NAT'L INST. OF STAND & TECH R.I.C.

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NBS SPECIAL PUBLICATION 658

U.S'. DEPARTMENT OF COMMERCE/National Bureau of Standards

Wind and Seismic

Effects

Proceedings of the

Eleventh Joint UJNR

Panel Conference
NATIONAL BUREAU OF STANDARDS

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HATIOHAL BUPEAO

Wind and Seismic Effects

^ "fT/l^ AUG 2 5 1983

Proceedings of the 11th Joint '^^'^()^%

Panel Conference of the U.S. -Japan

'^^^^a ^l) Cooperative Program in ^ ' c^:^

Natural Resources

H.S. Lew, Editor

Center for Building Technology

National Engineering Laboratory

National Bureau of Standards

Washington, DC 20234

U.S. DEPARTMENT OF COMMERCE, Malcolm Baldrige, Secretary

NATIONAL BUREAU OF STANDARDS, Ernest Ambler, Director

Issued July 1983
Library of Congress Catalog Card Number: 83-600550

National Bureau of Standards Special Publication 658

Natl. Bur. Stand. (U.S.), Spec. Publ. 658, 755 pages (July 1983)

CODEN: XNBSAV

U.S. GOVERNMENT PRINTING OFFICE

WASHINGTON; 1983

For sale by the Superintendent of Documents, U.S. Government Printing Office, Washington, DC 20402

Price $12.00

(Add 25 percent for other than U.S. mailing)
PREFACE

The Eleventh Joint Meeting of the U.S. - Japan Panel on Wind and

Seismic Effects was held in Tokyo, Japan on September 4-7, 1979. This

panel is a part of the U.S. - Japan Cooperative Program in Natural

Resources (UJNR). The UJNR was established in 1964 by the U.S. - Japan

Cabinet-level Committee on Trade and Economic Affairs. The purpose of

the UJNR is to exchange scientific and technological information which

will be mutually beneficial to the economics and v/elfare of both

countries.

These proceedings include the program, the formal resolutions, and

the technical papers presented at the Joint Meeting. The texts of the

papers, all of which were prepared in English, have been edited for

obvious errors and clarity.

It should be noted that throughout the proceedings certain commercial equipment, instruments or materials are identified in order to

specify adequately experimental procedure. In no case does such identification imply recommendation or endorsement by the National Bureau of

Standards, nor does it imply that the material or equipment identified

is necessarily the best for the purpose.

Preparation of the proceedings was partially supported by funds

from the Department of Housing and Urban Development

"' H. S. Lew, Secretary

U.S. Panel on Wind and

;» .. • Seismic Effects

i i i
SI CONVERSION UNITS

In view of the present accepted practice for wind and seismic technology, common units

of measurements were used throughout this publication. In recognition of the position of the

United States as a signatory to the General Conference on Weights and Measures, which gave

official status to the International System of Units (SI) in 1960, the table below is presented to facilitate conversion to SI Units. Readers interested in making further use of the

coherent system of SI units are referred to: NBS SP 330, 1977 Edition, The International

System of Units; and ASTM Standard for Metric Practice.

Table of Conversion Factors to SI Units

Customary Units

International

(SI) UNIT

Conversion

Approxi mate

1 P T\nf h i n r h ^ i n ^ 1 1 IC C 1 1111/ 1X inIII = 0u • n?S4 yjc -j^ m*III

foot fftl mpt P r f m

I ft1 U = 0\J m 0\J^KJ 304R m*Ml

Force npi,\/1" on r M ^ 11 \ U \ — H- • tO in

kilogram (kgf) newton (N) 1 kgf = 9.807 N

Pressure pound per square

Stress inch (psi)

kin npr *^niMrp

newt on/meter^ 1 psi = 6895 N/m2

inch (ksi) newton/meter2 1 ksi = 6895 x 103 N/m2

Energy inch* pound (in-lbf) joule (J) 1 in-lbf = 0.1130 J

foot'pound (ft-lbf) joule (J) 1 ft«lbf = 1 .3558 J

Torque

or Bending

Moment

pound 'i nch newton-meter (N»m) 1 Ibf.in = 0.1130 N.m

pound'foot (Ibf-ft) newton-meter (N-m) 1 Ibf.ft = 1.3558 N-m

Weight pound (lb) kilogram (kg) 1 lb = 0.4536 kg

Unit Weight pound per cubic foot

(pcf)

kil ogram per cubic

meter (kg/m^)

1 pcf = 16.018 kg/m3

Vel ocity foot per second

(ft/sec)

meter per second

(m/s)

1 ft/s = 0.3048 m/s

Acceleration foot per second

second (ft/sec^)

meter per second

per second (m/s^)

1 ft/sec2 = 0.3048 m/s^

3 Meter may be subdivided. A centimeter (cm) is 1/100 m and a millimeter (mm) is 1/1000 m.

* Exactly

i V
ABSTRACT

The Eleventh Joint Meeting of the U.S. - Japan Panel on Wind and

Seismic Effects was held in Tsukuba, Japan on September 4-7, 1979. The

proceedings of the Joint Meeting include the program, the formal resolution

and the technical papers. The subjects covered in the paper include (1)

the engineering characteristics of wind, (2) the characteristics of earthquake ground motions, (3) the earthquake response of structures, (4) the

wind response of structures, (5) recent design criteria against wind and

earthquake disturbances, (6) the design and analysis of special structures,

(7) the evaluation, repairing, and retrofitting for wind and earthquake

disaster, (8) earthquake disaster prevention planning, (9) storm surge and

tsunamis, (10) and technical cooperation with developing countries.

KEYWORDS: Accelerograph ; codes; design criteria; disaster; earthquakes;

hazards; ground failures; seismicity; solids; standards;

structural engineering; structural responses; tsunamis;

wind loads; and winds.

V
CONTENTS

Pag e

Preface iii

SI Conversion iv Abstract v Program of the Joint Meeting x List of Members , xv Formal Resolutions xx

Themes and Technical Papers

Opening Themes: ;

New Initiatives in Earthquake Hazards Mitigation ... 1 Charles C. Thiel

Earthquake Hazard Reduction Research at the National Bureau of Standards .... 15 Edgar V. Leyendecker, James R. Harris, Richard N. Wright, Edward 0. Pfrang

Outline of Research on Wind and Earthquake Effects in Japan ?2 Yoshijiro Sakagami

On the U.S. - Japan Cooperative Program of Large Scale Testing 23 Tadayoshi Okubo

Theme I: ENGINEERING CHARACTERISTICS OF WIND

Modeling the 1978 Tokyo Tornado that Overturned the Tozai Subway Train I-l Fiji Uchida, Ryozo Tatehira, Ichiro Tabe, Kazuyuki Ohtsuka

Extreme Wind Speeds at 129 Stations in the Contiguous United States 1-16

Michael J. Changery, Emil Simiu, James J. Filliben

Analysis of High Wind Observations from Very Tall Towers 1-31

Keikichi Naito, Isao Tabata, Noboru Banno, Katsumi Takahashi

Theme II: CHARACTERISTICS OF EARTHQUAKE GROUND MOTIONS (Including Engineering

Sei smol ogy)

A Proposal for a New Parameter in Assessing Seismic Disaster II-l

Michio Otsuka

Simulation of Earthquake Ground Motion and its Application to Dynamic Response

Analysis 11-13

Tetsuo Kubo, Makoto Watabe

Comparison of Vertical Components of Strong-Motion Accelerograms for Western

United States and Japan . 11-24

Tatsuo Uwabe

Earthquake Observation Systems in and around Structures in Japan 11-45

Kiyoshi Nakano, Yoshikazu Kitagawa

Theme III: EARTHQUAKE RESPONSE OF STRUCTURES (Including Hydraulic Structures and

Earth Structures)

Single-story Residential Masonry Construction in Uniform Building Code Seismic

Zone 2 III-l

G. Robert Fuller

vi
Page

The Performance of Lapped Splices in Reinforced Concrete Under High-level

Repeated Loading ..... III-23

Peter Gergely, Fernando Fagundo, Richard N. White

Dynamic Behaviors of Reinforced Concrete Frame Structures II 1-29

Keiichi Ohtani, Chikahiro Minowa

Seismic Response of Precast Concrete Walls III-41

James M. Becker, Carlos Llorente, Peter Mueller

Seismic Response Analysis of the Itajima Bridge v/ith Use of Strong Motion

Acceleration Records III-65

Masamitsu Ohashi , Toshio Iwasaki, Kazuhiko Kawashima

Field and Laboratory Determination of Soil Moduli III-87

William F. Marcuson, III, Joseph R. Curro, Jr.

An Experimental Study on Liquefaction of Sandy Soils on a Cohesive Soil

Layer III-119

Tatsuo Asama, Yukitake Shioi

Stress-strain Behavior of Dry Sand and Normally Consolidated Clay by Inter- laboratory Cooperative Cyclic Shear Tests III-133

Hiroshi Oh-oka, Kohjiroh Itoh, Yoshihiro Sugimura, Masaya Hirosawa

THEME IV: WIND RESPONSE OF STRUCTURES

Wind and Structure Motion Study for Pasco-Kennewick Bridge IV-1

M. C. C. Bampton, Harold Bosch, David H. Cheng, Charles H. Scheffey

Calculation of the Gust Responses of Long-Span Bridges IV-6

Nobuyuki Narita, Hiroshi Sato

THEME V: RECENT DESIGN CRITERIA AGAINST WIND AND EARTHQUAKE DISTURBANCES

Draft Seismic Design Guidelines for Highway Bridges ..... V-1

James D. Cooper, Charles F. Scheffey, Roland L. Sharpe, Ronald L. Mayes

THEME VI: DESIGN AND ANALYSIS OF SPECIAL STRUCTURES (Including Highrise Buildings,

Dams, and Nuclear Facilities)

Analysis and Design of Cracked Reinforced Concrete Nuclear Containment Shells

for Earthquakes VI-1

Peter Gergely, Richard N. White

Comparison of the Measured and Computed Responses of the Yuda Dam during the

July 6, 1976 and June 12, 1978 Earthquakes VI-9

Ryuichi lida, Norihisa Matsumoto, Satoru Kondo

THEME VII: EVALUATION, REPAIRING, RETROFIFTING FOR WIND AND EARTHQUAKE DISASTER

Report on the 1978 Miyagi -ken-oki Earthquake VII-1

Makoto Watabe, Yutaka Matsushima, Yuji Ishiyama, Tetsuo Kubo, Yuji Ohashi

Disastrous Ground Failures in a Residential Area over a Large-Scale Cut-and-Fill

in the Sendai Region Caused by the Earthquake of 1978 VII-22

Hitoshi Haruyama, Motoo Kobayashi

vi i
h'age

Damage Features of Civil Engineering Structures due to the Miyagi-ken-oki

Earthquake of 1978 VII-36

Tadayoshi Okubo, Masamitsu Ohashi, Toshio Iwasaki, Kazuguki Kawashima,

Ken-ichi Tokida

Damage to River Dykes Caused by the Miyagi-ken-oki Earthquake of June 1978 . . VII-62

Kazuya Yamamura, Yasushi Sasaki, Yasuyuki Koga, Eiichi Taniguchi

Damage to Port Structures by the 1978 Miyagi -ken-ohi Earthquake VII-79

Hajime Tsuchida, Setsuo Moda

Wood Diaphragms in flasonry Buildings VII-110

Mihran S. Agbabian

Mitigation of Seismic Hazards in Existing Unreinforced Masonry Buildings . . . VII-130

Mihran S. Agbabian

Development of a Universal Fastener for Wooden Building Roof Frames VII-143

Tatsuo Murota, Yuji Ishiyama

THEME VIII: EARTHQUAKE DISASTER PREVENTION PLANNING

The NASA/MSFC Experimental Facilities at Huntsville, Alabama VIII-1

John B. Scalzi, George F. McDonough, Jr.

Earthquake Hazards Reduction Research Supported in 1978 VIII-23

Charles C. Thiel, William A. Anderson, Michael P. Gaus, William Hakala,

Frederick Krimgold, Shih Chi Liu, John B. Scalzi

Functional Damage and Rehabilitation of Lifelines in the Miyagi-ken-oki

Earthquake of 1978 VIII-28

Kazuto Nakazawa, Eiichi Kuribayashi, Tadayuki Tazaki, Takayuki Hadate,

Ryoji Hagiwara

Social Aspects of Earthquake Mitigation and Planning in the United States . . VIII-113

Will iam A. Anderson

THEME IX: STORM SURGE AND TSUNAMIS

Orientation of Tsunami's Researches in Japan IX-1 Hiroshi Takahashi , Yukio Fujinawa

Wave Setup Caused by Typhoon 7010 IX-8

Hiroshi Hashimoto, Taka-aki Uda

Specification and Prediction of Surface Wind Forcing for Ocean and Current and

Storm Surge Models IX-19

Celso S. Barrientos, Kurt W. Hess

A Dynamical Model to Predict Storm Surges and Overland Flooding in Bays and

Estuaries IX-25

Chester P. Jel esnianski , Celso S. Barrientos, Jye Chen

Topics on Tsunami Protection along the Port Areas in Japan IX-29

Yoshimi Goda

yiii
Page

THEME X: TECHNICAL COOPERATION WITH DEVELOPING COUNTRIES

Seismic Risk Maps in the Southeast Asian Countries (Maximum Acceleration and

Maximum Partile Velocity) - Phill ippines, Indonesia and Indo-China X-1

Sadaiku Hattori

California School and Hospital Ceilings X-17

John F. Meehan

I:.

TX
PROGRAM OF THE ELEVENTH JOINT MEETING

OF THE

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS .

' SEPTEMBER 4-7, 1979

AT THE ' PUBLIC WORKS RESEARCH INSTITUTE

TSUKUBA SCIENCE CITY, JAPAN

TUESDAY - September 4 OPENING SESSION

11:00 a.m. Call to order by Dr. Ryuichi lida. Secretary, Japanese Panel, Director,

Planning and Research Administration Department, Public Works Research

Institute

Remarks by Mr. Shin-ichiro Asai, Engineer General, Ministry of Construction

Remarks by Mr. Justin L. Bloom, Counselor for Scientific and Technological

Affairs, Embassy of the United States of America

Remarks by Mr. Yoshijiro Sakagami, Chairman, Japanese Panel, Director-General,

Public Works Research Institute, Ministry of Construction

Remarks by Dr. E. 0. Pfrang, Chairman, U.S. Panel, Chief, Structures and

Materials Division, Center for Building Technology, National Engineering

Laboratory, National Bureau of Standards

Introduction of Japan Panel Members by Japanese Chairman and U.S. Panel

Members by U.S. Chairman

Election of Conference Chairman

Adoption of Agenda

1:20 p.m. New Initatives in Earthquake Hazards Mitigation — Charles C. Thiel,

E. 0. Pfrang

1:35 Earthquake Hazard Reduction Research at the National Bureau of Standards —

Edward 0. Pfrang, Edgar V. Leyendecker, James R. Harris, Richard N. Wright

1:50 Outline of Research on Wind and Earthquake Effects in Japan — Yoshijiro

Sakagami

2:05 Discussion

2:40 Modeling the 1978 Tokyo Tornado that Overturned in the Tozai Subway Train —

Eiji Uchida, Ryozo Tatehira, Ichiro Tabe, Kazuyuki Ohtsuka, Shigemi Fujiwhara

3:00 Extreme Wind Speeds at 129 Stations in the Contiguous United States —

Michael J. Changery, Emil Simiu, James J. Filliben, Celso Barrientos

3:20 Analysis of High Wind Observations from Very Tall Towers — Keikichi Naito,

Isao Tabata, Noboru Banno, Katsumi Takahashi

3:40 Discussion

4:10 Wind and Structure Motion Study for Pasco-Kennewick Bridge — M. C. C. Bampton,

Harold Bosch, David H. Cheng, Charles F. Scheffey
Calculation of the Gust Responses of Long-Span Bridges — Nobuyuki Narita,

Hiroshi Sato

Discussion ' • ' ' • ' :

• :

September 5 .. -feC ' .

> .p- :

A Proposal for a New Parameter in Assessing Seismic Disaster — Michio Otsuka

Strong-Motion Data Exchange Through National Data Bases — A. Gerald Brady,

Discussion

Simulation of Earthquake Ground Motion and Its Application to Dynamic Response

Analysis — Tetsuo Kubo, Makoto Watabe

Comparison of Vertical Component of Strong-Motion Accelerograms for Western

United States and Japan — Tatsuo Uwabe

Earthquake Observation System in and around Structure in Japan — Kiyoshi

Nakano, Yoshikazu Kitagawa

Discussion

Single-story Residential Masonry Construction in Uniform Building Code Seismic

Zone 2 — G. Robert Fuller

The Performance of Lapped Splices in Reinforced Concrete Under High-Level

Repeated Loading — Peter Gergely, Fernando Fagundo, Richard N. White

Discussion

Dynamic Behavior of Reinforced Concrete Frame Structures — Keiichi Ohtani,

Chikahiro Minowa

Seismic Response of Precast Concrete Walls — James M. Becker, Carlos Llorente,

Peter Mueller

California School and Hospital Ceilings — John F. Meehan

Seismic Response Analysis of the Itajima Bridge with Use of Strong Motion

Acceleration Records — Masamitsu Ohashi, Toshio Iwasaki, Kazuhiko Kawashima

Discussion ]:

Field and Laboratory Determination of Soil Moduli — William F. Marcuson III,

Joseph R. Curro, Jr., G. Robert Fuller ,., . ,• .

An Experimental Study on Liquefaction of Sandy Soils on a Cohesive Soil

Layer — Tatsuo Asama, Yukitake Shioi, Hideya Asanuma

Stress-strain Behavior of Dry Sand and Normally Consolidated Clay by Interlaboratory Cooperative Cyclic Shear Tests — Hiroshi Oh-oka, Kohjiroh Itoh,

Yoshihiro Sugimura, Masaya Hirosawa

Discussion •

Discussion of Task Committee Reports (* Chairmen of Task Committees)

a) Strong-Motion Arrays and Data — *Jerry S. Dodd, *Hajime Tsuchida,

Toshio Iwasaki, Keiichi Ohtani, Michio Ohtsuka, Makoto Watabe

xi
b) Large-scale Testing Programs — *H. So Lew, *Keiichi Ohtani, Tatsuor

Murota, Nobuyuki Narita, Setsuo Noda, Yashushi Sasaki, Makota Watabe

c) Repair and Retrofit of Existing Structures — *John B. Scalzi, *Kiyoshi

Nakano, Tatsuo Asama, Eiichi Kuribayashi

d) Structural Performance Evaluation — *G. Robert Fuller, *Makoto Watabe,

Tatsuo Asama, Yukitake Shioi

f) Disaster Prevention Methods for Lifeline Systems — *Charles F. Scheffey,

*Tadayoshi Okubo, Shigemi Fujiwhara, Toshio Iwasaki, Eiichi Kuribayashi,

Tatsuro Murota, Setsuo Noda

g) High Wind — *Celso Barrientos, *Shigemi Fujiwhara, Tatsuro Murota,

Keikichi Nalto, Nobuyuki Narita

THURSDAY - September 6

9:00 a-m. Draft Seismic Design Guidelines for Highway Bridges — James D. Cooper,

Charles F. Scheffey, Roland L. Sharpe, Ronald L. Mayer

9:20 Discussion

9:30 Analysis and Design of Cracked Reinforced Concrete Nuclear Containment Shells

for Earthquakes — Peter Gergely, Richard N. White

9:50 Comparison of Measured and Computed Response of Yuda Dam during the July 6,

1976 and June 12, 1978 Earthquakes — Ryuichi lida, Norihisa Matsumoto,

Satoru Kondo

10:10 Discussion

10:35 Report on the 1978 Miyagi-ken-oki Earthquake — Makoto Watabe, Yutaka

Matsushima, Yuji Ishiyama, Tetsuo Kubo, Yuji Ohashi

10:50 Disastrous Ground Failure in a Residential Area of Large-Scale Cut-and-Fill in

the Sendai Region caused by the Earthquake of 1978 — Hitoshi Haruyama, Motoo

Kobayashi

11:05 Damage Features of Civil Engineering Structures due to the Miyagi-ken-oki

Earthquake of 1978 — Tadayoshi Okubo, Masamitsu Ohashi, Toshio Iwasaki,

Kazuhiko Kawashima, Ken-ichi Tokida

11:20 Damage to River Dykes caused by the Miyagi-ken-oki Earthquake of June, 1978 —

Kazuya Yamamura, Yasushi Sasaki, Yasuyuki Koga, Eiichi Taniguchi

11:35 Damage to Port Structures by the 1978 Miyagi-ken-oki Earthquake — Hajime

Tsuchida, Setsuo Noda

11:50 Discussion

1:15 p.m. Wood Diaphragms Masonry Buildings — Mihran S. Aghabian

1:30 Mitigation of Seismic Hazards in Existing Unreinforced Masonry Buildings —

M. S. Agbabian

1:45 Development of a Universal Fastener for Wooden Buildings Roof Frames — Tatsuo

Murota, Yuji Ishiyama

2:05 Discussion

2:20 The NASA/MSFC Experimental Facilities at Huntsville, Alabama — John B. Scalzi,

George F. McDonough, Jr., Nicholas C. Costes

xii
2:40 Earthquake Hazards Reduction Research Supported in 1978 — Charles C. Thiel,

William A. Anderson, Michael P. Gaus, William Hakala, Frederick Krimgold,

Shih Chi Liu, John B. Scalzi

3:00 Functional Damage and Rehabilitation of Lifelines in the Miyagi-ken-oki

Earthquake of 1978 — Kazuto Nakazawa, Eiichi Kuribayashi, Tadayuki Tazaki,

Takayuki Hadate, Ryoji Hagiwara

3:20 Discussion

3:50 Social Aspects of Earthquake Mitigation and Planning in the United States —

William A. Anderson

4:10 Discussion

4:20 Seismic Risk Maps in the Southeast Asian Countries (Maximum Acceleration and

Maximum Particle Velocity) - Phillippines , Indonesia and Indo-China — Sadaiku

Hattori

4:50 Discussion of Task Committee Reports (* Chairmen of Task Commitees)

e) Land Use Programs for Controlling Natural Hazard Effects — *G. Robert

Fuller, *Eiichi Kuribayashi, Hitoshi Haruyaraa, Sadaiku Hattori

h) Soil Behavior and Stability during Earthquakes — *Jerry S. Dodd, *Toshio

Iwasaki, Yasushi Sasaki, Yukitake Shioi, Hajime Tsuchida

5:30 Adjourn

FRIDAY - September 7

9:00 a.m. Orientation of Tsunami Research in Japan — Hiroshi Takahashi, Yukio

Fu jinawa

9:20 Wave Setup Caused by Typhoon 7010 — Hiroshi Hashimoto, Taka-aki Uda

9:40 Specification and Prediction of Surface Wind Forcing for Ocean Current and

Storm Surge Models — Celso S. Barrientos, Kurt W. Hess

9:55 A Dynamic Model to Predict Storm Surges and Overland Flooding in Bays and

Estuaries — Chester P. Jelesnianski , Celso S. Barrientos, Jye Chen

10:10 Topics on Tsunami Protection along the Port Areas in Japan — Yoshimi Goda

10:30 Discussion

11:05 Task Committee Reports

11:05 a) Strong-Motion Instrumentation Arrays and Data

11:15 b) Large-Scale Testing Programs

11:25 c) Repair and Retrofit of Existing Structures

11:35 d) Structural Performance Evaluation

11:45 e) Land Use Programs for Controlling Natural Hazard Effects

12:55 p.m. Task Committee Reports

12:55 f) Disaster Prevention Methods for Lifeline Systems

xiii
1:05 g) High Wind

1:15 h) Soil Behavior and Stability during Earthquakes

1:25 Discussion

1:35 Adoption of Final Resolution

2:45 Adjourn

xlv
U.S. PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

JANUARY 1980

Dr. Edward 0. Pfrang (CHAIRMAN)

Chief, Structures and Materials Division

Center for Building Technology, NEL National Bureau of Standards ,

• Washington, DC 20234 I

Dr. H. S. Lew (SECRETARIAT)

Structures and Materials Division

Center for Building Technology, NEL .

' National Bureau of Standards - : / Washington, DC 20234

(301) 921-2647 , .. -

'- </': Dr. S. T. Algermissen : ; Office of Earthquake Studies i Denver Federal Center

Branch of Earthquake Tectonics, USGS i

.

:

Stop 978, Box 25046 . Denver, CO 80225 . . : (303) 234-4014 r.

t..

Dr. Celso Barrientos ; ; Research Meteorologist

NOAA r -i^r- . ii: 8060 13th Street , Silver Spring, MD 20910 - . ,-, (301) 427-7613 . , ;. ,. ; , ^.= ^,

Mr. Billy Bohannan . •

, {i;(v

Assistant Director

Wood Engineering Research J."'

' Forest Products Laboratory, USDA

" P.O. Box 5130 , •!

^ Madison, WI 53705 ' . • / (608) 257-2211 u -

Dr. Roger D. Borcherdt . i . Chief, Branch of Ground Motion

and Faulting . ,. Office of Earthquake Studies, USGS

345 Middlefield Road

Menlo Park, CA 94025

(415) 323-8111 ;

, .,

Dr. A. Gerald Brady

Physical Scientist

Office of Earthquake Studies, USGS

345 Middlefield Road

Menlo Park, CA 94025

(415) 323-8111

Mr. James D. Cooper

Structures and Applied Mechanics Division

Federal Highway Administration

Office of Research HRS-11

Washington, DC 20590

(202) 557-4315

Mr. Jerry Dodd

Bureau of Reclamation

P.O. Box 25007

Denver Federal Center

Denver, CO 80225

(303) 234-3089

Mr. G. Robert Fuller

Architectural and Engineering Division

Department of Housing and Urban Development

Washington, DC 20411

(202) 755-5924

Dr. Michael P. Gaus

Division of Problem-Focused Research

Applications

National Science Foundation

1800 G Street, NW Washington, DC 20550

(202) 632-5700

Mr. John J. Healy

Chief, Research and Development Office

Department of the Army

DAEN-RDM

Washington, DC 20314 , (202) 693-7287 , , . ..,

Dr. William B. Joyner

Office of Earthquake Studies

Branch of Ground Motion and Faulting, USGS

345 Middlefield Road , - . .- Menlo Park, CA 94025

(415) 323-8111

Mr. John W. Kaufman

Atmospheric Sciences Division

National Aeronautics and Space Administration

Marshall Space Flight Center, AL 35812

(205) 453-3104

Mr. James Lander • National Oceanic and Atmospheric

Administration

Environmental Data Service

National Geophysical and Solar-Terrestrial

Data Center

Boulder, CO 80302

FTS 323-6474 or (303) 449-1000, ext. 6474

XV
Dr. E. V. Leyendecker

Leader, Earthquake Hazards Reduction Group

Structures and Materials Division

Center for Building Technology, NEL National Bureau of Standards

Washington, DC 20234

(301) 921-3471

Dr. Shih C. Liu

Program Manager

Division of Problem-Focused Research

Applications

National Science Foundation

Washington, DC 20550

Dr. Richard D. McConnell

Office of Construction

Veterans Administration

811 Vermont Avenue, NW Washington, DC 20420

(202) 389-3103

Dr. William F. Marcuson, III Research Civil Engineer, WES-SH

Department of the Army

Waterways Experiment Station,

Corps of Engineers

P.O. Box 631

Vicksburg, MS 39180

FTS 542-2202 or (601) 636-3111, ext. 2202

Dr. Richard D. Marshall

Leader, Structural Engineering

Structures and Materials Division

Center for Building Technology, NEL National Bureau of Standards

Washington, DC 20234

(301) 921-3471

Mr. John F. Meehan

Department of General Services

Office of Architectural Construction

Sacramento, CA 95805

(916) 445-8730

Mr. Howard L. Metcalf

Deputy Director for Construction Standards

and Design

Office of the Assistant Secretary of Defense

(MRA&L) Room 3E763

The Pentagon

Washington, DC 20301

(202) 695-2713

Dr. Thomas D. Potter

Deputy Director

Environmental Data Service, NOAA

3300 Whitehaven Street, NW Washington, DC 20235

(202) 634-7319

Dr. John B. Scalzi

Program Manager, Division of Problem-Focused

Research Applications

National Science Foundation

1800 G Street, NW

Washington, DC 20550

(202) 632-0648

Mr. Charles Scheffey . -

• Director, Office of Research '

; - Federal Highway Administration • > Department of Transportation

Washington, DC 20590

(202) 557-5254 : , . -

Mr. Lawrence C. Shao

Assistant Director for General Reactor

Safety Research i; Division of Reactor Safety Research

Nuclear Regulatory Commission

Washington, DC 20555 ^• (301) 427-4442

Dr. Warren A. Shaw

Head, Civil Engineering Department • Civil Engineering Laboratory

Naval Construction Battalion Center

Port Hueneme, CA 93043

(805) 982-5407

Dr. Charles C. Thiel

Assistant Associate Director for Mitigation

and Research

Federal Emergency Management Agency

Washington, DC 20472

(202) 653-7860

ALTERNATES : > -

Mr. Michael Changery .

' National Climatic Center • - Federal Building

Asheville, NC 28787

(704) 258-2850, ext. 217 i f

Dr. Jerry Harbour i> Site Safety Research Branch

Office of Site Safety Research Branch

Nuclear Regulatory Commission

Washington, DC 20555 •

' (301) 427-4370

J Dr. R. B. Matthiesen

Research Civil Engineer

Branch of Seismic Engineering, USGS

345 Middlefield Road

Menlo Park, CA 94025

FTS 467-2881 or (415) 323-8111

xvi
Mr. Ronald J. Morony

Program Manager

Building Technology Research Staff

Depai tment of Housing and Urban Development

Washington, DC 20410

(202) 755-0640

Mr. Drew A. Tiedemann

Bureau of Reclamation

Engineering and Research Center

Denver Federal Center

Denver, CO 80225

(303) 234-3029

xvii
JAPAN PANEL ON WIND AND SEISMIC EFFECTS

MEMBERSHIP LIST

1980

Mr. Yoshijiro Sakagami , CHAIRMAN

Director General

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi , Tsukuba-gun

Ibaraki-ken 305

Mr. Tatsuo Asama

Director

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Dr. Nozomu Den

Head, Seismology and Volcanology Division

Meteorological Research Institute (MRI)

Japan Meteorological Agency (JMA)

1-1, Nagamine, Yatabe-machi , Tsukuba-gun

Ibaraki-ken 305

Mr. Shigemi Fujiwhara

Head, Typhoon Research Division

Meteorological Research Institute (MRI)

Japan Meteorological Agency (JMA)

1-1, Nagamine, Yatabe-machi, Tsukuba-gun

Ibaraki-ken 305

Dr. Yoshimi Coda

Director, Marine Hydrodynamics Division

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi

,

Kanagawa-ken 239

Mr. Hiroshi Hashimoto

Head, Coastal Engineering Division

River Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Dr. Sadaiku Hattori

Chief, Seismology Section

International Institute of Seismology and

Earthquake Engineering (I.I.S.E.E.)

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305

Mr. Masaya Hirosawa

Chief Researcher

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305

Mr. Toshio Iwasaki

Head, Ground Vibration Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Mr. Motoo Kobayashi

Head, The First Geographic Division

Geographical Survey Institute

Ministry of Construction

Kitazato 1-banchi, Yatabe-machi, Tsukuba-gun

Ibaraki-ken 305

Mr. Eiichi Kuribayashi

Director

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Mr. Tatsuro Murota

Chief, Building Aerodynamics Section

Structure Engineering Division

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305

Dr. Keikichi Naito

Head, Meteorological Satellite Division

Meteorological Research Institute (MRI)

Japan Meteorological Agency (JMA)

1-1, Nagamine, Yatabe-machi, Tsukuba-gun

Ibaraki-ken 305

Dr. Kiyoshi Nakano

Director

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305

xvlii
Dr. Nobuyuki Narita

Head, Structure Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi , Tsukuba-gun

Ibaraki-ken 305

Mr. Setsuo Noda

Chief, Earthquake Disaster Prevention

Laboratory : , Structure Division

Port and Harbour Research Institute

Ministry of Transport

3-1-3, Nagase, Yokosuka-shi

,

Kanagawa-ken 239

Dr. Tadayoshi Okubo

Deputy Director-General " ' '

'.' Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Mr. Keiichi Ohtani

Chief, Earthquake Engineering Laboratory

Second Research Division

National Research Center for Disaster

Prevention

Science and Technology Agency

3-1, Tennodai, Sakura-mura, Niihari-gun,

Ibaraki-ken 305

Mr. Yasushi Sasaki

Head, Soil Dynamics Division

Construction Method and Equipment Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Mr. Yukitake Shioi

Head, Foundation Engineering Division

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Mr. Hiroshi Takahashi

Head, Second Research Division

National Research Center for Disaster

Prevention

Science and Technology Agency

3-1, Tennodai, Sakura-mura, Niihari-gun

Ibaraki-ken 305

Mr. Hajime Tsuchida

Chief, Earthquake Resistant Structures

Laboratory

Port and Harbour Research Institute

Ministry of Transport

3-1-1, Nagase, Yokosuka-shi,

Kanagawa-ken 239

Mr. Osamu Ueda

Head, Earthquake Engineering Division

Earthquake Disaster Prevention Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

Dr. Makoto Watabe

Head, Structure Division

Building Research Institute

Ministry of Construction

Tatehara 1-banchi, Oh-ho-machi, Tsukuba-gun

Ibaraki-ken 305

SECRETARIAT :

Dr. Ryuichi lida, Secretary-General

Director

Planning and Research Administration

Department

Public Works Research Institute

Ministry of Construction

Asahi 1-banchi, Toyosato-machi, Tsukuba-gun

Ibaraki-ken 305

xix
RESOLUTIONS ON THE ELEVENTH JOINT MEETING

U.S. -JAPAN PANEL ON WIND AND SEISMIC EFFECTS

U. J.N.R -

,

September 4-7, 1979 '

. ^

The following resolutions for future activities of this Panel are hereby adopted:

1. The Eleventh Joint Meeting was an extremely valuable exchange of technical information

which was beneficial to both countries. In view of the importance of cooperative

programs on the subject of wind and seismic effects, the continuation of Joint Panel

Meetings is considered essential.

2. The exchange of technical information, especially revised codes and specifications

relevant to wind and seismic effects, and the promotion of research programs including

exchange of personnel and available equipment should be strengthened. The secretariats

of both sides shall be kept informed of such exchanges of information by individual

members

.

3. The Panel concurs in the provisions described in the "Implementing Arrangement" signed

by the Science and Technology Agency and the Ministry of Construction for Japan and the

National Science Foundation for the U.S. on August 10th, 1979. To insure successful

implementation of the Joint Research Program of Large-scale Testing of Structures, the

Panel should expand its existing Task Committee on the Large-scale Testing Program for

providing scientific and technical advice to participating institutions in the Program.

The co-chairmen of this Task Committee shall be appointed by the co-chairmen of the

Panel. Additional Task Committee members shall also be appointed by the Panel co-chairmen. The Task Committee should consider recommendations from other Task Committees of

this Panel for expansion of the joint research program such as life line componennts,

repair and retrofit and soil-structure interaction.

4. Considerable advancement has been made by various Task Committees. Cooperative research

programs should be promoted through the Task Committees, specifically in the area of

repair and retrofit of structures.

5. A new Task Committee on Tsunami and Storm Surge is hereby established. Co-chairmen and

members of this Task Committee will be appointed by the co-chairmen of the Panel.

6. The date and location of the Twelveth Joint Meeting on Wind and Seismic Effects will be

May 1980 in Washington, D.C. Specific dates and itinerary will be determined by the

U.S. Panel with concurrence by the Japanese Panel.

XX
NEW INITIATIVES IN EARTHQUAKE HAZARDS MITIGATION

Charles C. Thiel

Earthquake Hazards Reduction Coordination Group

Office of Science and Technology Policy

Executive Office of the President

Washington, D.C.

r., ^ , ABSTRACT

Under the National Earthquake Hazards Reduction Program a number of significant steps

have been taken to improve national earthquake mitigation policies and practices. The

recently established Earthquake Hazards Reduction Coordination Group within the Executive

Office of the President, is coordinating a number of efforts to improve building practices,

land use, insurance, preparedness, emergency response and post-earthquake recovery activities. An "Interagency Committee on Seismic Safety in Construction" has been established

to: — develop seismic design and construction standards for Federal projects; — develop guidelines to ensure serviceability following an earthquake of vital

facilities constructed or financed by the Federal government; and — develop guidelines that provide independent and State and local review of

seismic considerations in the construction of critical facilities constructed

and financed by the Federal government, where appropriate.

In a collateral activity a "Building Seismic Safety Council" has been formed within the

private sector to enhance the public safety by providing a national forum to foster improved

seismic safety provisions for use by the building community. The scope of the Council's

activities encompasses seismic safety of building-type structures with explicit consideration and assessment of the social, technical, administrative, political, legislative, and

economic implications of its deliberations and recommendations.
NATIONAL EARTHQUAKE HAZARDS REDUCTION PROGRAM

The National Earthquake Hazards Reduction Act [1] established the purpose "to reduce the

risks of life and property from future earthquakes in the United States through the establishment of an effective earthquake hazards reduction program." The Act specified that the

objectives of the program shall include:

1) The development of technologically and economically feasible design and construction

methods and procedures to make new and existing structures, in areas of seismic risk, earthquake resistant, giving priority to the development of such methods and procedures for

nuclear power generating plants, dams, hospitals, schools, public utilities, public safety

structures, high occupancy buildings, and other structures which are especially needed in

time of disaster;

2) The implementation in all areas of high or moderate seismic risk, of a system

(including personnel, technology, and procedures) for predicting damaging earthquakes and for

identifying, evaluating, and accurately characterizing seismic hazards;

3) The development, publication, and promotion, in conjunction with State and local

officials and professional organizations, of model codes and other means to coordinate information about seismic risk with land-use policy decisions and building activity; ^

4) The development, in areas of seismic risk, of improved understanding of, and capability with respect to, earthquake-related issues, including methods of controlling the risks

from earthquakes, planning to prevent such risks, disseminating warnings of earthquakes,

organizing emergency services, and planning for reconstruction and redevelopment after an

earthquake;

5) The education of the public, including State and local officials as to earthquake

phenomena, the identification of locations and structures which are especially susceptible to

earthquake damage, ways to reduce the adverse consequences of an earthquake, and related

matters

;

6) The development of research on

(a) Ways to increase the use of existing scientific and engineering knowledge to

mitigate earthquake hazards;

(b) The social, economic, legal, and political consequences of earthquake

prediction; and
(c) Ways to assure the availability of earthquake insurance or some functional

substitute; and .. .

7) The development of basic and applied research leading to a better understanding of

the control or alternation of seismic phenomena.

The Act further specified that an implementation plan be prepared to achieve these

objectives. In response, the National Earthquake Hazards Reduction Program [2] (NEHRP) was

prepared and transmitted by the President to the Congress. This program sets forth a wide

ranging, comprehensive set of activities. Of the tasks outlined, the highest priorities for

immediate action are: — The establishment of a focus—a lead agency—to provide national leadership

and to guide and coordinate Federal activities; — The determination of the interest of States for the development of State and

local strategies and capabilities for earthquake hazards reduction; — The completion of Federal, State and local contingency plans for responding

to earthquake disasters in the densely populated areas of highest risk; — The development of seismic resistant design and construction standards for

application in Federal construction and encouragement for adoption of

improved seismic provisions in State and local building codes; — The estimation of the hazard posed to life by possible damage to existing

Federal facilities from future earthquakes; — The maintenance of a comprehensive program of research and development for

earthquake prediction and hazards mitigation.

The Earthquake Hazards Reduction Group (EHRG) was established in September 1978 as an

adjunct to the Office of Science and Technology Policy, Executive Office of the President, to

serve as the interim lead agency for coordination of the NEHRP. The President's Reorganization Plan No. 3 of 1978, June 19, 1978, specifies that the Federal Emergency Management

Agency (FEMA) will assume lead agency responsibilities upon its activation. The balance of

this paper describes several initiatives taken under the NEHRP.

3
INTERAGENCY COMMITTEE ON SEISMIC SAFETY IN CONSTRUCTION

The Interagency Committee on Seismic Safety in Construction was formed in October 1978.

It is chaired by the head of EHRG and has participation from over 130 individuals from 17

Federal agencies and departments. The Committee has three principle objectives: — develop seismic design and construction standards for Federal projects; — develop guidelines to ensure serviceability following an earthquake of vital

facilities constructed or financed by the Federal government; — develop guidelines that provide for independent and State and local review of

seismic considerations in the construction of critical facilities constructed

and financed by the Federal government, where appropriate.

The detailed work of preparing specific provisions for consideration by the Committee is

underway in ten subcommittees and one task group. Since these groups are pivotal to the

federal Program, their objectives are described below:

1) Format and Notation

The Subcommittee is to develop an overall plan for the organization, arrangement, and

terminology for standards for all types of Federal facilities. Specific areas of activity

are to assure that scope and organization of subcommittees' reports and standards are

consistent with their mission; definitions and symbols are consistent for reports and standards of the Interagency Committee; levels of technology, resulting levels of risk, and uses

of reference standards are consistent; and, identify research needs, receive and synthesize

research needs recommendations of other subcommittees, recommend or endorse research programs of Federal agencies to Interagency Committee.

2) Standards for Buildings

The Subcommittee is to recommend seismic design and construction standards for Federal

buildings, their appurtenances and nonstructural components. Recommendations received from

other subcommittees will be reviewed for incorporation into the standards developed by this

Subcommittee. The assigned target date for completion of this task is 1980. After

appropriate testing and evaluation by Federal construction agencies, the results will be

reviewed, and the standards will be revised as necessary and presented to the full Interagency Committee for adoption. A procedure will be developed for periodically reviewing

the adopted standards for the purpose of improvement.

4
3) Existing Hazardous Buildings

The Subcommittee is to develop a strategy and technique for the identification of

existing seismically hazardous buildings and to develop hazard mitigation techniques and procedures to reduce these hazards. , . . ,

4) Lifelines

This Subcommittee is to identify and recommend Federal standards for seismic design,

construction and retrofit of energy, transportation, water and telecommunication systems.

The Subcommittee will study strategies for evaluating the seismic vulnerability of existing

lifelines and for improving their resistance to seismic effects and ease of repair. These

strategies will permit identification of those lifeline facilities important in the emergency, immediate recovery, and long-term economic recovery periods, and provide guidance for

appropriate levels of seismic protection for each type. The Subcommittee will establish

liaison with existing professional and industrial groups active in the seismic design. The

Subcommittee will classify lifeline components into three groups: 1) those elements which

behave seismically in a manner similar to buildings and installations already adequately

covered by codes and standards; 2) those unique facilities where existing lifeline engineering information exists to prepare a standard; and 3) those where new technology or

research will be required to provide a technically sound recommendation.

5) Risk Analysis

The Subcommittee is to recommend a suitable risk analysis methodology for policy and

technical decisions on seismic protection in Federal design and construction programs and in

capital investment decisions. This Subcommittee will recommend a risk analysis methodology

that is sufficiently general to include consideration of existing structures and new construction. Various levels of risk will also be considered. The methodology of risk analysis is to permit joint consideration of the seismic risk at a specific site with other types

of risk for the same facility such as damage from fire, floods or high winds.

6) Federal Grant, Lease and Assistance Programs

The Subcommittee is to develop strategies for implementation of the Interagency

Committee's aseismic standards for Federal programs such as those involving grants, loans,

leasing, aid, assistance or regulatory programs. The Subcommittee will deal with policy

matters necessary to enable application of direct aseismic standards to these programs,

in conjunction with State and local codes. The Subcommittee will develop recommendations

5
for implementation and for regulatory procedures which can be adopted by individual agencies.

Proper coordination will be maintained between all related regulatory organizations and other

Federal, State and local agencies involved in construction under Federal assistance programs;

to assure resolution of conflicts prior to adoption of mandatory aseismic standards. An

implementation procedure will be developed for periodically reviewing aseismic standards

for the purpose of revising and upgrading the standards, and for adopting future application

standards.

7) Evaluation of Site Hazards •

The Subcommittee is to establish guidelines, procedures, and criteria for the evaluation

of seismic risk and seismically induced geologic hazards to Federally funded, assisted, and

regulated construction sites. Areas of concern include procedures for site vulnerability

analyses, identification of special study zones, and seismic risk maps. Hazards that will be

considered include: seismicity, ground shaking, liquefaction, landslides, surface faulting,

and subsidence.

8) Tsunami and Flood Waves ' -

This Subcommittee is to develop standards for the hydrodynamic aspects of design and

construction against tsunami, seiche, and flood wave threats. The mission includes development of procedures for tsunami risk assessment and proposal of methods for mitigating

possible hazards associated with tsunamis, seiches, and flood waves. .

9) Post-Earthquake Serviceability

This Subcommittee is to develop guidelines to ensure serviceability following an earthquake of vital community facilities constructed or financed by the Federal government. The

guidelines are to be directed toward those functions within the facilities of a community

which are vital during the post-earthquake period. These facilities include emergency

installations such as hospitals, communication centers, police and firefighting control

systems, and computer centers and disbursement centers whose loss could have a disasterous

impact on the community. ^ >

'-'

10) Evaluation of Seismic Safety of Critical Facilities i

The Subcommittee is to develop guidelines for involving Federal, State and local governmental bodies in the evaluation of seismic safety of critical facilities that are owned,

6
operated, regulated, or financed by the Federal government. Critical facilities are those

whose seismically induced failure might increase the destructive impact and public consequences of earthquakes by accidental release of contained materials.

Task Group 1. Unregulated Federal Nuclear Facilities

The mission of this Task Group is to assess whether standards for unlicensed and

unregulated Federal nuclear facilities should be developed by the Interagency Committee.

BUILDING SEISMIC SAFETY COUNCIL

A key element in the achievement of increased public earthquake safety is the continued

development, evaluation and improvement of model seismic design provisions suitable for

incorporation into local building codes and practices. A major impediment to achieving

this objective has been that the diverse elements of the building community (professional

societies, and labor, trade, model code, voluntary standards, public interest and public

agency organizations) have not had a forum to develop such standards. This has been remedied

through the formation, in April, 1979, of the Building Seismic Safety Council as an independent, voluntary body to enhance the public's safety by fostering improved seismic safety

provisions. The Council was formed through the efforts of organizations representing all

segments of the building industry, including ASCE, AIA, SEAOC , NCSBCS, CABO, NBS , PC, AGC,

NIBS, ATC and EHRG. The Council's objectives are to:

a) promote the development of seismic safety provisions suitable for use throughout

the United States;

b) recommend, encourage, and promote the adoption of appropriate seismic safety

provisions in voluntary standards and model codes;

c) assess implementation progress by Federal, State and local regulatory and

construction agencies; ... , .. , -

d) identify opportunities for the improvement of seismic regulations and practices

and encourage public and private organizations to effect such improvements;

e) promote the development of training and educational courses and materials for use by

design professionals, builders, building regulatory officials, elected officials, industry

representatives, other members of the building community and the public;

f) provide advice to governmental bodies on their programs of research, development,

and implementation; and

7
g) periodically review and evaluate research findings, practice, and experience and

make recommendations for incorporation into seismic design practices.

The first specific task undertaken by the Council is to extensively review the Tentative

Provisions for the Development of Seismic Regulations of Buildings , [3] prepared by the

Applied Technology Council under contract to the National Bureau of Standards with the

National Science Foundation's support. This document contains tentative seismic design provisions for use in the development of seismic code regulations for design and construction of

buildings. The provisions represent the result of a concerted effort by a raultidisciplinary

team of nationally recognized experts in earthquake engineering. Design professionals,

researchers. Federal agency representatives, staffs from the model code organizations and

representatives from State and local governments throughout the United States were involved.

The provisions are comprehensive in nature and deal with earthquake resistant design of the

structural system, architectural and non-structural elements and mechanical-electrical

systems in buildings. Both new and existing buildings are included. They embody several

new concepts which are significant departures from existing seismic design provisions. An

extensive commentary documenting the basis for the provisions is included.

The Council has developed a review procedure depicted in figure 1 that is quite extensive to assure that the resulting procedures are widely considered on their technical,

administrative and applications merit. The NBS has agreed to provide technical management

of some of the tasks as indicated. BSSC will maintain responsibility for the overall effort

as well as the tasks indicated. After completion of this review, modification of the provisions, and test by trial designs the member organizations will be encouraged to adopt the

resultant provisions. It is important to note that the Council has not been formed to

replace the initiatives of its member organizations, but is to provide a national forum where

its members can develop seismic design provisions that are responsive to the public's interest. This is a unique undertaking in the building community for which there is no precedent

within the U.S. The opportunity for different groups (e.g., public officials, owners, architects) to consider the merits of proposed provisions before they are incorporated into a code

or national standard without the usual heated advisory climate should substantially improve

the process of reaching consensus views which provide appropriate and effective public

earthquake building regulations.

8
EARTHQUAKE RESPONSE PLANNING

From 1971 to 1976, the Federal Disaster Assistance Administration (FDAA) and its predecessor agency had conducted earthquake response planning studies [4] to respond to the special needs of earthquake victims. The NEHRP has reinstated that program. Concentrated in a

limited number of heavily populated areas in highly seismic zones, it consists of two phases:

a) an estimate of casualties and physical damage to critical facilities that might occur as

a result of one or more maximum credible earthquakes in the area; and b) the development of

response plans by the regional Federal agencies —coordinated by FDAA—to provide assistance

to supplement local and State resource needs identified by companion State and local plans.

In providing new emphasis to this program, a significant shift in approach was also

made. The affected States have been asked to assume a leadership role in the preparation of

both damage estimates and State and local plans, rather than just the latter. Considerable

effort will now be expended in marshalling the support of local jurisdictions and identifying

more specifically the resources that State and local organizations will provide —so that any

shortfall potentially can be filled by supplemental Federal assistance.

Tables 1 and 2 summarize the programs progress to date and expected future activities.

FY 79 activities have concentrated on starting a loss study for the Anchorage, Alaska area

and communities on Prince William Sound (containing critical facilities), under the leadership of the Alaska State Emergency Services. The initial phase of the effort, stressing the

geotechnical aspects and the encouragement of participation of local jurisdictions is

expected to be completed in early 1980. A somewhat similar effort, directed at planning for

coping with an earthquake and tsunami that might affect Honolulu, Hawaii, is expected to be

started this summer, with active participation of the State.

EARTHQUAKES AND FINANCIAL INSTITUTIONS

A study of earthquakes and their impacts on financial institutions has been undertaken

with three related objectives by EHRG with the cooperation of affected Federal agencies:

1) The viability of the financial system in the aftermath of a truly catastrophic

earthquake in a densely populated area. The system is known to have resilience, but the

degree of resilience in such an eventuality has never been examined.

9
2) The effects of a credible earthquake prediction upon the behavior of the public and

its consequent demands on the financial system. Preliminary indications point to some possibly heavy burdens to be placed on the system, particularly if the prediction time window is

of considerable length (i.e., more than six months).

3) The feasibility of effecting mitigation of the earthquake hazards through the financial institutions, without causing undue dislocations in the process. As in the case of fire

hazards —where financial institutions have played a significant role in encouraging better

codes and prevention programs —it may be possible to identify non-disruptive mechanisms to

cope with earthquake hazards reduction.

There are many kinds of financial institutions, both private and public. The private

institutions are financial intermediaries; the public ones, regulatory and subsidizing agencies as well as financial intermediaries. For private institutions, especially deposit-type

institutions, earthquakes exacerbate three ordinary financial concerns —liquidity, solvency,

and records maintenance and replication. These concerns impact on the public institutions,

whose roles require careful analysis to determine what public policy problems, if any, exist

and what action may be indicated. The study is expected to be completed by the end of 1979,

with policy recommendations by March 1980.

EARTHQUAKE PREDICTION EVALUATION COUNCIL '

'

The Council has been formed, as specified in the National Earthquake Hazards Reduction

Program. It is made up of members appointed by the Director of United States Geological

Survey. The Council is to advise the Director on issuing predictions as to the completeness and scientific validity of the available data. In evaluating predictions, the Council's

objectives are:

to recommend to the appropriate scientists any actions that might be desirable

or required to clarify or verify the basis for a prediction; <

to maintain an accurate record of predictions evaluated and evidence pertinent

to them; and — to provide the Director a timely and concisely written review of the evidence

relevant to a prediction of any potentially damaging earthquake and a» written

10
recommendation as to whether the evidence is sufficiently clear that an official prediction by the Director should be issued or, if not, what if any other

official position the Director should take. '

As of this date no predictions have been submitted for their evaluation.

RESEARCH PROGRAM

FY 1979 is the second year of substantial research under the National Earthquake

Hazards Reduction Program. PL 95-124 and the report Earthquake Prediction and Hazard

Mitigation Option for USGS and NSF Programs , [5] 1979 (Newmark Report) provide the basis for

these agencies programs. The main elements of the program paraphrased from the Newmark

Report are:

I Fundamental studies —research into the basic causes and mechanisms of earthquakes.

II Prediction—forecasting the time, place, magnitude and effects of an earthquake.

III Induced Seismicity—prevention or modification of an inadvertently induced or

natural earthquake.

IV Hazard Assessment —identification and analysis of the potential for earthquakes within

a region, their frequency and their effects. ;

V Engineering—design and construction of structures for acceptable performance during

and after an earthquake. '

. .

,

VI Policy Research —impacts of earthquakes on the community and options for dealing with

them.

'

EHRG is currently leading an interagency effort among NSF, USGS and NBS to develop

program options for Fiscal years 1981 through 1984. It is expected that FEMA once it is

fully operational will assume an active role in supporting research, particularly on related

natural disasters, which will strengthen the earthquake effort.

REFERENCES

[1] U.S. Congress, Public Law 95-124, October 7, 1977. Also printed as Appendix I of

Earthquake Hazards Reduction: Issues for an Implementation Plan , (Office of Science

and Technology Policy, Executive Office of the President, Washington, D.C., 1978).

[2] National Earthquake Hazards Reduction Program (Office of Science and Technology Policy,

Executive Office of the President, June 22, 1978).

11
Applied Technology Council, Tentative Provisions for the Development of Seismic

Regulations for Buildings , U.S. Government Printing Office, 1978, Number 003-003-

01939-9.

Department of Commerce, "A Study of Earthquake Losses in the Los Angeles, California,

Area," U.S. Government Printing Office, 1973, Number 0319-00026.

Department of Commerce, "A Study of Earthquake Losses in the San Francisco Bay Area,"

U.S. Government Printing Office, 1972, Number 4101-00011.

Department of Interior, "A Study of Earthquake Losses in the Puget Sound, Washington,

Area," U.S. Government Printing Office, 1975, Number 681-256/29.

Department of Interior, "A Study of Earthquake Losses in the Salt Lake City, Utah,

Area," U.S. Government Printing Office, 1975, Number 677-530/292.

National Science Foundation and Department of the Interior, Earthquake Prediction and

Hazard Mitigation Options for USGS and NSF Programs , U.S. Government Printing Office,

1976, Number 038-000-00332-1.

HATIONAL BUREAU OF STANDARDS BUILDING SEISHIC SAFETY COUNCIL

APPOINTS COWITTEE

TO REVIEW TRIAL DESIGNS PROCEDURES

DRAFT

FOR REVIEW

AND COWMENT

ICONbu^

SELECTE

p-RIAL IDESIGNS

COWITTEE CONSIDERS

COMMENTS. ETC. AND REFORMULATES

(IF NECESSARY) PROPOSED TRAIL DESIGN PROCEDURES

INDIVIDUAL RESPONSES

ETC.

FINAL COMtlENTS ETC.

RECOmENDED

TRIAL DESIGNS PROCEDURES

PREPARED ,

APPOINTS OVERVIEW

COWMITTEE APPOINTS COWITTEE AND SUBCOWIHEES TO RECONCILE

COfWENTS ON ATC 3-06 AND PROPOSE REVISION DOCUMENT

J-\.

REQUESTS BSSC MEMBERS TO PARTICIPATE REVIEW AND ADVISE NBS ON COMI^mEE COMPOSITION

AND WORK PLANS (BOTH)

REQUEST TECHNICAL REVIEW

AND COMMENTS ON ATC 3-06; SUBMISSION OF ANY TRIAL DESIGNS CONDUCTED; WORKSHOPS

OR MEETINGS HELP AS REQUIRED

CRITICAL

ISSUES DISCUSSION

AT ANNUAL MEETING

COMMITTEES AND SUBCOMMITTEES CONSIDER COMMENTS. SELECTED

TRAIL DESIGNS AND FORMULATES

iPROPOSED REVISIONS

REPORT TOl BOARD AND COUNCIL I

OVERVIEW

(BOTH)

INDIVIDUAL RESPONSES TO EACH COW'ENTER ON DISPOSITION

OF COMMENTS AND ACCOMODATIONS

OR CHANGES (IF ANY)

OPEN BSSC MEETING

DISCUSSION PANELS

ON PROVISIONS AND

TEST DESIGN METHODS

FINAL COtWENTS IN REBUTAL

BY INDIVIDUALS IF THEY ARE STILL UNSATISFIED

REPORT TO -» BOARD AND

COUNCIL

OVERVIEW

(BOTH)

DOCUMENT TO BSSC FOR CONSIDERATION IN TRIAL DESIGNS PREPARED

REPORT TO BOARD AND

COUNCIL

OVERVIEW CWWITTEE

RECOttlENDS TO BOARD PROCEDURES TO BE FOLLOWED

IN BSSC CONSIDERATION

OF RECCT»<ENDED DOCUMENTS

ICONS I DERATION, DISCUSSION

AND BALLOTING BY BSSC OF PROVISIONS FOR TRIAL

DESIGN APPLICATION BY

iBSSC MEMBERSHIP

DOCUMENT FOR CONDUCT

OF TRIAL DESIGNS

FIGURE 1: BSSC Workplan for Review of Seismic Design Provisions

12
, TABLE 1 ,

EARTHQUAKE LOSS STUDIES AND CONTINGENCY PLANS: PROGRAM PLAN

Legend: Loss Studies (risk analyses, damage estimates); NWM Response Plans (Federal, State, local)

Fiscal Years Region/State Population Area 1971-76 1979 1980 1981 1982 1983 1984

9 California San Francisco Bay

Loss stud

by F-S-L

/ comple

jlng; Fe ted 1972,d ping res followed

umed 1978

9 California Los Angeles Area

Study com ping; Fed Dieted 1 plan to 973 follow

follow SF

ed by S&L

pi an

10 Washington Puget Sound Area Study

1975

- .Federa 1 Plan

8 IWWSWSM

a Utah Salt Lake City Area Study

1976 Federal

Local p

Plan 8/78,

lans 1977-79

State Eir under F srgency PI 3AA preparin 7/77

3dness gnt

10 Alaska Anchorage Area mm

9 Hawaii Honolulu Area

4 S. Carolina Charleston Area Order o f these

1 Massachusetts Boston Area

may be reversed

Central U.S. Mississippi Valley

FY 8U Des Prbject

ign \V\\\\\V

2 New York** Rochester- Seaway swwwsn:

*At least Regions 4, 5, 6, and 7 (Kentucky, Tennessee, Mississippi, Illinois, Indiana, Missouri and Arkansas

are In the Zone 3--maJor damage--area ; major Zone-3 city: Memphis; Zone 2 Includes St. Louis, Cincinnati)

'Zone-3 extends from extreme western New York to northern tip of Maine, Includes Buffalo and Montreal

13
TABLE 2: EARTHQUAKE LOSS STUDIES AND CONTINGENCY PLANS: STATUS

Loss

Study

Federal

Response

State

Response

Local

Response

Functional

Study

San Francisco X X X X

Los Angeles X FY 80 X X Temporary

Housing Underway

Salt Lake A X X X

Puget Sound X X FY 81

Anchorage FY 79 FY 80 FY 80 Winter Housing

Underway

Honolulu FY 79 FY 81 FY 80

Central U.S.* FY 80 FY 82 FY 83

Charles tovm S. Carolina** FY 80 FY 81 FY 81

Boston** FY 81 FY 82 FY 82

Upstate

New York FY 81 FY 82 FY 82

To be preceded by a project design effort In FY 80. X — Completed

Order of these two items may be reversed

Newmark Report

Recommendation

1. Fundamental

Studies

2. Earthquake

Prediction

3. Induced

Seismicity

4. Hazards

Assessment

5. Engineering

6. Research for

Utilization

TOTAL

uses

2.9

15.9

1.2

11.1

31.1

NSF

6.0

16.6

0.8

23.4

Total

8.9

15.9

1.2

11.1

16.6

0.8

54.5

Option B

9.1

17.9

2.0

14.8

20.3

6.1

70.2

Option A

7.8

13.6

1.0

10.7

11.6

5.3

50.0

TABLE 3: FY 79 Research Expenditures (Amounts in Millions of Dollars).

14
EARTHQUAKE HAZARD REDUCTION RESEARCH AT

THE NATIONAL BUREAU OF STANDARDS

Edgar V. Leyendecker

James R. Harris

Richard N. Wright

Edward 0. Pfrang

Center for Building Technology

National Bureau of Standards

U.S. Department of Commerce

ABSTRACT

Current and planned Earthquake Hazard Reduction programs at the National Bureau of

Standards in Research and Standards Development are being conducted in order to achieve the

responsibilities assigned to NBS under the President's National Earthquake Hazards Reduction

Program. These responsibilities, to:

1) provide technical support to the building community in the development of seismic

design and construction provisions for building codes and national standards

2) provide technical support to the Federal agencies in development of seismic design

and construction provisions for Federal programs and

3) perform research on performance criteria and supporting measurement technology for

earthquake-resistant construction,

are being carried out in cooperation with the Federal and private sectors. This research is

complementary to that of the National Science Foundation (NSF) and U.S. Geological Survey

(USGS) . .o

15
INTRODUCTION

The National Bureau of Standards (NBS) Earthquake Hazards Reduction Program is part of

the national effort to reduce the risks to life and property from future earthquakes in the

United States. Under the President's National Earthquake Hazards Reduction Program [1]*,

issued June 22, 1978, to implement the Earthquake Hazards Reduction Act of 1977 [2], NBS has

been assigned the following continuing responsibilities:

° provide technical support to the building community, including the Federal

agencies involved in construction, "in continuing the development, testing,

and improvement of model seismic design and construction provisions suitable

for incorporation in local codes, standards (including Federal standards) and

practices .

° perform "research on performance criteria and supporting measurement technology

for earthquake resistant construction."

NBS has participated in the planning of Federal programs in earthquake hazard reduction

since 1971 and played a major role in providing input which was used in drafting the National

Earthquake Hazards Reduction Program.

OBJECTIVES

"

In order to meet its responsibilities under the President's National Earthquake Hazards

Reduction Program, NBS is concentrating its efforts in (1) Research for Standards and

(2) Standards Development. The objective of these areas are:

Research for Standards - Improve building design and construction practices by

conducting fundamental research necessary to establish consistent standards that

include reliable measures of the performance of buildings and their environment

during an earthquake.

Standards Development - Provide technical support in the "development of seismic

design and construction standards for consideration and subsequent application

in Federal Construction, and encouragement for the adoption of improved seismic

provisions in State and local building codes."

* Numbers in brackets indicate references.

16
Both of these objectives are consistent with past NBS efforts in the development of

technical bases for standards and on measurement technology and with the interagency effort

among NSF, USGS, and NBS to develop program options for Fiscal years 1981-1984 [3].

RESEARCH FOR STANDARDS

A research program at NBS to develop technical bases for performance criteria and the

supporting measurement technology is a new element of the national earthquake research program. The research is complementary to the U.S. Geological Survey (USGS) and the National

Science Foundation (NSF).

Earth sciences research led by the USGS and NSF provides greatly improved understanding

of ground shaking and displacement. To be suitable for use in building design, the knowledge

must be further developed to loading criteria for the various types of buildings and components. Structural, geotechnical , mechanical, architectural and economic studies supported

by NSF have provided substantial advances in abilities to model and predict the earthquake

responses of various built elements and their consequences. This knowledge must be synthesized to develop design criteria that will provide consistent safety among the various built

elements and various materials or technologies for each. Evaluation and measurement methods

consistent with practices of design, manufacture, construction and inspection, must be developed to implement the design criteria.

The NBS research program is essential to the NBS standards development role and will

complement the research conducted or funded by USGS and NSF and contribute substantially to

the implementation of the latter research. Also the NBS research program provides for an

enhanced Federal expertise in the area of earthquake engineering research. This expertise

is vital for effective Federal participation with the building community and during the

establishment of public policy.

Research for standards are divided into four tasks:

Task 1 - Research for Earthquake Loading Standards -

Develop methods for use in standards to characterize ground motion and the acceleration,

velocity, and displacement time-histories of this motion for use in engineering analysis,

planning and design. The effect of the earthquake source, transmission path, amplification

caused by local site conditions, and soil-structure interaction will be considered.

17
Task 2 - Research for New Building Standards -

Develop methods for use in standards to characterize the performance of buildings in

four areas :

i) Develop laboratory and in-situ techniques for characterizing the dynamic

properties of soils, develop analytical methods to evaluate soil failures, and

investigate the design of various types of foundations in order to develop criteria

for selection and design of appropriate foundations. • •..

ii) Develop procedures for characterizing the dynamic behavior of structures

and components up to the ultimate limit state and provide a basis for the formulation and validation of nonlinear and inelastic methods of analysis and design.

iii) Develop methods to characterize the response of non-structural systems

to earthquake loadings and develop test methods for evaluating the characteristics.

Methods to obtain loadings for use in evaluating non-structural systems as a transfer from the structural and site response information will be developed. Critical

response parameters will be formulated and test methods will be developed for

evaluating the characteristics of non-structural systems and components such as

machines, luminaires, piping, elevators, etc. System design conflict arising from

non-structural response (secondary hazards, such as fire) will also be examined.

iv) Develop the basis for performance criteria for occupant safety and

building functional requirements during and after earthquakes by relating casualties to both structural and non-structural response to earthquake motion and by : '

determining functionality requirements both during and after earthquakes for

various types of building uses. -

Task 3 - Research for Existing Building Standards -

Develop nondestructive methods to determine the properties of older buildings and

components in place, for use in determining realistic values of structural parameters and

evaluation of the strength of existing buildings and repairs. Analytical and laboratory

studies on components will be followed by field studies on whole buildings, taking advantage

of buildings scheduled for demolition where possible.

Task 4 - Research for Assessment of Standards -

Develop methods for assessing the impact of standards for improved seismic resistance.

Methodologies will be developed for assessing the benefits of earthquake hazard reduction

18
and defining risk for both building and community scales. The influence of building type and

configuration on serviceability and costs will be considered. Costs and benefits of improved

practices for seismic resistance associated with planning, design, construction and regulation will be included. Methods for assessing, the clarity, consistency, and completeness of

standards will also be improved.

Research in each of these tasks is planned to begin in Fiscal year 1981, depending on

approved funding. At the current level of funding, research is concentrating on Task 2 in

the development of procedures for characterizing the dynamic behavior of structures, particularly for masonry construction.

STANDARDS DEVELOPMENT

A program for the development of standards for buildings is an important element for

the implementation of the substantial research programs at USGS and NSF. Criteria for the

earthquake-resistant design of new construction used in many current Federal, State, and

local building codes, standards and practices do not reflect the current state of the art

and should be updated. These codes and the standards and professional practices underlying

them should not only represent our best knowledge, but be adaptable to different areas of

the United States according to differing seismic risks and the costs and benefits they

entail.

Under the leadership of the Office of Science and Technology Policy, an Interagency

Committee on Seismic Safety in Construction was organized in 1978. All Federal agencies

involved in construction, working through the Interagency Committee on Seismic Safety in

Construction, will develop seismic design standards for Federal building construction. NBS

is providing technical support to several of the subcommittees of the Interagency Committee

in the development of unified Federal standards, including chairing the subcommittee on

Building Standards [3]. . ;

• ":• ;.

The vast majority of the construction in the United States is undertaken by the private

sector and regulated by local government. To assist State and local governments, industry,

and the public in developing construction standards, criteria, and practices, the National

Bureau of Standards will work with other Federal agencies, the National Institute of Building

Sciences, professional organizations, model code groups. State and local building departments,

and the newly formed Building Seismic Safety Council. The Bureau will assist and cooperate

19
with these groups in continuing the development, evaluation, and improvement of model seismic

design provisions suitable for incorporation into local codes and practices. Incorporation

of these seismic design provisions into local codes is, of course, voluntary, but the provisions must be flexible and give consideration to costs and benefits, regional variation of

seismic hazard, and adaptation to local conditions. They must also be adequately tested.

The Building Seismic Safety Council formed in April 1979, when more than fifty professional societies, trade associations, model code organizations and public interest organizations met under the auspices of the National Institute of Building Sciences. The Council

provides a national forum to foster improved seismic provisions for use by the building

community. NBS will conduct technical studies in support of the Council. In turn, the

Council will base its recommendations for provisions in building standards and codes on

these studies.

Standards Development is divided into three tasks:

Task 1 - Seismic Design Provisions for Consensus Standards and Model Codes - -

Provide technical studies and draft provisions to assist the building community in

the development and implementation of seismic design provisions in State and local building

codes. This will be accomplished by working through the existing systems of voluntary standards and model building codes, making appropriate use of the assessments of the Federal '

building standards in Task 2.

