full-scale five-story

reinforced masonry research building in

accordance with the test plan and

acquiring data indicated. To observe

building response and adjust test

procedures and data measurements as

required to establish building behavior.

10.0 10.1

(Noland)

-0- Design Recommendations and Criteria

Development To develop and document

recommendations for the design of

reinforced masonry building subject to

seismic excitation in a manner conducive

to design office utilization. To

develop and document corresponding

recommendations for masonry structural

code provisions.

323
TABLE 1

Category Task % Title-Purpose

(Researcher) Complete

11.0 11.1 60 Coordination — To fully coordinate the

(Noland) U.S. research tasks. To enhance data

transfer among researchers and timely

completion of tasks. To schedule and

organize TCCMAR and Executive Panel

meetings. To establish additional

program policies as the need arises. To

stimulate release of progress reports

and dissemination of results. To

coordinate with industry for the

purposes of informing industry and

arranging industry support. To

interface with NSF and UJNR on overall

funding and policy matters.

324
TABLE 2 - RESEARCH TASK REPORTS

Task No. Author/s - Title

1.1-1: Atkinson and Kingsley, Comparison of the Behavior of Clay & Concrete

Masonry in .Compression, September 198 5.

I. 2(b)-l: Young, J.M., Brown, R.H., Compressive Stress Distribution of Grouted

Hollow Clay Masonry Under Strain Gradient, May 1988.

2.1-1: Hart, G and Basharkhah, M. , Slender Wall Structural Engineering

Analysis Computer Program (Shwall, Version 1.01), September 1987.

2.1-2: Hart, G. and Basharkhah, M., Shear Wall Structural Engineering

Analysis Computer Program (Shwall, Version 1.01), September 1987.

2.1-3: Nakaki , D. & Hart, G., Uplifting Response of Structures Subjected to

Earthquake Motions, August 1987.

2.1-4: Hart, G., Sajjad, N., and Basharkhah, M. , Inelastic Column Analysis

Computer Program ( INCAP , Version 1.01), March 1988.

2.3-1: Ewing, R., Kariotis, J. El-Mustapha, A., LPM/I , A Computer Program

for Nonlinear, Dynamic Analysis of Lumped Parameter Models, August

1987 .

2.3-2: Ewing, R . , El-Mustapha, A., Kariotis, J., Influence of Foundation

Model on the Uplifting of Structures, July 1988.

3 . la-1 : Scrivener, J., Summary of Findings of Cyclic Tests on Masonry Piers,

June 1986.

3 . lb-1 : Seible , F . and LaRovere , H . , Summary of Pseudo Dynamic Testing,

February 1987 .

4.1- 1: Limin , H . , Priestley , N. , Seismic Behavior of Flanged Masonry Shear

Walls, May 1988.

4.2- 1: Hegemier , G., Murakami, H., On the Behavior of Floor- to-Wall

Intersections in Concrete Masonry Construction : Part I:

Experimental

4.2-2: Hegemier, G., Murakami, H., On the Behavior of Flpor-to-Wall

Intersections in Concrete Masonry construction : Part II

:

Theoretical

6.2-1: Scribener, J., Bond of Reinforcement in Grouted Hollow-Unit Masonry:

A State-of-the-Art , June 1986

.

6.2-2: Soric, Z. and Tulin, L., Bond Splices in Reinforced Masonry, August

1987 .

8.1-1: Hart , G . , A Limit State Design Method for Reinforced Masonry

9.1- 1: Kariotis, J.C., Johnson, A.W., Design of Reinforced Masonry Research

Building , September 1987.

9.2- 1: Seible, F., Report on Large Structures Testing Facilities in Japan,

September 1985.

9.2-2: Seible, F., Design and Construction of the Charles Lee Powell

Structural Systems Laboratory, November 1986.

9.2-3: Seible, F., The Japanese Five-Story Full Scale Reinforced Masonry

Building Test, January 1988.

II. 1-1: TCCMAR, Summary Report: U.S. Coordinated Program for Masonry

Building Research, September 1985 to August 1986.

325

Theme V

Code Development Process and

Code Enforcement Responsibility

Minister-of-Construction's Agreement System in the

Building Standard Law of Japan

by

Tatsuo Murota

ABSTRACT

The Minister-of-Construction's Agreement

System in the Building Standard Law of Japan

is divided by five categories. They are concerned with the followings:

(1) Special building materials or methods of

construction unanticipated under the law.

(2) Materials. building elements or

equipments having property equal or

superior to that specified in the law or

cabinet orders.

(3) Structural design of tall buildings

exceeding 60m in height.

(4) Computer programs used for the structural

calculation of buildings.

(5) The bearing capacity of wooden shear

walls.

This paper describes the outline of those

categories of the Agreement System.

KEYWORDS

Building Standard Law. Ministry-ofConstruction's Agreement System

1. INTRODUCTION

The purpose of the Building Standard Law

(hereinafter Law) is to safeguard the life,

health and properties of the people by

providing minimum standards concerning the

site, structure, equipment and use of

buildings, and thereby to contribute to the

furtherance of the public welfare.

The minimum standards are provided in Chapter

II of the Law and they are supplemented by

the Building Standard Law Enforcement Order

(hereafter Cabinet Order) or notifications

issued by the Ministry of Construction.

Major purposes of the Minister— ofConstruction's Agreement System (hereinafter

Agreement System) are to make the minimum

standards responsible to technology

development and to supplement the

imperfection of the minimum standards.

The Agreement System are divided into five

categories according to the purposes:

(1) Recognition of special building materials or methods of construction unanticipated

under the Law.

In Chapter II of the Law. minimum standards

are provided concerning the site, structural

and equipment of buildings. Article 38 in

the Chapter refers to special materials or methods of construction as:

Article 38. The provisions of this Chapter

or those of orders or ordinances based

thereon shall not apply to buildings using

special building materials or methods of

construction deems that the said building

materials or methods of construction are equal or superior to those specified in the

said provisions.

The purpose of this Article is to correspond

to technology development by recognizing new

materials, new types of structures or

construction methods, etc. that are not

- anticipated under the Law. The first

category of Agreement System is based on this

Article, and has played an important role in

promoting technology development.

(2) Recognition of materials, building

elements or equipments having property

equal or superior to that specified in the

Law or Cabinet Order

Minimum standards concerning materials,

building elements or equipments in the Law or

Cabinet Order are provided in two ways: (1)

by showing examples or (2) by describing

minimum requirements. Either of the ways is

not perfect to provide minimum standards

correctly, and it is necessary to supplement

the imperfection by using the expression

"equal or superior to". A typical example of the former is seen in

Cabinet Order Article 108:

(Fire Preventive Construction)

Article 108. Fire preventive construction in

Article 2 item (8) of the Law shall be as described in each of the following items:

(1) Regarding walls whose studs and beds are made of non-combustible materials, the

said fire preventive construction shall be

as described in any of the following (a)

through (c):

(a) Metal lath with mortar finish of 1.5cm

or more In thickness:

(b) Cemented excelsior board or gypsum

board with mortar or plaster finish of

lcm or more in thickness:

(c) Cemented excelsior board with metal

board over mortar or plaster coating:

(2)

(3)

(4) In addition to those mentioned in the

preceding items, the said fire preventive

construction shall be that designated by

the Minister of Construction upon hearing

the opinion of the Director General of the

329
Fire Defense Agency as having fire

preventive property equal or superior to

that of the above-mentioned.

Article 108-2 is an example of the latter:

(Non-combustible Materials)

Article 108-2. Building materials having

non-combustibility as specified by Cabinet

Order under Article 2 item (9) of the Law

shall be those which are designated by the

Minister of Construction as having the

property described in each of the following

items (excluding item (2) in the case of

those used for external finish of buildings),

against the heat of normal fires.

(1) Materials which neither burn nor develop

deformation, melting, cracking or other

hampering fire protection:

(2) Materials which do not generate smoke or gas hampering fire protection.

computer programs are recognized to be valid

by the Minister of Construction, some parts

of structural calculation sheets prepared by

a qualified architect using the computer

program can be omitted in case of application

for confirmation.

(5) Recognition of the bearing capacity of

wooden shear walls

Cabinet Order Article 46 requests wooden

buildings to be provided with shear walls of

a certain extent of total shear strength. A

method for calculation of the shear strength

is also provided in the article.

According to the method, value of

"Multiplier" is necessary fop calculation

Multiplier is a physical quantity which is

determined by experiments as function of

shear strength and rigidity of bearing walls.

The fifth category of Agreement System is

concerned with recognition of the value of

Multiplier for bearing walls.

2. PROCEDURE OF APPLICATION FOR MINISTER-OFCONSTRUCTION'S AGREEMENT

The second category of Agreement System is

concerned with the recognition of materials,

buildings elements or equipments having

property equal or superior to that specified

in the Law or Cabinet Order.

(3) Recognition of structural design of

buildings exceeding 60m in height

Article 20 in the Law provides that the

safety of structure shall be confirmed

through structural calculation for buildings

exceeding a certain limit of scale.

In Cabinet Order Article 81-2. it is provided

that "Structural calculation for buildings

exceeding 60m in height shall be' made by

those methods which are recognized by the

Minister of Construction as being valid in

confirming the structural safety of the

building concerned".

The third category of Agreement System is

concerned with the provision.

(4) Recognition of computer programs used for

the structural calculation of buildings

In cases where a building owner intends to

construct any building as specified in Law,

he shall submit an application to the

building official for confirmation that the

plan conforms to the provisions of laws.

Article 1 of the Building Standard Law

Enforcement Regulations provided the form of

application for confirmations. and

also provides that the building

administrative agency may establish, by

regulations, provisions that, if the building

is designed by a qualified architect, all or some of the drawings and documents necessary

for confirmation may be omitted.

The fourth category of Agreement System is

conceded with the evaluation of computer

programs for structural calculation. If a

Procedures of application for the Ministerof-construction"s Agreement differ with the

categories. Details of the procedures are described in Reference 1), and so the

procedures only for category 1 and 2 will be

touched upon here.

(1) Procedure of Application for Category 1 Agreement Svstem

The flow of application is as follows (see

Fig. 1). The period from the beginning of

procedure 2 to the end of procedure 3 is

about four months, in general.

Q) Prior consultations

The applicant is suggested to have prior

consultations with the Building Guidance

Division of the Ministry of Construction,

regarding the application.

(2) Application for technical examination

The applicant is then required to apply to

the Building Center of Japan, for

technical examination on the proposed

technology.

(J) Application for recognition

Subsequently the applicant is required to

apply to the Minister of Construction for

recognition. In this case, the applicant

shall attach a document issued by the

Building Center of Japan certifying the

result of the technical examination in

procedure 2 .

(2) Procedure of Application for Category 2

The flow of application for Category 2

Agreement System is as follows (see Fig. 2). The period from the beginning of procedure 2

330
to the end of procedure 3 is about three

months, in general.

(T) Implementation of tests

Prior to the application, tests required

for the application shall be conducted at

the testing institutions designated by the

Minister of Construction

(5) Application for technical examination

Those materials which have passed the

above tests shall be subjected to

technical examination at the Building

Center of Japan.

(3) Application for recognition and

designation

Following the above technical examination,

the applicant is required to submit the

document certifying the result of the

examination above to the 'Minister of

Construction to apply for recognition and

designation. Those materials which have

been recognized to satisfy the specified

requirements, will be designated by the

Minister of Construction as materials

having specific property.

3. NUMBER OF AGREEMENT ISSUED BY THE MINISTER

OF CONSTRUCTION

Table 1 shows the number of Agreement's

issued by the Minister of Construction for

last five years (1984 to 1988 FY).

P REFERENCES

1) OUTLINE OF THE APPROVAL & CERTIFICATION

SYSTEM UNDER THE BUILDING STANDARD LAW. The

Building Center of Japan. 1986.
Table 1 Number of Agrement's issued by the Minister of

Comstruction in last five years

FIELD 1984 1985 1986 1987 1988

1 . High-rise Construction 0/1 Z4 ol AO 71ll 0701

o

L . KU Construction 00 OZ Oo DO CAOU

O

6 . Steel Construction 4U CO06 - oO oO

4

.

Wooden Construction A

U

o0

0

y 0

c

0

" 5. Foundation OA OQ00 OQ00 AQ4o C1ol

6. Computer Program 32 15 25 20 21

7. Base Isolation System 0 4 6 8 14

O

0 . nembrane Structure 5 12 12 19 13

y

.

Elevatory and Playing

Equipment

CA 00 OU 1 09 l££ 1 7Q

10. Plumbing 0 0 17 3 3

11. r ire Preventive

Construction

384 392 370 336 451

12. Fire Compartment 81 98 76 81 108

13. Separation Wall 11 6 11 18 17

14. Prefabricated Housing 154 132 85 138 158

TOTAL 849 889 861 1022 1257

332
Fig.l Application procedure for Category 1 Agrement System

Application

Certificate

(I) Certificate

® Prior Consultation

© Application

Building

Administrative

Agency Housing Burea

MOC

BC J

Examination

Commi ttee

Recognition

Commi ttee

Minister of

Construction

333
Fig. 2 Application procedure for Category 2 Agrement System

® Request for testing

© Test results -

-

(D Application

Applicant

© Report and Certificate

(D Application

© Certificate

Testing Institution

BC J

Examination

Committee

D Report and

Certificate

MOC

Housing Bureau

Notification to Building

Administrative Agency

334
History of Structural Regulations in Japanese Building Codes and

Outline of Wind-Resistant and Seismic Regulations

By

Yuji I SH I YAMA

'

Yuji OHASHT 2

SUMMARY dead or missing.

This paper presents a outline of Building Standard Law which prescribes the basic

building restriction of Japan and brief

histroy of Japanese building restriction

after Mei j i-Restorat ion when Japan set course for modern nation. And this paper

presents histories and current codes of

seismic design method and wind-resistant

design method. Opinions on the current codes are also summarized.

KEY WORDS : Building Restriction. Building

Standard Law, Seismic Design Method,

Seismic Force, Urban Building Law,

Wind Force. Wind-Resistant Design Method

1. OUTLINE OF JAPANESE BUILDING RESTRICTION

1 . 1 Feature of the land and cities of Japan

Japan is an earthquake-prone country

lying to the west of the circum-Pacif ic

earthquake belt. Approximately, 70% of the

land is either hilly or mountainous and the

remaining 30% is alluvial land that con- tains most of the urban areas. and is

subject to the influence of earthquakes.

Japan, therefore, has sustained much damage

due to earthquakes. throughout her long

h i s tory.

Japan is also situated in the path of

typhoons and many buildings are damaged by

typhoons every year.

Most of the urban areas in Japan can be

regarded as a continuation of old towns

«here wooden buildings are densely packed.

In such areas, dry air in winter, seasonal

strong winds and foehn phenomena that arise

in certain areas because of the mountainous

land, produces a very large degree of danger

of the occurence of fires and conflagrations.

Thus, it is highly necessary that buildings in Japan be able to safely withstand

attacks of earthquakes, typhoons and fires.

1 . 2 Purpose

Standard Law

and constitution of Building

Building Standard Law was enacted in 1950

considering the abovemen t i oned conditions,

and since then has been subjected to many

revisions in accordance with the advances in

technology and the conditions of building

disasters that have occurred.

The building standard

Building Standard Law.

Law Enforcement Order

Building Standard Law

t ions as Mini stry of

and Not i f icat ions of

Construction based on

statute consists of

Building Standard

as cabinet order.

Enforcement RegulaConstruction Order.

the Minister of

these regulations.

The objective of the statute is to "safeguard the life, health, and properties of

the people by providing minimum standards

concerning the site, structure, equipment,

and use of buildings, and thereby to con- tribute to the furtherance of the public

wel f are.

The statute comprises the "general provisions" on procedures for building con- struction, the "building codes" on standards

for safety and hygiene of all buildings in

the whole country, and the "zoning codes" on standards for ensuring the safety and appropriate environments in urban areas to be

applied to city planning areas based on City Planning Law, as well as penal provisions.

The urban areas in Japan have been subject to the attack of conflagrations in the

past.

The cause of conflagrations is not only

accidental fires but also the simultaneous

outbreak of fires in many places due to

a large-scale earthquake. The Great Kanto

Earthquake (1923) was typical of the latter,

with approximately 450.000 buildings being

destroyed by fire, and some 140.000 persons

1 Dr. Eng..

Dispatch Staff. Ministry of Construction.