Task 2 - Seismic Design Standards for Federal Building Construction -

Develop seismic design standards for Federal building construction, test and assess the

safety and economic consequences of their use, improve the standards following assessment and

implement in Federal programs.

Task 3 - Standards for Evaluation and Strengthening of Hazardous Existing Buildings -

Provide technical assistance in the development of methods for the evaluation of seismic

safety of existing buildings and in the development of practices and criteria for strengthening existing buildings or removing them from service.

Research in each of these tasks is planned to begin in Fiscal year 1981, depending upon

approved funding. At the current level of funding NBS is working with the Interagency Committee on Seismic Safety in Construction to develop uniform Federal criteria.

20
Efforts are now under way to assess the "Tentative Provisions for the Development of

Seismic Regulations for Buildings [4]," proposed by the Applied Technology Council under

contract to the NBS with NSF funding. This effort will be in cooperation with the Building

Seismic Safety Council, and will involve both the Federal and private sectors.

CONCLUSION

The role of the NBS under the National Earthquake Hazards Reduction Program includes

the responsibility to:

1) provide technical support to the building community in the development of seismic

design and construction provisions for building codes and national standards,

2) provide technical support to the Federal agencies in development of seismic design

and construction provisions for Federal programs, and

3) perform research on performance criteria and supporting measurement technology for

earthquake-resistant construction.

These responsibilities are being met by conducting (1) Research for Standards and

(2) Standards Development. Current efforts are concentrating on (1) developing procedures

for characterizing the dynamic behavior of structures, particularly masonry construction, and

(2) working with the Interagency Committee on Seismic Safety in Construction to develop

uniform Federal criteria and with the Building Seismic Safety Council to assess the newly

prepared Tentative Provisions for the Development of Seismic Regulations for Buildings.

Current and future plans are consistent with the interagency effort among NSF, USGS,

and NBS to develop program options for Fiscal years 1981-1984.

REFERENCES

[1] National Earthquake Hazards Reduction Program, Office of Science and Technology Policy,

Executive Office of the President, June 22, 1978.

[2] U.S. Congress, Public Law 95-124, October 7, 1977, Also printed as Appendix I of

Earthquake Hazards Reduction: Issues for an Implementation Plan, (Office of Science

Technology Policy, Executive Office of the President, Washington, D.C., 1978.

[3] Thiel, C. C. , New Initiatives in Earthquake Hazards Mitigation, 11th Joint UJNR

Conference, 1979.

[4] Applied Technology Council, Tentative Provisions for the Development of Seismic

Regulations for Buildings , SP-510, U.S. Government Printing Office, Washington, D.C.,

1978.

21
OUTLINE OF RESEARCH ON WIND AND EARTHQUAKE EFFECTS IN JAPAN

Yoshijiro Sakagaml

Public Works Research Institute

Note: Though Mr. Sakagami presented the research outline

as cited above, no written version could be obtained

for inclusion in these proceedings.

22
ON THE U.S. -JAPAN COOPERATIVE PROGRAM OF LARGE-SCALE TESTING

Tadayoshi Okubo

Public Works Research Institute

Ministry of Construction

1. INTRODUCTION

The Implementing Arrangement between the Science and Technology Agency (STA) and the

Ministry of Construction (MOC) of Japan and the National Science Foundation (NSF) of the

U.S.A. for cooperation in the U.S. -Japan Joint Earthquake Research Program involving LargeScale Testing, under the auspices of the Panel on Wind and Seismic Effects of the UJNR was

signed by Mr. Nobuo Kozu (STA), Mr. Hidenobu Takahide (MOC) and Dr. Jack T. Sanderson (NSF)

on August 10, 1979. (Appendix I).

The U.S. -Japan Cooperative Program on Large-Scale Testing was initiated in 1974 by this

Panel and has begun implementation. In this presentation, the author outlines the Program.

2. BACKGROUND

At the Sixth Joint Meeting of the UJNR Panel on Wind and Seismic Effects, held in Washington, D.C. May 15-17, 1974, the necessity of a joint research program which includes large

scale testing of structures was emphasized and the Panel produced the following resolution:

"3) that increased effort should be made in the near future to encourage joint research

programs, especially in the area of the mutual utilization of research facilities and the

exchange of researchers."

At the Seventh Joint Meeting of our Panel held in Tokyo, May 20-23, 1975, a special

session, "Theme IX: Joint Research Program Utilizing Large-Scale Testing Facilities" was

held and the exchange of information and opinion was carried out with two papers submitted to

the session: "A Proposal to the Japan - U.S. Joint Earthquake Engineering Research Project

Utilizing Large-Scale Testing Facilities," by M. Watabe, M. Hirosawa, and S. Nakata, and

"Earthquake Disaster Mitigation: A Joint Research Approach," by Charles T. Thiel. The Panel

concluded with the following resolution:

"7) Cooperative research programs including exchange of personnel and equipment should

be undertaken by both governments to address the following problems of mutual interest;

23
a) Strong motion Instrumentation arrays, at selected sites throughout the world,

b) Large-scale testing programs,

c) Repair and retrofit of existing structures, i.e., buildings, bridges, dams etc.,

d) Structural performance evaluation,

e) Land use programs for controlling natural hazard effects,

f) Disaster prevention methods for lifeline systems."

Based upon the above resolution, six Task Committees were established prior to the

Eighth Joint Meeting of the Panel and the Task. Committees energetically started their activities .

After the 1968 Tokachi-oki earthquake in Japan, a U.S. -Japan Cooperative Research Program

on Earthquake Engineering which emphasized the Safety of School Buildings was established

under the U.S. -Japan Cooperative Science Program for May 1973-October 1975. At the review

meeting of the Cooperative Research Program, held in Hawaii, August 18-20, 1975, the following

set of recommendations was drafted and signed by all official participants of the meeting.

"Recommendation No. 2. Establish a Cooperative Research Program in Earthquake Engineering with Emphasis on Large-Scale Testing of Structural Systems under the Sponsorship of the

U.S. -Japan Panel on Wind and Seismic Effects, UJNR Program. This program would concentrate

on controlled dynamic testing of full-scale buildings in the field and laboratory testing

of large-scale building systems using the available specially designed testing floors and

associated reaction walls and using a large-sized shaking table. These tests should be

designed to provide needed information on force-deformation, energy absorption and failure

characteristics of such systems. Correlative and analytical studies should be made to

improve mathematical modeling and computer-oriented dynamic analysis capability.

Recormnendation No. 3. Establish a Task Committee under the U.S. -Japan Panel on Wind and

Seismic Effects, UJNR Program. The assignment given to this Task Committee should be to make

detailed plans and recommendations for implementing Recommendation No. 2 which will ensure an

effective research program of maximum benefit to both countries within the constraints of the

sponsoring government agencies.

In view of the time scale for constructing the presently planned test facilities and

the major effort required to plan the program, the Task Committee should be appointed by the

U.S. -Japan Panel on Wind and Seismic Effects at the earliest possible date. One American and

one Japanese should be appointed as co-chairman to head this Task Committee."

24
At the Eighth Joint Meeting of our Panel, held in Washington, D.C., May 18-21, 1976, the

Task Committee on Large-Scale Testing Program concluded the following.

"1) Planning activities should be initiated through the combined efforts of U.S.

Federal Government and the Government of Japan for programs in large-scale testing of structures under the administration of the United States-Japan Panel on Wind and Seismic Effects.

2) Collaborative programs should be initiated in the areas of wind and seismic effects

so that technological advances can be utilized to best advantage. Excellent large-scale test

facilities in operation or proposed in Japan will provide a vehicle for advancement in technology. These large-scale test facilities should be combined with technological advances in

instrumentation, data acquisition and analysis, and measurements on structures in service so

that solutions of programs of mutual interest will be optimized.

3) The Task Committee exchanges the list of plans, including organization equipment

and status, on wind and seismic effects concerning these large-scale testing programs. It

is anticipated that the Task Committee can update the list periodically using inputs provided

by members of the U.S. -Japan Panel and individuals aware of large-scale testing programs

underway."

In succession, the Eighth Joint Meeting of the Panel adopted following resolution:

"2) The resolutions of the Seventh Joint Panel Meeting are important to the continued

cooperative program and are hereby incorporated as a part of these resolutions:

e) Promote cooperative research programs through the Task Committees including exchange

of personnel and available equipment. Publish reports of these Task Committees."

Meanwhile, based on a proposal, "Planning a Cooperative Research Program in Earthquake

Engineering with Emphasis on Large-Scale Testing of Structural Systems," submitted by the

University of California-Berkeley to the NSF which called for planning effort to be conducted under the auspices of the U.S. -Japan Panel on Wind and Seismic Effects, the Planning

Group was established and an adjustment of activities between our Panel and the Group was

necessitated. (The activity of the Planning Group is introduced later.)

Therefore, an unofficial meeting for the adjustment of the Panel and the Group activities was held on May 23, 1977, the day before the Ninth Joint Meeting of the Panel.

Professor H. Umemura, Japanese co-chairman of the Planning-Group , Dr. E. 0. Pfrang, U.S.

co-chairman of the Panel, Mr. K. Ohtani, Dr. M. Watabe and the author participated in this

meeting and concurred on the following.

25
"1) The cooperative research program on large-scale testing should be carried out under

the auspices of the UJNR Panel on Wind and Seismic Effects.

2) UJNR members of the Task Committee on Large-Scale Testing Programs should officially

participate with the Planning Group for Large-Scale Testing.

3) The final report to be prepared by the Planning Group should be submitted to the

UJNR Panel for review and suggestions relative to implementation of the cooperative test

program." , w .

.

,

Following this unofficial meeting, the Task Committee on Large-Scale Testing Programs

at the Ninth Joint Meeting of the Panel in Tokyo, May 24-27, 1977, concluded as follows:

"2) Future Programs

1) Official participation of UJNR members on the U.S. -Japan Planning Group

for large-scale testing chaired by Dr. J. Pengien and Dr. H. Umemura shall be

carried out.

2) The Planning Group and Task Committee shall coordinate their efforts.

3) The first joint meeting of the Planning Group with Task Committees is

expected to be held in the fall of 1977, in Japan.

4) The final report to be prepared by the Planning Group should be submitted to the UJNR Panel for review and suggestions relative to implementation of

the cooperative test program.

5) Japanese Committee members will endeavor to complete arrangements, in the

near future, for the official acceptance of the U.S. -Japan cooperative research

programs and participants."

The Ninth Joint Meeting of our Panel endorsed the conclusion of the Task Committee by

adopting the following resolution:

"4) Considerable advancement in various Task Committees was observed. These activities should be encouraged and continued. Correspondence between Task Committee chairmen

should be encouraged, and additional meetings should be held as required."

At the Tenth Joint Meeting of our Panel, held in Washington, D.C., May 23-26, 1978,

summaries of the first and second Joint Meeting of the Planning Group were reported to the

Task Committee Meeting and the following was agreed on at the Task Committee Meeting:

"The Task Committee on Large-Scale Testing Programs should keep the UJNR Panel on Wind

and Seismic Effects informed of developments in large-scale testing programs. Programs

26
proposed by the current planning group should be circulated to UJNR for review and comment

with appropriate recommendations returned to the planning group for consideration.

Also, the Tenth Joint Meeting of our Panel drafted the following resolutions.

"3. The Panel on Wind and Seismic Effects recognizes the importance of the U.S. -Japan

Cooperative Program on Large-Scale Testing and it urges early implementation of the program

under the auspices of this Panel.

4. Considerable advancement has been made by various Task Committees. Cooperative

research programs should be promoted through the Task Committees."

The Planning Group has held four meetings: Tokyo, Japan, September 5-10, 1977; San

Francisco, California, May 15-19, 1978; Tokyo, Japan, December 18-23, 1978; and BerkeleyCalifornia, July 9-14, 1979.

At the earlier meetings, six types of buildings — reinforced concrete, steel frame,

precast and/or prestressed concrete, steel-reinforced concrete, masonry, and wooden — were

discussed. Reinforced concrete and steel frame buildings, and the pseudo-dynamic test method

were selected for cooperative research. The summaries and resolutions of the four meetings

are shown in Appendix II.

3. COOPERATIVE RESEARCH PROGRAM RECOMMENDED BY THE PLANNING GROUP

The final meeting of the fourth meeting of the Planning Group recommended three research

programs. Details of these research programs are shown in Appendix III.

Recommended research on reinforced concrete building structures consists of a full-scale

seven-story reinforced building structure to be tested in the Large-size Structure Laboratory, Building Research Institute, Ministry of Construction, and associated tests to be conducted in Japan and the U.S. on the test specimens of element assemblies, structural walls,

planer frames, wall-frame assemblies and scaled models of the full-scale seven-story

structures

.

Recommended research on steel building structures consists of a full-scale seven-story

steel building test and complementary tests on structural components, beam-and-column

assemblages, braced and unbraced frame bents and small- and medium-scale models of the prototype structure.

27
Recommended research on the pseudo-dyamic test method consists of verification and

sensitivity studies of the numerical algorithm, verification of accuracy of hydraulic actuator control system and verification of adequacy of basic differential equation of motion.

4. IMPLEMENTING ARRANGEMENT

At the occasion of the third meeting of the Planning Group held in December, 1978, the

first draft of the implementing arrangement (at that time it was called a "Memorandum of

Understanding") was officially proposed by the U.S. side.

After surveying the first draft, the Japanese side made up a counter proposal (second

draft) and sent it to the U.S. side in May, 1979.

On June 19, 1979, members concerned of the Ministry of Construction and the Science and

Technology Agency of Japan and the National Science Foundation of the U.S. held a meeting in

Tokyo. Mr. J. Bloom, the Science Counselor of the U.S. Embassy, also attended at this meeting. The third draft (the official title was "Implementation Protocol") proposed by the U.S

side was discussed, however complete concurrence could not be reached.

During and after the fourth meeting of the Planning Group in July 1979, negotiation on

the third draft was being continued and during the negotiation the draft title was changed t

"Implementing Arrangement."

Finally in early August, 1979, a draft agreed to by both sides was codified as shown in

Appendix I and was signed on August 10. However, there are many detailed items which should

be agreed to by both sides in the "exchange of letters."

APPENDIX I

IMPLEMENTING ARRANGEMENT ; 4f BETWEEN THE SCIENCE AND TECHNOLOGY AGENCY AND t THE MINISTRY OF CONSTRUCTION OF JAPAN

AND THE NATIONAL SCIENCE FOUNDATION OF THE UNITED STATES OF AMERICA

FOR COOPERATION IN THE U.S. -JAPAN JOINT EARTHQUAKE ' RESEARCH PROGRAM INVOLVING LARGE-SCALE TESTING,

UNDER THE AUSPICES OF THE PANEL ON WIND AND SEISMIC . EFFECTS OF THE U.S. -JAPAN NATURAL RESOURCES

DEVELOPMENT PROGRAM (UJNR) '

;

The Science and Technology Agency (STA) and the Ministry of Construction (MOC) of Japan

and the National Science Foundation (NSF) of the United States agree to cooperate In the

28
implementation of the joint earthquake research program involving large-scale testing (hereinafter referred to as the "Joint Program"), to the extent possible within the existing laws

and regulations of the respective countries, under the conditions as set forth below:

1) The purpose of this Joint Program is to improve scientific knowledge and engineering

practices in the earthquake-resistant design of structures by conducting cooperating earthquake research involving full-scale building structural tests in Japan, and their components

and assembly tests in both countries on the basis of equity and mutual benefit.

2) MOC and NSF agree to serve as the principal points of contact between the Japanese

side and the U.S. side and to provide the required mutual liaison on matters of implementation of the Joint Research Program.

3) STA and MOC, the Japanese Party, and NSF, the U.S. Party, (hereinafter referred to

as "The Parties") agree that the Joint Program shall be implemented with the endorsement of

the UJNR Panel on Wind and Seismic Effects. The Parties may establish, through separate

exchange of letters, whenever necessary, appropriate bilateral implementation procedures.

4) The Parties agree that the research tasks to be supported under this Implementing

Arrangement shall include: (1) tests of full-scale reinforced concrete building structures

and their components and assemblies; and (2) study of pseudo-dynamic test procedures.

5) To assure the success of this Joint Program, the Parties agree to provide access

to the Program by all U.S. and Japanese research institutions to the maximum extent possible,

in accordance with their standard procedures and with the endorsem.ent of the UJNR Panel on

Wind and Seismic Effects. The participating institutions may include, but are not limited

to, in Japan, the National Research Center for Disaster Prevention, the Building Research

Institute, the Public Works Research Institute, and universities; in the United States, the

National Bureau of Standards, U.S. universities, industrial organizations and other governmental or private laboratories or agencies. Participant institutions shall make their facilities, including the Tsukuba Test Facility of MOC, available for use in the Joint Program

where such availability and the terms of use are approved by an appropriate authority.

6) The Parties shall make every effort to secure funds necessary to implement the

Joint Program. Any obligations under this Implementing Arrangement are subject to the

availability of funds to the Parties. The Parties retain the executive and administrative

authority and responsibilities for activities under this Joint Program in accordance with

the agencies' established policies and normal procedures. The Parties shall also consult

29
with each other to make necessary adjustments of annual implementation plans in accordance

with the laws and procedures of both countries. It is anticipated that the parties shall

share support for the total program on an approximately equal basis. '

' '

'

7) The Parties agree that the results of the test program shall be made available to

the UJNR Secretariat of each country and to the scientific community through the normal

publication channels in each country and in accordance with the normal procedures of each

agency

.

8) The Parties shall, as mutually agreed, conduct as annual project review, including

an assessment of scientific and technical achievements for the past year and plans for the

coming year. Additional reviews may be held at the request of either party to address the

terms and conditions of this Implementing Arrangement and other matters of mutual interest.

9) Details which are not prescribed in this Implementing Arrangement shall be decided

between the Parties. "''

10) Modifications of this Implementing Arrangement may be made by consultation between

the Parties.

11) This Implementing Arrangement will take effect on August 10, 1979, and may be terminated by either party giving the other party advance written notice of at least six month

Such termination shall not affect projects underway at the time of termination. Unless so

terminated, this Implementing Arrangement shall be for a term of two years.

FOR JAPAN: FOR THE UNITED STATES OF AMERICA: ''''

Nobuo Kozu

Deputy Director-General

Research Coordination Bureau

Science and Technology Agency

Jack T. Sanderson

Assistant Director for

Engineering and Applied Science

National Science Foundation

J

Hidenobu Takahide

Counsellor for Engineering Affairs

Minister's Secretariat

Ministry of Construction

August 10, 1979

Tokyo, Japan

30
APPENDIX II

A. First Planning Group Meeting

The first planning group meeting, U.S. -Japan Cooperative Research Program Utilizing

Large-Scale Testing Facilities, was held in Tokyo during the period September 5-10, 1977, to

discuss future research programs in earthquake engineering utilizing large-scale testing

facilities

.

During this meeting, the individuals from academic and government institutions and from

private industrial groups exchanged information and views on the following topics:

1) Locations and performance characteristics of existing medium- and large-scale

testing facilities in both countries with example tests illustrating capabilities.

2) Locations and performance characteristics of medium- and large-scale testing facilities under construction or planned for future construction in both countries with planned

programs

.

3) Possible large-scale tests which might be conducted under the recommended cooperative research program including estimates of efforts involved.

4) Implementation of the recommended cooperative research program and level of effort

possible considering funding procedures, restrictions, and limitations in both countries, and

5) Selection of large-scale tests and supporting smaller-scale tests to be conducted

under the recommended cooperative research program.

At the conclusion of the first planning group meeting, the following resolutions were

adopted:

1) The first major test to be carried out under the recommended cooperative research

program should be conducted on a full-size multi-story building to determine its seismic

behavior utilizing the Large-Size Structures Laboratory, Building Research Institute, Tsukuba

New Town for Research and Education. It is intended that the number of stories on the test

structure be the maximum permitted by the facility (8-10 stories) unless the type of

construction or building practice requires a lower number of stories.

2) Six different types of buildings representative of good current practice should be

considered for possible testing, namely, a reinforced concrete building, precast and/or prestressed concrete building, a steel frame building, a steel-reinforced concrete building, a

masonry building, and a wooden building.

31
3) A preliminary design of each building type should be made prior to the second

planning group meeting so that a choice can be made at that time as to which type will be

tested first.

4) The building to be tested first should include non-structural elements such as

curtain walls, partition walls, piping, etc.

5) The building should be tested by a pseudo-dynamic procedure so as to develop and

define realistic seismic behavior.

6) After initial testing, the building should be repaired and retested to assess the

effectiveness of repair procedures.

7) Complementary tests, such as component and model tests, should be conducted in

various institutions prior to and/or parallel with each major full-scale test.

8) Preliminary long-range plans should be made for a possible full-scale test on the

same building type selected for the first major test utilizing the super large-scale twodimensional shaking table.

9) Long-range plans should include the possibility of testing large-scale structures

other than buildings and experiments to study the effect of wind on buildings and civil

engineering structures.

10) Concerned government agencies in both countries are urged to complete arrangements

for implementing the preceding resolutions. It is recommended that the U.S. -Japan Cooperative Science Program and the U.S. -Japan Panel on Wind and Seismic Effects UJNR Program,

coordinate their efforts towards implementing the large-scale testing program.

11) The second planning group meeting, U.S. -Japan Cooperative Research Program Utilizing

Large-Scale Testing Facilities, should be held during the period May 15-19, 1978, in San

Franscisco, California, USA.

B. Second Planning Group Meeting ">

^

,

The second planning group meeting, U.S. -Japan Cooperative Research Program Utilizing

Large-Scale Testing Facilities, was held in San Francisco during the period May 15-19, 1978,

to continue planning for future research programs in earthquake engineering utilizing largescale testing facilities. r . • ,

After opening remarks by J. Penzien and M. Watabe, preliminary design considerations

of full-scale and support tests were presented by H. Aoyama and G. Corley, A. Mattock and

32
T. Okada, R. Hanson and B. Kato, C. Culver and S. Nakata, and A. Mattock, and V. Bertero and

M. Ozake for building Types 1-6 (No. 1 - Reinforced Concrete, No. 2 - Precast/prestressed

Concrete, No. 3 - Structural Steel, No. 4 - Masonry, No. 5 - Timber, No. 6 - Mixed Steel/

Reinforced Concrete), respectively. Extensive discussion toward developing final designs

and testing procedures and toward prioritizing structural types followed these presentations.

Considering such factors as (1) need based on understanding seismic performance of current

construction, (2) need based on developing improved seismic performance in future construction, (3) economic benefits, (4) safety considerations, (5) test feasibility, (6) mutual

benefit to both countries, and (7) cost and manpower advantages of conducting full-scale

tests on a cooperative basis, it was decided that the immediate future planning should

recognize the following priority listing of structural types: No. 1 - Reinforced Concrete,

No. 2 - Structural Steel, No. 3 - Precast/prestressed Concrete, No. 4 - Mixed Steel/

Reinforced Concrete, No. 5 - Masonry, and No. 6 - Timber. It was agreed however that

detailed planning of the full-scale tests for both the first and second priorities, namely,

reinforced concrete and structural steel, should be completed and that plans for component

testing involving these and other types should be developed as deemed appropriate. A geaeral

flow chart of the first full-scale test project was developed as shown in Figure 1.

K. Otani and M. Watabe reported on the "Current Status of BRI Facilities" and "Design

Earthquake Intensity," respectively, and S. C. Liu and T. Okubo discussed plans for implementing the program under the auspices of the U.S. -Japan Panel on Wind and Seismic Effects,

UJNR Program. It was proposed that the overall program be implemented in accordance with

the flow chart shown in Figure 2, that the UJNR Panel on Wind and Seismic Effects be used as

the implementation and coordination mechanisms for the joint program, that the funding

agencies have administrative management responsibility for the joint program, and that the

program utilize scientists and engineers from academic, government, and industrial organizations in both countries as appropriate to the conduct of specific projects.

At the conclusion of the second planning group meeting, the following resolutions were

adopted:

1) Due to safety and economic consideration, it is urgent that full-scale pseudodynamic tests be conducted on various building types for the purpose of verifying and

improving seismic performance.

33
2) In the interest of improved efficiency and research productivity, it is highly

desirable that full-scale pseudo-dynamic tests on buildings be conducted by Japan and the

United States on a cooperative basis.

3) Immediate plans for conducting full-scale pseudo-dynamic tests on a cooperative

basis should be developed in accordance with the following priority listing of building

types: No. 1 - Reinforced Concrete, No. 2 - Structural Steel, No. 3 - Prestressed/Precast

Concrete, No. 4 - Mixed Steel/Reinforced Concrete, No. 5 - Masonry, and No. 6 - Timber.

This listing should remain flexible, however, to accommodate the availability of test

facilities and equipment and to meet other special conditions which may arise prior to and

at the time of testing.

4) Complete detailed structural designs of the reinforced concrete and structural

steel buildings should be prepared for presentation at the third planning group meeting.

The designs of support tests for both types should also be presented at this meeting.

5) Because of their great importance, extensive full-scale tests should be conducted

on non-structural members together with the structural full-scale tests. Plans for these

tests should be developed prior to and be presented at the third planning group meeting.

6) A subgroup should be established to develop procedures and techniques for laboratory simulation of seismic excitations.

7) The implementing agencies of both governments should proceed immediately to request

funding for the procurement of special test equipment (e.g. hydraulic actuators) and instrumentation still required by the overall proposed test program.

8) Due to the importance of a balanced exchange of research personnel prior to and

during the testing program, every effort should be made on both sides to bring about such

an exchange . •

'

9) Administrative and management arrangements and guidelines for implementing the above

resolutions should be completed by the government agencies involved as soon as conveniently

possible

.

10) The third planning group meeting, U.S. -Japan Cooperative Research Program Utilizing

Large-scale Testing Facilities, should be held during the period December 18-23, 1978, in

Japan

.

34
C . Third Planning Group Meeting , ,

.

The third planning group meeting of the U.S. -Japan Cooperative Research Program

Utilizing Large-Scale Testing Facilities was held in Tokyo during the period December 18-23,

1978. The meeting continued discussions to develop future research programs in earthquake

engineering utilizing large-scale testing facilities.

The summary and resolutions of the first and the second planning group meetings held in

Tokyo in September, 1977, and in San Francisco, in May, 1978, were presented and confirmed.

During the third planning group meeting, representatives from academic institutions,

governmental agencies and private industrial firms exchanged information and views. Preliminary design considerations for full-scale tests and their supporting tests were presented by

representatives from both U.S. and Japan for four structural types;

No. 1 Reinforced Concrete ,. .

No. 2 Structural Steel

No. 3 Precast/Prestressed Concrete

No. 4 Mixed Steel/Reinforced Concrete

Preliminary discussion were also presented on masonry and timber types.

For the first two of these structural types, U.S. and Japan representatives discussed in

depth and reached certain conclusions on experiment designs. In addition to working sessions

on structural design, a session on loading procedures for full-scale tests was held. A major

discussion in the loading procedures session centered on the feasibility of the pseudodynamic test method in which sufficient numbers of electro-hydraulic jacks are used.

At the conclusion of the Third Planning Group Meeting in Tokyo, the following resolutions were adopted:

1) The goal of the joint program is to improve seismic safety practices through studies

to determine the relationship among full-scale tests, small-scale and component tests, and

analytical studies. ; .. .

2) The joint program shall be designed and conducted to:

a) achieve clearly stated scientific objectives; .

b) represent total building systems as realistically as possible;

c) balance the simplicity and economy of test specimens with the need to test

structures representing real situations;

d) maintain a balance among small-scale, component, and full-scale tests;

35
e) utilize previously performed experiments and studies to the extent

practical;

f) represent the best design and construction practice in use in both

countries

;

g) check the validity of newly developed earthquake-resistant design

procedure s

;

h) maintain flexibility to accommodate new knowledge and conditions as

successive experiments are completed; and

i) assure the practicality of program results.

3) This program should be initiated in 1979 jointly and cooperatively in both the U.S.

and Japan.

4) To implement this program, the establishment of the following committees and working

subpanels is recommended for inclusion in the governmental MEMORANDUM OF UNDERSTANDING:

a) Joint Executive Committee for the purpose of providing scientific advice

to participating institutions in this program and to appoint subpanels other than

stated below to perform tasks as agreed necessary;

b) subpanel for execution of the full-scale and supporting tests for each

structural type; and

c) subpanel for assessing the feasibility and validity for use in this

program of pseudo-dynamic loading techniques.

5) To implement this joint program, quick and positive response by both governments as

to funding and staff arrangements is requested. Strong emphasis is placed on funding the

loading systems needed to assure adequacy of the facilities to perform the planned experiment s

.

6) The planned order of testing is first the reinforced concrete structures, and second

the steel structure. Precast-prestressed concrete structure and mixed steel-reinforced concrete structure are the next priorities. Masonry and timber structures should be studied

further for inclusion in this program.

7) Additional tests and analyses found to be required beyond the planned program should

be conducted to assure that research results can be applied in the practical design of buildings.

36
8) All activities of the joint program (full-scale tests, support tests, analytical

studies, etc.) should be conducted cooperatively with balanced participation from both

countries to the extent possible.

9) The fourth planning group meeting, U.S. -Japan cooperative research program utilizing

large-scale testing facilities, should be held in the U.S. during the period July 9-14, 1979.

D. Fourth Planning Group Meeting

At the conclusion of the fourth planning group meeting in Berkeley, the Resolutions of

the first three meetings were re-confirmed and additional resolutions were adopted as

follows:

1) In view of the importance of improving seismic safety practices through studies to

determine the relationship among full-scale, small-scale, component tests, and analytical

studies, the final report of the planning group should be completed and be transmitted to the

appropriate agencies of government as soon as possible2) It is recommended that the program of cooperative research as set forth in the

planning group final report be initiated in 1979 in both countries.

3) The level of funding provided to the cooperative research program should permit the

fulfillment of goals and objectives as defined in the planning group final report.

4) To insure successful execution of the recommended cooperative research program,

full coordination of all research activities carried out in both countries is essential;

therefore, the following committees are recommended:

a) Joint Technical Coordination Committee to provide scientific and technical

advice to participating institutions in the program. This committee should meet

at least once a year during the program.

b) Joint subcommittees to provide advice on the execution of research related

to reinforced concrete buildings, steel buildings, pseudo-dynamic loading techniques, and other major areas of activity. These committees should meet as frequently as needed.

5) To assure and enhance the cooperative effort in this joint research program,

adequate exchange of research personnel from both countries is needed. It will be desirable

for one exchange researcher to be associated with each participating research organization in

each country.

37
6) Following successful completion of the recommended cooperative research programs

related to reinforced concrete and steel buildings, similar investigations should be carried

out on other structural types as suggested in previous resolutions taking full advantage of

the experience gained in the use of the pseudo-dynamic testing procedure. Appropriate

planning for this activity should be initiated at an early date.

7) Long-range plans should include the possibility of testing large-scale structures

and systems, other than buildings, in this cooperative program. '

8) The planning group encourages support of the recommended cooperative large-scale

test program by the UJNR Panel on Wind and Seismic Effects. : c'>.

APPENDIX III

I. RECOMMENDED RESEARCH ON REINFORCED CONCRETE BUILDING STRUCTURES

It is recommended that a full-scale seven-story reinforced concrete building structure

representing good current practice be tested in the Large-Size Structures Laboratory, Building Research Institute, Tsukuba New Town for Research and Education, Japan. This test structure should have a layout as indicated in the plan and elevation views of Figures 3 and 4,

respectively, and it should include non-structural elements. The tests should be conducted

using a procedure intended to simulate dynamic response to prescribed seismic excitations.

A pseudo-dynamic loading procedure based on actuator on-line system should be considered for

this purpose. As indicated in Figure 3, the structure should be loaded only in the direction

of the shear wall.

After an initial series of tests limiting the response to moderate damage levels, the

full-scale structure should be repaired and then retested under severe simulated seismic

conditions to near collapse conditions. The objectives of this final test are to determine

large deformation and failure characteristics and to assess the effectiveness of repair

procedures

.

It is recommended that a series of coordinated experiments associated with the fullscale tests be conducted in Japan and the United States on the following test specimens:

(1) element assemblies, (2) structural walls, (3) planar frames, (4) wall-frame assemblies,

and (5) scaled models of the full-scale seven-story structure. The purpose of these tests

38
are to investigate the range of validity of each type in interpreting the seismic response

of full-scale buildings. Both quasi-static and dynamic loadings should be used in conducting

these experiments.

The correlation of the following three groups of experimental parameters are considered

important: „.

1) Methods of loading such as dynamic (earthquake-simulation and harmonic) and static

(pre-programmed reversals and pseudo-dynamic).

2) Scales ranging from smallest practical (around 1/10) to that which uses regular

concrete and standard reinforcing bars (say, 1/3 scale to full scale).

3) Complexity of test assemblies required to give reliable information on the

performance of full-size structures.

It is suggested that the test specimens include: (1) two- and three-dimensional models of

the full-scale structure; (2) two- and three-dimensional simple element assemblies; and (3)

simple and complex element assemblies including walls and frames. In conducting this series

of tests, the importance of including non-structural elements should be determined.

The results of all associated tests in both countries should be fully correlated with

each other and with the results of the tests made in Japan on the full-scale seven-story

structure. The objectives of these correlation studies will be to (a) identify the relative

merits of static loading and dynamic loading and of component tests, small-scale tests, and

medium scale tests in predicting prototype performance under moderate to severe seismic

conditions, and (b) to improve structural details for earthquake resistant design.

It should be noted that the overall investigation recommended represents the first

attempt anywhere in the world at carrying out a fully integrated test program having both

of the above objectives. It, therefore, should prove of great value in setting long-range

priorities for future test programs based on benefit-cost considerations.

Prior to testing the full-scale structure, a thorough analysis of its seismic performance should be carried out in both Japan and the United States. The best available mathematical modelling and computational procedures should be used. In developing the mathematical

model of the structure, use should be made of existing component test data in both countries

supplemented by the results of the associated test previously described. Through these analyses, seismic excitations to be used in the full-scale tests as well as in the associated

tests can be selected.

39
Researchers from both Japan and the United States should participate in the full-scale

tests, the associated tests, and the correlation studies.

II. RECOMMENDED RESEARCH ON STEEL BUILDING STRUCTURES

There is a recognized need to establish the relationship among full-scale tests,

reduced-scale and component tests, and analytical studies of steel building structures

designed to resist earthquake ground motions. In recognition of this need, it is recommended

that a full-scale seven-story steel building, a series of components and assemblages, and

several scaled models of the building be tested. The full-scale structure should be designed

by the latest U.S. and Japanese codes and it should be tested in the Large-Size Structures

Laboratory of the Japanese Building Research Institute in Tsukuba New Town. The test structure should be representative of an office building and the structural system should consist

of moment-resisting exterior frames with a braced interior core. Plan and elevation views

of such a building are shown in Figures 5 and 6, respectively. The interior core should be

braced by either concentric or eccentric K braces. It is suggested that the floor system be

a metal deck with composite floor slab. Non-structural elements such as curtain walls and

partitions should be added to the structural system at some point in the testing program.

The structure should be tested in a manner which simulates realistic earthquake conditions.

The recommended test program for the full-scale building should include free vibration

tests and simulated seismic loading tests at three intensity levels, namely working stress

level, post-buckling of braces level, and post-moment frame yielding level. In order to

maximize the amount of knowledge that may be gained, a multi-stage test program including

bare frames, frames with interior braces, and frames with non-structural elements should be

considered. '" '

As part of the overall investigation, it is recommended that a series of complementary

quasi-static tests be conducted on (a) structural components including girder-to-column connections, column-to-footing connections, individual braces, and floor slabs, (b) beam-andcolumn assemblages, both two- and three-dimensional, and (c) braced and unbraced framed bents

with and without slab interconnections. Further, it is recommended that small- and mediumscale models of the prototype structure be tested under simulated earthquake ground motions

on shaking tables and with the pseudo-dynamic test procedures. Some of the test specimens

may be full-size while others may be half- or third-scale. These associated tests, which are

40
to be carried out by researchers in both countries, should be carefully planned and be made

fully complementary. The results of these tests should greatly assist in predicting the

response of the prototype test structure at all levels of load application. The various

quasi-static, pseudo-dynamic tests and shaking table tests should be correlated with each

other and with the prototype structure tests.

The entire program should be well coordinated and researchers from both the U.S. and

Japan should cooperate fully during the course of the investigation. It is recommended that

researchers from both countries work together in each others' laboratories to the extent

possible and that U.S. investigators take part in the testing of full-scale building in

Japan

.

III. RECOMMENDED RESEARCH ON THE PSEUDO-DYNAMIC TEST METHOD

The most realistic method for assessing the seismic behavior of structures is to test

them using earthquake simulators. However, many structures of interest are either too large,

strong or massive to be tested on available shaking tables. Two approaches have been suggested for quasi-static testing of such structures. One approach assumes a first mode shape,

such as an inverted triangle, and lateral forces are applied in a cyclic manner with the

forces at each floor level being proportional to their first mode contribution. Since the

actual conditions depend on the type of structure, the size of structure and the selected

input earthquake accelerogram, the selected test sequences must be idealized. A second, more

sophisticated approach assumes the force-deformation characteristics of the structure and a

nonlinear dynamic analysis is performed for a given input earthquake accelerogram to determine the response history of the structure. These computed response histories are used as

input data to control displacements, for the subsequent tests of the structure. The major

shortcoming of these two types of quasi-static test methods is that the nonlinear forcedeformation characteristics of the structure must be known or assumed prior to starting the

test.

The computer-actuator on-line (pseudo-dynamic) test procedure has been suggested as the

most realistic method for simulating the dynamic earthquake responses of test specimens which

are too large for existing shaking tables. In the on-line test procedure, the computer uses

the preceding test data to select the next step in the test sequence. Thus, the test data

are used in controlling the progression of the test in much the same way that current

41
building damage predicates the response of that building during the remaining portion of the

earthquake. A significant advantage of the on-line procedure is that it is possible to pause

during the test to record visual observations as well as to recalibrate or relocate instrumentation. .

^

:

Since about 1973, various investigators at the University of Tokyo, in particular, the

Institute of Industrial Science, have been developing and evaluating algorithms for on-line

test procedures. Similar efforts in the United States have not maintained the same continuity of effort.

At the Third Planning Group Meeting of the U.S. -Japan Cooperative Research Program

Utilizing Large-Scale Testing Facilities, concerns were expressed on the capability of controlling up to eight degrees-of-f reedom simultaneously and also on the required accuracy of

the prescribed displacements and measured force outputs necessary to maintain accurate nonlinear dynamic responses of the building. Based on these discussions, there are three

distinct areas which need additional research and/or verification efforts before the on-line

procedure can be used with full confidence. The first of these three areas is verification

and sensitivity studies of the numerical algorithm used to connect the dynamic equations of

motion and the test specimen for the computer control. The second major aspect is the verification that the hydraulic actuator control systems can operate accurately in multiple

displacement control operation with velocity control of the actuator movement during simultaneous motion at each of the load points in order to eliminate inaccurate structure/building

system feedback to the measured forces. This verification would require a physical test in

which a specimen with force-deformation characteristics similar to those expected in the

full-size test structure are used. The third aspect which needs verification is that the

basic differential equation of motion is an adequate representation of the actual dynamic

response of the building system. This could be verified only by comparing on-line static

test results with shaking-table test results. ,i

Analytical studies and experimental verification tests should be carried out in Japan

and the U.S. in a coordinated effort so that the pseudo-dynamic test method can be used

with confidence for the full-scale testing.

42
[Reject and Proportion Full-Scale Test Structure

Dynami

c

Analysis

Shaking Tab le Reduced-Scale Tests] | Component Tests |

, i -

r

I Construction of Full-Scale Test Structure

U.S.A. ^

First Test Series

Repair

i Second Test Series

Recommendations I Japan {

Fig. 1. Flow Chart of First Full-Scale Test Project.

Japan <- 'U.S

PARTICIPATION FROM GOVERNMENT AGENCIES: MINISTRY OF

CONSTRUCTION, STA

FUNDS

STA GOVERNMENT AGENCY

ACADEMIC

RESEARCHERS

INDUSTRY

RESEARCHERS

UJNR WIND

AND SEISMIC

EFFECTS PANEL

PARTICIPATION FROM

^GOVERNMENT AGENCIES

NBS, NSF, DOT, VA,..

UNDER AUSPICES

LARGE-SCALE

TESTING GROUP

I

GOVERNMENT RESEARCHERS^

INDUSTRY RESEARCHERS

ACADEMIC RESEARCHERS

FUNDS

NSF

PROJECT #1

Fig. 2. Flow Chart of Implementation Plan.

43
EARTHQUAKE FORCES

DESIGN CRITERIA

1. DESIGN BY 1976 UNIFORM

BUILDING CODE

2. CONCRETE STRENGTHSTONE AGGREGATE

A 20.7MPA-SLABS AND BEAMS

B. 276MPA- COLUMNS _ AND WALLS

500 MM X 500 MM TYP.

3. REINFORCING STEEL ASTM A6I5

|T 300 X 500 MM BM T T 300 x 500 MM BM

' II J_l

250 MM

nT 300 X 500 MM BM

I I

I

3^ 6M OMM

Fig. 3 Plan of Reinforced Concrete Test Structure

44
DIRECTION QF LOADING

A X- .1—1. •H- -Xin

id

B X- HBRACE

-H- -T

C X6M

o

<

enm

H2

BRACE

o

<

G3'

] Ji

<

D!R£C^O^i CF

DECKING

R6M 6M

id

_-r 1

4- i

id

Fig. 4 Seven-story Office Building 18 m X 16.5 m X 22.5 m

45

MODELING THE 1978 TOKYO TORNADO

THAT OVERTURNED THE TOZAI SUBWAY TRAIN

Eiji Uchida

Shigemi Fujiwhara

Meteorological Research Institute

Ryozo Tatehira

Japan Meteorological Agency

Ichiro Tabe

Kazuyuki Ohtsuka '

Teito Rapid Transit Authority

ABSTRACT

A high wind developed in the evening of February 28, 1978, over areas from Kawasaki City

(Kanagawa Prefecture) to Kamagaya City (Chiba Prefecture). The wind caused extensive damage

(for example, overturning the Tozai subway train, houses, cars, vessels etc.)

Judging from weather conditions which were detailed by data from weather maps, automatic

records of wind direction and speed, of air pressure, radar echo patterns, AMeDAS data, and

disaster distributions, we envisage that this high wind was associated with a tornado formed

within a tornadic cyclone having 5 ~ 6 x 10 km diameter.

We estimate that this tornado was generated within a coverging and unstable area near a

squall line preceding a cold front, the diameter of the tornado core is estimated as 100 -

200 m (radius 50 ~ 100 m), the maximum wind velocity 60 ~ 80 m/s and the propagation speed

25 m/s.

A numerical simulation experiment was carried out in relation to meteorological elements

of the tornado (core radius, the maximum-wind velocity, propagation speed) under an assumption of a single tornado having a straight path and a uniform velocity. The most adaptable

values of tornado elements, in regard to the kinematical analysis of the overturned train

were located within 75 —100 m in core radius, 70 80 m/s in maximum-wind velocity and

25 — 35 m/s in propagation speed.

These results from the engineering aspect nearly coincide with those of the estimated

values and the reliability of this model was thereby roughly verified.

I

I-l
L. INTRODUCTION

In the evening of February 28, 1978, a high wind developed over the areas from Kawasaki

City in Kanagawa Prefecture to Kamagaya City in Chiba Prefecture. The wind caused extensive

damage (for example, overturning the Tozai subway train at about 21H 34M local time). Since

the disaster occurred in a densely populated area in Tokyo, the overturning of the subway

train was widely reported. It was a spectacular accident, but no one was killed, and the

injured list grew to be no more than twenty-one.

Using the available meteorological data, which consists of weather maps, automatic

records of wind directions and speed, radar, and AMeDAS (Automated Meteorological Data Acquisition System) data, the structure of the high wind was analysed.

This analysis is used in a simple mathematical simulation of the engineering aspects for

the overturning of the train vehicles. ,

2. METEOROLOGICAL DATA AND DAMAGES BY THE HIGH WIND

The subway train was composed of ten vehicles, each 20 m in length (200 m in total

length). The rear two vehicles were overturned close to 21H 34M local time. The last rear

vehicle weights 27.6 ton, and the next to last one weighs 36 ton. The direction of overturning was in the right side (hill side) of the train moving direction, corresponding roughly to

a side of shifting direction of the high wind area regarding as the railway as shown in

Figure 1 and 2.

The total damage area was extended over about 30 km long and 500 ~ 1500 m wide. Within

the area, some extensive damage occurred, including injuries to the 21 people (refer to

Figure 3 and reference [1] (1979).

The weather map on that day showed a developing cyclone moved northeastward cross the

Japan Sea, and after the passage of its warm front over the accident site, a tornado occurred

within the unstable and convergence region of squall line in front of the cold front. The

thunderstorms were active in this region (see Figure 4). AMeDAS data showed that high wind

belt ran through the damage area during the evening as illustrated in Figure 5. Two typical

examples of high wind records were obtained by automatic wind recorders within the disaster

area. One of them is shown in Figure 6. In this figure, an instantaneous high wind and a

1-2
cyclonic wind direction change were recorded. However, as the cyclonic wind direction

changed during a 30-minute period, the scale of wind direction did not correspond to one of

wind speed. However, the change of wind direction endorses the existence of a tornadic

cyclone

.

When these wind records were compared with past high wind data, including tornadoes,

this high wind was viewed to have the typical characteristics of a tornado from the point

of view of an instantaneous high wind, although the maximum pressure drop was recorded as

only about 2 mm within the disaster area (see Figure 7).

On the other hand, Tokyo weather radar detected a hook echo at 21H 50M local time just

after the accident, as shown in Figure 8 schematically. This hook echo is suggested like

to be stemming from the tornadic cyclone, and the existence of tornado itself is also

endorsed.

Fujita (1979) [2] proposed the relationships between downburst, tornado, and hook echo

such as those classified in Figure 9, which includes the probability of tornado generation

about 30% for the occurrence of downburst and about 40% for that of hook echo.

Therefore, the high wind in question was estimated synthetically as a tornado under the

following consideration: the characteristics of wind (the instantaneous high wind, and the

cyclonic change of wind direction suggested as a tornadic cyclone), a hook echo and damage

situations (very narrow regions and damage reports suggesting a tornado flew over). There

was unfortunately no positive, visual observation as to whether the high wind was definitely

a rotating vortex or not because it was night time.

3. STUDIES RELATING TO THE TOKYO TORNADO

Muramatsu (1979) [3] analyzed a -50°C disturbance areas along the squall line by using

CMS (Geostationary Meteorological Satellite) data. This disturbance, its diameter 40 to

50 km, corresponds to the tornadic cyclone mentioned above (see Figure 10).

Fujita (1978) [4] depicted precipitation areas, stream lines, frontal lines, and a

radar echo pattern through mesoscale analysis as shown in Figure 11 and 12. Soma (1978) [5]

performed a detailed analysis about damages and suggested evidence of a tornado from the

damage features of the Ooi wharf, etc. He also proposed a three tornadoes model from time

sequence verification for disaster generations.

Miyazawa, Mitsuta, Shoji, etc. (1979) [1] also presented reports.

1-3
4. MODELING AND RELATING A TORNADO TO THE OVERTURNED VEHICLES

Based on the foregoing observational results, a tornado model concerning the overturned

train vehicles is conventionally proposed. It is to be noted that the model would rather

stress to simulate its engineering characteristics than physical one. In addition, the tornado model may be considered as a complex of several vortexes or a complex of a normal high

wind and a tornado. However, because of the deficit of unique and definite data concerning a

complex of several vortexes as a simple step, a single tornado model of a Rankine vortex [6]

type is assumed, and its moving state is also assumed in straight forward motion with uniform

velocity. This seems to compare reasonably with the observations around the train accident

site, especially in regarding to the overturning of the two vehicles.

In order to define the modeling scale, we adjust the following three elements within an

applicable range for observations of train overturning:

Core radius of the tornado (A)

(radius of maximum wind position in a Rankine vortex)

50 ~ 100 m estimated by the disaster survey.

Maximum wind velocity of the tornado (V^^j^^)

60 - 80 m/s estimated by wind records of Sunamachi drainage office and

No. 10 signal office.

Propagation speed of the tornado (V-^)

25 m/s estimated by damage survey.

The reliability of this model is to be verified by the fact of the overturning of the train.

Using the above three element values with extended ranges. A: 25 ^ 100 m, Vmax: 60 ^

80 m/s, V-j^: 15 ^ 35 m/s, the perpendicular component of wind concerning the train is calculated (as shown in Figure 13).

It took about 7 (± 1) sec until the trains two rear cars were overturned after the

driver operated a break facing with some dangers.

On the other hand, calculated wind distributions are composed of a group of circles

expanding to both sides, and higher wind areas are located in the right hand side of

tornado's trajectory (see Figures 14, 15 and 16).

We can hypothesize the overturning of the vehicles as the following steps in the

Figures:

1-4
1) The first vehicle received a wind-shock when it contacted the 40 m/s circle which

is calculated from the kinematical survey for a train vehicle [1].

2) At that time, the driver applied the brake.

3) Successive relative positions between the train and the tornado were depicted each

second

.

4) Next, the cases in which the two rear cars were fallen into within more than 80 m/s

wind areas were picked up which is the threshold value of overturning of a vehicle from the

kinematical survey [1]

•

The final results are summarized in Table 1. In the table, the optiumum examples are

as follows:

Core radius (A): 75 ~ 100 m.

The maximum wind velocity (Vmax) : 70 ~ 80 m/s.

Propagation speed (V-|^): lb- 35 m/s.

Comparing the observation data, we can verify the tornado model clearly by the overturned

train without entering into a more complex tornado model.

ACKNOWLEDGMENT

The authors are greatly indebted to acknowledge Prof. T. T. Fujita of Chicago University, Professor Mitsuta of Kyoto University, Dr. S. Miyazawa (Chief forecaster of JMA)

,

Mr. T. Muramatsu of JMSC , Professor S. Arakawa of Met. College, Dr. M. Shimada of Kofu Local

Met. Obs., Mr. S. Ooishi of Tokyo District Met. Obs., Mr. H. Tokuue of Obs. Dept. (JMA) for

their valuable suggestions and helpful opinions to our research.

REFERENCES

[1] Report on the Counter-Measure Research for Tozai Subway Train Accident (in Japanese),

Committee of the Counter-Measure Research for Tozai Subway Train Accident, March, 1979,

Teito Rapid Transit Authority. . , ,

[2] Fujita, T. T., 1978: Workbook of Tornadoes and High Winds for Engineering Application,

Satellite and Mesometeorology Research Project No. 165, 17-60.

[3] Muramatsu, T. , 1979: Analysis for Squall-Line and Tornado by Use of QMS, Radar and

AMeDAS System. (in Japanese) Tenki, 26, No. 7, 399-411.

[4] Soma, K. , 1978: On Tornadoes and Similar Phenomena (in Japanese), No. 5 Symposium on Wind-Proof Structures.

[5] Ishizaki, H., Mitsuta, Y. , Kawamura, S., Murata, T., Kimoto, E., and Tahira, M.: Studies of a Tatsumaki at Toyohashi, December 7, 1969, A, pp. 481-500

1-5
Fig. 1 Overturned vehicles viewed from N ishifunabashi side

1-6
Fi,. 3 A schematic sketch of damage d.str.but.on caused by a tornado on Feb. 28, 1978.

Black spot areas show severe damage areas".

1-7
!n'Feb!'l8/l97rr' °^ ^^^^^^"^ ^"'^ accompanying front

1-8
<'fS'0£(:f;iiJi;s,V, 2 )l28ll20B.'f- 3 I fl 4 1J#)

Fig. 5 Maximum wmd distribution during Feb. 28, 20H~Mar.l, 04H(k')cal time) over Kanto district

1-9
I-IO
ROTATING TSM DOWNBURST TSM OTHER TSM TOWERING CU

HOOK ECHO BOW ECHO STRONG ECHO WEAK ECHO

9 Four types of convect.ive mother (Parent, Ed.) cloud which could start tornadoes

J a

10 Movement of squall line and cell

I-ll
Fig. 12 Surface weather map at 22 H 00 M of Feb. 28, 1978 ( local time)*\

1-12
Fig. 14 A calculated locus of train at Vi = 15m/s, A = 75m, Vmax = 80m/s

1-13
Fig. 15 A calculated locus of train at V, = 25 m/s, A = 75 m, Vmax = 80 m/s

Table 1 A comparison of simulation results, A means the best adaptable. B means adaptable.

C means roughly adaptable. D means slightly adaptable. E means a few related. A means no relationship.

Vmax

(m/s) 60 70 80

Vi (m/s) 15 25 15 25 35 15 25 35

25 A A A A A A A A A

A 50 A A A A A A A A A

(m) 75 A A CO A © © ® ®

100 A (g) ® ® ®

1-14
Fig. 16 A calculated locus of train at V, = 35 m/s, A = 75 m, Vmax = 70 m/s

1-15
EXTREME WIND SPEEDS AT 129 STATIONS

IN THE CONTIGUOUS UNITED STATES

Michael J. Changery

National Climatic Center

Asheville, North Carolina

and

Emil Simiu and James J. Filliben

National Bureau of Standards

Washington, D.C.

ABSTRACT

The purpose of this report is to present information on recorded and predicted wind

speeds at 129 airport stations in the contiguous United States at which reliable records are

available over a number of consecutive years. This information is provided to serve as basic

documentation from which appropriate decisions can be made on values of design wind speeds to

be specified in building codes and standards, and on special projects. Included in the

report are: recorded wind speeds and anemometer elevations; predicted wind speeds based on

probability distributions of the largest values; estimates of the sampling errors inherent

in the predicted wind speeds; a description of the statistical procedure used in the analysis of the data; and a discussion of the results of the analysis.

KEYWORDS: Building (codes); probability distribution functions; statistical analysis;

storms; structural engineering; wind pressure; wind speeds.

1-16
1 . INTRODUCTION

The purpose of this report is to present information on extreme wind speeds at 129

airport stations in the contiguous United States at which reliable wind records are available

over a number of consecutive years.

This information consists of:

1.1 Extreme yearly wind speeds, and the corresponding wind directions, recorded at each of

the 129 stations. These data were obtained by the National Climatic Center from the original

records. Thus, reading errors of original records and errors of transcription that have been

determined to be present in Local Climatological Data (LCD) monthly and annual summaries have

been eliminated. The vast majority of the originally recorded data consisted of fastest-mile

speeds. These have been listed without modification in the report. However, at a few

stations, some of the recorded data consisted of fastest observed one-minute speeds. These

have been transformed into fastest-mile speeds using a relation given in Section 2.1. It

is these fastest-mile speeds that have been listed in the report in lieu of the orginally

recorded fastest observed one-minute data. The stations and dates at which fastest-minute

speeds were originally recorded are listed in Section 2.1.

1.2 Anemometer elevations at which the largest yearly wind speeds were recorded.

1.3 Largest yearly wind speeds reduced to an elevation of 10 m above ground (corrected

speeds). These were obtained by using an expression given in Section 2.4.

1.4 Results of the statistical analysis of the corrected wind speed data. These results

include

:

* for each of the 129 sets of data, the predicted wind speeds corresponding to

various return periods, based on the assumption that the Type I probability

distribution of the largest values is a valid description of the extreme

wind speeds

* for those sets of data that are best fit by a Type II probability distribution

of the largest values, the predicted wind speeds corresponding to various

return periods, based on that distribution

1-17
* estimates of the lower bound of the standard deviation of the errors inherent

in the predicted speeds

* estimates (obtained by the method of moments) of the standard deviation of the

errors inherent in the predicted speeds

Extreme wind speed predictions have been included for mean recurrence intervals of up to

1,000,000 years. However, in the writers' opinion, physical considerations suggest that predictions corresponding to mean recurrence intervals beyond a few hundred years should be

regarded with caution.

A brief description of the procedure used in the analysis of the data is presented in

Section 3. Section 3 includes a summary, and Sections 3.2 and 3.3 a discussion of the

results of the statistical analysis. A sample of the information described under items 1

through 4 above is included in Section 4 of the report.

It is noted that at a number of stations the extreme yearly wind speed data may not

provide a reliable basis for predicting extreme speeds. The results of the statistical

analysis for these stations should therefore be regarded with caution. Stations for which

such caution is in order are listed in Appendix I.

2. WIND SPEED DATA

2.1 Fastest Observed One-Minute Speeds

It was indicated in Section 1 that the vast majority of the original data used in this

report consisted of fastest-mile wind speeds, i.e., speeds averaged over a time interval (in

seconds) t = 3600/v£, and Vj = fastest-mile wind speed in miles per hour. However, at the

following stations the original recorded maximum annual wind speed data consisted of fastest

observed one-minute speeds during the periods indicated below:

Atlanta, Georgia (1961 through 1963)

Indianapolis, Indiana (1962-1963)

Boston, Massachusetts (1954 through 1958)

Lansing, Michigan (1955 through 1958)

Sault Ste Marie, Michigan (1956 through 1965)

According to Reference 1, studies of the relationship between fastest observed oneminute to fastest-mile wind speeds undertaken at four weather stations "showed the mean

regression between the two types of observation to be

1-18
Vf = 9.55 + 0.999 v - , . (1) r m

where = fastest-mile speed in miles per hour and v^^ = fastest-minute speed in the same

hour as the fastest-mile, in miles per hour. Since the slope is very near unity and the

mean difference very near 10, it has been assumed for some time that adding 10 mph to the

fastest-minute would give an approximation to the fastest-mile." It is this relation which -

in the absence of other information - has been used in this report.

While the writers are not certain that Eq. 1 provides a correct relation between v^ and

Vjjj, they note that it results in estimates of V£ that are conservative from a structural

safety point of view. ,

, ... . •

2.2 Roughness Conditions at Airport Stations

In an attempt to ensure that the terrain roughness conditions are uniform among all the

sets of data being analyzed, only airport stations have been considered herein. In principle, it may be assumed that at such stations open exposure conditions prevail. Nevertheless, the mere fact that wind speed measurements are taken at an airport station does not

necessarily ensure that the wind c limatological conditions reflected by these measurements

are identical, from the standpoint of the terrain exposure, to those prevailing at a different airport. For example, it is noted in Reference 2 that the estimated 50-year wind at

Chicago Midway Airport is about 15 mph less than at the Chicago O'Hare airport. The probable reason for this difference is that the terrain around the Chicago Midway Airport is

relatively heavily built-up. Similar considerations might explain to some extent the difference between the estimated 50-year winds at the Washington National Airport and the

Baltimore-Washington International Airport, which are estimated in this report to be 66 mph

and 75 mph, respectively. Thus, in interpreting airport data for the purpose of developing

wind maps, it is appropriate to take into account the possibility that, at the airport of

concern, the terrain exposure conditions might differ somewhat from those defined as "open"

(e.g., in Reference 3). ,! .

2.3 Variation of Wind Speed with Height Above Ground

To ensure the micrometeorological homogeneity of the data at any given station it is

necessary to reduce all the wind speeds recorded at that station to a common elevation. The

elevation chosen for this purpose is 10 m above ground.

1-19
The mean wind profile near the ground in homogeneous terrain is given by the well-known

logarithmic law, which may be written in the form:

In z_ U(z) = ^_o_ U(10) (2) In IQ_

where z = height above ground and z^ = roughness length, both expressed in meters. In open

terrain, z^ may vary from, say, 0.03 m to 0.10 m. In this report the reduction of the data

to an elevation of 10 ra is based on the assumption z^ = 0.05 m. It can be verified that the

errors inherent in the assumption z^ = 0.05 m — when in fact the values z^ = 0.03 m or z^ =

0.10 m were correct — are small (of the order of 1% or 2%).

An approximation to Eq. 2 is given by the power law

U(z) = (^) ""^10^ ~ (3)

where, for open terrain conditions, it is generally assumed a = 1/7 (3). It is noted that

Eq . 2, and therefore its approximate equivalent given by Eq. 3, is valid for mean wind speeds

averaged over a relatively long time interval, e.g., one hour. The question thus arises of

expressing the variation with height of the fastest-mile wind speed, which is averaged over a

relatively short time (30 to 90s or so).

To obtain an approximate expression for the fastest-mile wind profile, note that it may

be assumed, approximately,

V -"

2

where Upj^ = peak wind speed, = fastest-mile speed, and U = hourly mean speed (see, e.g..

Reference 4, p. 62). The expression for can, in open terrain, be written as

Upi^(z) = U(z) + 3 u

2 (5)

-17 . . .. where u = r.m.s of longitudinal velocity fluctuations, and

1-20
.1/2 U(10)

/ u

'2 = In 10 (6)

where is expressed in meters (see Reference 4, pp. 45 and 54).

It can be verified by using Equations 2, 4, 5, and 6 that, within the anemometer elevation range of interest in this report, it is possible to write approximately

Uf^(lO) ^ U(10) (1 + z-10 0.02)

Ufni(z) U(z) "ToT ,

-

where z is expressed in meters. The errors inherent in Equation 7 are of the order of -1 to

3%, the higher errors being on the conservative side (i.e., yielding slightly higher fastest

mile values at 10 m above ground than would be obtained by a more "exact" expression). Eq

.

has been employed to obtain the corrected speeds at 10 m ground in this report.

.3. STATISTICAL ANALYSIS

3.1 Objective of Statistical Procedure

Probabilistic considerations, as well as available empirical evidence suggest that the

asymptotic probability distributions of the largest values with unlimited upper tail are an

appropriate model for the behavior of the largest yearly wind speed. There are two such

distributions, known as the Type I and Type II distributions of the largest values, whose

cumulative distributions functions, F-|.(v) and F-|--£(v), respectively, are of the form

F-|-(v) = exp [-exp (- v -y

)

] ;- °o < v < ;

r . a

— <»<|J<oo.Q<(j<oo

and ^ll(^) = ^^P [- (v ~ ^] ; u < V < « ; (8)

a

_ oo < y < oo. 0 < a < Y > 0

'

in which y, o, and Y are location, scale, and tail length parameters, respectively.

Actually, the Type I distribution may be shown to be a Type II distribution with y = oo (see

Reference 4, p. 422); however, it is convenient to refer to it separately.

The data were analyzed using — with minor modifications — a computer program listed

in Reference 5. For convenience, the main features of the procedure used in the analysis of

the data are summarized in this section.

1-21
The procedure consists of three distinct stages. In the first stage the value of y

(Eq. 8) is determined which yields the closest fit to the observed data set (recall that y =

" corresponds to an extreme value Type I distribution). The "closest fit" criterion used in

this stage is the so-called maximum probability plot correlation coefficient criterion. The

probability plot correlation coefficient is defined as

Z(X- - X) [M-(D) - M(D)]

ro = Corr(X,M) = — ^ :- (9)

{ZUi - X)2 z:[Mi(D) - M(D)]2} 1/2

in which X = IX^/n; M(D) = IM^(D)/n; n = sample size; and D = probability distribution

tested. The quantities X^ are obtained by a rearrangement of the data set: X^ is the

smallest, X2 is the second smallest; and X^^ the ith smallest of the observations in the set.

The quantities M^(D) are obtained as follows. Given a random variable X with probability

distribution D and given an integer sample size n, it is possible from probabilistic

considerations to derive mathematically the distributions of the smallest, second smallest,

and generally the ith smallest values of X in a sample of size n. There are various quantities that can be utilized to measure the location of the distribution of the ith smallest

value X^ (e.g., the mean, the median, or the mode). It is convenient to use the median as

a measure of location in Eq. 9 — these medians of the distribution of the ith smallest value

being denoted by M^(D).

If the data set was generated by the distribution D, then aside from a location and

scale factor, X^ will be a approximately equal to M^(D) for all i, and so the plot of X^

versus M^(D) [referred to as probability plot] will be approximately linear. This linearity

will, in turn, result in a near unity value in r^. Thus, the better the fit of the distribution, D, to the data, the closer r^ will be to unity.

The procedure just described makes use of 46 extreme value Type II distributions defined

by various values of y from 1-25 in steps of 1, from 25-50 in steps of 5, from 50-100 in

steps of 10, from 100-500 in steps of 50, from 500-1,000 in steps of 250, and y = For

any given data set, 46 probability plot correlation coefficients are computed corresponding

to these distributions, and the distribution with the maximum probability plot correlation

coefficient is chosen as the one which best fits the data. The final result from this first

stage is a value, YQp,- , of y corresponding to the estimated best fitting distribution.

1-22
The second stage in the procedure consists of estimating the location and scale

parameters, p and a, respectively, in Eqs. 7 and 8 for the observed data set and for the

determined optimal value, Ygp]- > as determined in stage 1. Estimates of the location and

scale follow directly from the basic probability plot approach. If a least-squares line is

fit to the probability plot corresponding to YQp(- , then the computed intercept and slope of

the fitted line serve as estimates for the unknown location and scale parameters, y and a.

In terms of the and Mj^(D) , these estimated location and scale values, m and o, are as

follows:

Z(X- - X)[Mi(D) - M(D)]

^^^^

2:[Mi(D) - M(D)]2

U = X - a M(D) (11)

The third and final stage in the procedure determines the predicted wind speed, v>,, for

various intervals N of interest. The estimate for v^^ is

V„ = M + OGy^ (1-1) ' (12)

1^ ^^opt ^

in which y^^^ = the optimal value of y (as determined in stage 1); u and o are the estimates

of the location and scale parameters, u and a in Eqs. 7 and 8 (as determined in stage 2); and

Gyv (p) ~ the percentage point function of the best fitting extreme value distribution,

^'opt

If Yqpj- ^ °° (i.e., if a member of the extreme value type II family provides the best fit),

then

^XYopt " P^~^^^ •

" ^^^^

If Yqpj- = °° (i.e., if the extreme value type 1 distribution provides the best fit), then

G„„ (p) = -ln(-ln p) (14)

^''opt

In effect, the procedure described in this section is an automated equivalent of probability

paper plotting in which 46 types of probability paper, corresponding to 46 extreme value distributions, would be used and in which fitting would be carried out on the basis of the

least-squares method, rather than by eye.

1-23
3.2 Estimation of Sampling Errors

As indicated in Section 1, the computer output of Section 4 includes estimates of the

standard deviation of the sampling errors, i.e., errors that are a consequence of the limited

size of the data sample from which the Type 1 distribution parameters are estimated. Two

such estimates were used. One estimate is based on the method of moments and has the following expression given by Gumbel in Reference 6 (pp. 10, 174, and 228):

SD(vN) . [jl! + l-1396(i^_0,1722)^ + l.Ky - 0.5772)2]l/2^ ^5)

6 6 -^Cj n

in which SD(vj^) = the (estimated) standard deviation of the sampling error in the estimation

of the N-year wind

y = -In [ -In (1 - j.)] (16)

N

a = the estimated value of the scale parameter; and n = the sample size.

A lower bound for the estimated sampling error is given by the following expression:

SD(.j^(vj^) (0.60739y2 + 0.514y + 1 . 10866) ^ - (^y)

n

where the notations are the same as in Equation 15. Equation 17 is commonly referred to as

the Cramer-Rao lower bound [7]. " '

3.3 Summary of Results

The results of the analyis are summarized in Table 1, in which the following notations

are used

:

n = sample size ,'

"

X = sample mean

s = sample standard deviation

^max ~ sample maximum

^opt ~ value of optimal tail length parameter (see Section 3.1)

= estimated extreme wind corresponding to a n-year return period, based

on Type I distribution '

ppcc = probability plot correlation coefficient (see Section 3.1) for Type I "

distribution

1-24
v^Q = estimated 50-year wind speed . .

SD(v^q) = estimated standard deviation of sampling error for 50-year wind speed

3.4 Type I Versus Type II Distribution

Of the 129 stations listed in Table 1, 15 stations [marked with the superscript (c) in

Table 1 and listed in Appendix 1] have been noted to have largest yearly speed records that

may not provide a reliable basis for predicting extreme winds. The remaining 114 stations

may be divided into three categories characterized by the value of the optimal tail length

parameter Y^p^. , as shown in Table 2.

Table 2 Classification of Stations According to Value of Y^p^

Category Range of Y^pj. Number of Stations Percentage

I 13 1 Yqp^ < " 89 78%

II 7 1 Y^p^ < 13 . ' 11 10%

III 2 1 Yqp^ < 7 14 12%

The sample size for the stations of Table 2 varies between n = 10 and n = 45.

It is noted that the percentage of Table 2 are in qualitative agreement with those found

from the analysis reported in Reference 8, in which all sample sizes were n = 37. This tends

to confirm the hypothesis advanced in Reference 8 to the effect that, for stations in wellbehaved wind climates, the best fit of a Type II (rather than Type I) distribution to a set

of extreme wind data might be attributed to a sampling error in the estimation of the tail

length parameter. This hypothesis does not exclude the possibility that stations exist for

which a Type II distribution might provide an appropriate description of the wind climate;

however, according to the results of both Reference 8 and Table 2, the number of such

stations, if they exist, is very likely to be small. Thus, it appears justified to assume,

as in Reference 8, that the Type I distribution of the largest values provides in general a

better description of the wind climate than Type II distributions with small values of the

tail length parameter (say, 2 ^ Y ^ 12).

3.5 Largest Wind Speed in a Sample of Size N and the N-Year Wind

It is shown in Reference 9 (see also Reference 4, p. 423) that, if a variate X has a

Type I distribution, the mode of the largest value in a sample of n values of X is very

nearly equal to the value of the variate corresponding to the mean return period n (recall

1-25
that the mode of a variate X is the value of that variable most likely to occur in any

given trial). It can be seen from Table 2 that, for most sets for which Y^pj. is large, the

ratio Vjjjax'^^n indeed close to unity.

4. WIND SPEED AND DIRECTION DATA AND COMPUTER OUTPUT (SAMPLE STATION)

BISMARCK, N.D. (1940-1977)

THE SAMPLE NUMBER OF OBSERVATIONS =38.00 - ,

THE SAMPLE MEAN =58.30

THE SAMPLE STANDARD DEVIATION = 5.57

THE SAMPLE MINIMUM =49.26

THE SAMPLE MAXIMUM = 68.94 .

DATE

06/02/40

03/15/41

07/07/42

07/12/43

08/08/44

05/06/45

ANEMOMETER

ELEVATION (FT)

43.

43.

43.

43.

43.

43.

FASTEST MILE WIND SPEED

AND DIRECTION (RECORDED

AT ANEMOMETER ELEVATION)

57. NW

65. NW

66. NW

72. W

72. S 61. NW

CALCULATED FASTEST MILE

WIND SPEED AT 10 M ABOVE

GROUND (CORRECTED SPEED)

55.

62.

63.

69.

69

.

58.

12/08/73

05/01/74

01/11/75

02/02/76

05/28/77

20. 20.

20.

20.

20.

45. NW

45. W

54. NW

47. NW

45. SW

49.

49.

59.

51 . 49.

1-26
RETURN PERIOD PREDICTED EXTREME WIND ESTIMATED STANDARD ESTIMATED STANDARD

(IN YEARS) BASED ON EXTREME VALUE DEVIATION SAMPLING DEVIATION SAMPLING

TYPE I DISTRIBUTION ERROR CRAMER-RAO METHOD OF MOMENT

2.0

3.0

4.0

5.0

57.45

59.84

61 .36 62.49

.83

1.01

1.16

1.27

.83

1 .05 1.24

1.40

10.0 65.83 1.63 1 .89

50.0 73.17 2.48 3.05

100.0 76.27 2.85 3.55

1000.0 86.53 4.09 5.23

10000.0

100000.0

1000000.0

96.77

107.01

117.25

5.34

6.60

7.86

6.92

8.61

10.31

REFERENCES

[1] H. C. S. Thom, "Prediction of Design and Operating Velocities for Large Steerable

Radio Antennas," Large Steerable Radio Antennas - Climatological and Aerodynamic

Considerations, Annuals of the New York Academy of Sciences, Vol. 116, Art. 1, pp. 90-

100, New York, N.Y., 1964.

[2] H. C. S. Thom, "Engineering Climatology of Design Winds, with Special Reference to the

Chicago Areas," Proceedings, Symposium on Wind Effects on High-Rise Buildings, Northwestern University, Evanston, IL. , March 1970.

[3] American National Standard Building Requirements for Minimum Design Loads in Buildings

and Other Structures A58.1, American National Standards Institute, New York, N.Y. , 1972.

[4] E. Simiu and R. H. Scanlan, Wind Effects on Structures: An Introduction to Wind

Engineering, Wi ley-Inter science , New York, N.Y. , 1978.

[5] E. Simiu and J. J. Filliben, Statistical Analysis of Extreme Winds, NBS Technical

Note No. 868, National Bureau of Standards, Washington, D.C. , June 1975.

[6] E. J. Gumbel, Statistics of Extremes, Columbia University Press, New York, N.Y., 1958.

[7] F. Downton, "Linear Estimates of Parameters in the Extreme Value Distribution,"

Technometrics , Vol. 8, No. 1, February 1966, pp. 3-17.

[8] E. Simiu, J. Bietry and J. J. Filliben, "Sampling Errors in Estimation of Extreme

Wind Speeds," Journal of the Structural Division, ASCE, March 1978, pp. 491-501.

[9] E. Simiu and B. R. Ellingwood, "Code Calibration of Extreme Wind Return Periods,"

Journal of the Structural Division, ASCE, March 1977, pp. 725-729.

1-27
Table 1 Summary of Results

I

.

Birningham, Alabama 34 46.6 0.139 62. 3 1.00 .99085 65 4

2

.

Montgomery, Alabama 28 ^ J . J 76. 7 1 20 3 .95241 68 cJ

3

.

Prescott, Arizona 1

7

52.2 0.169 66. 0 0,96 .95262 76 7

/ 4 . Tucson, Arizona 30 51.4 0.167 77, 7 1,09 °° ,97064 75 5

5

.

Yuma, Arizona 29 48.9 0.157 65. 1 0,98 °° .98549 70 5

6

.

Fort Smith, Arkansas 26 46.6 0,150 60. 7 0,98 .95078 65 5

7

.

Little Rock, Arkansas 35 4 6.7 0.206 72.2 1.03 1

3

,98663 73 5

8

.

Fresno, California 3 7 34 .4 0. 140 46. 5 ,99385 4 7

-1J

9

.

Red Bluff, California 33 52.1 0,141 67. 3 0.97 oo .96970 7 2 10. Sacramento, California 29 46.0 0.223 67. 8 0.97 a> .97737 74 6

1 1

.

San Diego, California 38 34 .

5

0.130 46. 6 1 .02 oo .97803 47 2 12. Denver, Colorado 2 7 49.2 0.096 62. 3 1 .02 a. .98755 62 3

1 3

.

Grand Junction, Colorado 3

1

52.7 0.102 69. 9 1.07 1

0

.98060 67 3

lA

.

Pueblo, Colorado 3 7 b2 ,8 0.118 79. 0,98 as .98362 83 4

1 5

.

Hartford, Connecticut 38 4 5.1 0.151 66. 8 1,08 8 .98738 64 4

16. Washington, D.C. 33 48.3 0.135 66. 3 1 ,04 .98466 66 4

1 7

.

Jacksonville, Flori^a^^^

Key West, Florida^^^

Z8 48.6 0.206 74. 4 1.04 1

2

.98068 76 6

18. 1

9

51.0 0.337 89. 5 1 .06 6 .96796 99 13

1 9

.

Tampa, Florida^^' 1

0

4 9.6 0.163 65. 1 1.05 OS .99226 74 9 20

.

Atlanta, Georgia 42 4 7.4 0.195 75. 5 1 .06 70 .996 1 3 73 5

21

.

Macon, Georgia

c . u n , ( c

)

Savannah, Georgia

28 45.0 0. 169 59. 7 0.96 CO .97855 66 5 22

.

32 4 7.6 0.202 79. 3 1.13 6 .97840 74 6 23. Boise, Idaho 38 4 7.8 0.111 61

.

9 1 .01 a. .99 W5 62 3 24. Pocatello, Idaho 39 33.3 0.128 71.6 1 .02 a> .97231 72 4 25. Chicago, Illinois 35 4 7.0 0. 102 58

.

b 1 .00 ,98819 60 3 26

.

Moline , Illinois 34 54 .

8

0.141 72. 1 0.98 ,98027 76 4

27

,

Peoria, Illinois 35 52.0 0.134 70

.

2 1 .02 350 .99462 71 , 4 28 Sprlngfieid, Illinois 30 54 2 0 111 70

.

6 9 .98082 70

29 . Evansville, Indiana 37 46.7 0. 1 30 61. 3 1 .00 96240 63 3

30. Fort Wayne, Indiana 36 53.0 0. 125 69

.

0 1 .00 98389 7

1

4 31

.

Indianapolis, Indiana 34 55.4 0 .200 9 3 . 0 1.12 4 97030 86 6 32

.

Burlington, Iowa 23 56.

U

0.164 71. 9 0,95 96831 8

1

6 33. De s Moines, Iowa 27 57.7 0.147 79

.

9 1 .04 98917 8

1

5 3A . Sioux Citv, Iowa 36 57.9 0.157 88

.

1 1.10 1 [ 9 86b0 83 5 35

.

Concordia Kansas 16 57.6 0.160 73

.

7 0 97 9669 3 85 g 36. Dodge City, Kansas 35 60 .

6

0,099 7 1

.

5 0.95 968^3 77 3

37 . Topeka, Kansas 28 54.5 0.150 78. 8 1 .08 7 97 76 7 7 7 5 38

.

Wichita Kansas 37 58.

1

0.146 89 1.13 y 97908 80 - . . 5 39. Louisville, Kentucky 32 49 . 3 0.136 65. 7 1.00 ,98870 68 4

AO. Shreveport , Louisiana \ I 44.6 0.121 53. 4 1 00 ,9 7409 60

' A 1 . Portland Maine 37 48.5 0.179 72. 8 1 04 7 7 2 5 A2

.

Baltimore, Maryland 29 55.9 0.123 71. 2 0 99 ,97826 7 5 4 43

.

Boston Mas sac huse t t s^ ^

Nantucket Massachusetts^^^

42 56 .

3

0.172 81

.

4 1 05 1 6 .99598 8 3 5

44. 23 56 .

7

0.141 71. 3 0 9 7 .9 7294 79 45. Detroit, Michigan 44 48 .

9

0 . 1 40 67. 6 1 01 .98953 7 3 46

.

Grand Rapids, Michigan 27 48 .

3

0.209 66.8 0.94 .96946 76 6

4 7 . Lansing, Michigan 29 53.0 0.125 67 .0 0.98 .98360 7 1 4 48

.

Sault Ste Marie, Michigan 37 48.4 0. 159 67 .0 0.99 .98835 69 4 A9 . Duiuth, Minnesota 28 50.9 0.151 69.6 1 .01 .99249 72 5 50

.

Minneapolis, Minnesota 40 49 .2 0.185 81,6 1.14 4 .97068 73 5

5 1

.

Jackson, Mississippi 29 45.9 0.155 64.4 1 .03 40 .98574 66 4 52

.

Columbia, Missouri 28 50.2 0. 129 62.4 0.97 « .94496 67 4 53 . Kansas City, Missouri 44 50.5 0.155 75.2 1 .06 500 .99408 72 4 54

.

St. Louis, Missouri 19 47.4 0. 156 65.7 1 .06 to .97461 68 6 55

.

Springfield, Missouri 37 50.1 0.148 71.2 1 .04 50 .98845 70 4 56

.

Billings, Montana 39 59.4 0.135 84.2 1 .06 oo .99193 81 4 57 . Great Falls, Montana 34 59.0 0.110 74.2 1 .00 .97365 77 4 58

.

Havre , Montana 17 58.0 0.159 77.7 1 .03 40 .97252 84 8 59

.

Helena, Montana 38 55.2 0.118 71.2 1.14 .98203 73 4 60

.

Missoula, Montana 33 48.3 0.122 70.9 1.14 4 .96352 64 3 61. North Platte, Nebraska 29 62.0 0. 108 74.4 0.96 .97417 80 4 62 Omaha , Nebraska 42 5 5.0 0.195 104.

U

1 ,28 3 .94035 83 6 63. Valentine, Nebraska 22 60.6 0.142 74.1 0.95 .95559 84 6 64. Ely, Nevada 39 52.9 0.117 70.1 1.02 .99318 70 3 65. Las Vegas, Nevada 13 54.7 0.128 70.1 1 .05 13 .98687 75 7 66. Reno, Nevada 36 56.5 0.141 76.6 1 .00 .99298 79 4 67. Winnemucca, Nevada 28 50.2 0.142 62.6 0.95 .95996 69 5 68. Concord, New Hampshire 37 42.9 0.195 68.5 1 .08 9 .98926 66 5 69. Albuquerque, New Mexico 45 57.2 0.136 84.3 1 .09 7 .98635 78 4 70. Roswell, New Mexico 31 58.2 0. 153 81.6 1.03 .98116 83 5 71 . Albany, New York 40 47.9 0.140 68.5 1 ,06 6 .97576 66 4 72. Bingharaton, New York 27 49.2 0.130 63.8 1 .00 ,99016 67 4 73. Buffalo, New York

New York, New York^"^'

34 53.9 0.132 78.6 1 .U 8 ,97908 73 4 74. 31 50.3 0.143 61.4 0,93 ,9300 69 4 75. Rochester, New York 37 53.5 0.097 65,4 0.99 ,98338 68 3 76. Syracuse, New York

Cape Hatteras, N. Carolina^"^'

37 50.3 0.121 67.2 1,03 .98827 67 3 77. 45 58.0 0.214 103.0 1 . 14 6 .98108 91 5 78. Charlotte, N. Carolina 27 44.7 0. 168

I- 64.6 -28

1 .05 .97374 65 5
Table 1 Summary of Results (Continued)

79. Greensboro, N. Carolina 48

80. WiluingCon, N. Carolina-'^' 26 81. Bismarck, North Dakota 38 82. Fargo, North Dakota 36 83. Williston, North Dakota 16 84. Cleveland, Ohio 35 85. Columbus, Ohio 26 86. Dayton, Ohio ' 35 87. Toledo, Ohio 35 88. Oklahoma City, Oklahoma 26 89. Tulsa, Oklahoma 35 90. Portland, Oregon 28 91. Roseburg, Oregon 12 92. Harrisburg, Pennsylvania 39 93. Philadelphia, Pennsylvania^'^' 23 94. Pittsburgh, Pennsylvania 18 95. Scranton, Pennsylvania 23 96. Block Island, Rhode Island^^' 31 97. Greenville, South Carolina 36 98. Huron, South Dakota 39

99. Rapid City, South Dakota 36 100. Chattanooga, Tennessee 35

101. Knoxville, Tennessee 33 102. Memphis, Tennessee 21 103. Nashville, Tennessee 34 104. Abilene, Texas 34 105. Amarillo, Texas 34 106. Austin, Texas 35 107. Brownsville, Texas'*^' 35 108. Corpus Christi, Texas'*^' 34 109. Dallas, Texas 32 no. El Paso, Texas 32 111. Port Arthur, Texas^"^' 25 112. San Antonio, Texas 36 113. Salt Lake City, Utah 36 114. Burlington, Vermont 34 115. Lynchburg, Virgina 34 116. Norfolk, Virginia^'^^ 20

117. Richmond, Virginia 27 118. North Head, Washington 41 119. Quillayute, Washington 11 120. Seattle, Washington 10 121. Spokane, Washington 37 122. Tatoosh Island, Washington 54 123. Green Bay, Wisconsin 29 124. Madison, Wisconsin 31 125. Milwaukee, Wisconsin 37 126. Cheyenne, Wyoming 42 127. Lander, Wyoming 32 128. Sheridan, Wyoming 37 129. Elkins, West Virginia 10

42.3 0.,180 66..8 1

,

.07 49.9 0.,218 84.,3 1 .14 58.3 0.,096 68.,9 0..96 59.4 0..185 100,.5 1 .17 56.5 0.,117 69..3 1..00 52.7 0.,125 68..5 1,.00 49.4 0.,133 61

.

.3 0,.96 53.6 0.,142 72..0 1

,

.00 50.8 0,.177 82,.2 1 .13 54.0 0.,110 69..3 1

,

.03 47 .9 0..145 68,.3 1

,

.05 52.6 0.,196 87..9 1

,

.16 35.6 0,.169 51

,

.1 1 .14 45.7 0. 164 64..4 1

,

.00 49.5 0. 115 62.,4 1

.

.00 48.4 0. 120 59..6 1..00 44.6 0. 107 54..2 0..99 61.4 0.,142 86..2 1..06 48.5 0 .226 71 .9 0 .95

61 .4 0,.132 78 .8 0 .96

61 .0 0 .087 70 .5 0 .96 47.8 0 .218 75 .9 1 .04 48.8 0,.141 65 .9 1 .01 45.4 0 .137 60 .7 1 .04 46.8 0 .171 70 .2 1 .06 54.7 0 .192 99 .9 1 ,27

bl .0 0 .117 80 .7 1 ,03 45.1 0 .122 58 .0 1 .00 43.7 0 .185 6b .1 1 .04 54.5 0 .288 127 .8 1 .45 49.1 0 .132 66 .8 1 .03 55.4 0 .087 66 .7 J ,00 53.1 0,.181 81 .0 1 ,09 47 .0 0 .183 79 .5 1 ,14 50.6 0,.142 69 .0 1 .00 45.7 0 .160 66 .5 1 .05 40.9 0..149 53,.4 0,.96 48.9 0,.182 68,.9 1,.03 42 .2 0 .152 61 .3 1 .08 71,5 0,.141 104,, 4 1,.09 36,5 0,.085 41,.9 1,.01 41,9 0..080 49 .3 1,.00 47,8 0..133 64,,6 1

,

.01 66,0 0..106 85,.6 .99 56,6 u

,

.212 103,.0 1 .24 55,7 0..190 80..2 1..00 53,7 0,.121 67..9 0,.97 60,5 0,.093 72..6 0..98 61,2 u..160 80. 0..96 61,5 0..116 82..0 1..04 51,1 u

,

.160 68..5 1

,

.08

6 .98239 63 4

4 .97311 80 7 .96635 73 3

5 .97527 89 6 .98686 75 6 .97810 70 4 .97303 67 4 .98306 74 4

11 .98360 75 5 30 .99422 71 4 150 .98115 67 A

4 .96907 81 7

2 .95775 53 6 .98849 66 4 150 .99008 66 4 .98221 65 5 .98344 58 3

7 .97591 85 5 .98512 78 6 .97919 83 4 .92675 75 3 12 .98779 76 6 .98045 68 4 10 .97829 63 5

8 .98665 69 5

3 .93065 82 6 .97987 80 4 .97715 60 3 20 .99440 66 5

2 .87186 92 9 30 .99392 67 4 .97912 69 3 11 .99098 80 6

3 .96353 70 5 .99384 70 t 14 .98801 66 4 .95527 57 4 .99284 74 7 35 .98752 60 4

3 .94329 98 5 .95120 45 3 .91433 51 4 .97885 65 4 .98811 85 3 4 .94101 88 8 45 .98207 85 6 .98670 71 k .96510 75 3 .97563 88 6 .97227 81 4 13 .98313 75 11

(a) Corresponding to a Type I distribution

(b) Estimated by method of moments

(c) At this station the largest yearlv wind speed data may not provide a reliable

basis for ^^redicting extreme winds - see Appendix 1

1-29
APPENDIX 1

STATIONS AT WHICH THE LARGEST YEARLY WIND SPEED DATA MAY NOT

PROVIDE A RELIABLE BASIS FOR PREDICTING EXTREME WINDS

As indicated in some detail, for example, in Reference 4 (p. 84), in a hurricane-prone

region most of the speeds in a series of the largest annual winds are considerably lower than

the extreme speeds associated with hurricanes. It may then be argued that in hurricane-prone

regions the series of the largest annual speeds may not, in certain cases, provide useful

statistical information on winds of interest to the structural designer.

For this reason, caution is in order in using the results of the statistical analysis

for the following stations at which hurricane winds may occur:

Jacksonville, Florida • .

Key West, Florida \ .;

Tampa, Florida

Savannah, Georgia

Boston, Massachusetts . -

New York, New York

Cape Hatteras, North Carolina ' ..

Wilmington, North Carolina

Philadelphia, Pennsylvania

Block Island, Rhode Island

Brownsville, Texas ;

Corpus Christi, Texas

Port Arthur, Texas

Norfolk, Virginia

1-30
ANALYSIS OF HIGH WIND OBSERVATIONS FROM VERY TALL TOWERS

Keikichi Naito

Isao Tabata

Noboru Banno

Katsumi Takahashi

Meteorological Research Institute

Tokyo, Japan

-

. , ABSTRACT '

, .

This paper describes the analyzed results of high winds, such as typhoons, observed in

Japan and its vicinity from tall towers which vary in height from 200 to 400 meters. The

results show that the vertical profile of the average wind velocity is well simulated by the

power law: the obtained values of power are 0.12, 0.23, 0.46 and so on, and depend upon the

local topography upwind. When the upwind roughness of the ground surface, or the unobstructed

"clearness" which represents the reciprocal character of upwind roughness, are introduced to

model the characteristics of the strong wind, the power decreases with larger degree of

clearness, and vice versa. The power is expected to be around 0.12 or 0.15 in extreme clearness. This seems to agree with what is called the seventh-root formula obtained in wind tunnel experimentation. The power is considered usually to be less than 1.0, but will be over

1.0 in severely obstructed situations. The turbulent intensity decreases generally with the

—0 1 increase of height. It shows the following height dependency: for example it is 0.52 Z

for good clearness, and 1.02 Z ^'^^ and 2.14 Z ^'^^ for intermediate, or worse clearness,

respectively, where Z represents the height in meters. These results show that the profile

of turbulent intensity depends upon the effect of the local terrain. The gust factor shows

characteristics somewhat similar to those of turbulent intensity. The energy spectrum of

turbulence is simulated quite well by the -5/3 power law, but the local topography affects

height variation in the distribution of small spectral peaks.

KEYWORDS: High wind analysis; gust winds; tall towers; wind turbulence spectrum.

1-31
1 . INTRODUCTION

There have been only a limited number of high wind observations taken in Japan and its

vicinity by the use of very tall towers. Under these circumstances, therefore, observations

of typhoon 17 (T17/70) on the Iwojima Island in 1970, typhoon 2 (T02/70) on the Okinawa

Island in 1970 and the severe gust at Kawaguchi in spring 1970 are very valuable for wind

analysis. These tower locations are denoted I, 0, and K, respectively, for simplicity. At

location I, a Loran tower height is 410 meters high, whereas at location 0 and at location K,

respectively, broadcasting tower (165 m) and the NHK broadcasting tower (313 m) were used.

Each tower is equipped with aerovanes at several heights, where observations were continuously undertaken for about one and a half years. Since aerovanes were attached to the tower

side only at the one direction of azimuth, the effective observations were confined to the

strong winds mentioned above. Figure 1 shows the observation locations.

On the campus of the Meteorological Research Institute at the Tsukuba Scientific City,

a meteorological observation tower height of 213 meters has been installed and meteorological

instruments and data analysis facilities have been gradually installed. Winds have been

observed for one and a half years. During this period, such a strong wind as a typhoon has

not been observed. This circumstance results in a wind over 10 m/s at the top of the tower

being specified as a strong wind for the data taken.

This paper describes the analyzed results based on these strong winds. For vertical

profile of wind speed, data of ten-minute averages were used, and the power law obtained by

simulation was examined. The characteristics of turbulence, turbulent intensity, gust factor and spectrum are investigated.

2. VERTICAL PROFILE OF STRONG WIND ^ -vi.-J:

2.1 Typhoons and a Severe Spring Gust [1]

For the vertical profile of strong wind, an investigation will be made concerning the

following power law,

u - ZP (1)

where u and Z represent wind velocity and height, respectively. Figure 2 illustrates

examples of the vertical profile. They are well simulated by the power law and the same

1-32
holds in all other cases. The obtained value of the power p has no large variation at the

same location, but the difference of the power p is seen clearly among those at the different

locations. Table 1 shows the average power at each locations.

For a relatively weak wind, p should depend upon the atmospheric stability or the vertical gradient of air temperature. For a strong wind, however, the stability would be almost

neutral and then p should be affected largely by a topographical difference.

Figure 3 shows the topography of the three locations. The tower at the location 0

stands on an isolated and small hill, where a plateau and a long stretch of hills rich in

trees are to the wind direction. For the power p at the location 0, the height from the

plateau surface is taken as Z. When the height from the base of the tower is adopted as Z,

a significant kink appears in the vertical distribution. However, for the use of the height

above mentioned, this kink disappears and hence fits the power law very well as a whole. K

is located in the Kanto Plains with a small topographic undulation, a grass field, and a

small residential area to the wind direction. ...

At the location I, there are no obstructions to the wind, whereas at the location 0,

there are many obstructions. The condition of location K is somewhere between these two.

These topographical conditions yield the following results: At the least rough surface

(1-location) , the value of p is smallest, whereas at the roughest surface (0-location) , it

is largest. In the intermediate surface (K-location) , it is between these values. These

results will be examined in detail in the following sections. ,

.

2.2 Strong Wind at the Tsukuba Science City

This section describes the analyzed results of strong wind over 10 m/s at the top of

the tower for a year from October 1977 to September 1978. The ten minute averages of wind

speed were obtained hourly. It is noted as illustrated in Figure 4 that the power law

applies well. Table 2 shows the monthly statistics of p. It is rather insufficient to have

reliable statistics in July due to a small number of strong wind samples. The standard deviation is not always small, but it is easily seen that the value of p is small in May, June

and August and that it is large in October and November. This change of p should be

examined

.

The change of p is considered usually due to the change of the atmospheric stability.

Unfortunately it has become clear there is a problem in the accuracy of the measurement of

1-33
air temperature at the meteorological observation tower. At present an effort has been paid

to improve the accuracy, but the analyzed result of the atmospheric stability is not yet

ready to show. However, in the case of a strong wind, vertical variation of air temperature

is not an influential factor as explained in the previous section, whereas local topography

upwind may be an important factor.

The Tsukuba Scientific City Area is generally a flat terrain, but there are many blocks

of buildings of research laboratories and pine groves which obstruct the air flow near the

ground. These circumstances yield the results that the wind profile obtained at the 200

meter tower might be affected by near areas with obstructions of over 20 meters in height.

A direction where such obstructions exist is termed in this paper an "obstructed azimuth"

for convenience; an unobstructed approach is called a "clear azimuth." ...

All the horizontal direction is divided into the obstructed azimuth, 0, and the clear

azimuth, C, at the center of the meteorological observation tower (see Figure 5). In this

figure the distance of 2 km from the center is undertaken. In the obstructing area, buildings of over 35 meters or even 30 meters in height are very few, thereby the distance of

2 km is almost sufficient. In fact the area within 31 km from the center does not show any

significant difference from that within 2 km in this case. It should be noted further that

obstructions over short distances influences the wind quite a bit.

There is an interesting result found in comparison of the azimuthal characteristics

shown in Figure 5 with the windrose obtained each month. In May, June and August when the

value of p is small, the wind comes from the clear azimuth (C-azimuth) very frequently and

in the other months the wind does from the obstructed azimuth (0-azimuth) frequently. This

fact is illustrated in Figure 6 showing the windroses in June and October, and corresponds

to the correlation between the upwind topography and the power p as mentioned in the previous section. In addition it should be noted that both the topography and the power in May,

June and August resemble those of the case at the location K.

So far this correlation has been examined rather qualitatively, but it is desired to

investigate more or less quantitatively. Therefore the term "clearness" is introduced in

regard to upwind topography. Clearness is defined as the percentage frequency of wind

coming from the C-azimuth and is obtained by comparing the windrose with the azimuthal

characteristics as shown in Figure 5. Clearness seems to be reciprocal to surface roughness.

1-34
At present clearness is evaluated, very roughly, to around 85 percent in June and 28

percent in October. In Figure 7, the dots show the monthly clearness C and power p. As a

whole, it is not clear in this figure that there is a good correlation between the clearness

C and the value of p, but in the region of C over 60 percent, p decreases with the increase

of C, as shown in solid line. For the value of C over 85 percent an extrapolation is made

in the figure with a broken line.

It is very interesting to note that, at the C of 100 percent, p is in the region ranging

from 0.12 to 0.15, as indicated by the broken line. As mentioned in the previous section, at

the location I the local topography is extremely open to the wind and p is 0.12. This gives

good agreement with the extrapolation of C at 100 percent.

Furthermore there is a well known result which is called the seventh-root formula

obtained by a wind tunnel experiment: the p value of l/7(= 0.142) occurs in the generally

accepted power law profile for the tubulent boundary layer of a flat plate in a wind tunnel

[2]. Obviously this experiment corresponds to the case at the C of 100 percent and verifies

the extrapolation. With these considerations, the trend of solid and broken lines would be

an understandable and acceptable fact.

As a matter of fact the definition of clearness in this section would be beyond satisfaction. It should be defined by taking into account the distance between observation

point and obstruction position, the height of obstruction, the size of obstructing area and

so on. Thus the value of C under 60 percent in Figure 7 would be under-estimated if compared

with that obtained by the use of the satisfactory definition of clearness. For this reason,

the position of dots with C under 60 percent should be shifted to the right considerably.

Another thing to note with the aid of the standard deviation in Table 2: the value of

p is possible in some cases to be over 1.0 even in such flat terrain as the Tsukuba area.

This should be pointed out here, since the value of p is usually considered to be less than

1.0. [2]

3. TURBULENT CHARACTERISTICS

The fluctuation of wind speed was observed by the instantaneous measurement of the

aerovane propeller rotation as mentioned in Section 2.1. -In estimating the turbulent characteristics, ten minute sampling was adopted in the case of a significant trend in wind,

whereas in the rest of the cases thirty minute sampling was adopted. In the observation

1-35
described in the Section 2.2, supersonic anemometers are used for the turbulence. However,

the observation system has not been completed yet for strong wind and analysed results are

not ready to show in this paper.

3.1 Turbulent Intensity

Turbulent intensity is defined as follows:

standard deviation of wind speed fluctuation

Turbulent intensity = —— —————— ——— ____ . j , average wind speed u

The values of this turbulent intensity are averaged at each observation height and

plotted against the height as is shown in Figure 8. The turbulent intensity decreases with

the height of observation. It shows surprisingly linear dependency with height in this

figure. Hence the turbulent intensity in a vertical profile is possibly simulated by the

following power law: ~ s

i--- - '

£H. = rz-q (2) u

where r and q are constant.

Table 3 shows empirical equations obtained by the use of equation (2) and Figure 8.

The values of r and q increase with the order of locations I, K and 0, as indicated in this

table. These values are clearly affected by the surface roughness or the upwind clearness

as seen at the vertical profile of the average wind speed. The power law obtained experimentally should be examined in details for various terrains.

There is another point to be noted in Figure 8. The height variation at the location

shows an inversion at above 200 meter height. This phenomena was noticed only at the

location I, but there is not reason to cause this inversion in the profile of average wind

and the surface roughness. It might be caused by the internal waves in the atmosphere, but

' j'"'

it is not verified at all at present.

3.2 Gust Factor

Gust factor is defined by the following expression:

maximum wind speed Uj^

Gust factor = average wind speed u

1-36
Analysed results show that all the characteristics of gust factor are similar to those of

turbulent intensity. Generally the gust factor is large at low altitudes and small at high

altitudes. Moreover there appears quite the same inversion at the location I as that of the

turbulent intensity. The obtained values of the gust factor disperse very much, but it could

be seen in Figure 9 that the gust factor tends to decrease with the increase of average wind

speed.

3.3 Energy Spectrum of Turbulence

Figure 10 shows one of the examples of the calculated spectra. The trend of this example is quite general. In the range of higher frequency especially more than 0.02 Hz, the

so-called -5/3 power law holds. However, a further examination in detail shows spectral fine

structures. In Figure 11 frequencies of spectral peaks over 2dB high locally are plotted at

the observation heights. The spectral fine structures in this figure were obtained by the

use of observations arbitrarily picked up. In this figure, in the case of the I location,

the distribution of the spectral peaks remains rather unchanged at all the altitudes. In the

case of the K location, the range of the peaks seems to be somewhat dispersed at the low

height. At the location 0, it is clear that the range of the peaks is much dispersed at the

lower altitudes. These facts seem to show the effect of the surface roughness as seen in the

vertical profile.

4. CONCLUSION

Strong winds, such as typhoons, were analyzed based on the observations by the use of

tall towers ranging from some 200 meters to 400 meters. The following was concluded for the

average wind speed.

1) In the vertical distribution, the power law fits very well.

2) The powers are 0.12, 0.23, 0.46 and so on, which depend upon the upwind terrains.

In the case of small roughness in the surface or open clearness in the upwind, the power is

small and vice versa.

3) The minimum value of the power seems to agree with what is called the seventh root

formula empirically obtained by a wind tunnel experiment. In the worse case of clearness,

the power is possible to be over 1.0, beyond our expectation.

1-37
The following was concluded in the turbulence characteristics.

4) The turbulent intensity decreases generally with the increase of height by showing

the following height dependency: for example, it is 0.5 Z"0*31, 1.02 Z"0-'^2^ ^^d 2.14 •'^^

in the case of good, intermediate and worse clearness, respectively, where Z is the height in

meters

.

5) Turbulent intensity distribution, as described in (4), is clearly affected by the

terrain. Moreover, at high altitudes unusual inversion can be present in some cases.

6) Gust factors show a similar trend to turbulent intensity, except that the former '

disperse more than the latter. • - • '.ji,,' ,

7) The energy spectrum is expressed by the -5/3 power law, but fine spectral peaks are

affected by the terrain. -

"

ACKNOWLEDGMENTS

The authors wish to express their thanks to Dr. Tsutomu Takashiraa, Meteorlogical *

Research Institute, for his many useful suggestions. ' '

REFERENCES

[1] Naito, K., Takahashi, K., Kinase, R. , Toyama, N., and Ito, S., 1973, Strong Winds

Observed from Very Tall Towers, Chokoso Tower Shiryo, Japan Meteorological Agency,

No. 5, pp. 20-32 (In Japanese).

[2] Sutton, 0. G., 1953, Micrometeorlogy , McGraw Hill, New York, p. 238.

1-38
IWojima (1) Kawaguchi(K) Okinawa{0)

0.12 Z0.2 3 Z0.46

Z

Table I Power law obtained over the dIfferehT terrains.

Table 2. Statistics ot the exponent p in the power law

Obtainea In Tsukuba Science City

Monfh Oct. ,'77 Nov„'77 Dec.,'77 Jon.,'78 Feb., '78 Mar,'78 Apr. ,'78 May,'78 Jun.j78 (Jul,'78] Aug.,78 Sep. ,'78

Sample

NurTiber_

P

44 7 0 9 2 132 1 4 i 7 7 1 1 9 735 185 (15) 1 29 3 5

0450 0.482 0.396 0404 0399 0391 0.372 0.290 0.258 (0418) 0.271 0 369

S.D.p 0.087 0258 0 159 0.242 0.201 0.183 0.225 0283 0 182 (0170) 0 128 0.417

^ I Average value of p X-2 Standard deviation of p

Iwojinna ( 1

)

KawaguchKK) Okinawa (0)

-0.31 0.52 Z

-0.42

1.02 Z

-0.54

2.14 Z

Table 3 Power law lor turbulent intensity.

1-39
Fig 1 Locations of the observation points

i-CJtJ

fiun t T17^70 SO 3C 40 30 20 50 40S0

^60

5

Mean wind speed (it\/s) Mean wind speed (m/s)

10 20 3O405O Mean wind speed (m/s)

Fig. 2 Vertical profiles of the mean speed of strong Wind

1-40
5 I ^ zoo\

^ 'oo\

wind direction An isolated and drcj!nr hii 40m2_

Okinav/a

5 (Km)

Fig. 3 Topographical conditions and observed wind directions

around the towers.

10' 1 I I I M I 1 1 10" ' 1

I I M I 1 1 3 4 5 6 8 10 20 3 4 5 6 8 10 20

Mean wind tpttd(m/i) Mtan wind tpttd(m/s)

Fig. 4 Examples of the vertical profile of the mean wind speed

larger than lOm/s at the tower top in Tsukuba Science City.

1-41
180°

MT : Meteorological observation tower. C : Clear azimuth viewed from M T. 0 : Obstructed azimuth viewed from MT. Shaded area : Blocks of buildings and pine groves.

Fig. 5 Azimuthal pattern of the clearness, C,

and obstruction,©.

Fig. 6 Examples oi monthly windrose and azimuthal pattern

of the clearness and obstruction.

1-42
1.0

0.8 i 0.6

0.5

0.4

0.3

0.2J

0.

1

10

I I I I I I I I « 20 30 40 60 80 100 (%)

C(Clearnes8)

Fig. 7 Correlotion between the -clearness C

and the power P.

0 : Okinawa

K : Kawaguchi

1 Iwojima

Turbulent Intensity

Fig. 8 Height variation of turbulent intensity

1-43
'•t5 « 0 - o e

3

HEIGHT Un)

. • 165

134.5

* 103

* 72

17.5

T7002/OKIfslAWia

9 •

16 20 24 28 32

MEAN SPEED (m/s)

36

Fig 3 Correlation between the gust factor and the mean wind speed.

^.05 o

^ .03

.02

.01

.OOG

.003

.002

OOI

h (m) 405

U (mA

12 1

y —

:

1 Iwoj

1

i

i mo

\

,\

.002 .Ol .02 05 . I

Fig. 10 An example of calculated

energy spectrum of turbulence.

1-44
380-

9— 200-

«* I

4) I

X

X XX

X XX

XX XX

' X XX

XX X

XX XX

X XX

Iwojima

I

'

I " " I

1—' ' I " "I 1—[—

Kawoguchi X

XO X

O Xo X

I I T—I ' I " "I 1—

r

Oklnowo

X XX

X XX X

X X X X X

X XX X X

' ' I "... 1 r '

I ""I 1

'- 003 005 01 02 05 .1 2 .3 ( Mt) Frequency

Fig. 11 Local peaks in

speed fluctuations

1-45

the spectra of wind

at various heights.

A PROPOSAL FOR A NEW PARAMETER IN ASSESSING SEISMIC DISASTER

Mlchio Otsuka

International Institute of Seismology and Earthquake Engineering

Building Research Institute

ABSTRACT • .

:

Out of the recent progress in modern seismology, especially in the rapidly increasing

body of knowledge about the nature of what the fault generating process looks like, a prospect of practical earthquake prediction in the near future seems reasonable.

Although a general methodology is not at hand for complete forecasting, long-term

estimating of earthquake occurrences in limited areas based on past seismicity studies has

already been put to practical use in many reported instances. Though this technique is

still insufficient for prediction, since it lacks temporal accuracy about the occurrence of

the earthquake, it should prove invaluable and be reflected in disaster reduction strategy.

What is needed still is the methodology to associate the expected fault geometry to the

plausible disasters.

The purpose of this paper is, then, to search for a technique to evaluate the influence

of the impending earthquake based only on a geometrical estimate of a 'should be' fault and

without any dynamic information.

Through repeated trials, the author has arrived at the conclusion that the intensity

of the seismic vibration is closely related to the solid angle extended from the observation

site toward the periphery of the fault.

KEYWORDS: Earthquake prediction; fault dynamics; seismic disaster parameters.

II-l
1 . INTRODUCTION

There are at least two categories among modern seismological research items which seem

to be promising for future earthquake engineering. One among them is the study of the focal

process of an earthquake and the other is earthquake prediction.

a) Focal Process

Integration of seismograms to depict what sort of dynamic process has caused that event

in relation to the tectonic system of the focal area has come into popular practice among

seismologists. This technique can easily be extended to guess the pattern of a seismogram

which should be recorded at any arbitrary site. This feature seems to be of great value

since it is expected to produce invaluable data for disaster reduction strategy. However, a

lack of knowledge to enable the prediction of the detailed focal process in the present stage

causes a bottleneck for the practicality of this approach.

b) Earthquake Prediction

Earthquake Prediction is divided into two categories depending on the duration in time

after issuance of warning until the occurrence of the earthquake: long-term, and short-term

prediction. It goes without saying that short-term prediction is much more effective for

reducing loss of human lives and property. From the standpoint of building construction

with longer lifespan, however, long-term prediction is useful as well and should be taken

into account also. The complicating circumstance is that for long-term prediction, no

detailed information is available. What is needed for the technique to make the best use of

long-term prediction, therefore, is that it reflects the meager knowledge about the impending

earthquake to disaster reduction strategy.

It may be premature to discuss the successful use of this information since a considerable amount of inaccuracy cannot be avoided now, but it is the firm belief of the author that

every possible effort should be made to minimize the possibility of earthquake disaster by

making the maximum use of even the small amount of knowledge now on hand.

2. AN EXAMPLE OF LONG TERM PREDICTION — SEISMICITY

STUDY OF KURIL-HOKKAIDO REGION

Figure 1 shows the chain of the focal domains of the events in recent years compiled

by Utsu (1972 a, b). Utsu remarked that focal domains of each earthquake are strictly

II-2
bounded and do not superpose each other. He also pointed out that domain C looks likely to

experience an occurrence of a fairly large earthquake since this area has not experienced any

since 1894. ...

Shimazaki (1972) associated crustal deformation observed at the eastern part of Hokkaido

to the compression of continental lithosphere subjected to the stress of the Pacific Plate.

As was anticipated, an earthquake with magnitude 7.4 occurred on June 7, 1973, off Nemuro

Penninsula situated at the eastern tip of Hokkaido.

This is an example of successful long-term prediction of earthquakes based only on gaps

of seismicity and of crustal deformation. The technique exemplified herein is one regarded

to be the most reliable among various methods proposed toward the prediction of earthquakes

through many successful reports (Sykes (1976), Borovik, et al. (1971), Ohtake, et al.

(1977)).

It must be noted, however, that information obtained from this kind of study is not

about the detailed mechanism of the focal process, but about the vague location of the focal

region.

3. PARAMETER M AND A

The trial is customarily made to associate the seismic damage with the magnitude M of

that earthquake and with the distance A from the epicenter. Various formulas are successfully proposed in this regard proving that this technique is promising for application to

seismic disaster assessment. However, it should be noted that each of the two parameters

has its own inadequateness judging from seismological view point.

a) Magnitude M

Since magnitude M of the earthquake is basically determined based on the maximum

amplitude of seismograms, it appears to be the most effective measure to assess a seismic

disaster. Unfortunately, however, it has been shown that the magnitude scale is accompanied

by an inherent skew to represent the real potentiality of seismic ground motion originating

from the frequency characteristics of seismographs. It seems fatal, moreover, that tendency

of skew is more pronounced in the region of "large earthquakes" to which earthquake engineering should mainly be concerned.

1 1-3
b) Epicentral Distance A ,

Epicentral distaace A is defined as the distance between the epicenter and the site.

Association of seismic disaster with epicentral distance is only varied when a major part of

seismic energy is radiated from the hypocenter. According to the modern understanding of

seismology, hypocenter is just a point from which the fracture started and the major part of

seismic energy is radiated from a rather wide region around the fault.

In addition, parameters M and A involve some physical confusion. Use of these two

parameters for assessing the seismic disaster does not suit the type of long-term prediction

of earthquakes described above. •.. : t,.

4. NEW PARAMETER

Seismic disaster or intensity of earthquake is regarded as a complicated function of

motion. It is hardly possible to express it with a limited number of parameters, but as a

primary step to the study, a parameter is developed for expressing the acceleration in

place of M and A.

For that purpose we start from a simple a priori assumption connecting seismic peak

acceleration and source geometry.

1) Point Source ,;j , : /

As shown in Figure 2, a simple dynamic situation is considered that a single force F

is exerted on a point P. It is assumed that seismic peak acceleration observed at 0 is

expressed by i :

a = Ki ' (1)

r ^

' , . .-.

where r is the distance between P and 0, and k is a constant.

2) Dipole Source

As is shown in Figure 3, when dipole force is exerted on P instead of a single force,

the peak acceleration at 0 is expressed as

a = kp/ ^ - 1 \ = k cose . (2)

\r - A cos6 r + A cosO/ r^

2 2 .

II-4
expanding the Eq. 1. The quantity Fl is the dipole moment and 6 represents the azimuth of

vector PO in relation to the dipole reference.

3) Dipole Shell Source

Let us consider a group of dipoles lined up so as to construct a shell with positive and

negative surfaces as is shown in Figure 4. The force exerted per unit area of the surface is

designated as F' and surface element as dS. Peak acceleration can be obtained as Eq. 3 by

integrating Eq. 2.

a = ! kill cose dS (3)

r2

where integration ranges over the entire surface.

While, using the relation r^ dw = cos 9 dS (dw is solid angle extended by dS) , Eq. 3 can be

written

a = / kF'l dw = kF'l n (4)

where Q represents the total solid angle extended from 0 toward the entire surface.

The strain e that ground material exerts is given by

e = D

where D denotes the dislocation due to the formation of the fault. Denoting the rigidity of

the material around the focal area as y , F' can be expressed as

£

merging Eq. 5 with Eq. 4, we get

a = k\x DU ' ' (6)

Eq. 6 signifies that if p and D are constant for an earthquake, a is proportional to H

Taking logarithm of Eq. 6, we get - . _ -

.

log a = C + log n (7)

II-5
Eq. 7 shows that when plots are made of peak accelerations a against the solid angle toward

the fault plane, both on logarithmic scale, they should be aligned on a straight line with

gradient of unity.

5. MIYAGIKEN-OKI EARTHQUAKE ,4.

Figure 5 shows schematically the fault geometry of Miyagiken-Oki Earthquake worked out

by Seno et al. (1978) which occurred on June 12, 1978. Reported peak accelerations for this

earthquake are plotted against the solid angle extended to the fault plane from the site on

log-log sheet in Figure 6. It is observed in Figure 6 that plots are aligned on a single

line with a gradient of unity as expected which shows that our a priori postulation is not

physically absurd.

6. VOLCANIC FRONT

Grouping of the plots in Figure 6 into two parts is made by fitting a line to show the

approximate trend. Solid and open circles show the sites which exhibited peak accelerations

more and less than average, respectively. Replotting the solid and open circles to the

original sites given in Figure 7, it is remarkable that the distribution of plots show an

extremely interesting feature from a geophysical view point. That is, the position of the

border between the two symbols looks very similar to the so-called volcanic front line,

across which seismic wave dissipation characteristics are believed to be markedly different

because of difference in heat flow.

For reference sake, the same grouping is applied to the traditional technique using

epicentral distance, and plots are made in Figure 8. Note that no physically interesting and

systematic interpretations are available. ,,v, .

7. CONCLUSION

It should be admitted that several clear drawbacks do exist for the technique proposed

here. For example, information on the minute structure of fault generating dynamics is

entirely too thin, and a dipole source is not appropriate from modern seismological understanding. These are merely two points, among others, which need further refinement.

II-6
REFERENCES

Borovik, N. , Mansharina, L. , and Treskov, A. , On the Possibility of Strong Earthquake

in Pribaykalia in the Future, Izv. Acad. Sci. USSR, Phys. Solid Earth, English Translation, (1) 13-16 (1971).

Ohtake, M. , Matsumoto, T. , Latham, G. V., Temporal Changes in Seismicity Preceding

Some Shallow Earthquakes in Mexico and Central America, Bulletin International Institute of Seismology and Earthquake Engineering 15, 105-123 (1977).

Shimazaki, K. , Focal Mechanism of a Shallow Shock at the Northwestern Boundary of the

Pacific Plate: Extentional Feature of the Oceanic Lithosphere and Compressional

Feature of the Continental Lithosphere, Phys. Earth Planet. Inter. 6, 397-404 (1972).

Sykes, L. , Aftershock Zones of Great Earthquakes, Seismicity Gaps, and Earthquake

Prediction for Alaska and Aleutians, U. Geophys. Res. 76, 8021-8041 (1971).

Utsu, T., Large Earthquake Near Hokkaido and the Expectancy of the Occurrence of a Large Earthquake Off Nemuro, Rep. Cook. Comm. Earthquake Prediction, 7, 7-13 (1972 a)

Utsu, T., Aftershocks and Earthquake Statistics, TV, J. Fac. Sci. Ser. 7, 4: 1-42

(1972)

.

II-7

Fig. 5 Schematic representation of the Miyagiken Oki earthquake fault

II-9
solid angle (rod.) (Q,)

Fig. 6 Peak acceleration against the solid angle

11-10
Fig. 7 Distribution of magnitude of peak acceleration

(grouped by a-^ relation)

11-11
Fig. 8 Distribution of magnitude of peak acceleration

(grouped by a-A relation)

11-12
SIMULATION OF EARTHQUAKE GROUND MOTION AND

ITS APPLICATION TO DYNAMIC RESPONSE ANALYSIS r Tetsuo Kubo

Makoto Watabe

Structural Engineering Department

Building Research Institute

Ministry of Construction

ABSTRACT ..

Through the relation of Fourier transformation, two types of stochastic modeling of

earthquake ground motion are introduced. Twenty samples of synthetic motion, in each case,

are generated by use of these two modelings simulating a certain recorded motion. Using

these motions, characteristics of synthetic motions such as the cumulative energy distribution (the integration of square acceleration), the maximum elastic response and the maximum

elasto-plastic response are evaluated. From a statistical viewpoint, the results are compared with those obtained from the recorded earthquake motion in an attempt to make use of

synthetic motions for an engineering application.

KEYWORDS: Dynamic response analysis; ground motion; modeling; synthetic earthquake motion;

waveforms

.

11-13
INTRODUCTION

For aseismic design of a structural system, there would be two alternatives to determine

an earthquake input to the system. One is to use a waveform of motion, and the other is to

use a design spectral curve such as that in the literature [1]. When using the former procedure, the following two manners can be used to determine a waveform for structural design.

The one is to determine a waveform either by use of real strong motion waveforms such as

those recorded at El Centro, California and at Hichinohe Harbour, Japan with consideration

upon such as subsoil conditions, or by use of real waveforms of intermediate-size earthquakes

recorded at the prospective construction site. It has been recognized that earthquake

motions, even recorded at the same site location, have often yielded quite dissimilar characteristics to dynamic response properties. Hence, there would arise a question "Could the

specific motion of a past earthquake be representative of a future ground motion at a certain

site location?".

The other is to determine a waveform by use of a stochastic process having appropriate

properties in a statistical sense. Recently by a sei smological approach, time trace of an

earthquake motion can be calculated determining the source mechanism and physical properties

of wave propagation path. However, considering complexity and irregularity in the initiation

of seismic wave and its transmission paths, stochastic modeling of earthquake motion is a

realistic and feasible approach for an engineering application.

Under certain mathematical conditions, a time function corresponds uniquely to its

Fourier transform through the relations

where f(t) and F(cj) represent a time function and its corresponding Fourier transform,

2 respectively, and j denotes the imaginary unit, i.e. j = -1 . Fourier transform in Eq. 1 can

GENERATION OF SYNTHETIC EARTHQUAKE MOTION PROCESS

(1)

F(t) (2)

be expressed by a pair of an amplitude function and a phase function through

11-14
F((o) = A(w) exp [-j$(a))] (3)

where

A(a)) = |F(a))| • (4)

= -tan"^ [Im[F(oo)]/[Re[F(a))] ] (5)

are Fourier amplitude spectral function and Fourier phase spectral function, respectively.

Eq. 1 through 5 indicate that time function f(t) and a pair of functions A(u)) and $(cj) relate

uniquely with each other through the relation of Fourier pair.

Provided that f(t) is a real time function such as an earthquake accelerogram studied in

this paper, Eq. 2 can be written in discrete form by

N/2

f(t) = Z cos (oj^t - (6)

n=0 . - _ . :

•

where

a^^ = Aa)/2Tr (n = 0 and N/2), a^^ = A^^ Aoj/tt (otherwise)

^n "

'^('^n)' % " ^^"^ " 2tt/Tq, and •p^ = $(aj^)

(7)

Let Fourier amplitude and Fourier phase spectral functions obtained from recorded earthquake accelerogram fpCt) be denoted by Aq(w) and 'I'q(oj) respectively. The following two

stochastic modelings of synthetic earthquake motion are introduced.

The first, called the Modeling I hereafter, is to determine Fourier phase spectrum by

analysis of a real earthquake motion and to determine Fourier amplitudes by uniformly distributed random numbers, i.e. .. -

N/2

g^Ct) = Z a^^ cos(a)^t - $^^) (i = 1,2, ... ) (8)

n=0

in which i denotes the i-th individual member from the ensemble and a^^ are random numbers

distributed uniformly over the range of 0 and 1.

The second, the Modeling II, is to establish Fourier amplitude spectrum from a real

earthquake motion giving random phase angles distributed uniformly in the interval of 0 and

2Tr to Fourier phase spectrum; thus

11-15
N/2 " ^

g^(t) = Z a^^ cos((o^t - ^^^) (1 = 1,2, ... ) (9)

n=0 •

. . .

in which 41^^^ represent random phase angles uniformly distributed over the range between 0

and 2ir.

SYNTHETIC EARTHQUAKE MOTIONS

Modeling I Since both the mean value and the auto-correlation function are given as

functions of time [2], the process reveals nonstationarity

.

The amplitudes in Eq. 8 are given by random numbers in the range of 0 and 1. When comparing the synthetic motions with a recorded motion, one should make an appropriate modification upon amplitudes of the generated waveform. Many intensity scales have been proposed to

estimate a seismic intensity. Among them, the scale introduced by Arias [3] is selected in

this investigation. Determining the intensity scale of a synthetic motion equal to that of a

real motion, the waveform of synthetic earthquake motion fj^(t) is given through the relation

fi(t) = c^ gi(t) (10)

where c^ is a scaling factor obtained from

f^2(t) dt / \ g.2(t) dt (11)

^\Jo U o

in which fQ(t) represents the waveform of a recorded earthquake motion. -

'

Modeling II Both the mean value and the auto-correlation function are time invariant.

Average over time of any member from the ensemble equals its corresponding average across the

ensemble [2]. It leads to the conclusion that, in addition to being stationary, the process

is ergodic.

It is easily shown that the intensity of process g^(t) in Eq. 9 is essentially equal to

that of recorded motion f^(t). Since the process has been concluded stationary, it should be

multiplied by an appropriate deterministic time function to reveal adequate rise-up and

decaying of intensity of waveform during shaking. Let the energy of waveform at time t be

defined through - . . ;

•

11-16
:.(t) = \ £^2(3) ds ,

' (12)

J o

and suppose synthetic earthquake waveform f^(t) is expressed by the product of process g^(t)

and deterministic time function C(t),

fi(t) = C(t) gi(t) (13)

Assuming that the energy of a synthetic earthquake waveform at time t would be expected to

equal the corresponding energy of the real earthquake waveform, deterministic intensity function ?(t) is given as follows

C(t) = / Td ^ Vt)/Eo(TD) (14)

or in a normalized form,

e(T) = /^eo(T) (15)

in which t and e^ denote normalized time variable t/Tjj and normalized energy of waveform

E^{t)/E^(,Tj^), respectively.

In this investigation, considering conditions upon a form of function C(t), an intensity

function is determined through the following manner. In Figure la, the curve of a function

for energy of a waveform is shown. Divide axes x and y into m and n pieces, respectively,

and obtain the intersecting points with the curve. Connecting these points in succession by

straight lines, approximate the curve with the pieces of straight line. Consequently in this

study, the deterministic intensity function is obtained by a step function as shown in Figure

lb.

Earthquake Waveform The real waveform, the East-West component at Hachinohe Harbour

recorded during the Tokachi-oki earthquake of May 16, 1968, is selected. Figure 2 presents

the time trace of the real accelerogram. Twenty samples of synthetic motion, in each case,

are produced by use of these two modelings simulating the recorded waveform. In Figures 3

and 4, time traces of synthetic earthquake motion are shown. It can be recognized that plots

of motions from the Modeling I reveal quite similar features with one another, while they

show higher fluctuation than the plot of the real earthquake motion.

11-17
CUMULATIVE ENERGY DISTRIBUTION

Energy distribution of waveform associated with time for a synthetic motion is examined

by terms of the so-called "normalized cumulative energy function,"

ei(T) = E.(t)/E.(Ti3) (16)

where t and denote normalized time variable t/Tj-, and the energy of waveform defined previously by Eq . 12, respectively. For motions from the Modeling II, the energy function is

expected to coincide with that obtained from the real motion with certain amount of fluctuation. Hence, examination is carried out only for motions from the Modeling I.

Cumulative energy functions both for synthetic motions and for the real motion are presented in Figure 5, in which x and y axes represent normalized time variable t and normalized

cumulative energy distribution e^, respectively. Six curves in the figure show the cumulative energy functions for cases described as follows; the thick solid curve for that of the

recorded motion, the thin solid curve for the mean value of those obtained from the twenty

samples of synthetic motion, two dot-dashed curves for the mean value plus or minus ten

times the standard deviation of those obtained from the synthetic motions, and two dashed

curves for the maximum and the minimum of those obtained from the synthetic motions.

The results indicate that energy distribution functions for waveforms from the Modeling

I reveal quite similar plots with one another. Consequently, forms of envelope curve which

specifies intensity varying with time for synthetic motions are similar with one another. It

can be observed that the general form of envelope plots for synthetic motion is similar to

that obtained from a real earthquake motion.

MAXIMUM ELASTIC RESPONSE SPECTRA

Figure 6 and 7 show the results of maximum elastic response spectra for motions of the

selected component from both the Modeling I and the Modeling II, respectively. Axes x and y

represent the natural period of an oscillating system and the maximum pseudo velocity

response associated with a fraction of the critical damping of 0.05 in the system. In these

figures, the thick solid curve, the thin solid curve and two dashed curves represent the

results of the maximum response subjected to the recorded motion, of the mean value of maximum responses subjected to the synthetic motions, and the maximum and the minimum of maximum

11-18
responses subjected to the synthetic motions, respectively. The shaded zone shows the range

where responses fall in value within the interval of the mean value plus or minus one standard deviation taken across the responses obtained from the twenty samples of synthetic

motion. • .

. Modeling 1 Similarity between the maximum velocity response spectrum and Fourier amplitude spectrum suggests that the plot of the thin solid curve will be flat. The results in

Figure 6 reveal a fairly flat form. The magnitude of responses in the intermediate range

of periods is smaller for the synthetic motions than that for the real motion. For motions

from the Modeling I, one cannot observe the dominant period of motion included which is one

of the characteristic features of earthquake motion in engineering sense. Therefore, the

Modeling I is concluded not adequate for engineering practice.

Modeling II In Figure 7, the thin solid curve coincides with the thick solid curve

with good agreement over the entire range of period. It is an interesting result that the

responses subjected to a recorded motion usually fall in the shaded zone representing the

interval of the mean value plus or minus one standard deviation obtained across the synthetic motions. The twenty responses subjected to the synthetic motions associated with a

certain natural period of an oscillating system have a tendency to fall in value around the

mean value obtained from these twenty responses. Though distribution of responses cannot

be concluded Gaussian by x examination with confidential level of 50 percent, numerical

analysis yields the result that the maximum response subjected to a synthetic motion is not

greater than the mean value plus one standard deviation with probability of 85 percent,

while in case of Gaussian random distribution the corresponding probability is 84.13 percent.

MAXIMUM ELASTO-PLASTIC RESPONSE SPECTRA •

The responses of a linear system subjected to a stochastic process can be theoretically

determined by variations of statistical quantities associated with the input process. For a

ori-linear system, in general cases, one is forced to conduct numerical analyses. In this

j case, input excitation is essentially expressed explicitly as a waveform of motion. For

synthetic motions from the Modeling II, maximum elasto-plas tic response spectra are evaluated and are compared with those obtained from a real motion. ;

Presented in Figure 8 is a schematic rule for the Degrading Tri-Linear hysteresis model

[ employed in this investigation which is one of representative models for a reinforced

TI-19

1
concrete system. Cracking strength k^, and yielding strength ky are assumed 100 gal and 300

gal, respectively, and the ratio of the second stiffness to the initial one is 1/4.

Multiplying amplitudes of waveform by scaling factors 1.640 and 2.187 so as resulting peak

accelerations of the real motion to be 300 gal and 400 gal, and assuming a fraction of the

critical damping of 0.05, the maximum displacement responses are obtained both for synthetic

motions and for the recorded motion. Using logarithmic scale diagrams, the results are shown

in Figures 9a and 9b. The legends in these figures such as the thick solid curve, the thin

solid curve, the dashed curves and the shaded zone are identically similar to those in

figures showing the results of elastic response spectra.

Observation upon these figures leads to the evidence that the response subjected to a

recorded motion and the mean value of responses subjected to the synthetic motions coincide

well with each other over the entire range of natural periods. It should be noticed that

responses obtained from a real motion usually fall in the shaded zone. On an average, in

sixteen or seventeen cases out of twenty, the responses subjected to synthetic motions do

not exceed in value greater than the mean value plus one standard deviation obtained from

the responses subjected to the synthetic motions.

CONCLUDING STATEMENT

Through the correspondence of a time function with its Fourier transform, two types

of stochastic modeling of synthetic earthquake motion are introduced. Using this modeling,

synthetic earthquake motions are produced which simulate a certain selected recorded motion.

In an attempt to use these synthetic motions for engineering practice, characteristics such

as the cumulative energy distribution, the maximum elastic response and the maximum elastoplastic response are obtained, and are compared with those obtained from the recorded motion.

Dynamic properties of synthetic motions are summarized in the following:

1) Motions from the Modeling I have nonstationarity in intensity with time, and reveal

quite similar cumulative energy distributions with one another. Having quite similar forms

with one another, the general form of envelope functions from synthetic motions coincide

well with that from the recorded motion. In the maximum elastic response analysis, certain

dominant periods Included in the recorded motion cannot be synthesized by the Modeling I.

2) For motions from the Modeling II, the mean value of responses both in elastic analysis and in elasto-plastic analysis coincides with the response obtained from the recorded

11-20
motion with good agreement over the entire range of periods of an oscillating system. In

both analyses, the maximum response subjected to an individual member from synthetic motion

falls in the value not greater than the mean value plus one standard deviation obtained from

analysis across the synthetic motions with a probability of higher than 80 percent.

Although motions from the Modeling I have been concluded to be inadequate for an engineering application, they reveal quite a specific feature in intensity with time. It is

necessary that a mathematical model of synthetic earthquake motion be developed that considers the phase angles of waveforms for extensive application of synthetic motions to

engineering practice.

ACKNOWLEDGMENT

The authors wish to express their sincere thanks to Mr. Norio Suzuki, Graduate Student,

University of Tokyo, Tokyo for his great contribution to this paper. For other components o

real earthquake accelerograms, similar analysis and observation has been carried out [2,5].

The first-named author is indebted to Professors H. Umemura and H. Aoyama, University of

Tokyo, Tokyo for their continuous guidance and encouragement through the study.

REFERENCES

[1] Newmark, N. N. , and Rosenblueth, E., "Fundamental of Earthquake Engineering," Prentice

Hall, pp. 229, 1971.

[2] Kubo, T., and Suzuki, N. , "Simulation of Earthquake Ground Motion and Its Application

to Response Analysis," Trans. Archi. Inst. Japan, No. 275, pp. 33-43, January 1979 (in

Japanese).

[3] Arias, A., "A Measure of Earthquake Intensity," in "Seismic Design for Nuclear Power

Plants," ed. R. J. Hansen, MIT Press, pp. 438-483, 1970.

[4] Fukada, Y., "On the Load-Deflection Characteristics of R/C Structure," Proceedings

40th Kanto District Symposium, Archi. Inst. Japan, pp. 121-124 (1969) (in Japanese).

[5] Kubo, T., Suzuki, N. , Aoyama, H. , and Umemura, H. , "Simulation of Synthetic Earthquake

Motions for Response Analyses," Proceedings 49th Kanto District Symposium, Archi. Inst Japan, pp. 197-204 (1978) (in Japanese).

11-21
Fig. 1 Deterministic intensity function.

HCICHINOHE HARBOUR. TOKOCHI-OKI

5.0 10.0 15.0 TIME (SEC.

)

Fig. 2 Time trace of the recorded notion,

the EW coii(>anent, Hachlnohe Harbour.

HQCHINOHE. Slf1U_0TED MOTIONS

300.0 COMPONcNl mm-

-300.0

300.0 COMPONENT EW

r.o Tors-- - -15.0 joTo

I IME (SEC . ) Fig. 3 Tine traces of notions from the Model- ing 1, the EW cocqponent, Hadiinohe Harbour.

hqchin'ohe. simul(^ted motions

300.0 , COni^ONENI E«

-300.0

300.0 COMPONENT E

300.0 COMPONENT E

5.0 10.0 15.0 20.0 TIME (SEC.)

Fig. 4 Tine traces of notions from the Model- ing II, the EW ccBiponent, bachinohe Harbour

11-22
11-23
COMPARISON OF VERTICAL COMPONENTS OF STRONG-MOTION

ACCELEROGRAMS FOR WESTERN UNITED STATES AND JAPAN

Tatsuo Uwabe

Port and Harbour Research Institute

Ministry of Transport -

ABSTRACT ...

Characteristics of the vertical component of 187 strong-motion accelerograms recorded

in the Western United States were studied. According to the analysis of the ratio of maximum

vertical acceleration to maximum horizontal acceleration, the average ratio of vertical peaks

to horizontal peaks is 0.48. Significant correlations between maximum horizontal acceleration, epicentral distance, and earthquake magnitude could not be found. The vertical maximum

acceleration occurs near the occurrence time of the horizontal maximum acceleration. But,

the coincidence of vertical and horizontal peaks is very rare. At the time when the safety

factor against sliding and overturning for gravity-type structures is smallest, the ratio of

the vertical acceleration to the resultant horizontal acceleration is less than one-third. A

comparison was made of the ratios of vertical peaks to horizontal peaks for Western United

States and for 574 strong-motion accelerograms from Japan. The result of this comparison is

as follows: The ratio of vertical peaks to horizontal peaks in the Western United States is

larger than that in the Japanese recordings. The difference between the accelerograms of the

two countries is believed to be caused by dissimilar instruments, dissimilar installation

conditions, and differences in focal depth and site conditions. This difference should be

kept in mind when the digitized accelerograms of both countries are used together.

KEYWORDS: Earthquake engineering; earthquake resistant structures; earthquakes; ground

motion.

11-24
INTRODUCTION

Strong-motion accelerograms consist of three perpendicular components of ground

acceleration: vertical and two horizontal. It is generally recognized that vertical

acceleration time histories usually have smaller maximum amplitudes and higher frequency

components as compared with horizontal acceleration time histories. Hence, the effects of

vertical accelerations on stability of civil engineering structures and buildings are often

relatively unimportant. However, case studies of earthquake damage records indicate that

the vertical motion may have severe influence on the stability of a soil embankment or piled

blocks. [1] Some accelerograms whose vertical components were larger than their horizontal

components have been recorded. It is, therefore, important to investigate the characteristics of vertical components of accelerograms recorded during past earthquakes.

In order to collect more reliable data for the seismic design, engineering features of

vertical motions including peak amplitude, ratios of vertical peaks to horizontal peaks and

predominant frequency were investigated in the study by analysis of 187 strong-motion accelerograms recorded in the western United States. Difference in occurrence time between the

vertical and horizontal peaks also were analyzed.

As the number of accelerograms is not sufficient, accelerograms recorded in foreign

countries are used. But there are differences between accelerograms of two countries.

Therefore, it is necessary to compare accelerograms of two countries before these data are

used together. From this viewpoint, a comparison was made of the ratios of vertical peaks

to horizontal peaks for the western United States and for 574 strong-motion accelerograms

from Japan.

STRONG-MOTION DATA

The strong-motion accelerograms used in this study were derived from the "Catalogue of

Strong-Motion Earthquake Records, Volume I, Western United States, 1933-1971." [2] These

accelerograms have been extracted from the Volume II tapes [3] which contain corrected accelerograms uniformly processed by the California Institute of Technology. [4] The accelerograms have been band-pass filtered to retain only those components between 0.07 Hz (in case

of 70- and 35-mm film records, the cutoff frequency of 0.125 Hz [5]) and 25 Hz.

11-25
These data were obtained from 49 earthquakes whose magnitude range from 3.3 to 7.7. The

data contain 187 records recorded on the ground or in the basements of buildings. Of 187

records, 90 percent correspond to the magnitude range of 5.0 to 6.9. Eighty-one percent of

the data are from shallow focal depth (less than 20 km) earthquakes. Fifty-nine percent of

the data were recorded on "alluvium" sites, 24 percent of the "intermediate" sites, and 11

percent on the "hard rock" sites. [6] Sixty-five percent of the records are for epicentral

distances less than 50 km, 18 percent between 50 and 100 km, and 17 percent are more than

100 km.

'

Six different types of strong-motion accelerograms that have a variety of natural

periods, sensitivities, and recording media are described in Table 1. [7,8] Some of these

are or have been used in the United States, and some are used in Japan.

CHARACTER OF VERTICAL ACCELERATION : .

' to ;^ f

Ratio of Maximum Vertical Acceleration to Maximum Horizontal Acceleration

Figure 1 presents the maximum vertical accelerations (^^^^^^ °^ western United States

strong-motion data plotted versus the maximum horizontal accelerations (Aj^max^ ^ log-log

grid. Circles denote the San Fernando (1971) earthquake record, and dots indicate other

strong-motion data for the period in 1931-1971. In this figure, ratios of the vertical peaks

to the horizontal peaks (^^^-k^ ^hvaax^ also indicated by straight lines. The average of

the ratios is 0.48, and the standard deviation is 0.19.

The maximum horizontal acceleration in Figure 1 means the larger one of the two horizontal components, but the real horizontal acceleration at every moment is the resultant of the

two components, and the maximum resultant acceleration in the horizontal direction is always

larger than the largest horizontal acceleration in a single component. Therefore, the ratio

Nrmax^^hmax w^l<^ become smaller, if the maximum horizontal resultant acceleration were used

instead of the maximum horizontal acceleration.

Factors Affecting Ratio of Maximum Vertical to Horizontal Acceleration

According to Figure 1, some of the data show large ratios of A^^^^/ Aj^j^^^^ . If it can be

determined what factors tend to cause large ratios, this may be useful in the development of

site-specific ground motion parameters for earthquake-resistant design. The influences of

11-26
various factors Including horizontal acceleration, epicentral distance, earthquake magnitude,

subsoil condition, and focal mechanism on the ratio ^^^Qx^^hmax ^^^^ examined on an empirical

basis .