Peru Office. JICA. ( Japan International Cooperation Agency ) Av. Salaverry 3150. San Isidro.

Lima. Peru (Apartado No. 1104171)

2 Senior Research Engineer.

Structural Engineering Department. Building

Research Institute, Ministry of Cnstruction

Tatehara-1. Tukuba-City. lbaraki. Japan

335
1 . 3 Outline of the regulations of Building

Standard Law

1.3.1 General provisions (Procedures for

building construction)

© Building confirmation

Generally in the case where a building is

to be constructed, the building client must

receive the cofirmation of a building official of the municipality. as to whether the

plan for the building is in conformity with

the laws relating to the site and structure

of the building, and the building equipment.

© Design and supervision by architect

The design and the supervision of con- struction works of buildings exceeding a certain scale must be performed by qualified

architects.

(D Inspection after the completion of cons truct ion

Once the construction work have been

completed, the building client must submit

a notification of this effect, to the building official. The said building must receive

an inspection by a building official, as to whether the building is in conformity with

the law relating to the building site,

structure and the building equipment. A certificate of inspection is then granted

when it is confirmed.

© Periodic inspection and periodic report

system

The owners of certain buildings such. as hospitals, hotels, department stores. theaters, collective dwellings and offices ex- ceeding certain scales, and the owners of

escalators and elevators and other building

equipments must have a qualified person

inspect the conditions of such buildings and

building equipments, every 1 to 3 years, and

must report the results of investigations to the municipality.

1.3.2 Bui lding codes

Building codes stipulate standards for

structural safety, fire prevention and sanit at ion.

external forces such as fixed load, imposed

load. snow load, wind force, soil pressure,

water pressure, seismic force, etc.

In order to ensure the safety of a building. Building Standard Law Enforcement

Order determines minimum standards according

to the structural type. It also provides

that structural calculations must be performed to confirm the safety of large buildings. The range of building for which structural calculations must be performed is as

f ol lows.

type of construction scale

wooden structure n > 3 or A > 5 0 0 m' others nS 2 or A Z 200 rrf (where n:number of stories of the building

A:total floor area of the building)

The method of the structural calculation

was revised in 1980, and has been in force

since June 1, 1981. Adding to the conventional allowable stress design method for

elastic range, this new method of structural

calculation determines design method considering ultimate stress including the stress

within a plastic range. In addition, story

drift, rigidity factor and eccentricity

factor should be checked to ensure a balanced structural plan.

The Enforcement Order prescribes the

loads. the external forces, the allowable

stress of building materials necessary for

structural calculation, and also the methods

of calculation that are necessary for structural safety in accordance with the height

and structure of buildings.

Regarding the structural calculation, to cofirm structural safety. at least five

kinds of loads and external forces must be

checked. namely. fixed load, imposed load, snow load, wind force and seismic force.

Values of allowable stress are provided

for the common materials such as timber,

steel, concrete, etc. Dual allowable stress

system is adopted, namely, allowable stress

for permanent loads and allowable stress for

temporary loads are provided.

Regarding structural safety, the codes

lay down the standards for detailed structure on wooden construction, masonry con- struction, reinforced masonry construction,

steel-frame construction, reinforced con- crete construction, steel encased reinforced

concrete construction and no reinforcing bar

construction and structural calculation.

In Building Standard Law. the basic

idea concerning structural strength is that

structures must be safe against loads and

Values of allowable stress for permanent

loads are half to one third of ultimate

stress of materials, values of allowable

stress for temporary loads are equivalent

to yield level or elastic limit of ma- terials.

The stress generated in each of the elements by the combinations of loads and

external forces, does not exceed the allowable stress.

336
The combinations of loads and external

forces are shown in table 1.

Regarding seismic design, adding to the

allowable design method. limitations on story drift, rigidity factor, eccentricity

factor and ultimate lateral shear strength

of each story are provided.

As for fire prevention, the codes stipulate that roofs and external walls of buildings in a specified area be made fire

preventive construction. that department

stores and other buildings for special use be made fireproof construction, and that

fire compartments be created in large scale

buildings according to a specified floor

space. Standards for emergency facilities,

including hallways, stai rcases. smoke-exhaust

equipment, etc. are also stipulated.

To secure sanitation, the codes lay down

standards for natural lighting and ventilation of rooms. structure of lavatories,

structure of elevators and escalators, etc.

1.3.3 Zoning codes

The zoning codes prescribe relationship

between sites and roads, land use designated

districts, ratio of total floor area to site

area, height of a building, restrictions on buildings in fire protection districts. etc.

2. HISTORY OF BUILDING RESTRICTION OF JAPAN

2. 1 Before Building Standard Law

With the end of the Edo Period and the

Meiji Restoration, Japan set the course for

a modern nation. Abolishing the insulation

policy of the Edo Era. the newly established

Meiji government carried out various meas- ures to open the country and catch up with

western advanced nations.

Among them was restructuring the capital

city of Tokyo. At that time, most houses in

Tokyo were of wooden construction and.

therefore vulnerable to disasters, especially fires. The government took various

steps to make housing in Tokyo fireproof and

renovate the old urban structure in Tokyo.

One of the initial actions was the "Ginza

Brick Construction Program" in the wake of

the 1872 large fire in Ginza. The program

was designed, to make housing in Ginza fireresistant through the brick construction

introduced from Western Europe and to re- consruct streets.

In the further effort for full-fledged

restruction of Tokyo. Tokyo City and Ward

Revision Ordinance. the first city planning legislation in Japan, was enforced in

1 888.

Meanwhile, in 1886. prefectural governments which have ports established regulations on hygiene of Nagaya apartment houses

against the background of the spread of

infectious diseases, such as cholera.

Around 1900, building restrictions on hygiene and f i reproof ness were set up in

many prefectures.

In these restrictions,

structural regulations.

there were few

With the outbreak of World War I . the

Japanese economy made outstanding progress,

causing concentration of the population and

industries in urban areas and making it necessary to establish urban planning and

building regulations in big cities other

than Tokyo.

In 1919. City Planning Law and Urban

Building Law was enacted for six major

cities. The two laws served as the starting

point of modern building regulations in

J apan.

The two laws provided for such systems

as building lines regulations and zoning

systems. They also stipulated standards for

buildings in terms of f i reproof ness. safety

in structural strength, and guarantee of

hygiene.

Regarding structural safety, the codes

laid down the standards for detailed structure on wooden construction, masonry construction, steel-frame construction, reinforced concrete construction and no reinforcing bar construction and structural

calculation.. Fixed load and imposed load were provided in regulations of structural

c a 1 cu 1 a t i on.

In accordance with progress in con- struction technology. and economic and industrial growth. Urban Building Law underwent several revisions. In 1923. the Great

Kanto Earthquake rocked the Kanto Area,

destroyinga huge number of buildings. Fires

caused by the earthquake burned down wooden

hous-es in the region. killing more than

100.000 people.

In reflection on the massive damage.

Urban Building Law set up seismic force and

made it obligatory to calculate the seismic

strength of building. Regulations on safety

of buildings in Japanese legislation have

since centered on f i reproof ness and earthquake-proof ness.

337
In the 1930s. the Japanese economy was subject to the military policy. As a result,

building restrictions through Urban Building Law were suspended in 1943. excepting

some on f i re-proof ness. and the war-tine

standards on building structures were established to conserve goods and aaterials.

After the World War II. Urban Building

Law was restored. But in accordance with the

enactment of the Constitution of Japan.

Building Standard Law was enforced, abolishing Urban Building Law. ( Technically

speaking, Building Standard Law was basically the extension of Urban Building

Law. Regarding structural calculations,

revised standard of the war-tine standard

was adopted.

)

The application area of Building Standard Law is the whole nation, not United

districts as in Urban Building Law. City

Planning Law, a sister law of Urban Building

Law, renains unchanged after the war.

2 . 2 Changes in building restrictions by

Building Standard Law

In 1978. the Miyagikenoki Earthquake

caused extensive danage to buildings, making

it urgent to revise seisnic design nethod in the law.

And in 1981. the drastic revision took

effect with dynanic considerations on seisnic calculation. For instance, the seisnic

force, which had been set at nore than 0. 2 of seisnic coefficient, was fixed in accordance with natural periods of buildings. The

revision also called for two-stage designing

-designing against the seisnic force equivalent to self weight of a building, in addition to the conventional allowable stress

designing against seisnic force of 20% of

the self weight of the building. Moreover,

retained ultinate lateral shear strength was

to be calculated with consideration for ductility of a building, not allowable stress.

Revisions and addition have been nade to standards for detailed structure in accordance with progress in structural naterials.

such as steel products, and development of new structural forms, including wood frame

construct ion.

Nearly 40 years have passed since

Building Standard Law was enacted in 19S0.

During the period. Japan has made remarkable

progress in econony and building technology.

And Building Standard Law has been revised

several tines. Following are major revisions

in the structural codes.

The codes underwent najor revisions,

while taking into account progress in building technology and damages to buildings

caused by earthquakes.

The Tokachi Earthquake in 1968 inflicted

heavy damage on reinforced concrete buildings, cons idered highly earthquake-resistant,

leading to review of earthquake-proof ness of

reinforced concrete buildings.

As a result, the structural codes of

Building Standard Law stipulated ductility

of buildings with the revision of reinforcement for shearing of pillars of reinforced

concrete buildings in 1971.

At that time, the development of computer

technology also promoted structural analysis

knowhow. Through computerized seismic analysis, construction technology of highrise

buildings was developed around 1970.

Against the background of these factors,

the Ministry of Construction led the development project for new seismic design method

from 1972 to 1977 with a view to thoroughgoing revisions of seismic design method in

Building Standard Law.

3. OUTLINE OF SEISMIC DESIGN AND ITS HISTORY

3. 1 History of seismic design method

After the Meiji Restoration in 1868. the

government abandoned the 230 year isolation

from foreign countries and resumed overseas

trade. .The government invited many researchers, scientists and engineers to Japan in order to develop Japanese industries as well

as to absorb Western culture as soon as possible. v

The Imperial University of Technology

was founded in 1877. Education of building

engineering started then.

The Yokohama Earthquake (M*5.4) occurred

in 1880 and caused moderate damage. The

earthquake was not very severe, but was strong enough to frighten resident researchers from overseas countries. In the same year the Seismological Society of Japan was founded by the effort of Dr. John Milne who

was a mineralogist from England.

The Architectural institute of Japan

< A I J ) was founded in 1886.

In 1891. the Nobi Earthquake (M=7. 9) oc- curred and caused very severe damage. In the

next year(1892) the Earthquake Investigation

Committee was founded and research on seismology was accelerated. The San Francisco

Earthquake (M«8. 3) occurred in 1906. In 1914

Riki Sano proposed the concept of seismic

338
coef f i c i en t.

The 1923 Great Kanto Earthquake occurred

near Tokyo and killed more than one 100.000

people. Most of the deaths occurred because

of the fires which spread after the earthquake. Next year (1924). one article was supplemented into Urban Building Law which

had been already enforced at that tine. The

article says, "Horizontal seisiic coefficient shall be more than 0.1." That neans

each building shall be designed and constructed so that it can resist a lateral force

of at least 10 % of its weight. In 1925 the

Earthquake Investigation Committee was re- chartered as the Earthquake Research Institute.

After the adoption into law of the horizontal seismic coefficient, many researchers

tried to develop a practical method for calculating stresses in the structure caused by

seismic forces. There were only a few engi- neers who could analyze the stresses of

a structure subjected to the horizontal

forces. In 1933. the year when Santikuoki

(M*8. 4) and Long Beach (M*6. 25) earthquakes

occurred. Kiyoshi Mu to proposed the wellknown Muto* s D value method to solve this

problem.

There had been only one allowable stress

for each material which corresponded to allowable stress for permanent load or about

one-third of the yield level of the material.

During the World War II. because of the

shortage of construction materials, a dual

allowable stress system was introduced into

Japan. That is. the stress caused by permanent loads was to be less than the allowable

stress for permanent loads, and the stress

from temporary loads was to be less than the

allowable stresses for temporary loads.

In 1950. after the World War II. Building

Standard Law and its enforcement order re- placed the old regulations. At that time the

horizontal seismic coefficient became 0.2

and the allowable stress for temporary loads

became twice the allowable stress which had

been used.

The height limitation of buildings had

been 31 meters or about 100 feet since adoption of Urban Building Law. The objective

of the height limitation was to restrict

the volume of the bu'ldings in urban area.

In 1963 the limi'ition was abolished. In 1964. the Niigata Larthquake(M»7. 5) occurred

which caused the tilting of buildings

because of liquefaction of saturated sand.

In 1968. Kasumigaseki Building. the first

highrise building in Japan, was completed.

After the San Fernando Earthquake (M»6.4)

in 1971. a five year national research project for establishing a new seismic design

method was carried out from 1972 to 1977 by

the Ministry of Construction, with the co- operation of the Building Research institute,

Public Works Research Institute. universities, private companies and many other

organizations. The new seismic design method

was proposed in 1977.

In 1978. the Miyagikenoki Earthquake

(M*7. 5) hit the Sendai area and its occurrence accelerated adoption of the new seismic des ignmethod. The new seismic design

method, which had already been proposed, was reviewed and evaluated for use as a practical design method. The new seismic design

method has been included in Building

Standard Law Enforcement Order. Notifications by the Minister of Construction,

etc. which constitute the current seismic

design code since 1981. Though the code

introduced most of up-to-date knowledge of

earthquake engineering, there still remain

many items to be improvedas the damage

by the Mexico Earthquake in 1985 (M=8. 1)

suggested.

3 . 2 Outline of the current seismic design

code

The purpose of the current seismic design

code is that buildings should withstand

moderate earthquake motions with almost no damage, and should not collapse nor harm

human lives during severe earthquake motions. In order to satisfy the purpose,

buildings are to be designed using one or more of thefoll owing design procedures.

1) Structural requirements : Buildings shall

meet the relevant structural requirements

specified by Building Standard Law Enforcement Order, Notifications of the

Minister of Construction. Specifications of

the Architectural Institute of Japan., etc.

2) Stresses : The stresses caused by the

lateral seismic shear for moderate earthquake motions shall not exceed the allowable

stresses for temporary loads.

a) Story drift : The drift of each story of

the building caused by the lateral seismic

shear for moderate earthquake motions shall

not exceed 1/200 of the story height. This

can be increased to 1/120, if nonstructural

members will have no severe damage at the

increased story drift.

4) Eccentricity, stiffness, etc.: The stiffness eccentricity Re(see note of Table 3) of

each story shall be less than 0.15. The lateral stiffness ratio Rs(see note of Table 4)

of each story shall be greater than 0.6.

339
5) Ultimate lateral shear strength : The ultimate lateral shear strength of each

story shall not be less than the specified

ultimate lateral shear.

The above design procedures are applied

to each building according to its structural

type, floor area, height, etc.

The lateral seismic shear

story is determined by

Qi of i-th

sidering not only strength but also ductility, because a two step design, i.e.. allowable stress design and ultimate strength

design, has been introduced.

3.3.2 Unfavorable opinions

In spite of the above favorable opinions

on the current design code. there are several demerits or items to be improved as follows:

Qi = Ci Wi (1)

where Ci is the lateral seismic shear co- efficient of the i-th story and Wi is the

weight of the building above the i-th story.

Lateral seismic shear coefficient of the

i-th story is given by

Ci ' Z Rt Ai Co (2)

where Z is the seismic hazard zoning co- efficient (F i g. 1). Rt is the design spectral

coefficient (Fig. 2). and Ai is the lateral

shear distribution factor(Fig 3). The standard shear coefficient Co is 0.2 for moderate

earthquake motions and 1.0 for severe earthquake motions.

For buildings higher than 31 m, or with

irregularities exceeding the specified limits, the calculated ultimate lateral shear

strength of each story must not be less than

the specified ultimate lateral shear Qun ;

Qun = Ds Fes Q (3)

where Ds is the structural coefficient. Fes

is the shape factor, and Q is the shear for severe earthquake motions which is given by

Eqs. (1) and (2) letting Co * 1.0.

The structural coefficient Ds is determined by the energy absorbing capacity of

the structure (Table 2). The shape factor

Fes is the product of the factor Fe(Table 3)

and the factor Fs (Table 4).

A. Shape factor Fes:

® In order to satisfy the requirements

related to the shape factor, engineers are

apt to intentionally eliminate shear walls

and braces which have been proven to be very

effective against earthquake motions.