• -

The average vertical to horizontal acceleration ratio for records whose horizontal peaks

are less than 50 gal is 0.51. When horizontal peak ranges 50 to 100 gal, the average ratio

is 0.49. When the horizontal peak is more than 100 gal, the average ratio is 0.46. From the

trend, it can be seen that the average ratio decreases slightly as the maximum horizontal

acceleration increases. Significant correlations between the ratio of vertical to horizontal

peak and epicentral distance, earthquake magnitude, subsoil condition or focal mechanism

could not be found.

Frequency Characteristics of Vertical Motion

It is generally recognized that the vertical motion contains more relatively high

frequency components. The predominant frequency is taken as the frequency at which the peak

amplitude of the Fourier Spectrum occurs. The source of the Fourier Spectra used in this

study is Volume IV of the series "Analysis of Strong-Motion Earthquake Accelerograms" published by the Earthquake Engineering Research Laboratory of the California Institute of

Technology. [9] The predominant frequencies of the vertical accelerations were plotted

against those of the horizontal accelerations in Figure 2. The predominant frequencies of

the vertical accelerations are less than 12 Hz and those for horizontal accelerations are

nearly all less than 7 Hz. The predominant frequencies of the vertical acceleration are

not always larger than that of the horizontal acceleration, but a few predominant frequencies of vertical accelerations show relative high values.

Figure 3 presents the predominant frequencies of vertical acceleration plotted against

epicentral distance. Small dots show the data recorded on the "alluvium" sites, circles

denote data on "hard rock" sites, and large dots show "intermediate" sites. The result is

that the upper bound of the predominant frequency decreases with the increase of the epicentral distance.

Difference In Occurrence Times Between Vertical Peaks and Horizontal Peaks

There are few strong-motion accelerograms whose maximum vertical and horizontal component occur at the same time. Therefore, discussion based on the maximum values of accelerograms treat the extreme cases. This is on the safe side from a design viewpoint. But, for

11-27
rational earthquake-resistant design, it is desirable to know the ratio of vertical to horizontal component at the time when the structures become most unstable during earthquakes.

Figure 4 presents the frequency distribution of the difference in occurrence time

(units; seconds) between vertical and horizontal motion. The positive sign denotes that

the maximum vertical accelerations occur after the maximum horizontal accelerations, and the

negative sign means opposite. This figure shows that although the vertical peaks occur near

the occurrence time of the horizontal peaks, the coincidence of vertical and horizontal

maximum acceleration is very rare.

From the viewpoint of the seismic design, it is desirable to know the ratio of vertical

to horizontal components at the time when the structures become most unstable during the

earthquakes. Consider the stability of a rigid block on a solid level plane analyzed by a

seismic coefficient method. A resultant force due to gravity and seismic force and its

direction are expressed by the following equations.

R = mg ^ (1 ± k^)2 + V.^ (1)

K = tane = ^— (2)

1 ± ^

where

R = instantaneous resultant force acting on the block

m = mass of the block

g = acceleration of gravity

k^ = instantaneous seismic coefficient in the vertical direction

kj^ = instantaneous seismic coefficient in the horizontal direction

K = instantaneous resultant seismic coefficient

0 = angle between directions of resultant force and gravity

The block becomes more unstable against sliding and overturning with increase of the

resultant seismic coefficient which tends to maximize as k^^ increases and as k^ becomes more

negative. From the digitized records, the resultant accelerations (Aj.^^) of the two components in the horizontal direction are calculated and the resultant seismic coefficients at

each instant are obtained from Eq. 2. Figure 5 presents, for the 187 western United States

records, the vertical acceleration (A^) and the horizontal acceleration (Aj.j^) at the time

11-28
when the resultant seismic coefficient reaches a maximum. Usually the maximum resultant

seismic coefficient occurs at the time when the horizontal acceleration is almost maximum,

but the corresponding vertical acceleration at that time is not large. Figure 5 shows that

the ratios A^/A^.^ are always almost less than one-third at this time.

COMPARISON WITH JAPANESE STRONG-MOTION DATA

Figure 6 presents comparison between vertical and horizontal peaks of strong-motion

data in Japan. [10] The data contain 574 records which have been recorded on the ground.

These data, recorded by SMAC-B2 strong-motion accelerographs , are uncorrected. The average

of the ratio Aym^x'^^hmax 0.33 and the standard deviation is 0.18. Comparison of Figures

and 6 indicate that the average ratio from Japanese data is smaller than that from U.S. data

and that the ratios of Japanese data decrease more clearly in accordance with increase of

horizontal peak accelerations than that of U.S. data. The differences between Figures 1 and

6 might be associated with the differences in the installation conditions, dissimilar instru

ments, and differences in focal depth and subsoil conditions. Each of these possible causes

is discussed below.

Installation condition. All Japanese records in Figure 6 were recorded on the ground,

but most of the United States data are from accelerographs installed in the basements of

buildings. It was pointed out by Housner [11] that the motion in the basement floors of the

buildings is almost the same as the motion on the ground without large coupling between the

ground and the structures. Also, in the recent paper by Boore, Oliver, Page, and Joyner

[12], it is pointed out that the peak horizontal acceleration is smaller on the average at

the base of large structures than at the base of small structures. In order to determine

whether there is any difference between the A^^^/Aj^j^^j^ ratios for data obtained on the

ground and on the basement floor in Japanese data, the relevant strong-motion data recorded

in Japan are presented in Figures 7 and 8. Figure 7 shows the records in 1978 of the Izu

Oshima-Kinkai earthquake and Figure 8 presents the records in 1978 of the Miyagi-Ken-Oki

earthquake. [13] Circles denote the records on the ground, and dots show the records on

the basement floors. In case of the 1978 Izu Oshima-Kinkai earthquake, the average ratio of

records on the ground is 0.33, and that of records on the basement floors is 0.45. In the

event of the 1978 Miyagi-Ken-Oki earthquake, the mean ratio of data on the ground is 0.37,

11-29
and that of data on the basement floors is 0.43. Though the data are not sufficient, it

seems that the ^^g^x^^hmax ^^^tio of the records on the ground is smaller than that of the

record on the basement floors.

Type of instruments. The Japanese data in Figure 6 were recorded by SMAC-B2 accelerographs. This SMAC-B2 type of instrument has different frequency response characteristics

than the accelerographs used in the United States. Table 1 shows the characteristics of the

accelerographs in the United States and SMAC-B2 type of instruments. [7,8] The frequency

response characteristics of SMAC-B2 accelerographs shown in Figure 9 indicate that this type

of instrument has diminished response at frequencies above 2 Hz. For example, the amplification associated with the frequency of 10 Hz is one-third. On the other hand, the data in

the United States are corrected in the frequency range of 0.07 (or 0.125) to 25 Hz. This

fact is important, because peak ground accelerations are often associated with high frequency

components of ground motion. A correction program for SMAC-B2 records was reported recently

by lai, Kurata, Tsuchida, and Hayashi [14] but has not been used on the data examined here,

because it is not the main object of this report to investigate. To further examine the

effects of different frequency characteristics, the United States accelerograms were converted to SMAC-B2 type accelerograms, using the filter in Figure 9 and fast Fourier Transform

(FFT) method. The filter is

H(f) = _i (3)

1 -

(-f-) + 2h^ (X) i

where

H(f) = filter

f = frequency (Hz)

f = natural frequency (= > 0.14 Hz) -'

h^ = damping (= 1.0)

The accelerograms transformed to frequency domain by FFT were filtered through Eq . 3, and

were transformed to time domain. Figure 10 presents peaks for the converted accelerograms

plotted against those for original United States accelerograms. Dots show the maximum horizontal acceleration and circles show the maximum vertical acceleration. The vertical peaks

in the U.S. records are attenuated more by filtering through Eq. 3 than are the horizontal

11-30
peaks. The post-filtering maximum vertical accelerations are plotted against the maximum

horizontal accelerations in Figure 11. The average ratio is 0.43, and the standard deviation

is 0.17. The comparison between Figure 1 and 11 indicates that the ratios ^^^ax^^hmax

decrease slightly as a result of filtering through Eq. 3. In the Japanese data in Figure 6,

there is a trend of decrease A^gx^^hmax ^hmax increases. The average ratio for postfiltering records in Figure 11 is 0.48, when the horizontal peak is less than 50 gal. When

the horizontal peak ranges from 50 to 100 gal, the average ratio is 0.42. When the horizontal peak is more than 100 gal, the average ratio is 0.39. The trend of decrease '^vjj,ax^^hmax

with increase of A^^jj^^^^ in Figure 11 is clearer than in Figure 1.

Focal depth. The focal depths of the records in the western United States, as described

in the section on strong-motion data, are shallow; usually less than 20 km. On the other

hand, the focal depths of earthquakes in Japan are deep, usually more than 20 km. Figure 12

shows ^yf^^x^^hmax plotted versus the focal depth. Circles show the records for Japan, and

dots show filtered acceleration data for the western United States. The line in this figure

connects the average ratio of each focal depth. The trend is that the ratio decreases as

focal depth increases.

Subsoil condition. Figure 13 presents the horizontal and vertical peak acceleration

data plotted versus epicentral distance. The left part of Figure 13 is the horizontal acceleration, and the right part is the vertical acceleration. Dots show the strong-motion data

recorded on the alluvium sites in the San Fernando earthquake, and circles show the data

obtained in Japanese ports from the earthquakes whose magnitude range from 6.4 to 6.7.

Figure 13 indicates that the horizontal accelerations recorded in Japanese ports whose subsoils consist of very soft layers are larger than those in the San Fernando earthquake, and

that the difference between the vertical accelerations of both data are relatively small.

As the vertical peaks are almost the same, the larger horizontal peaks will come lower

ratios of the vertical to horizontal peaks. This difference of subsoil condition between

the two countries might be one of the reasons why the A^^^/Aj^^^^ ratio of the data in

Japan is smaller than that of the records in the western United States.

CONCLUSIONS

From the analysis of vertical components

United States, the following conclusions were

of strongdrawn :

;-motion accelerograms in the western
a) The average ratio of the maximum vertical acceleration to horizontal acceleration

is about one-half.

b) Significant correlations between this ratio and maximum horizontal acceleration,

epicentral distance, or earthquake magnitude could not be found.

c) The predominant frequency of the vertical component is less than 12 Hz. Significant

relationships could not be found between the predominant frequencies of vertical and horizontal components.

d) The predominant frequency of the vertical component decreases with increasing

distance.

e) The vertical maximum acceleration occurs near the occurrence time of the horizontal

maximum acceleration. But, the exact coincidence of vertical and horizontal peaks is very

rare.

f) With respect to sliding and overturning for gravity-type structures, at the time

when the angle of the resultant force vector with the vertical is largest, the ratio of the

vertical acceleration to the resultant horizontal acceleration is almost always less than

one-third.

The result of a comparison of ratios of vertical peaks to horizontal peaks for western

United States and Japan is as follows:

a) The ratio of vertical peaks to horizontal peaks in accelerograms for the western

United States is larger than that in the Japanese recordings.

b) The difference between the data from the two countries is consistent with variations to be expected because of dissimilar instruments, dissimilar installation conditions,

and differences in focal depth and site conditions.

ACKNOWLEDGMENT

This report was prepared during the author's visiting research at the U.S. Army Engineer

Waterways Experiment Station (WES) sponsored by the Office, Chief of Engineers (OCE). It is

part of ongoing work at WES in Civil Works Investigation Studies 31246 entitled "Dynamic

Stresses and Permanent Deformations of Earth Structures in Response to Seismic Loadings,"

which is monitored for OCE by Mr. Ralph W. Beene. His visiting research at WES was recognized as part of the research personnel exchange program of U.S. -Japan Panel on Wind and

Seismic Effects pursuant to the resolution of the tenth meeting of the panel.

11-32
The author wishes to express his thanks to Dr. Paul F. Hadala, Chief of the Earthquake

Engineering and Geophysics Division (EE&GD), Dr. Ellis L. Krinitzsky, Chief, Engineering

Geology Research Facility, and Mr. Frank K. Chang and Dr. Arley G. Franklin of EE&GD who

reviewed the manuscript and offered many useful suggestions. He also expresses his appreciation to Dr. William F. Marcuson, III, of EE&GD for his efforts on this visiting research

in WES.

REFERENCES

[1] Fujino, Y. , Sasaki, Y. , and Hakuno, M. , "Slip of a Friction-Controlled Mass Excited by

Earthquake Motions," Bulletin of Earthquake Research Institute, University of Tokyo,

53 (1978).

[2] Chang, F. K., Miscellaneous Paper S-73-1, Report 9, U.S. Army Engineer Waterways

Experiment Station, CE , Vicksburg, Mississippi (1978).

[3] Trifunac, M. D. , and Lee, V. E., "Routine Computer Processing of Strong-Motion Accelerograms," EERL 73-03, California Institute of Technology (1973).

[4] Hudson, D. E., Brady, A. G. , Trifunac, M. D. , and V jayaraghavan. A., "Strong-Motion

Earthquake Accelerograms, Corrected Accelerograms, and Integrated Ground Velocity and

Displacement Curves, Vol. II, Part A," EERL 71-50, California Institute of Technology

(1971).

[5] Trifunac, M. D. , Brady, A. G. , and Hudson, D. E., "Strong-Motion Earthquake Accelerograms, Corrected Accelerograms, and Integrated Ground Velocity and Displacement Curves,

Vol, II, Part G," EERL 73-52, California Institute of Technology (1973).

[6] Trifunac, M. D. , and Brady, A. G. , "On the Correlation of Seismic Intensity Scales with

the Peaks of Recorded Strong Ground Motion," Bulletin of the Seismological Society of

America, 65 (1975).

[7] Halverson, H. T. , "The SMA-1 Strong-Motion Accelerograph, " Fourth Symposium on Earthquake Engineering, Roorkee, India (1970).

[8] Hudson, D. E., "Ground-Motion Measurement," Earthquake Engineering (Prentice-Hall,

Inc., New Jersey, 1970), pp. 107-125.

[9] Hudson, D. E., Trifunac, M. D., Udwadia, F. E., V jayaraghavan, A., and Brady, A. G.

,

"Strong-Motion Earthquake Accelerograms, Fourier Amplitude Spectra Volume IV, Part

A-Y," Earthquake Engineering Research Laboratory, California Institute of Technology.

[10] Uwabe, T. , Noda, S., Kurata, E., and Hayashi, S., "Characteristics of Vertical

Component of Strong-Motion Accelerograms," Ninth Joint Meeting of U.S. -Japan Panel

on Wind and Seismic Effects, UJNR (1977).

[11] Housner, G. W., "Interaction of Building and Ground During an Earthquake," Bulletin of

the Seismological Society of America, 47, 3 (1957).

[12] Boore, D. M., Oliver, III, A. A., Page, R. A., Joyner, W. B., "Estimation of Ground

Motion Parameters," U.S. Geological Survey, 78-509 (1978).

[13] Strong-Motion Earthquake Observation Council, "January 14, 1978 Izu Oshima-Kinkai

Earthquake," No. 13, "June 12, 1978 Miyagi-Ken-Oki Earthquake" No. 14, National

Research Center for Disaster Prevention, Science and Technology Agency (1978).

11-33
[14] lai, S., Kurata, E. , Tsuchida, H. , and Hayashi, S., "Integration of Strong-Motion

Accelerograms," Tenth Joint Meeting, U.S. -Japan Panel on Wind and Seismic Effects

UJNR ( 1978)

.

11-34
Table 1 Characteristics of Strong-motion Accelerographs

. Cha racteristics

Period, sec Sensi tivi ty nun/0.1 g

Recording range, g's Damping

% Critical Damping Mechanism

Recording Speed, om/sec

Recording Medium

Recording Drive Recording Duration

Time Marking

Size, in. (cm)

Weight, lb

(kg)

USCGS

Standard

0.013-0.085

5.5-19.

7

0.01-1.0

60

Magnetic

1

Photo Paper

DC motor

1.25 min

2/sec

13x20xii5 (33x51x111)

135

(61)

Teledyne

AR-240 Teledyne

RFT-250 TeledyneRMT-280

0.055-0.065

5.0-7.5

0.01-1.0

55-65

Elect romagnetic

2

Photo Paper

DC motor

7 sec after last strong motion

2/sec at + 1%

11x16x16

(36x41x41)

60

(27)

0.05

1.9

0.01-1.0

60

Electromagnetic

1

70 mm (Type II Perforated)

DC motor

5 sec after last strong motion

2 sec at + 27.

8-3/4xl0-l/2x

19-1/2 (22x27x

50) 30

(14)

0.05

+ 200 cps FM deviation/+ 1 g

0.01-1.0

60

Electromagnetic

3-3/4 in. /sec (9.53 cm/sec)

1/4-in. Magnetic

Tape

DC motor

7 sec after last strong motion

2/sec at + 27.

9x15x19

(23x38x48)

42 (19)

Klnemetrics

SMA-1

0.4

1.9

0.01-1.0

60

Magnetic

1

70 mm film

AC motor

25 mln

2/sec + 1%

8xRxl4

(20x10x36)

25

(11)

New Ze^.land MO 2

0.03

1.5 Horizontal

0.01-1.0

60

Oil Paddle

1.5

35 mm film (Unperforated)

DC motor

4 7 (70) sec

Trace interrupt

5 and 50 cps 0.1%

7x7x17

(18x18x43)

20

(9)

Akashl

SMAC-B2

0.14

6.5

0.006-0.5

100

Air Piston

1

Waxed Paper

Hand-Wound Spring

3 mln

1/sec

15x21x21

(54x54x37)

220

(100)

Figure 1. Maximum Vertical Acceleration Versus Maximum

Horizontal Acceleration

1 S 10 50 100 500 1000 MAXIMUM HORIZONTAL ACCELERATION lA^^^,, CM/SEC^) 11-35
0 2 4 6 6 10 12 14 PREDOMINANT FREQUENCY OF HORIZONTAL ACCELERATION. H;

Figure 2. Predominant Frequency of Vertical Acceleration and Horizontal Acceleration

1 1 1 1 1 1 1 II. 1 1 1 1 1 1 1

LEGEND

• ALLUVIUM

0 HARD ROCK

• INTERMEDIATE

1 1 1 1

0

o

o o

•

•

o

••

o

o•

1 1

o

o

1 1 1 1

. °

" • '

"• • •• . o •

.•°o-..-

• 1 \ •r

'

•

•

>

1

•

r 1 oi III!

2 5 10 50 100 500 1000 EPICENTRAL DISTANCE, km

Figure 3. Predominant Frequency of Vertical Acceleration Versus Epicentral Distance 11-36
+ T »—

H

[w~'\^/^yY^1/v. VERTICAL ACCELERATION

^^-XV] P\y^ HORIZONTAL ACCELERATION

\\}\^ VE

1 T = DIFFERENCE IN -OSEC

RTICAL ACCELERATION OCCURRENCE TIME. SEC

m n 11

1

1 -10 -9-8-7-6-5-4-3-2-1 0 I 2 3 4 5 6 7 9 9 10 DIFFERENCE IN OCCURRENCE TIME. SEC

Figure 4. Frequency Distribution of Difference in Occurrence Time Between Vertical and Horizontal Peaks
Figure 6. Maximum Vertical Acceleration Versus Maximum Horizontal

Acceleration for Japanese Recordings

11-38
Figure 7. Maximum Vertical Acceleration Versus Maximum Horizontal

Acceleration for Japanese Recordings

11-39
1000

500

u 100

1978 MIYA

JUN 12,

1 I.I

GI-KEN-OKI

1978 - MAG

1 . 1 EARTHQUAKE

JITUDE: 7.4

1 1 1 1 1 1 1 1 1 I Z

-

LEGEND

O GROUND

• BASEMENT OR 1 ^' FLOOR / • / O

-

-

-

/ O /

/ / ° / • /

°

>

O

3 •

/ •

• /o "a 5

• *Ju °°

'

/»c6» o CO

^ » •—« oo« V

1 1 1

/\

1 1 IK

f o«o o • • <• •

1 1 1 1 1 1 1

1 1 1 1 1 1 1

1 5 10 50 100 500 1000 MAXIMUM HORIZONTAL ACCELERATION lA^jj^^^. CM/SEC^I

Figure 8. Maximum Vertical Acceleration Versus Maximum Horizontal

Acceleration for 1978 Miyagi-Ken-Oki Earthquake

11-40
FREQU ENCY , HZ

0 1 1 10

0.8

0.6

0.2

0.1 1 FREQU ENCY.

Figure 9. Characteristics of SMAC-B2 Accelerograph

Figure 10. Calculated SMAC Acceleration Versus Observed Acceleration

11-41
1000

<' 100

1J 50

1 1 1 MM 1 1 1 1 1 1

1

1 1 1 1 1 \y.

-

-

• / /

/ * /

- -

/ • *« / •

•

.'V. '

'

•

*/ *

* / r ' ;

/ • •

-7 V

1 1 1

1 1 1 h 1 1 1 1 1 I 1 1 1 1 1 1 1 1

1 5 10 50 100 500 1000 CALCULATED SMAC - HORI ZONTAL ACCELERATION lA. .. CM/SEC ^1

Figure 11. Calculated SMAC Vertical Acceleration Versus Calculated

SMAC Horizontal Acceleration

11-42
_L _L

STRONG MOTION EARTHQUAKE RECORDS

O JAPANESE PORT, 1963-1974 WESTERN U.S., 1933-1971 (CONVERTED SMAC ACCELERATION)

_L _L _L

20 60 FOCAL DEPTH, hm Figure 12. Ratio of Vertical Peak to Horizontal Peak Versus Focal Depth

11-43
I I I I "I I r I M u

SAN FERNANDO EARTHQUAKE

M = 6.6 ALLUVIUM

o EARTHQUAKE RECORDS IN JAPANESE PORT

M = 6.4 ~ 6.7

o t

o o o

HORIZONTAL ACCELERATION

J I

I I I I I I I L

VERTICAL ACCELERATION

I I I I I

I I I I I I 50 100 200 50 100 EPICENTRAL DISTANCE, km

500 1000

Figure 13. Maximum Acceleration Versus Epicentral Distance for Western United States

and Japan

11-44
EARTHQUAKE OBSERVATION SYSTEMS

IN AND AROUND STRUCTURES IN JAPAN

Kiyoshi Nakano

Yoshikazu Kitagawa

Building Research Institute

Ministry of Construction

ABSTRACT

Ever since the observation system for underground earthquake motions was undertaken in

1934, efforts have been made to obtain exact knowledge of the characteristics of ground

motions, the dynamic interaction behavior between building and subsoil, and the earthquake

motions input to structures.

The number of observation sites has increased year by year. There are now more than 200

in number, and much valuable data has been accumulated. In this paper the state-of-the-art

of underground earthquake observation in Japan is reviewed.

KEYWORDS: Earthquake observation systems; simulation models; spectral analysis; transfer

function; underground amplitudes.

11-45
1 . INTRODUCTION

In Japan, there is an extensive history of underground earthquake observation. In 1934,

earthquake observations were performed from the bottom of wells in the center of Tokyo by

means of mechanical type accelerographs (Ishimoto's type) to make comparisons with observations on the ground surface. In the 1950' s, the advantages of earthquake observations at the

bottom of bore-holes were emphasized, and a bore-hole seismometer of the electromagnetic type

was successfully developed.

Around 1965, with the advent of high-rise buildings, long-span bridges, and nuclear

power facilities, a remarkable increase in research and investigation has taken place with

special reference to earthquake observation projects. The observation of earthquake motions

in and around a structure began about the same time to yield data from the analyses of soilstructure interaction problems. Instrumentation has improved year by year, underground

observation projects have increased in number, and there are now more than 200 in Japan.

2. THE PRESENT STATE OF EARTHQUAKE OBSERVATION '

In order to make inquiry about the state-of-the-art on the observation of underground

earthquake motions, the authors obtained information from universities, institutes, construction companies, the Port and Harbor Research Institute, the Center of the Disaster Prevention

Agency, electric power companies, the P.W.R.I., and other organizations. The earthquake

observation systems of the organizations mentioned above are classified into two types;

type A is a set of observations of a structure and its surrounding subsoil, while type B is a

set of observations of the subsoil only.

The locations and instrumentation of some observation stations obtained from the inquiry

are listed in Table 1 (A) and (B). [1] In Table 1 (A), S means an underground observation

point separate from a construction. These observation systems are also reported in the

publications of some organizations such as the Architectural Institute of Japan.

2.1 Purpose of Earthquake Observation

The purposes for performing underground observations in the field of earthquake

engineering are many and vary according to the interest of the individual organizations which

carry out each project. A review of various projects is still in progress, but they may be

summarized now as follows:

11-46
1) to Investigate the characteristics of earthquake motions,

2) to investigate changes in depth of wave shape, and maximum acceleration in relation

to ground conditions,

3) to investigate input motions to structures and their spectral characteristics,

4) to investigate the interaction effects between ground motions and the responses of

super-structures, piles and underground structures,

5) to investigate the characteristics of earthquake motion within rock or rock-material

which may be regarded as the so-called base.

6) to investigate the propagation characteristics of earthquake motions (amplitude and

phase) along a line by installing a horizontal array of seismometers on the ground surface or

underground.

2.2 Earthquake Observation System

Object of Observation — Earthquake observations within soft soil deposits are quite

large in numbers. This is due to the majority of earthquake observation projects being

planned in association with structures in urban areas which are developed on areas of soft

soil deposits. As seen in Table 1 (A) and (B), the earthquake observation systems are classified into two types. The systems in type A amount to approximately half the total number,

indicating that the soil-structure interaction problem is one of great interest. At the same

time the underground seismometers are usually installed along double vertical lines; one just

below the structure and the other apart from the structure by a distance which is greater by

approximately twice the structure's width, which may be considered enough to eliminate influences from the presence of the structure. The systems in type B are installed in order to

investigate the characteristics of ground motions and of underground structures and are

performed together with the measurements of micro-tremors and the seismic prospecting test

in the usual manner.

A percentage breakdown of the number of type A earthquake observations which have been

recorded is as follows: 51% - university, apartment, office, and hotel buildings; 11% -

electric power facilities; 10% - bridges and banks; 10% - dams; 9% - models (reactors and

other rigid structures); 9% - tanks, towers, and miscellaneous.

Location of Underground Seismometer — The deepest seismometer in type A and B is

usually installed at the upper boundary of the base layer (rock or rock-like formation) with

11-47
N-value of the penetration test more than 50 or shear wave velocity more than 400-800

m/sec. Table 2 shows the statistics of the number of seismometers in the projects of Types

(A) and (B), and Table 3 shows the statistics of the greatest depth of observation. As for

the distribution of the depths of underground observation, the embedded depths of seismometers range as follows: at ground surface (included rock), 5%; at a depth between ground

surface and 20 m below the surface, 50%; at a depth of between 20 m and 60 m, 30%; at a depth

of between 60 m and 100 m, 7%; at a depth deeper than 100 m, 3%; where the maximum depth is

about 3500 m. ' .

'

'

System of Seismometer — In general, a system of electromagnetic seismometers consists

of pickups, amplifier and recorder. Pickups of servo type (32%) and moving coil type (62%)

are usually employed. The former is advantageous in general since the uniform frequency and

phase characteristics can be maintained from 0 up to approximately 300 Hz.

Amplifier — Amplifiers such as a logarithmic amplifier, auto-gain-control amplifier,

binary-gain amplifier, and so on are tried and used in order to enlarge the dynamic range of

records. The auto-gain amplifier is the one most frequently used. ,

Recorder — The recorder of analog magnetic-tape type is replacing the conventional

optical type. The delay device utilizing integrated circuit memory is becoming more popular

year by year. In some delay devices, the digitized record which is stored in integrated circuit memory not only actuates the earthquake observation system, but also automatically

establishes adequate gain of the amplifier after the actuation. The delay time is usually

in the range of 3 to 5 seconds.

For any of the observation systems a crystal clock and a simple voice recorder such as

those used in public radio broadcasting are employed as time-mark devices to make timeinterconnections between the results of observations from the different recorders of an

earthquake motion.

Consequently, the uniform frequency characteristics of the overall system is usually

maintained from 0.3 up to 30 Hz with long span stability of instruments, and the measurements

of acceleration are overwhelmingly large in number. Table 4 shows the statistics of objective quantities to be aimed at for measurement. It should be noted that great importance is

attached to acceleration as the basis of earthquake engineering. However simultaneous

observation of acceleration, velocity and displacement have tended to increase in number

recently.

11-48
3. EXAMPLE OF OBSERVATION

Among the observation systems listed in Table 1 (A) and (B), the B.R.I. , P. W.R.I, of the

Ministry of Construction, and the University have carried out observation at several places,

for example at numbers 6, 48, 49 and 42 in type A, and at number 16 in type B. The details

of the earthquake observations of these five places are described in the following

paragraphs.

3.1 The University Building (No. 6) in Type A [2]

The subsoil below the building is Kanto loam layer, alternative layers of fine and sandy

loam, alternative layers of fine sand and silty or clayey sand, and a gravel layer, in that

order from the surface. The water-level is 8.8 m below the ground surface. The soil profile

and penetration test results (N-values), and the result of the seismic prospecting test are

shown in Figure 1

.

The building is a six-story reinforced concrete structure with a one basement level. It

is 15.8 m X 84 m in plan, as shown in Figure 2. The west part of the building was first constructed during 1962-1963 with three additions by 1970. Each construction stage is indicated

in Figure 2. The main lateral resisting elements of this building are the walled-type reinforced concrete frames, which have a walled girder (spandrel) and a walled column, in the

longitudinal direction, and reinforced concrete shear walls in the transverse direction. In

both directions there are slightly reinforced concrete and concrete block partition walls

that are considered to be non-structural members. The building is supported by individual

footings without piles at 5 m below the level.

The fundamental natural period of the building determined by forced vibration tests is

0.29 sec in the longitudinal direction, and 0.32 sec in the transverse direction. The rocking vibration is predominant in the transverse direction.

Electromagnetic seismometers were used to measure the earthquake motion under, around,

and inside the building. The locations of the seismometers are shown in Figure 3. Seismometers A, B, C, D and E are placed along the vertical line from the top of the building to

the soil layer 42 m below the surface. Seismometers F, G and H are arranged along the vertical line also, and are 60 m south of the above group in a horizontal direction. Seismometer

K is arranged at a depth of 82 m below ground level. Each underground seismometer was

11-49
installed at the bottom of a bore-hole, which was refilled after the installation so that

seismic waves would not be affected by the bore-hole. Examples of the acceleration records

are shown in Figure 4.

' "

. O i f '

'

3.2 The Test Model Building (No. 48) in Type A [3] . ,

The subsoil is composed of loam, clay soil and fine sand. The soil profile, the penetration test results (N-values), and the result of the seismic prospecting test are shown in

Figure 5.

The foundation of the test model building is reinforced concrete with a mat foundation.

The super-structure of the model building consists of a concrete slab of 1.0 m which is

supported by four steel pipe columns. The model building is 2.0 m high, and 4.0 m in plan,

as shown in Figure 6.

The fundamental natural frequency of the soil-foundation-building system obtained by the

vibration test is 3.41 Hz. The vibrator was set on the foundation during the test. It

should be noted that vibration due to the relative displacement is predominant.

The location of the seismometers on the model and in the soil is shown in Figure 7. The

pendulum of the seismometers is heavily damped (h t 10) and the output voltage which has a

natural frequency of 3.5 Hz is proportional to the acceleration. Typical examples of waves

in the soil-foundation building are shown in Figure 8.

3.3 The Apartment House (No. 49) in Type A [4]

The subsoil is composed of soft alluvial silt and clay deposits, the depth of which is

about 25 meters. The soil profile, the number of blows in the standard penetration tests

(N-values), and the velocity of the S waves are shown in Figure 9. - :. :

The building is an eleven-story apartment house made of steel reinforced concrete, as

shown in Figure 10. It is supported on reinforced concrete bored piles. This type of construction is called the Benoto method. The main structural components of the building are a

walled frame in a transverse direction and a rigid frame in a longitudinal direction.

The forced vibration tests confirm that the fundamental natural frequencies of the

structure are 2.2 Hz in a transverse direction, and 3.6 Hz in a longitudinal direction,

respectively.

Servo-mechanism type accelerometers are installed at two places. The accelerometers

are set in the building pile and soil at one place, and in the soil at the other place. The

11-50
four accelerometers , one placed at the top of the building and three in the soil, are connected to the double integrator in order to obtain the record of displacement. The earth

pressure gauges and pore water pressure gauges of a differential transformer type are

installed both under and beside the foundation of the building. The location of the instruments is shown in Figure 11. The frequency characteristics of the accelerometer , of which

the undamped natural frequency is 5.0 Hz, is almost flat (more than 95% relative sensitivity)

in the overall range from 0.3 to 30 Hz. Examples of acceleration records are shown in

Figure 12.

3.4 The Spherical Tank (No. 42) in Type A [5]

The soil profile and the S-wave travel-time curve measured at the boring hole for

installation of the underground seismometers are shown in Figure 13.

The tank is a spherical shell supported by 12 steel pipes (700 x 9) with a pair of

steel-bar bracings (2 x 800) between them, and the heights to the crown and the circumference

and the diameter of the shell are 20.4 m, 11.5 m and 17.9 m, respectively, as shown in Figure

14.

The electromagnetic accelerometers (Tuss type, f = 5 Hz , h = 0.6) are installed at 5

observation points along two vertical lines, as shown in Figure 14. An example of the acceleration records is shown in Figure 15.

3.5 The Ukishima Park (No. 16) in Type B [6] '

'

The earthquake observations at Ukishima Park represent one of the four underground

observation projects carried out at four stations around Tokyo Bay. The subsoil is classified into several layers as follows; reclaimed soil, silt, clay and sand. The soil profiles

and the penetration test results are shown in Figure 16 together with the location of the

seismometers.

Four seismometers have been installed, one on the ground surface, one at a depth of 27 m

below the ground surface, one at a depth of 67 m, and one at a depth of 127 m, respectively.

Each accelerometer, whose pendulum is heavily damped (h v 10) with a natural frequency of

4.0 Hz, has three components; one in the vertical direction, and two in the horizontal

( irection. An example .if the records triggered is shown in Figure 17. . ;

11-51
At some sites mentioned above, the digitized earthquake observation data have been

published by the B.R.I, and the P. W.R.I. [7] of the Ministry of Construction, Japanese

Government. r, ..

On the other hand the analyses are widely made on the records observed for moderate

earthquake motions in accordance with the purpose of observation. The most representative

examples of analyses are as follows:

Distribution of Underground Amplitude — The amplification of layered subsoil, in which

the depth to the layered boundary is changed, is one of the basic research purposes, and is

investigated with the quantities of the maximum acceleration, the root mean square amplitude

and the amplitude after band-limited filtering. The maximum amplitudes observed on two vertical lines, i.e., building-subsoil line and soil line, are generally normalized by those at

the ground surface or at the base layer, and the average amplitude ratio of maximum accelera

tions with standard deviations is investigated.

Spectral Analysis — Fourier spectra, power spectra and response spectra are usually

used with a smoothing technique such as the window types of Manning, Perzen and so on. The

spectral techniques of the maximum entropy method and auto-regressive method in the information theory also have come into use recently. In addition, the running spectra technique is

frequently used in order to examine time variations of spectra.

Transfer function — The transfer function is calculated as the spectral ratio between

two spectra of different records. Consequently, a considerable amount of variation is seen

in the observed spectral ratios. The fluctuation increase is seen in the spectral ratios

between the top and bottom of the building, between the base of the building and the ground

surface, and between the ground surface and the deeper soil layers. The characteristics of

phase difference between the ground surface and the deeper soil layers have begun to be

analyzed recently. ... , . ,

.

Simulation Models •— The response analysis is generally carried out by means of one

dimensional wave propagation theory, lumped mass model and Finite Element model. Through

comparisons between the observed and the computed values, the validity of the models is

examined. The average equivalent damping coefficient and shear strain of soil deposits are

also estimated with various treatments of models. .

,

11-52
r

• • . r. : 0 ; CONCLUDING REMARKS •

The present state of current earthquake observation projects in Japan have been

reviewed in this paper. The number of observation sites is increasing year by year, and

much valuable data have been accumulated. In relation to the increase of data, the disclosure of the data obtained at any observation site, for cooperation and coordination

between earthquake observation projects would be expected, and the free use of this data

would be fruitful in the progress of the research. However, more systematic, more orderly

planned, and better equipped measurement systems are required. Hereafter possible technical

improvements should be extended widely in underground observations for moderate and strong

earthquake motions as follows: stability and accuracy of instruments; wider range measurements in amplitude and frequencies; more intentional choice of the site; proper overall

array arrangement of instruments; automatization of data processing and so on.

ACKNOWLEDGMENT

In preparing this paper, the authors would like to express their thanks to the

researchers who responded with information about earthquake observation system through the

mail survey. The authors also wish to express their thanks to Dr. T. Tanaka, of the Earthquake Research Institute for many valuable discussions, and to Mr. T. Kajima and

Miss M. Makishima of B.R.I, for providing the illustrations.

REFERENCES

[1] Kitagawa, Y. , "Observation System for Underground Earthquake Motions in Japan," Building Research Institute Research Paper No. 73, Building Research Institute, 1977.

[2] Osawa, Y. , Kitagawa, Y. and Ishida, K. , "Response Analysis of Earthquake Motions

Observed in and around a Reinforced Concrete Building including Building Subsoil

System," Fifth W.C.E.E., 1973.

[3] Osawa Y. , Kitagawa, Y. and Irie, Y., "Evaluation of Various Parameters on Response

Analysis of Earthquake Motions Including Soil Building Systems," Sixth W.C.E.E., 1977.

[4] Ohta, T. , Niwa, M., Ando, H. and Ujiyama, M., "Study on Seismic Responses of Soft

Alluvial Subsoil Layers by Simulation Analysis," Sixth W.C.E.E., 1977.

[5] Minami, T., Osada, K. and Osawa, Y., "Earthquake Observation on a Spherical L.P.G.

Tank and the Surrounding Ground," Proceedings, of Fourth J.E.E.S., 1975.

[6] Iwasaki, T. , Wakabayashi, S. and Tatsuoka, F., "Characteristics of Underground

Seismic Motions at Four Sites around Tokyo Bay," Eighth UJNR, 1976.

11-53
"Report on Digitized Earthquake Accelerograms in a Soil-Structure System," B 1976.

"Strong-Motion Earthquake Records Obtained at Undergrounds," P.W.R.I., 1976.

11-54
Table 1 (A) Earthquake Observation Project in Type A.

Type or

Pui Idine

Type of fonn t rue t 1 on

Type or F'ouiida t i on

0 t'giin 1 za t ion Opening

Date Observal ion Points Objectiv! Quantity

Crav i ty Electric Power

Deve lopmen t Co

.

1959

(J) CL (Rock) AGO. Ve 1 .

Rorkfi 11 Flee trie Power

tiovt, lnpm»nt Co

.

1 '»6 1 D im Top A C C.

Rockfill

a;.202m I15ra Electric Power

Deve lopment Co

.

1961 Dam Toi'

(?) GL (Rock) ACC.

\' e 1

.

HeTCtor 4

Foot ing

RC. S Architectural Inst,

of Japan

RF, BF

;s) GL,0L-llni,-l7m A 0 C,

Da/n Rockfill

l = 355n h^l28iii Electric Power Development Co. FL +'lU9m. 'iftSm. 52'tm

(|) CL (Hock) ACC.

6 Univera 1 ty Bui ldin«

PC

6F

indlvidu;»l Univ. of Tokyo 196I4 RF.'iF. lF.CL-23m,-I*2m 0 GL,CL-23'n.-'*lni. CL-5m,-82m ACC.

7 High school

Fui IcJlng 2FPile (l=13m) Kyoto Univ. 1965

R F 2F IF

(5) GL , GL- Im , - 10m , - 20m V e 1 .

8 S true ture Mold No,l

IF Architectural Inst, of Jap.Tn. Kajima Corp. 19^5 PF.IF.DVF(|) CL.GL-2.2m.-'*.^*<n.-li*.'.m.-l9.7'n ACC.

9 St rue ture

Model No.

2

2Fmat

Architectural Inst, of Japan. Knjima Corp. 1965 PF 2F , 1 F

,

d) 'ct,CL-20m ACC.

10 Unl vers 1 ty

Bui Iding

5F*B1F

pile (l-19m) Tokyo Inst . of Tech. niF © GL-10m.-llOoi.-50i. • ACC.

] 1 Tower

S81m Ministry of Const. 19(^7 r.L»eiiii,CL ® t>L ACC.

) 2 Rigid

S t rue ture Mndel No.

3

PC Architectural _Inst. of Japan, Shireizu const.

BF, IF, DIF @ GL-Jm.-lim.-SOm ACC.

Orf - ce Duilding

S Kajima Corp. HF.23F,13F,D2F (HMAC)

D2F,r.L-80m acc.

Apartment

Bui I ding

SRC

1 5F steel pile Uaseda Univ. 1968 ilF (SHAC),GI,-'ini.-l<l4m Apr

15 Model Bldg. with PC panne I

RC

IF

H steel pile (l=10m)

Nippon Steel Corp., Univ. of Tokyo 19^>8 RF.1F,GL-5"1

Gl-3ra.-l5TM acc.

Model

Bu ilJ ing

RC

Biat Japan Electric Association 1968

RF. 3F,2F. IF A 0 0.

1 7 Fill1 = 1 5V3m h = 38111 Electric Power

Dove lopmen t Co

.

1968 FLtl200m. »1225m, I2'i0n acc,

Ve 1 .

Unl vers i ty Building

SRC

9Fpile (l=12m) Tohoku Univ

.

1969 RF,7F,3F,IF acc.

Bi-rtC tor

rtui id in^

PC fokvn El"ctrir P«v-r Co. 107.1 HF. 5F. Foo t i nK. OL-1'* 31. -2-t 111.-5"'' ACC.

M .'Wl rMllding

Kyjto Univ. 1 97 f> TIF.UF, IF

;; CL- ACC

21 Offlcft Bui Id ing

s40F+B3F

nat Kajima Corp. 1970

1.0F. 10F.20F. 10F,P2F (S-IAC) @ CL-20in,-90oi ACC.

tlniversity Building

RC5F+B1F Nibon Univ. 1970 DIF ® GL-lm.-l'.m.-IOm ACC.

21 Apnrtment

Bullrling

S0te*l H pile (Wl2m)

Chiba Univ.. Nippon S^eeI Corp. 1970

1 1F.6F. lK.OL-J2m.-3')m © CL- In. -12111. -30m ACC.

Reactor

Hui Id ing

RC

5F Blf

n.-i t Toky- Flertric Power Co. 1970 RF.5F.3F.BF Q El,-I,.Jiii.-ri>i..-lim ACC.

11-55
H-. I M t ni;

1'^

PC pile ( 1 = 12iii) Iniv. ^: fukyu, Tdisei Corp. 197 J

. 1 f .01 -'iM, - 1 Lm. -. '.I, (o) CL.GL-'tm,-12m,-2lim

Objective Quantity

Bill Irting

RC

5Fcont inuous Ministry of Const. 1971 RF,l|F,BF,CL-2.5m,GL-10.2m,GL-50m

(D GL-2.5m,-10.2m

CL-2. 5m ACC.

D i s .

27

I lectric rower Facilities

RCteel pile

Chugoku Electric

Powe r Co

.

1971 Turbine frajne GL*8 . Im, -3. 8m

steel pile GL- 1 5 . Im , -23 . 35m © CL-l.0m,-23. 50 ACC.

28 Of rice

6u i 1 r1 1 ng

RClOF steel pile Central Research inst. of Electric Power Industry 1971 RF,6F,1<F.2F,B1F,B3F CL-30m(tip of steel pile) © GL,Gl^200m ACC.

29 Office

Bui Iding

S3F

1 nd 1 vidua

1

Nippon Telegraph A Telephone

Public Co. 1971 RF,3F.1F (SMAC)

CL-5oi.-l5m ACC.

30 Apartment

B'ji Iding

SRC

lliF PC pile Ministry of Const. 1972 lIiF.8F.Ba9epoint © GL-9m,-16ro,-26m,-60ni ACC.

31 Hif^ school building

RC

RC pile(l = 32«') Architectural Inst,

of Japan 1972 RF.lF,GL-32oi ACC.

Vel.

P a rt h RvM

32 Exper iment Dui Id ing

RC5F.B2F Shimizu Const. 1972 RF,lF.B2F,GL-20m ACC.

33 Apartment

Dui Iding

Searth drill pile (l=llm) Shimizu Const. 1972 11F,1F,GL-I0m ACC.

3l< Apartment

Bui Id ing

SRC

lOF pier Kajima Corp. 1973 BF,BF,GL-10n),-20m © GL-2m,-20m ACC.

35 Steam Power Station

RC

5Fsteel pipe pile (l=20in) Chugoku Electric Power Co., Ohbayashi-gumi Ltd. 1973

FL-38m, -55m. -23.75m © GL-73",-70m ACC.

36

Pe.TC t or Building

RC5F.B1F Chugoku Electric Power Co. Kajima Coro. 1973 RF.5F,lF,BF,CL-35m,-70m © GL*0.3in ACC.

37 Office Building

with Tower

Building SRC,15F*B5F Tower S, h^liiOm mat Tokyo Electric Power Co.

Kajimrt Corp. 1973 Tower GL +200ir , +107m

,

Building 15F, B5 ( S.MAC ) ACC,

Vel.

3» S25m

steel pipe pile (l=38m) Shimizu Const. 1973 Top , Foot ing,CL-2m, -18m, -39m © GL-2m, -18m,- 38m, -,71m ACC.

39 Office

Bui Iding

S30 F mat Ohbayashi-gumi Ltd. 1973 RF, 1 5F, 1F,B2F,CL-I7m ACC.

i.0 Office building

SRC 12F+B2F Kiimaga 1 -gumi General Contract Co. 1973 12F,6F,B2F © &L-19.6m ACC.

D i s .

1.1 A| a r L me lit Bui Id ing

RCISfBlF

pier (l=20m) Kajima Corp. 197'< RF,9F,BlF,GL-2llm(tip of pier) © GL-lm,01.-27m ACC.

112 Tank

S

20TI1 steel pipe pile (l=31m)

Univ. of Tokyo 197'i Top, Footing © GL,GL-17.'lm,-'l3.'<m ACC.

•tj Ch 1 mne y

S

2 30m oteel pipe pile (l^liOm)

Tokyo Electric power Co., Mitsubishi Heavy Industries Co.

197'< Cl.*222m, »112m, »0. 5m &L-10m.-31ra.-38m ACC.

D i s .

UU Plant tower

S50m . 30m steel pipe pile (l=ii5*r) Mitsui Petrochemical Co.,

Taisei Corp. 1974

T'<9,T30,F1 ,F2,

GL-10ci,-l4 5m,-85m © GL ACC.

D 1 s.

Founda t ion

of Shaking TuMe

SIFmat

The National Research Centt.

for Disaster Prevention

Science A Tech. Agency

197'i BF © GL,GL-10m,CL-JtOm ACC.

Iiic:h school

RC

3F

rr pi 1 <- Mitsui Const. iPV't RF.IF © rL.-lm,-7. '.m,-6f.m ACC.

i?

L.ih.j ra t ory

. ' 1 1 I '1 - SRC TK'f'F Central Research Inst, of

tl ' rir P.wr Iii lM.il. y )97i P".7F.6F.'iF. IF, lF,KlF.B3r ACC.

liR

y-j .

- i Bu i Id Lng

SIFmat Ministry of Ooii^t. 1975

!<F, FuuLii'j © CL-0. 5ai » 2 ,-15m,-ltlm ACC.

4 'J Apai tment

Bui Idins

SRC

IIF

pier (1 : 26.11

)

Ministry of Const., Terhriology

rc-iiter for National L.->tid Development, Kajima Torp.

1971

111 ,lF,C.L-nm (pier GL-1 3TM, -26m

)

GL-2m,-13m © CL-2m,-5.5m,-9n,-13m,-30m,-100, ACC.

D i s .

11-56
Table 1 (B) Earthquake Observation Project in Type B.

Orgnni zat Ion Opening

Date

Obj ec tive Quantity

1 K'lah I ro . Hokkaido Hokkaido Univ. CL, CL-20m ACC.

2 !!,,.n^m:^t^ucho. Tr,k>'n j ima Corp. 1966 r.L , GL~?i*-n . -UOm ACC.

Tocniikomal , Hokkaido Hokkaido Univ. 1966 CL, GL-DOm. -50m ACC.

U Shinjiiku, Tokyo Takenaka Co. 1967 GL, CL-1.3ni. -13ni. -27ni, -61. 6m ACC

5 Sapporo , Hokka Ido Hokkaido Univ. 1967 CL. GL-60m ACC.

6 Saljo, Ehinio Shikoku Electric Power Co, 1968 CL , Gl^lOm , -20m ACC.

Yokohama , Kanagawu Takenaka Co

.

1969 GL-kai, -I8m, -31m

8 O-itHciachi , Tokyo Vaseda Untv

,

1969 GL, GL-26b>. -66ni acc.

9 Abeno , Ohsaka Hukui Univ. 1970 GL, GL-lOm, -30m acc.

10 Kiyoae , Tokyo Obbayashi -g^ujni Ltd

.

1970 CL . GL- 10m. -20ra, -iBOro A C C

D 1 s .

1] Ki o i chf> , Tokyo Taisel Corp. 1970 CL-2m, -33m> -83m ACC.

D , s .

12 Kobe, Hyoffo Kikken Sekkel Ltd. 1970 CL-2.2in, -3'*. 5m. -83.9m ACC.

13 Nishiuwa. Enime Power Co. 1970 EL+lOm, -lOni, -30m ACCShinjuku, Tokyo Kajlma Corp, 1970 GL, GL-2i*.3m ACC.

15 Vatsu, Chiba Ml tiistry of Const

.

1970 GL, CL-70m. -IJOrr) ACC.

16 K^iwasaici , Kanagawa Hinlatry of Const. 1970 GL, GL-27ra. -67m, -127tn ACC.

17 KaTinonzakl , Kauagawa Ministry of Const. 1971 GL, GL-80m, -120m ACC.

16 Ynkohama , Kanagawa Pujita Corp. 1971 GL-1.5m, -13.2m, -60,2m ACC.

1

' at^y ..lid , Chlb^ 1 h'_' N.-» t vo:-.» I !!-• rch Ccii t e : for DisaaLnr Prfrvcn tion bclence 4 Technology Agencyl'*7i GI. \ e 1 .

H.« 1 iiy..ni;icho , Ch i^- j ih- N..tti>n.-»1 tt—*-.^Tch C-^T.t'T lor

Technolojfv A{^<'n^y

\' e 1 .

21 Chishlnacbo, Ohsafca Fukui Univ. 1971 GL, CL-20m, -33TM, -60m \CC.

Mori noinlya , Ohsaka Pukul Univ. 1972 GL, CL-.?2m, -60a ACC.

23 Mliiato, Tokyo Ohbay«9ht-5>.itiil Ltd

.

1972 GL, GL-lflm, -26.5m ACC.

Yokohama , Kanafcawa Kajima Corp. 1972 GL, CL-Simi ACC.

11-57
7^ Ul. ,r,L - 1 0 . 5m. - 3 1 . 5m Objective

Quantity

26 Kawdg^ich 1 , Salt ama Ministry of Const. 1972 CL-7ra . - 1 i»ni . - 2 1 m . - 30m , 3in , -60ni ACC,

27 Kitasuna, Tokyo Ministry of Const. 1972 GL.GL-Sm, -20m, -300, -l*8m, -60ni ACC.

28 Wakayama Toanenryo Industrial Co., 1972 GL 3, CL-31m, -5701 ACC.

V e 1 .

29 Cbiyoda, Tokyo Shlmizu Const

.

1972 GL-lm, -25n>. -50n ACC.

30 Totnlyainacho , Chiba TbB Nat lona 1 Research Center for Disas t er Preven t Ion Sc lence A

Technology Agency

1972 GL-50m V e 1 ,

31 Tomlyamacho , Cb Iba The National Research Center for Disaster Prevention Science A

Technology Agency

1972 Ve 1 .

32 Ajoa^asaki , Hyogo Pukul Univ., Ohaaka Inst . of Technology

1973 CL, GL-15m, -300, -60m A 0

33 Toyosu, Tokyo Shlmizu Const

.

1973 GL-2ai, -18m, -38m, -17ni ACC.

Kouto, Tokyo Ministry of Transport 1973 GL, GL-20m. -50m, -90m ACC.

35 FXmabashi, Chiba Ministry of Transport 1973 GL, GL-20m, -^Om, -90id ACC.

3^ 1973 GL , GL-20ni , - 50m , -90m ACC.

37 Taisho , Ohoaica Minis try of Transport 1973 CL, GL— 20m, — 50m, —90m A ^ 0.

38 Urayasu, Chiba Nuto Institute of Structural

Mechanics

1973 GL, GL-lOm, -kl.hm ACC.

D j s .

39 Naraahlno, Chiba Takenaka Co. 1973 CL-2 . 5" ACC.

^0 Iwatsuki , Saitajna The National Research Center for Diaaster Prevention Science A Tech jIo logy Agency

1976 GL-3510m

CL-80in ACC.

Ve 1 .

Ul Ohsaka , Ohaaka Pukui Univ. Ohsaka Xnst . of Technology

197^* GL, GL-60m ACC.

U2 Hachlohjl, Tokyo Vaseda Unlv

.

197'* GL, GL-l^in, -25TM ACC.

Yokoheuna , Kanagawa Ministry of Const. 197^* GL-5m. -15m. -38m. -l60m ACC.

Tokyo Internat lonal Airport , Tokyo Ministry of Transport 197^* GL 6, -50TM, -67m ACC.

f*^ Hachinohe , Aomorl Hachinohe City, Nippon Steel Corp.

1975 GL, GL-2m ACC.

U6 Kawasaicl , Kanagawa The Nat lona 1 Research Center for

r>ls."»ster Pr**vont ion sc lence A

Tecri.ioloKT A.-rency

1975 CL-600m Ve 1 .

U7 PuW-ui. Pukul rukul Univ. 1976 CL, CL-lBm, -eOm, -175ffl (Rock) ACC.

.\ 1 uw.^yuma , f'likui Kukui Univ. 1976 CL ACC.

11-58
Number of Seismomenters Percentage

\ lo)

< 2 23

3 - A A5

5-6 20

7-8 " 7

^ 9 5

Table 2 Statistics of Number of Seismometers

Depth

(m)

Percentage

C/o)

< 20 57

20 - AO 17

AO - 60 1

1

60- 80 3

80 -100 A

100 -200 2

^200 1 Outcrop

Rock 5

Table 3 Statistics of Largest Depth of Seismometers

1

Objective Quantity Percentage

Acceleration

(cm/see

)

87

Velocity

(cm/sec

)

8

Displacement

(cm) 5

Table 4 Statistics of Objective Quantity for Measurement 11-59
DEPTH 39^

ETER

^100

sasMc

S€TER

300

EXCAVATION LOAMJ _

!<W«\NTO LCm_

L0A>'1 , SAND C3AND

RNE SANU

r/v ^00- MtTtR

- bOO-

- 600

- 700-

:i]5M0

';>EOLpOiCAL ^e^fV-y4gE

10 X) 50

RNE SAND - CLAY

FINE SAND

GRAVEL

_SILI.

FINE SAND

FINE oANU, CLAY

CLAY , 'jANO

SILT

0

10

20

¥

H

u3 20 acD«oeoo80olooo ^ aDO«OaD080DioOO

i8

DENSITY

lOi

20

40

SH-VELDCITY

-.8

< CM

p-vELxmr

Fig. 1 Soil Profile and Underground Structure

h 1962-1963 - ADDITION

196^ - 1965

1968 r

ADDITION

1970

f "'1^1 1

'

I -

1

112.000. - ^ ^ eoooa _^ 20.Qon

^

STAGE 1(1963- ) STAGE n(1965- ) ST>^ B (1968- ) —-STAGE W(1970- }

GL

SXXi ,

ABCDE

-j m

iJK

J(GU

R4^

TEST

BUILDING

<MO

00 a>

T-23)

E

(-42)

rl (560 GLl

1(530-5)

G, (560-23)

30.000-

H

(560-41)

30.000-

(-82)

Fig. 2 Plan and Construction Stage

Fig. 3 Location of Seismometers

11-60
G»oloacBl l>f«y N Mlur Mxlly of P Mi* VIMoaty o( S «<o»» PoMor s ruio Melon (frWt (Ivn/wcl (km,i.<l 0-, p_|0> 1 iL_ffl_4!L yi 2 ?

I I 02 (U. Ot. Ot - -

HO,

J

I ^20

vet

i Lfe

Fig. 4 Example of Acceleration Records Fig. 5 Soil Profile and Underground Structure

PLAN

35011003001100550

400C ELEV. 55C 290C .5»

200.75Q5O 50.750200 XJOOXXW iOOO XXXJtOOO 8000

15m I

Fig. 6 Plan and Elevation Fig. 7 Location of Seismometers 11-61
(GAL) ZOOr

100

iOO

200

-y^|^ *^^<^.< V

SFB -2

(13) Ta

4**

(6) -05II5

(9) -05«30

1

10,0 tsjd iseci

•

,

, -""^

(5) -15V15

(2) -ilY15

ijoi -o;yx

(16) FSZ

,6) -yszis

Fig. 8 Example of Acceleration Records

(m^ Soil 0 N 50 0 VS C^/S) 500 GL

30

40

I

I

i

•mended 5

[13 CLAY d LOAM E3^JJ^\

EH] SILT CD SAND

Fig. 9 Soil Profile and Underground Structure

Fig. 10 Plan and Section Fig. 1 1 Location of Seismometers

NBI lONHL PROJECT EQ

Fig. 12 Example of Acceleration Records 11-62
Fig. 13 Soil Profile and S-Wave Travel Curve

L-P-G--TANK (ICHIHfRAl LOC • i (1974. 5-91 EW

L.P.G--TANK (ICHIH*RA) LOC • i (l374.5-aj n5

L-P-G- - l\ANK, (ir.HIHARA) LOC - 3 (1974.5-91 EW

Ldc. 5

I

20.4 m

• Tuss Type /Iccelerometer (T= 0.2sec J X Displocement-meter (T= 5 sec )

Ljx.3 — -6m

Loc.4

Loc. 2 •-I7'.4m

L-P-G--TANK (ILHIHACJA) LOL • 3 (19C4.5-9) N5

L'P-G--TANt\ UCHJHA9A1 LOC-4 (1974.5-91 EW

-33mLoc. I ,.434r 80m H

Fig. 14 Location of Seismometers

L.P.G--TANK (ICHIHA9A) LDC-4 (1974.5-91 n5

L.P-G--TAN\ (ICHIHARA) LOC-5 (l974-5-9i Ew

L,.l^-r,.-TANK. IIrMIH»RAI LOr-'^ (1974-5-9' N5

Fig, 15 Example of Acceleration Records

11-63
DEPTM GEOtOGICAL N -VALk^

(w) I I x tg K

w2Ca

ec

8C

9C XX

110-

12GWC

15Cn>clalm>d Soil SQn<^ Sill CklyOT Silt

Silly CWy

Sill SUty Fln» Sand

Muddy Sond

Sllty Fln« Send Sandy Silt Sond

Sllty Rn» Sand

Sand

Sllty Sand

Silly ftn» Sena

Fig. 16 Soil Profile and

Location of Seismometers

GAL Om NS .A

(^AL OmEW

L'

- 27m UO

- 67mUD

-127m UD

Fig. 17 Example of Acceleration Records

11-64
SINGLE-STORY RESIDENTIAL MASONRY CONSTRUCTION

IN UNIFORM BUILDING CODE SEISMIC ZONE 2

G. Robert Fuller

U.S. Department of Housing and Urban Development

Washington, D.C.

ABSTRACT : : ,

The U.S. Department of Housing and Urban Development's (HUD) Minimum Property Standards

(MPS) require that all masonry construction in Seismic Zone 2 have partial reinforcement in

accordance with the Uniform Building Code (UBC). Since there is a lack of behavioral data on

the resistance of single-story masonry houses to earthquake forces, the housing industry has

objected to the added cost of providing this reinforcement. HUD therefore contracted with

the University of California at Berkeley to determine the behavior of masonry structures

under seismic loads, by testing full scale specimens on the shaking table. This paper is a

report of the related results of material, timber roof connections and shaking table tests.

HUD also engaged the Applied Technology Council (ATC) of Palo Alto to develop design and construction criteria for adoption in the HUD-MPS. Upon completion of that contract in early

1980, a separate report will be presented to the UJNR Panel.

KEYWORDS: Minimum property standards; partially reinforced masonry; residential; roof

diaphragms; seismic resistance; single-story; shaking table tests.

III-l
INTRODUCTION - .

"Partial reinforcement" is required for all residential masonry construction in Seismic

Zone 2, if compliance with HUD's Minimum Property Standards (MPS) [1]* is to be maintained.

The Uniform Building Code (UBC) [2] is cited in the MPS as being the acceptable criteria.

There is, however, a lack of data on the resistance of single-story masonry houses, subjected

to earthquakes. Justification of partial reinforcement was therefore questioned by many in

the housing design and construction industry, particularly in the western states of Nevada,

Arizona, Utah, and Colorado, where masonry is the predominate form of construction.

Consequently, HUD contracted with the Earthquake Engineering Research Center (EERC) of

the University of California at Berkeley to test the seismic behavior of full-scale, singlestory residential masonry buildings on the shaking table. This paper therefore presents the

findings from these tests, plus cyclic tests conducted on timber roof to masonry wall

connect ions

.

As previously stated, HUD's MPS requires compliance with UBC; masonry construction in

Seismic Zone 2 must be partially reinforced. This MPS requirement applies to all housing

produced under HUD programs or covered by HUD mortgage insurance.

The Minimum Property Standards are considered as minimum requirements for the design and

construction of adequate housing. Where local codes or standards permit lower requirements

than the MPS, the latter shall apply. Local Acceptable Standards (LAS) to modify the MPS,

however, may be issued by HUD Field Offices, with concurrence by HUD Headquarters.

In August 1974, the HUD office in Phoenix, Arizona issued an LAS to cover the design and

construction of masonry buildings. Specifically, unreinforced masonry was not permitted, and

masonry walls of single-story residences located in Seismic Zone 2 would require at least the

following amounts of vertical reinforcement: [3]

1) No. 4 bars at all building corners, at wall ends, at all door and window openings,

and in wall sections 2'-0" (0.6 m) wide or less. Openings larger than 12*0" (3.7 m) in width

require special analysis.

2) No. 4 bars spaced at not over 12'0" (3.7 m) o.c.

3) All reinforcement to be matched with dowels embedded in foundation walls or slabs.

* Numbers in brackets indicate references.

III-2
The promulgation of this standard was questioned by representatives of local industry,

on the grounds that compliance with the requirements would lead to increased costs and

unnecessarily high factors of safety. To address this problem, the research program was

organized to yield the following information: [3]

1) Dynamic behavior of simple masonry structures subjected to simulated earthquake

motions; ., ,. , .

2) Reinforcement requirements, if any, for adequate transverse and in-plane resistance

of typical masonry house construction for level of seismic activity associated with Seismic

Zone 2, ,. , , ,, , ^ ... ,

3) Adequacy of typical roof-masonry wall connection details to withstand earthquake

forces . ; , . ., ; : ...

OBJECTIVE

The overall objective of this investigation was to subject typical masonry houses to

carefully controlled simulated earthquakes and to observe their structural behavior. Partially reinforced and unreinforced wall panels, fabricated of hollow-core concrete masonry

units and hollow clay bricks, were tested. [3] ..

:

Four house specimens, similar to that shown in Figure No. 1, were tested, each was

16' X 16' (4.9 m x 4.9 m) in plan and had 8' - 8" (2.6 m) high walls. Walls were connected

at the top by a timber truss roof diaphragm with trusses arranged parallel or perpendicular

to the horizontal motion of the shaking table. To account for the reduction of mass resulting from the scaled-down plan dimensions, concrete slabs were bolted to the roof. [4] ,

Each structure was tested with a series of horizontal base motions of increasing

intensity. The three simulated earthquakes utilized in the program were the 1952 Taft,

1949 El Centro, and 1971 Pacoima Dam accelerograms. All houses were designed so that simultaneous transverse and in-plane response of both reinforced and unreinforced panels could be

measured. [5]

Another objective of the research program was to analyze the behavior of typical timber

roof connections used in single-story masonry residential construction when subjected to

cyclic loads. To assess the adequacy of these connections in transferring roof inertia

loads, five basic types of roof-to-wall connections (nineteen models) were subjected to

displacement controlled load tests, both in-plane (along the wall) and out-of-plane . The

III-3
five types of connections contained loadbearing and nonloadbeari ng wall sections and either

gabled truss or flat roof construction. By examining the mode of failure and code allowable

loads on bolted and nailed connections, the margin of safety inherent in current code

requirements could be determined. [6]

Although inadequate seismic resistance of poorly designed or constructed masonry

buildings has been given heavy emphasis in post-earthquake survey reports, no experimental

studies had been reported. The overall purpose of the study was to evaluate the earthquake

resistance of typical masonry dwelling construction in less seismically active regions of

the U.S. Once this was accomplished, reinforcement could be determined. Therefore, during

the project's planning phase, an important aspect was to model structures which would be

simple in concept and yet contain the most significant components, such as wall panels with

and without openings, and with wall-to-footing and roof-to-wall connections.

MATERIAL PROPERTIES

Since the final objective of the research was to arrive at reinforcement requirements,

the number of parameters had to be limited. Therefore, the types of masonry units, mortar,

grout and reinforcement had to be kept to a minimum. In addition, all materials had to be

commercially available and typical of those commonly used in Arizona. [3]

Concrete Block Units

House specimens No. 1, 2 & 4 were constructed of standard two-core hollow block with

nominal dimensions of 6" x 4" x 16" (15 cm x 10 cm x 41 cm). Two different shapes of block

were used; corner return and standard blocks in House No. 1, and only standard units in

House No. 2 and 4. Dimensions and properties of typical concrete masonry units are shown

in Table No. 1 (House No. 4) [7] and masonry strengths are shown in Table No. 2. [5] Typica

units are shown in Figure 2. [3]

Hollow Clay Brick Units

House No. 3 was constructed of 4" x 6" x 12" (10 cm x 15 cm x 30 cm) standard hollow ''

clay bricks. Two shipments of brick were received at different times; Shipment No. 2 had ''^

III-4
38% voids (Figure 3). Five bricks were selected at random and were tested for strength.

The average strength of bricks in Shipment No. 1 was 4970 psi (34268 kN/m ) and in Shipment

No. 2, 4060 (27994 kN/m^). (Table No. 2).

Compression tests were conducted on two- and ten-unit high prisms and diagonal compression tests were conducted on 36" (91.4 cm) square panels. The average strength of two-unit

prisms was 1220 psi (8412 kN/m^) for Lot 1 and 2280 psi (15721 kN/m^) for Lot 2. The corresponding values for ten-unit prisms were 2120 psi (14617 kN/m ) and 1710 psi (11790 kN/m )

,

o respectively. Critical tensile strength was determined to be 47.2 psi (325 kN/m ) and 81.5

psi (562 kN/m ) for the two shipments. [8]

:

' Mortar

Type S mortar, as specified by UBC , was used in all houses. Volume proportions were

1: 1/2: 4 1/2 for cement: lime: sand. For each batch, two mortar samples in 2" x 4"

(5.1 cm X 10.2 cm) cylinders were taken and air cured in the laboratory under the same

conditions as masonry specimens. The resultant, average, 29-day mortar strengths are shown

in Table No. 2. [5]

^

^ . - : Grout

A fine sand and cement grout was used in all partially reinforced wall panels. One part

of Type I cement and three parts of top sand were mixed with water to achieve the desired

consistency. Control specimens were cast in typical concrete block cells and stored in the

laboratory under the same conditions as wall panels. Compressive strengths were determined

by cutting the filled cells into 2" (5.1 cm) cubes and testing on a standard testing machine.

A minimum of 2,000 psi (13,790 kN/ra ) strength was specified and exceeded by all samples as

shown in Table No. 2. [5]

, .1 * Reinforcement

Partial reinforcement was provided in Houses No. 1, 2 and 3 by medium grade steel #4

bars. House No. 4 was tested twice with no reinforcement; then 2 - #3 bars were placed in

each wall. In House No. 1, average yield (fy) and ultimate strengths were 54,000 psi

(372,330 kN/m^) and 80,000 psi (551,600 kN/m^), respectively. The corresponding values for

III-5
House No. 2 were 59,250 psi (408,529 kN/m'^) and 86,000 psi (592,970 kN/m^) , respectively.

No attempt was made to obtain additional properties of steel, since stress levels were not

proved significant in determining structural response. [3]

Joint reinforcement of 9 gage, truss type steel wire was provided in the top two bed

joints of each wall and corner unit of Houses No. 1 and 4. No joint reinforcement was

provided in Houses No. 2 and 3.

SHAKING TABLE TESTS

Test Specimens

The first test structure was considered exploratory to observe behavior of masonry

buildings under realistic earthquake loads. Limitations of types of masonry components and

size of the shaking table resulted in design and fabrication of the structure shown in Figure

No. 4. Four 8' (2.4 m) long wall panels and four corner L-shaped units were used. Two wall

panels and two corner units were reinforced with #4 bars and all reinforcement was lapped 20"

(51 cm) with dowels embedded in the footings. The footings were then attached to the shaking

table to prevent horizontal translation and rotation.

At the beginning of the testing program, only translation of the walls was prevented.

During later stages, however, in-plane walls were bolted down to prevent observed uplift.

Footings under all four corners were also prestressed against the shaking table.

In House No. 1 the roof was anchored by three symmetrically placed bolts to the

masonry wall panels and corner units. Overall structural integrity was provided by the roof

diaphragm, which consisted of trusses and plywood sheathing.

To account for scaling down of the overall building plan and reduction in weight, six

concrete slabs were bolted to the plywood sheathing. A prototype total roof load of 20 psf

(964 N/m ) was assumed, resulting in a total roof weight of 15,000 lbs (6,804 kg). [3]

Compared to current construction practice, the geometry of the first test specimen was

over simplified. It was therefore decided, after consultation with professional engineers

and home developers in Arizona, to revise the second test house configuration.

House No. 2 consisted of four independent wall panels with either door or window

openings. In-plane walls were 16' (4.9 m) in length and transverse walls were 14' (4.3 m)

long. Two walls were unreinforced ; the other two walls had a #4 bar in each end cell.

III-6
Double steel angles were used for lintels above doors and windows, and large openings had

steel beams with the top flange trimmed and a steel plate welded to the bottom flange.

Corner units were omitted to prevent interaction between in-plane and transverse walls,

and wall elements were designed to represent center parts of longer wall panels in a typical

dwelling. [3] r

-

•'

. •

'

In order to compare clay brick to concrete block wall construction, House No. 3 was

constructed almost identically in House No. 2. Some minor adjustments were made to the

dimensions to account for the difference in length of the two material units. [7]

Another modification was to introduce continuous bracing into the roof system to

minimize the effect of flexible roof trusses when transverse to the direction of motion.

This reduced, but did not eliminate, rocking of the roof trusses which apparently absorbed

part of the energy from the concrete slabs. ., .

House No. 4 was different from all previous test specimens. A rectangular plan was

used and separate 4'-0" (1.2 m) wide wall panels were constructed. Initially no reinforcement was provided. Then, after two tests were conducted, two #3 bars were placed and grouted

in each of the four wall panels. There were two objectives considered: [7]

1) To establish a lower bound of critical base acceleration for a completely reinforced

house . >: • . .

2) To determine the influence of large out-of-plane displacements on relatively slender

piers having different reinforcement details. , , ,. .

Shaking Table Test Procedures

Each of the first two houses was subjected to a series of single horizontal component

base motions. These motions varied in intensity from just barely perceptible (peak base

acceleration less than 0.05 g) to extremely severe (base acceleration over 0.5 g). Three

different types of earthquakes were simulated within this range of intensities:

1) The N-S component of the 1940 El Centro,

2) the S69E component of the 1952 Kern County (Taft), and

3) the S74W component of the 1971 San Fernando earthquake recorded at Pacoima Dam. [3]

Response of the structures was recorded by an array of displacement and acceleration

measuring devices. Displacements were measured by direct current transducers or potentiometers and accelerations by strain gage or force-balance acce lerome ters . Displacements of

III-7
the structure relative to the shaking table were measured from a rigid frame inside the

structure. This rigid frame was well braced and extremely stiff, with natural frequencies

considerably higher than significant response frequencies of the test structure.

Each house was tested under a wide spectrum of conditions. In addition to the variations of applied base motions, roof truss orientations as well as base fixity of in-plane

wall footings were changed. [3] After initial testing, cracked masonry elements were also

repaired with a fiberglass based surface bonding mortar, and then retested.

The basic philosophy was to subject the structures to a wide range of earthquake

motions. By measuring and visually observing the structure's behavior, it was possible to

identify initiation of significant structural distress and to compare simulated seismic

effects with what could be expected in a Zone 2 area. [8]

All tests conducted on the first two houses had only a horizontal component. Then,

after initial failure and subsequent repair of the unreinforced panels, four runs on the

third house had vertical input motions.

This third structure was subjected to simulated earthquakes over a period of two weeks.

Characteristic features and intensities of applied motion were varied, and the roof structure was rotated 90° for several of the tests. After initial failure, a surface bonding

cement was used to repair the walls. The parameters of the first 17 tests were identical

to those of House No. 2, and the results were very similar. But then in Run 18 (Pacoima/

300) and Run 19 (Pacoima/400) , transverse unreinforced walls failed. These same walls had

survived similar testing in House No. 2. After these walls were repaired, vertical table

motions of 50% to 60% of horizontal intensity were introduced. In general, this caused a

10% - 15% increase in recorded amplitudes. [8]

For comparable peak accelerations, the Pacoima record seemed to be more with a particular base motion and with the roof trusses oriented either parallel or transverse to the

table motion. A total of 29 test runs were conducted on the specimen in a two-week period,

roughly an equal number for each roof orientation. Evaluation of test results indicated the

following: [3]

1) Response of test specimens to base motions was complex, and affected by roof orientation, in-plane wall base fixity, and cracking of unreinforced walls.

III-8
2) Overall response limits were dependent upon roof orientation and vibratory characteristics. In-plane response was governed by inertial effects transmitted from the roof.

Out-of-plane response depended upon constraints and vertical load provided by the roof.

3) Both out-of-plane walls were capable of significant transverse displacement in the

"cantilever" mode without structural distress. Inasmuch as transverse walls, because of

their much smaller rigidity, resisted little of the roof inertial load, their performance

was governed by the amount of transverse moment capacity.

4) Nominal amounts of vertical reinforcement prevented signficant damage.

5) Surface bonding material used to repair cracked unreinforced walls appeared to be

effective. Following repair, test specimens had to be subjected to increased base motions

before cracking reoccurred. [3]

House No. 2

This test specimen was also tested over a two-week period. The roof was first oriented

with the trusses in line with the table motion; about midway in the test sequence the roof

system was rotated 90°. In all, 32 tests were run.

Because of different geometry of in-plane walls. House No. 2 was stiffer than House

No. 1. Therefore, smaller displacements were recorded for similar base motions. Other

signficiant observations were: [3]

1) Transverse walls displaced consistently greater than in-plane walls. Relative

displacement of the roof was greater than out-of-plane wall displacement. The roof diaphragm

was very flexible when trusses were perpendicular to the table motion.

2) Because of torsional aspects, inherent in the design response of the structure,

simultaneous twisting and bending were observed in the two panels perpendicular to table

motion. However, during both roof orientations, transverse displacements did not cause

substantial cracking.

3) The unreinforced in-plane panel failed during Test 32 (Pacoima motion with 0.519 g

peak acceleration), when it was not in a loadbearing configuration. This indicates that the

roof orientation had a large influence on the measured response of the walls.

4) Horizontal cracking at the midheight of the unreinforced transverse wall occurred;

caused by lateral inertial forces. [3]

III-9
House No. 3 ' i ',' J'^"

The following observations were noted during the testing of the subject specimen: [8]

1) The Pacoima record appeared to be the most critical for the same peak acceleration.

Failure occurred in the out-of-plane , unreinforced wall at 0.386 g. Both unreinforced panels

underwent significant structural damage at 0.480 g, but the house did not collapse.

2) In addition to sliding relative to shear walls, the roof deformed unevenly and

overall torsional motion took place. The roof structure did not appear to be very rigid.

3) Structural response appeared mostly at frequencies of between 2 Hz and 7 Hz. For

the rigid structure, the table motion contributed greatly. Acceleration peaks also shaped

the overall response.

4) Based on the results of these tests, 0.35 g peak base acceleration was approximately

where strong, nonlinear behavior and structural distress began. There seemed to be no fundamental difference between the seismic response of concrete block and clay brick.

5) Introduction of vertical motion increased displacements by 10% to 15% for similar

accelerations.

6) Reinforcement consistently made a difference between structural integrity and

failure. However, the reinforcement was stressed minimally, while the grouted cavities withstood the forces.

7) Repairing cracked masonry elements with surface bonding restored some of the

flexural strength of panels. However, surface bonding was only partially successful where

panels were subjected to significant in-plane shear or shear combined with flexure. [8]

I.

House No. 4

- <(

The initial tests require careful examination, since these were conducted on totally

unreinforced structures. The structure was subjected to a base motion similar to the El

Centro accelerogram, with both horizontal and vertical peak accelerations.

Some cracking occurred in all the panels, but the structure did not approach failure.

Transverse panels and piers fared better than in-plane walls. These transverse panels,

however, did eventually fail as a consequence of shear, since they couldn't accommodate large

diaphragm displacements. [7]

III-IO
The Taft signal with horizontal and vertical peaks of 0.287 g and 0.219 g was the second

strong base motion used. Crack patterns indicated that both in-plane walls developed lateral

load resistance largely from friction. Transverse walls were able to withstand significant

displacements and unreinforced piers deflected or arched at about midheight. Rigid attachment of the roof structure to these elements apparently improved their behavior.

Final tests were conducted after wall panels had cracked and after reinforced cells

were grouted. Reduction of displacement amplitudes indicated an overall strengthening and

stiffening of the structure.

Other important observations from the tests are as follows: . .,

.

1) For in-plane walls, a horizontal acceleration of below 1/3 g did not cause failure.

Signs of failure were observed, however, at this value or at a slightly higher intensity.

The critical level of excitation appeared to be slightly higher for transverse walls which

were unreinforced.

2) Transverse walls accommodated large displacements because of transverse flexibility

of plywood roof diaphragm. In-plane walls, however, tended to develop uplift modes in

responses to overturning moments.

3) Doweling of vertical reinforcement to foundations did not appear to be required for

transverse panel strength. Dowels, however, had a signficant effect on in-plane walls when

accelerations exceeded 0.45 g. [7] . . • • ;

•

- Shaking Table Tests - Conclusions

A large number of simulated earthquake tests were conducted on four masonry houses

containing both unreinforced and reinforced components. Although the test structures had a

similar configuration, roof truss orientations and base motions at various intensities were

different. An overall evaluation of the shaking table tests yields the following: [5]

1) Strong base motions of peak accelerations greater than 0.3 g caused significant

cracking in unreinforced elements. Partial vertical reinforcement provided a definite beneficial effect, for base motions up to 0.6 g, by inhibiting cracking and by limiting permanent

displacements. No structural failure was recorded for partially reinforced walls.

2) Unreinforced wall transverse strength was closely related to bond strength between

mortar and masonry units. Cracked transverse walls experienced large out-of-plane

TTI-11
displacements under base motions on the order of 0.3 g. Because of arching and support

restraint, however, walls did not become unstable. This stability of unreinforced masonry

walls needs further study.

3) Two #4 bars at 14'-0" (4.3 m) spacing in transverse walls were adequate to resist

all force levels up to 0.6 g peak acceleration.

4) The strength of in-plane unreinforced walls was governed by overturning; but nominal

shear stresses only reached a peak of about 30 psi (207 kN/m ). Therefore, the dowelled

vertical reinforcement at each of the 8'-9" (2.4 m) wind piers in House No. 1 and 16'-0"

(4.9 m) long walls in Houses No. 2 and 3, was sufficient to resist overturning and prevent

failure

.

5) Detailed evaluation of test data is still underway, and specific reinforcement

requirements for masonry houses in Seismic Zone 2 have not yet been determined. The amount

of reinforcement is dependent upon the safety factor considered appropriate for residential

construction. Low safety factors would not require reinforcement, whereas safety factors

over 3 may require at least partial reinforcement. A more detailed analysis of the test

results, including the safety factor effects, will be required. [5]

CYCLIC LOAD TESTS OF TIMBER ROOF CONNECTIONS

Timber Roof Connection Specimens

This investigation was conducted to determine strength and cyclic behavior of typical

roof connections used for masonry dwellings constructed in less seismically active areas of

the U.S. Five types of roof connections were cyclically loaded by dynamic actuators. Specimens were subjected to either in-plane (parallel to wall) or out-of-plane (transverse to

wall) horizontal loads. Both loadbearing and non-loadbearing models of gabled truss and flat

roof construction were tested. [6]

The study clearly distinguished between in-plane behavior of connections with inertial

effects transmitted from the roof diaphragm oriented parallel to the wall, and transverse

behavior when the force was applied normal to the wall. In addition, bolt size was considered a variable. Each model had an 8'-0" (2.4 m) long masonry wall panel with bolts,

III-12
4'-0" (1.2 m) apart, embedded in the wall. Wall heights were varied and both hollow concrete

block and hollow clay brick, were used. Wall panels were reinforced with 2 - 4# bars at each

end

.

The first two specimens represented a portion of bearing wall with four truss joists,

at 2'-0" (61 cm) on center, nailed to a single 2" x 6" (5.1 cm x 15.2 cm) top plate. Two

truss joists were also secured to the top plate by metal framing anchors (see Figure No. 5).

The third connection specimen represented a gable end of a roof system with trusses

parallel to the wall. "Truss-ties", at 6'-0" (1.8 m) on center, were used to connect the

bottom chord to the wall segment. In addition, the bottom chord was attached to the top

plate through solid wood blocking at 4'-0" (1.2 m) on center (see Figure No. 6).

Specimens four and five represented flat roof construction with ledger boards bolted

to the wall face. The nominal thickness of the ledger was 3" (7.6 cm) for one test and 2"

(5.1 cm) for the other. Both groups were subjected to horizontal forces transverse to the

wall. Some tests were run with an added vertical roof load of 180 plf (2.65 kN/m) simulated

by placing weights above the ledger. [6] j -. , .

•'; • '

Timber Roof Connection Test Conclusions

On the basis of measured strength and observed behavior of tested connections, the

following was concluded:

1) Cyclic shear forces caused local deformations resulting in simultaneous slip and

pull-out of nailed timber connections. Therefore, safety factors based on code provisions

had to be substantially reduced.

2) The truss rafter to top plate and masonry wall connections tested in the first two

groups of specimens adequately resisted Seismic Zone 2 forces.

3) The size of bolts in the first three groups of tests did not affect connection

strength. Minimum 1/2" (12.7 mm) diameter bolts, with embedment lengths equal to or greater

than UBC code requirements, are recommended.

4) Further testing of the ledger connection with horizontal forces parallel to the wall

is recommended. Anchorage devices would not be effective initially. Therefore, the parallel

forces would be resisted solely by anchor bolts. -

III-13
5) Response characteristics measured in the fourth group of tests (flat roof to ledger

connections) were not significantly affected by either the use of hollow clay masonry or by

the addition of the gravity roof loads.

CONCLUSION

This research at the University of California at Berkeley, conducted under HUD contract,

was just completed in August 1979. Results of the shaking table and cyclic roof connection

tests will be published by HUD's Office of Policy Development and Research. Copies of the

final reports will be available from the National Technical Information Service (NTIS),

Springfield, VA 22161, USA.

Films were also taken of each of the shaking table tests and, after editing, should be

available for viewing at future UJNR Panel meetings. A slide presentation will also be prepared, to be used for training HUD field personnel. ^ '

Planning of the experimental research program was coordinated, under a separate HUD

contract, with the Applied Technology Council (ATC-5) Advisory Panel. ATC will also provide

design and construction recommendations for single-family masonry dwelling construction in

U.S. Seismic Zone 2 areas. The final criteria is intended to be incorporated in HUD's Minimum Property Standards (MPS) or in the MPS Manual of Acceptable Practice (MAP).

REFERENCES

.1 .. ^

'

[1] U.S. Department of Housing and Urban Development, "Minimum Property Standards, One- and

Two-Family Dwellings," HUD 4900.1, 1973 Edition, Washington, D.C.

[2] International Conference of Building Officials (ICBO), "Uniform Building Code," Vol. I, 1973 Edition, Whittier, CA.

[3] P. Gulkan, R. L. Mayes, and R. W. Clough, "Experimental Study of Reinforcements for "' Single-Story Masonry Houses - Vol. 1," (Draft) (1979).

[4] P. Gulkan, R. L. Mayes, and R. W. Clough, "Response of Single-Story Brick Masonry House

to Simulated Earthquakes," Draft for Fifth International Brick Masonry Conference,

October 1979 (Brick Institute of America, McLean, Virginia, 1979).

[5] P. Gulkan and R. L. Mayes, "Experimental Investigation of Reinforcement Requirements

for Simple Masonry Structures in Moderately Seismic Areas of U.S.," Draft for Second

U.S. National Conference on Earthquake Engineering, Stanford University, August 1979

(Earthquake Engineering Research Institute, Berkeley, CA, 1979).

[6] P. Gulkan, R. L. Mayes, and R. W. Clough, "Strength of Timber Roof Connections '

• Subjected to Cyclic Loads," Draft Report No. EERC-79 to HUD, Earthquake Engineering

Research Center, Berkeley, CA (1979).

III-14
Earthquake Engineering Research Center, "Interim Report on Summary of Experimental

Observations Concerning Behavior of Fourth Masonry House Subjected to Simulated

Earthquakes," Berkeley, CA (February 1979).

Earthquake Engineering Research Center, "Interim Report on Summary of Experimental

Observations Concerning Behavior of Third (Brick) Masonry House Subjected to Simulated

Earthquakes," Berkeley, CA (October 1978).

III-15
Figure 1. TYPICAL TEST STRUCTURE ^

III-16
-e 15 5/8" *

STANDARD UNIT

HOUSES I AND 2

CORNER RETURN UNIT

HOUSE I

Figure 2. CONCRETE MASONRY UNITS

SHIPMENT 1

i l-!/2 11-1/2

HEIGHT 3-5/8"

Figiire 3. BRICK UNiTS FOR HOUSE 3

III-17
FOOTING

1°: C2

ELEVATION

L 1/2 X 4 x5 BOLTED

TO FLAT BAR ONLY

FOOTING 2^

1

1 l/2"x 5 1/2" B

FLAT BAR ^^"^---4^

ROD

STRESS ^

W2

#4 REBARS

DIRECTION OF

MOTION

Wl

PLAN

WALL FOOTING TO

SHAKING TABLE

CONNECTION DETAIL

C3

ALL WALLS

8-8" HIGH

W4 C4

2 0 2 0 8 0 2 0 20

16 0

20

2'0"

80

16 0

2 0

20

Wl, W2 REINFORCED 2--?^4REBAR

C I, C3 REINFORCED 3-#-4RE8AR

PLAN

Figure A. HOUSE NO. 1

III-18
"1.

16" < >

1/2" PLYWOOD (EDGES W/Sd & 8"0.C.)

(INTERIOR W/6d (5) I2"0.C.) 2x6 FASCIA (2-l6dNA!LT0 JOIST)

2x4 JOIST (3-l6d NAILTO PLATE)

A35 SIMPSON STRAP (4' 0.C.W/N8 NAILS)

ANCHOR BOLTS (8"EMBEDMENT, 2"H00K)

1x4 FRIEZE BOARD (l-8d TOENAIL/EACH JOIST)

2x6 TOP PLATE (SECURE W/2 ANCHOR BOLTS)

1/2" GPDW (W/ I 3/8"NAILS 0 6"0.C. 2 STRIP (W/GdO I6"0.C. HOLLOW CONCRETE BLOCKS (6"x4"x 16") REBAR (SPLICE TO DOWEL FROM FOOTING,

BOTH ENDS OF WALL AND GROUTED)

FOOTING (BOLTED TO FLOOR)

Figure 5. TYPICAL TIMBER ROOF TRUSS CONNECTION TEST SPECIMEN

-PLYWOOD SHEATHING

OUT-OF- PLANE

DIRECTION

1

eWtrusstie-x

g>6'-0"e.c. \

»

n

M

h

li• -

TYP. TRUSS

DRYWALL NAILER TOP PLATE

-MASONRY WALL

DRYWALL CEILING

Figure 6. GABLE END ROOF CONNECTION TEST SPECIMEN

III-19
Table I. Dimensions and Properties of Concrete Masonry Units

width Height Length Minimum

Thickness

in

Gross Area

Net Area Compressive

Strength

psi

Oven

Dry Weight

Concrete Weight

in in in Face Shell Web in' Net Gross

% Area Area lb lb/ft

5.61 3.63 15.56 1.03 1.06 87.4

(1018A kN/u 53.5 2778 1477

»2) 10.84 110.5

(U.2cra) (9.2cm) (39.5cm) (2.6om)(2.7cm) (563.9cra^) (1915A kN/m^) (A. 92kg) (1770 kg/m-

(1) Values given in the table represent the average results from tests or measurements of 5 units.

Table II. Materials for the Test Structures ^

House

No. Type of Masonry

Unit

Masonry

Strength

psi

Mortar

Strength

psi

Grout f Strength ^ psi ksi

Rein forced Panels

(Fiq. 1)

1 6x4x16 in. Cone, block 1,530 2,650 6,350 54.0 Wl, W2, CI, C3

2 6x4x16 in. Cone, block

1,983 4,730 3,360 59.3 B, Bl

3 6x4x12 in. Clay brick

4,970 2,336 2,536 59.3 B, Bl

4 6x4x16 in. Cone, block

2,778 1,944 2,324 54.0 See Note (3)

J

Notes: (1) Masonry strength is based on net area of units.

(2) Type S mortar of the UBC was specified for all houses.

(3) House 4 was tested tested twice with no reinforcement in any of

the four walls before two #3 bars were grouted in each wall.

III-20
FIGURE 6 HOUSE NO.

3

III-21
FIGURE 6 HOUSE NO. 3 (CONTINUED)

III-22
THE PERFORMANCE OF LAPPED SPLICES IN REINFORCED CONCRETE

UNDER HIGH-LEVEL REPEATED LOADING

Peter Gergely

Cornell University, Ithaca, N.Y,

Fernando Fagundo

University of Puerto Rico, Mayaguez, P.R.

Richard N. Whi te

Cornell University, Ithaca, N.Y.

ABSTRACT

The results of two series of experiments on lapped splices in reinforced concrete is

reported; one on half-scale beams and the other on full-scale beams. The effects of repeated

loading and transverse reinforcement on splices in constant moment regions have been studied

to date.

KEYWORDS: Beams; bond; concrete; design; lapped splices; reinforced concrete; seismic

design; splices; testing.

III-23
INTRODUCTION

The lapped splice provisions of seismic design codes vary greatly and are generally

conservative. Lapped splicing is ordinarily not allowed in regions of a structure where

repeated yielding may occur. This restriction most likely originates from research on cyclic

loading in bond tests where serious degradation may occur, or from observations of splice

failures in seismically loaded structures. The effect of transverse steel, which can greatly

enhance splice performance, is not considered explicitly in splice design in codes for either

seismic or nonseismic areas.

Recent research [1] has resulted in expressions for the design of lapped splices in

nonseismic areas in which the effects of transverse reinforcement, cover, and bar spacing are

explicitly accounted for. The suggested provisions [1] specify a maximum amount of transverse steel (stirrups) above which additional steel is not effective. The limit is given by

where Aj^^. is the area of transverse steel crossing the expected splitting crack, s is the

transverse steel.

The research summarized here has studied the effects of high-level repeated flexural

loading (0 to Maximum to 0) and the amount and distribution of transverse steel along the

splice. Beams tested under reversed loading are not discussed in this paper. The amount of

transverse steel used in the beams are expressed in terms of the maximum amount given above.

The half-scale specimens had a span of 6 ft (1.27 m) and a cross section of 6 in. by

10 in. (152 by 254 mm), whereas ths corresponding dimensions for the full-scale tests were

21 ft (6.40 m) and 9.5 by 16 in. or 12 by 20 in. (241 by 406 or 305 by 508 mm). All beams

were loaded at the third points and the splices were in the constant moment region. The

main reinforcement consisted of two #4 (13 mm) bars in the small scale tests and two #8 or

#10 (25 or 32 mm) in the large scale tests. Both longitudinal bars were spliced in every

beam. Since the only two bars were used, they were in the corners of the stirrups.

(1)

spacing of stirrups, d^ is the diameter of the main bars, and f is the yield strength of the

TEST PROGRAM

III-24
Three small-scale beams were tested under monotonic loading and three similar beams

under repeated loading. All splices were 28 d^ long. Four beams had no transverse steel

whereas two beams had the maximum effective amount.

Twelve large-scale beams were tested under repeated loading and two under monotonically

increasing loading. All beams had web reinforcement ranging from 16% to 210% of the maximum

effective amount given by Equation 1. The distribution of stirrups along the splice was

variable; three main types of stirrup arrangements were used: (1) one near each end of the

splice and one at the center, (2) two stirrups near each splice end, and (3) closely spaced

stirrups along the splice. The splice lengths were 30 d^^ or 36 d^.

The concrete strength was nominally 4000 psi (28 MPa).

TEST RESULTS

Only selected preliminary results are discussed here; a forthcoming report will contain

a detailed presentation of the test results and a critical evaluation of their significance.

The effect of the amount and distribution of stirrups on splice behavior becomes evident

as several pairs of tests are compared. Four half-scale tests are summarized in Table 1.

The transverse steel is in terms of the maximum effective amount given in Equation 1. It is

seen that this amount was sufficient to allow the beam to reach twice the yield deflection

under monotonic loading (beam B), but it was insufficient for cyclic loading (beam D). Beams

without transverse steel are obviously unsatisfactory even for monotonic loading; beam A

failed at 90% of the expected yield load. • ': -

;

Beam Load Steel ^^"^b Failure Cycle

A M 0 28 0.90 Py 1

B M 100% 28 2 Dy I

C R 0 28 0-8 5 Py 7

D R • 100% 28 0.90 P^ 25

M = monotonic, R = repeated, P^ = yield load,

Dy = yield deflection measured at mid-span

Table I. Comparison of half -scale beam tests

III-25
Several sets of full-scale beams with #8 (25 mm) or #10 (32 mm) bars are compared. One

set of three beams had 1/3, 1, and 2 times the maximum effective amount of transverse steel.

The first of these was insufficient to go beyond the yield strength of the beam, whereas the

second design permitted inelastic deformation to about twice the yield displacement but for

only a few cycles. The beam with about twice the maximum amount of transverse steel did not

fail after many cycles at over twice the yield displacement.

The amount and distribution of stirrups has a great effect on the rate of penetration

of damage towards the center of the splice. A beam with #10 (32 mm) bars and #3 (9.5 mm)

stirrups at 3 in. (76 mm) spacing (giving 1.15 times the maximum effective amount) reached

about twice the yield displacement. The yielding of the main bar spread about 12 in. (305

mm) along the splice which was 30 d^ long. Two stirrups at each end yielded. Both the bar

strains and the stirrup strains increased continuously, especially at the peak load level,

where failure occurred after seven cycles.

The third beam had #4 (12.7 mm) stirrups at 3 in. (76 mm) which provided 2.1 times the

maximum effective amount of transverse steel. It did not fall even after many cycles at

about twice the yield deflection. The strains in the stirrups tended to equalize as cycling

continued beyond bar yield. However, the stirrup strains remained below about half the yield

level. . . ( ;.,.• 1 .1-;.

The distribution of stirrups along the splice is more important for cyclic loading than

for monotonic loading. Concentration of stirrups near the ends of the splice is beneficial

up to the yielding of the bar, but beyond that stage, damage (mainly cracking around the bar)

progresses along the splice during cycling and additional stirrups are brought into action.

Thus, it is essential to have closely spaced stirrups along the splice. The concrete cover

is much less important for repeated high-level loading than for monotonic loading below

yield.

The study of the effect of load history generally confirmed previous investigations in

bond. Repeated loading up to about 75% of yield had little effect on splice behavior. However, deflections and strains along the bar and in the stirrups increased with cycling above

about 80% of yield.

High-level load cylces cause local damage around the splice that affects subsequent lowlevel behavior. This effect can be seen in Figure 1 where bar and stirrups strains are shown

at 40% of the flexural yield load before and after cycles at P^, 0.8 P^, and P^. After a 111-26
single loading to Py the bar and stirrup strains increased substantially when reloaded to

0.4 Py (compare curves 1 and 3). However, 11 further cycles at 0.8 Py caused little additional damage (cycles 3 and 15). Another loading to Py caused comparatively more damage

because subsequent bar strains were higher as comparison of cycles 15 and 21 shows. This

figure illustrates the conclusion that a force level of greater than about 80% of yield is

detrimental to splice behavior.

The investigation has shown that the use of about twice the maximum effective amount of

stirrups specified in the suggested code revisions by ACI Committee 408 [1] with the stirrups

closely spaced along the splice, will permit repeated loading to at least twice the yield

deflection. The research is continuing and several additional variables are being studied.

Pilot tests indicate that reversed loading or the presence of substantial shear forces both

seriously impair the load-carrying capacity of splices unless a large amount of closely

space stirrups are used.

ACKNOWLEDGMENTS

This investigation has been supported by the National Science Foundation; however, any

opinions, findings, conclusions, or recommendations expressed herein are those of the authors

and do not necessarily reflect the views of the NSF

.

REFERENCES _ -

[1] ACI Committee 408: "Suggested development, splice, and standard hook provisions for

deformed bars in tension," Concrete International, Vol. 1, No. 7, July 1979.

III-27
III-28
DYNAMIC BEHAVIOR OF REINFORCED CONCRETE FRAME STRUCTURES

Keiichi Ohtani

Chikahiro Minowa

Earthquake Engineering Laboratory

National Research Center for Disaster Prevention

Science and Technology Agency

. .. , ^ , ABSTRACT

This study presents earthquake simulator test of reinforced concrete structures conducted at the Earthquake Engineering Laboratory of the National Research Center for Disaster

Prevention at the Tsukuba New Science City.

Two types of one-story, single-bay reinforced concrete frames in actual size were built

on the shaking table. The two reinforced concrete frames had different slab weights; about

60 tons for the first tested frame and about 130 tons for the second one. Using this largescale shaking table, both reinforced concrete frames were subjected to simulated and modified

earthquake ground motions with intensities large enough to cause inelastic behavior and

dynamic property changes to the frames.

Test results of both frames are described, as is the yielding of reinforcing steel bars,

the varying properties of the overall frame responses, natural frequencies, and damping

ratios

.

Finally, computer simulations to evaluate inelastic responses and the correlation with

measured performance are described using bi-linear or tri-linear analytical models.

KEYWORDS: Earthquake acceleration-displacement analysis; reinforced concrete structures;

shaking table simulation; structural testing.

III-29
1. INTRODUCTION

The nonlinear responses of reinforced concrete building structures have been studied

theoretically and experimentally. It may be necessary to verify the analytical correlation

with the earthquake simulator test, using the computer simulation methods. However, concerning experimental studies, the test specimens were the element tests or the reduced structure

models. The large-scale shaking table of the National Research Center for Disaster Prevention has a capacity large enough to excite the actual size reinforced concrete structures.

Therefore, the earthquake simulator test of actual-size reinforced concrete structure was

planned. One-story, single-bay reinforced concrete frames were adopted for the reasons of

economy and safety. . .

Two frames were tested. The test of the first frame was executed in 1977 and the test

of the second was in 1978. Both test frames had no walls, so these structures exhibited the

moment yield patterns. The column section and height of both test frames were identical.

In order to obtain different natural frequency, the second frame was given a slab weight

about twice as much as the first frame.

In another way, it is a primary interest to verify that our shaking table would cause

damage to actual-size reinforced concrete structures with the simulated earthquake motions.

2. TESTING FACILITIES AND INSTRUMENTATION

The shiking table as used for testing can move in one horizontal direction with electrohydraulic driving system. This table is 15 m x 15 m in size, 160 tons in weight. The maximum displacement and velocity that can be achieved are ± 3 cm and 37 cm/sec. This shaking

table may be used to subject a specimen weighing up to 500 tons. The natural frequency of

the table weight and the oil spring would be considered as 20 Hz approximately. The operational frequency range would be 0 - 50 Hz

.

A total of about 45 transducers were employed to measure the structure responses in the

tests of the two frames, respectively. The table and the roof acceleration were measured by

the accelerometers of strain gage type in three directions. The table and the roof displacements were measured by IDS type position meters in excitational direction. In order to fix

the position meters, the steel frame was constructed on the shaking table foundation, as

shown in Figure 2 and 9. The story displacement is obtained by subtracting the table

III-30
displacement from the roof in the use of the electric amplifier. The wire strain gages

were arranged on reinforcing steel bars (D19) of each columns, as shown in Figures 4 and 10.

In addition to these measurements, the behaviors of the steel frame were observed by accelerometer in order to verify the accuracy of displacement measurements. The test data were

recorded with 14 channel data recorders. With 16 mm movie cameras and video systems, the per'

forraances of cracks grown in columns were watched and recorded for the tests of both frames.

When the earthquake ground motion is simulated, by the reason of the limitations of the

shaking table stroke of 3 cm, the low frequency components of the original earthquake motions

may be cut off in order to produce high intensity shakes. Therefore, the table inputs for

the earthquake motion tests are obtained as the outputs from the one-degree of freedom

systems 's responses to the original earthquake motions. The damping constant of this system

should be settled as 0.707. The natural frequency of this system would be selected in accordance with the dynamic properties of the test frame. These natural frequencies were decided

as 3 Hz for the first frame, 2 Hz for the second frame. The SMAC record of HACHINOHE E-W in

Tokachi-Oki earthquake 1968 was adopted as the table input for the earthquake motion tests.

The Fourier spectra of original ground motion and measured shaking table motions of both test

frames are shown in Figure 1. •

.

: '.

3. TEST OF THE FIRST FRAME

The first frame was designed and constructed in accordance with the old A.I.J. Calculation Code for Reinforced Concrete Structure. The general view of the first frame is shown

in Photo. 1. The elevation and setting plan of this frame are shown in Figures 2 and 3,

respectively. The details of reinforcing steel bar installations and the section of columns

and girders are shown in Figure 4. ,

The total height of this frame was 4.2 m and the column height was 2.8 m. The widths

of frame are 8 m in long axis, 6 m in short axis. The long axis of this frame was corresponding to the excitational direction of the shaking table. Each column had a section of

45 cm X 45 cm, and the roof girders in both axes was 60 cm in height, 35 cm in width. The

footing of each column was 3 m x 3 m in plan, 80 cm in height. The footings were connected

by the tie beams in both axes. The section of tie beam was 80 cm in height, 40 cm in width.

The roof slab was 40 cm in thickness, and was strengthened by a beam of 60 cm x 25 cm in long

axis. The slab reinforcing steel bars of DIO were installed with 20 cm x 20 cm mesh in both

III-31
sides. The reinforcing steel bars of the roof girders and tie beams were anchored in

columns. Each footing was fastened to the shaking table by sixteen 2-inch steel bolts. The

total weight of the first frame was about 130 tons, and the weight of the roof was about 60

2 tons. The dead load axial stress of the columns was considered to be 7.5 kg/cm approxi2 mately. The 28 day compressive strength of concrete was 215 kg/cm . The tensile strength

of the steel bar (D19) was estimated as 5.59 ton/cm . The yield point for the steel bars

2 was estimated as 3.78 ton/cm .

The test of the first frame was started one month after the removal of slab supports.

Ahead of the vibration test, the micro-tremor of this frame was observed and the dominant

frequency was estimated about 5.8 Hz. •

'

At first, the free vibration frequency and damping of the frame without cracks was

measured by rectangular displacement table motions in amplitude of 1 mm. The frequency of

the frame was estimated as 4.76 Hz in this free vibration test. As the second step of the

test, the resonance test was carried out. The intensity of sinusoidal excitation was approximately 15 gals. During the excitation of 4.5 Hz sinusoidal motion, near the resonance frequency of this cracked frame, the properties of the frame changed. Therefore, the cracks

at columns were enlarged by this test. After this excitation, the properties of the frame

were continued to change with every sinusoidal excitation. Finally, the peak frequency

and the damping of the resonance curve was determined in 3.5 Hz, 0.02. However, the free

vibration frequency after the resonance test was measured 3.96 Hz. Then, the static tensile

test was carried out, and the frame spring was estimated as 30 ton/cm by loading the frame

laterally at the roof. Adopting this spring and the roof mass, the natural frequency would

be calculated to 3.5 Hz.

The first run of the simulated earthquake ground motion was carried out with an intensity of about 0.3 G. The Fourier spectrum for this run is shown in Figure 1. The chart

record at this run is shown in Figure 8. The maximum accelerations measured at the table

and the roof were about 0.34 G and 0.8 G. The maximum story displacement obtained was

approximately 2.9 cm. The strain gages at the foot of the column exhibited the yield of

reinforcing steel bars. From the instance of yielding, the frequency of the frame began

to decrease. The estimated running gain for this run is shown in Figure 7. Judging from

this gain, it is clear that the frame underwent nonlinear behavior and the frame frequency

decreased from 3.5 Hz to 2.5 Hz. This frequency change occurred from main shocks in about

111-32
5 seconds. The diagram of the acceleration-story displacement relation, which might be

considered the approximate relation between story shear and story displacement, is shown in

Figure 6. This diagram also described the degrading properties of the frame stiffness.

The free vibration frequency and the damping after this run was observed as 3.07 Hz and

0.07. During this run, the cracks of the column grew, but a failure of concrete was not

found.

The second run of the simulated earthquake ground motion was carried out with an intensity of about 0.5 G. During this excitation, the cracks were noticed and a few, small broke

pieces of concrete fell down. The maximum accelerations were approximately 0.78 G for the

table, 1 G for the frames. The maximum displacement was calculated as 4.19 cm. The free

vibration frequency and the damping after the second run were 2.25 Hz and 0.08. The results

of free vibration for this frame are shown in Figure 5.

4. TEST OF THE SECOND FRAME

This frame was designed and constructed in accordance with A.I.J. Code. The general

view of this frame is shown in Photo. 2. The elevation is shown in Figure 9. The details o

the reinforcing steel bar installations and sections of columns are shown in Figure 10.

The total height of the second frame was 4.4 m. The height and the section of columns,

and the frame widths corresponded to those of the first frame. The section of roof girders

in both axes were 100 cm x 45 cm in order to obtain a higher rigidity than the first frame.

The roof slab had the dimensions of 11 m in long axis, 9 m in short axis, and 40 cm in thick

ness. The roof was strengthened by three beams of 100 cm x 35 cm; two in short axis, one in

long axis. The footing of each column had the same dimensions in plan as the first frame,

and thinner thickness of 60 cm than the case of the first frame. The footing was connected

by the tie beams in long axis. There were no anchor steel bars in the columns. The slab

reinforcing steel bars of D13 were installed with 20 cm x 20 cm mesh in both sides. Each

footing was fastened in the same manner as the first frame. The total weight of the second

frame was about 180 tons, the weight of the roof which was considered twice as much as the

case of the first frame, was 130 tons approximately. The dead load axial stress in a column

was estimated about 15 kg/cm . The compressive strength of concrete on the test day was

194 kg/cm^. The yield point of the steel bar D19 was estimated as 3.35 ton/cm^, and the

2 tensile strength was 5.03 ton/cm . III-33
The test of this frame was carried out in four weeks after the removal of slab supports.

The micro-tremor of this frame was observed before the vibration test. The dominant frequency was estimated about 4.8 Hz. During the first measurement of vibration test, the free

vibrations were induced in the same way as in the test of the first frame. The dominant frequency was measured 4.0 Hz

•

The simulated earthquake ground motion was executed with the full stroke of the shaking

table. The Fourier spectrum of the table acceleration is shown in Figure 1. The maximum

acceleration and displacement were 0.43 G and 2.5 cm at the table, 0.45 G and 4.9 cm at the

frame, respectively. The maximum story displacement was calculated as 4.25 cm. All strain

gages on the reinforcing steel bars at the foot and head of columns increased as the steel -

yielded and went over the measurable range at the first shock of the excitation. Thus, the

strain measurement became impossible. From this excitation, the cracks at the top of columns,

which are shown in Photo. 3, were caused. At the foot of the columns, pieces of the concrete

broke off.

Using the table and the frame acceleration records, the estimated running gain was

obtained with a frame time of 2 sec and a shift time of 0.5 sec, which is shown in Figure 11.

By the running gain for this excitation, it was obvious that the second frame underwent nonlinear behavior. The frame frequency decreased from 4 Hz to 1.5 Hz during the simulated

earthquake motion test. The diagram of acceleration-story displacement relation which is

shown in Figure 12, also expressed the degrading properties of the frame stiffness. .

5. COMPUTER SIMULATION AND ITS CORRELATION WITH EARTHQUAKE MOTION TEST

The computer simulations were carried out for the first run test of the first frame and

for the second frame.

Judging from the diagram of the acceleration-story displacement relation, as shown in

Figure 6, the restoring force characteristics in the first frame are seen to possess a

bi-linear relation and story displacement. Moreover, the stiffness degradation would be

found in the estimated running gain and the diagram of acceleration-story displacement.

Thus, for the computer simulation, the bi-linear peak oriented mechanism, which is depicted

in Figure 13, was adopted. The time history correlation between computer simulations and

test results are shown in Figure 17. The computed relation between acceleration and story

displacement is shown in Figure 14. III-34
Judging from Figure 12, the second frame would be considered to have the tri-linear

restoring force characteristics. Therefore, the tri-linear peak oriented mechanism was

adopted for the second frame. The computation model is depicted in Figure 15. The XE3 was

the cross point of the skelton curve and the straight line which passed the origin parallel

to the second gradient of the skelton curve. The time history correlations between the

computer simulation and the test results are shown in Figure 18. The computed diagram is

shown in Figure 16.

Both computer simulations produced close correlations to the test results in the

acceleration time histories.

III-35
Photo. 2 General View of the Second Frame

Photo. 3 Damage of the top of column {2nd Frame)

III-36
F00BIE8 SPECTRA HACHINOtt E-W(l«.8(*.l

Fig. 1 Fourier Spectra

for Simulated Earthquake

Groiind Motion

SHUIK TABLE Q

r

1 juia;

j

1 °

^EICIIAlm DIUCTIOI—

»

Fig. 3 Setting Plan

of the First Frame

FMNt FOB IBAmOICCR

3^ UMf [MC TULC

Fig. 2 Elevation of the

First Frame

0 icr I**'*" 76HI h'Q <S*

FREE VISBATIOMS IIDUCEO BT STEP 01 SL PACEHtal

S

Fig. 5 Records of the

Free Vibration Tests

^ GIRDER

SHAKING TABLE

8000

Fig. 4 Structural Details

of the First Frame

Fig. 6 Relation between

Acceleration and Story

Displacement

111-37
Fig. 8 Chart Record of the 1st Run (1st Frame)

CSTinAUD MWIIIIG UK

Fig. 7 Estimated Rvinning Gain

(1st Frame)

ESTinirEi niiiiiHe uin

FRAME TIME 2sec SHIFT TIME OSsec.

lOseo

f-t It 1 1 n n t M ! 1

1

Ml, r

Fig. 9 Elevation of the 2nd Frame

Oo

8 ©

<NI o.^

C7)

|450

column

2-D19

6-D16

Fig. 11 Estimated Running Gain

(2nd Frame)

Fig. 10 Structural Details

of the 2nd Frame

III-38
ACCELERATION

(SSton)

Fig. 12 Relation Between

Acceleration and Story

Displacement

plOOOgal

Fig. 15 Computational p^g. Compured Diagram

Model for the 2nd Frame f^j- ^he 2nd Frame

III-39
HEASUKED TABLE ACCELESATIOH SECOItO

RESPOMSZ ACCBLERATIOM — lOOOgal

dott lin«t M«aur«d

full llnai coaput*d

XZ1:1.8CB d«npln9iS« OEli^ton — RESPONSE STORY DISrUkCEMEMT kj tlton/cB 5cm

Fig. 17 Time History Correlations for the First Frame

RESPONSE ACCELERATION —500g«l

RESPONSE STORY OISPIACEMENT

dott lln«i Masut«d

full lin«t coaput«d

XEli0.3ca. XE2il.3ca, X£3i2.S7c

0Eli30ton, 0B2tS0ton

tlton/cs

Fig. 18 Time History Correlations for the Second Frame

III-AO
SEISMIC RESPONSE OF PRECAST CONCRETE WALLS

James M. Becker

Carlos Llorente

Peter Mueller

Massachusetts Institute of Technology

Cambridge, Massachusetts

ABSTRACT

Large panel precast concrete structures have been widely utilized in major seismic

regions throughout the world. The seismic behavior of such structures is strongly dependent

upon the characteristics of both the horizontal and vertical connections. The limiting

behavior of precast systems, however, is basically dependent upon the horizontal connection.

The influence of horizontal connections can be studied in terms of the behavior of a simple

wall —a vertical stack of panels having only horizontal connections.

This paper reports on research into the seismic behavior of simple precast concrete

walls. The research was carried out through the development of computer-based modeling

techniques capable of including the typical behavioral characteristics associated with horizontal joints. The model assumes that all nonlinear, inelastic behavior is concentrated in

the connection regions and that the precast panels remain linear elastic. This assumption

allows the precast panels to be modeled as statically condensed 'super-elements' and the

connection regions as interface elements. The above modeling technique allows for nonlinearinelastic seismic analysis that is capable of handling both rocking type motions throughout

the height of the structure and slippage due to shear in the plane of the connection.

A series of parametric studies are presented to illustrate the potential influence of

rocking and slip on precast walls using both regular reinforcement and post-tensioning

.

These studies demonstrate the period elongation associated with the non-liner elastic rocking

phenomenon. Shear slip is found to occur only when friction coefficients are extremely low

or when the normal forces across the connections are low. This latter case occurs only in

low buildings or in the upper floors of tall buildings.

III-41
The paper concludes with a brief discussion of the design implications of these result

Particular attention is paid to the problems stemming from the force concentrations associated with rocking and shear slip.

KEYWORDS: Dynamic analysis; friction; post-tensioning ; precast concrete; seismic response;

shear walls.

III-42
INTRODUCTION

Large panel precast concrete structures have been widely utilized in major seismic

regions throughout the world. This form of construction raises many unusual problems in

terms of asei smic design because of the connection regions inherent in precast construction.

These connection regions are one of the main sources of the nonlinear and inelastic behavior

associated with the seismic response of these structures. At the same time, the likely concentration of nonlinear and inelastic behavior in the connection region facilitates the nonlinear modeling of these structures. This paper reports on research that uses a nonlinear

modeling approach to study the influence of the horizontal connection on the seismic response

of precast concrete walls.

The paper begins with the description of precast configurations typical of American

construction practice. An isolated or simple wall is abstracted from these typical precast

configurations and the behavioral basis for its nonlinear modeling is discussed. A finite

element model, using contact type elements for connections and linear-elastic substructures

for wall panels, is presented. The seismic response predicted by this model is presented in

the form of a series of parametric studies. The results of these studies are discussed in

light of their aseismic design implications.

SEISMIC BEHAVIOR OF SIMPLE PRECAST WALLS

The cross-wall structure illustrated in Figure 1 is typical of many of the precast

systems currently being used by the American precast concrete industry. The system is typified by the use of one-way precast prestressed floor planks spanning between the load-bearing

cross walls. As a basis for understanding the seismic behavior of these systems, they can be

idealized as being constructed of three basic structural elements (see Figure 1). The primary lateral load-carrying element is a simple wall, typified by an isolated load-bearing

cross wall. The simple wall is constructed of a vertical stack of solid panels and therefore

has only horizontal connections. The other lateral load-carrying element is a composite or

coupled wall. This element consists of simple walls joined together by vertical connections

or lintels. The third element is the floor system acting as a series of horizontal

III-43
diaphragms and tying together the various wall elements. It is important to note that the

horizontal connections found in both simple and composite walls also act, in general, as

connections in the precast floor systems.

Precast floor systems often have significantly reduced in-plane stiffness [2,4,10,19]

when compared to cast-in-place systems. This reduced stiffness and the difficulty in achieving adequate tensile continuity between floor elements poses problems in both dynamic analysis and design of large panel structures [19,24]. However, in order to concentrate the

ensuing discussion on the modeling and behavior of the simple wall, it will be assumed

that the floors as well as the foundations behave in a rigid fashion.

As is seen in Figure 1, the simple wall is often found as a primary lateral loadcarrying element in cross-wall systems. However, the interest in the behavior of simple

walls extends beyond this explicit use, for it serves as a bound for the behavior of composite walls (see Figure 2) and also allows for the detailed investigation of the possible

influence of the horizontal connection on seismic response.

Because of normal precast construction procedures and the effects of shrinkage and

creep, it is reasonable to consider horizontal connections to act as precracked planes. Thus

it is to be expected that crack opening will be concentrated in the region of the horizontal

connections. In this sense, the simple wall may be thought of as a precracked, reinforced or

prestressed concrete beam with known crack locations. Crack opening and closing along a

horizontal connection is usually referred to as rocking. However this term must be clarified

because the motion is not that associated with the rocking of rigid bodies; rather a smooth,

continuous opening and closing occurs in which plan sections no longer remain plane in the

regions immediately adjacent to the connection (Figure 3).

In the modeling of large panel precast concrete structures it is typical to consider

that solid panels remain linear elastic [15,17,20]. This assumption is reasonable in light

of the expected role of these elements in the structure's seismic response. Typical wall

panels can be assumed to behave in an elastic-brittle fashion in regards to in-plane normal

forces and whatever nonlinear-inelastic behavior that may exist can be effectively lumped

into the behavior of the connection regions.

The assumption of elastic-brittle behavior is based upon the general lack of confinement

of panel concrete in areas of maximum expected compressive forces. The typical single or

double layer of wall reinforcement (usually not tied together) cannot be expected to provide

I I 1-44
confinement for the concrete. However, the weakness to be expected in the connection regions

may protect the walls from compressive failure through their inability to transfer the necessary forces. This makes it possible to assume that the nonlinear-inelastic behavior is concentrated in the connection region and that the panels remain linear elastic.

Vertical reinforcement across the horizontal connection plane is generally kept to a

minimum, especially when compared to cast-in-place construction. Many techniques have been

developed for facilitating the necessary connection of vertical reinforcement (grouted voids,

splice sleeves, welded details, use of vertical connection regions, post-tensioni ng , etc.).

Few if any allow for the economic development of the amount of vertical steel normally

expected of load-bearing cast-in-place shear walls. While this lack of vertical reinforcement may seem to contradict normal aseismic design practice it may have two potential benefits. First, as suggested by Abrams and Sozen [1], it may provide a softer wall, in which

the seismic forces will be less than those associated with a stiffer, more highly reinforced

wall, and second the compressive force that can be developed across the potentially vulnerable horizontal connection is directly related to the amount of tensile steel in the wall.

The limited experimental data [8,9,12] on the compressive strength of the platform type

horizontal connections typical of American systems (Figure 4) indicates significantly reduced

strength when compared to the wall panel (in general less than 50-60%) and reduced stiffness

(50% or less of the wall's elastic modulus). To the author's knowledge, no experimental data

has been published on the cyclic behavior of horizontal joints under compression and/or shear

in a range that is meaningful to the ultimate seismic behavior of these joints.

The transfer of shear along the horizontal connection may be accomplished by the

following mechanisms: (a) coulomb friction, mN, (b) shear friction, associated with transverse reinforcement across the crack, yA^fy, and (c) by mechanical means such as keying and

metal details. The first two mechanisms may be accompanied by significant shear displacement

(slip) and may exhibit ductile behavior [7,13,14,25,26]. The latter mechanism requires

smaller shear displacements and may exhibit ductile or brittle behavior, depending on the

detail used.

It is typical to assume that the ultimate shear strength of a connection, considering

only friction transfer mechanisms, is the product of the friction coefficient, u, and the

sum of the normal force, N, and the steel yield force, A^f^. However, it should be noted

that the latter contribution ( uA f ) to the ultimate strength in shear can only be developed

s y III-45
when there is relative slip in the cracked plane. Thus the shear resistance is initially

given by the coulomb friction component (pN), which requires no relative slip to be active.

In general, the lower bound for the friction coefficient is considered to be around 0.7

[3,22,25]. However, under the unique conditions of smooth panel edges and grouted or dry

packed interfaces, along with cyclic loading, a lower value may be expected [14]. Values as

low as 0.4 [5] and even 0.2 [7] have been suggested.

Regardless of the particular shear transfer mechanism, global shear slip along the

entire length of the connection may be possible (Figure 3). It should be noted that such

global slip represents an unconfined yield mechanism when only coulomb friction is involved.

The actual displacement is only controlled by reversals of acceleration or by the stiffness

of secondary elements if present.

Modeling of Behavior , . >

;

While the connections in large precast panel buildings pose difficult design problems,

the concentration of nonlinear and inelastic behavior around connections and penetrations

facilitates the modeling of these structures. Based upon the discussion in the previous

section, the following modeling assumptions were made in order to study the seismic response

of the simple wall:

• Precast panels remain linear elastic —this assumes that the panels are either

solid or that they only have minor penetrations and that they are adequately

reinforced.

All nonlinear and inelastic behavior occurs in the connection regions.

The floor system acts as a series of rigid diaphragms.

The foundation is rigid.

The response of the simple wall was modeled using finite element techniques [3,6,15,18,

20]. The finite element idealization used is illustrated in Figure 5. Due to the assumption

of linear elastic wall panels, the degrees of freedom of a wall panel, which are not needed

to interconnect it to other components, can be eliminated through static condensation. This

leads to the repetitious use of 'super elements' or substructures (Figure 5).

i

i

III-46
Connection Modeling

There are two main sources of nonlinear and inelastic behavior in horizontal connections: opening of the connection and the associated shifting of the neutral axis and slippage along the connection plane (Figure 6). Both the connection panel interfaces along which

opening and slip most likely occur, and the connection material between the interfaces are

modeled by an interface or contact element [21] of infinitesimal height (Figure 7). Thus the

properties both the connection panel interface and the connection material are condensed

together to provide the constitutive relationship for the contact element. For simplicity

the connection material was assumed to be linear elastic in compression and shear (with

reduced modulus). Across the connection panel interface zero tensile strength (vertical

reinforcement is modeled separately) and a rigid-plastic coulomb friction shear transfer

mechanism were assumed. The combined stress-deformation relationship of the contact element

is thus nonlinear-elastic in compression and tension and elasto-plastic in shear (Figure 8).

Physical slippage in the connection-panel interface occurs when the shear stress exceeds

the shear strength. The shear strength is given by the coefficient of friction times the

compressive stress across the connection (Figure 8), thus coupling the shear behavior to the

behavior normal to the connection. This implies that the level at which slip occurs at a

specific point will not remain constant, but will vary with the level of overturning moments.

This problem was handled by assuming that the slip level remained constant within a given

time step.

The deformation displacement relation for the contact or interface element is as follows

(Figure 6): ^ {^}- . • —

where Av and Au are the axial and shear deformations of the connection respectively, and

0 (1 + r) 0 (1 - r) 0 -(1 - r) 0 -(1 + r)'

(1 - r) 0 (1 - r) 0 -(1 - r) 0 -(1 + r) 0

u^ = [u^ Vj^ U2 V2 U3 V3 u^ v^]

r = 2s/L.

The restoring nodal forces Fj-(t) for a typical element are evaluated as follows

' ' III-47

Fr(t) =
and the stress are given by

aft'! o(t - At) + Ao

'K ^ J T(t - At) + At

wh PTP G ( t ^ axial stress at time t

shear stress at time t

Aa increment in axial stress during time interval At

At increment in shear stress during time interval At

b thickness of connection

L length of interface element

The nodal restoring forces F^(t) were evaluated numerically using two integration points per

element. Four and eight interface elements corresponding to 8 and 16 integration points,

respectively, were used along a connection, and the smaller number was found to provide

sufficient accuracy. The vertical reinforcing steel was modeled using elasto-plastic truss

elements between panel edges. In the case of ungrouted post-tensioning bars a truss element

connects the top of the structure with the base. In this case the bar is only allowed to

yield in tension; however, yielding of such a bar seldom, if ever, occurs. Reinforcing and

post-tensioning steel was assumed to provide only vertical continuity; its contribution to

shear transfer was not considered.

Time Step Integration . ,

The dynamic analysis was carried out using an integration scheme based on the central

difference method. This method is computationally very efficient within each time step; in

conjunction with diagonal mass and damping matrices, there is no need for solving a system of

equations. A global stiffness matrix does not need to be assembled and all operations can be

performed at the element level. Unfortunately, the method is only conditionally stable, that

is for a time step . ^

At^ &t„r = ^ '

where T^ is the smallest period of the system. Because precast panel buildings are fairly

rigid structures, the time step required for numerical stability can be very small. In the

III-48
parametric studies reported in the next section, a time step in the order of 1/1200 =

0.000833 sec was required. Mass distribution is also a potential problem, since, to avoid

infinite frequencies, masses must be allocated to all degrees of freedom.

SEISMIC RESPONSE

Using the analytic model previously described, a series of parametric studies have been

carried out to examine the influence of horizontal connections on the seismic response of

simple walls. Specifically, the effects of such behavioral aspects as rocking and shear

slip on the magnitude of the seismic force levels and deformations are examined. Studies

were carried out on both five-and ten-story walls, with different reinforcement patterns for

providing vertical continuity and for various earthquake records.

The simple wall was constructed out of panels 7.32 meters wide by 2.44 meters high and

20 cm thick. (Figure 9) A connection region with a height of 30.5 cm was assumed. The

wall panels had a composite modulus of 25.2 x 10^ kPa (v = 0.15) and the connection modulus

was one half that of the wall panels. Shear slip was assumed to have a secondary stiffness

of one hundredth that of the connection's primary shear stiffness. The lumped floor mass was

68.3 tons (51.7 tons at roof level) (see Figure 10) and the gravity load for each floor level

was 670.4 kN (507.1 kN at the roof level).

Six different cases of vertical continuity across the horizontal connection were

considered. The first is a linear elastic case, which serves as a reference for response

comparisons. For this baseline case the ten-story structure had a fundamental period of

0.62 seconds, while the five-story structure had a fundamental period of 0.17 seconds. Two

cases are post-tensioned identically by ungrouted bars, but have different friction coefficients (p = 0.2 and 0.4). In both the 5- and 10-story walls a uniform prestress of 1400 kPa

was applied. The three remaining cases have vertical continuity provided by Grade 40 reinforcement, either evenly distributed along each connection level (0.25% and 1.00%), or

concentrated at the ends of the panel (0.5%). All three reinforced cases have the same

friction coefficient (u = 0.4).

The majority of the parametric runs were made using an artificial earthquake generated

to match the Newmark-Blume-Kapur response spectra for 2% damping. To provide a comparison

for these results, several analyses were made using the El Centro (N.S. Component, 1940) and

Taft (Kern County N21E, 1952) records. The response spectra of all three earthquakes for 5% III-49
critical damping (which was used in the analyses) are given in Figure 11. Table 1 presents

a summary of the various parametric studies upon which the following discussion of seismic

response will be based.

Figure 12 presents three different cases that are intended to illustrate the basic

characteristics of both the rocking and the shear slip phenomena. All three cases refer

to the ten-story structure being subjected to the artificial earthquake with a peak acceleration of 30% g. The horizontal connection was modeled with eight contact elements (16

integration points) for these three cases. Case 1 is the linear elastic reference case.

The only difference between the present wall and a cantilever beam is the reduced stiffness

of the connection. In this case the maximum base shear was 39 54 kN. As was to be expected,

the axial strain and shear stress distribution corresponds to normal beam theory. The axial

strains and shear stresses are given for the time of maximum overturning moment.

In Case 2 tension across the connection is solely transferred through ungrouted posttensioning bars. The shear friction coefficient is 0.4. The maximum base shear has dropped

to a value of 2476 kN. This decrease in maximum base shear is attributable to an effective

elongation in the fundamental period of the wall and an ensuing shift in the response spectra. This elongation of the fundamental period is easily observed by comparing the time

histories of Case 1 and Case 2. The elongation of the fundamental period is caused by a

rocking motion. This rocking motion is evident from the axial strain distribution presented

in Figure 12 for the time of the maximum crack length. In the tension region the axial

strain is obtained by dividing the crack opening by the height of the connection. As can be

easily seen, plane sections no longer remain plane in the region of the connection. The

shear stress distribution shows effects of local shear slip relieving potential stress concentrations at the tip of the crack [11]. Note that in the region immediately adjacent to

the crack tip the shear stress is directly proportional to the compressive axial strainCase 3 is the same as Case 2 with the exception that the shear friction coefficient is

now 0.2. In this case global shear slip is occurring at the time of maximum crack opening.

Global shear slip has two significant effects. First shear slip provides both a source of

energy dissipation and a force isolation mechanism. This can be observed in the time

history of Case 3. The higher frequency components observed in Case 2 are missing and the

maximum base shear is reduced to 1940 kN (the product friction coefficient times dead load

plus post-tensioning is 1787 kN) . Second the shear stress distribution is now basically

III-50
proportional to the compressive stress distribution and becomes a triangular stress block in

accordance with the simplified assumption of linear elastic behavior in compression. The

fact that the peak compressive and shear stresses coincide and are concentrated near the

connection ends, when global slip occurs, enhances the changes of shear compression type

failure in the connection region.

It Is interesting to note that in all three cases the maximum compressive axial strains

are almost identical. The major difference is in the concurrent shear associated with these

compressive strains. Also note the magnitude of the shift in the location of the neutral

axis between the first case and the latter two cases. Figure 12 has illustrated the potential basic differences between precast and cast-in-place walls. In the following sections

the influence of horizontal joint behavior on the seismic response is discussed in greater

detail. , , . "

.

. ....

The Effect of Rocking Motion on Response

As was mentioned above, when a precast wall undergoes a pure rocking motion (i.e., no

global shear slip) it exhibits a softening behavior with little or no energy dissipation.

(It is assumed that the energy dissipated by local shear slip is negligible). As long as

the connection material properties remain nonlinear-elastic, rocking is a nonlinear-elastic

phenomenon which lends only to an apparent elongation of the wall's fundamental period.

Results typical of rocking motion are summarized in Table 1(a) and In Figure 13. The

results in Figure 13 have been normalized with respect to the linear elastic response for

25% peak acceleration. As can be observed, there is a decrease in the maximum force levels

attained. This decrease in forces can be explained by noting that any elongation in the

fundamental period will cause a downward shift in the acceleration spectrum (Figure 11),

leading to lower forces. The opposite is true for deformations, for which an Increase in

period will only cause Increased deformations relative to the elastic response. This trend

can also be observed in the results for both the El Centro and Taft earthquakes (see Figure

11 and Table 1(d). It must be noted, however, that for different earthquakes or for stlffer

buildings the period elongation associated with rocking could lead to the opposite results,

that is an increase in force levels. . . ,

.

A major concern In the use of ungrouted post-tensloning systems In aseismlc design

is the possible yielding of the post-tensloning bars. Figure 14 shows the time history for

III-51
the post-tensloning bar closest to the walls edge in the 30% g rocking case (note: this is

the same as Case 2 in the earlier discussion). Even for this extreme case, where the nonlinear-elastic material assumption for the horizontal connection is of questionable validity,

the post-tensioning bar is not stressed close to its yielding level.

The Effect of Shear Slip on Response

The magnitude of the friction coefficient, together with the compressive resultant due

to gravity loads, post-tensioning (if present), and the stressing of mild reinforcement (if

present), determine whether global shear slip will occur. The effect of global slip is to

limit seismic force levels and to dissipate energy as has been previously noted [5,20]. To

illustrate this global slip and shear limiting phenomenon. Figure 15 presents selected

response time histories for both the post-tensioned (p = 0.2), and the mild reinforcement

(1.0%) cases. Note the uniform peaks of the shear force time history of level 4 in the mild

reinforcement case and how each peak is associated with some level of global slip. This same

isolation phenomenon can be observed, to a smaller extent, at the base of the post-tensioned

case. As can be seen in Figure 16, the cases with the lower friction coefficient (y = 0.2)

experience global slip along all connection levels and consequently have the lower force

levels. The post-tensioned cases with the higher friction coefficient (u = 0.4) undergo

pure rocking motion only with localized slip occuring; thus their force levels are the

highest. It is interesting to note, in Figure I6c, that the peak global slips, for the

y = 0.2 post-tensioned cases, are located from the third through the fifth levels. This can

be attributed to a combination of higher mode contributions and a decreasing shear strength

with height.

The shear strength of the three mildly reinforced cases for both 5- and 10-story walls

lies between the shear strength of the two post-tensioned cases in the lower levels, and

below in the upper levels. As a result, these three cases experienced global slip in the

upper levels and only localized slip in the lower levels (see Figure 16 a & b). The resulting base shears and moments lie between the two post-tensioned cases. The magnitude of

global slip indicates that mild reinforcement will be engaged in developing a shear friction

mechanism and that this should be accounted for in future modeling. The yielding of the

mild reinforcement also contributed to a diminished force level due to some limited energy

dissipation.

III-52
To illustrate the effect of rocking and global shear slip on the overall response, the

time histories for both roof deflection and overturning moment are presented in Figure 17 for

the 10 story post-tensioned ( iJ= 0.2) case subjected to the El Centro earthquake (25% g) . In

addition to these time histories, the figure shows two load-deformation (base overturning

moment-roof deflection) plots for selected segments of the time histories. The load deformation plot for the response between seconds 1 and 2.1 shows the possibility for hysteretic

damping through global shear slip. This is illustrated by the full hysteresis loop. It must

be noted, however, that this form of inelastic deformation represents an unconfined yield

mechanism and could lead to problems not dissimilar to those of soft story systems. In

particular, degradation of a horizontal joint directly threatens the overall stability.

The second load-deformation plot shows the response between seconds 4 and 5. This plot

demonstrates the nonlinear elastic nature of pure rocking motion. During this period no

global shear slip occurs, and thus the load-deformation plot exhibits a closed hysteresis

loop, in which no energy is dissipated.

Effect of Rocking and Slip on Force Distribution

Rocking motion and global shear slip have a significant effect on the distribution of

forces along the horizontal connection and within the precast wall panels. As can be seen

in Figure 16 a and c, cracked lengths on the order of 70% of the total connection width were

obtained for the 10-story wall subjected to earthquakes with a peak acceleration of 25% g.

When coulomb friction is assumed, the transfer of shear occurs only in the unopened portion

of the connection in direct relationship with the normal stresses.

The relationship between shear and normal stresses is illustrated in Figure 18. It

presents a series of time histories for the 10-story post-tensioned (y = 0.2) case subjected

to the El Centro earthquake (25% g). This particular case was run using eight contact elements (16 integration points) for the horizontal connection. At the bottom of Figure 18 the

base shear and cumulative base shear slip time histories corresponding to those in Figure 17

are presented. A^sove the shear stress time histories for the two extreme integration points

(pts. 1 and 16) and one integration point near the center (pt. 8) are shown. Superimposed on

the shear time histories are the envelopes of shear strength (i.e., shear slip level ya(t)).

When global slip occurs, the shear strength envelopes and the shear stresses coincide. This

condition occurs only three times in the ten seconds of the analysis. Figure 19 presents the

III-53
shear stress - shear deformation hysteresis loops for the same integration points. Because

the axial stress varies and is not necessarily in phase with the shear stress, the hysteresis

loops in Figure 19 do not show the simple elastoplastic behavior associated with constant

axial stress levels. -

The concentration of the peak shear and axial stresses near the connection ends leads to

a thrusting action that can create severe biaxial stress conditions in the connection as well

as in the panel corners. Of particular concern is the upper corner of the panel, in which

significant tensile stresses are found, indicating the need for care in the design of panel

reinforcement. This thrusting condition is illustrated in Figure 20.

Figure 20 presents the flow of principal stresses in the bottom two panels for the posttensioned case discussed above (10-story wall. El Centro, 0.25 g, PT, p = 0.2). The shown

deformed configuration occurs one cycle after the wall has undergone the maximum global slip

in the opposite direction. The stresses shown by dashed lines are not expected to occur in

the actual structure. These stresses can be attributed in part to the distribution of mass

in the model. Upon opening, the corner of the wall panel should be essentially stress free;

however, because there are lumped masses and consequent inertia forces in these wall panel

corner nodes, significant stresses are present.

CONCLUSIONS .

'

.

'

This paper has examined the role of the horizontal connection in the seismic response

of precast concrete panel buildings. The study of a simple precast wall allowed for the

isolation of the effects of nonlinear and inelastic horizontal connection behavior.

The seismic response of a simple precast wall is found to be mainly governed by a

nonlinear-elastic rocking phenomenon. The opening of the horizontal connection, associated

with rocking, leads to a progressive softening of the structure with increasing excitation

level. Thus, rocking leads to an elongation of the apparent fundamental period of the wall,

with a consequent increase or decrease in the seismic response, depending upon the nature of

the earthquake. . , ,

It was found that global slippage between two panels would occur only in limited situations. These situations included extremely low coefficients of friction, the upper levels of

non-post-tensioned walls and shorter walls. It was concluded that global slippage could not

normally be counted upon for seismic force isolation. It should be noted that in all of the

III-54
reported work the possible effects of vertical acceleration have not been considered. Such

effects may have a significant influence when considering behavior dependent upon gravity

loads

.

Rocking coupled with local and global shear slip leads to significant force concentrations in both the corners of panels and the ends of connections. It is felt that such force

concentrations, unless specifically designed for, will lead to a progressive deterioration of

the panel corners and connection ends in a significant earthquake. The concern for these

force concentrations, along with a hesitancy to use a unconfined yield mechanism to limit

forces, suggests caution in relying on beneficial effects of global shear slip mechanisms in

the design of simple walls. It is recommended that, unless specific design provisions are

made, the force levels in horizontal connections should be kept significantly below ultimate

capacity for design earthquakes.