(2) If the requirement for stiffness eccen- tricity Re is not satisfied, the ultimate

lateral shear strength should not merely be

increased but should be increased so that

the torsional vibration will be reduced.

d) Current procedure for estimating Fes

does not seem to reflect the real behavior

of buildings. Fes should be determined not

only by the elastic stiffness but also by

the behavior after yielding.

® Evaluating stiffness of partition walls

and other nonstructural members is very

difficult. so that Fes estimated using such

stiffness may not indicate real characteristics of the buildings.

® In order to only satisfy the requirements for stiffness eccentricity and for

lateral stiffness ratio, engineers are apt

to unnecessarily increase sizes of columns.

® In order to control the stiffness, slits

along columns and walls are often provided,

which may not improve the total behavior of

the building.

B. Structural coefficient Ds;

3 . 3 Opinions on current seismic design code

3.3.1 Favorable opinions

Generally speaking. the current seismic

design code has been favorably accepted by

most engineers, mainly because of the following reasons:

(D Architects as well as engineers have

come to recognize the importance of structural planning and many buildings have been

designed so they are in good structural

balance.

® Many buildings have been designed, con0 Ultimate lateral shear strength is only

evaluated as the sum of lateral strength of

columns. walls. braces, etc. The deformation at which the lateral strength is attained, however, should be considered.

<Z> Structural behavior of the building can not be expressed only by the unique value of

Ds. ® Ds is evaluated at each story, but it should be evaluated as a unique value for

the building, considering the total behavior

of the building.

340
® The way to determine the structural co- efficient Ds depends lainly on analyses for

single degree of freedoa systeis. Therefore

Ds evaluated by the current nethod does not

necessarily express the real behavior of

buildings which should be regarded as multi

degree of freedoa systems.

C. Other iteas;

0 The seisaic hazard zoning coefficient Z ranges between 1.0 and 0. 7 . It does not

sufficiently reflect seisaicity in Japan

which seeas to be aore diversified.

® Not included is the importance factor

which is considered in aost seisaic regulations in the world. The factor should be

1 nc 1 uded.

(3) Regulations for substructure are not

sufficiently provided, coapared to super

structure.

® Daaage by earthquakes shows that non- structural aeabers. building equipaent. furniture, fences, etc. aay be aore hazardous

than aain structure which is already designed against earthquakes. Therefore aore provisions for nonstructural iteas should be

provided.

® Consideration of dynaaic behavior of

buildings is well covered in the code, but

nonlinear behavior of structures after

yielding is not sufficiently covered.

© Dynaaic analysis/design aethod. e.g.,

aodal analysis and tiae history analysis,

should be included.

® In the case of extension for old buildings, there is difficulty in satisfying the

current requ i reaen t s.

D. Overall concept;

In addition to the above coaaents. there seeas to be a aore crucial point concerning

the overall concept for the current design

code. That is "The current design code has

stipulated too auch in detail.* This happened in spite of the fact that at the

beginning stage of drafting the current code

it was confiraed that design philosophy

should be clarified. and design detail

should not be specified in the code so that

it could be deterained by each engineer.

Soae related opinions are as follows:

(D Regulations in detail aake aost engineers just follow the stipulated design procedures without considering the overall

structural design. Many engineers are apt

to rely too auch on coaputer prograas which

can handle any design routes from the beginning to the end of the structural design

without aan-nachine interaction.

(2) Factors such as Ai. Ds. and Fes should

not be, specified in the code but be appropriately determined by engineers.

© An ideal case nay be that the code

should only clarify the criteria of probability of daaage by earthquakes and experts

on structural designing would then just

report to building officials that the criteria is satisfied, in order to get building

perai t. ® Current design procedures require nuch aore tiae for designing buildings and also

for checking docuaents for building pernit

than the previous code.

3.3.3 Closure

The current seisaic design code, which is

often called the new seisaic design aethod.

has been enforced since 1981 and has been

accepted favorably by aany engineers and

building officials. The above aentioned

opinions, however. indicate . aany useful

suggestions not only to iaprove the current

code in Japan but also to consider in developing new codes in other countries.

4. OUTLINE OF I 1 ND-RES I STANT DESIGN AND ITS

HISTORY

4. 1 History of wind-resistant regulation

In 1886. the Architectural Institute of

Japan ( A I J ) was founded. In 1890, Tatsuzo

Sone introduced translated paper on windresistant design of Eiffel-Tower. In 1906.

Riki Sand presented paper on wind force.

Those were the first papers on wind-resistant design of buildings in Japan. _

.

In Taisyo-Era, several typhoons struck

Japan, and did considerable daaages to wooden

buildings. Consequently Syozo Uchida and

others proposed draft regulation for wooden

bui ldings.

Urban Building Law. which had been

already enforced at that tiae, had structural

calculation regulation. But it didn't have

wind. force regulation.

Because of daaages by typhoon. wind

force regulation was provided as follows in

Enforcement Regulation of Urban Building Law

of Tokyo,

wind force

75kg/m* (hS6a) (where h : height

100kg/rrf (h>6m) froa ground)

341
In 1934. the Muroto Typhoon. which had

the largest magnitude, struck Kansai district, western part of Japan. A lot of

buildings suffered very severe damages, and

there were heavy casualties. These damages

promoted reseach on wind-resistant design

of build i ngs.

Yosikatsu Tsuboi. Kiyosi Mu to and others

introduced foreign research papers on windresistant design. Hantaro Nagaoka who was a famous physicist criticized wind-resistant

design method of buildings.

Japan-China War occurred in 1937. and

Japan became under war codition.

AIJ proposed draft standard of structural

calculation for steel structures. In this

draft standard. wind force regulation was provided as follows.

W = C x q x A W : wind force (kg)

C ; coefficient of wind force

q : veloci ty-presser (kg/mm 2

) height q for building for tower

h£15m 100 200

15m<h£30m 130 260

30m<h£50a 160 320

50m<h 200 400

where A : action area ( m*

)

h ; height from ground (i)

During the World Warn. Y.oshiro Taniguchi

researched into wind resistant design of

hangers for airplanes.

vided nearly equivalent to 1 20 4 /"h. This

recommendation was not compulsory.

Research on dynamic effect of wind had

been done. 'Draft Standard of Loads for

Buildings*. which was proposed by AIJ in 1975, presented data for dynamic effect of

w i nd.

The regulations of structural calculation of Building Standard Law Enforcement

Order was revised in 1981. At that time,

veloci ty-presser q was revised as follows,

q = 50 ^ h ( h g 16* )

1 20 4 /" h ( h > 16m )

In 1981. AIJ proposed "Recommendation of

Loads for Bu i Id i ngs" . i n which dynamic effecl

of wind and statistics were introduced. This

recommendation also was not compulsory.

4. 2 Present regulation of wind force

4.2.1 Regulation of Building Standard Law

Enforcement Order

The regulation of structural calculation

of Building Standard Law Enforcement Order

adopts allowable stress method. Concerning

the wind resistant design, the regulation

says that the stress caused by the com- bination of permanent stresses and temporary

stress by wind force shall be less than the

allowable stresses for temporary load. The

allowable stresses for temporary load are equivalent to yield level or elastic limit

of materials.

Because of the shortage of construction

materials. war-time standards on building

structure was provided. In this standards,

wind force was as follows.

q 40/~ h q : ve 1 oc i ty-presser (kg/im 2) h ; height from ground (m)

After the World War II. Japan Standard

3001. which was the regulation of structural

calculation and was revised standard of

war-time standard. was provided. In this

standard, velocity presser q became 60/" h.

Building Standard Law was provided

instead of Urban Building Law, in I960.

Japan Standard 3001 was adopted as the

regulation of structural calculation in

Building Standard Law Enforcement Order.

Notification by the Minister of Construction on reduction coefficient for wind

force was provided in 1952.

AIJ proposed "Technical Recommendation

for Highrise Building" in 1964. In this

recommendation, ve loc i ty-presser q was proWind force is provided as follows.

q x A wind force^ (kg)

coefficient of wind force

ve 1 oc i ty-presser (kg/mm-)

q = . 60 f h ( h £ 16m ) 120 4 /"h ( h > 16m ) action area ( nf

)

4.2.2 " Recommendation

Buildings" by AIJ

of Wind Force for

Regarding the wind force. "Recommendat ion

of Loads for Buildings " by AIJ says as

f ol lows.

P = qx Cr x Gr XA

P ; wind force (kg)

q ; veloc i ty-presser (kg/mm 2

) Cr : coefficient of wind force

Gr ; gust influence coefficient

A : action area ( nf

)

where

q = (1/2) 2 • p • V: 2

p : density of air

Vz : design wind velocity

342
V: = Vo • E • R Va ; normal wind velocity

see fig. 4 (mean return period

of 50 years)

E ; wind velocity distribution

coef f i c i en t see table 5.

R : return period conversion

coefficient of wind velocity

R = 0. 61-0. lOloge C log* (t/(t-l) )

]

t : return period

Cf is given by wind tunnel test.

Gr is given froa table 6 or

by precise analysis.

4.3 Opinions on current wind resistant

design code

4.3.1 Opinions

Opinions on current wind resistant design

code are as follows.

<X> Current wind resistant design code is

provided for the buildings, area of which

exposed to wind are coaparat i vely saall, and

rigidity of which are coaparat i vely high.

Consideration to dynaaic effect of wind is

necessary for buildings, area of which exposed to wind are large. and rigidity of

which are low.

@ Current wind resistant design code is

provided uniformly for all over country, and

isn't based upon statistic data. and the

roughness of the ground of the construction

site of building isn't considered.

® Fornula of velocity pressure changes at height of 16a. That is irrational.

® Although daaages of substructure by

strong wind were larger than super-structure

regulations for substructure are not sufficiently provided.

® Current wind resistant design code is

based on the allowable stress method, but it is necessary to include fatigue design,

deforaation design, rigidity design, etc.

4.3.2 Closure

Current wind resistant design code is

provided basically forty years ago. and have

not adopted results of later studies, it is

necessary to introduce resent studies of

wind force in code.

Table 1 Combination of stresses for structural calculation

Kind of Stress Conditions estimable

regarding loads and

external forces

General case Case of heavy

snow district

Permanent stress Normal time G + P G + P + S

Temporary stress Snow season G + P + S G + P + S

Storm G + P+W G+P+S+W

Earthquake G + P + K G+P+S+K

where, G.P.S.W and K represent respectively the following stresses

(stresses combined of axial stress, bending monent, shear and others):

G : stresses caused by fixed load

P : stresses caused by imposed load

S : stresses caused by snow load

W : stresses caused by wind force

K : stresses caused by seismic force

343
Table 2 Structural Coefficient Ds*

Type of Frame

Behavior of

Members

(1) Ductile

moment frame

(2) frame other than

listed in (1) & (3)

(3) **

A. Members of

excellent ductility

0.3 0.35 0.4

B. Members of

good ductility

0.35 0.4 0.45

C. Members of

fair ductility

0.4 0.45 0.5

D. Members of

poor ductility

0.45 0.5 0.55

* Values can be decreased by 0.05 for steel encased reinforced concrete

buildings and for steel buildings. ** Frame with shear walls or braces for reinforced concrete and steel encased

reinforced concrete buildings, and frame with compressive braces for steel

buildings.

Table 3 Shape Factor Fe and Stiffness Eccentricity Re

Re* Fe .

less than 0.15 1.0

0.15 i Re i 0.3 linear interpolation

more than 0.3 1.5

* Re = e/re

where e is the eccentricity of the center of

stiffness from the center of mass, and re is the

elastic radius, i.e. square root of the torsional

stiffness divided by the lateral stiffness.

Table 4 Shape Factor Fs and Lateral Stiffness Ratio Rs

Rs* Fs

more than 0.6 1.0

0.3 ^ Rs 1 0.6 linear interpolation

less than 0.3 1.5

* Rs - r/f

where r is the lateral stiffness, which is defined

as the value of the story height divided by the

story drift caused by the lateral seismic shear for

moderate earthquake motions, and r is the mean

value of r.

344
Table 5 (a) Class of roughness of ground surface

Class of roughness

of ground surface

Condition of ground surface near the construction

site of the building

I flat ground with almost no obstacles

n district with some obstacles like crops or with thin trees and low rise buildings

m district with dense trees and low buildings or with thin middle or high rise buildings

IV district with dense middle or high rise buildings

Table 5 (b) Distribution coeficient E

Class of roughness

Height of ground surface

from ground

I n m IV

ZSZb 1.12 1 0.87 0.72

z b < zs

z

c 1.20(Z/10) 0-' (Z/10)°- ,s 0.80(Z/10)°- 2 0.52(Z/10)°- 8

z 6 < z 1.62

Table 5 (c) Z b . Z e

Class of roughness of

ground surface

I n ffl rv

Zt 5 1 0 1 5 3 0

Zg 2 0 0 2 5 0 3 5 0 4 5 0

Table 6 Gust influence coefficient Gf

Class of roughness G,

of ground surface

I 2.1 n 2.2

m 2.4

IV 2.8

345
Ei3 • a z 1.0 §11 '• b z 0.9

••C Z»0.8

2=0.7 h Okinawa

Fig. 1 Seismic Hazard

Zoning Coefficient, Z

Oi-W/Wn

W» Weight above

i-th story

n Weight above

ground level

T » Fundamental natural

period (sec.)

Fig. 3 Lateral Shear

Distribution Factor, Aj_

1-02 <oT-U

2

,-0.2(^-1)2

Soil Profile Type 3

(Soft)

Soil Profile Type 2

(Medium)

Soil Profile Type 1 (Hard)

1.0 2.0 (sec)

Fundamental Natural Period ,T

Fig. 2 Design Spectral Coefficient,

Through the precise analysis of the structure, the foundation,

the soil, etc., the value of can be reduced to 0.75 of the

value given by this figure. But the value shall not be less

than 0.25.

346
347
Development and Enforcement of U.S. Building Regulations

by

James G. Gross and Richard N. Wright

ABSTRACT

U.S. practices for development and

enforcement of building standards and

codes are described for the guidance of

those who seek to develop or apply

improved building products or practices.

Trends, such as mitigation of effects of

natural hazards, conservation of energy,

water and environment, and efforts to

reduce barriers to beneficial innovations or trade, lead to deniands for changes in

building products and practices.

Understanding of the U.S. building

regulatory system and the roles of

participating organizations is essential

to introduction of new products or

practices and to improvement of the

building regulatory system.

KEYWORDS: Accreditation; Building;

Building Code; Certification; Product

Approval; Regulation; Standard.

1. BUILDING AND BUILDING REGULATICNS

1.1 Introduction

Historically, building regulations were

created and enforced to satisfy the

objectives of protecting the "health" and

"safety" of building occupants and the

public in general [1]. In recent years,

a third objective "welfare" has been

added and has received increased

emphasis. Furthermore, during recent

years, building regulations have become more stringent. New provisions hare been

added, with very few deletions.

Enforcement of more stringent regulations

raises concerns for adverse effects on

the cost of building construction [2]

.

The Constitution of the United States

leaves the authority to regulate building

design and construction to the states.

In the past, this authority usually has

been exercised by political jurisdictions

at the local level, such as cities and

counties. Recently, there has been a

broad movement by the individual states, as described in section 2, to reclaim

authority for the development and

implementation of statewide building

codes. International efforts to increase

trade in building products and services

seek elimination of use of standards and

certification systems as barriers to

international trade by establishing a

framework of cannon rules t [3] . These are expected to lead to international

harmonization of building standards,

regulations, product approval systems,

and requirements for professional

licensing. These efforts are likely to

require changes in U.S. building

regulations and regulatory practices.

This paper begins with a description of

the building process, its principal

actors, and the role of standards in the

building regulatory system and in the

building process. Interfaces between the

building regulatory system and the

building process are described.

Technical and policy trends leading to

changes in building regulations and

regulatory processes are noted.

1.2 Building Process

A simplified, schematic diagram of the

building process is shown in figure 1.

The early phases, feasibility study and

program analysis, decide where the

building will be located, how much space

will be provided and what characteristics

National Institute of Standards and

Technology, Gaithersburg, MD 20899

348
are required for the space. The

principal actors in this activity are the owner and the architect. It is important

to obtain adequate professional inputs in

programming. Specialists, such as the

engineering geologist and the structural

engineer, may guide in site selection and

in selection of a structural scheme well

adapted to the site conditions,

respectively.