The work presented in this paper is based on the limited experimental data available

about the behavior of panelized buildings. This data calls for simple and therefore also

questionable assumptions with respect to material behavior (non-degrading shear-slip relationship, etc.). Accordingly, the computer studies presented are limited in their scope.

However, it is felt that the basic behavioral aspects presented are correct and point to the

major problems in the aseismic design of panelized buildings. While the presence of weak

horizontal joints is a drawback relative to cast-in-place shear walls, this does not need to

be true for vertical connections. Studies indicate [16,23] that the presence of vertical

connections in panelized buildings may even improve their seismic performance. It is obvious

that, for such construction to achieve wide acceptance in seismic regions, extensive experimentation is required along with continuing theoretical studies.

ACKNOWLEDGMENTS , ,

This research on the Seismic Resistance of Precast Concrete Panel Buildings is being

conducted at the Massachusetts Institute of Technology and has been sponsored by the National

Science Foundation under Grant ENV 75-03778 and Grant PFR-7818742. In addition, the authors

would like to acknowledge the assistance of Joseph Burns, Maria Kittredge, and Sally Brunner

in the preparation of this paper.

III-55
REFERENCES

[1] Abrams, D. P., and Sozen, M. A., "Experimental Study of Frame-Wall Interaction in Reinforced Concrete Structures Subjected to Strong Earthquake Motions," Civil Engineering

Studies, Structural Research Series No. 460 , University of Illinois, Urbana.

[2] Anicic, D., "Experimental Determination of the Stiffness of the Floor Diaphragam Loaded

by Horizontal Forces," unpublished, personal communication, October, 1977.

[3] Becker, J. M., and Llorente, C, "Seismic Design of Precast Concrete Panel Buildings,"

Proceedings of a Workshop on Earthquake-Resistant Reinforced Concrete Building

Construction , Vol. 3, University of California, Berkeley, July, 1977. .r

[4] Bernander, K. G. , and Liefendahl, T. , "Shear Force Transmission Through Joints in

Precast Hollow Floors Subjected to Bending in the Plane of the Floor," Nordisk Betong , No. 6, 1976, pp. 23-26 (in Swedish).

[5] Brankov, G., and Sachanski, "Response of Large Panel Buildings for Earthquake Excitation in Nonelastic Range," Proceedings of the Sixth World Conference on Earthquake

Engineering, New Delhi, India, 1977.

[6] Frank, R. A., "Dynamic Modeling of Large Precast Panel Buildings Using Finite Elements

with Substructuring, " MIT Publication No. R76-36 , Department of Civil Engineering,

Massachusetts Institute of Technology, Cambridge, Massachusetts, August, 1976.

[7] Hanson, N. W. , "Seismic Test of Horizontal Joints," Design and Construction of Large

Panel Concrete Structures , Supplemental Report C, prepared for U.S. Department of

Housing and Urban Development, Portland Cement Association, January, 1979.

[8] Iyengar, S. , and Harris, H. G., "Behavior of Horizontal Joints in Large Panel Precast

Concrete Buildings," Report No. M78-1, Structural Models Laboratory, Department of

Civil Engineering, Drexel University, Philadelphia, Pennsylvania, February, 1978.

[9] Johal, L. S., and Hanson, N. W. , "Horizontal Joint Tests," Design and Construction of

Large Panel Concrete Structures, Supplemental Report B, prepared for U.S. Department

of Housing and Urban Development, Portland Cement Association, November, 1978

[10] Lewicki, B., Building with Large Prefabricates , 1st ed., Elsevier Publications Co.,

London, 1968.

[11] Llorente, C, Becker, J. M. , and Roesset, J. M. , "The Effect of Non-Linear-Inelastic

Connection Behavior of Precast Panelized Shear Walls," paper presented at the ACI

Convention, Toronto, Canada, April, 1978, submitted for inclusion in Session Volume

on Mathematical Modeling of Reinforced Concrete Structures (ACI 442). '

[12] Lugez, J., and Zaravcki, A., "Influence of Horizontal Joints on the Resistance of

Prefabricated Panel Elements of Bearing Wall," Cahiers du Centre Scientifique et Technique du Batiment , No. 103, Paris, October, 1969.

[13] Mattock, A. H. , "Shear Transfer Under Monotonic Loading, Across on Interface Between

Concretes Cast at Different Times," Final Report, Part 1 , Department of Civil Engineering, University of Washington, Seattle, Washington, September, 1976.

[14] Mattock, A. H. , "Shear Transfer Under Cyclically Reversing Loading, Across an Interface Between Concretes Cast at Different Times," Final Report, Part 2 , Department of

Civil Engineering, University of Washington, Seattle, Washington, June, 1977.

[15] Mehlhorn, G. , and Schwing, H. , "Tragverhalten von aus Fertigteilen Zusammengesetzten

Scheiben," Forschungsberichte aus dem Institute fur Massivbau der Technischen

Hochschule , No. 33, Darmstadt, 1976.

III-56
[16] Mueller, P., and Becker, J. M. , "Seismic Characteristics of Composite Precast Walls,"

Proceedings of Third Canadian Conference on Earthquake Engineering , Montreal, Canada,

June, 1979.

[17] Petersson, H. , "Analysis of Building Structures," Chalmers University of Technology,

Department of Building Construction, Gothenburg, 1973.

[18] Petersson, H. , and Backlund, J., "SERFEM: A computer program for in-plane analysis of

plates by the finite element method," Chalmers University of Technology, Department of

Building Construction, Gothenburg, 1974.

[19] Polyakov, S. , Design of Earthquake-Resistant Structures , 1st ed., MIR Publishers,

Moscow, 1974.

[20] Powell, G., and Schricker, V., "Ductility Demands of Joints in Large Panel Structures,"

ASCE Fall Convention, San Francisco, October 1977, Preprint 3022.

[21] Schafer, H. , "A Contribution to the Solution of Contract Problems with Aid of Bond

Elements," Computer Methods in Applied Mechanics and Engineering , Vol. 5, 1975.

[22] Suenaga, Y., "Study of Rationalization of Horizontal Dry Joints in Box-Frame-Type

Precast RC Structures," Bulletin of Faculty of Engineering , Yokohama National

University, Yokohama, Japan, March 1976.

[23] Thiel, V., "The Role of Vertical Connections in the Seismic Response of Precast

Composite Walls," thesis presented to the Massachusetts Institute of Technology, at Cambridge, Massachusetts, in 1979, in partial fulfillment of the requirements for the

degree of M.S. in Civil Engineering.

[24] Unemori, A. L. , Roesset, J. M. , and Becker, J. M. , "Effect of In-Plane Floor Slab

Flexibility or the Response of Shear Wall Buildings," paper presented at the ACI

Convention, Houston, Texas, October/November, 1978, submitted for inclusion in Session

Volume on Mathematical Modeling of Reinforced Concrete Structures , (ACI 442).

[25] Verbic, B., "Behavior of Joint Elements of a Building Constructed of Large Panels by

the System 'Vranica' Under Quasistatic Cyclic Loading," Institute za Materijale i Konstruckcije , Sarajevo, April, 1977.

[26] White, R. N. , and Gergely, P., "Shear Transfer in Thick Walled Reinforced Concrete

Structures Under Seismic Loading," Final Report , NSF Grant No. ATA73-03178, Department

of Structural Engineering, Cornell University, Ithaca, New York, May, 1978.

III-57
TABLE 1 SUMMARY OF PARAMETRIC STUDIES -

Case Shear Moment Deflection Cracked Length Global Slip

(klO (kI^I-m) (cm) % Level (mm) Level

10/og peak accel. T 1 1 ft ±DZ / J z9 Q 7 . y / 29.0 0 15%g peak accel. ZXHy 91 A Q A^ A 9/i 50.0 0 20%g peak accel. 1 /U / 9 ^ A 1 A D . 0 J 64.0 0 25%g peak accel. Z jZ J

9 9 7 9^ jZ / JO 0 . y 0 70.0 0 30%g peak accel. Z4 / U

9 ^ 0 An JoZUU o. bU 78.0 0

(a) Rocking Study., lU btory Wall , .PI , y - U. 4, Artificial Earthquake •

Case Shear Moment Deflection Cracked Length Global Slip

(kN-m) (cm) % Level (mm) Level

Linear Elastic 3295 47000 5.50

r i , \i - 0.1 29950 6.06 65.0 0 2.9 Q

ri , y - U.

4

0 9 9 9 Z jZ J 32736 6.96 70.0 0 K/L, . Z ^^ , y — U.'t 1 Q 1 ly 1

J

28146 6.08 71.1 0 4.7 Qy K/ u , i . u/o , y

— u , H on 9 ZU jZ 31991 5.86 62.8 0 8.1 y

p/r* 11 — n A K/ L, , . J/o , y - U . 41 QQA 30511 5.85 70.8 0 5.2 Qy

(b ) Reinf oirceinent oLUuy, lu Story Wall, Artificial Earthquake, '.25%g

Case Shear Moment Deflection Cracked Length Global Slip

(kN) (kl^-m) (cm) % Level (mm) Level

Linear Elastic 1533 12598 0.54

ri , y - U .

Z

T 1 9 ^ 9422 0.57 13.9 0 0.9 1 Fi , y - U.

4

T A £1 7 14o / 11751 0.53 29.0 0 R/L , • Z3^ , y = (J . 4 1211 10490 0.88 38.1 1 3.4 4 K/C , 1 . , y

= U .

4

1185 9590 0.81 24.3 0 3.4 4 R/C , 0 . 5/i , y

= 0.4 1242 9778 0.83 31.6 0 3.4 4

(c) Reinforcement Study, 5 Story Wall, Artificial Earthquake, 25%g

Case Shear Moment Deflection Cracked Length Global Slip

(kN) (kl^-m) (cm) % Level (mm) Level

Artificial y

= 0.2 1894 29950 6.06 65.0 0 2.9 3 Artificial y = 0.4 2323 32736 6.96 70.0 0 El Centre y = 0.2 1860 29035 6.90 70.2 0 6.7 5 El Centre ' y = 0.4 2640 36000 8.44 72.0 0 Taft y = 0.2 1790 29700 6.28 68.0 0 2.6 4 Taft y = 0.4 2012 31640 7.45 75.2 0 (d) Earthquake Study, 10 Story Wall, Post-Tensioned , 25%g

III-58
COMPOSITE

(q)

FIGURE 1 BASIC STRUCTURAL ELKIENTS

OF LARGE PANEL PRECAST

CONCRETE BUILDINGS

FIGURE 2 TYPICAL PRECAST WALL

CONFIGURATIONS (Shading

denoces a simple wall)

ROCKING SHEAR SLIP

W////////////////////.

DRY PACK

VERTICAL

CONTINUITY REINFORCING

PRECAST PLANK

—BEARING STRIP

FIGURE 3 ROCKING AND SHEAR SLIP

MECHAJIISMS

FIGURE 4 TYPICAL PLATFORM

CONNECTION

NODES - CONDENSED OUT

NODES - RETAINED AU I

FIGURE 5 FINITE ELEMENT MODEL FIGURE 6 CONNECTION ELEMENT

III-59
1

CONNECTION ELEMENT

|u' 2' CONTACT ELEMENT l'

3

4' {au) = (au'}

AU

FIGURE 7 CONTACT ELEMENT FIGURE 8 SHEAR STRESS-SHEAR DEFORMATION AND SLIP RELATIONSHIP

•—5l,7tons

68 3 tons par floor

i ; :

LUMPED MASS

AT NODES

4.27 tons 8.54tons

FIGURE 9 10-STORY SIMPLE WALL FIGURE 10 MASS DISTRIBUTION IN SIMPLE WALL

(wall thickness = 0.20m)

300

.05 .10 I NATURAL PERIOD, SEC

FIGURE 11 RESPONSE SPECTRA, 5% DAMPING,

l.Og PEAK ACCELERATION

III-60
CASE I • LINEAR ELASTIC

CASE 3 • /I =0.2 . f,= 0

FIGURE 12 COMPARISON OF AXIAL STRAIN AND SHEAR STRESS AT TIME OF MAXIMUM

BASE SHEAR, AND BASE SHEAR TIME HISTORIES, 10 STORY, P.T.,

ARTIFICIAL 0.30g.

III-61
PEAK ACCELERATION %g

FIGURE 13 EFFECT OF ROCKING IN A SIMPLE WALL, 10 STORY

P.T. , y = 0.4, ARTIFICIAL EQ,

"6 12 YIJLD LEVEL

TIME SEC

FIGURE 14 RESPONSE OF TYPICAL POST

TENSIONING BAR, y = 0.4,

ARTIFICIAL 0.3g.

REINFORCEMENT- 1.0% ,/x = 0.4

E

Eo

UJ -5

UJ -104

2

0

to -2-4 1400

1 g 0 <

I UJ

' -700

^ E uil > Q-

-14004

2

POST-TENSIONED, /a =0.2

0 5 10

5 10 TIME (SEC)

10

5

0

-5

-104

2

0

=i « -2

-4 1400 5 7D0

0

UJ UJS -700

-1400

4

< a. °

trt _2-4

1400 f 700

0 m -700

-1400

5 10 TIME (SEC)

FIGURE 15 SELECTED TIME HISTORIES, 5-STORY WALL, ARTIFICIAL 0.25g

III-62
POST-TENSIONEO, ^'02

REINFORCEMENT - I.Orfc . ^= 0.4 REINFORCEMENT- 25% , U'O

CONCENTRATED REINF- 5%, u

EL CEHTRO .PT '^•02

TAFT.PT fJ.'OZ

SHEAR SLIP (mm) SHEAR SLIP (rr

RElNfOHCEMENT-l 0 % , ^. 04 — REINFORCEMENT- 25% , D4

CONCENTRATED REf*- 5%. /x- 04

POST-TENSiONEO. /J.' 04

-- POST-TENStONEO. IJ.-02

y 5

ARTIFICIAL, PT ^-02

ARTIFICIAL. PT ;x'0 4 EL CENTRO. PT ;i»0 2 EL CENTRO .PT /i.'04 y TAFT. PT ^•0.2

TAFT. PT- fJ^O.A

CRACKED LENGTH!'/. o< pan«l «iaih) CfiACK LENGTH I'/'Of o' POn«l width) CRACKED LENGTH I •/. ot pawl «i(11h)

(O) EFFECT OF REINFORCEMENT10 STORY

(b) EFFECT OF REINFORCEMENT—

5 STORY

(C) EFFECT OF EARTHQUAKE

10 STORY

III-63
III-64
SEISMIC RESPONSE ANALYSIS OF THE ITAJIMA BRIDGE

THROUGH USE OF STRONG MOTION ACCELERATION RECORDS

Masamitsu Ohashi

Toshio Iwasaki

Kazuhiko Kawashlraa

Public Works Research Institute

Ministry of Construction

ABSTRACT

In analyzing the seismic behavior of highway bridges constructed on soft soil deposits,

it is important to take into account soil-structure interaction effects. In this paper, the

seismic response of a bridge pier-foundation is analyzed from earthquake acceleration records

taken simultaneously from the pier crest and the ground surface near the bridge. Four motions

were used in the analysis, i.e., two were induced by two earthquakes with magnitudes of 7.5

and 6.6, and two by their aftershocks. In the former two earthquakes, the maximum accelerations were 186 and 441 gals on the ground surface, and 306 and 213 gals on the pier top.

Analyses of frequency characteristics of the motions showed that the predominant frequencies

of the pier-foundation were always almost identical to the fundamental natural frequency of

the subsoil. Analytical models were formulated to calculate the seismic response of the

pier-foundation assuming the subsoil and pier-foundation to be a shear column model with an

equivalent linear shear modulus and an elastically supported beam on the subsoil, respectively. Bedrock motions were computed from the measured ground surface motions and then

applied to the bedrock of the analytical model. The seismic responses of the pier-foundation

were thus calculated and compared with the measured records and produced a good agreement.

KEYWORDS: Bridge-pier foundations; bridge seismology; earthquake frequency characteristics;

ground surface accelerations; foundation structure response.

III-65
INTRODUCTION

In the past, numerous highway bridges have suffered extensive damage due to strong

motion earthquakes. Seismic damage to bridges consisting of simply supported girders or

trusses rested on massive piers and abutments were commonly caused by foundation failures

resulting from excessive ground deformations and/or loss of stability and bearing capacity

of the foundation soils. As a direct result, the substructures often tilted, settled, or

sometimes overturned, and these large displacements of the supports caused relative shifting

of the superstructures, induced failures of the bearing supports, and even caused dislodging

of the spans from their supports.

It has been well recognized from these evidences that the influences of surrounding subsurface ground are very important for the seismic responses of foundations deeply embedded in

the ground, and considerable interests were concentrated on the soil-structure interaction

effects of such structures through model experiments and theoretical analyses. However very

limited researches have been undertaken in studying seismic responses of actual foundations

during high intensity seismic excitations since little data have been available for such purposes. This investigation presents the results of observation of seismic motions on a bridge

pier and on the ground surface nearby with high accelerations which were measured during four

earthquakes, and also correlates the pier motions measured with the pier motions analyzed

using the ground surface acceleration records.

STRUCTURAL AND SITE CONDITIONS OF THE BRIDGE

The Itajima Bridge studied is a five-span simply-supported plate girder bridge as shown

in Figure 1. The strong-motion acceleration observations have been conducted since 1966 at

a crest of one of the piers and on the free field ground surface located about 400 m apart

from the pier. Two SMAC-B2 type accelerographs are set at both sites to measure accelerations in the longitudinal, transverse and vertical directions of the bridge axis.

Ground surveys were performed at both sites and information on soil profile, N-value of

standard penetration test, and shear wave velocities were obtained. Figure 2 shows the soil

profiles and N-values at both sites. It is recognized from the results that the ground conditions are essentially the same between the two sites, i.e., the soil profiles consist of

upper soft alluvium loam to fine sand formations with the averaged N-value of approximately

III-66
7 and lower stiff diluvium gravel formations with the averaged N-values of 30 or more. The

shear wave velocities of the upper and lower formations were estimated to be approximately

130 and 480 m/sec, respectively. Figure 3 shows the geological structures around the Itajima

Bridge. The stiff diluvial gravel formations are recognized to be underlaid by mesozoic cretaceous formation. According to Figures 2 and 3 it can be recognized that the ground conditions are continuous and almost identical between the two sites. The gravel formations were

assumed to be the bedrock at the sites in calculating seismic responses of the ground and

foundation in the following paragraph.

ANALYSES OF STRONG-MOTION ACCELERATION RECORDS

Strong-Motion Acceleration Records

Four simultaneous strong-motion acceleration records have been obtained at the Itajima

Bridge as summarized in Table 1 which were induced by four earthquakes, i.e., main and after

shocks of both the Hyuganada Earthquake of April 1, 1968 and the Bungosuido Earthquake of

August 6, 1968, which are designated herein as A, B, C and D Earthquakes, respectively.

Figures 4, 5, 6 and 7 show the acceleration records thus obtained in the longitudinal and

transverse directions of the bridge axis. In the D-Earthquake, accelerations on the pier

could not be recorded, unfortunately, and only the maximum value of acceleration was

obtained. It should be noted here that although the seismic response accelerations developed at the pier crest were very high, superstructures and foundations of the Itajima Bridge

suffered no structural damages through any of these four earthquakes.

Figure 8 represents amplifications of maximum accelerations between the ground surface

and the crest of the pier. It can be recognized from this result that the amplifications of

maximum accelerations are very much different between the Hyuganada Earthquake (A- and BEarthquakes) and the Bungosuido Earthquake (C- and D-Earthquakes ) , i.e., the amplification

factors are in the range of 1.1 - 1.7 in the case of A- and B-Earthquakes , whereas they are

in the range of 0.4 - 0.6 in the C- and D-Earthquakes. The frequency characteristics of the

motions were than investigated. Predominant frequencies of the records are summarized in

Table 2, based on the power spectra presented in Figure 9.

III-67
Characteristics of Ground Motions

Based on Figure 9 and Table 2, it can be understood that the ground motions induced

during the A-Earthquake have the predominant frequencies of 1 . 5 Hz and 4.5 Hz in the longitudinal motion, and 1.3 Hz and 4.0 Hz in the transverse motion. The frequencies of 1.5 Hz

and 1.3 Hz are the most predominant ones whereas the frequencies of 4.5 Hz and 4.0 Hz are the

secondary predominant ones. The ratio between the most predominant frequency f-^ and secondly

predominant frequency f2 are approximately 3.0 in both longitudinal and transverse motions.

The geological situation at the site consists of two formations with significantly different

stiffnesses, i.e. , alluvium loam silt formation and diluvium gravel formation. In such

case that subsoils with shear wave velocity Vg and thickness H are rested horizontally on the

stiff formation, the natural period T^^ of the subsoils can be estimated by Eq. 1 for the

n-th mode of shear vibration.

T„ = ' (1)

The ratio between T-^ and T2 (T-|^/T2) are always evaluated as 3.0 in such case, which well

explains the characteristics of ground motions developed during the A-Earthquake. It can be

understood therefore that the predominant frequencies observed in the ground motions would

represent the fundamental and second natural frequencies of the subsoils. Since there are

not any predominant frequencies beyond 1.3 1.5 Hz and 4.0 J" 4.5 Hz in the ground motions, it

is considered that the incident motions in the bedrock themselves have these frequencies as

their predominant frequencies.

According to the ground condition presented in Figure 2, the fundamental natural period

T]^ of the subsurface ground at the free field observatory can be estimated by Eq . 2 taking

the diluvium gravel formation as bedrock.

Ti = Z ^ = 4(^ + ^+ ^ + 1l5) = 0.53 sec (2)

i ^si ^25 145 125

Consequently the fundamental natural frequency fj^ of subsurface ground is estimated as 1.9 Hz

which is considered to be the natural frequency during low amplitude vibration. Assuming

that the difference -of fundamental natural frequency fj^ between 1.9 Hz derived by Eq . 2 and

III-68
the values shown in Table 2 depends on the strain dependence of soil modulus, the ratio of

shear wave velocity which would be developed during the A-Earthquake V^^^ and that during the

low amplitude vibration V^^ would approximately be estimated as follows.

^ = 1-3 ^ 1>5 = 0.68 >/> 0.79 (3) 1-9 so , .

Consequently the ratio of soil stiffness during the A-Earthquake G-j^ and that corresponding to

low amplitude strain level G^ can be estimated as follows. —G. V_. 2 (-31) = 0.46 0.62 (4) Go V

2 so

According to laboratory experiments concerning the strain dependence of soil stiffnesses,

such a reduction of soil stiffness is reasonable.

The explanation of ground motion characteristics for the A-Earthquake can also be applicable for the ground motions of the B- and C-Earthquakes , i.e., the ground motions of the

B-Earthquake have predominant frequencies of 1.9 Hz and 5.1 Hz in the longitudinal motion,

and 1.5 Hz and 4.9 Hz in the transverse motion, which shows that the ratios between two predominant frequencies are approximately 3.0 in both directions. The ground motions of the

C-Earthquake have predominant frequencies of 1.4 Hz, 3.7 Hz, 4.2 Hz and 4.8 ^ 5.2 Hz in both

the longitudinal and transverse motions. Although there are many apparent predominant frequencies, it would be considered that the frequencies of 1.4 Hz and 4.3 Hz represents the

fundamental and second mode frequencies, respectively, which again shows that the ratio

between two predominant frequencies are approximately 3. The frequencies of 3.7 Hz and

4.8 >^ 5.2 Hz are considered to be the predominant frequencies of the incident motions in the

bedrock.

Characteristics of Pier Motions

It is understood from the power spectra of Figure 9 that in the A-Earthquake the most

predominant frequencies of the pier motions are 1.5 Hz in the longitudinal direction and

1.3 Hz in the transverse direction, which are approximately equal to those of ground motions.

Also all of the four acceleration records during the A-Earthquake are narrow banded waves

with a single frequency predominant. It is also understood from Table 2 that in the B-Earthquake the predominant frequencies of the pier motion are approximately 1.6 Hz in the

III-69
transverse direction, which is very close to the predominant frequency of the ground motion

of 1.5 Hz. In the longitudinal direction, however, the pier motion has several predominant

frequencies. The most predominant one is approximately 1.8 Hz which is again very close to

the predominant frequency of the ground motion of 1.9 Hz. As is the case of A-Ear thquake

,

the acceleration records on the ground in both directions and the acceleration record on the

pier top in the transverse direction are narrow banded waves with a single frequency predominant. On the other hand, the records during the C~Earthquake are somehwat different

from those for the A- and B-Earthquakes described as above. The most predominant frequencies of the pier motions are approximately 1.4 Hz in both the longitudinal and transverse

directions which are different from the most predominant frequencies of the ground motions,

i.e., approximately 3.7 Hz in both directions. It should be noted here, however, that the

ground motions have a predominant frequency close to the 1.4 Hz although they are not the

most predominant.

It can be recognized from these results that the acceleration records of pier always

contain the motions in the range of 1.3 - 1.8 Hz as the most predominant ones in both the

longitudinal and transverse directions, which is considered to correspond to the lowest

natural frequency of the subsurface ground. The frequency response functions were then

computed between the pier motions and the ground motions as shown in Figure 10. It can be

recognized from Figure 10 that amplifications of the frequency response functions are rather

small for the range of frequency of 4 Hz or more because of low pass filter effects of the

foundation. The amplifications vary almost evenly between 0.5 Hz and 4 Hz although the

apparent natural frequency of the foundation is not evaluated.

It would be deduced from these considerations that the most predominant frequency that

is always contained in the motions of the pier is significantly influenced by the lowest

natural frequency of the subsurface ground so that the pier vibrates in accordance with the

motion of the subsurface ground nearby.

Nonsteady Vibrations of Pier and Ground

In order to investigate the time dependence of predominant frequencies of the pier and

ground motions, running power spectra were computed. One of the results is displayed in

Figure 11 for the A-Earthquake . Based on the running power spectra changes of predominant

frequencies of the motions in time domain were estimated as shown in Figure 12. For the

I I 1-70
A- and C-Ear thquakes , it can be observed that the frequencies are rather low at the main

parts of the motion where a large amplitude of vibration occurred. For instance, in the case

of the transverse motion of the A-Earthquake, the predominant frequency is approximately 2 Hz

around 4 seconds in time. The predominant frequency then decreased to 1.3 Hz around 12 - 18

seconds, and again recovered to 1.7 Hz around 30 seconds. It should be noted here that the

predominant frequency developed at the beginning and ending of the motions are very close to

that estimated by Eq. 2 by employing the shear wave velocity of subsurface ground at low

strain amplitude.

It should also be noted here that the time dependence of the predominant frequencies

described as above are almost identically developed in both the pier and the ground motions.

Such characteristics of the time dependence of motions suggest that the nonsteady responses

of the foundation were not derived from degradation of the soil stiffness just around the

foundation but were caused by the nonsteady responses of subsurface ground nearby.

SEISMC ANALYSIS PROCEDURE OF FOUNDATION

A discrete analytical model as shown in Figure 13 was formulated to calculate earthquake

responses of the pier foundation. The equation of motions of the system can be written as

(Mp + Me) Up + CpUp + + C^C^ " ^) + " Hg) = £ (5)

where.

mass matrix of the foundation

Me mass matrix of surrounding soils

^P

damping matrix of the foundation

^P

stiffness matrix of the foundation

Ce damping matrix expressing radiatlonal dampings

Ke stiffness matrix expressing springs between foundation and surrounding soils

^p. %

absolute displacement, velocity and acceleration vectors of foundation

absolute displacement and velocity vectors of subsurface ground

in which the subsurface ground motions of and jj are assumed to be specified. Denoting as

III-71
M = Mp + Mg

C = Cp + : ' (6)

K = Kp +

Eq. 5 can be written as . ..

M iip + C ^ + K Up = Ug + Ug (7)-

The vector u can be conveniently decomposed into a quasi-static displacement vector u_,„ —p —ytt and a dynamic displacement vector JJp^, i.e., . -

= iipd + i^ps (8)

By definition of quasi-static displacement in the form as , , i

K £ps - Ke£g = 0 (9)

u can be written as

^ps = :^~^^g = i^g (10)

Substitution of Eqs. 8 and 9 into Eq. 7 gives

Miipd + C Upd + K Upd = M K^Ug + (C^ - C K3) Ug (11)

Usually the damping term on the right hand side of Eq. 8 is less significant when compared

with the inertia terms so that it can be dropped from the equation without introducing significant errors. Then Eq. 11 can be written as

Miipd + C Upj + K Up^ = M K^g (12)

Eq. 12 can be solved by modal-superposition procedures provided that the damping matrix

on the left hand side of the equation is assumed to be triangularized in the same manner as

the mass and stiffness matrices in the form of the critical damping ratio.

III-72
,

, . CALCULATION OF PIER MOTIONS

The pier motions were computed by using the analytical procedure described in the preceeding paragraph based on the measured ground motions for the A- , B- , and C-Earthquakes,

and they were compared with the measured motions.

The bedrock motions were computed from the measured ground surface motions by the

de-convolution procedure taking account of the strain dependence of the shear moduli and

hysteretic damping ratio of the subsoils as shown in Figure 14. The subsoils and foundation

were idealized by a one-dimensional shear column model with equivalent linear soil properties

and one-dimensional elastic beam supported elastically by the surrounding subsoils, respectively. The weight of a girder supported by the pier was idealized as an additional mass

lumped at the crest of pier. The lowest natural frequencies of the pier and surrounding

subsoils thus estimated are shown in Table 3.

The response accelerations of pier were then calculated based on Eq. 12 by applying the

bedrock motions at the bottom of the shear column model of subsoils. The comparative plots

of both the theoretical and measured accelerations at the crest of pier are shown in Figures

15, 16 and 17 for the A- , B- , and C-Earthquakes, respectively. The damping ratios assumed

in the analyses are shown in Table 3. It is recognized from the results that fairly good

agreements are obtained for the motions in the A- and B-Earthquakes. However, the correlation for the motion in the C-Earthquake is appreciably less satisfactory and further precise

investigations are needed to clarify the frequency characteristics of the foundation.

CONCLUSIONS

Based on the results presented, the following conclusions may be deduced:

1) Seismic responses of the deeply embedded foundation are significantly influenced by

the effects of surrounding subsurface soils. The most predominant frequencies which are

always contained in the motions of the pier crest are prescribed by the lowest natural frequency of the subsurface ground so that the pier foundation responses in accordance with the

motion of the subsurface ground nearby.

2) The nonsteady responses developed on the pier foundation are almost idenitical with

those observed on the ground surface, which suggest that the nonsteady responses of the pier

III-73
foundation are not derived from degradation of the soil stiffness just around the foundation

but are primarily caused by nonsteady responses of the subsurface ground nearby.

3) Seismic response accelerations of the foundation can be calculated with fairly good

accuracy by the analytical procedure presented herein from the free-field ground accelerations measured near the foundation for earthquakes which induce ground accelerations at the

bridge site with the most predominant frequencies lower than the fundamental natural

frequency of the foundation.

REFERENCES

[1] Iwasaki, T. , "Earthquake-Resistant Design of Bridges in Japan," Bulletin of Public

Works Research Institute, Volume 29, Public Works Research Institute, Ministry of

Construction, May 1973.

[2] Kuribayashi, E. and Iwasaki, T. , "Dynamic Properties of Highway Bridges," Fifth World

Conference on Earthquake Engineering, Roma, December 1972.

[3] Iwasaki, T. , Wakabayashi, S., Kawashima, K. and Takagi, Y. , "Strong-Motion Acceleration Records from Public Works in Japan (No. 3)," Technical Note of the Public Works

Research Institute, No. 34, Public Works Research Institute, Ministry of Construction,

1978.

[4] Yamahara, H. , "Ground Motions During Earthquakes and the Input Loss of Earthquake

Power to an Excitation of Buildings," Soils and Foundations, Vol. 10, No. 2, 1970.

[5] Duke, C. M., et al. , "Strong Motions and Site Conditions — Hollywood," Bulletin of

the Seismological Society of America, Vol. 60, No. 4, 1970.

[6] Matsushima, Y., "Structure-Soil Interaction for Finite Element Hysteretically Damped

Systems, Journal of Architectural Institute of Japan, No. 199, 1972.

[7] Iwasaki, T. and Kawashima, K., "Seismic Analysis of Highway Bridge Utilizing StrongMotion Acceleration Records," Proceedings of Second International Conference on Microzonation, San Francisco, California, USA, 1978.

'IT

I I 1-7

4
Table 1 Strong Motion Acceleration Records at The Itajlma Bridge

Earthquake

No. Earthquake Date Rlchter

Magnitude

Eplcentral

Distance [km]

Maximum Acceleration ( Cal ) Pier Motion Ground Surface Motion

Longitudinal Transeverse Longitudinal Transverse

1 The Hyuganada Earthquake 1968.4.1.09:42 7.5 101 209 306 169 186

2 The Hyuganada Earthquake

(Aftershock) 1968.4.1.16:13 6.3 99 38 66 34 42

3 The Bungosuldo Earthquake 1968.8.6.01:17 6.6 11 199 231 441 353

4 The Bungosuido Earthquake

(Af tershock) 1968.8.6.13:21 5.3 19 100 63 227 172

Table 2 Predominant Frequencies of Strong Motion Acceleration Records Measured on Pier Crest and Ground Nearby [Hz]

Earthquake

No. Earthquake Longitudinal Transverse

Ground Pier Crest Ground Pier Crest

1 The Hyuganada Earthquake 1.5 1.5 1.3 1.3

2 The Hyuganada Earthquake

(After shock)

1.9 1.8,2.0,2.3,3.7 1.5 1.6

3 The Bungosuido Earthquake 3.7 1.4 3.7,4.4 1.4

4 The Bungosuido Earthquake

(Aftershock) 4.2 4.4

Table 3 Lowest Natural Frequencies and Damping Ratio Assumed

In Seismic Analyses

Earthquake

Surrounding

Ground Caisson-Pier

Foundation Hystretic Damping

Ratio of Subsoils(%)

Radiational

Damping Ratio(%)

1st 2nd 1st 2nd The Hyganada Earthquake

(A-Earthquake) 1.6 4.9 2.

a

7.8 10 20

The Hyuganada Earthquake

(After Shock) (B-Ear thquake) 1.9 5.7 3.1 8.3 5 20

The Bungosuido Earthquake

(C-Earthquake)

1.6 4.7 2.7 7.7 8-12 20

III-75
BRIDGE PER SUOEO

Fig. 1 General View of The Itajima Bridge

Fig. 2 Soil Profile at the Pier Site and Ground Observatory Site

III-76
iTajima Accelerometer

Malt : Mesozoic Cretoceous

Fig. 3 Geological Condition Arround The Itajima Bridge

Longitudinal

Transverse

Its I to

III t

Ji 1 .0

•t o 20.0 2t 0

(a) AT CREST OF PIER

Longitudinal

Transverse

*' °

r

(b) ON GROUND SURFACE NEARBY

Fig. 4 Acceleration Records At the Itajima

Bridge During the Hyuganada Earthquake

of April 1, 1968 (A-Earthquake)

111-77
Longitudinal

Fig. 5

Acceleration Records at the Itajima

Bridge During An Aftershock of the

Hyuganada Earthquake of April 1, 1968

(B-Earthquake)

lOI.)

J» 0

^ 1» s 0.0

s 39 0 to 'CO Transverse

(a) AT CREST OF PIER

Longitudinal

'S 0 10.

0

•0 0 €JCC>

(CILI

0 1>.0 0.0

-IJ.O 14 0

Transveise

(b) ON GROUND SURFACE NEARBY

Longitudinal

M.» M.fl Transverse

** 9 M.9

(a) AT CREST OF PIER

Longitudinal

n fi M.f n-* <Hci Transverse

<o 0 n 9 M IMO

(b) ON GROUND SURFACE NEARBY III-78

Fig. 6

Acceleration Records at the

Itajima Bridge During the

Bungosuido Earthquake of

August 6, 1968 (C -Earthquake)
Transverse

Fig.

so 10 0

(a) ON GROUND SURFACE NEARBY

7 Acceleration Records At The Itajima Bridge During

An Aftershock of the Bungosuido Earthquake of

August 6, 1968 (D-Earthquake)

q:

LlJ a.

o

o

<

<

500

300

200

1 00

70

50

30

20

1 0

EARTHQUAKE No.

1 ) 1968.4.1 .09:42

2 ) 1968.4.1 . 16 : 13

3 ) 1968.8.6.01 : 17

4 ) 1968.8.6.13: 21

"1

\ Mil I I r/

A /

/

/

A y'' A

Q LONGFTUDINAL

DIRECTION

/\ TRANSVERSE

DIRECTION

J \ \ I I I I I I J I L

10 20 30 50 70 100 200 300 500

MAX. ACCELERATION ON GROUND SURFACE NEARBY [GAL]

Fig. 8 Amplifications of Maximum Acceleration Between

The Pier Crest and Ground Surface Nearby

III-79
Fig. 9 Power Spectra of Acceleration Records at the Itajima Bridge

j Otfcction

j

u ? 5 4 5 6 Ftequency ( M, )

Gffxrd SltIcu

Pier

Frequency (Hz)

(a ) Longirucinol Motion

; |Dlr«cilQn

|

j I Gfoufid Suffoc

aJ

1/1

2 3 1 5 6 7 8

FrBquenc/ 1 Hi ] loi Longirudinai Mofion

(1) The Hyuti^nada Earttiquake (A-EarCliquake) An Aftershock of The Hyuganada Earthquake

(B-Earthquake)

Oreciion

- -OortSrfoce —Pter

II 2 3 4 5 e 7 8

FrequBOcy {Hi)

(bl TronsvfliSe Mrtian

(3) The Buiigosuldo Earthquake

(C-Earthquake)

12 3 4 5 6 7 Ffequancy C Hz ] lb I TfQoivflfsB Motion

(A) An Aftershock of The Buugosuldo Earthquake

(D-Earchquake)

III-80
4 6

Frequency [ Hz ]

(a) The Hyuganada Earthquake

(A-Earthquake)

10

4 6

Frequency [ Hz ]

(b) An Aftershock of The Hyuganada Earthquake

(B-Earthquake)

Tronsverse

Longitudinal

\tlA\ ^ r''' 'i/'i '"W\

11

1 1 1 1 L. 1 2 4 6 8

Frequency [ Hz ]

(c) The Bungosuido Earthquake

(C-Earthquake)

10

Fig. 10 Frequency Response Functions Determined by Measured Strong Motions

Between Pier Crest and Ground Surface

III-81
0 5 10 15 0 5 io I Frequency CHzJ Frequency CHzJ £

( a ) Longifudinal Motion (b ) Tronsverse Motion t'

0 5 10 15 0 5 10 15 Frequency C Hz j Frequency r Hz 3 (o) Longitudinal Motion (b) Transverse Motion (2) GROUND SURFACE

Fig. 11 Running Power Spectra of Strong Motions During

The Hyuganada Earthquake (A-Ear thquake)

III-82
N 4 X

t. 3 o

0

— Pier Cresf —Ground Surface -.4

1 1 1

4

3

2 k

1

10 20

Time [ Second ] (a) Longitudinal Direction

30

1

\ Pier Cresf Ground Surface

10 20

Time C Second 3 (b) Transverse Direction

30

(t) THE HYUGANADA EARTHQUAKE ( A-Ea rthqua ke

)

Pier Cresf Ground Surface

10 Time [ Second]

(a ) Longitudinal Direction

20

5 3 M 2

O)o- 1 0) 0

- Pier Crest

- Ground Surface III.

10 Time [ Second ]

( b ) Transverse Direction

20

(2) AN AFTERSHOCK OF THE HYUGANADA EARTHQUAKE ( B-Ea rthqua ke

)

10Time C Second ] (a) Longitudinal Direction

10 20

Time C Second ]

( b ) Transverse Direction

(3) THE BUNGOSuiDO EARTHQUAKE ( C-Earthqudke)

Fig. 12 Time Dependence of Predominant Frequencies of

Pier and Ground Motions

III-83
»• Gravels with Silts

Silty Fine Sand

Sandy Clay

0..°; o

O Gravels with Clay

' O Gravels with Cobble

(a) Caisson-Pier and Surrounding Subsoils (b| Analytical P^el

Fig. 13 Analytical Model of Subsoils and Foundation

SHEARE STRAIN V

Fig. 14 Sheare Strain Dependence of Shear Modulus and

Hysteretic Damping Ratio Assumed In Seismic

Analyses

III-84
COMPUTED MEASURED

10 12 14 16 TIME (SECONOI M 2? 24 26 28

(a) Longitudinal Motion

i „

1

•

s

jL i . A 1

- MEtSUKD

• f r V 1

) Z 4 ( 1 n 12 M TIME —* r1 SECONOI K

(b) Transverse Motion

Fig. 15 Correlation of Seismic Response Acceleration At the Pier Crest

(The Hyuganada Earthquake; A-Earthquake)

III-85
80

AO

0

-40 -

ID 12 TIME (SECOND)

(a) Longitudinal Motion

2046

COMPUTED

MEASURED

14 16 18 TIME ISECONDI

(b) Transverse Motion

20

Fig. 16 Correlation of Seismic Response Acceleration At The Pier Crest

(An Aftershock of The Hyuganada Earthquake; B-Ear thquake)

i'

— COMPUTED — MEASURED

0 2 4 6 10 12 14 16 TIME ISECONDI

20 22 24 26 26 30

Fig. 17 Correlation of Seismic Response Acceleration At The Pier Crest

(The Bungosuido Earthquake; C-Ear thquake)

III-86
FIELD AND LABORATORY DETERMINATION OF SOIL MODULI

William F. Marcuson III

Joseph R. Curro, Jr.

U.S. Amy Engineer Waterways Experiment Station

Vicksburg, Mississippi

. ;

. . • ABSTRACT

A field geophysical Investigation was performed to determine the shear and Young's

moduli as a function of depth for a site in southern Ohio. This investigation included

crosshole, downhole, and surface refraction investigation techniques. A supplementary set

of laboratory resonant column tests was performed with the Drnevich resonant column device.

Laboratory undisturbed specimens were excited in both the longitudinal and torsional modes

to obtain both Young's and shear moduli as a function of strain. Both laboratory and field

data are presented, compared, and discussed.

An idealized soil profile for boring 821-UD is presented. This soil profile includes

recommended design values of moduli and damping. At this boring location, bedrock lies

approximately 36 ft (11 m) below the ground surface. The 36 ft (11 m) thick soil deposit

has been subdivided into four layers whose shear moduli range from 3.4 x 10-^ to 86 x 10^ psi

(23.4 X 10^ to 592 x 10^ kPa). The moduli increase as a function of depth. The Young's

moduli for this soil profile range from 10 x 10^ to 225 x 10^ psi (59 x 10^ to 1551 x 10^

kPa) and also increase with depth. The internal damping of the soil deposit was found to

range from 3 to 5 percent for low dynamic strain amplitudes and is constant with depth. The

bedrock has a shear modulus of 125 x 10-^ (861 x 10-^ kPa) and a Young's modulus of 370 x 10-^

psi (2551 X 10^ kPa).

KEYWORDS: Field testing; laboratory testing; dynamic properties; shear modulus; damping;

wave velocities; geophysical; resonant column test.

III-87
INTRODUCTION

In order to evaluate the dynamic response of soil or a structure founded on soil,

knowledge of stress-strain properties of the foundation material is required. When using

current computer analysis techniques, the initial elastic moduli (very low strains, generally

10~^ cm/cm or less) are required along with the variation in moduli as a function of strain.

These initial moduli (shear and Young's) can be determined in situ using various

geophysical techniques. Or, resonant column tests can be conducted in the laboratory on

undisturbed samples to obtain moduli as functions of strain level.

Since the moduli discussed above would be needed for dynamic foundation design of four

proposed buildings and a test loop structure [13], a test program was formulated that consisted of field and laboratory investigations to provide the required moduli. In addition,

Poisson's ratios were determined from the investigations to aid in the dynamic design.

Therefore, the primary purpose of this project was to determine values of shear and

Young's moduli for the foundation materials underlying the five proposed structures (designated as X-340-1 through X-340-4 and X-774), located as shown in Figure 1. Additionally,

building X-340-4 was arbitrarily selected as an example case to illustrate the procedure

used for determining dynamic design values.

The purpose of this paper is to compare dynamic soil properties determined in situ and

in the laboratory and to discuss the rationale used in obtaining design values. ^

.

j

, ;.

SITE DESCRIPTION '

' b

The construction site is located in the vicinity of Portsmouth, Ohio, near the confluence of the Ohio and Scioto Rivers. Overburden soils at the site were primarily Pleistocene

river- and lake-bed deposits. Figure 1 is a geologic map of the site [13]. The map outlines

the aerial distribution of the glacial lake-bed deposits and early alluvial deposits. The

lake-bed deposits, consisting of fat, thinly laminated clays, are largely confined to the old

Scioto River channel.

The area outlined as lake deposits in Figure 1 shows the occurrence of fat, laminated

clay either at the surface or at some depth beneath the surface. Lean clays, silts (alluvium),

and a fill material which was placed during the period 1953 to 1955, lie above the fat clay

III-88
in various thicknesses over a large portion of the outlined area. Early river alluvial

silty clay and silt occupy that portion of the valley fill outside the fat clay boundaries.

Up to 60 ft (18 m) (averaging 35 to 40 ft (10.5 to 12.0 m)) of early river-and lake-bed

deposits in the old valley still remain at the site. These deposits are mostly fat, laminated clays, lean clays, and fine silts, with a 2 to 10 ft (0.6 to 3 m) thick layer of

weathered clayey gravel at the base. A 1 to 10 ft (0.3 to 3 m) thick layer of residual

soil and colluvium mantles a large part of the valley. Rock, consisting of shale and sandstone, underlies the soil [13].

. . FIELD INVESTIGATION

The field investigation was comprised of two phases: (a) drilling and sampling and

(b) geophysical investigation.

o Drilling and Sampling

During this phase of the field investigation, 27 holes were drilled at the site for in

situ seismic test use and to obtain undisturbed samples for laboratory testing. The holes

were drilled in sets of three with one set near each end of the four proposed building locations and one set at the test loop location. The three holes were oriented to form an L-type

configuration with the distance from the boring at the vertex of the L to the boring at the

end of each leg being approximately 20 ft (6 m) . Undisturbed samples were obtained from the

boring at the vertex of the L for each set with a 5 in. (12.7 cm) Hvorslev fixed piston

sampler. The remaining holes were drilled with a Waterways Experiment Station (WES) modified

fishtail bit with upward baffles. After the holes had been drilled to the desired depth (2 to

5 ft (0.6 to 2 m) into rock), 3-in. (7.6 cm) ID PVC (polyvinyl chloride) pipe was installed

in the boreholes, and the annular space between the casing and walls of the borings was

grouted with a mixture of cement, bentonite, and water that, after setting up, had the

approximate consistency of soil.

111-89
Geophysical Investigation

The geophysical investigation conducted at the site consisted of nine series of in situ

seismic tests. Each series was comprised of crosshole, downhole, surface refraction seismic,

and surface vibratory tests.

Crosshole Test Procedure ' ' ' '

The crosshole tests discussed herein were conducted using a portable 24-channel

refraction seismograph. A recorder displayed the data while operating at a paper speed of

approximately 35 ips (89 cm/sec). Timing lines were displayed on the oscillogram at 10 sec

intervals. The test procedure to obtain shear wave (S-wave) data was a modified technique

that used the same vibratory source and control package as that described by Ballard [1].

The primary difference in the two procedures was concerned with the data acquisition package,

i.e., no signal enhancement was used. Rather, the vibratory source was swept through a frequency range while monitoring the output of the two geophones placed at the same elevation in

the receiver boreholes (the vibrator was producing vertical oscillations, thereby transmitting a vertically polarized shear wave). When an acceptable response was received at a specific frequency, this frequency was then interrupted by a tone burst generator to send a

specific number of cycles of that frequency to the receiver units. The source geophone was

displayed simultaneously along with both receiver geophones on the oscillograph. In this

manner, the origin of the source pulse could easily be recognized, and the time difference

that pulse and the signal arrival at each of the two receivers could be determined.

The tests to obtain S-wave data were conducted by first placing the source and receivers

at the same elevation near the top of the boreholes, then pulsing the source unit several

times recording both transmitted and received signals. After a satisfactory record had been

obtained, the units were then repositioned 5 ft (1.5 m) deeper. The procedure above was

repeated at this and each succeeding 5 ft (1.5 m) depth until reaching the bottom of the

boring. Since these holes were relatively shallow, their verticality could be checked by

visual observation using either flashlight or mirror sources. In all cases, the boreholes

were vertical within +_ 3 in. (+ 7.6 cm), thereby eliminating the need for a formal deviation

survey

.

III-90
The compression wave (P-wave) data were obtained in a similar manner, except that the

vibrator and its associated instrumentation were not used. Instead, exploding bridgewlre

(EBW's) were used at each 5-ft (1.5-m) increment as the P-wave source.

Data obtained from the crosshole tests were the times required for P- and S-waves to

propagate from the source to a point of detection. These times were then divided into the

distance between source and receiver geophones to provide apparent velocities. If a nearby

higher velocity layer exists, the wave will refract and travel along that layer, thus traveling along a faster path than the direct distance path. A computer program [2] for crosshole

seismic interpretation, based on Snell's law of refraction, is used to determine true velocities by accounting for zones of high velocity contrast.

Downhole Test Procedure

This test, designed to provide data for determination of vertically oriented P- and

S-wave velocities, was conducted using the same seismograph and recorder as those used for

the crosshole tests.

The receiver geophone was placed at a depth of 5 ft (1.5 m) in a borehole, and a

vertical hammer blow to a steel plate positioned approximately 1 ft (0.3 m) from the mouth

of the boring on the ground surface was used as the P-wave source. A geophone adjacent to

the hammer Impact point provided zero time. S-wave tests were then conducted by placing a

large wooden plank on the ground surface and striking the plank at each end thereby reversing

polarity of the horizontally polarized S-wave which facilitates identification. The procedures for obtaining P- and S-wave data were then repeated with the receiver geophone at 5 ft

(1.5 m) deeper Increments until the bottom of the borehole was reached.

Surface Refraction Seismic Test Procedures

Conventional surface refraction seismic tests were conducted at the site using the

24-channel unit. Refraction lines run at the site were 240 ft (73 m) in length. Geophones

were spaced at 10 ft (3 m) intervals with the first geophone positioned 5 ft (1.5 m) from an

explosive source. Both forward and reverse transverses were made at the site so that true

velocities, in addition to apparent velocities, could be determined and depths to refracting

Interfaces computed.

III-91
Surface Vibratory Test Procedure

In order to provide a necessary input parameter (near-surface S-wave velocity) to the

crosshole computer program, surface vibratory tests [5] were conducted near each crosshole

set. Rayleigh (R) waves may be generated using a controlled energy source such as a vibrator. In this case, the electromagnetic vibrator used to conduct the crosshole tests was also

used as the R-wave source. The vibrator was operated at several frequencies, and the R-wave

was monitored by the geophones used for the surface refraction seismic tests, which were

placed in a line at intervals of 1 ft (0.3 m). In this manner, the velocity of the R-wave

can be determined from a plot of R-wave phase arrival time versus distance for each specific

frequency. Then, a corresponding wavelength can be calculated by dividing the frequency into

the velocity. Wave velocities thus derived are considered to be average values for an effective depth of one half the wavelength [10]. S-waves, for practical purposes, can be

considered to have the same velocity as R-waves.

Since the vibrator used at this site was of limited force output (50 lb (222 N)), the

depth of investigation was limited to approximately 5 ft (1.5 m). These data could then be

compared with those obtained for the shallow crosshole test position.

Elastic Moduli and Poisson's Ratio Determination

The P- and S-wave velocities determined from the above-mentioned test procedures should

be converted to elastic parameters such as shear modulus G, Young's modulus E, and Poisson's

ration v for use in foundation design. This conversion can be accomplished using theory of

wave propagation in elastic media [8] and the total mass density P of the medium.

FIELD TEST RESULTS

The results determined from the in situ seismic tests conducted at the proposed location

of building X-340-4 are presented and discussed below. It appeared feasible to the authors

to select one location as being "typical" and to discuss the test results obtained there.

Proposed Building X-340-4 Location

Two series of in situ seismic tests were conducted at this location, one near each end

of the proposed building location as shown in Figure 2.

III-92
Crosshole Tests

These tests consisted of two crosshole sets. Borings 821, 822, and 823 comprised one

set near the south end of the proposed building, whereas the second set, near the north end

of the building, was made up of borings 824, 825, and 826 as shown in Figure 2.

Borings 821 and 824 were used as the seismic source boreholes for their respective sets,

and borings 822, 823, 825, and 826 were employed as receiver holes for their respective sets.

The seismic source and receiver (geophone) placements used in the performance of the crosshole test are shown in Figure 3. True P- and S-wave velocities determined from the crosshole

data are presented alongside the receiver and source locations. Note that only one P- and

S-wave velocity is shown for each test elevation although true P- and S-wave velocities were

determined from the seismic source borehole to each of the receiver boreholes for each crosshole set. In comparing the velocities from the source to each receiver, the spread was so

narrow (< 10%) that it was thought advantageous to average the two P-wave velocities and

likewise, the two S-wave velocities. .

Downhole Tests

Two downhole tests located as shown in Figure 2 were conducted, one using boring 821 and

the other using boring 824. The results of the downhole tests conducted in borings 821 and

824 are presented in Figures 4 and 5, respectively. As shown, average and incremental P- and

S-wave velocities were determined from the downhole data.

Surface Refraction Seismic Tests

These tests consisted of four traverses (two lines), the orientation and location of

which are shown in the test layout. Figure 2. The time versus distance plots for these traverses, designated as S-15 through S-18, are shown in Figures 6 and 7. These figures also

present apparent velocities, true velocities, and associated depths to refracting layers.

Surface Vibratory Tests

One surface vibratory test was conducted at boring 821 and another at boring 824. The

results of these tests yielded average R-wave velocities of 375 and 235 fps (114 and 72 mps)

III-93
for the near-surface material (< 5 ft (< 1.5 m) depth) at borings 821 and 824, respectively.

These velocities were used as true S-wave velocity inputs for the near surface to the crosshole computer program [2].

Data Interpretation

P- and S-Wave Velocity Zoning , ^ , , ,

The P- and S-wave velocity results determined from the crosshole, downhole, surface

refraction seismic, and surface vibratory tests at this building location were analyzed and

interpreted to produce the P- and S-wave velocity profiles shown in Figures 8 and 9, respectively. The P- and S-wave zones will not necessarily align with one another. This is primarily due to a large Increase in P-wave velocity as percent saturation of the soil reaches

99-100 percent, whereas S-wave velocity is not affected.

Referring to Figure 8, five velocity zones were delineated from the P-wave data. The

velocities presented for the zones were usually weighted averages based on data quality. A

1400 fps (425 mps) velocity was established for the near-surface zone, which averaged about

7 ft (2.1 m) in thickness. The second velocity zone averaged 4900 fps (1500 mps) and was

about 20 ft (6 m) thick. This zone was underlain by a lower velocity layer (2800 fps (850

mps)), which ranged from 5 to 13 ft (1.5 to 4 m) in thickness and is best delineated by the

uphole seismic data. The next two velocity zones (7100 and 9400 fps (2165 and 2865 mps))

could be collapsed into one. However, since the top of the 9400 fps (2865 mps) zone coincides with the top of shale, it was believed advantageous for analysis purposes to define top

of rock and to keep the zones separate. The 7100 fps (2165 mps) zone Interpreted from the

crosshole and downhole tests in borings 824 through 826 is probably indicative of a silt,

sand, and gravel mixture cemented to form the conglomerate noted in the laboratory classification of samples from boring 824 in this depth range. - •

Referring to Figure 9, five S-wave velocity zones were interpreted from the S-wave data.

A 350 fps (105 raps) velocity was established for the near-surface zone and averaged about

7 ft (2.1 m) in thickness. The underlying zone had a 500 fps (150 mps) average and a thickness of about 15 ft (4.6 m) . The third zone averaged 700 fps (215 raps) and was approximately

12 ft (3.7 m) thick. The two deeper zones (1750 and 2000 fps (535 and 610 mps)) could probably have been combined; however, for the same reason given for the P-wave zoning, they were

111-94
not. It will be noted that the value of wet unit weight applicable for each zone is also

presented in Figure 9. These values were obtained from the undisturbed samples used in the

resonant column tests and from laboratory tests conducted on samples from borings near the

in situ seismic test locations [13].

Shear and Young's Moduli and Poisson's Ratio (G, E, v)

The interpretation of the data was continued by establishing zones from the velocity

profiles and determining values of (G), (E), and (v) for these zones. This was accomplished

by overlaying the P- and S-wave velocity profiles (Figures 8 and 9). This overlay produced

seven zones. Then, values of G, E, and v were computed for each zone using the P- and S-wave

velocities and wet unit weights that fell into each zone. After analyzing G, E, and v values

for the various zones, it was evident that the controlling factor was the S-wave velocity

zoning because the P-wave velocity variations had little effect on E and v values and of

course, no effect on G values. Therefore, the seven newly established zones could be reduced

to the five original zones shown on the S-wave profile (Figure 9). The final field interpretation for this building location showing the five zones with their associated G, E, and v

values is presented in Figure 10. -.^

LABORATORY INVESTIGATION

Undisturbed samples taken in the field were transported by truck to the WES. Once in

the WES laboratory, the samples were opened and classified. The samples were evaluated to see

which were representative of field conditions according to the boring logs and general knowledge of the site [13] and which samples could be tested in the laboratory based on lack of

evidence of disturbance. Resonant column tests were performed on the representative samples

found to be suitable for testing, i.e. those which did not contain particles of gravel or

roots .

, ' ,'i J

Resonant Column Test Equipment

The equipment used in this series of resonant column tests is known as the Drnevich '

torsional resonant column apparatus. Resonant column apparatus normally excite the specimen

I 11-95
only torsionally; however, this equipment differs in that it also includes an electromagnetic

oscillator for longitudinal vibration of the specimen. A more detailed description of this

equipment is given elsewhere [4]. , :

' t.

'JIM J -a -.-v.. Test Procedures

All tests were performed on a 2.8 in. (7.1 cm) diam by 7 in. (17.8 cm) long undisturbed specimens. These specimens were isotropically consolidated so as to correspond to the

in situ mean principal effective stress or octahedral stress, Oocf situ mean principal effective stress was computed assuming normally consolidate conditions [13] and a coefficient of earth pressure at rest of 0.5.

After primary consolidation was complete, the drainage valve was closed and the B valve

was checked and recorded [3]. The specimen was subjected to sinusoidal excitation in both

the longitudinal and torsional modes. Longitudinal and torsional vibrations were not conducted simultaneously. The frequency of vibration was varied until resonance was obtained

and data were recorded such that the modulus and strain amplitude values could be computed.

In order to obtain the variation of modulus as a function of strain amplitude, the driving

force was increased and the general test procedure was repeated. '''•Z " '

The procedure above is a brief description of that used by the WES. A more detailed

testing procedure, along with a complete discussion of the analysis and calculations required

to obtain shear and Young's moduli from the raw data, is given elsewhere [4]. ' '- ujL

Resonant Column Test Results

Undisturbed samples from various depths in borings 800 UD, 803 UD, 806 UD, 809 UD,

812 UD, 815 UD, 818 UD, and 821 UD were tested in the resonant column device [3]. A total

of 37 specimens were tested. ...

Reference 3 contains plots of modulus and damping versus dynamic strain amplitudes for

all the resonant column tests. In order to familiarize the reader with the format of the

resonant column data and its interpretation, the test results obtained from boring 821 UD

samples will be discussed. The location of boring 821 UD is shown in Figure 2.

Sample 2 was taken at a depth of 2.7 to 4.2 ft (0.8 to 1.3 m) and was considered to

represent the surficial layer to a depth of 14 ft (4.3 m) . This material was classified

in the laboratory as a CH clay and tested in the resonant column device at a mean effective

III-96
principle stress of 1.7 psi (11.7 kPa) . Figure 11 presents the results in terms of dynamic

modulus versus dynamic O-peak strain amplitude for the resonant column tests conducted on

this specimen.

Figure 12 presents the damping ratio as a function of dynamic strain amplitude for the

resonant column tests conducted on sample 2. The damping ratio varies from approximately 5

to 10 percent over the strain range of 10~^ to 10"^ in. /in. (rad/rad) (cm/cm). In general,

the damping determined in the torsional mode is approximately two thirds of that determined

in the longitudinal mode. As expected, damping generally increases with increasing strain

[6,7,9,12].

Sample 9 was taken at a depth of 20 to 21.8 ft (6.1 to 6.6 m) and was considered to

represent a clay layer from a depth of 14 to 22 ft (4.3 to 6.7 m). A resonant column test

was performed on the specimen taken from sample 9 which was consolidated to a mean effective

principle stress of 6 psi (41.4 kPa) . The results of this test are presented in Figures 13

and 14

.

Sample 11 was taken at a depth of 24.5 to 26.6 ft (7.5 to 8.1 m) and was considered to

be typical of the 5 ft (1.5 m) thick layer between a depth of 22 and 27 ft (6.7 and 8.2 m)

.

This material was a silt and was tested in the resonant column apparatus at a mean effective

principle stress of 9.5 psi (65.5 kPa). Shear and Young's moduli and damping ratios as a

function of dynamic strain amplitudes are presented in Figures 15 and 16.

Sample 13 taken at a depth of 29.2 to 30.5 ft (8.9 to 9.3 m) was considered to be

representative of the soil layer between a depth of 27 and 31 ft (8.3 and 9.4 m) . This

sample was classified a CL clay and tested in the resonant column device under a mean principal effective stress of 10.75 psi (74 kPa) . The dynamic modulus and damping ratio as a

function of dynamic strain amplitude are presented in Figures 17 and 18, respectively.

Below a depth of 31 ft (9.4 m) the soil is extremely gravelly. Samples obtained were

disturbed to such an extent as to be considered unsatisfactory for testing. Rock was found

at a depth of about 36 ft (11 m).

DISCUSSION OF LABORATORY TEST RESULTS

The resonant column data obtained from specimens trimmed from samples taken in boring

821 UD have been presented. A soil profile will be developed for this boring in terms of

shear modulus. Young's modulus, and internal damping ratio.

III-97
Shear Modulus

Figure 19 summarizes the shear modulus data obtained from testing four specimens taken

from samples of boring 821 UD. Also shown in Figure 19 is a soil profile depicting the four

layers that were observed in the boring. Note that the near-surface clay layer is 14 ft

(4.3 m) thick, and has a shear modulus at low strain levels of about 3.5 x 10-^ psi (24 x 10-^

kPa) (10^ kPa = 1 MPa) . As the shear strain is increased to a dynamic strain level of about

10~3 rad/rad, the shear modulus decreases to about 1.7 x 10^ psi (11.7 MPa). Immediately

below this 14 ft (4.3 m) thick surficial layer is an 8 ft (2.4 m) thick layer of clay.

From the resonant column test results, it appears that this material has a shear modulus of

about 2.8 x 10^ psi (19.3 MPa) at very low strains. This modulus decreases to about 1 .0 x

10^ (6.9 MPa) as the strain level increases to 10"-^ rad/rad. Below a depth of 22 ft (6.7 m)

there exists a 4 ft (1.2 m) thick silt layer. This material exhibits a shear modulus of

17 X 10-^ psi (117 MPa) at low strain levels. As the strain increases to a dynamic strain

amplitude of about 10"^ rad/rad, dynamic shear modulus decreases to a value of 5.6 x 10^ psi

(38.6 MPa). Below the 4 ft (1.2 m) thick silt layer there exists a 5 ft (1.5 m) thick

clay layer between a depth of 26 and 31 ft (7.9 and 9.4 m) . This layer exhibits a shear

modulus of about 5.6 x 10-^ psi (38.6 MPa) at low strain levels. This modulus decreases to

about 2.8 X 10^ psi (19.3 MPa) as the strain level becomes greater than 10"^ rad/rad.

In reviewing Figure 19 several specific items should be noted. These are:

a) As the strain level increases from 10"^ to lO"-^ rad/rad, the shear modulus is

generally reduced by about 50 percent, except in the silt layer at a depth of 22 to 26 ft

(6.7 to 7.9 m). In this layer the shear modulus is reduced to about one third of its

original low strain value.

b) One would expect that the modulus would be a function of the mean principal

effective stress. Thus, the modulus should increase as a function of depth, other things

being equal. This is not the case in boring 821 UD. The clay layer at a depth of 14 to

22 ft (4.3 to 6.7 m) is not stiffer than the surficial clay.

c) The 5 ft (f.5 m) thick clay layer immediately below the silt has a modulus that

is significantly lower than the silt layer at a depth of 22 to 26 ft (6.7 to 7.9 m) . This

is not unusual but should be noted because it may be important in any future dynamic analyses

or calculations.

III-98
Young's Modulus

Figure 20 shows Young's modulus (E) versus dynamic strain amplitude as a function of

depth in boring 821 UD. The 14 ft (4.3 m) thick, surficial layer has a Young's modulus at

low strain levels of about 26 x 10^ psi (179 MPa). This value decreases to approximately

15 X 10^ psi (103 MPa) when the strain level increases to 5 x 10"^ in. /in. (cm/cm). The clay

layer between a depth of 14 and 22 ft (4.3 and 6.7 m) is very similar to the surficial layer.

This 8 ft (2.4 m) thick clay layer has a Young's modulus of about 26 x 10-^ psi (179 MPa) at

a strain level of 4 x 10"^ in. /in. (cm/cm). As the strain increases to a level of approximately 10"^ in. /in. (cm/ cm), the Young's modulus decreases to 9.7 x 10^ psi (66.9 MPa). The

Young's modulus in the silt layer that lies at a depth of 22 to 26 ft (6.7 to 7.9 m) is

approximately 56 x lO-^ psi (386 MPa) at strain levels below 10"^ in. /in. (cm/cm). As the

strain level increases above 10~^ in. /in. (cm/cm), this Young's modulus decreases to approximately 20 X 10-^ psi (138 MPa) when the strain level is about 4 x 10"^ in. /in. (cm/cm). The

5-ft- (1.5-m-) thick clay layer located at a depth of 26 to 31 ft (7.9 to 9.4 m) in boring

821 UD has a Young's modulus at low strain levels of about 43 x 10-^ psi (296 MPa). Again,

as the strain increases from a level of around 10"^ to 10"^ in. /in. (cm/cm), the Young's

moduli decreases to approximately 16 x 10-^ psi (110 MPa).

As one looks at this collection of data, it is again apparent that the silt layer at a

depth of 22 to 26 ft (6.7 to 7.9 m) is relatively stiff compared with the other layers. The

two clay layers above 22 ft (6.7 m) have similar characteristics, with the layer between 14

and 22 ft (4.3 and 6.7 m) being more sensitive to strain than the surficial layer.

- Damping Ratio

Figure 21 presents the damping ratio versus dynamic strain amplitude as a function of

depth for boring 821 UD. In dynamic analyses and in the calculation of dynamic response of

soils, the predominant energy input comes from shear waves. Consequently, shear or torsional

damping is the most important consideration, and only shear damping will be discussed here.

The longitudinal damping data were sometimes atypical [3] and should be considered suspect

(see Figure 14). Close study of Figure 21 reveals that damping is relatively constant as a

function of depth for boring 821. For strains less than 10"^ rad/rad the damping ratio is

III-99
on the order of 3 percent. As the strains increase above 10~ , then the damping ratio

approaches 12 percent. Based on WES experience, this appears to be reasonable and consistent .

The collection of data presented elsewhere [3] was obtained from 35 resonant column

tests. In general, shear and compression moduli decrease as a function of increasing strain

[4,6,7] while damping increases.

COMPARISON OF FIELD AND LABORATORY VALUES . ; , y ., ,

Figure 22 shows a boring log for boring 821 UD along with laboratory value s of shear

modulus. Young's modulus, and damping as a function of depth at low 10 ^) strain levels.

The subdivision of the soil profile was based on soil type and laboratory visual classification. These values have been discussed previously.

For comparison purposes, the values that were determined in the field by geophysical

techniques are shown immediately to the right of the laboratory values. The reader is

reminded that the subdivision of the soil profile is based on field techniques and results

from changes in P- and S-wave velocities and is independent of material types or visual

classification. Consequently, it is not surprising that the soil profile is not subdivided

identically using data obtained in the laboratory and field.

Generally, moduli values obtained in the laboratory are less than in situ determined

values. In fact, the values obtained for laboratory tests may be as low as one half the

field values [11]. The fact that the data obtained during this investigation compare so

poorly is surprising. No consistent trend can be obtained from a comparison of the data

presented in Figure 22. For some zones (soil layers) the laboratory values are approximately

equal to the field values. For other zones the laboratory values are greater than the field

values and vice versa. Because of the scatter in the data, a discussion of the data comparison and the selection of design values are presented below. j

Shear and Young's Moduli or.'.jna',)

Near-Surface Material

. r', '•.:-<r' '»

In comparing the laboratory and field values, the thickness of the near-surface layer

was determined to be 14 ft (4.3 m) in the laboratory. This was based primarily on the boring

Ill-lOO
logs; however, in the field, using velocity data, the near-surface layer was chosen to be

only 8 ft (2.4 m) thick. The laboratory-determined shear modulus of 3.5 x 10-^ psi (24 MPa)

compares very favorably with the field-determined value of 3.3 x 10^ psi (22.7 MPa). For

practical purposes, these numbers are the same. The Young's modulus computed from the field

data was determined to be 9.7 x 103 psi (67 MPa). This value seems reasonable; however, it

is about one third of the laboratory-determined value, which is 26 x 10^ psi (180 MPa). This

value is judged to be too high. For the top layer, values of shear and Young's moduli

selected for design were 3.4 x 10-^ and 10 x 10^ psi (23.4 and 69 MPa), respectively.

Second Layer

Immediately below the near-surface layer is a second clay layer that ends at 20 to 22 ft

(6 to 6.7 m) depending on whether the field seismic data or the laboratory classifications of

samples obtained from the boring are accepted. The shear modulus as determined in the laboratory for this layer is 2.8 x 10^ psi (19.3 MPa). This value is slightly less than half of

the shear modulus (6.6 x 10^ psi) (45 MPa) that was determined in the field. Because the

soil is predominantly clay to a depth of 20 ft (6 m) , and because the shear modulus should

increase with increasing depth, a design shear modulus of about 5 x 10^ psi (34 MPa) was

selected. The values of Young's modulus range from 19.7 to 24.4 x 10^ psi (136 to 168 MPa)

depending on whether one chooses the field or laboratory values. These values are relatively

close (within about 20 percent of each other), and this is considered good agreement. A

value of 20 x lO^ psi (138 MPa) has been selected for design purposes.

Third Layer

The third layer selected for design purposes combines the two layers within the depth

interval of 22 to 31 ft (6.7 to 9.4 m) in the laboratory-determined values and is assumed to

be at a depth of 20 to 30 ft (6 to 9.1 m) . If one averages the shear modulus values of

16.7 X 103 and 5.56 x lO^ psi (115 and 38 MPa), one obtains 11.1 x 10^ psi (76.5 MPa). This

value is in close agreement with 12.6 x 10-^ psi (87 MPa) that was obtained in the field, and

consequently a value of 12 x 10^ psi (83 MPa) was assigned for design purposes. In the

laboratory, averaging the Young's moduli for the two layers between 22 and 31 ft (6.7 and

9.4 m) yields a value of about 50 x 10-^ psi (344 MPa). This is considerably higher than

III-lOl
Young's modulus of 27.1 x lO-' psi (256 MPa) as determined by geophysical methods. Based on

engineering judgment, a value of 40 x 10^ psi (276 MPa) has been selected for design purposes, n

Fourth Layer

From a depth of 30 to 36 ft (9.1 to 11 m), the soil consisted of sandy silts with

gravels and consequently was not suitable for laboratory testing. Shear and Young's moduli

of 86 X 10-^ psi (593 and 1550 MPa), respectively, were assumed for design purposes. These

values are based on the field information.

Foundation Rock

Finally, the shear and Young's moduli of the foundation rock were determined to be

125 X 10-^ psi (860 and 2550 MPa), respectively. These values are based solely on the field

geophysical data.

Damping Values

If analytical calculations are to be conducted, a damping value of 3 to 5 percent ., ,

should be used for low strain calculations. This value is based on resonant column data and

is valid only to a depth of 31 ft (9.4 m) because the soil below that depth was not tested.

However, because the damping appears to be uniform with depth, extrapolation to the bottom

layer (30 to 36 ft (9.1 to 11 m) in depth) does not seem unreasonable. ., ;,. ^. ,j

SUMMARY

Based on the data and the analysis presented, the soil at boring 821 UD can be subdivided and zoned as shown in Figure 22 for the purpose of dynamic analysis of building or

machine foundations. This figure shows four soil layers above rock. The near-surface layer

is 8 ft (2.4 m) thick, the second layer is 12 ft (3.6 m) thick, the third layer is 10 ft

(3 m) thick, and there is a 6 ft (1.8 m) thick layer immediately overlying the rock. This

is an idealized soil profile that would be suitable for design calculations. The values of

shear modulus. Young's modulus, and damping ratio for each of these layers to be used in

design are listed in the right hand three columns of Figure 22. :

III-102
When there exists considerable scatter in the laboratory and field data obtained for a

specific soil layer, WES usually places more emphasis on the field data. The reasons for

this are :

- <

a) The field investigation usually involves the use of several procedures (such as the

crosshole, downhole, surface refraction, and vibratory techniques), which produce redundant

data.

b) The laboratory investigation is usually limited to resonant column testing.

c) The data obtained in the laboratory is influenced by such things as sample

disturbance, equipment limitations and boundary conditions, etc.

d) Furthermore, the laboratory data is only as valid as the specimen tests is representative of the soil layer. In this regard the field investigation is influenced by a volume

of soil measured in terms of cubic meters not cubic centimeters as in the laboratory.

ACKNOWLEDGMENTS

Grateful appreciation is expressed to those colleagues at WES who assisted in the data

acquisition phase of this study and to those who offered constructive criticism during the

documentation stage. The support of the Energy Research and Development Administration

(ERDA) and their permission to published this paper are gratefully acknowledged.

REFERENCES

[1] Ballard, R. F., Jr., "A Method for Crosshole Seismic Testing," Journal Geotechnical

Engineering Division, American Society of Civil Engineers, Vol. 102, No. GT12, December

1976, pp. 1261-1273.

[2] Butler, D. K. , Skoglund, G. R. , and Landers, G. B., "CROSSHOLE: An Interpretive

Computer Code for Crosshole Seismic Test Results, Documentation, and Examples,"

Miscellaneous Paper S-78-8, U.S. Army Engineer Waterways Experiment Station, CE,

Vicksburg, Mississippi, July 1978.

[3] Curro, J. R. , and Marcuson, W. F., Ill, "In Situ and Laboratory Determinations of Shear

and Young's Moduli for the Portsmouth, Ohio Gasious Diffusion Add-on Site," Miscellaneous Paper S-78-12, U.S. Army Engineer Waterways Experiment Station, CE, Vicksburg,

Mississippi, August 1978.

[4] Drnevich, V. P., Hardin, B. 0., and Shippy, D. J., "Modulus and Damping of Soils by

the Resonant Column Method," Dynamic Geotechnical Testing, American Society for Testing

and Materials , STP 654, Philadelphia, Pennsylvania, September 1978, pp. 91-125.

[5] Fry, Z. B., "A Procedure for Determining Elastic Moduli of Soils by Field Vibratory

Techniques," Miscellaneous Paper No. 4-577, U.S. Army Engineer Waterways Experiment

Station, CE, Vicksburg, Mississippi, June 1963.

III-103
[6] Hardin, B. 0., and Drnevich, V. P., "Shear Modulus and Damping in Soils: Measurement

and Parameter Effects," Journal, Soil Mechanics and Foundations Division, American

Society of Civil Engineers , Vol. 98, No. SM6, June 1972, pp. 603-624.

[7] Hardin, B. 0., and Drnevich, V. P., "Shear Modulus and Damping in Soils: Design

Equations and Curves," Journal, Soil Mechanics and Foundations Division, Amer ican

Society of Civil Engineers , Vol. 98, No. SM7 , July 1972, pp. 667-692.

[8] Kolsky, H. , "Stress Waves in Solids," Dover Publications, Inc., New York, N.Y., 1963.

[9] Marcuson, W. F., Ill, and Wahls, H. E. , "Effects of Time on the Damping Ratio of

Clays," Dynamic Geotechnical Testing. American Society for Testing and Materials , SPT 654, Philadelphia, Pennsylvania, September 1978, pp. 126-147.

[10] Maxwell, A. A., and Fry, Z. B., "A Procedure for Determining Elastic Moduli of In Situ Soils by Dynamic Techniques," Miscellaneous Paper No. 4-933, U.S. Army Engineer

Waterways Experiment Station, CE, Vicksburg, Mississippi, October 1967.

[11] Richart, F. E. , Jr., Anderson D. G. , and Stokoe, K. H. , II, "Predicting In Situ

Strain-Dependent Shear Moduli of Soil," Proceedings, Sixth World Conference on Earthquake Engineering, New Delhi, Vol. 6, January 1977, pp. 6-159 to 6-164.

[12] Seed, H. B. , and Idriss, I. M. , "Soil Moduli and Damping Factors for Dynamic Response

Analyses," Report No. EERC 70-10, Earthquake Engineering Research Center, University

of California, Berkeley, December 1970.

[13] Taylor, H. M. , Jr., et al., "Title I Design Foundation Investigation for Static

Loading, Gaseous Diffusion Add-On Plant, Portsmouth, Ohio," Miscellaneous Paper S-77-

20, U.S. Army Engineer Waterways Experiment Station, CE, Vicksburg, Mississippi,

November 1977.

III-104
Fig. 1 Proposed Building Locations and Geologic Map of the Site

X-774

ELEVATIONS ON BASE OF FAT CLAY , MSL

|5| SHALE, MISSISSIPPIAN (CUYAHOGA)

^777t ALLUVIAL DEPOSITS OF LEAN CLAYS AND SILTS (PLEISTOCENE) LAKE BED DEPOSITS (FAT LAMINATED

CLAY) AT OR BENEATH THE SURFACE

(PLEISTOCENE)

1000 FT

305 M

NOTE: ELEVATIONS (REFERENCED TO MEAN SEA LEVEL) AND GRID COORDINATES ARE IN FEET. TO CONVERT FEET TO METRES

MULTIPLY BY 0.3048. E 8,000

III-105
NORTH

E 9850 E9830

t-97.5 m824 UD

S-18 O

DOWNHOLE

TEST LOCATION

826

825

S-17

N 8650 N 881 5 N 8630 N 8575

I I I I

-343 mS-16—

^823 O S-15 822° DOWNHOLE TEST LOCATION 7821 UD

N 7525 N 7710 N 7545 N 7470

I \ \ I

E 9890

LEGEND

825 o BOREHOLE USED FOR CROSSHOLE TEST AND NO.

N 8650 COORDINATE, FT SHOTPOINT

S-18, SURFACE REFRACTION SEISMIC TRAVERSE

NO. AND DIRECTION

NOTE: THE ELEVATION (TOP OF BORING 821 WAS 671 FT. WITH THIS INFORMATION

THE READER CAN CONVERT DEPTH TO

ELEVATION AND VICE VERSA, AS REQUIRED.

Fig. 2 In Situ Seismic Test Layout, Proposed Building X-340-4 Location

NORTH

-343 m

674 —

tr 654 -

634 -

(BORINGS 824 UD, 825, 826)

235

140

£ 140

o

Oj 215 >^ 215 <^ 640

640

380

3| 1570

3 1480 53 i 1395 5

o

3 1130 si> 3 1 1 30 %

<

2095 5

(BORINGS 821 UD, 822, 823) —

3 2095

665 : 2510

TOP OF SHALE

LEGEND O -SEISMIC SOURCE AND GEOPHONE LOCATIONS

USED FOR CROSSHOLE TESTS

NOTE: ALL VELOCITIES ARE MPS

115

190

135

135 205

410

570 570

G 415 3 1135 3 1410 G 1540 3 1480 3 1480 3 2240 3 2240

Fig. 3 P- and S-Wave Velocities Determined from Crosshole Tests

Proposed Building X-340-4 Location

III-106
oclties

,

631-

Fig. 4 Arrival Time Versus Depth from Downhole Test, Boring 821,

Proposed Building X-340-4 Location . ,

Fig. 5 Arrival Time Versus Depth from Downhole Test, Boring 824,

Proposed Building X-340-4 Location

III-107
0 10 20 30 1*0 50 60 TO Distance, m Fig. 6 P-Wave Arrival Time Versus Distance, Proposed Building

X-340-4 Location, S-15 and S-16

North

Legend: G Traverse S-l6

X Traverse S-15

V Average velocity

30 1*0 Distance, m

Fig. 7 P-Wave Arrival Time Versus Distance, Proposed Building

X-340-4 Location, S-17 and S-18

III-108
North ^

31*3 m

67"* _

651* -

Borings 82I1

, 825, 826

1,10**

llt30*

631.-

3765**

0

1r

0

17ltO* 0

- 0

675* 0

2235*

380 1570

11*80 1395 1130 ~

1130

2095 —

2095 2510

1*25

1500

850

Borings 821, 822, 823

1*70*

2165'

_ltl5 1135

ll*10

151*0

11*80 "11*80

22l*0

22l*0

T

ltlO*»

1355*

760*

2865

151*5**

2950*

Legend: O Seismic source and geophone locations used for crosshole

tests

.

Note: All velocities are mps . No asterisk

denotes velocities from crosshole tests.

One asterisk denotes velocities from dovmhole tests. Two asterisks denote velocities from refraction seismic tests

Fig. 8 P-Wave Velocity Profile from In Situ Seismic Tests,

Proposed Building X-340-4 Location

31*3 m

North

t 651*-

631*-

Borings 82U, 825, 826

9,^*

155*

190* \560* 1

YO 105 Y

= 202 kgm/m

ll*0

ll*0

ll*0 215

215

61*0

61*0 665'

150

215

535

610

196 kgm/m

3 190

135

135

Y

= 192 kgm/m^ 205 w

1*10 = 208 kgm/m^

Y

= 232 kgm/m^ 570

Borings 821, 822, 823

125* T160* I.

235*

505*

Legend: O Seismic source and geophone locations used for crosshole tests.

Wet unit weight kgm/m

Note: All velocities are mps. No asterisk denotes velocities

from crosshole tests. One asterisk denotes velocities

from downhole tests.

Note: error m • for ex. 202 kgra/m^ should be 20.2 kN/m^ , etc. w

Fig. 9 S-Wave Velocity Profile from In Situ Seismic Tests,

Proposed Building X-340-4 Location

III-109
Fig 10. Shear and Young's Moduli Profile frora In Situ Seismic Tests,

Proposed Building X-304-4 Location

6lkBoring Lab

821* Class.

Borings

82lt 825 826

65"* -

631*-

CLAY. CH CUAY. CL CLAY. CH 2

CLAY. CL

CLAY. CH

r CL AY EY LSAND. SC CLAY, CL

CONGLOM- ERATE

3h3 mG = 22.75 MPa E = 66.88 MPa ^0.!+7

G = I15.5O MPa E = 135.83 MPa ^=0M

G = 86.87 MPa E = 255.80 MPa v=0.1*8

G = 592.28 MPa E = 1551.37 MPa ^_2j.^h

North

Borings

821

822 Lab Boring

823 Class. 821 —

G = 862.55 MPa E = 251*9.77 MPa v=0.1t8

III-llO
1 1 1 [

1 1 1 1 ! 1 1 11 1 1 1 1 1 1

1

1 1 1 1 1 1 1 1 M 1 1

LEGEND

0 LONGITUDIN

0 TORSIONAL

1 1 1 11 M 1

AL

-

D Q ^v.^*- YOUNG' S MODULUS. E

-

-

1 1 1 1 1 1 1

1

^SH EAR MOC ULUS. G

1 1 1 1 1 1 1 1

0

o

°

1 1 1 1 1 1 1

1

1 1 1 1 1 1 1 1 1 1 1 1 1 1 M

10 = 10-" DYNAMIC STRAIN t , , RAD/RAD, CM/CM Fig. 11 Dynamic Modulus Versus Dynamic Strain Amplitude from Resonant

Column Tests, Sample: 821 UD No. 2-1, a =11.7 kPa

B = 0.18, Nonsaturated

"T 1 II I I n I I

I I 1

1

T t I I

I I 1

1

1

LEGEND

n LONGITUDINAL

o TORSIONAL

T I

I

I I I I

« 15

I I I I I I I I I

I I

I I I I I I I 1 1 I I I I I I l_ I I I I I I I I I I I I I I I I I ,0"* 10"^ 10 * 10"^ DYNAMIC STRAIN , RAD/RAD, CM/CM

to

Fig. 12 Damping Ratio Versus Dynamic Strain Amplitude from Resonant

Column Tests, Sample: 821 UD No. 2-1, a ^ = 11.7 kPa,

B = 0.18, Nonsaturated

III-lll
-

1 [

[

1—III] I 1 1 ]

1 TTT 1 1 1

1

1 1 1 1

'

LEGEND

LONGITUDIN

0 TORSIONAL

AL

-

-

Nv YOU NG'S MODULUS, E

-

1 1 1 1 1 1 1

1

1 1 1 1 1 1 M

0 o 0

u

/ SHEAR MODU LUS. G

1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 ^ 1 1 1 1 1 1 1

DYNAMIC STRAIN f j , RAD/RAO, CM/CM Fig. 13 Dynamic Modulus Versus Dynamic Strain Amplitude from Resonant Column Tests, Sample: 821 UD No. 9-1. a = 41 4 ]cPa B = 0.68 ' oct "^J^a.

1 1 1

1

1 1 1

1

1 1 1 11 1 1 1 1 1 1

1 M 1 1 1 1 1

1 1 M 1 LEGEND

LONGITUDIN*

0 TORSIONAL

1

1 1 1 1 II 1 1

L

n

1 1 1 1 M 1 1 1 1 I 1 1 1 1 1

D °

1 1 1 1 1 1 1 1

D —

1 1 1 1 1 1 1 1 ; 1 1 1 1 1 II 10"' 10"" DYNAMIC STRAIN f j , RAD/RAD, CM/CM

Fig. 14 Damping Ratio Versus Dynamic Strain Amplitude from Resonant

Column Tests, Sample: 821 UD No. 9-1, o = 41 4 kPa

B = 0.68

III-112
1 1—I

I

I I 1

1

1

1 1—I

I

I I 1

1

n I I {III! n I I I M 1

1

1 I

1—I

I

I M I LEGEND

LONGITUDINAL

o TORSIONAL

5

I I I J I

I I I I I I

DYNAMIC STRAIN < , , RAD/RAD, CM/CM

Fig. 16 Damping Ratio Versus Dynamic Strain Amplitude from Resonant

Column Tests, Sample: 821 UD No. 11-1, a^^^^ = 65.5 kPa,

B = 0.83

III-113
1 1 1 1

1 1 1

1

1 1

1 1

1 1 1 1 1 1 11

1 1 1 1

1 1 1 I

1 1 1 1 LEGEND

0 LONGITUDIN/l

o TORSIONAL

1 1 1 1

1 M 1

L

D

-0 "

" D 0

n

1 1 1 1 1 1 1 1

1 1 1 1 1 1 1 1

00

o -a 0

1 1 1 1 1 1 1 1

1 1 1 1 1 1 1 1

1 1 1 1 1 1 1 1 10"^ 10"' 10 DYNAMIC STRAIN ( , , RAD/RAD, CM/CM Fig. 18 Damping Ratio Versus Dynamic Strain Amplitude from Resonant

Column Tests, Sample: 821 UD No. 13-1, a^^^ = 74 kPa,

B = 0.91

III-114
Fig. 19 Dynamic Shear Modulus Versus Dynamic Shear Strain

Amplitude from Resonant Column Tests, Boring 821 UD

~] I

I

I I I I ~i I I I ] I I ~1—I Mill T 1 1—I I I I 1

CL»Y X

in 2 J 0

I 5 io-«

Ill >- a

J I I I I I I I

-1 1—I—I I I I I

_i I I I I I I I J I

I I I I I I

-|—I I I I I "T 1 1—TTT

10-2

3=1 1 20

5 10 <

Z «

I I I

I I I I I 10-6 ro-5

1

1

1

I M I I

J I I

I I I I I

~l I

I

I I I I

. 40

20

J I I I I I I I I ] I I I I I I I I I I I _l _l t I I I I I 10-5 10--* 10-3 DYNAMIC SHEAR STRAIN AMPLITUDE^ R A D/ R A D

III-115
Fig. 20 Dynamic Young's Modulus Versus Dynamic Strain

Amplitude from Resonant Column Tests, Boring 821 UD

III-116
Fig. 21 Damping Ratio Versus Dynamic Strain Amplitude

from Resonant Column Tests, Boring 821 UD

1 1 1 11 1 11 1 1 1

1 1 1 1

1

1 1 1

1 1 1 1

1

1 1 1 11 1 1

1

o

6

1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1

1 1 1 1 M 1 1 1 1 1 1 M 1 1 1 1 1 1

1 1 1

1

1 1 1 1 1 1 1 1

1 1 1 1 1 1 1 1

—0 -o

1 1 1 1 1 1 1

1

1 1 1 1 M M 1 1 1 1 1 1 1 1

llj

g 10"* <

o: 10

10-2

1 1 1 1

1 1 1

1

1 1 1 f M 1 1 1 1 1 1 M 1 1 1 1 1 1 1 1 1 1

1 1 1 1 1 1 1 1

1 1 1 1 1 1 1 1

6> 0 o1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 10-3

1 1 1 1

1 ! n 1 1 1 11 1 1

1

1 1 1 1 M 1 1 1 1 1 1 M 1 1

1 1 1 1 1 1 1

1

oo r o o

1 1 1 1 1 1 1

1

1 1 1 1 1 1 M 1 1 1 1 1 1 1 1 10-8 10-5 10-"* 10-3 DYNAMIC STRAIN AMPLITUDE , RAD/RAD

10-2

III-117
Fig. 22 Field and Laboratory Comparison of Shear and Young's Moduli

with Resultant Design Values for Boring 821 UD

_|Wt,

E

J::::

_ g

V./0

1 CR TOP OF f

r LOR ET A L. (REF 1

YOUNGS MODULUS SMEAR MODULUS YOUNGS MODULUS

III-118
AN EXPKRIMENTAL STUDY ON THE LIQUEFACTION OF

SANDY SOILS IN A COHESIVE SOIL LAYER

Tatsuo Asama

Yukltake Shioi

Public Works Research Institute

Ministry of Construction

,Ti.,t l, ABSTRACT . ; . : r

Experimental and theoretical researches on liquefaction have been made by many researchers. There is a general understanding that liquefaction occurs when the shear stress reaches

a critical value which is determined by the type of soil, its density, its normal stress, and

the like. The previous research shows that an essential factor should be stress levels in

the soils. But the authors observe that liquefaction chiefly depends upon strain levels of

soils. Especially where a sand layer lies on a soft cohesive soil layer, the strain can be

amplified by the response of the cohesive soils.

KEYWORDS: Cohesion in soil; liquefaction; sandy soils; strain levels; stress levels.

a •ni 1 .... •)•

III-119
1. INTRODUCTION

There are a number of research works on the liquefaction of saturated sandy soils. However, upon examination of the results of researches against the phenomena occurring in the

structures or soils, they are not always correlated with each other. For example, the level

of acceleration and duration and vibration causing liquefaction in the experiments are not

observed at the sites of liquefaction. Further, the acceleration is not in correspondence

to the damage to the structures in the vicinity. Still further the soils generating liquefaction are not distinguishable clearly from those generating no liquefaction. In the Fukui

Earthquake, liquefaction occurred in the gravel soil along the Kuzuryu River, while in the

Niigata Earthquake, slight liquefaction occurred in the dune zone. Further, the phenomena

of water and sand spouting were seen after the main shock of the earthquake.

Where large scale road construction works are carried out, soils having a high possibility of liquefaction are distributed extensively as in the estuary of a large river or a

coastal or reclaimed area. Great is the influence of liquefaction of a sand layer at the

time of an earthquake upon the design parameters of a large foundation provided in such a

place. It is then necessary to re-examine the captioned phenomenon to clarify its mechanism

so that adequate measures can be taken.

Before starting the investigation, the process of generation of liquefaction was pursued

theoretically

.

Principal elements of an earthquake wave will be the waveform, amplitude, cycle, and

duration. These elements represent one or another form of energy respectively. The earthquake wave may be taken as a phenomenon of an energy accumulated in the crust being transmitted as a variable dynamic energy in all directions through the destructive phenomenon at

the seismic center. As such earthquake wave motion comes to the surface, it causes damage to

structures. The damage occurs generally in the forms of destruction of structures and loss

of stability such as overturn, sliding and sinking. The damage due to liquefaction of the

soil corresponds to the latter. By liquefaction, the soil loses the bearing force. But, the

loss of bearing force is due to elevation of the pore water pressure and decrease of the

effective stress of the bearing ground. The cause is the shear strain of the soil. As seen

from Figure 1, when a shearing strain is brought into the saturated soil by the action of a

shearing force t^, there is generated an excessive pore water pressure Oq against the

III-120
decrease in the volume if there is no change in the balance of water. To compensate for the

I external force Pj^ , the effective stress a-^ decreases to a-^, and so the shear resistance.

t When the shearing deformation proceeds further to the stage of o-^ at zero, it represents

fluidization or liquefaction of the natural ground. In the case of the foundation of a

structure, the shear resistance of the ground decreases with decreasing effective stress so

that the bearing force is shortcoming to cause tilting or sinking.

On the other hand, the shearing strain is a kind of work quantity so that it is necessary to look for the source of the corresponding energy. In general, waves are apt to shift

from a hard medium to a soft medium but scarcely from a soft medium to a hard medium. Thus,

the waves transmitted from the crust are apt to be stored near the surface where relatively

soft soil layers are distributed. Then, it is assumed that the energy thus stored gives a

shearing strain to the upper saturated sand layers and thus causes liquefaction. From such

a point of view, review was made of the soil conditions of the points at which liquefaction

occurred in the past earthquakes, and it was found that they had invariably highly strain

following soils such as clay or silt distributed thereunder. The case of Fukui Basin is

illustrated in Figure 2. Thus, here, the mechanism of liquefaction will be examined according to the foregoing assumption.

2. OBJECTS OF I^fVESTIGATION

It is intended, in order for reasonable design of the foundation of a structure on a

saturated sandy soil, to clarify the mechanism of the phenomenon of liquefaction of the

soil at the time of an earthquake and also to examine the load upon, and the properties of

strain of, the foundation. ,„ i . .

'

3. METHOD OF INVESTIGATION

In 1977, vibration experiments by means of models were carried out in order to test the

validity of the aforementioned assumption. .

The experiments were made with the six models shown in Figure 3. Model 1 was designed

for a cohesive layer beneath a sand layer. It was prepared with a grouting material and was

intended for investigation of the performance upon vibration. Model 2 was intended for

investigation of any change in the vibratory performance of said cohesive layer under the

influence of a surcharge. Model 3 was Intended for investigation of the vibratory

III-121
performance of a cohesive layer the upper part of which was restrained by rigid panel.

Model 4 was intended for investigation of the vibratory performance of the cohesive layer

having a dry sand layer placed thereover in reference to the case of the saturated sand

layer of Model 5. Model 5 was intended for investigation of the vibratory performance of

a saturated sand layer on the cohesive layer and the process of liquefaction. Model 6 was

designed to follow the elevation of pore water pressure and the process of liquefaction with

a rubber coat laid over the saturated sand for assumption of an impermeable layer (Photos 1

and 2). , .

The materials used in the models and the layout of instruments are shown in Figures 4

and 5 . , - - ,

>

For the model of the cohesive layer, plastic materials were used in a thickness of

40 cm. They were compounded so as to give a natural frequency of 7 8 Hz . The sand was

that occurring at Takahagi, Ibaraki Prefecture in a proportion of grain sizes adapted to

flow. As measuring instruments, 7 accelerometers (A) were installed on a seat and in the

cohesive and sand layers as shown in Figure 5. For measurement of the changing soil pressure, meters (D) were installed on the side of the respective vibration boxes at the portions

corresponding to the cohesive and sand layers. To catch the symptom of liquefaction, 2 pore

water pressure gauges (P) were installed on those parts of the side wall which corresponded

to the sand layer. For measurement of the strain of cohesive and sand layers, a phosphorbronze plate having 8 strain gauges (S) affixed to it was installed at the center of the

respective vibration boxes (Photo 3). , v .

.

The shaking table was of the specification shown in Table 1.

The inputs were in the forms of irregular waves, sine waves and shock waves, and they

were caused to act over the range of 20 gal to 500 gal depending on the respective objects.

The irregular wave was of white noise containing frequency components of 5 ^ 40 Hz . It was

reproduced by a data recorder and introduced into the control panel of the shaking table. ,

For the sine waves, frequencies of 5, 10, 15, and 20 Hz were used for Models 1^3, and •

those of 10 and 20 Hz for Models h ^ 6. The shock waves were produced by a method giving

an instantaneous displacement to the shaking table control unit manually, and the subsequent

free vibrations were measured (Table 2). ' ' ' i-'b.^

III-122
Recording was made by a data recorder, pen recorder, or magnetic oscillograph. Analog

data recorded on a magnetic tape were converted into actual response waveforms by AD conversion. Sampling was 0.005 seconds. -

Waveforms thus obtained were arranged as below.

With respect to the irregular waveforms, there were obtained power spectrums Gx(f) and

Gy(f) and cross spectrum Gxy(f) of the waveforms by the accelerometer . Gx(f) is related to

the input waveform, and Gy(f) to the ground response waveform. From the relationship between

Gx(f) and Gxy(f), the frequency response function H(f) and phase difference 6(f) were calculated according to the formulas " '

;

. i.

H(f) = ^y(f) (1)

Gx(f ) c. . , . ;

6(f) = tan-1 ( I(Gxy(f))

) (2)

'R(Gxy(f))'

where R(Gxy(f)) represents the real portion of Gxy(f), and I(Gxy(f)) the imaginary portion of

Gxy(f).

For the sine waveforms, the maximum response values, response magnifications, and the

phase differences were obtained from the data of measurement by the accelerometer (A), soil

pressure meter (D), pore water pressure gauge (D), and embedded strain gauge (S).

For the shock waveforms, free damping vibrations and damping coefficients were obtained

so far as practicable from the free damping waveforms.

4. EXPERIMENT RESULTS : . -' -

Vibration tests of the models were conducted in the order of irregular waveform, sine

waveform and shock waveform. However, for each vibration test, the models or sand were not

changed so that the model grounds or, more particularly, sand grounds in the later tests were

subject to the compacting and other effects given in the preceding tests.

Responses of the model grounds to the irregular waveforms were arranged in use of the

spectrums shown in Figure 6. In Figure 6, the upper diagram shows the power spectrums Gx(f)

and Gy(f) of the input and response and the cross spectrum Gxy(f) between them. The middle

III-123
diagram illustrates the frequency response function H(f), and the lower diagram illustrates

a curve of phase difference (f). Marked o in the diagrams are the response values upon input

of sinusoidal waveforms.

The values of prevailing frequency of vibration at the positions of accelerometers in

the respective models as calculated from such response curves were as given in Table 3. In

Model 1, the frequency of vibration of the layer of plastics failed to come to the expected

value of 7 ^ 8 Hz , and this was found, from the analysis executed later, to be due to the

restraining effect of the side wall. '

The process of liquefaction by irregular waveforms is illustrated in Figure 7. In ^

-

Model 5, the liquefaction occurred with the irregular waveform at 500 gal and sine waveform

of 10 Hz at 400 gal. At a higher level of acceleration or with a sine wave of 20 Hz, 400 gal,

a phenomenon close to liquefaction appears. In Model 6, liquefaction occurs with an irregular waveform of 200 gal and a sine waveform of 10 Hz, 300 gal.

The values of maximum response magnification of the respective grounds by sine waves

were given as shown in Table 4. The accelerometers A2, A3 and A6 were positioned at the

center of the respective test tanks.

Free vibrations by shock waves tended to damp quickly so that it was difficult to read

the succeeding waveforms except in Model 2 where a damping frequency of vibration at 6.6 Hz

and a damping constant of 0.025 was observed.

Subsidence of the upper sand layers in the shaking tests was about 1 cm in Model 1 or

about 5.5 cm in total in Model 5 including 1.9 cm due to liquefaction. In Model 6, it

totaled 3.6 cm including 2.3 cm due to liquefaction.

5. DISCUSSION •' '

The experiments were designed to resolve the problems described in the summary and had

a character of preliminary examination. The results were not always satisfactory, but we

were able to find some tendencies.

In Model 1, there was observed a good agreement between the irregular waveform and the

sine waveform with respect to the results of vibration of the lower cohesive layer, as shown

in Figure 8. This means that the frequency of vibration obtainable from the function of

III-I24
transfer by the irregular waveform is usable for calculation of the natural frequency. It

may be used in the experiments to be conducted hereafter. Marked o in the diagrams are the

values of response upon input of sine waveforms.

From the frequency of response at 10.9 Hz in Model 1, the ground constants of the

cohesive layer may be estimated as

(Propagation speed) Vg = Ahfg = 4 x 40 x 10.9 = 1744 cm/sec,

(Shear coefficient) G = • p/g = 1744^ x 1.0 x 10"^/980

=3.1 kg/cm^, and

(Elastic modulus) E = 2(1 + v)G = 2(1 + 0.5) x 3.0 = 9.3 kg/cm^

where h = Height of the cohesive layer; •

' ' '

. -

fs = Frequency of resonance; •

' '

p = Density of the cohesive layer; "

:

g = Gravity; and .

•

. i^^f r

V = Polsson's ratio. -Si' " • J . -

Upon input of these ground constants to a multiple reflection model a resonance appeared

at about 11.0 Hz as shown in Figure 8. '

•

In Model 2, water was laid in a depth of 30 cm over the cohesive layer. The results of

measurement showed the values of 24.1 Hz and 20.7 Hz respectively at the positions of A2 and

A3 as shown in Table 3. In Model 1, they were equal at 10.9 Hz so that it was found that the

weight effect would cause greater influence onto the vibration of lower layer. This was the

same in the case of sine waveforms and would, therefore, be reliable. But, how to express it

in a dynamic model is a problem to be resolved hereafter.

In Model 3, the upper surface of the cohesive layer was restrained by a rigid plate, but

its effect was not much and the frequencies of resonance were distributed from 9.6 Hz to 11.1

Hz. In this case, the response to sine waveform agreed well with that to irregular waveform.

Model 4 had a dry sand layer placed on the cohesive layer and could, therefore, be

regarded as a composite of Models 2 and 3. It was taken as a reference to Models 5 and 6.

In the vibration tests, no liquefaction occurred. Further, as seen from Table 5, there was

noted a trend of lower frequency of resonance with higher levels of acceleration. This trend

was noted more or less in each of the other models.

Model 5 may be regarded as a saturated Model 4. In the vibration tests, liquefaction

occurred with an irregular waveform at 10 Hz, 500 gal. In this case, the strain gauges on III-125
the phosphor-bronze plate showed a corresponding movement. Further, the accelerometers A6

and A7 exhibited response vibrations at lower response magnification so that it could be

assumed that there was no complete loss of shear resistance.

The frequencies of vibration prevailing before and after liquefaction are shown in

Table 6. Before and after the liquefaction, the values are more or less decreasing, but the

decrease is not substantial. From this, it is considered that although in the tests each

model had the vibrations applied successively the effect of the previous vibration in the

respective vibratory performances would not be great.

Further, the level of acceleration at liquefaction was as high as 400 to 500 gal or a

far greater value than that at the actual liquefaction. Thickness, permeability, surcharge

load, etc. of the sand layer may be considered to be the causative factors.

Model 6 has an impermeable film provided over Model 5. In the vibration with the

irregular waveform, liquefaction occurred at an acceleration of 200 gal. This is very different from the value of 500 gal in the case of Model 5, although both models had the grounds

prepared in the same way. In the case of the sine waveform, liquefaction occurred at 10 Hz,

300 gal. As to the reasons of the liquefaction being caused relatively readily, elevation of

the pore water pressure due to the presence of the impermeable film and increase of the

strain in the sand layer due to reflection of the input waves at the impermeable film may be

considered.

With respect to the prevailing frequencies of vibration before and after the liquefaction, they changed little; although there was more or less elevation after the liquefaction,

and they were similar to those in the case of Model 5. ... .... .

6. CONCLUSION

The foregoing experiments were conducted as part of the study designed for examination

of the resistance to earthquake of foundations installed on sand layers over cohesive soils.

Nevertheless, from the experiments, the following discovered.

1) The saturated sand layer on a cohesive soil is very much subject to the influence of

the vibratory energy accumulated and expanded in the cohesive soil.

2) It is presumed that upon liquefaction of the saturated sand layer, the vibratory

energy is converted, for the greater part, into the moment of work. ; ' —

III-126
3) When the upper soil has a high strength, vibration of the lower cohesive soil is

restrained so that the vibration as a whole is reduced.

4) In vibratory analysis of a cohesive soil, the weight of upper soils is by no means

negligible. With presence of the weight, there is an elevation of the value of apparent

elastic modulus of the lower cohesive soil.

5) The presence of an impermeable layer on the saturated sand layer exerts a great

influence upon the liquefaction.

Some other phenomena deserving study were also noted, and they will be examined through

a succeeding series of experiments. As the next stage, it is planned to examine the effects

of thickness of the saturated sand layer as well as the cohesive soil, responses of the

foundations installed in such soils, etc.

7. ACKNOWLEDGMENTS

The authors are grateful for Mr. Toshio Iwasaki of the Public Works Research Institute,

Dr. Fumio Tatsuoka of the Tokyo University and Dr. Eiji Yanagizawa of the Tohoku University

for their valuable advice in planning the experiments. Execution of the experiments owes

greatly to the assiduous supports of Messrs. Takeaki Tomozawa, Omichi Ukon and Takeshi

Onomura, Long Span Bridge Design Center. We should also like to thank Messrs. Ichiro

Watanabe and Yoji Yamamoto, Public Works Research Institute, for management of the experiments. ..

1 Si fft.

1 f- 1

"1

IFig. 1 Effects of shear deformation

of soil

/o 20 jn dfi CO o-

+0 00

-J.ti

-8.)S

/o20-

-// fs

-II 3i 30-

r

«)-

so6l>-

, - ton

Fig. 2 Typical Column of geology at Fukui

III-127
Nitto SS layer

(Clay assumed)

V/ater

Clay assumed

0.1 1.0

Takahagi Sand, Ibaraki Prefecture

Specific gravity : 2.7 Maximum grain size: 1.0 mm Minimum grain size: O.OSiran 60% Grain size: 10% Grain size: 0.6 mm

0 . 3 mm Uniformity coefficient: 2,0

Model 1

Top plate(Veneer

^ £"t - 1 Omm , \i= ^ g )

Clay assumed

Model 2 o To

Dry sand

Clay assumed

Model 3

vSAturstad sard

Clay assumed

Model 4 Rubber (t=l. Omm, {~w=0.75kg) V

Clay assumed

Model 5 Model 6

Fig. 3 Kinds of models

To

Fig. 4 Physical properties

80, (cm)7060

50

AO

30

20

10

"13"" A7

D-4o

D-3o

A5

A°4

A-3 "J D-2 oA-20

D-1 o

S-8

S-7 A P-2

S-6 A P-1

S-5

S-4 S-3

S-2

S-1

//////////// 1 //////>,/;'//!////////////f////^ <Legend> O ; Accelerometer

o ; Soil pressure meter A ; Pore water pressure gauge

I : Embedded strain gauge

Fig. 5 Arrangement of measuring

instruments

111-1^8
PHASE fiNOLE

funs; KKOt-t

-ISO0

Fig. 8 Response frequency

characteristics

(Model 1)

III-129
Photo 3 Model box and measuring device

III-130
Table 1 Specifications of shaking table

Performance

Actuator

Max. exciting force: ?.5 ton-G(45 kip) or more.

Max. amplitude: lOOmra (4 Inches) or more.

Max. speed: 76 cm/sec(30 inches/sec) or more.

Frequency range: DC'\'100 Hz.

Low frequency oscillator

Oscillating waveform: Sine wavcj rectangular wave and triangular wave.

Oscillating frequency: 0.01 Hz^lOOO Hz or more.

Output voltage: ±10V (when fixed).

Hydraulic

source

Pressure: 210 kg/cm'.

Flow rate: 31 H/min or more to satisfy the foregoing specification.

Shaking table

Linearity: ±0.05%.

Drift: 0.5%/C (full scale).

Noise: 0.5% of full scale or less.

Controllable min. amplitude: ±0.31mm or less.

Accuracy: 10%.

Table 2 Experimental conditions

Measurement components

Number

of records Model Input waveform Nominal Input acceleration Accelerometer

Soil pressure

meter

Pore water pressure

gauge

Embedded strain

gauge

Number of components

Irregular waveform 100 gal A-1 D-1 S.-1 3

1 Sine waveform 100 gal A-2

A3

D-2 S-2

S-3

11 4 Shock waveform Shaking table A-A

A-5

S-4 3

Irregular waveform 100 gal A-1 D-1 S-1 3

2 Sine waveform 100 gal

A-2

A3

D-2 S-2

S-3 11 4 Shock waveform Shaking table A-i

A-5

S-4 3

Irregular waveform 100 gal A-1 D-1 S-1 3 3 Sine waveform 100 gal A-2

A3

D-2 S-2 S-3 11 4 Shock waveform Shaking table A-4

A-5

S-4 3

Irregular waveform 20, 50, 100, 150 gal A-1 A-6 D-3 P-1 S-1 S-5 3

« Sine waveform 20, 50, 100 gal A-2 A-7

A3

D-4 P-2 S-2 S-6 S-3 17 6 Shock waveform Shaking table A-4

A-5

S-4 3

Irregular waveform 100, 150, 250, 500 gal A-1 A-6 D-3 P-1 S-1 S-5 3

5 Sine waveform 200, 'lOO, 500 gal

A-2 A-7 A-3

D-4 P-2 S-2 S-6

S-3

17 6

Shock waveform

Shaking table and movable

wall Impact A-4A-5

S-4 5

Irregular waveform 100, 200, 300 gal A-1 A-6 D-3 P-1 S-1 S-5 3

6 Sine waveform 100, 200, 300 gal A-2 A-7 A-3

D-4 P-2 S-2 S-6

S-3 17 6

Shock waveform

Shaking table and movable

wall impact A-4

A-5

S-4 5

Note) In Che tests by sine and shock, waveforms In Models 5 and 6, measurement was toade after occurrence of liquefaction.

III-131
Model No. 1 2 3 A 5 5' 6 6'

Acted acceleration 100 100 100 100 100 250 100 300

A2 10.9 2A,1 11.1 21.5 20.0 23.2 18.0 20.5

A3 10.9 20.7 10.3 20.7 18.5 17.3 18.0 17.8

A4 10.9 24.1 9.6 21.1 18.5 17.3 18.0 19.0

A5 10.9 20.7 10.9 20.8 18.3 16.6 18.0 16.8 A6 20.9 18.0 16.3 18.0 17.5

A7 21,1 18.3 17.8 18.0 17.2

Note) 5 and 6 for input of 100 gal.

5' for Input of 250 gal, and 6' for input of 300 gal.

Table 3 Prevailing frequencies

of vibrations in the

ground

Table 4 Response magnifications

in ground (at 10 Hz)

Model No. 1 2 3 4 5 5' 6 6'

Acted

acceleration

100 100 100 100 200 400 200 300

A2 2.7 1.7 2.7 1.3 1.5 0.8 1.4 0.5 A3 3.2 1.6 3.4 1.3 1.7 1.1 1.6 0.8 A4 2.0 1.1 2.4 1.3 1.5 1.1 1.5 0.9 A5 3.1 1.3 2.9 1.2 1.6 1.1 1.4 0.8 A6 1.4 2.0 0.9 2.0 0.5

A7 2.9 1.9 0.7 1.6 0.1

Note) 5' having liquefaction at 10 Hz, 400 gal.

6' having liquefaction at 10 Hz, 300 gal.

Input

acceleration

Instrument

Prevailing frequency of vibration

20 gal 50 gal 100 gal 150 gal

A2

22.6 Hz 22.1 Hz 21.5 Hz 21.7 Hz A-3 22.6 22.9 21,7 19.9

A422.6 22.0 21.1 21.3

A5

22.6 22.0 20.8 20.0

A-6 22.8 22.1 20.9 20.2 A-7 22.8 22.1 21.1 20.6 A-2^A-5 are accelerometers installed in the clay

layer, and A-6 and -7 are those installed in the sand layer.

Table 5 Input acceleration versus

frequency of resonance

Table 6 Prevailing frequencies

of vibration before and

after liquefaction

^\Model

^v^ndition

Instrument^s^^

Prevailing frequency of vibration

Before liquefaction

Under

liquefaction

Af ter liquefaction

A-2 A-3 16.75 Hz

20.8 Hz 20.0 Hz 16.5 Hz A-4 18.5 Hz 20.75 Hz 18.0 Hz

A-5 16.75 Hz 19.0 Hz 15.5 Hz A-6 17.2 Hz 16.25 Hz

A-7 16.75 Hz 16,0 Hz III-132
STRESS-STRAIN BEHAVIOR OF DRY SAND AND NORMALLY CONSOLIDATED

CLAY BY INTER-LABORATORY COOPERATIVE CYCLIC SHEAR TESTS

Hiroshi Oh-oka

Kohjiroh Itoh

Yoshihiro Sugimura

• ^ • Masaya Hirosawa

Building Research Institute

Ministry of Construction

(,:/ , ABSTRACT - • '

'

' "-

Inter- laboratory cooperative cyclic shear tests were conducted to obtain fundamental

information about stress-strain behavior of dry sand and normally consolidated clay by using

various kinds of dynamic shear apparatus which have been developed in Japan.

The tests were carried out in order to examine the characteristics of test apparatus and

test procedures under as nearly identical conditions as possible. The results obtained by

these tests are compared and discussed.

KEYWORDS: Damping ratios; dynamic soil properties; shear-strain testing of sand and clay;

shear modulus; stress-strain soil behaviors; test procedures.

III-133
INTRODUCTION

Different soil deposits behave differently in an earthquake. When soil deposits support

foundations which in turn support structures, it follows that earthquake damage to the structures will vary according to the type of soil on which the structure is built. Therefore, it

is important to clarify basic stress-strain behavior of various soils by subjecting them to

cyclic shear stresses, and through analysis of shear modulus and damping ratios.

Up to this time, shear modulus and damping ratios have been measured with several types

of apparatus. However, little information has come to the writers' attention concerning the

test data which examined the characteristics of test apparatus and test procedures.

The ad hoc committee, "Dynamic Properties of Soils," headed by Professor Yoshiaki

Yoshimi of the Tokyo Institute of Technology was founded in 1974 to gather fundamental

information about stress-strain behavior of sand and clay using several types of apparatus

(Yoshimi, et al. , 1975, 1976, 1977). Some of these test results are reported here.

CHARACTERISTICS OF THE APPARATUS k ; ri, .iS'

The laboratory apparatus for dynamic shear tests can be classified into the following

three groups.

1) Shear stress is applied to the surface of the specimen; shear stress and strain are

measured directly. n

2) Principal stresses are applied periodically on the surface of the specimen; shear

stress and strain are measured directly. ' "

3) The physical quantities such as elastic wave velocity or resonant frequency of

soil specimens are measured; shear moduli and damping ratios are calculated from them.

The First Group

a) Simple shear apparatus - ^

Dynamic simple shear apparatus of the Roscoe type (Roscoe, 1953), which was employed by

Peacock (1968) and Finn (1971), has scarcely been used in Japan. On the other hand, the

NGI type devices or the new Kjellman type device (Hara and Kiyota 1977) have been

developed for simple cyclic shear tests.

III-134
b) Ring torsion apparatus

With the ring torsion apparatus, the ring-shaped specimen is subjected to dynamic shear

stresses with inertia torque working on weights which are placed on the specimen. Shear

strain is nearly uniform in the specimen, because the height of the specimen is proportional

to the radius (Yoshimi and Oh-oka , 1973). With both the simple shear and ring torsion apparatus, cyclic shear stresses can be applied under plane strain condition. However, one has

to estimate the effective stresses acting on the side surfaces of the specimen with Jaky's

equation, i.e., Kq = 1 -sinj5'

.

c) Torsional shear device ' ' ^ ' " '

'

The advantage of this apparatus lies in being able to clearly control and measure the

horizontal effective stresses (Iwasaki, Tatsouka, and Takagi, 1978). However, strain occurs

not under plane strain conditions but with a little lateral strain in the strict sense if

lateral strain is permitted.

The Second Group , . . . ' ' -

The dynamic triaxial compression apparatus, in which the strain of the specimen is

usually symmetric with respect to the vertical axis. The stress condition is extremely

clear except at the zone near the upper and lower ead of the specimen. It is necessary,

however, that shear strain be estimated with axial strain and dynamic Poisson's ratio which

is presumed or measured during shear. In the triaxial compression test, the effective mean

principal stress at compression is different from that at extension, if the lateral confining

pressure is constant. Also the effective mean principal stress changes during shear in

spite of its undesirability.

In this respect, special devices have been adapted for the dynamic compression apparatus

as follows :

~ '

' ' ''

a) An apparatus which is able to apply principal stresses which alternate axial and

lateral stress with a phase difference of 180° (Tamaoki 1976). Through this improvement,

the effective mean principal stress either becomes constant or the magnitude of the fluctuar tion becomes smaller.

'

b) An apparatus with which the dynamic Poisson's ratio can be measured during cyclic

shear (Suzuki, Sugimoto, Babasaki, and Kakita, 1976) (Yasuda and Sakai, 1976).

III-135
The Third Group , ;

a) Supersonic pulse apparatus . _

A supersoaic pulse is generated at one side of the specimen, and velocities of P-wave

and S-wave propagating through the specimen are measured at the other side (Sugimoto, Suzuki

and Terada, 1975). Shear modulus and Poisson's ratio are calculated from these velocities.

Shear strain is estimated to be very small in this test type. ^ ,, . . ;,

b) Resonant-column apparatus

The specimen is fixed at the bottom, torsional vibration is applied at the top (Koori,

1975). By changing the frequency of vibration, resonant frequency is found and with this,

shear modulus is calculated. Damping ratio is estimated by free vibration at the resonant

frequency by use of the same specimen. . . • < .

The devices above were developed in the research facilities to which each member of the

"Committee on Dynamic Properties of Soils" belong. i . . ^

TEST CONDITIONS OF INTER-LABORATORY COOPERATIVE CYCLIC SHEAR TESTS ...

Special attention was given to the following points:

1) In the series of tests on Toyoura sand, specimens were prepared uniformly to obtain

identical skeleton structure of soil specimens as well as the identical dry densities. Specifically, the air-dried saad grains were dropped into the mold from a constant height,

through a flexible tube with a constricted outlet.

2) In the clay tests, artifically consolidated clay samples were prepared for specimen

uniformity, and to minimize sample disturbance during the sampling procedure.

The Tests for Toyoura Sand j , - -

1) Common conditions : i-. '

a) Air-dried Toyoura sand gathered from the same spot was used. The physical

properties of the sand are as follows; specific gravity of 2.64, effective diameter of '

'

0.15 mm, uniformity of 1.4, maximum dry density of 1.628g/cm-^, minimum dry intensity of

1.330g/cm3. The photographs of sand grains are shown in Photo 1.

b) The relative density of specimens was about 70%, and the effective mean principal

stress was 1.0 kg/cm^ during cyclic shear. _.

III-136
2) Special conditions • ' "

a) The frequencies used were about 0 Hz (static test), 1 Hz, and 4 Hzj primarily 1 Hz.

b) Inherent apparatus characteristics were various as mentioned before, and

as detailed hereafter.

The Tests for Normally Consolidated Clay

1) Common conditions

a) The samples were prepared one at a time with identical marine clay (liquid

limit of 102%, plastic limit of 38%) being consolidated one-dimensionally under an

o overburden pressure of 1.0 kg/cm .

b) The degree of saturation of the specimens was about 99%, i.e., specimens

were almost totally saturated.

c) The specimens were all cylindrical or disk-shaped.

d) The specimens were reconsolidated for about 24 hours under a mean principal stress of 0.59 kg/cm prior to cylic shear. (The mean water contents after

reconsolidation were about 63%.

)

e) A confining pressure of 0.59 kg/cm^ equal to the reconsolidation stress was

maintained during shear.

f) The stress wave shape was sinusoidal; the frequency was 1 Hz except in

the resonant-column test.

g) During cyclic shear, 100 stress cycles were usually applied to the specimens.

2) Special conditions

a) The unconfined compressive strength on the prepared samples was scattered from

0.7 kg/cm^ to 1.0 kg/cm^. Therefore, some difference was produced in the strength

characteristics of block samples distributed to each facility.

b) The initial pore water pressure was usually set to zero, but cyclic shear tests

in which the initial pore water pressure was not zero were also performed.

c) The recognized differences in the inherent characteristics of the apparatus

are as follows:

III-137
1. In the resonant-column and the dynamic simple shear tests, stress conditioas

during the specimen reconsolidation are equal to KQ-consolidation stresses, which

were applied to the prepared block sample beforehand. While in the dynamic triaxial

compression tests, the specimen has to be consolidated under all-around pressure, in

order to be subjected to fully reversed cycles of shear stresses. Consequently,

consolidation stresses change from those under KQ-conditions to those under allaround pressure conditions. In such cases, the specimens may be disturbed to some

extent.

2. In the triaxial compression test, maximum shearing stress acts on a plane

inclined to the surface of the specimen by 45°, but not in the resonant-column

and the dynamic simple shear tests. Therefore, for anisotropic soils, shear

test characteristics may reflect the difference of the dynamic properties of

the soils.

3. In the dynamic simple shear apparatus, lateral stresses acting on the side

surface of the specimen cannot be measured. Therefore, it is necessary to

evaluate them with ^' obtained from drained shear test or consolidatedundrained shear test using Jaky's equation, which may allow some error

\ («S = 40.1° for sand, «5' = 38.5° for clay).

4. The methods to obtain equivalent shear modulus and damping ratio differ

according to the apparatus adopted. In the simple shear and the triaxial compression tests, they are directly determined with stress-strain curves. In the

triaxial compression test, however, it is necessary to measure or assume

Poisson's ratio. In the resonant-column test and the supersonic pulse test,

equivalent shear modulus is calculated indirectly in terms of resonant frequency and

S-wave velocity.

5. The strain state in the simple shear test, i.e. , the plane strain condi- j

tion, is different from the axi -symmetrical strain condition in the triaxial

compression test.

6. There are two types of dynamic triaxial compression apparatus. In one type,

lateral total stress is kept constant during cyclic shear; while in the other,

lateral total stress changes periodically with the same amplitude and a 180°

phase difference, comparing with axial total stress.

III-138
STRESS STRAIN CURVE AND DYNAMIC POISSON'S RATIO

OF DRY SAND AND ALMOST SATURATED CLAY

Shear stress-strain curves obtained from cyclic shear tests are shown in Figure 1

through Figure 3. Typical data of Toyoura sand obtained from ring torsion shear tests are

shown in Figure 1 (a) (b) (c) (d) according to shear strain levels (Hatanaka, 1976). The

clay test results in the simple cyclic shear tests and the dynamic triaxial compression

tests are also shown in Figure 2 (a) (b) (Hara and Kiyota) and in Figure 3 (a) (b) (Suzuki

and Sugimoto) , respectively.

It is shown in Figure 1 through Figure 3 that the shapes of the shear stress-strain

curves are very similar to each other, irrespective of soil types and test types. However,

it can be seen in Figure 1 that for sand, the loop in the second cycle is different from

the loop in the tenth and the 100th cycle. On the other hand, for clay, a similar tendency

is never seen as shown in Figure 3.

Dynamic Poisson's ratios were measured for dry sand and almost saturated clay. In the

supersonic pulse test, P-waves and S-waves were generated independently at the top of specimen, and their velocities were calculated by measuring the time when the waves reached the

bottom of the specimen. Dynamic Poisson's ratio vj is calculated with the following

equation (Sugimoto and Suzuki).

(V /V3)2 - 2 ^d =-—^ T (1)

2{(Vp/V,)2 - 1} . ;.„. , „^^, ,

, ,

where, Vp, Vg are velocities of P-wave and S-wave , respectively. Dynamic shear modulus is

also calculated with the following equation:

G = . - . ^ ^ - ^^^^

g

s ^

where, yj- is the wet unit weight of specimen, and g is acceleration of gravity. In the

triaxial compression test, dynamic Poisson's ratio is estimated by measurement of dynamic

lateral strain as well as dynamic axial strain. Lateral strains are estimated by measuring

the diameter change of specimen electrically (Suzuki, Sugimoto, Babasaki and Katita, 1976)

or measuring the volume change of specimen by measurement of the change of the water height

in the cell electrically (Yasuda and Sakai, 1976).

III-139
Figure 4 shows the Poisson's ratio of Toyoura sand. It can be seen in Figure 4 (a) that

dynamic Poisson's ratio is almost independent of initial void ratio and effective mean principal stress though the data are scattering. It is also recognized in Figure 4 (b) and (c)

that dynamic Poisson's ratio is almost independent of strain level. From the experimental

result mentioned above, dynamic Poisson's ratio may be estimated at about 0.35 for air-dried

Toyoura sand, and this value was used to calculate shear strain from axial strain.

Figure 5 shows the Poisson's ratio of almost saturated clay. The data scatter in the

range from .492 to .518, mainly .500 + 0.06. If the clay is saturated thoroughly, Poisson's

ratio becomes about 0.5. However, the value of 0.45 should be used as the dynamic Poisson's

ratio of the clay, because the prepared clay is not saturated in the strict sense and because

the data obtained with the electric diameter transducer are apt to be somewhat larger than

those obtained with the other methods as shown in Figure 4. The difference of dynamic

Poisson's ratio results in a small difference of shear modulus when the shear modulus G is

evaluated with elastic modulus E and Poisson's ratio v^j in using the following equation:

(3),

where ^

E = (4)

e

e is axial strain, and is deviator stress

EQUIVALENT SHEAR MODULI OF SAND AND CLAY : . ..^

Typical results of shear moduli obtained from the cyclic shear tests are shown in

Figure 6 through Figure 8. Figure 6 (a) (b) shows the relationship among shear modulus,

shear strain y (single amplitude) and number of stress cycles N for sand obtained from the

ring torsion shear test and the simple cyclic shear test, respectively. It is seen in both

figures that the shear modulus of sand increases as the shear strain decreases and as the

number of stress cycles increases. Especially, it can be noticed that the shear modulus

increases considerably with the number of stress cycles for the strain level more than

111-140
5 X 10" strain. This tendency is recognized most clearly from Figure 7 (a) (b), which shows

the increase of shear modulus of sand with the number of stress cycles for each strain level.

On the other hand, the results obtained for normally consolidated clay in the simple

cyclic shear test and the dynamic triaxial compression test are shown in Figure 8 (a) (b),

respectively. It can be seen that for clay, shear modulus decreases slightly with the number

of stress cycles, being different from the tendency for sand.

DAMPING RATIOS OF SAND AND CLAY

Typical results of damping ratios of sand and clay are shown in Figure 9 and Figure 10

(a) (b), respectively. Figure 9 shows the data obtained from the ring torsion shear test.

It can be seen from this figure that the damping ratio of dry sand decreases considerably as

the number of stress cycles increases. It is also recognized that the damping ratio

increases as the shear strain increases.

Figure 10 (a) shows the relationship between the damping ratio and the number of stress

cycles obtained in the simple cyclic shear test for clay. As shown in this figure, the damping ratio of normally consolidated clay decreases slightly as the number of stress cycles

increases

.

On the other hand, the damping ratio increases slightly as the number of stress cycles

increases in the dynamic triaxial compression test as shown in Figure 10 (b). This tendency

is contrary to the result obtained from the simple cyclic shear test. However, in either

case it may be found that number of stress cycles has little effect on the damping ratios of

clay, which is completely different from the case of dry sand.

COMPARISON OF INTER-LABORATORY TEST RESULTS

ON SHEAR MODULUS AND DAMPING RATIO ' " '

'

'

In order to compare test results under as similar conditions as possible, appropriate

measures were taken in connection with special test conditions. Concerning the special condition (a) mentioned before, the value of the shear modulus G is divided by the unconfined

compressive streagth q^^, i.e., G/q^ were compared directly, for there were some differences

among the block samples distributed to each facility. This step was based upon the assumption

that shear modulus was proportional to q^.

III-141
In regard to special condition (b), the data of damping ratio of the specimen with

non-zero initial pore water pressure were used; those of shear modulus were not used. This

is because the supplementary test results indicated that the initial pore water pressure had

little effect on damping ratio. However, the shear modulus of the clay with non-zero initial pore pressure was smaller than that with initial pore pressure under the same effective

confining stress (Suzuki et al . , 1976).

The results obtained from the inter-laboratory cooperative cyclic shear tests are shown

in Figure 11 through Figure 14 with respect to shear modulus and damping ratio against single

amplitude of shear strain.

The Results of Shear Modulus

It is recognized from Figure 11 and Figure 12 that the results of shear modulus can be

roughly classified into two groups. Especially for sand, the test results of simple shear,

ring torsion, torsion shear, resonant-column and supersonic pulse belong to one group, which

almost coincides with Shibata' s equation (Shibata and Soelarno, 1975). Dynamic triaxial compression test results belong to the other group, which coincides with the curve presented by

Seed and Idriss (1970) Shibata' s equation was drawn in Figure 11 by assuming the shear modulus obtained from supersonic pulse test as the initial shear modulus, i.e., shear modulus at

very small shear strains.

The difference of the test results of these two groups is caused by the inherent

characteristics of apparatus and test procedures, because the influence of the special

conditions (a) (b) on the test results were already eliminated.

Cornforth pointed out that the angle of internal friction ^ of sand obtained from

static shear tests under plane strain conditions was larger than that obtained from triaxial

compression tests (Cornforth, 1964). While, according to the inter-laboratory cooperative

test results, shear modulus obtained with the apparatus where shear stress is applied to the

surface of the specimen under nearly plane strain condition, is larger than that obtained

with the triaxial compression apparatus. This result shows a similar tendency of the results

obtained by Cornforth. Consequently, it may be supposed that the difference of the test

results is based on the difference of strain state, i.e., whether axi- symmetrical strain

system or nearly plane strain system of the test.

111-142
Additionally, attention should be paid to the fact that, in the trtaKlal compression

test for clay, close agreement is obtained between shear modulus measured under constant

cell pressure and that measured under cell pressure which was changed periodically during

shear (Tamaoki, 1976, 1977). It can also be noticed, that the measured values by several

triaxial compression apparatus almost agree for both sand and clay.

The Results of Damping Ratio

The results of damping ratio are shown in Figure 13 and Figure 14 plotted versus

shear strain. It can be seen from Figure 13 that the damping ratio of sand agrees well,

irrespective of the test types, with respect to the values in the tenth cycle. For

normally consolidated clay, however, the data are distributed over a wide range as shown

in Figure 14 and classification cannot be conducted clearly. It is also recognized that

the result obtained from the simple cyclic shear tests differs from that obtained by the

resonant-column test, and that there is some differences among the results obtained with

the several triaxial compression apparatus. However, the data are distributed within the

range presented by Seed and Idriss (1970) as shown in Figure 14.

It should be noticed that for clay, there is no significant difference with respect to

shear modulus and damping ratio between the results of the fresh specimen and those of the

stage tested specimens mentioned by Silver and Park (1975).

A supplementary test was performed for the clay specimen which was subjected to cyclic

stresses of larger amplitude than that intended at the time (Suzuki et al . , 1976). The

result indicates that shear modulus becomes smaller and damping ratio becomes larger than

that of the fresh specimen, respectively at the same shear strain level.

CONCLUSIONS

The following conclusions may be made on the basis of the inter-laboratory cooperative

cyclic shear tests on dry sand and normally consolidated clay. '

.

1) The results for Toyoura sand, the relative density of which was about 70% under an

effective mean principal stress of 1.0 kg/cm^, are as follows:

a) Concerning the relationship between shear modulus and shear strain in the

tenth cycle, the results were divided into two groups. The first group included

the results of simple shear, ring torsion, torsion shear, resonant-column and

III-143
supersonic pulse test, which almost coincided with Shibata's equation. The second

group was composed of the results from dynamic triaxial compression type tests,

which coincided with the curve shown by Seed and Idriss.

b) Concerning the relationship between damping ratio and shear strain in the

tenth cycle, the results agreed well with each other regardless of test types and

procedures.

c) The number of stress cycles had significant effects on shear modulus and

damping ratio of dry sand at large strain levels. .

d) Dynamic Poisson's ratio can be measured electrically during cyclic shear -. .

,

testing.

2) The results for artificial, normally consolidated clay, the water content of which

was about 63% under the effective mean principal stress of 0.59 kg/cm , are as follows:

a) Concerning the relationship between shear modulus and shear strain, the .

'

results were similar to those described for sand. Shear moduli obtained from the

tests of simple shear, resonant-column and supersonic pulse tests were larger than

those obtained from the cyclic triaxial compression tests at the same shear strain

level . -jfiii, • • ..

b) Concerning the relationship between damping ratio and shear strain, the

results were scattered in a wide range, being different from those for sand. The

results were distributed, however, within the range presented by Seed and Idriss.

c) The number of stress cycles had little effect on shear modulus and the

damping ratio of clay. -'i -^j

d) No significant difference was recognized between the results of the fresh

specimen and those of the stage tested specimen.

e) In the triaxial compression test, close agreement was obtained between , "

shear modulus measured under constant cell pressure and that measured under cell

pressure which was changed periodically during shear. v

ACKNOWLEDGMENTS

The authors are grateful to Professor Yoshiaki Yoshimi of Tokyo Institute of Technology

for his guidance to "The Committee on Dynamic Properties of Soils," and to the members of

the Committee for their cooperation.

I I 1-144
The work described in this paper was partially supported by the research funds of the

Ministry of Construction. ' '

;, NOTATION , .

= relative density • :

E = elastic modulus

e^ = initial void ration

G = equivalent shear modulus

g = acceleration of gravity

h = equivalent damping ratio

= coefficient of earth pressure at rest

N = number of stress cycles

= unconfined compressive strength

Vp = velocity of P-wave

Vg = velocity of S-wave .

Y = shear strain

Y^- = wet unit weight

e = axial strain

= dynamic poisson's ratio

' = angle of internal friction in terms of effective stress

= deviator stress

OjjjQ = effective mean principal stress

T = shear stress , , . .. ;

REFERENCES

[1] Cornforth, D. H. , "Some Experiments on the Influence of Strain Conditions on the

Strength of Sand," Geotechnique Vol. XIV, No. 2, pp. 143-167, 1964.

[2] Finn, W. D. L. , Pickering, D. J. and Bransby, P. L. , "Sand Liquefaction in Triaxial and

Simple Shear Tests," Journal of the Soil Mechanics and Foundations Division, ASCE , Vol.

97, No. SM4, Proc. Paper 8039, pp. 639-659, 1971.

[3] Hanai, T. , "Dynamic Behavior of Soil Obtained in Dynamic Triaxial Test, Reports of

Committee on Dynamic Properties of Soils," Building Research Institute, Ministry of

Construction (in Japanese), 1976, 1977.

III-145
[4] Hara, A. and Kiyota, Y., "Dynamic Behavior of Sand at Very Small to Large Shear

Strain," Proceeding of the 11th Annual Meeting of Japanese Society of Soil Mechanics

and Foundation Engineering, pp. 331-334 (in Japanese), 1976.

[5] Hara, A. and Kiyota, Y. , "Dynamic Shear Tests of Soils for Seismic Analyses,"

Proceedings of the Ninth International Conference on Soil Mechanics and Foundation

Engineering, Tokyo, Vol. 2, pp. 247-250, 1977.

[6] Hatanaka, M. , "Dynamic Behavior of Sand Obtained in Ring Torsion Shear Test, Reports

of Committee on Dynamic Properties of Soils," Building Research Institute, Ministry

of Construction (in Japanese), (1975, 1976).

[7] Iwasaki, R. , "Seismic Response Analyses of Ground by Using Data Obtained in Interlaboratory Cooperative Tests, Reports of Committee on Dynamic Properties of Soils,"

Building Research Institute, Ministry of Construction (in Japanese), 1977.

[8] Iwasaki, T. , Tatsuoka, F. and Takagi, Y. , "Shear Moduli of Sands Under Cyclic Torsional

Shear Loading," Soils and Foundations, Vol. 18, No. 1, pp. 39-56, 1978.

[9] Koori, Y. , "Shear Moduli and Damping Ratios Measured by Hardin Oscillator," Reports of

Committee on Dynamic Properties of Soils, Building Research Institute, Minstry of

Construction (in Japanese), 1975, 1976, 1977.

[10] Peacock, W. H. and Seed, H. B. , "Sand Liquefaction Under Cyclic Loading - Simple Shear

Conditions," Journal of the Soil Mechanics and Foundations Division, ASCE, Vol. 94,

No. SM3, Proc. Paper 5957, pp. 689-708, 1968.

[11] Roscoe, K. H. , "An Apparatus for the Application of Simple Shear to Soil Samples,"

Proceedings of the Third International Conference on Soil Mechanics and Foundation

Engineering, Vol. 1, pp. 186-191, 1953.

[12] Seed, H. B. and Idriss, I. M., "Shear Moduli and Damping Factors for Dynamic Response

Analyses," Earthquake Engineering Research Center Report, No. EERC 70-10, University

of California - Berkeley, 1970.

[13] Shibata, T. and Soelarno, D. S., "Stress-Strain Characteristics of Sands Under Cyclic

Loading," Proceedings of the Japan Society of Civil Engineers, No. 239, pp. 57-65

(in Japanese), 1975.

[14] Silver, M. L. and Park, T. K. , "Testing Procedure Effects on Dynamic Soil Behavior,"

Journal of the Geotechnical Engineering Division, ASCE, Vol. 101, No. GTIO, Proc. Paper 11671, pp. 1061-1083, 1975.

[15] Sugimoto, M., Suzuki, Y. and Terada, K. , "Study on Dynamic Properties of Soils - Comparison of Laboratory Test Results with In-Situ Test Results," Proceeding of the

Tenth Annual Meeting of Japanese Society of Soil Mechanics and Foundation Engineering,

pp. 297-300 (in Japanese), 1975.