The design phases, schematic design and

design development, are concerned with

the identification and elaboration of a

solution meeting the requirements

developed in programing. Principal

participants in design include the owner, architect, structural engineer, mechanical engineer, foundation engineer, etc.

The contracting phase, involving

preparation of contract documents and

bidding or negotiation for selecting the

contractor, requires preparation of

detailed working drawings and

specifications by the design team or potential contractors, and the

preparation and evaluation of

contractors' offers. The construction

phase involves the actual construction

and its review for compliance with the

plans and specifications by the designer, as the owner's agent, and the building

regulatory authority.

The final part of the building process,

occupancy, has as principal actors, the

persons using the building, and those who

maintain or remodel it.

1.3 Building Regulatory System

Major interactions of the building

regulatory system with the building

process are also shown in figure 1. The

principal elements of the building

regulatory system are: (1) the building

regulations denoted as "codes," and (2)

the regulatory agency or agencies which

enforce and interpret the codes.

The regulations speak directly to certain

aspects of the building and the building

process:

« o Land use, defining the natural

and man-made environments

,

o Design, controlling the perfor- mance of the building in the

natural and human environments,

o Construction, controlling how

the design is implemented, and

o Occupancy, controlling how the

building is used, maintained,

remodeled, or removed.

The usual interactions of the building

regulatory system and the building

process are shown in figure 1. The codes

themselves are referred to by the

planners and designers during the

programming and design stages to assure

that these regulations are complied with.

The building regulatory system reviews

the planning and design when appropriate

documents describing the designer's

conception are made available to the

building regulatory system. This

interaction is denoted "plan review" in

figure 1. Following approval, after any

necessary modifications , field

enforcement by the building regulatory

system promotes compliance with the

design during construction. A view of

interactions and roles in the building

regulatory system is shown in figure 2.

1.4 Aspects of Building Regulation

Standards and codes may be either

performance or prescriptive in nature

[4] . A performance standard expresses

the desired attributes of the product or

process in question: any scheme which

results in achievement of these

attributes meets the standard. This

approach promotes economy and innovation

since it focuses on the users'

requirements. A prescriptive standard is

a specific definition of an acceptable

process or product. The desired

attributes often are unstated: the

process or product may meet the standard

but may not meet the users' needs.

Prescriptive standards have the advantage

349
that they are objective. It is

relatively straightforward to evaluate

and determine whether or not the standard

is complied with. However, prescriptive

standards do not assure satisfaction of

the users' needs, let alone in an optimal

fashion. Performance standards have the

disadvantage that it is conceptually more difficult to evaluate whether they are complied with. As indicated in figure 3,

a laboratory-type activity may be

required to evaluate compliance. Efforts

have been made to standarize the process

of evaluation of performance standards

[5].

Performance standards should not be seen

as the antithesis of prescriptive

standards. Both have their place.

Performance standards are essential for

description of desired attributes of the

building. It is most meaningful to the user to say: "the building must have

less than one percent probability of

unserviceability from earthquake per

year." Prescriptive standards are useful

statements to the designer or to the

builder of exactly what is wanted.

Ideally, prescriptive standards will be

related to performance standards so that

the performance intended is explicit.

land use regulations may be intended for

protection from natural hazards or

protection of property values. The major

elements of land use regulations are:

o The general plan, which is a

prescriptive standard for a

geographical area describing

how land use is to develop.

o Zoning regulations, which

describe the conditions for

land use for various purposes.

One set of zoning regulations

may be applied to a variety of

general plans.

o Subdivision regulations prescribe in detail how land is to

be used when subdivided for a

specific purpose. Applications

of land use regulations for

natural hazards mitigation are described in [6] : economic

consequences of overly

stringent land use regulations

are discussed in [2]

.

Design and construction standards and

codes deal with the various physical

systems of the building, such as the

structure, mechanical systems, plumbing

and electrical, or with specific aspects

of performance such as fire safety.

Section 2 provides a detailed description

of the development of building codes.

This is representative of development of

other standards and codes.

Housing codes control the conditions of

occupancy of buildings. These codes

generally are concerned with public

health under normal conditions of use,

rather than mitigation of natural

hazards.

Mobile homes, housing units built on

permanent chassis, are uniquely subject

to Federal building regulations.

Originally mobile homes were considered

to be often-transported housing and

regulated under transportation

authorities of the various states.

However, it became cannon for a mobile

home to be transported only once to a

building site where it remained

throughout its useful life as lower-thannormal cost housing. In recognition of

their public importance, Congress

established a national system of

regulation for mobile homes in 1974 [7]

.

Other manufactured homes are factorybuilt, in search of improved quality and

economy, but are similar to conventional

houses when erected on site. Efforts are continuing [8] to create a regulatory

system for factory-built housing that

will provide a national market to

manufacturers while imposing consistent

requirements on conventional and factorybuilt housing.

Rehabilitation of existing buildings and

preservation of historic buildings pose

special regulatory requirements. First,

350
current building standards and codes may

not apply to the materials and systems of

older, existing buildings. Second, it

may be unduly expensive to require the

same performance level of existing

buildings as is required of new ones. Technical studies of the National Bureau

of Standards in cooperation with the

National Conference of States on Building

Codes and Standards [9] led to the

development of improved technical bases

for regulations, improved regulatory

process, new technology and evaluation

tools, and cost-effective decision

methods. The National Institute of

Building Sciences developed rehabilitation guidelines [10] incorporating this

work: these guidelines, in turn, are

referenced in state and local building

codes.

Licensing for professional services in

the building process also is a state

responsibility in the United States.

Such licensing includes design services

of architects and engineers, and

licensing of contractors. Professional

licensing in the U.S. requires graduation

from an accredited, professional academic

program, experience under the supervision

of a licensed professional, and

successful completion of a written

examination. The 1988 U.S. -Canada free

trade agreement, and similar, anticipated

agreements with the European Connunity

[11] have caused the National Society of

Professional Engineers [12] , and similar

organizations for other licensed

professionals, to explore national

systems for professional licensing.

2. BUILDING CODES

2.1 Introduction

There are many types of codes that fall

into the broad category of building

codes. The various types are reviewed in

section 1. For purposes of illustration

and discussion, this section will deal

only with building codes. Mich of the

discussion relative to the development

and administration of building codes also

applies to the other building-related

codes.

2.2 Status

There are now 39 states that have

statewide building codes, in various

forms, which are implemented in different

ways [13] . Thirty- five are based upon

one of three model codes, and four are

state written. Some states allow

entirely voluntary use at the local

level, while twenty-two others require

mandatory usage for the entire state

(Alabama, Alaska, California,

Connecticut, Florida, Indiana, Kentucky,

Massachusetts, Michigan, Minnesota,

Montana, New Jersey, New Mexico, New

York, North Carolina, Ohio, Oregon, Hnde

Island, Utah, Washington, West Virginia,

Wisconsin) . Eleven states (Arkansas

,

Idaho, Illinois, Iowa, Kansas, Maryland,

Mississippi, Nevada, New Hampshire,

Oklahoma, South Carolina) require

mandatory application of the state code

only for state buildings. Five states

have legislation under consideration to

adopt statewide codes.

Building regulation and enforcement have

been widely cited in the literature [2] as "adding significantly to the cost of

building construction." The two

principal constraints vhich have been

identified and cited for increasing the

cost of construction are : (1) some building codes have requirements that are excessive, or are so prescriptive as not

to permit alternate designs which would

satisfy the regulatory objective; and (2)

the time delay involved to carry out the

building regulatory process. The first

deals primarily with the technical

content of the building codes, and the

second deals primarily with the process

of administering building codes

.

Several national organizations have been

formed to improve both the code content

and the process for building code

enforcement. Three of these organizations are associations of building

officials; each promulgates a model

building code (i.e., the Building

351
Officials and Code Administrators

International (BOCA) , the National

Building Code; the Southern Building Cods

Congress International (SBCCI), the

Standard Building Code; and the

International Conference of Building

Officials (ICBO), the Uniform Building

Code). These "model" codes ate not, in a

strict sense, a code. A code is a

standard or other set of conditions and

requirements which is made mandatory by

government bodies, either through direct

legislation or through administrative

regulation, based on authority granted by

a legislative body. A model code is a

standard or other set of conditions and

requirements that is reconniended for use

as a code or regulation. It is developed

and promulgated with the intent that it

be adopted for regulatory application.

The Council of American Building

Officials (CABO) is a consortium of the

three model code organizations, with its

offices in the Washington, DC metropolitan area. It promulgates the One and

Two Family Dwelling Code and the Model

Energy Code. The One and Two Family

Dwelling Code is widely used for the

regulation of single family attached and

detached dwellings. It is also used by

many states as the basis for regulating

modular housing. Modular housing is

housing which is made up of prefabricated

volumetric boxes produced in a

manufacturing plant and moved to the site

where the boxes are arranged and attached

to each other so as to form a complete

structure . The Model Energy Code

represents a codification of the American

Society of Heating, Refrigerating and Air

Conditioning Engineers (ASHRAE) standards

addressing energy conservation for

buildings. It is used as the basis for

energy conservation in the model building

codes as well as most of the states of

the Nation.

The model building codes cited above are gaining in acceptance and use by state

and local governments. Of the 39 states

with statewide building codes, 35 are based on either the National, Standard, or Uniform model codes. The major cities

of the United States also appear to be

moving toward the adoption and use of the

three model codes. The Association of

Major City Building Officials (AMCBO),

made up of the 31 cities of over 500,000

population, reports that 80 percent of

these cities now use the model codes as

the basis for their regulations, although

many of these cities amend the models.

The model building codes are not uniform

in format, scope, or content; however,

there is a great deal of similarity in

scope and technical content.

A major advantage of model building codes

is that each of the promulgating

organizations, BOCA, ICBO, and SBCCI, has

well organized code revision procedures

for keeping the model code up to date and

responsive to the latest advances in

technology. A code change cycle takes a

year, so it is' necessary for proposed

code changes to be submitted nearly a

year ahead of the time that a final

decision is made to change the code. The

process includes two public hearings. At

the first, an appointed code changes

comoaittee hears testimony and studies the

documentation submitted with the proposed

change. Following this hearing, the code

changes ccmnittee makes recaimendations

for approval, approval as modified, or

disapproval. These recaimendations are published and distributed to the

membership and other interested parties.

Anyone may challenge the reconnendations

in writing with supporting documentation.

Challenges are heard at the final hearing

before the full membership and there

finally acted upon. Unchallenged

reccnnendations of the code changes

committee are automatically ratified by

the membership. The model codes are

issued in their entirety on a three-year

cycle with supplements issued during

intervening years. The latest complete

edition is 1988 for the Standard and

Uniform Model Codes and 1987 for the

National Model Code. The last complete

One and Two Family Dwelling Model

Building Code was issued in 1989. It,

too, is printed on a three-year cycle.

352
2.3 Administration and Fhforrarent o Adoption and change procedures.

The efficient application of building

regulations requires technically sound,

clear, and correct documents, and

efficient processes for applying these

regulations. It is the opinion of many

experts that the inefficient enforcement

of building codes contributes more

unnecessary costs to buildings than do

ineffectual technical constraints. The

essential elements of an effective

regulatory process include:

o Clear governmental authority,

responsibility, and accountability with little or no

conflict or duplication,

o Efficient process to eliminate

conflicts, overlaps, duplication and delays,

o Qualified administrative and

technical personnel,

o Certification and training

programs for building code

enforcement personnel, and

o Sufficient funding to provide

for the above.

The building code enforcement process

generally includes three major steps:

o Plan review and permit

issuance,

o Inspection during construction,

and

o Approval for occupancy and use.

In addition, activities which do not

affect each building but are essential to

the overall building code administration

and enforcement include:

o Pre-permit consultation,

o Procedures for appeals, and

These , elements of the process, if

cumbersomely and inefficiently administered, can cause excessive cost to both

the applicant and enforcement official.

Excessive time delays in conduct of the

process affect the overall time for

construction, and thus total construction

cost. While pre-permit consultation is

not a mandatory part of the enforcement

process, it is receiving increased

attention as an opportunity early in the

design process to resolve differences in

understanding the building code requirements and as a means to significantly

smooth out and reduce the time required

for plan review.

A recent development is one-stop

permitting . This idea is being embraced

by designers and building officials alike

to simplify the permit process. The

following benefits accrue to the

participants.

o An applicant goes to only one

location for services,

o Duplicate requirements are reduced, which avoids

unnecessary confusion,

o Time is saved by both code

users and the building

department staff, and

o User uncertainty and misunderstandings are reduced or

eliminated.

In 1982, the President's Commission on

Housing Report [2] called for "development of models of efficient building

department administration appropriate to

various size States." The National

Conference of States on Building Codes

and Standards (NCSBCS), under contract to

the Federal Trade Commission, has

developed four model administrative

procedures for effective state building

code administration of residential

buildings and a "model one-stop permit

processing for one and two family

353
dwellings." NCSBCS has developed model

processes for nonresidential construction as well.

A recent study by the Business

Roundtable, and reported in their report

E-l, "Administration and Enforcement of

Building Codes and Regulations," provides

excellent background dealing with many of

these issues [14]

.

3. BUILDING STANDARDS

3.1 Introduction

Standards, which are referenced or

incorporated, provide the bases for most

of the technical requirements in building

codes. While most of these standards are materials specifications or test methods,

there are numerous design standards,

particularly structural design standards,

which are referenced in the codes. Many

of these same standards are widely used

as a basis for design contracts and

construction specifications as well as building regulations. Because of this

broad regulatory and contractual use,

standards provide the principal vehicle

for the transfer of advancements from

research in building technology.

3.2 Standards Used in Building Codes

There are nearly 2,000 standards

referenced in U.S. building codes. A

typical U.S. building code references 300

to 400 primary standards. In a National

Bureau of Standards report [15] these

standards are listed for major city and

state codes. Although this report is more than ten years old, the situation is

much the same today, with the exception

that more states and major cities now

rely on the model building codes. The

codes studied included the three model

building codes (National Building Code,

Standard Building Code, and Uniform

Building Code), 20 state building codes,

and the building codes of the 30 largest

cities. The standard referenced most

often was the American Concrete Institute

Standard 318, Building Code Requirements

for Reinforced Concrete , which was

referenced 123 times in 36 codes. The

second most referenced standard was the

American Institute of Steel Construction

Specifications for the Design.

Fabrication, and Erection of Structural

Steel for Buildings , which was referenced

62 times in 34 codes. Both of these

standards are structural design

standards, which indicates the widespread

reliance on such standards for regjQatory

purposes. Approximately 2,000 standards were referenced by one or more of the

codes studied. Further, these primary

standards in turn reference many

secondary and tertiary standards.

Compliance with a referenced standard

often is mandated by the building code.

There is a heterogeneous array of

standards development organizations in

the United States engaged in standardsmaking activities. The standards

referenced in building codes referred to

above were developed by 91 different

organizations.

The American Society for Testing and

Materials (ASTM), the largest U.S.

standards organization, develops

approximately 8,500 standards. Six types

of standards are developed: "test

methods , " "specifications , " "practices ,

"terminology," "guides," and

"classifications." ASTM publishes a 3-

volume compendium of 663 standards [16]

specifically for building regulatory use.

3.3 Standards Development Processes

Most national standards in the United

States are produced through a "voluntary

system" made up of government and

industry, producers and consumers,

institutions and individuals. The system

is called "voluntary," since

participation of interested parties is on a voluntary basis; and the standards

produced are available for voluntary use

in building regulations and in building

contracts, when so used, ccmpliance with

the standard is mandatory.

The voluntary standards are produced by

two general processes, leading to

354
"industry" standards and "consensus"

standards. "Industry" standards are

those promulgated by a given industry or

industrial group. An industry standard

reflects an agreement by the promulgating

industry and is controlled by that

industry. Such standards are often

produced under limited participation

processes. Frequently, however, the

standard reflects the view of some affected parties as well.

"Consensus" standards are most often

developed by standards organizations

devoted to standards writing. Some

professional societies and industry trade

associations also develop consensus

standards as part of their activities

and, in doing, ensure the opportunity for

broad representation and participation.

Consensus standards reflect the view of

all interested and concerned parties.

Due to the rigorous procedural

requirements and broad participation in

the process, consensus standards

generally receive the highest level of

recognition by regulators and others

involved in the building process. They

are more readi ly accepted and used than

are industry standards developed under

limited participation processes.

The American National Standards Institute

(ANSI) serves as the coordinating body

for consensus standards produced by other

recognized, authoritative organizations.

It encourages the development of

standards through the consensus process,

and it is concerned with the well-being

of the U.S. standards generating system.

It also coordinates U.S. activities

related to international standards

participation, particularly the

International Organization for Standards

(ISO) and International Electrotechnical

Conmission (IEC). The principal

criterion for American National Standards

is the concept of "consensus" which must

be used during development. Consensus is

defined by ANSI as:

"Substantial agreement reached by

concerned interests according to the

judgment of a duly appointed

authority, after a concerted attempt

at resolving objections. Consensus

implies much more than the concept

of a simple majority but not

necessarily unanimity.

A standards developer may be accredited

by ANSI to use one or more recognized

methods of developing evidence of

consensus. These are (1) Accredited

Organization Method, (2) Accredited

Standards Committee Method, and (3)

Accredited Sponsor Using Canvass Method.

Each of these methods is further defined

and has as a guiding principle the

development of consensus for approval,

revision, reaffirmation, and withdrawal

of ANSI standards. American National

Standards are produced under a due

process, which means that any person,

organization, company, government agency, or individual with a direct and material

interest has a right to participate. Due

process allows for equity and fair play.

Minimum due process requirements include:

o Openness

Participation shall be open to

all persons who are directly

and materially affected by the

activity in question. Timely

and adequate notice shall be

provided of the initiation and

development of a new standard

or a substantively revised

standard.

o Balance

The standards development

process shall have a balance of

interest and shall not be

dominated by any single

interest category.

o Interest Categories

The interest categories

appropriate to the development

of a consensus in any given

standard activity are a

function of the nature of the

355
standard being developed, but

must include at least producer,

user, and general interest.

Written Procedures

Written procedures shall govern

the methods used for standards

development and shall be

available to any interested

person.

Appeals

An identifiable, realistic, and

readily available appeals

mechanism for the impartial

handling of substantive and

procedural complaints must be

in the written procedures

.

Notification of Standards

Development

Standards activities shall be

announced in suitable media, as appropriate, to demonstrate

provision of opportunity for

participation by all directly

and materially affected

persons.

Consideration of Views and

Objections

Prompt consideration shall be

given to the written views and

objections of all participants.

Consideration of Standards

Proposals

Prompt consideration shall be

given to proposals for

developing new standards or revising or withdrawing

existing standards.

Records

Records shall be prepared and

maintained to provide evidence

of compliance with the written

procedures.

4. PRODUCT APPROVAL

4.1 Introduction

Product approval and product acceptance

have different meanings to different

users. Product acceptance, on the part

of designers, contractors, and owners,

suggests the need for assurance that the

product will perform as expected under

conditions of intended use. Regulators

of building construction, on the other

hand, look at product approval from the

viewpoint of assurance that the

regulatory needs of the jurisdiction will

be complied with. Product approval, in

response to regulatory needs, is the use

for which the following discussion

generally applies.

There is no coherent national product

approval system in the United States. A

recent study of the National Institute of

Building Sciences [17] j describes the

current procedures. Product approval, by

and large, is dependent upon the final

judgment of the building official.

Taking into account the seriousness of

the consequence of a potential failure,

he makes his judgment based upon the

information available. Building

officials in the United States rely on

others to provide assistance in

evaluating products for their compliance

with the intent of the building code.

The essential elements contributing to

product approval are standards, testing,

and certification that the product or

service supplied meets the applicable

standard or other criteria. In the case of conventional products meeting

recognized standards, the problem is

somewhat less difficult than approval of

innovative products for which standards

have not been developed. The regulator

relies on laboratory reports, experience

and certification, be it selfcertification or certification by another

third party, and evaluation systems of

others. The model building code

organizations and their umbrella

organization, CABO, provide evaluation

services which are widely relied upon by

the building officials. Although there

356
is a lack of uniformity in approval of

conventional products vhich are produced

to a standard, it is in the area of

innovative products where standards do

not exist that the most difficulty lies.

Lack of standards and performance

criteria discourages innovation and

places the regulatory official in a

difficult position of accepting or

rejecting innovation. If there is doubt,

he must primarily be concerned for the

regulatory requirements and health and

safety of the occupants.

4.2 Evaluation Services

Evaluation systems of each of the model

codes and the National Evaluation System

(NES) of CABO are highly regarded and

widely relied upon by building officials

for making judgment relative to the

compliance of conventional products with

the model building codes and for

evaluation of innovation. These

organizations evaluate submissions of

proponents describing products with

supporting documentation. Based upon

information submitted and additional data

which may be called for, determination is

made as to the compliance with specific

requirements of the model cedes. Reports

when issued are distributed to the

menbership of the respective model code

organizations and are available to other

interested parties upon request. Each of

the model code organizations issues

hundreds of reports including those

developed under the NES of CABO. A broad

range of conventional and unusual

products are covered. In recent years,

evaluation services have been formalized

and are gaining in credibility and

acceptance. The model code organizations are developing acceptance criteria for use by proponents of building products

and systems to provide guidelines on

required performance to meet the code

intent. As of January 1989, ICBO has

issued 17 sets of acceptance criteria

addressing such broad subjects as

criteria for expansion anchors in

concrete, criteria for quality control

manuals, and criteria for residential

chimney liners. These criteria are

developed and adopted in conjunction with

public bearings conducted by the

evaluation committees of ICBO. They may

be revised from time to time through the

same public hearing process.

4.3 laboratory Accreditation

It is recognized generally that national

consensus standards provide for the

essential qualities of the products

falling under the scope of the standard.

Testing, however, does not have the same

credibility, particularly testing by

laboratories lacking recognized qualification for the testing. Laboratory

accreditation has come into being to meet

the need of further validation and

assurance of a laboratory's capability.

There are many different programs

available for the competence of testing

laboratories in the United States.

Laboratory accreditation serves to meet

the needs of:

o Laboratories seeking independent evaluation of technical

proficiency and national

recognition of their

competence,

o Industry requiring the services

of competent testing

laboratories

,

o Governmental regulatory agencies requiring evaluation of

public safety,

o Purchasing authorities requiring assurance of product

conformance to specifications

(standards), and

o Consumers seeking assurance of

product safety and performance.

Although there are many laboratory

accreditation programs in the United

States, only a few are national in scope

that relate to construction. The three

major national programs are briefly

discussed below.

357
NVLAP

The National Voluntary Laboratory

Accreditation Program (NVLAP) was

established in 1976. It is administered

by the National Institute of Standards

and Technology and is available on a

voluntary basis to all laboratories be

they public or private. In the area of

construction technology, NVLAP currently

has programs that address thermal

insulation, carpeting, construction

testing services, solid- fuel room

heaters, acoustical testing services, and

seals and sealants.

A2LA

The American Association for Laboratory

Accreditation (A2LA) Program was

established by the American Council of

Independent Laboratories, in 1978, by a

group of concerned individuals to develop

a management system to verify a wide

variety of testing laboratories. A2LA

grants accreditation in a field of

testing and, in seme cases, on specific

standards. Fields of testing currently

recognized which relate to construction

regulation are construction materials,

geotechnology, and thermal testing.

AASHID

In 1988, the American Association of

State Highway and Transportation

Officials (AASHID) initiated their

laboratory accreditation program. This

program addresses soils, concrete and

asphalt materials. It is standards

specific and implements ASTM Standards

E994 [18] and E548 [19]. The criteria

for the evaluation of laboratories in

specific fields of construction materials

testing are based on the following ASTM

standards:

ASTM E3666-83 Standard Practice tor

Evaluation of Inspection and Testing

Agencies for

Bituminous Pavement

Materials

ASTM E&561-86 Standard Practice for

Quality Control

Systems for an

Inspection and

Testing Agency for

Bituminous Pavement

Materials

ASTM D3740-80 Standard Practice tor

Evaluation of Agencies Engaged in the

Testing and/or

Inspection of Soil

and Rock as Used in

Engineering Design

and Construction

ASTM C1077-87 Standard Practice far

Laboratories Testing

Concrete and Concrete

Aggregates for Use in

Construction and

Criteria for Laboratory Evaluation

These ASTM standards include criteria on

the quality assurance system of the

laboratory, qualifications and training

of laboratory personnel, evaluation of

laboratory equipment, and evaluation of

the ability of laboratory personnel to

conduct tests according to the standard

methods of test for which accreditation

has been requested.

4.4 Certification

Certification programs play an important

part in quality control/quality assurance programs of the United States as related

to building products and related

services. The extent of these services

is not well known and the effectiveness

of certification in assuring quality and

compliance with standards has not been

analyzed comprehensively. Some certification programs appear to be well run,

well cTocumented, and very reliable.

Others appear to provide little that

contributes to quality control and

quality assurance. Certification is a

process by which a given product is

asserted to meet a specific standard or

358
other specific criteria. Certification

generally takes one of two forms, either

certification by a third independent

party or self-certification by the

producer or proponent of the product.

NBS Special Publication 703 [20]

identifies 109 private sector organizations in the United States which engage

in product certification activity. The

directory includes organizations which:

administer programs and certify that

products meet some criteria; administer

programs using an independent third party

certifier; or serve as an independent

third party certifier. This report

identifies only product-related programs.

It does not address programs certifying

services or professional skills, nor doss

it include programs which do not specify

criteria or standards which a product

must meet in order to be approved. An

extensive effort was made to make this

document complete when it was prepared in

1985. It is interesting to rote that 73,

out of a total of 109, of the organizations identified are agencies which

certify construction products, firm this

information, it appears that the

construction industry is the sector most

dependent upon certification programs for

regulatory compliance and quality

assurance.

The American National Standards Institute

has two relevant standards: 234. 1,

covering third party certification

programs, and Z34.2, covering selfcertification. ANSI also has a program

for accrediting certifiers. This program

for accrediting certifying agencies had

only two agencies in the program at the

end of 1988, both construction product

certifiers. As of April 1989, a third

organization has been accredited, and two more certifying agencies have applied and are being processed for accreditation.

All five of these agencies are primarily

concerned with construction products.

5. PRESSURES KR CHANGE

5.1 Introduction

Building regulations and regulatory

practices are under steady pressure to

incorporate advances in knowledge from

research. In areas such as earthquake

and wind engineering, energy conservation, and indoor air quality, provisions

are updated to incorporate recent results

and enforcement procedures are modified

to test and assure compliance. However,

two additional main trends are causing

major changes in building regulations and

regulatory practices. One is the effect

of advanced computation and automation - changing what we build and how we build

to take advantage of the growing

information-processing capabilities based

on microelectronics. The other is

international harmonization of building

regulations and regulatory practices,

stimulated by the European Community

effort to create a single European market

by the end of 1992, and motivated by an

international desire to reduce non- tariff barriers to beneficial trade in

construction products and services.

5.2 Advanced Computation and

Automation

Advanced computation and automation will

provide great changes in what we build

and how we build [21] . Building standards and codes, written on paper, can be

replaced by computer programs that will

incorporate the requirements of the code,

provide automatic data processing

capabilities in applying the code in

specific instances, and guide the regulator in evaluating the building material,

component or system under consideration

[22]. Traditional, printed standards and

codes can be replaced by "kncwlfidge-based

expert systems." These techniques facilitate updating standards and codes to

introduce new technologies. Moreover,

they facilitate checking revisions for

consistency with other provisions

.

Establishment of a computer-based

information system for a project, and the

development of information interface

technologies allowing automatic exchange

of data between the project information

system and the information system of the

359
regulator, will allow regulatory

evaluation of a project without physical

exchange of plans and specifications

between the designers and the regulatory

authority.

Advances in computers, making them

smaller, faster and cheaper, allow much

improved measurements of the properties

of building materials , components and

systems in the factory or in the field to

assist in the enforcement of building

regulations

.

Most of the above efforts presently are at the research stage, but improved

textual representations of standards and

building codes are already available in

computer-based information systems.

5.3 International Harmnrri7.aHnn

The European Ccnraunity (EC) has work

underway to permit free flow of goods,

services, capital, and people within

western Europe by the end of 1992. Plans

to implement this objective are underway

in European construction. The EC is

actively developing a program for

standards, regulations, certification and

testing which will make products and

services acceptable in all of the EC

countries.

This action was brought about by the

Single European Act adopted in 1986. The

basis for the Act is that any product

legally manufactured in an EC member

state must, in principle, be admitted to

the markets of all the other members.

This requires unification and acceptance

of various systems of building standards,

building regulations, product testing and

certification. The existing European

national standardization bodies have

underway an immense effort to harmonize

existing standards and develop new ones

so the construction products will be

acceptable to all the countries. The

standards being developed will be

European Ccranittee for Standardization

(CBM) standards, not International

Organization for Standardization (ISO)

standards [11]

.

In all of this effort there is little

U.S. involvement or input. However, a

profound effect on U.S. building regulations and regulatory practices can be

anticipated. The General Agreement on

Tariffs and Trade and bilateral

agreements, such as the U.S. -Canada free

trade agreement of 1988, call for the

elimination of non- tariff barriers to

trade. Therefore, U.S. building

standards, building regulations, and

product approval systems must be

harmonized to allow trade between the

United States and other countries in

building products and services. This

process will lead to products and

services meeting U.S. requirements being

eligible, without further testing or

approval, for international trade, and

comparable access of foreign products and

services to the U.S. market.

6. SUfttKY AND TRENDS

Health, safety and welfare regulations

for buildings are fragmented and vary

greatly, both in technical requirements

and their administration. There is

widespread interest in improving the

regulation of buildings, with much

attention given to this subject by

building officials, design professionals,

government at all levels, and the public.

6.1 Building Code Development and

Adoption

o Adoption of model building

codes'

Many national organizations and

ccnmissions have reccnmended

the adoption of one of the

three major model building

codes -- National, Standard, or

Uniform. These model codes are technically similar, are

updated annually, and reissued

every three years. The

adoption of model building

codes without amendment, to the

extent feasible, is encouraged

by many building industry

organizations.

360
Develop uniform

model codes

format for

The National Institute of

Standards and Technology and

the National Institute of

Building Sciences , among

others, have studied the

development of a uniform format

for use in the model building

codes. The American Society of

Civil Engineers and the

American Institute of

Architects have reconmended

that the model code organizations develop a uniform format

to facilitate model code usage

and updating so as to be

technically and administratively harmonious and as uniform as is practicable.

o Adoption of statewide building

codes

A task force, composed of the

National Association of Home

Builders, NCSBCS, and the model

code organizations , has

proposed that each state adopt

consistent building regulation

on a statewide basis with

enforcement at the local level.

This is envisioned as a means of improving construction

productivity.

o Recommend changes for updating

building codes

When potential improvements in

the technical content of building codes and their usage are

noted, architects, engineers,

contractors, and owners are being encouraged to submit

proposed changes to the model

code promulgating body to

update code content and to

facilitate its use.

6.2. Code Administration and EnfoicaiBTt

Architects, engineers, and contractors

are concerned with the processes used for

administration and enforcement of

building regulations at the local level.

Particular items of concern include time

required to receive permits, recognition

of pre-permit counseling, one-stop

permitting, and procedure for appeals

from the building official's decision.

Model administrative

enforcement processes

and

Model processes from NSBCS are available for implementation.

o Appeals boards

Design professionals are being

encouraged by their professional societies to seek

membership on code appeals

boards.

o Training of building officials

Training and certification of

building officials at all

levels is being implemented by

model code organizations and

states with support by

contractors and professionals.

o Use of computers and other aids

Automation of the building

administration and enforcement

process to improve the quality

and reduce the time required

for building regulation is

rapidly growing.

6.3 Product Approval

State, local, and Federal gpvemmiLs and

the private sector are studying the

development of a national product

approval system meeting the requirements

given in section 4. A product approval,

based on nationally or internationally

recognized standards, should be available

locally for acceptance of products in

national or international trade. This

will require: (1) developing an active

U.S. advocacy role in international

361
standards activities, and (2) establishment of a coherent system for acceptance

of innovative building products (those

for which there are no national standards or products for vhich nonstandard use is

proposed) and Improvement of the

acceptance and quality assurance of

products for which there are applicable

international standards.

6.4 Automated Exchange of

Construction Information for

Regulatory Review

Machine representation of standards and

codes is under development to allow

efficient, automated review of compliance

of designs. Moreover, these techniques

could be used to assure consistency vhen

standards or codes are modified to

improve provisions. The family of

national and international standards

required, for automatic exchange of con- struction data (drawings, specifications,

etc.) between the diverse computing

systems of owners, designers, and

regulators is being developed to improve

the economy, accuracy and timeliness of

plan review. These capabilities also

will facilitate transmittal of plans and

specifications to manufacturers,

fabricators at the construction site, or

the transmittal of changed conditions

from the site to engineering and

regulatory offices.

7. REFERENCES

1. Sanderson, R.L. , Codes and Code

Administration. Building Officials

Conference of America, Chicago,

1969.

2. The Report of the President's

Conniission on Housing. Washington,

1982.

3. Cooke, P.W., "A Review of U.S.

Participation in International

Standards Activities," NBSIR 88-

3698, National Bureau of Standards,

January 1988.

4. Working with the Performance

Approach in Building. Publication

64, International Council for

Building Research, Studies and

Documentation, Rotterdam, January

1982.

5. Dikkers, R.D. , "Model Documentation

for Building Regulation," TN 775,

National Bureau of Standards, May

1973.

6. Wiggins, J., et.al., "Land Use

Planning and Natural Disaster

Mitigation, 11 in Building Practices

for Disaster Mitigation. BSS 46,

National Bureau of Standards,

February 1973.

7. National Manufactured Housing

Construction and Safety ' Standards

Act of 1974, P.L. 93-383; 88 Stat.

633, 42 U.S.C. , Section 5401.

8. Field, C.G., "The Changing Building

Regulatory System," Pacific Rim

Conference of Building Officials,

April 1989.

9. Gross, J.G. , Pielert, J.H. , Cooke,

P.W., "Impact of Building

Regulations on Rehabilitation - Status and Technical Needs," TN 996,

National Bureau of Standards, May

1979.

10. National Institute of Building

Sciences, Rehabilitation Guidelines

1980 . Department of Housing and

Urban Development, 1980.

11. Cooke, P.W. , "A Summary of the New

European Community Approach to

Standards Development," NBSIR 88-

3793; National Bureau of Standards,

June 1988.

12. Stephens, K. , "U.S. , Europe Draft

Accord on Acceptance of Engineering

Credentials," En

gineering Times . National Society of Professional

Engineers, March 1989.

362
13. "Directory of State Building Codes

and Regulations , " National

Conference of States on Building

Codes and Standards, " 1987.

14. "Administration and Enforcement of

Building Codes and Regulations,"

Report E-l, Construction Industry

Cost Effectiveness Project, The

Business Roundtable, 1982.

15. Vogel, B.M. , "Standards Referenced

in Selected Building Codes," NBSIR

76-1140, National Bureau of

Standards, 1976.

16. "ASTM Standards in Building Codes,"

American Society for Testing and

Materials, 1988.

17. "Building Product Approval/Acceptance Processes," National Institute

of Building Sciences, June 1987.

18. "Standard Guide for Laboratory

Accreditation Systems," ASTM E994-

84.

19. "Standard Practice for the

Preparation of Criteria for Use in

the Evaluation of Laboratories and

Inspection Bodies," ASTM E548-84.

20. Breitenberg, M.A. , "Private Sector

Product Certification Programs in

the United States," SP 703, National

Bureau of Standards, 1985.

21. Wright, R.N. , "Ccnputer Integrated

Construction," Proceedings, International Association of Bridge and

Structural Engineers, February 1988.

22. Wright, R.N., Lyons, J.W., "Machine

Representation of Standards,"

Standardization News . ASTM, August

1986.
THE BUILDING PROCESS

AND THE

REGULATORY SYSTEM

Review Enforcement

Figure 1. The Building Process and Regulatory System

THE REGULATORY SYSTEM

Figure 2. Roles in the Building Regulatory Process

THE REGULATORY SYSTEM

WITH SUPPLEMENTAL EVALUATION

Figure 3. Supplementary Evaluation in the Building Regulatory System

364
Appendix:

Task Committees

A-J Reports

Report of Task Committee on

(A) STRONG-MOTION INSTRUMENT ARRAYS AND DATA

Date : May 18, 1989

Place : Public Works Research Institute

Ministry of Construction

.1, Asahi, Tsukuba-shi, Ibaraki 305

Attendees: Japan Side -

U.S. Side -

H. Tsuchida (Chairman)

K. Ohtani

M. Seino

T. Kuwabara

K . Tamura

A. G. Brady (Chairman)

J. F. Lander

( PHRI

)

(NRCDP)

(MRI)

(PWRI)

(PWRI)

(USGS)

(Univ. of Colorado)

Following the request in item 8 of the Final Resolutions adopted at

the 20th Joint Meeting held in Gaithersburg in 1988, the "Mission

* Statement" of this Task Committee was drafted for adoption by the Panel as

follows

;

MISSION STATEMENT

1. Objective

The objective of the task committee is to coordinate the provision of

strong-motion earthquake data to researchers and to the practicing

engineering community. Task committee members are involved with

instrumentation; recording, processing, and analyzing strong-motion data;

research directed to ensuring high-quality data disseminated to users; and

the analysis of ground motion and structural motion in dynamic response to

earthquakes

.

2. Scope of Work

The T/C coordinates strong-motion related research, processing

strong-motion data, and disseminating information on the recorded behavior

of structures. In addition to regular exchange of data and publications,

the T/C's technical approach includes:

* Plan and conduct T/C workshops and meetings, held generally in

conjunction with UJNR joint meetings; often with IASPEI/IUGG

conferences and with World Conferences on Earthquake Engineering.

* Conduct related technical sessions at professional society

meetings.

367
* Create a procedure for the dissemination of significant

strong-motion digital data with regard for the rights and

expectations of the owner, the users of data, and the earthquake

engineering community.

* Create an up-to-date information exchange on computer processing

procedures and related research so all users of processed data are

aware of valid ranges and practical limitations.

* Develop a research program for appropriate instrumentation and

dynamic analysis of structures with base isolation or other

response-reducing systems, under earthquake excitation.

* Develop a research program using temporary arrays and data derived

to study site response.

3. Accomplishments

* A workshop on strong-motion earthquake recording was held at USGS

in Menlo Park, CA, August 1987.

* The US side presented the Japanese side with a manual and floppy

disks containing a global strong motion data catalog and an access

program prepared by NOAA's National Geophysical Data Center in May

1989.

r * A Workshop was held in Tokyo on 6 August 1988 on the Processing of

Strong-Motion Records, under the sponsorship of IASPEI, and

International Association of Earthquake Engineering (IAEE). The

workshop was scheduled to coincide with the 9th World Conference on

Earthquake Engineering (9WCEE), 2-9 August 1988 in Tokyo and Kyoto.

Task Committee chairmen and members were responsible for the

organization and review of preparatory processing results by

participants. A pre-proceedings containing these results was ready

for the workshop; a final Proceedings of the .workshop has been

published. A total of 52 attended; 31 Japanese, 21 others.

* A Workshop was held on 1 August 1988 in Tokyo on the effects of

surface geology; T/C members provided significant involvement.

* All Japanese T/C members and several US members attended the 9WCEE

in Tokyo and Kyoto, 1988. Many papers addressing the objectives of

the T/C were presented.

* Dr. A.G. Brady lectured to three groups of Japanese earthquake

engineers, including PWRI and BRI, during his stay in Japan for the

9WCEE

.

368
* Data tapes and reports on the Whittier Narrows, CA earthquake of 1

October 1987 were delivered to the Japanese T/C at the 9WCEE.

4. Future Plans

* The T/C Chairmen will investigate funding sources for a 3rd

Workshop on Processing Strong Motion Records, concentrating on

array data.

* Coordinate the analysis of records from arrays: downhole 3-D arrays

(geophysics and earthquake engineering) and structural arrays

(structural engineering).

* The T/C recognizes the efforts of the International Working Group

on the Effects of Surface Geology and encourages wider

participation by Panel members.

5. Information Exchange

* will be performed within the domestic community through technical

meetings, conferences, seminars, and workshops.

* will be performed through participation in post-earthquake field

investigations with concentration in areas where strong-motion

stations have recorded data.

* will be performed with the counterpart side through T/C workshops;

visitor exchanges, seminars, lectures; and participation in annual

Joint Panel meetings.

* The most crucial information exchange is rapid description of

recovered strong motion (and subsequently digital strong-motion

data on tape or disk) from significant earthquakes, for example

Whittier Narrows, CA of 1 October 1987, M5.9; Supersition Hills, CA

24 November 1987, M6.6; Eastern Chiba Offshore, 17 December 1987,

MB 6.0; South Coast of Honshu, JAPAN, 18 March 1988, MB 5.5.

6. Impact

* Most managers who process strong-motion recordings use techniques

based on, or closely associated with, the work of this T/C and its

members. Foreign networks fitting this category include:

AUSTRALIA - Australian Seismological Center, Bureau of Mineral

Resources, P0 Box 378, Canberra ACT 2601.

CANADA - Pacific Geoscience Center, P0 Box 6000, Sidney, BC. V8L

4B2.

ITALY - The Italian National Commission for Research and Development

369
on Alternative Energy Sources (ENEA), Rome.

TURKEY _ University of Bogagici, 80815 Bebek, Istanbul.

YUGOSLAVIA - Institute of Earthquake and Engineering Seismology,

Skopje.

* Use of seismological and strong-motion data, in convenient form, in

countries other than where recorded.

* Dissemination of digital data from significant earthquakes

increases the research data base; application of this data in

design practice aids in reduction of earthquake hazards.

Barriers

* Difficulty of funding for T/C joint endeavors.

* Joint research efforts with close interaction between participants

are difficult unless visiting research appointments can be made.

370
Report of Task Committee on

(B) LARGE-SCALE TESTING PROGRAM

Date : May 17, 1989

Place : Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki-ken 305

Attendees ; Japan Side - K. Ohtani (Chairman) (NRCDP)

Y. Yamazaki (BRI)

S. Nakata (BRI)

H. Yamanouchi (BRI)

Y. Koga (PWRI)

K. Yokoyama (PWRI)

U.S. Side - H. S. Lew (Chairman) (NIST)

G. R. Fuller (HUD)

Following the request in item 8 of the Final Resolutions adopted at the

20th Joint Meeting held in Gaithersburg in 1988, the "Mission Statement" of this

Task Committee was drafted for adoption by the Panel as follows,-

"MISSION STATEMENT"

1 . Objective

The objective of the Task Committee is to develop performance data of

full-scale dynamic properties of structures to better resist wind and seismic

loads through both laboratory testing of prototype structures and field testing

of structures in situ. Improved full-scale test data will validate the results

of small-scale model tests and substantiate the results of computer analyses of

structural behaviour.

2 . Scope of Work

The Task Committee develops its research agenda in coordination with other

appropriate T/Cs, such as T/C C and D, for full-scale evaluation of buildings,

and other structures except bridges

.

* Plans and conducts workshops and joint meetings to identify research

topics and develops joint test programs

.

* Coordinates research projects carried out by various laboratories in

the U.S. and Japan. Facilitates publication of research results and

implementation of findings in codes and standards

.

* Facilitates exchange of research personnel, technical information and

available testing facilities.

3 . Accomplishments

* The second phase of the U.S. -Japan Coordinated Program for Masonry

Building Research has been funded by the National Science Foundation

for U.S. researchers. Both analytical and experimental research,

including component and subassemblage tests are being carried out at

several U.S. universities.

* The Japan side published the technical document entitled "Design

371
Guidelines for Medium Rise Reinforced Masonry Buildings". Taking the

opportunity of the publication of the document, the Reinforced Masonry

(RM) Building Promotion Society of Japan was established in May 1989

to further the RM building systems and disseminate technical

informations

.

* The 4th Joint Technical Coordinating Committee on Masonry Research

( JTCCMAR) was held in October 1988 at San Diego, CA. Progress on

individual research projects were presented by principal investigators

and future coordination of the Joint Program was discussed.

* The results of a full-scale 5-story masonry building test performed

in Japan were published in the Journal of the Masonry Society in 1988.

* Precast seismic structural systems was selected as the fourth topic in

the Joint U.S. -Japan Program on Large-Scale Testing during the 20th

Joint Panel Meeting. A meeting of the U.S^ Planning Group was held at

the University of California at San Diego on July 1988. The scope of

precast structure research program in the U.S. and the planning for a

T/C workshop were discussed.

* The first workshop of the Precast Seismic Structural Systems (PRESSS)

was held on November, 1988 at San Diego, CA. The purpose of the

workshop was to develop a comprehensive U.S. research agenda for

PRESSS. Over 35 research ideas were presented; they are being

evaluated by the executive committee to develop a comprehensive

multi-year research program. On the U.S. side, the proposal will be

submitted to the National Science Foundation for funding in 1990.

* On the Japan side, two preliminary meetings were held to discuss the

research program for precast seismic structural systems . The

Government of Japan approved funding for the PRESSS research program

to begin in 1989.

4 . Future Plans

* The testing of a five-story full-scale masonry structure will be

carried out at the University of California at San Diego beginning

October 1989. The U.S. Technical Coordinating Committee on Masonry

Research will be held during the first week of July 1989.

* The 5th meeting of the JTCCMAR is scheduled in October 1989 at

Tsukuba, Japan. The joint masonry program will be reviewed including

the testing requirements for a full-scale 5-story masonry structure

in the U.S.

* A Joint U.S. -Japan PRESSS Program Coordinating Meeting is planned

for October 1989 in Japan. The scope and schedule of the Joint

Coordinated Program of Precast Seismic Structural Systems will be

discussed in detail during this meeting.

* Continue exchange of information of the following topics between

appropriate U.S. and Japan organizations.

a. Seismic performance of composite and mixed construction;

a U.S. -Japan Joint Workshop is being planned,

b. Application of high strength 'construction materials, such as

concrete and steel, to structures in high seismic zones,

c. Application of base isolation and active controlled systems to

372
critical structures located in high seismic and wind zones , and

d. Testings of large-scale structures other than buildings and

bridges

.

* Exchange of information on large-scale testing facilities and

large-scale testing programs should be encouraged.

5 . Information Exchange

* Technical reports and research documentation were exchanged with

participating organizations in both countries.

* The results of the 5-story full-scale test of masonry structure in

Japan were published by the Journal of the Masonry Society thereby

widely disseminating the T/C's research.

6 . Impact

* The results of the past joint research projects on reinforced concrete

and steel structures have been published in U.S. technical journals.

A number of key findings have been incorporated into U.S. and Japanese

building codes and design guidelines

.

* As a results of joint research projects, the state of practice and

understanding in earthquake engineering has been improved in both

countries . The results from the joint program have further stimulated

research in both countries

.

* Facilitate joint research projects utilizing available testing

facilities in both countries.

7. Barriers

* There are no major technical barriers for the attainment of the Task

Committee mission. However, different building practices and needs

require careful planning for effective joint research projects.

* Securing funding to support exchange of research personnel is

difficult in both countries. Funding agencies of both countries

should be informed about the technical benefits resulting from the

joint panel activities and encouraged to participate researcher

exchanges and joint research programs.

373
Report of Task Committee on

(C) REPAIR AND RETROFIT OF EXISTING STRUCTURES

Date: May 18, 1989

Place : Public Works Research Institute

Ministry of Construction

1 , Asahi , Tsukuba-shi , Ibaraki 305

Attendees: Japan Side - M. Hirosawa (Chairman)

M. Fujiwara

S. Nakata

(BRI)

(PWRI)

(BRI)

U.S. Side - H. S. Lew (Acting Chairman) (NIST)

Following the request in item 8 of the Final Resolutions adopted at

the 20th Joint Meeting held in Gaithersburg in 1988, the "Mission

Statement" of this Task Committee was drafted for adoption by the Panel as

follows

;

Mission Statement

1 . Objective

The objective of the Task Committee is to encourage research and provide

technology transfer activities in materials, components, and total building

systems related to evaluating the performance of existing building

subjected to environmental and earthquake forces. The research topics

provide data and information for the repair and retrofit of existing

buildings.

2. Scope-of-Work

The T/C work includes information exchanges and planning and hosting

workshops and seminars. Workshops are generally held in conjunction with

the annual Panel's meeting. They provide T/C members with an opportunity

to review related research and to direct plans for future research, share

and implement results. The T/C conducts related technical sessions at

professional society's meetings. The T/C studies new materials and methods

to accomplish repair and retrofit operations on buildings e.g., fiberglass

reinforced plastics, and exploits use of automation and robotics for repair

and retrofit. Work focuses on coordinating research projects in Japan and

the U.S. to minimize duplication and to maximize benefits.

3. Accomplishments

Several accomplishments were achieved during the past two years:

* Held the 4th Workshop on Repair and Retrofit of Existing

Structures at the 19th Joint Meeting, JAPAN, 1987.

374
* Held a workshop to develop a research agenda for repair and

retrofit of different types of buildings, September 1987, in San

Francisco. The workshop proceedings were published.

* FEMA published cost data for various methods of repair and

retrofit of existing buildings.

Future Plans

* A joint workshop is being planned on the design approaches and

construction methods for repair and retrofit of buildings, May

1990 in the U.S. The workshop will focus on the use of newly

developed methods for repair and retrofiting of buildings.

* In the U.S., FEMA will publish methods to strengthen existing

buildings.

* In Japan, the Guidelines for Evaluating and Strengthening Existing

Buildings will be revised, and the Guidelines for Evaluation of

Earthquake Damage and Repair Methods of Buildings will be

published.

Information Exchange

* Publications of T/C workshops were widely distributed and made

available to the design professionals through technical

presentations at workshops and at other national meetings of

professional societies.

* Exchange of information with the JAPAN-side has been achieved

using the published workshop reports and individual's research

papers

.

Impact

* Contributed to development of US and Japan design criteria for

repairing reinforced concrete, steel, and other buildings.

* T/C recommendations were incorporated into building codes and

professional practices as illustrated by rehabilitation programs

used by the City of Los Angeles to upgrade existing masonry

buildings to resist earthquake forces.

* Influenced U.S. and Japan design practices to upgrade existing

structures

.

Barriers

* Greater progress need be made to implement available research

results. One approach is to create' a central coordinating group

who would be in contact with researchers, professionals, and

decisionmakers. This group would maintain a library of the T/C

activities and of researchers participating in related meetings

and workshops. A data bank of reports could be established and

made available by electronic methods and accessed by personal

375
computers. This barrier would be eliminated when sufficient funds

are identified to support this endeavor including the

identification of appropriate person/organization to lead the

effort.

Securing funding to support the workshop participants is difficult

in both countries.

376
Report of Task Committee on

(D) EVALUATION OF STRUCTURAL PERFORMANCE

Date : May 18, 1989

Place: Public Works Research Institute

Ministry of Construction

1 , Asahi , Tsukuba-shi , Ibaraki 305

Attendees : Japan Side - T. Murota (Acting Chairman) (BRI)

Following the request in item 8 of the Final Resolutions adopted at

the 20th Joint Meeting held in Gaithersburg in 1988, the "Mission

Statement" of this Task Committee was drafted for adoption by the Panel as

follows

;

MISSION STATEMENT

1 . Objective

Thousands of structures in each country are considered inadequate to

resist probable high winds or earthquakes. An important aspect of disaster

mitigation is determining the capacity of existing structures to resist

wind and seismic forces, so necessary repair, strengthening, and retrofit

can be planned. To provide adequate performance evaluation, each country

must coordinate development of condition assessment, screening, and

structural analysis methodologies. Structures must then be analyzed and

instrumented, and evaluated after disasters.

2. Scope of Work

* Develop a complete inventory of "benchmark" structures that have

been analyzed and instrumented for measurement of wind and seismic

response

.

* Coordinate development of a uniform system of screening and complex

structural analysis of wind and seismic resistance and capacity of

structures in each country.

* Develop advanced sensor technology, instrumentation, and "expert"

systems to provide condition assessment of existing structures.

* Evaluate masonry and precast concrete buildings, and

seismic-isolation systems, and include these findings in each

country's catalog of "benchmark" structures for post-disaster

performance evaluation.

* Plan and conduct workshops at future UJNR Joint Panel Meetings,

H. Yamanouchi

K . Kurashige

(BRI)

(MRI)

U.S. Side G. R. Fuller (Chairman)

K. C. Mehta (Temp. Member)

(HUD)

(Texas Tech. Univ.)

377
cooperatively with Task Committees "C" and "B".

3. Accomplishments

* Exchanged several post-earthquake investigation reports prepared byeach country, such as Japan Sea, Mexican, and Whittier Narrows

Earthquakes

.

* Presented Applied Technology Council ATC-14 Report on "Evaluating

the Seismic Resistance of Existing Buildings," which enables

engineers to analyze the seismic resistance of existing buildings

and to identify structural weaknesses and areas needing

strengthening (7/88).

* Exchanged documents on "Measurements of High Winds on High Towers

at Kashima Coast (3/87), and "Report of Observation at

Meteorological Observation Tower No. 2" by Meteorological Research

Institute in Japan.

* Received Japanese MOC first and second year reports on state-of-art

and catalog of buildings with seismic isolation or wind damping

systems. The U.S. Side is exploring ways to have the documents

translated into English (5/89).

* The U.S. Side sponsored an "International Workshop; on Sensors and

Measurement Techniques for Assessing Structural Performance." A

research agenda for sensor technology and measurements related to

structural performance was developed (9/88).

* A research workshop sponsored by NSF and coordinated by UJNR Panel

members was held in Washington, DC to develop a U.S. research

agenda for masonry. A working group was related to structural

performance of masonry buildings. Evaluation techniques,

nondestructive testing, and condition assessment were identified as

priority research needs (8/88).

* As a part of the U.S.TCCMAR Program, completed a project on

"Nondestructive Evaluation of an Existing Masonry Building" in

Williamsburg, VA (6/88).

* Meetings of U.S. TCCMAR and The Masonry Society (TMS) was held in

Salt Lake City, Utah in January 1989. Discussed nondestructive

testing methods and analysis of masonry buildings for resistance to

wind and earthquake forces.

4 . Future Plans

* Develop a reporting system and data collection form to catalog

existing structures in the U.S. and Japan that have been evaluated

for wind and seismic resistance or have been instrumented. Compile

a computer data file for exchange between countries and report its

progress at each UJNR Joint Panel Meeting.

* Encourage and enlist the participation of private industry,

consulting engineering firms, universities, and other governmental

378
agencies involved in the instrumentation, evaluation, and condition

assessment of existing structures.

* Encourage each country to develop sensor technology and

instrumentation of structures, for making condition assessments and

for determining the long term envir'onmental and aging effects on

building components and materials.

* Translate and disseminate two Japanese MOC Reports on seismic

isolation systems and wind damping systems into English.

* Plan with T/C "C" a workshop to be held in conjunction with 22nd

UJNR Panel Meeting in US and ASCE Structures Congress in Baltimore,

MD in May 1990.

5. Information Exchange

Exchange methodologies for evaluation of structural performance and

reports of actual post disaster evaluations of structures.

Provide updated data of "benchmark" structures evaluated and

instrumented

.

Continue the collection of related technical papers and literature,

and encourage development of an annotated bibliography of

information for periodic exchange.

Prepare papers and reports for annual UJNR Joint Panel Meetings and

associated workshops.

Exchange reports on U.S. ASCE Structures Congress held in San

Francisco, CA, in May 1989, Sixth U.S. Conference on Wind

Engineering held in Houston, TX in March, 1989, and similar

meetings in the U.S. and Japan dealing with evaluation of

structural performance of existing structures.

Obtain and evaluate the Texas Tech Report by Dr. Kishor Mehta on

structures damaged by tornadoes and hurricanes.

Obtain and exchange documents and tornado probability maps

developed under a US-NSF project.

Obtain complete information on California Division of Mines and

Geology (CDMG) instrumentation of approximately 130 structures (90

to 100 buildings, plus bridges and other structures.).

6 . Impact

* Enables the development of revised standards and building code

provisions to prevent destruction from high winds and earthquakes.

* Provides engineers with the capability to conduct condition

assessmen t of existing structures for the purpose of developing

cost effective repair, retrofit, strengthening or demolition

programs

.

379
* Will increase the confidence level of the public, regulators,

policy makers and engineers, to be assured that the majority of

major structures will perform adequately to resist high wind and

earthquake forces.

. Barriers

* Difficulty in identifying priorities and inadequate resources to

document load carrying capacity of existing buildings subjected to

highly probable extreme winds and earthquakes.

* Difficulty in developing probability risk maps for tornadoes,

typhoons , and hurricanes

.

* Limited resources for instrumenting buildings, for conducting

post-disaster evaluations, and for participating in foreign travel

for coordination of activities.

380
Report of Task Committee on

(E) NATURAL HAZARD ASSESSMENT AND MITIGATION THROUGH LAND USE PROGRAMS

Date : May 17, 1989

Place : Public Works Research Institute

Ministry of Construction

T, Asahi, Tsukuba-shi, Ibaraki 305

Attendees : Japan Side - K. Kawashima (Chairman) (PWRI)

S. Noda (PHRI)

K. Yokoyama (PWRI)

K. Tokida (PWRI)

K. Hasegawa (PWRI)

H. Matsumoto (PWRI)

U.S. Side - A.G. Franklin (Acting Chairman) (WES)

N. Raufaste (NIST)

A.G. Brady (USGS)

Following the request in item 8 of the Final Resolutions adopted at

the 20th Joint Meeting held in Gaithersburg in 1988, the "Mission

Statement" of this Task Committee was drafted for adoption by the Panel as

follows

;

Mission Statement

1 . Objective

Natural hazard assessment and mitigation activities represent the analysis

and application of a broad range of geological , seismological , and

engineering research to the problems of the earthquake resistant design of

buildings, land use planning and zoning, disaster preparedness, and

mitigation of damage and life loss. The Task Committee emphasizes the

improvement of methods for natural hazard assessment and techniques for

applying these assessments. Specific technical areas of concern are:

1. Probabilistic Earthquake Hazard Assessment

2. Attenuation of Seismic Waves

3. Prediction of Strong Motion

4. Seismic Zonation

5. Earthquake Losses

6. Delineation of Seismotectonic Elements

7. Application of Hazard Assessment to Seismic Provisions of Building

Codes

Estimation of natural hazard losses are important to the Committee because

these establish, in part, the quantitative bases for mitigation activities.

2. Scope of Work

* Plan and conduct T/C workshops and meetings. These meetings and

workshops are generally held in conjunction with annual Panel on

381
Wind and Seismic Effects joint meetings.

* Conduct cooperative research programs.

* Exchange research personnel

.

* Exchange information on a timely basis.

Accomplishments

* Papers have been presented at Panel on Wind and Seismic Effects

technical meeting during the past several years.

* Exchanged publications throughout the year. About 20-30 papers are

exchanged per year.

Future Plans

* Organize a workshop in Japan during 1990, for the purpose of

discussing and exchanging ideas on the state-of-the-art knowledge

and practice on earthquake hazard and risk (loss) assessment.

* Improve knowledge about researchers exchange programs by

disseminating information about fellowship programs and about other

financial assistance; and by exchanging of .information on

institutions, research programs, and related laboratory facilities

where opportunities for visiting researchers exist.

Information Exchange

* Between United States and Japanese sides through annual joint

meetings, T/C meetings and workshops.

* Visits by researchers and joint investigations of earthquakes.

* With the U.S. and Japanese communities through publications,

technical meetings, conferences, seminars and workshops.

Impact

* Increased public awareness of the importance of hazard and risk

assessment

.

* Improved public safety through implementation of results reported

at Panel on Wind and Seismic Effects meetings and workshops. As an

example of implementation is the use of probabilistic ground motion

maps in U.S. in Applied Technology Council and Building Seismic

Safety Council seismic design studies.

Barriers

* Lack of available funds for T/C workshops and travel to annual

meetings of Panel on Wind and Seismic Effects.

382
Report of Task Committee on

(F) DISASTER PREVENTION METHODS FOR LIFELINE SYSTEMS

Date : May 17, 1989

Place : Public Works Research Institute

Ministry of Construction

1, Asahi, Tsukuba-shi , Ibaraki 305

Attendees

:

Japan Side

K. Tamura (Observer)

T. Kuwabara ( " ) H. Sugita ( " )

Y. Sasaki (Chairman)

T. Murota

K. Tokida

( PWRI

)

(BRI)

(PWRI)

( PWRI

)

( PWRI

(PWRI)

U.S. Side A. G. Brady (Acting. Chairman) (USGS)

Following the request in item 8 of the Final Resolutions adopted at

the 20th Joint Meeting held in Gaithersburg in 1988, the "Mission

Statement" of this Task Committee was drafted for adoption by the Panel as

"follows

;

MISSION STATEMENT

1 . Objective

Lifeline systems such as gas, oil, water, and sewage pipelines; and power,

communication, and transportation systems are crucial to the survival and

health of a city or community. Earthquakes affecting such systems cause

severe social and economic disruptions to communities and cause human

suffering to the residents. T/C "F" focusses on improving 1) the behavior

of lifeline systems during earthquakes and 2) engineering and other seismic

countermeasures such as damage estimation techniques and inspection

procedures. These technologies will provide users with improvements to

existing standards of practice for safety and serviceability of lifeline

systems

.

2 . Scope of Work

Plan and conduct T/C workshops and meetings; they generally are

held in conjunction with the annual UJNR Joint Meeting. The

purpose is to exchange the state-of-art knowledge and practice, and

to identify cooperatively key opportunities for exchange and

studies which effectively implement research results.

Conduct related technical sessions at professional society

meetings

.

Create a research program on a selected lifeline system (similar in

concept to the Large-Scale Testing Program). For example, through

a joint demonstration, produce criteria for emergency evaluation

and on-line monitoring systems for lifeline operation.

383
* Conduct special activities such as translating, printing, and

distributing MOC PWRI's publication, Manual for Repair Methods for

Civil Engineering Structures Damaged by Earthquakes.

* Develop performance standards and a manual on repair, restoration,

and retrofit of lifelines.

Accomplishments

* Researchers of both sides presented more than ten papers on

lifeline systems at past Joint Panel meetings and discussed the

state-of-art and identified research needs.

* A U.S. - Japan Joint Workshop on lifeline earthquake engineering

was held in May of 1989 in Japan.

* Exchanged technical reports on disaster prevention for lifelines,

such as buried pipeline systems, and relevant specifications.

* Translated the MOC document on repair methods,

* NIST (formerly NBS) and PWRI have carried out cooperative research

on seismic loading effects on structural components of lifeline

systems such as column-like structures and buried pipes using

large-scale models.

* PWRI instrumented ductile pipes in soft ground at Sodegaura in

Chiba and started observation as one of the U.S. - Japan

cooperative research program.

Future Plans

* Instrument ductile pipes at different orientations to faults at

Parkfield, CA (sponsorship by NSF & NCEER) 5/89

* Continue observation of instrumented buried pipes and exchange data

* Distribute to US engineering community the English version of PWRI

Manual on Repair Methods for Civil Engineering Structures Damaged

by Earthquakes 10/89

* Develop a knowledge base for seismic disaster mitigation 5/90

Develop an engineering manpower database 5/90

* Conduct Joint Workshop on Disaster Prevention for Lifeline Systems

in the U.S /91

Information Exchange

* Domestic community: through technical meetings, conference

sessions, seminars, and workshops.

Between U.S. and Japan: through T/C workshops, and annual Joint

384
Panel meetings

.

* Visits by researchers and participation in technical meetings such

as NCEER-sponsored activities and post-earthquake field

investigations by EERI.

* Short-term (3 to 12 months) researcher exchanges to participate in

on-going lifeline research such as performed at the University of

Michigan and PCA in the U.S. and at BRI in JAPAN.

Impact

* Increased public awareness of problems with lifeline earthquake

engineering using results from workshops e.g., "Abatement of

Earthquake Hazards to Lifelines", Denver CO, Nov 1986.

* Contributed to increase field engineers' technical knowledge on

repair methods, through Seminars with lectures by T/C members using

the MOC Manual, in nine major cities in Japan including Tokyo and

Osaka; more than two thousand participants attended seminars.

* Contributed to the assessment of design and operating standards for

lifeline systems through T/C member participation in the Building

Seismic Safety Council.

* Contributed to the technical development program of the Technical

Council on Lifeline Earthquake Engineering (TCLEE/ASCE) via T/C

members participation in the committee work.

Barriers

* Insufficient funds to conduct annual Task Committee Workshops and

to publish reports.

* Lack of participation by U.S. industrial organizations and

utilities in research or in implementation of findings due to

caution on taking risks and a reluctance to adopt new technologies

and practices.

385
Report of Task Committee on

(H) SOIL BEHAVIOR AND STABILITY DURING EARTHQUAKES

Date May 18, 1989

Place Public Works Research Institute,

Ministry of Construction

1, Asahi, Tsukuba-shi, Ibaraki 305

Attendees Japan Side K.

T.

Y.

T.

S.

M.

Y.

H.

J.

Tokida(Chairman)

Iwasaki

Koga

Matsumoto

Noda

Okahara

Yamazaki

Matsumoto (Observer)

Koseki (Observer)

(PWRI)

(PWRI)

(PWRI)

(PWRI)

(PHRI)

(PWRI)

(BRI)

(PWRI)

(PWRI)

U.S. Side - A.

W.

C.

G. Franklin (Chairman)

E. Roper

E. Smith

(WES)

(CORPS)

(MMS)

Following the request in item 8 of the Final Resolutions adopted at

the 20th Joint Meeting held in Gaithersburg in May, 1988, the work of this

Task Committee was drafted for adoption by the panel as follows;

Mission Statement

1 . Objective

To enhance technology for predicting the dynamic behavior of soils,

for establishing procedures to analyze dynamic soil-structure interaction

and for modifying the dynamic behavior of foundations and earth

structures, to assure their safe performance during earthquakes. Government

agencies with responsibility for public works have a duty to the public to

provide for public safety and to make economical use of public funds.

Consequently, they have a need for more accurate methods of predicting soil

behavior, and for improved, economical methods of controlling soil

behavior, to assure the seismic safety of public works and to minimize the

costs of pconstruction by avoiding unnecessary over-design.

2 . Scope of Work

* At the annual joint panel meetings, present technical papers

describing technological developments and the state of the

art and practice related to soil behavior and stability during

earthquakes.

* Exchange relevant information and technical data relating to field

performance, research and methods of practice.

* Plan and conduct T/C workshops in coordination with proposed or

ongoing cooperative research programs.

386
* The task committee will plan and conduct programs of cooperative

research, by means of coordination of research tasks, exchange of

researchers between U.S. and Japanese government laboratories,

exchange of results, and technology transfer through joint

publication of research results and recommended practice. By

bringing greater resources to bear on problems of mutual interest,

and by coordinating research tasks to make the best use of the

respective strengths of the two sides' research capabilities, the

benefits realized from available research funds can be amplified.

* Exchange visiting researchers. This may be done in connection with

cooperative research programs, and independently of them, while will

further the aims of the cooperative research programs, technical

information exchanges, familiarization with methods of practice on

the respective sides, and mutual understanding and cooperative

relationships between workers on the two sides.

Accomplishments

* U.S. researchers presented 4 papers, and Japanese researchers

presented 10 papers, on soil behavior and stability during

earthquakes, at the 20th and 21st joint panel meetings 5/88,5/89

* Mr. Yoshitake Hachiya, of the Port and Harbor Research Institute,

completed a one-year visit as a guest researcher in Pavement

Systems at the U.S. Army Engineer Waterways Experiment Station. .3/88

* A T/C (H) Workshop on Remedial Treatment of Potentially

Liquefiable Soils was held at Jackson Lake, Wyoming 5/88

* Mr. Osamu Matsuo, of the PWRI, completed a one-year guest researcher

assignment, performing research in Dynamic Behavior of Offshore

Structures, at Southern California University 8/88

* Mr. Nario Yasuda of the PWRI is a guest researcher at the U.S.

Bureau of Reclamation, working on Seismic Behavior of

Soils 1/89

* Proceedings of the Workshop on the In-Situ Testing, held at San

Francisco, California in 1985 was published and disseminated to the

Panel members 4/89

* A planning meeting for cooperative research on preventive

measures against soil liquefaction was held prior to the 21st joint

meeting 5/89

Future Plans

""' Identification of research tasks on preventive measures against

soil liquefaction that are of common interest, in accordance with

the results of the planning meeting 9/89

* Develop plans for a cooperative research program to verify soil-pile

interaction models for marine structures subject to earthquakes,

387
measurement of sea floor earthquake motions, and the stability of

gravity based structures during seismic events 9/89

* Initiate research tasks on the development of in-situ assessment

methods for soil liquefaction, under the cooperative research

program . 10/89

" Hold the 1990 U.S. - Japan Workshop on Design and Treatment

Methods for Mitigating Liquefaction Induced Damage, in Japan. .. 10/90

5. Information Exchange

* T/C workshops, exchange of visitors, and participation in annual

joint panel meetings, including presentation of technical papers.

* Exchange of researchers between U.S. and Japanese government

laboratories.

* Visits by researchers and participation in technical activities

such as post-earthquake field investigations.

* With the professional community in the U.S. and Japan; by

publication of papers in journals; publication of research reports

by the respective agencies; participation in professional society

meetings and conferences; seminars; and workshops.

6 . Impact

Through the exchange of researchers and the diffusion of earthquakerelated technical data, experience, and information on methods of practice,

the state of technology in geotechnical earthquake engineering has been

raised on both sides. Through the cooperative research program on in-situ

testing of soils which was carried out after the 1983 Nihonkai-Chubu

Earthquake, improved and more accurate methods of using in-situ tests for

evaluating the liquefaction potential of soils were achieved; these have

been in general use in the United States since 1985.

7. Barriers

* Insufficient funds allocated to earthquake engineering research.

* Administrative restrictions on foreign travel.

388
Report of Task Committee on

(I) STORM SURGE AND TSUNAMI

Date: May 17, 1989

Place : Public Works Research Institute

Ministry of Construction

Tsukuba, Ibaraki 305

Attendees : Japan Side - T. Uda (Chairman) (PWRI)

K. Tanimoto (PHRI)

K. Kurashige (MRI)

M. Okada (MRI)

N. Oyagi (NRCDP)

U.S. Side - J. F. Lander (Acting Chairman) (Univ. of Colorado)

W. E. Roper (COE)

Following the request in item 8 of the Final Resolutions adopted at

the 20th Joint Meeting held in Gaithersburg in 1988, the "Mission

Statement" of this Task Committee was drafted for adoption by the Panel as

follows

;

MISSION STATEMENT

1 . Objective

Both storm surge and tsunami are hazards capable of inflecting damage of

disastrous proportions. Storm surges are associated with hurricanes and

typhoons where high winds and the mounding of water under the low

barometric pressure of the storm's eye can cause extensive flooding and

severe wave action. Tsunamis are predominately caused by underwater

earthquakes and, to a lesser extent volcanic activity. Depending on the

distance from the source tsunamis arrive within minutes to up to a day

after generation.

The main objective of this T/C is to mitigate these hazards through shared

technologies, research, information, and cooperative work.

2. Scope of Work

Exchange results of research on storm surge and tsunami occurrence,

generation, propagation, and coastal effects. This includes

observations on historical, current, and theoretical tsunamis. Of

particular interest is the effort by US and JAPAN to acquire deep

ocean tsunami measurements.

* Exchange results and status of anti-storm surge and tsunami;

activities including analysis of the problem, planning, warning,

and engineering approaches.

* Exchange information on important planned and ongoing projects

relating to storm surge and tsunamis.

389
* Exchange information on development of technologies such as

computer programs to predict travel times, run-up heights, and wave

characteristics and analysis; improved instrumentation, and use of

satellite communications for detection and warning.

* Facilitate dissemination through exchange of literature, technical

reports at joint meetings, special workshops, joint projects, and

direct interaction among participants.

3 . Accomplishments

* Exchanged digital bathymetric data with the Japanese side; US side

received a computer program to calculate tsunami travel time

charts

.

* Exchanged computer program forecasting storm surge run-up.

* Exchanged technical reports on related subjects.

4. Future Plans

* Develop exchanges of data and information on activities related to

deep ocean tsunami measurements.

* A Second Tsunami Workshop (proposed location, Hawaii in October or

November 1990) to address advances since the first Workshop and to

bring together computer modelers and potential users of modeling

results is planned.

* Explore and promote joint undertakings such as instrument testing,

joint publications, scientist exchange, satellite communications,

and assistance in contacts within the US and JAPAN.

* NOAA has adopted the travel time program developed by Mr. Okada

(MRI/JMA) using NOAA gridded bathymetric data to run on NOAA

computers. The first ever computed travel time charts have been

prepared for the Caribbean Sea using this program developed by Mr.

Okada. A copy of the published paper will be sent to the Japanese

side

.

* The activity area of the Task Committee will be expanded to include

members with strong interest on the engineering aspects of storm

surge and tsunamis in U.S. side.

* The U.S. side has compiled detailed information on all tsunamis

affecting the U.S. and possessions. It is being printed and will

be exchanged this summer.

5 . Information Exchange

* Continue exchange of publications, bibliographic data, technical

reports, and personal contacts.

6 Impact

390
* Increased cooperation among scientists' working in these fields.

* Hasten implementation of techniques developed from the US to Japan

and vice versa

* Facilitate wider dissemination of information and technology.

Barriers

* Insufficient funding for government and private sector

organizations to participate in T/C workshops and the Panel's

annual joint meetings.

* The tsunami problem is less serve in the US than in JAPAN which

leads to a smaller community of researchers. There are fewer US

researchers focusing on tsunamis at any or several locations; thus

highlighting the importance of T/C periodic meetings and workshops.

* There are different approaches and degree of significance in

dealing with problems in the US and JAPAN e.g., in US the hazard is

from a remote source which leads to emphasizing developments of

warning and evacuation schemes while in JAPAN the engineering

approach is favored due to more locally generated tsunamis.

391
Report of Task Committee on

(J) WIND AND EARTHQUAKE ENGINEERING FOR TRANSPORTATION SYSTEMS

Date : May 18, 1989

Place : Public Works Research Institute, Ministry of Construction

1-banchi Asahi, Tsukuba-shi, Ibaraki-ken 305, Japan

Attendees : Japan Side - S. Saeki (Chai rman) (PWRI)

T. Fujitani (MRI)

T. Iwasaki (PWRI)

Y. Sasaki (PWRI)

M. Fujiwara (PWRI)

K . Yokoyama (PWRI)

M . Okahara (PWRI)

K. Kawashima (PWRI)

K. Tokida (PWRI)

K, Minosaku(Observer) (PWRI)

U.S. Side - H. R . Bosch(Acting Chairman) (FHWA)

A. G. Franklin (WES)

Following the request in item 8 of the Final Resolutions adopted at

the 20th UJNR Joint Panel Meeting held in Gaithersburg in 1988, the

"Mission Statement" of this Task Committee was drafted for adoption by the

Pane I as foil ows

;

[Mission Statement]

1 . Object i ve

The objectives of this Task Committee are to plan, promote, and foster

research on the behavior of highway bridges when subjected to wind and

seismic forces and to disseminate research results and provide

specifications and guidelines based on the task committee's findings.

Surface transportation systems for movement of goods and people play a

392
vital part of commerce and intercou-rse between people. Highway bridges are

especially influenced by the forces of wind and earthquakes because of

their open exposure to those forces.

2. Scope-of-Work

The scope of the work is applicable mostly to highway bridges without any

limitation on their size and function: such as existing bridges and new

bridge designs; whole system of bridges; and/or to single components of a

bridge. The mission is performed through:

conducting workshops, exchanging researchers,

developing methods for design evaluations and test procedures,

inspection techniques, rehabilitation and maintenance specifications

and policies,

and performing cooperative research programs and other relevant

cooperative administrative activities.

3. Accompl ishmen ts

* Conducted coordinated research studies on the seismic performance of

bridge piers and columns. Details of this work involved determing the

performance of reinforced concrete piers and columns subjected to dynamic

cyclic loading, performing model tests on the failure of reinforced

concrete piers, testing full-scale concrete columns and for the behavior of

concrete filled steel tubes.

* Held the fourth bridge workshop in San Diego, U.S. in 1988 and the fifth

bridge workshop in Tsukuba, Japan in 1989, dealing with the determination

and evaluation of the performance and strengthening of bridge structures,

structural monitoring, repairs, safety and non-destructive evaluation.

Special attention was given to the seismic and wind loadings of modern cabl

-stayed bridges, vibration suppression, base isolation and dynamic control

techn i ques

.

* Both sides presented technical papers at the UJNR Joint Meeting and

participated in technical study tours such as factories, construction sites

and instrumented structures.

4 . Future Plans

* Conduct the 6th bridge workshop just prior to the 22nd Joint Panel

393
meeting in the U.S.

* Continue the coordinated experimental research study on the seismic

performance of bridge piers and columns, and continue emphasis on base

isolation.

* Develop a coordinated research study on seismic, aeroelastic, and

aerodynamic response of cable-supported bridges with emphasis on cable

inspection, vibration and corrosion protection.

* Develop a coordinated research study to compare the seismic design

criteria for bridges in JAPAN and the U.S.

* Develop a coordinated research study on seismic response control, system

identification techniques, and non-destructive evaluation of bridge

s tructures

.

5 . Information Exchange

* Exchanged technical reports, research program documentation, construction

logs, design plans, and assorted photographs and TV videos.

* Mr. S. L'njoh from PWRI worked in University of Southern California and Mr

Ray Zelinsky from Transportation Bureau of California State worked in PWRI

to engage in research studies relating to this Task Committee.

6 . I mpact

* Greater uniformity was achieved in wind engineering test procedures and

modeling, resulting in more efficient solutions to the aerodynamic bridge

problem

.

* The workshops and study tours in the two countries have facilitated

technology transfer, as an example, in the area of cable protection and

vibration suppression of cable-stayed bridges.

* kith respect to the earthquake protection of bridges, the free exchange

of literature, instrumentation technology, and earthquake response data has

led to the improvement of design and retrofitting guidelines and to

calibration and verification of specifications.

* A set of priorities for research needed from a more international

viewpoint were established.

394
7. Barriers

There are no major technical barriers to achieve the Task Committee mission.

However, financial barriers, especially funding needed to participate in

workshops and panel meetings and for conducting actual bridge tests and

large scale model tests, limit the effectiveness of the work in this Task

Committee' s work.

OU.S. GOVERNMENT PRINTING OFFICE: 1990-2 61-986:04107 395

NBS-114A (REV. 2-8C) U.S. DEPT. OF COMM.

BIBLIOGRAPHIC DATA

SHEET (See instructions)

1. PUBLICATION OR

REPORT NO.

NIST/SP-776

2. Performing Organ. Report No 3. Publication Date

January 1990

4. TITLE AND SUBTITLE

Wind and Seismic Effects —Proceedings of the 21st Joint Meeting of the U.S. -Japan

Cooperative Program in Natural Resources Panel on Wind and Seismic Effects

5. AUTHOR(S)

Noel J. Raufaste, Editor

6. PERFORMING ORGANIZATION (If joint or other than NBS. see instructions)

NATIONAL INSTITUTE OF STANDARDS AND TECHNOLOGY

(formerly NATIONAL BUREAU OF STANDARDS)

US. DEPARTMENT OF COMMERCE

QMTHERSBURG, MD 20899

7. Contract/Grant No.

8. Type of Report & Period Covered

Final

9. SPONSORING ORGANIZATION NAME AND COMPLETE ADDRESS (Street. City. State. ZIP)

SAME AS ITEM #6 ABOVE,

10. SUPPLEMENTARY NOTES

| | Document describes a computer program; SF-185, FIPS Software Summary, is attached.

11. ABSTRACT (A 200-word or less factual summary of most significant information. If document includes a significant

bibliography or literature survey, mention it here)

The 21st Joint Meeting of the U.S. -Japan Panel on Wind and Seismic Effects was held

at the Public Works Research Institute, Tsukuba, JAPAN, from May 16-19, 1989. This

publication, the proceedings of the Joint Meeting, includes the program, list of members

panel resolutions, task committee reports, and 33 technical papers.

The papers were presented under six themes: (I) - Wind Engineering, (II) - Earthquake

Engineering, (III) - Storm Surge and Tsunami;,. (IV) - U.S. -Japan Cooperative Research

Program, (V) - Code Development Process and Code Enforcement Responsibility, and

(VI) - Armenia Earthquake (oral presentation).

12. KEY WORDS (Six to twelve entries; alphabetical order; capitalize only proper names; and separate key words by semicolons)

accelerograph; Armenia; bridges; codes; concrete; design criteria; disaster; earthquakes

geotechnical engineering; ground failures; inelastic; lifelines; liquefaction; masonry;

repair and retrofit; risk assessment; seismicity; soils; standards; storm surge;

13. availability structural engineering; tsunami; and wind loads

Unlimited

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15. Price

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