[16] Suzuki, Y. , Sugimoto, M., Babasaki, R. and Kakita, E., "Measurement of Poisson's Ratio

by Means of Dynamic Triaxial Apparatus," Proceeding of the Eleventh Annual Meeting of

Japanese Society of Soil Mechanics and Foundation Engineering, pp. 415-418 (in

Japanese), 1976.

[17] Tamaoki, K. , "Dynamic Behavior of Soil Obtained in Dynamic Triaxial Test with Lateral

Stress Changed," Reports of Committee on Dynamic Properties of Soils," Building

Research Institute, Ministry of Construction (in Japanese), (1976, 1977).

[18] Yasuda, S. and Sakai, U. , "Dynamic Behavior of Soil Obtained in Dynamic Triaxial Test,"

Proceeding of the Eleventh Annual Meeting of Japanese Society of Soil Mechanics and

Foundation Engineering, pp. 423-426 (in Japanese), 1976.

III-146
[19] Yoshimi, Y. and Oh-oka, H. , "A Ring Torsion Apparatus for Simple Shear Tests,"

Proceedings of the Eighth International Conference on Soil Mechanics and Foundation

Engineering, Moscow, pp. 501-506, 1973.

[20] Yoshimi, Y. et al. , "Reports of Committee on Dynamic Properties of Soils," Building

Research Institute, Ministry of Construction (in Japanese), 1975, 1976, 1977.

III-147
III-148
Fig. 1. Stress strain curves obtained from ring torsion shear test

(after Hatanaka)

III-149
Fig. 1. Stress strain curves obtained from ring torsion shear test

(after Hatanaka)

III-150
Fig. 2. Stress strain curves obtained from cyclic simple shear test

(after Hara and Kiyota)

III-151
0.4

0.3

c

o

CO *o

Q.

O

E

D

C

>» O

(a) Supersonic pulse

*

. G A

O G° ^ —A

0.2 -

Ql

AB

( Suzuki a Suglmoto )

o O.S ^^/cm"^

A T.O K5/cm2

1,5 Kt/cm2

Q7I 0.72 073 0.74 075 076

Initial void ratio , 60

077 0.78 079

05

0.4

0.3

Q2

0.1

(b) Triaxial

f = 1 Hz

oN« I Q N«IO

A N 100

0

0.001

%

O

O

00 O

A

-A"

Dr=70%

( Suzuki a Suglmoto ) 'I ' l l —I L 1 I lit.

0.005 0.01 0.05 0. Axial strain (single amplitude)

0.5

0.41-

0.3

0.2

(C) Triaxial

. f = 03Hz A-

-A—

Q Co = 0.71

0.1 H Ae<, = o.68 (Yasuda a Sakai )

O 60=0.65

Axial strain ( single amplitude)

Fig. 4 Dynamic Poisson's ratio of dry sand

-3

IH-152
0.51 h D

o

(/) o

Q.

E

o

c

O

0.52

. A

O50

0.49

(o A X + Mark of specimen

Triaxial

o

O A

oo

A^

+ 0+

o

A

4d —

^V?." V o "a

°

0 20 40 60 80

Number of cycles , N

100

Fig. 5. Dynamic poisson's ratio of saturated clay

(after Suzuki and Sugimoto)

111-153
1200

Shear strain , If (Single amplitude)

J 0.005 0.01 0.05 0.1 0.5 1.0 Shear strain (single amplitude), /(%)

Fig. 6. Shear modulus versus shear strain relationship

for dry sand (a) (after Hatanaka) , (b) (after Hara

and Kiyota)

III-154
CME

1000

0 702

0 731 0 71 7 0 7 35

0 740

v>

n 733

dul 500

o 0 726 E 0 715

0 760

k.o 0 731 0) n 739 SI 0 742 cn

I

,»..-• 0.700 (No. 5)

Initial Voi d Ra t i 0

0.730 ( 0.716 ( 0.735 ( 0.736 ( 0.732 ( 0.722 ( 0.727 ( 0.710 ( 0.72! ( 0.727 ( 0.721 (

No. 3) No . 2

)

No . 1 ) No. 4) Mo. 6) No. 8) No. 9) No. 7) No. 12) No . 1 1 ) No. 10)

Final Void Ratio

(Q) Ring torsion shear

10 100

Number of cycles , N

Fig. 7. Effect of number of stress cycles on shear

modulus for dry sand (a) (after Hatanaka)

,

(b) (after Hara and Kiyota)

III-155
<M6

>^

CP

o

E

o

a> sz(O

200

100

0

(a) Simple shear

/» 0.015 %

O O "O .Q-o-oc—

/= 0.038 %

-a-o-o-ooo-

/= 0. 25 %

1.16 %

-1 I I I I I I '

No. I

No. 2

'No. 3

>No.4

J I 5 10 50 100

Number of cycles , N

Number of cycles , N

Fig. 8. Effect of number of stress cycles on shear

modulus for normally consolidated clay

(a) (after Hara and Kiyota) , (b) (after Suzuki

and Sugimoto)

III-156
Damping

ratio

h (%)

40

30 -

20 -

10

10

a =1 . 70 kq/cm

e =0.715-0.760

o Cycle 2

• Cycle 10

© Cycle 100

Ring torsion shear

oo

8

9

( Seed 1970 )

0^=1.0-1.5 kg/cm

10

-3 10 10

Sliear strain,/" (Single amplitude)

Fig. 9. Damping ratio versus shear strain relationship for dry sand

(after Hatanaka)

111-157
20

.9

*—o

c

Q.

E

o

10

No. 4

(0) Simple shear

I I I I 5 10

Number of

1.16 %

/=0.25%

/= 0.038%

/= 0.015 %

I 50

!es , N

100

Fig. 10. Effect of number of stress cycles on damping

ratio for normally consolidated clay

(a) (after Hara and Kiyota) , (b) (after Suzuki and

Sugimoto)

III-158
Fig. 11. Shear moduli versus shear strain relationship obtained

from inter- laboratory cooperative test for dry sand

1250 w

3

"OO

E

o

"5 ^ 750

«- o

c ^

5 500

cro

u

E

o

c

>»

250

, ^ ^^

Shibata's equation (1975)

"^r^.-^ ^>c'5imple shear

Super

'

' ^s;. sonic ^Sv '^'\rR"^9 torsion

pulse shear

Seed (1970)

Torsional shear

Toyoura sand

6mo rlO kg /cm*

Dr=70 7o

10th cycle

Triaxiai ^

compression(lHz)

Static triaxiai

'"^>xxC9mpression

Q 2x10"'* 10"* 10"' 10'^ 5XI0"2

Shear strain (single amplitude),

III-159
Fig. 12. Shear moduli versus shear strain relationship obtained

from inter-laboratory cooperative test for normally

consolidated clay

250

200^

150

00

50

Super sonic pulse

Resonant column

Simple shear

Triaxial comprssion

(cell pres. = const.)

Normally consolidated clay

Qu = 0.7 ~ 1.0 kg/cm^

(5mo= 0.59 kg/cm^

Zero back pressure

/'Triaxial comprssion

*<>>x (cell pres. i; const.)

^-2 2X10"^ lO""* 10'' 10"

Shear strain (single amplitude),/'

III-160
30

01 . ._ 2x 10'^ 10"* 10"^ 10'^

Shear strain (single amplitude ),

Fig. 13. Damping ratios versus shear strain relationship obtained

from inter-laboratory cooperative test for dry sand

III-161
Fig. 14. Damping ratios versus shear strain relationship obtained

from inter- laboratory cooperative test for normally

consolidated clay

30

o

o

c

Q.

E

o

o

20

Resonant column

Simple shear

(cell pres. Triaxial = const.

)

compression

-

Seed 8. Idriss

OL

2x

Normally

consolidated

clay

6mo = 0.59Kg/cm2

10r5 \-4 10

-3

Shear strain ( single amplitude ) . ^

III-162
WIND AND STRUCTURE MOTION STUDY FOR PASCO-KENNEWICK BRIDGE

M. C. C. Bampton

Battelle Pacific Northwest Laboratories, Richland, Washington

Harold Bosch, David H. Cheng, and Charles F. Scheffey

Federal Highway Administration

'ii. : Washington, B.C.

ABSTRACT

A program to study the responses to the natural wind of the 2507 ft (763 m) PascoKennewick Bridge is described using an automatic data collection system. The study consists

of three phases: preliminary investigation, data collection and data analysis. The study is

sponsored by the Federal Highway Administration and will last two years.

KEYWORDS: Bridges, aerodynamic response, bridges; cable-stayed bridges.

IV-1
INTRODUCTION

Traditionally wind tunael tests on scaled section models have been used to evaluate the

aerodynamic stability of long-span bridge designs. This procedure is generally considered to

provide conservative (safe) results. However, there are still two areas of uncertainty to

wind tunnel tests and the interpretation of their results. The first of these is related to

differences in turbulence characteristics between the real world and the wind tunnel and the

effect of these differences on critical wind velocities for the initiation of unstable structural aeroelastic oscillations. The second area of uncertainty is related to the amount of

energy the structure absorbs from the aerodynamic forces imposed by the wind.

To provide prototype information on atmospheric turbulence and the actual response of a

bridge to the turbulence, the Federal Highway Administration is sponsoring a study of wind

induced motions of the new, cable-stayed bridge between Pasco and Kennewick, Washington. The

specific objective of the study is to determine the responses of an existing, inland, longspan, cable-stayed bridge to the natural wind. This will be accomplished by making detailed

measurements of the wind structure and bridge motions during a two year period. The contract

for this study, which has been awarded to Battelle Pacific Northwest Laboratories in

Richland, Washington, runs through March 1981.

PASCO-KENNEWICK CABLE-STAYED BRIDGE

The Pasco-Kennewick Cable-Stayed Bridge was opened to traffic September 16, 1978. It is

currently the longest span, cable-stayed bridge in the United States. The total length of

the bridge is 2503 ft (763 m) and its center span covers 981 ft (259 m) . The width of the

bridge is 81 ft (25 m) . At the middle of the center span there is a 60 ft (18 ra) vertical

clearance between the bridge and the river. The concrete towers from which the cables are

hung extend 244 ft (74 m) above the river surface.

J: Pasco, Kennewick and Richland, collectively referred to as the Tri-Cities, are situated

on the Columbia River in South-Central Washington State, just north of the Washington-Oregon

border. This part of the State has a semi-arid climate. The prevailing winds in the vicinity of the bridge site are either southwesterly, or northeasterly, and have an average speed

of 7.9 mph (3.5 m/s). Approximately 4 percent (approximately 342 hrs/vr'i of the time the

IV-2
speed exceeds 32 mph (14 m/s). At 50 feet above the ground a gust exceeding 60 mph is

expected about once every two years. Higher gusts are expected at lesser frequencies.

In the reach of the river between Pasco and Kennewick, the Columbia flows from the westnorthwest to east-southeast. The new, cable-stayed bridge is oriented in the south-north

direction. Therefore, the prevailing winds and the vast majority of high winds will approach

the bridge at an angle from its upriver side. The upwind fetch in this direction is relatively unobstructed with a surface roughness typical of residential and small commercial

buildings. There are no major structures to affect the turbulence. -•

On the downriver side of the cable-stayed bridge there is an old (completed in 1922)

cantilevered-truss bridge which the new bridge is supposed to replace. The two bridges are

parallel and the separation between the two bridges is approximately 60 ft (18 m) . The

effect of the close proximity of these two structures on the wind and the response of the new

bridge to it are unknown. There are plans for an eventual demolition of the old bridge, but

no date has been set. There are no other structures or topographic features that would cause

atypical winds or potentially affect the structural response of the new bridge.

RESEARCH PROGRAM

The research to be conducted by Battelle, under the aegis of the Federal Highway Administration, generally falls into three areas: preliminary investigations, data collection,

and data analysis.

Preliminary investigations include: review of previous research on wind induced structural motions with particular emphasis on motions of long-span bridges; study of the reports

of wind tunnel tests of a scaled section model of the Pasco-Kennewick Cable-Stayed Bridge;

qualitative evaluation of air flow patterns at various locations along the bridge to determine if there is an interaction between the cable-stayed and the old bridge; and evaluation

of the adequacy of the instrument system available for use in the measurement or data

collection phase of the study. The results of the preliminary investigations will be used

to determine specifics of the data collection and analysis phases of the research, such as

instrument locations and sampling rates.

The data collection phase of the research includes: physical check-out and calibration

of the instrument system provided by the Federal Highway Administration; selection of optimum

locations for installation of the instruments; preparation of appropriate mounting brackets

IV-3
and installation of the instruments and collection of wind and bridge motion data- Many of

the details of the data collection program are yet to be determined including sampling rates

criteria for initiation of sampling, and the duration of individual sampling periods. These

details will initially be determined on the basis of results of the preliminary investigations and will be changed if results of data analysis indicate changes are appropriate.

The primary wind instruments for this study are 3-dimensional propeller anemometers,

commonly referred to as Gill anemometers. They provide a nice compromise between the rugged

instruments required for high wind speeds on the one hand and the response sensitivity

required for turbulence measurements on the other. The three propeller configuration permit

full representation of the wind vector and their 3 ft (1 m) distance constant permits resolu

tion of turbulent eddies that are small compared to the major structural components of the

bridge.

The primary instruments to be used in measurement of bridge motions are paired, servotype accelerometers , manufactured by Terra Tech, Inc. The accelerometers will be attached t

the deck structures. The acceleroraeter signals are to be filtered to remove the effects of

extraneous high-frequency motions that might be generated by vehicular traffic.

Anemometer and accelerometer signals will be recorded automatically on magnetic tape by

a digital recording system using predetermined signal levels to Initiate recording. The

instrument system will be checked bi-weekly by Battelle staff to determine its operational

status

.

Data analysis will be conducted in two steps. The Initial step will be a screening

analysis to determine the general wind and bridge motion characteristics during each measure

ment period. The second step will be a detailed analysis to determine the specific characteristics of the wind turbulence structure and bridge motions. All data collected during

the study will be given a screening analysis; the results of the screening analysis will be

used to identify data sets to be examined more closely in detailed analysis. '

The screening analysis of the wind data will include: computation of mean wind speeds,

directions and vertical angles, and characterization of the variability of the wind by computation of root-mean-square and maximum departures from mean values. Screening analysis of

the bridge motion data will include computation of root-mean-square and maximum

accelerations. . .

• . :

:

IV-4
The detailed analysis will include examination of frequency distributions for the wind

velocity components and bridge accelerations, and spectral analysis of the turbulence and

motion data. The temporal characteristics of the wind will be determined for each anemometer

location. Power spectra will be computed to determine the distribution of wind energy in a

frequency domain. Spatial characteristics of the wind will be examined using data from pairs

of anemometers. The response of the bridge structure to the wind will be determined through

spectral analysis of the accelerometer data. Bending and torsional mode shapes and frequencies will be determined for the bridge. The damping characteristics of the bridge will also

be estimated.

The results of the screening and detailed analyses of the wind and bridge motion data

will be combined to estimate the critical wind conditotis that initiate structural aeroelastic

oscillations. These estimates, along with a general discussion of the interpretation of the

data, and a description of the procedures by which data collection and analyses are performed, will be presented in the final report at the end of the study. The nuraercial results

of the screening and detailed analyses will be included in the final report as appendices.

In addition, the output of the study will include the original data tapes and documented

computer programs used in the data analyses. .

.

SUMMARY

In summary, a program to monitor the dynamic responses of the 2507 ft (763 m) PascoKennewick Bridge to the natural wind is described. This research is only a part of an overall field monitoring program nationwide sponsored by the Federal Highway Administration for

the purposes of verifying the recent analytical as well as the section model test results,

and of increasing the database needed for delineating the characteristics of the natural

wind

.

IV-5
CALCULATION OF THE GUST RESPONSE OF LONG-SPAN BRIDGES '

Nobuyuki Narita . '. "

Hiroshi Sato - •

Structure Division .;

Structure and Bridge Department

Public Works Research Institute

Ministry of Construction '

- •• ' >

ABSTRACT '

Discussed in this paper are the random responses of long-span bridges due to the turbulence of natural wind, which are important to the fatigue problem of structural materials and

to the serviceability of the bridge for the public. Both the vertical bending and the torsional modes of vibration are considered in performing the calculations, the analytical portion of which is described in a frequency domain and in a time domain as well.

The calculated vibrations are compared with those obtained by field observations conducted at two bridges in order to insure the validity of the method. The calculated results

agreed reasonably with the observed ones. It is found that, by setting more rational values

for the parameters, gust responses of long-span bridges can be predicted more precisely by

the use of the method described.

KEYWORDS: Aerodynamics of bridges; bridge stability; bridge vibration; gust responses; longspan bridges.

IV-6
INTRODUCTION

It is necessary to insure the aerodynamic stability and the safety of long-span bridges

against strong winds right from the early stage of their design. The dynamic phenomena of

structures due to wind include self -excited vibrations, vortex-excited vibrations, and

buffeting. For the first two phenomena, stability problems can generally be confirmed by

means of wind tunnel tests with section models under uniform smooth flow.

Through the tests the critical wind speed for self -excited vibration shall be increased

high enough to estimate the structural safety of the bridge, and the vibrational amplitude

induced by vortex-shedding shall be decreased small enough to determine the serviceability

of the bridge for the public by changing the geometrical configuration of the deck, the handrail, and the curb.

The real structures, however, are three dimensional and are exposed to the turbulent

natural wind. Then, the relationship between the aerodynamic behaviors of structures

observed in the wind tunnel tests and those of real structures under gusty wind should be

examined thoroughly. Note a few things have been left unclear till now about the turbulence

effects of the aerodynamic behavior of bridges. The followings are qualitatively recognized

through the recent boundary layer wind tunnel tests [1, 2, 3,] on the aero-elastic fullmodels :

1) The critical wind speed of flutter increases in the turbulent flow,

2) The vibrational amplitude of vortex-excitation is suppressed in the turbulent flow,

and

3) The intensity of buffeting vibration increases in the turbulent flow.

The effects 1) and 2) show that the dynamic phenomena obtained from the wind tunnel

tests in a uniform smooth flow are on the safe side, but the effect 3) requires further

investigation of buffeting, whose characteristics cannot be measured correctly in wind tunnel

tests under the smooth flow.

Natural wind data and bridge response data at four specific bridges (Kanmon Br. , Suchiro

Br., Hirado Br., and Suigo Br.) have been collected and processes at the Structure Division

of the Public Works Research Institute in order to insure the validity of the wind resistant

design applied to the bridge and to establish a more rational wind resistant design method

for the buffeting problem. [4] On the other hand some computer programs for calculating the

IV-7
buffeting of long-span bridges have also been developed at the Structure Division, Public

Works Research Institute. [5] In the first part of this paper the calculation methods for

vertical bending vibration and the torsional one are described. In the second part these

methods are applied to the real long-span bridges in Japan, and the results are compared to

the observed responses in the last part.

NOTATION . • :

• v , .

A Indicial admittance

A^ Aerodynamic admittance

Width of bridge model

Bf Width of full-scale bridge

Cq Drag coefficient

C-^ Lift coefficient

Cj^j Aerodynamic moment coefficient ^

Cf Cl + aCp .

C Unsteady aerodynamic coefficient

ex. Ct Lift in phase with vertical bending displacemer

'-'r] - [ :

zJ, Moment in phase with torsional velocity

Y ,

' - ' D Drag

f Frequency

h Impulse response

I Polar moment of inertia per unit length of spair

L Lift

1 Span length of the bridge

.1 . -(/.. . M Aerodynamic moment

m Mass per unit length of span

p External force per unit length of span

q Generalized coordinate

r Order of vibration mode

R Cross-correlation coefficient of wind

S Power or cross spectral density

t Time

IV-8
u

u

w

X

1/Z

a

S

n

<l>

u

p

$

(1)

Mean horizontal component of wind speed perpendicular to the bridge axis

Fluctuating horizontal component of wind speed perpendicular to the bridge

axis

Fluctuating value of vertical component of wind speed

Spanwise coordinate of bridge

Admittance

Angle of attack

Structural damping constant

Vertical bending displacement

Torsional displacement

Amplitude of forced vibration during the measurement of unsteady aerodynamic coefficients

Contribution factor of lift to moment

Air density

Vibration mode

Circular frequency

Natural circular frequency of r-th order mode

Expression of Fourier transform

; of motion can be expressed as follows when vertical bending vibration and

are considered;

_d_ dt

Complex conjugate

CALCULATION METHOD

FUNDAMENTAL EQUATIONS

(1)

/m$2 dx

nr

q *r + 0)^ q nr nr

(2)

(j)r

IV-9
where n(x, t) = Zq^j.(t)$^j.(x) (3)'

r

*(x, t) = Zq^j.(t)$^j.(x) ' (4)

r

Pl(x, t) = Pl^(x, t) + Pl^(x, t) + Pll(^'

Pj^(x, t) = PMn(x. t) + PM(t,(x, t) + Pj^(x, t) (6)

In the above equations, the self -excited forces, p-j^^ p^^^ pj^^ and pj^^ are

assumed to be

Pii (x, t) = / j(x, t) —(t - T)dT (7)-

where i = L, M (8)

j = n, * (9)

dAij(t) ^ 1

dt 2ti Zij(w)

do)

IpU^CClRCuj) + iCLl(a)))

ipU'Bf(CMR(a)) + iCmICw))

2 n

^ ipU'Bf(CLR(w) + iCLl(a)))

ipU^B|(%R(a)) + iC^lCo)))

2 <t> ct>

(10)

(11)

(12)

(13)

(14)

In Eqs. 11 through 14, the unsteady aerodynamic forces [6] can be measured by the forced

vibration methods and are assumed to be linear to the amplitude of the motion.

The aerodynamic forces due to the turbulence of natural wind p^^j^ and pj^ are assumed to

be as follows:

IV- 10
PLl(x, t) = j° ^ "(5^. t)hu(t - T)dT

+J

^ W(X, t)h^(t - T)dT

p^(x, t) = yBfPLL(x, t) [7]

(15)

(16)

The impulse responses and can be obtained by inverse transform of aerodynamic

admittance Ad. Some of the aerodynamic admittances are given as follows:

1) To relate u to p-^^

o Adu(k) = |pBfU^ . aT(k) [8] ; , (^7)

f 'If: where T(k.) is Horlock. s function [9], and k is tt • —

o lAd^COr = (PBfUCp)'x'(C) [10] (Ig)

where X^(S) = —|-^[75-l + e "^ij'

(^9)

considering the correlation both of cherdwise direction and of

vertical direction,

i-rfZC-l + e (Tcp^''^"" J (20)

fBf considering the correlation of vertical direction, and C = ——

U

In the equations 19 and 20, the decay factor of cross-correlation

coefficient is assumed to be 7.0.

2) To relate w to p-^^

1

^ y'BfU— T^^y l^^J (21)

o Ad^(k) = iPBfU^ <Kk) [11]

where 4)(k) is Sear's function, and k = it f *BfU

IV- 11
The approximations of (t)(k) are as follows;

a + (ira + l)k + Zttk'

1 + 2irk

o |Ad„(Or = (ipSfU*^ X^(C) [10] . (24)

where X^(C) is the same as equations 19 and 20. ''

FUNDAMENTAL EQUATIONS IN A MATRIX FORM

When equation 3 - equation 10, equation 15, and equation 16 are substituted into

equation 1 and equation 2, they can be rewritten as equation 28 and equation 29 using

transform and assuming the following orthogonality.

/*' $„<,dx < = 0(r s) . (25)

Jo f 0(r = s)

^..dx < = 0(r s)

J 0 *^ = 0(r = s)

/„ %r %sdx ?^ 0 (27)

(26)

0(r = s)

+ 25nr%r<i-) + <r )%r(-)

i J— q^r('^)/'' %r (x)dx + ^q^iMf' $n(^)*rir('^)d>

„u..2^ dx[zL,(u)) J' Zl^(co) -'^

+y*^^{u(x, a))Ad^j(a)) + w(x, u))Ad^^(to) } (28)

IV- 12
_i_ Zq i(aj)/^ $^i(x).r(x)dx + __L„q^r ('^> /

^ '^r ('<)dx

+ \iBfJ'\a(yi, u)Ady(a)) + w(x, u))Ad„(cj) }<f^^(x)dx (29)

Then, the relation between q^j., q^^ and L^,, M^., are obtained as follows.

(w)q (cj) ^

ZX^ (w)q .(cj) ^ nr

nr ^ (t)ri L (u)) r (30)

where

% (a))q (oj) +

ZX^ _(a))q^.(u)) ^ ^(^)

1 nr 1

Xl M = (-03^ + 2? a, (iuj) + (.2^ ) - —i . — (X

nr rn,$2 dx ZL(a>)/

Jo ri 0

(31)

)dx (32)

^(t)ri /Jm4'|j- dx (w) / $^i(x)$,r(x)dx

(33)

%<t)r(") =

+2?^j.aj^r(ia)) + (d^j. > - j.;i%r dx Zm

1 $f dx (34)

%nri(")

;Jl*^r d^ (o)) (35)

Lr(a)) = /i^i^^r dx

{u(x, cj)Ad^(a)) + w(x, oj)Ad„(a)) I't'^r^^^dx (36)

Mj-(a)) = /o"|r dx

yBf/ {u(x, u))Adu(u) +w(x, oj)Ad„((x)) }$^j.(^)dx (37)

lV-13
Equations 30 and 31 can be written in a matrix form as follows:

'Xi. Xi..

Xl.,2

Xi.ok Xi.«;ki Xi «>kn

Xm¥'2

(I

q^k Lk

M,

(38)

0

Xwiink , q«>n

.

Mn,

CALCULATION METHOD IN A FREQUENCY DOMAIN

The power spectral density of q^^. and q^j. can be calculated from the inverse matrix

in Equation 38, from the complex conjugate of the inverse matrix, and from the power and

cross spectral densities of L^ and M^..

One example of the calculations of cross spectral densities of L^ and M^ is as follows

^^r • ^s

St M

= lim-£ ^

pBf fl l(x, w)*.^(x)dx • fl I(x, u))*.j.(x)dx = lim I I T-» T

'J^

m^^j. dxili^j. dx

where

uBf /o So Sl(x, y; a))*^^(x)*^j.(Y)dxdy

(39)

l(x, 0)) = u(x, a))Ady((o) + w(x, (D)Adw(u)) ^^q^

Sl(x. y; 0)) = iim.U*. u)) l(x, 0)) ^^^j

IV- 14
The cross spectral density St(x, y; u) can be calculated from the power spectral

density of fluctuating lift at reference point Sj^^jCaj), using the following assumption [7].

Sl(x, y; 0)) = Sl(,(u)R(x, y; u) (42)

where

l(x* a))t(xo (jo) (43) T

The power spectral densities of n and (j) can be calculated from those of q^^ and q^j-,

using Equations 3 and 4. Then the gust responses of the vertical bending mode and the torsional one can be calculated in a frequency domain from the wind data.

CALCULATION METHOD IN A TIME DOMAIN

When the gust responses are calculated without neglecting the effect of coupling different modes to the self-excited forces, it is quite difficult to obtain the impulse responses

analytically which relate L^. and Mj. to q^j. and q^j--

In this paper, the impulse responses are obtained numerically from the inverse transform

of the inversion of the matrix in Equation 38.

On the other hand, the aerodynamic forces due to the turbulence of natural wind can be

calculated from Equations 15 and 16. q^^j. and q^^ can be obtained from the convolution of

the impulse responses and the aerodynamic force. Then the gust responses of the vertical

bending mode and torsional one can be calculated in a time domain from Equations 3 and 4.

These are some of the assumptions and parameters in the calculation process of the gust

responses. Comparisons between the calculated results and observed ones are required in

order to insure the validity of the assumptions and to set the proper value for the

parameters. This study is just proceeding at Structure Division, Public Works Research

Institute. In this paper, some examples of the calculated gust responses are described.

The calculation methods in a frequency domain and in a time domain are applied to a cablestayed girder bridge and to a suspension bridge, respectively.

SOME EXAMPLES OF CALCULATED GUST RESPONSES

IV- 15
CALCULATION IN A FREQUENCY DOMAIN

The gust responses of the Suehiro Bridge were calculated in a frequency domain. The

Suehiro Bridge, constructed in 1976, is a cable-stayed girder bridge with the span of 110 +

250 + 110 m.

The gust responses were calculated with regard to the standard deviation at the center

of the main girder by using observed wind data and compared with observed gust responses.

Twelve sets of observed data were chosen in which the mean angle of attack ranges from - 1°,

to 1°, and the mean horizontal wind speed perpendicular to the bridge axis ranges from 7 m/s

to 14 m/s and the intensity of turbulence ranges from 0.16 to 0.25. The values of the

parameters were determined as in Table 1 by considering aerostatic and aerodynamic characteristics obtained from wind tunnel testing in a uniform smooth flow. The calculated gust

responses are shown in Figures 1 and 2, and they are compared with the observed ones. From

the results, it is concluded that the gust responses in the vertical bending mode were predicted well, but as for torsional mode, the parameters such as the contribution factor of

lift to moment should be evaluated more precisely. 'v , '^v

CALCULATION IN A TIME DOMAIN

The gust responses of the Kanmon Bridge were calculated in a time domain. The Kanmon

Bridge, constructed in 1974, is a suspension bridge with the span of 178 + 712 + 178 m.

The mean horizontal wind speed perpendicular to the bridge axis used in the calculation

was about 20 m/s. In order to consider the effect of spanwise correlation, some wind data

series whose correlations to one another are nearly zero were made to act on each segment

of the bridge deck. [14] These data series were made from one series by providing a large

lag.

The mean wind speed was calculated for each segment of wind data series, to consider

the effect of its increase and decrease. The values of parameters are shown in Table 2.

The gust responses were calculated for the five cases in Table 3, and they were compared with

the observed ones in the same table regarding the standard deviation at the center span.

Although more investigations are required concerning the parameters, the calculated results

agree to some extent with the observed ones. The prediction of gust responses in a time

domain is expected to become more accurate by evaluating the parameters more precisely.

IV-16
CONCLUSION

The gust responses of a long-span bridge in the vertical bending mode and in the torsional one can be calculated using the foregoing process, and the more precise evaluation of

the described parameters will make the prediction of gust responses more accurate. To further this point of view, the following improvements are required.

1) As for parameters whose more rational values can be obtained in a theoretical way or

in an experimental way, use these values. .. <

2) As for the parameters whose more rational values cannot be possibly obtained at present time, get the values to fit the observed responses of the bridge, and use them.

ACKNOWLEDGMENT

The authors are indebted to the Japan Highway Corporation and to the Tokushima Prefecture for obtaining the valuable wind data and the responses at the Kanraon Bridge and the

Suehiro Bridge, respectively. The authors also wish thanks to Dr. N. Uehara of the Nikken

Consultants, Inc. and to Mr. S. Yuzawa from the Structure Division, Public Works Research

Institute, for fine work in programming the calculation and in operating the computer

respectively. i ; ' .

REFERENCES

[1] Davenport, A. G. , Isymov, N. , and Miyata, T. , "The Experimental Determination of the

Response of Suspension Bridges to Turbulent Wind," Proceedings of Third International

Conference on Wind Effects on Buildings and Statistics, 1971.

[2] Melbourne, W. H. , "West Gate Bridge Wind Tunnel Tests," August 1973.

[3] Irwin, H. P. A. H. , Schuyler, G. D. , "Experiments on a Full Aeroelastic Model of

Lions' Gate Bridge in Smooth and Turbulent Flow," Laboratory Technical Report,

National Technical Report, National Research Council Canada, October 1977.

[4] Okubo, T. , Narita, N. , Yokoyama, K. , "On the Wind Response of the Kanmon Bridge,"

7th Joint Meeting, UJNR, Tokyo, May 1975.

[5] Narita, N. , Yokoyma, K. , "On the Gust Response of Long-span Suspension Bridges,"

6th Joint Meeting, UJNR, Washington, D.C., May 1974.

[6] Okubo, T. , Narita, N. , and Yokoyama, K., "Some Approaches for Improving Wind Stability

of Cable-Stayed Girder Bridges, Proceedings of the 4th International Conference on Wind Effects of Buildings and Statistics, 1975.

[7] Konishi, I., Shiraishi, N. , and Matsumoto, M. , "Studies on Wind Resistant Design of

Long-Span Suspension Bridges, Fac. Engineering, University of Kyoto, 1973 (in

Japanese )

.

IV- 17
[8] Shiraishi, N. , and Matsumoto, M., "Studies on Gust Responses of Long-Span Suspension

Bridges, Fac. Eng., University of Kyoto, 1976 (in Japanese).

[9] Horlock, J. H. , "Fluctuating Lift Forces on Aerofoils Moving Through Transverse and

Chordwise Gusts, Trans. ASME, Journal of Basic Engineering, December 1968.

[10] Daveneport, A. G. , "Buffeting of Suspension Bridge by Storm Winds , " Proceedings ASCE,

Vol. 88, ST3, June 1962.

[11] Konishi, I., Shiraishi, N. , and Matsumoto, M., "Fundamental Approach to Aerodynamic

Random Response of Structural Sections, The Third Symposium on Wind Effects on Stuc~

tures, 1974 (in Japanese).

[12] Fung, Y. C., "The Theory of Aeroelasticity , " J. Wiley, 1955. . ::'=t c <

[13] Liepmann, H. W. , "On the Application of Statistical Concepts to the Buffeting

Problem, J. Aeronaut. Sci. 19, 1952.

[14] Konishi, I., Shiraishi, N. , and Matsumoto, M., "Studies on Wind Resistant Design of

Long-Span Suspension Bridges," Fac. Eng., University of Kyoto, 1975 (in Japanese).

Table 1. Parameters

Table 2. Parameters

Aerodynamic admittance; Ad^; neglected

Ad^; Eq. (2l) with approximations in Eq. (22)

Life curve slope; measured in smooth flow , v-i;' '

Decay factor of cross-correlation coefficient; 7

Unsteady aerodynamic coefficient; measured in smooth flow

Structural damping constant; 0.003 (0.02 in logarithmic damping)

Vibration mode; vertical bending; 6 modes torsion; 4 modes

Contribution factor in Eq. (16)); measured in smooth flow

(Cm(0)/Cl(0))

Aerodynamic admittance; Ad^; Eq. ( 17

)

Adw; Eq. (21) (Both in complex)

Life curve slope; measured in smooth flov" - ^

Number of divided segments of the span; 2

Lag of wind data series; 8 sec. •; ,,

'

Time length of the divided segments of wind data series; 8 sec.

Unsteady aerodynamic coefficient; measured in smooth flow

Structural damping constant; 0.005 (0.03 in logarithmic damping)

Vibration mode; vertical bending; 5 modes torsion; 2 modes

Contribution factor ( fi in Eq. (16)); measured in smooth flow

(Cm(0)/Cl(0))

IV- 18
Table 3.

Start Time of the Datn Series U(m /s) ffiit m/s) a. (ni/s) <7,(cm) < in V;i(l) "i (cm) a,' < in Vad)

1976, 9, 13

1: 59: 53 18.6 1.2 1.0 1.40 2.48 2.85 1.74

2: 00: 33 19.9 L.I 0.9 2.46 3.56 2.47 3. 32

2: 01: 13 20.8 1. I 1.0 2. 10 2.93 2.43 2. 29

2: 01: 53 19.6 1.4 1 . 0 3. 50 3. 19 2.31 2.74

2: 02: 33 22. 5 1.7 1.0 3. 08 2. 53 3. 36 3.77

u ; horizontal tomponeiit. of wi nil speod perpendicular to hridgo axis

w ; vertical c ompon nn t, of w ind sp(?ed

II ; mean of u.

a ; standard deviation (observed)

I?

; vertical bending response at the center of the main girder

(p ; torsional response at the center of the main girder

o ; standard deviation (calculated)

Fig. I Fig. 2

IV-19

DRAFT SEISMIC DESIGN GUIDELINES FOR HIGHWAY BRIDGES

James D. Cooper

Charles F. Scheffey

Federal Highway Administration

Washington, D.C.

Roland L. Sharpe

Ronald L. Mayes

Applied Technology Council

Palo Alto, California

. ABSTRACT ,.

.

Seismic design guidelines for highway bridges in the United States have been under

development since 1977. The draft guidelines represent the collective thinking of a distinguished group of academicians, consultants and highway bridge engineers. The guidelines were

formulated and based on both the observed performance of bridges during past earthquakes and

on recent research conducted in the United States and abroad. They are currently undergoing

evaluation by bridge designers and based on their comments are subject to change. The final

version of the guidelines will be available in 1981.

KEYWORDS: Bridges; seismic design; design guidelines; bridge.

V-1
INTRODUCTION

Significant earthquake engineering research has been initiated by many organizations

during the past several years. Structural analyses, response predictions, building codes,

and specifications have been refined and updated to provide the engineer with the background

and capabilities necessary to design and construct modern buildings to adequately resist the

forces developed during periods of strong earthquakes. However, little research had been

conducted in the United States prior to the 1971 San Fernando earthquake to insure the satisfactory seismic performance of highway bridges.

Researchers are now focusing their attention on the problem that exist in developing a

rational earthquake-resistant design methodology for highway bridges. It is interesting to

note that in 1971, 21 countries of the world specified some form of a seismic design coefficient for use in the design of buildings. Of the 21 countries, only four (India, Japan,

Turkey, and the United States) had requirements specifically for the design of highway

bridges, and it was not until just recently that researchers or practicing engineers in any

of the 21 countries took into consideration the dynamic response characteristics of bridge

structures in developing their seismic codes and design criteria.

CODE DEVELOPMENT IN UNITED STATES " •^

Until relatively recent times almost all consideration of earthquake forces on structures and relevant code considerations have been concentrated in building construction as

opposed to bridge construction.

The 1906 San Francisco earthquake that caused an estimated $50 million of property

damage and with subsequent $350 million in fire damage was considered ill fortune, and the

city rebuilt in almost identical fashion. It was the 1925 Santa Barbara earthquake with

several million dollars in damage that gave impetus to consideration of earthquake design

provisions in building codes. In 1927 the simple Newtonian concept that lateral earthquake

force is proportional to mass was included in the Uniform Building Code. The Long Beach

earthquake of 1933 with at least $50 million in damage emphasized the need for earthquake

design; since then changes have continued in the various California building codes.

The first requirment in the United States for the inclusion of seismic loading in the

design of highway bridges was in the AASHO, American Association of State Highway Officials

V-2
(now American Association of State Highway and Transportation Officials, AASHTO) 19 58-59

interim to the 1957 Specifications which were unchanged until the 1975 Interim Specifications. In 1971 the AASHTO Bridge Committee was prepared to present a new earthquake criteria

for adoption. Prior to adoption, the San Fernando earthquake occurred and demonstrated that

the proposed revisions were inadequate and did not seem to apply.

The 12th edition (1977) of the AASHTO "Standard Specifications for Highway Bridges" presented a new approach for designing highway bridges to withstand earthquake forces. Article

1.2.20 of the Specification requires, "In regions where earthquakes may be anticipated,

structures shall be designed to resist earthquake motions by considering the relationship of

the site to active faults, the seismic response of the soils at the site, and the dynamic

response characteristics of the total structure." The specification is generally based on

the 1973 Earthquake Design Criteria for Bridges for the State of California. The California

criteria were developed for California conditions and subsequently modified to allow for

their use in other areas of the United States where possibly damaging seismic motions can

occur. The impact of the application of the specification, particularly in regions outside

California, on design complexity, design and construction costs, and construction complexity

was unknown at the time of adoption.

Consequently, a study entitled "Bridge Seismic Design Criteria" was initiated in 1977

to: (1) evaluate current criteria used for seismic design; (2) review recent seismic

research findings for design potential and use in a new specification; (3) develop new and

improved seismic design criteria; and (4) evaluate the impact of the proposed criteria on

construction and costing. .

DRAFT SEISMIC DESIGN GUIDELINES

A working draft of preliminary seismic design guidelines for highway bridges has been

developed by a Project Engineering Panel of Applied Technology Council (ATC) of Palo Alto,

California. Leading experts in seismic design from universities, private consulting firms

and AASHTO were assembled by ATC to serve on the Project Engineering Panel. The Panel has

provided continued guidance into the development of the guidelines. The purpose of the

guidelines is to establish design and construction criteria for bridges subject to earthquake

V-3
motions in order to minimize the hazard to life and provide the capability for bridges

to survive during and after an earthquake, with essential bridges to remain functional.

The design earthquake motions specified in the guidelines are selected so there is a

low probability of their being exceeded during the normal lifetime expectancy of the bridge.

The probability of the elastic design force levels being exceeded in 50 years is in the range

of 80 to 90 percent. However, the design earthquake ground motion by itself does not determine risk; the risk is also affected by the design rules and analysis procedures used in connection with the design ground motion. Bridges and their components which are designed to

resist these motions and which are constructed in accordance with the design details contained in the guidelines may suffer damage, but should have low probability of collapse due

to seisraically induced ground shaking. ' '

'

The principle used for the development of the guidelines are: ri

1) Small to moderate earthquakes should be resisted within the elastic range of the

structural components without damage.

2) Realistic seismic ground motion intensities are used in the design procedure.

3) Exposure to shaking from large earthquakes should not cause collapse of all or part

of the bridge. Where possible, damage that does occur should be readily detectable and

accessible for inspection and repair.

A basic premise in developing the bridge seismic design guidelines is that they be

applicable to all parts of the United States. The seismic hazard exposure varies from very

small to rather high across the country. Therefore, for purposes of design, four Seismic

Performance Categories (SPC) are provided to which bridges are assigned. Additionally,

bridges are classified as to their relative importance—either as an essential bridge or

all others. Differing degrees of complexity and sophistication of seismic analysis and

design are specified for each SPC. Category D bridges include those designed for the highest

level of seismic performance with particular attention to methods of analysis, design and

quality assurance. Category C bridges include those where a slightly lower level of seismic

performance is required and therefore the potential for damage is slightly greater than the

SPC D. Seismic Performance Category B bridges include those where a lesser level of seismic

performance is required and a minimum level of analysis and specific attention to support

V-4
design details are provided. Category A bridges include those where no seismic analysis is

required, but attention to certain design details for superstructure support is provided.

DRAFT GUIDELINES EVALUATION

The adequacy and impact of the draft seismic design guidelines for highway bridges is

to be evaluated during the next year by redesigning selected bridges and then comparing both

general design and construction feasibility and costs with that of the originally designed

bridges. Based on the evaluations by selected State highway departments and private consultants, the guidelines will be revised and then presented to AASHTO for their consideration

and adoption. The guidelines will be finalized in early 1981.

V-5

ANALYSIS AND DESIGN OF CRACKED REINFORCED CONCRETE

NUCLEAR CONTAINMENT SHELLS FOR EARTHQUAKES

Peter Gergely

Cornell University, Ithaca, N.Y.

and

Richard N. White

Cornell University, Ithaca, N.Y.

, ,. ABSTRACT , • ,

,

Cracking reduces the shear stiffness of reinforced concrete cylindrical containment

shells. An extensive experimental program on specimens carrying biaxial tension and cyclic

shear produced highly nonlinear shear-slip curves, relatively high sliding shear strength,

and low shear stiffness. The effects of this behavior on seismic response and design are

discussed.

KEYWORDS: Concrete; containment vessels; cracking; dynamic analysis; hysteresis; nuclear

structures; reinforced concrete; seismic effects; shear; stiffness; testing.

VI-l
INTRODUCTION

One of the critical loading conditions in the design of cylindrical reinforced concrete

containment vessels is the combination of internal pressure and membrane shear from seismic

forces. Vertical and horizontal cracks form as a result of the pressure and this leads to a

significant reduction in shear stiffness of the structure. The seismic shear forces must be

transmitted across horizontal cracks by the combination of interface shear transfer, dowel

action of the reinforcement, and axial forces in inclined steel (if inclined steel is used).

The seismic loading also produces inclined cracks which further reduce shear stiffness.

The cyclic shear transfer mechanism across cracks (also called sliding shear) has been

studied at Cornell University with an extensive series of tests on uniaxially or biaxially

tensioned reinforced concrete specimens subjected to simultaneous cyclic shear. The accompanying analytical studies have investigated the dynamic behavior of typical containment

structures with nonlinear shear-slip characteristics at cracks.

Typical PWR (pressurized water reactor) containment vessels in the U.S. have a cylindrical wall thickness of about 50 in. (about 1.25 m) and usually a hemispherical dome with a

thickness of about 30 in. (about 0.75 m). The thickness of the steel liner is usually 1/4 or

3/8 in. (6 or 10 mm) and it is anchored to the concrete on a grid of about 12 in. (0.3 m).

The wall is commonly reinforced with two orthogonal layers of #18 (57 mm diameter) bars near

each face (Fig. 1). Inclined (diagonal) bars are often added, especially near the base of

the wall, to help carry seismic membrane shear forces. One of the principal purposes of the

investigation has been to evaluate the seismic response of containment shells without diagonal bars.

FAILURE MODES

The primary objective in the design of containment shells is to preserve the integrity

or leak-tightness of the liner. Normally the stresses in the liner are low; for example, for

a peak seismic shear stress of 200 psi (1.4 MPa) in a 50 in. (1.27 m) thick concrete shell

the stress in a 3/8 in. (9.5 mm) liner is only 1.4 ksi (9.8 MPa), assuming linear elastic

behavior. However, when internal pressurization causes horizontal and vertical cracks, the

VI-2
shear stiffness of the concrete decreases and the stress in the liner increases. Two main

types of structural failure modes are possible that conceivably could lead to excessive

strains in the liner: diagonal tension failure (controlled by the reinforcing), and sliding

shear failure (controlled by both concrete and reinforcing).

Sliding Shear Failure

Seismic shear must be transmitted across horizontal cracks that may be open to about

0.030 in. (0.75 mm) as a result of internal pressurization. The transfer mechanism involves

interface shear transfer of concrete (also called aggregate interlock or shear friction) and

the dowel action of vertical bars. High-level cyclic shear tends to degrade the concrete

surfaces and produce relatively large slips. Dowel action of the bars must then carry an

increasing share of the total shear force and splitting along the bars may develop. As both

of these mechanisms develop, the shear stiffness decreases, which in turn affects the dynamic

response of the structure and increases the stresses and strains in the liner. The sliding

shear behavior of cracked concrete has been the subject of an extensive research project at

Cornell University which is summarized below.

Diagonal Tension Failure

Prior to diagonal cracking the stresses in the orthogonal reinforcing bars are produced

by pressurization and by tensile forces generated at the cracks by the interface shear transfer mechanism. Seismic shear may cause diagonal cracks and an equal increase of force in the

orthogonal steel, as shown in Fig. 2a, when the cracks are at 45°.

Thus the force in the orthogonal steel is composed of the membrane force due to pressurization and the increase in force caused by seismic shear after the formation of diagonal

cracks. It is customary to design the steel on a factored load basis with the steel operating at 0.9 yield stress.

When the factored seismic shear is larger than the capacity of sliding shear transfer,

inclined steel must be used for the excess shear which is not carried by the orthogonal

steel acting in conjunction with the concrete. '

Diagonal cracking changes the load transfer mechanism as can be seen in Fig. 2b.

Diagonal compression acts between the diagonal tension cracks and this involves compressive

stresses across the pre-existing orthogonal cracks (not shown); this stress intensity is VI-3
equal to the shear stress for 45° cracks. Thus the width of the orthogonal cracks tends to

decrease as the seismic shear stress increases, which enhances the interface shear capacity

and reduces the tension in the orthogonal bars. '

The equilibrium conditions described above can be used to design the reinforcement.

Strains beyond yield; especially in the diagonal bars, do not necessarily represent any

impending failure of the shell. A compatibility analysis could result in a prediction of

deformations and forces in the concrete, the reinforcement, and in the liner; however, such

an analysis hinges on the thorough knowledge of the behavior of all constituent elements and

their interaction.

The element shown in Fig. 2a is taken at the maximum tangential shear stress position of

the cylindrical shell. The elastic shear stress distribution around the cylinder is sinusolidal but yielding or softening at the peak stress region would clearly increase the stresses

elsewhere around the circumference. This force redistribution represents an excess capacity

which is currently not considered in design.

Diagonal tension failure obviously begins in a region of high shear. However, it is not

clear how diagonal cracks can propagate around the cylindrical shell because the shear forces

are theoretically zero at points 90° from the points of maximum shear force. A complete

strain and force analysis of the entire vessel may be necessary to evaluate the failure mechanism, which would involve a combination of membrane and radial shear. '

'

'

TEST PROGRAM

Extensive research at Cornell University on uniaxially loaded specimens with large bars

[1] and recent biaxially tensioned specimens [2] with //6 (19 mm) bars have shown that relatively high cyclic shear can be transmitted across cracks but the shear distortions are

appreciable.

(1) Jimenez, R., Gergely, P., and White, R.N., "Shear Transfer Across Cracks in Reinforced

Concrete," Report 78-4, Structural Engineering, Cornell University, August 1978, 357 pp.

(2) White, R.N., Perdikaris, P.C., and Gergely, P., "Strength and Stiffness of Reinforced

Concrete Containments Subjected to Seismic Loading: Research Results and Needs," Structural

Mechanics in Reactor Technology, Berlin, August 1979.

VI-4
Fig. 3 summarized the results of one series of tests on 6 in. (15 mm) thick biaxlally

stressed specimens. The steel ratio p refers to the smaller amount of steel; in the other

direction twice as much steel was used. The steel stress f^ is the applied initial stress

of 0, 0.3fy, 0.6fy, or 0.9fy, where f^ is the yield stress, nominally 60 ksi (414 MPa).

Specimens were loaded with either monotonically increasing shear to failure, or with fully

reversing shear. Shear was applied after the specimens were cracked by applying biaxial

tension to the bars. It is seen that the reduction due to cyclic shear is about 15-20%;

usually 10 cycles were applied at 50 psi or 0.34 MPa load increments beginning at 125 psi

(0.86 MPa). All specimens failed by diagonal tension indicating that sliding shear failure

is probably not a primary failure mode. '

However, the shear distortions in cracked concrete are quite high and appreciable slip

occurs along cracks and from the effects of diagonal cracking, when the shear is reversed.

A set of typical hysteresis curves is shown in Fig. 4 for a specimen stressed biaxlally to

0.6fy. The effective shear rigidity is only a small fraction (less than 10%) of that for

uncracked concrete after a few shearing cycles. This low stiffness affects the dynamic

behavior of the containment shell. It should be noted that recent tests on larger specimens

2 ft. *0.61 m) thick and 5 ft. (1.52 m) square at the Portland Cement Association have given

similar results. •

The splitting of concrete along bars has not yet been fully assessed. In the uniaxial

tests [1] splitting occured in some specimens with #11 or #14 bars (36 and 43 mm) but this

behavior depends strongly on the orientation of the bars and a slight unintentional component of the tensile bar force paralled to the crack can initiate splitting. In the tests at

the Portland Cement Association cracks were observed along nearly every bar but only a few of

these penetrated the entire thickness. It is not known whether a splitting plane developed

through all bars parallel to the concrete surface.

DYNAMIC ANALYSIS

The hysteretic shear-slip relationships found in the experimental research were used

in dynamic analyses of containment structures to evaluate their effects on dynamic response.

VI-5
in particular on the maximum shear stress and maximum slip at a horizontal crack. The rotational inertia was included and idealized hysteretic shear-slip curves were assigned to each

horizontal crack [3]. '•

A number of idealizations were used in which the main parameters were the number of

horizontal cracks and the shear modulus between cracks. Typical results for an artificial

ground acceleration scaled to 0.4g maximum ground acceleration show that the peak shear

stresses are reduced by about 30% for nonlinear stiffness characteristics as compared with an

uncracked elastic idealization. The maximum shear stress ranged between about 200 and 300

psi (1.38 and 2.07 MPa) for the several cases analyzed so far. The maximum slips at the

cracks were only about 0.02 in. (0.51 mm). The displacement time history near the top of the

vessel had more high-level peaks in the nonlinear analysis than in the uncracked linear analysis. In all analyses the liner was neglected. r- - . .

The experimental and analytical work is continuing and additional parameters, such as

the effect of diagonal bars, are being studied.

ACKNOWLEDGMENTS

This study was supported by the Nuclear Regulatory Commission and the National Science

Foundation; however, any opinions, findings, conclusions, or recommendations expressed herein

are those of the authors and do not necessarily reflect the views of the NRC and NSF. The

contributions of several current and past graduate research assistants, including R. Jimenez,

J.K. Smith, P. Perdikaris, and C.H. Conley, are acknowledged. -lisb lo-i;

(3) Smith, J., Gergely, P., and White, R.N. , "The Effects of Cracks on the Seismic Analysis

of Reinforced Concrete Nuclear Containment Vessels," Report No. 368, Structural Engineering,

Cornell University, April, 1977.

VI-6
liner

C 0

C 0

-t

o

Tl2 in.

I 0.30 m

Fig. 1 Typical cross section of wall with #l8 (57 nim) bars.

Fig. 2 Free-body diagrams for diagonally cracked elements: (a) Forces

in orthogonal steel caused by seismic shear, (b) Diagonal

compression forces.

VI-7
-40 0 40 80 120 160 200

Average Shear Strain, y (RADxlO"^)

4 Typical shear distortion curves for a specimen tensioned biaxially to 0

VI-8
COMPARISON OF THE MEASURED AND COMPUTED RESPONSES OF THE YUDA

DAM DURING THE JULY 6, 1976 AND JUNE 12, 1978 EARTHQUAKES

Ryuichl lida

Norihisa Matsumoto

Satoru Kondo

Public Works Research Institute

:. ABSTRACT

Two accelerograms were recorded at the foundation and dam crest of the Yuda Dam, an

arched gravity dam, during the July 8, 1976 and June 12, 1978 earthquakes. The frequency

response functions for dam crest to foundation were calculated from these accelerograms.

The ground acceleration was used as an input, and the dynamic response was calculated using

two-dimensional FEM analysis to compare with the measured response.

The essential conclusions may be enumerated as follows:

1) The displacements at the crest of Yuda Dam were calculated from recorded accelerograms using a digital filtering procedure in which the cutoff frequency of the weighting

function was slightly lower than the fundamental frequency of the dam.

2) According to the frequency response functions obtained from two accelerograms, the

performance of the dam body may be considered to be linear elastic. The water level of the

reservoir influences the frequency response function.

3) The computed response of the dam crest using the foundation accelerogram as an

input from the two-dimensional FEM analysis coincides with the measured one. Comparing

Westergaard's method and Chopra's method in the evaluation of hydrodynamic pressure, Chopra'

method gives the response which contains more low frequency components than Westergaard's

method. However, there is not much difference between them. Westergaard's added mass when

computed by neglecting the compressibility of water provides a satisfactory approximation

of the hydrodynamic pressure.

KEYWORDS: Arched gravity dam; dynamic analysis; earthquake accelerogram; frequency response

functions; hydrodynamic pressure.

VI-9
1- INTRODUCTION

In the design of concrete dams subject to earthquake loads, linear elasticity of

concrete can be assumed when the strain of concrete does not exceed certain established

values. Because significant parameters such as elastic constants and damping ratio can be

determined by laboratory and field tests, the most critical point in analysis procedures of

concrete dams lies in the evaluation of reservoir effects. One method of accounting for the

reservoir effects is to represent the complete dam/reservoir system as an assemblage of

finite elements. This approach requires much computational effort and is not very practical

for a dam/reservoir interaction problem. Westergaard ' s added mass concept [1] is another

method used to evaluate the dam/reservoir effects. In this theory, the dam is assumed to

be rigid, and the hydrodynaraic pressure is represented as an effective mass added to the

structure. A third approach is Chopra's substructure concept [2] in which results of the

continuum-reservoir solution, including hydrodynamic interaction, are used as input. At

present, these latter two methods are considered to provide the most practical design procedures for the dynamic analysis of concrete dams. These two methods can be best verified by

comparing analyzed results with actual behavior of a full-scale dam.

- - i>cv <::.i i The accelerograms of the July 8, 1976 earthquake (simply described as the 1976 earthquake hereafter) and the June 12, 1978 earthquake (simply described as the 1978 earthquake

hereafter) were recorded at the foundation and crest of Yuda Dam. The maximum values of

acceleration at foundation for the 1976 earthquake and the 1978 earthquake are about 5 cm/s^

and 25 cm/s respectively, being five times different from each other. The water level of the

reservoir is El. 219.8 m (76 percent of dam structural height) during the 1976 earthquake and

El. 231.9 m (89 percent of dam structural height) during 1978 earthquake. As the effects on

dam body due to the difference of input acceleration amplitude and water level are of interest, the recorded accelerograms were analyzed and the results of two-dimensional FEM analysis

were compared with the measured response.

VI-10
I

2. GENERAL DESCRIPTION OF YUDA DAM

Yuda Dam is an 89.5-m high arched gravity dam which was completed in November 1964 by

the Ministry of Construction of the Japanese Government (Figure 1). The location map of

Yuda Dam is shown in Figure 2, in which two epicenters of the 1976 and 1978 earthquakes are

shown, too.

The general information and layout data for the dam and reservoir are shown in Table 1.

The plan and sections through the dam at different locations along the axis are shown on

Figures 3 and 4. The multipurpose dam provides for flood control, municipal water, irrigation, and power.

Two strong-motion accelerometers were installed at Yuda Dam. One is located at the cen

ter of the dam, at the crest, and the other in the tunnel on the left abutment (see Figure

3). Both accelerometers are at the same elevation and of the electromagnetic type. The sen

sitivity of the accelerographs is above 90 percent in the frequency range of 0.3"^ 30 hertz.

They are capable of measuring high-frequency components of vibration with good accuracy.

3. ANALYSIS OF THE 1976 EARTHQUAKE

3.1 The 1976 Earthquake and Its Accelerograms

The earthquake motions were recorded at 8:48 p.m., July 8, 1976. The epicenter of this

earthquake was located offshore from Iwate Prefecture at 40°14'N. latitude and 142°26'E.

longitude, and the hypocenter at 30 km. The distance from the epicenter to the damsite was

168 km. The magnitude of the earthquake was 5.9 on the Richter scale. The intensity scales

in the vicinity of the damsite were Intensity Scale IV (the Japan Meteorological Agency).

The reservoir water level was El 219.8 m when the earthquake occurred.

The recorded accelerograms at the foundation and at the crest of the dam are shown on

Figure 5.

The maximum values of acceleration in the horizontal upstream/downstream, cross canyon

and vertical direction are 4.6, 5.7, and 4.5 cm/s respectively at the foundation, and 58.4,

10.7, and 7.5 cm/s^ respectively at the crest. At the crest of the dam, the vibration is

amplified, especially in the upstream/downstream direction. Assuming the time acceleration

histories of the foundation and crest to be f(t) and g(t), respectively, Fourier spectra F('^

for the foundation and G{^^) for the crest will be given by:

VI-11
F(a)) f(t)e-tot dt (1)

G(a)) —00

CO g(t)e -ut dt (2)

and the acceleration frequency response function Z(u) for the dam crest to foundation is

expressed as follows:

Since Yuda Dam is a three-dimensional structure, the earthquake ground motion in the

upstream/downstream direction, for example, affects the crest motion not only for the

upstream/downstream direction but also for the cross canyon and vertical ones. For simplicity, however, the frequency response function was calculated for each component, neglecting

these couplings of motion. Figure 6 shows the Fourier amplitude spectra of the 1976 earthquake. Figure 7 is the frequency response function for dam crest obtained from Eq. 3. The

results shown in Figures 6 and 7 indicate that the first fundamental frequency of the dam is

about 5 Hz, two horizontal input accelerograms in the upstream/downstream and cross canyon

direction have similar frequency components, most of which are lower than the fundamental

frequency of the dam, and the frequency components of the vertical input accelerogram scatters in wider band than those of horizontal accelerograms.

The displacement response for the dam crest relative to the foundation will be calculated later in this report. In order to compare the computed response with the measured one,

it is necessary to obtain the displacement from the digitized measured accelerograms. The

original accelerograms have baseline error, so that it should be eliminated from original

digitized accelerograms.

The baseline correction procedure, which is briefly summarized in the flow chart shown

on Figure 8, was applied to the original accelerograms. This procedure is similar to one

proposed by M. D. Trifunac [3]. The main feature of this procedure is the digital filtering

procedure and its transfer function H(f) is expressed as follows [4]:

Z(u)) = G(a)) / F((ij) (3)

VI-12
1 f < f c

H(f) H f^ - f l:fT ^' (A)

0 f > f T

' ^

where: = filter roll-off termination frequency

= cutoff frequency

The weighting function h(t) associated with H(f) is given by:

^ a

J —

h(t) = \

ei'^t j^(^) (5)

where: u = 2Trf

u = circular frequency ^

f = natural frequency i

For discrete or time sampled data, the weighting function h^ is given by:

cos 2 Xp - cos 2ir„ XTM hn = ^—5 ^L_i_ (6)

n = 0, ±1, ±2 ±N

X^ = X(, + Xj^ ' '

' ' ...

where: X^ = f^^ / fg

fg = 1 /

Aj. = equally spaced time interval

In Eqs. 4 through 6, f^ and f^. should be determined according to the frequency range of

the long-period noise introduced into the data by the warping of the record paper and the

digitization process. Because the recording speed of the electromagnetic-type accelerographs

is 4.0 cm/s, which is faster than that of standard strong-motion accelerographs, the accelerogram of Yuda Dam contains higher frequency errors than that of standard accelerographs, and

VI-13
therefore, a higher cutoff frequency will be required. Since the ground motion must contain

long-period displacement, a high-frequency cutoff was not desirable. Also, low-frequency

error components are not all filtered out when f^ = 0.05 hertz.

The fundamental frequencies of the dam in the three different directions are higher than

4.5 Hz, so frequency components of relative accelerograms less than 2.5 Hz will be small.

Figure 9 shows the relative displacements for dam crest relative to foundation obtained from

the procedure shown in Figure 8 using f^ = 2.5 and f j = 3.0. The maximum displacement at the

crest is approximately 0.05 cm. .

3.2 Dynamic Analysis

A two-dimensional finite element analysis was conducted to determine the dynamic •:

response of the Yuda Dam to the July 8, 1976 earthquake using the computer program developed

by P. Chakrabarti and A. K. Chopra [5]. Prior to the analysis, two procedures were added to

this computer program — one to obtain the acceleration response and the other to utilize

Westergaard ' s added mass approach.

Yuda Dam is a curved concrete gravity dam, as shown on Figure 3. It was designed by the

trial-load method of analysis and should be analyzed as a strictly three-dimensional structure. However, a two-dimensional analysis would be permissible, because (1) Yuda Dam is

thick compared to a normal arch dam, and (2) the three-dimensional analysis requires much

more computer effort to compute the several different cases required and thus is uneconomical .

Material properties used in the analysis are shown in Table 2. The unit weight and the

elastic modulus of concrete were determined from laboratory tests. The damping ratio (percent of the critical) was determined from the field f orced-vibation tests of other arch dams

[6, 7].

The hydrodynamic effects of the reservoir on the dam were evaluated using two methods:

1) Westergaard ' s added mass concept, and • • • -

. : ...

2) Chopra's substructure concept. " • a . i..

In Westergaard ' s approach, the following assumptions were made: , ti," ' . - S-

* Water is assumed to be linearly compressible and its internal viscosity is ^

neglected

.

* The motion of the dam/reservoir system is considered as two-dimensional.

VI-14 .
• The upstream face of the dam is vertical. ... ^ ,

• ,

' The reservoir bottom is horizontal, the water extends to infinity in the upstream

direction, and the hydrodynamic pressure decreases to zero at infinity.

* The dam is rigid. • . . . .

With these assumptions, the small amplitude motion of the reservoir is governed by the

two-dimensional wave equation. ^

, - ,,

,

2

3 4' + 3 "t* = w 3

^ y ^

3 m^ 3 y^ § ^ 3 t^ ' -

'

where cj) = velocity potential

w = unit weight of water

g = acceleration of gravity

k = bulk modulus of elasticity of water

If the fundamental period of the reservoir is less than that of the ground motion, the

hydrodynamic pressure for Eq. 7 for above boundary conditions is

n

^ = - 8a wh Ztt t Z _l_sin_S^LX (8)

IT 2 T 1.3,5 ^2c^ 2h

where T = period of earthquake ground motion

a = acceleration amplitude of ground motion

h = depth of the reservoir

y = depth from the water surface

C„ = A - (9) ^ n2 g k t2

In Chopra's approach, the above first four assumptions are the same. However, the dam

is considered to be deformable. The complete system is considered as two substructures —

the dam, represented as a finite element system, and the reservoir, as a continuum of

infinite length in the upstream direction governed by the wave equation.

Eigenvalues and mode shapes were calculated at two sections — one at the maximum section, and the other block 5 section (Figure 10). In computing Westergaard 's added mass, the

fundamental period of the ground motion was assumed to be 2.0 seconds, because the predominant period of the displacement is about 2.0 seconds.

VI-15
According to the results of the analysis of the recorded accelerograms, the natural

frequency of the crest of Yuda Dam in the upstream/downstream direction is 4.9 hertz, which

coincides with the fundamental frequency of block. 5 sections. The maximum section has a

lower fundamental frequency than the results of recorded accelerograms. Although Yuda Dam

should be treated as a three-dimensional structure, the block 5 section could be considered

to represent the whole structure, thus allowing the analysis of Yuda Dam as a two-dimensional

problem. For this study, the response of the dam will be obtained using block 5 section.

Next, the acceleration and displacement response of block 5 section subjected to the

recorded ground motion of Yuda Dam is described. High-pass filtered accelerograms were used

as the input accelerogram.

The dynamic response of Yuda Dam was calculated for the following seven cases:

Case No. Foundation Motion Hydrodynaralc Pressure

1 Horizontal Reservoir empty

2 Horizontal Westergaard ' s theory

3 Horizontal Chopra's theory

4 Horizontal plus vertical Reservoir empty

5 Horizontal plu s vertical Westergaard ' s theory

6 Horizontal plu s vertical Chopra's theory ( a* = 0.0)

7 Horizontal plus vertical Chopra's theory (a* = 0.8 5)

* a = a refraction coefficient at the bottom of the reservoir, or

a = ^ L where k = — i- k + 1 C • W

where W^. = unit weight of the foundation rock "

'

Cj. = P-wave velocity of the foundation rock ' >

W = unit weight of water "•'

'

C = wave velocity of water '

- ;

VI-16
The calculated accelerograms in the horizontal and vertical direction at the crest of

the dam are shown on Figures 11 and 12, respectively. The calculated displacements are shown

on Figures 13 and 14. Included on Figure 11 through 14 are the measured results for the

convenience of comparison.

Some of the characteristics drawn from Figures 11 through 14 are:

1) The accelerations and displacements calculated by assuming the reservoir empty were

smaller than the measured ones.

2) The effects of vertical ground motions were small. There is not much difference

between case No. 6 and No. 7 in which the refraction coefficient at the bottom of the reservoir is different from each other. The fundamental modes of the gravity dam are horizontal

bending motions. Although the vertical components higher than 15 Hz do contribute to the

response of the dam, the input vertical accelerogram high frequency components.

3) The displacement and acceleration results, using Westergaard's and Chopra's

approach, were in good agreement with the measured results.

The two assumptions of Westergaard's theory — (1) that the period of ground motion is

longer than that of the reservoir, and (2) that the dam is rigid, may be permissible for the

1976 earthquake. As to (1), the predominant periods in the stream direction of acceleration

and displacement are 0.5 and 2.0 sec, respectively.

As for assumption (2), the relative displacement of the dam below the water level was

quite small when compared to the absolute displacement of the foundation.

Comparison of the computed and measured responses in the time domain were described

above. Next, the comparison in frequency domain corresponding to Figure 10 will be shown.

Figure 15 shows the frequency response functions of FEM analysis.

Figure 15 indicates that whereas the horizontal input acceleration contributes the horizontal response acceleration for wide band frequency in Westergaard's theory, it does not

contribute except around 7.8 Hz of resonance frequency of the reservoir in Chopra's theory.

The vertical input acceleration contributes the vertical response for dam crest only in the

frequency band higher than 15 Hz. The maximum values of response magnification ratio in the

horizontal upstream/downstream direction in Figure 15 are from 40 to 50, although the maximum values of measured response are about 30 to 40 in Figure 7. This is because, in Figure

7, the peaks of frequency response function are truncated by the application of Parzen's '

spectral window. The acceleration response function of the measured accelerogram in the

VI-17
upstream/downstream have peaks at 5.0, 9.0, 10.5, 11.8 and 15.0 Hz, which is similar to that

of Westergaai d

' s theory than that of Chopra's theory. The frequency response function of

Chopra's theory contains fewer high frequency modes compared to the measured frequency i

response function. ^ : ; .•

4. ANALYSIS OF THE 1978 EARTHQUAKE

4.1 The 1978 Earthquake and Its Accelerograms j .

This earthquake (= 1978 Miyagi Prefecture Earthquake, Magnitude 7.4) occurred on June

12, 1978, with its epicenter at 38°09'N latitude and 142°10'E longitude, and its hypocenter

40 km deep. The damsite and location of the epicenter are shown in Figure 2. The distance

between damsite and epicenter is 169 km, which is approximately equal to that of the 1976 '

earthquake. The plot of digitized accelerograms recorded at Yuda Dam during this earthquake

is shown in Figure 16. The ratios of maximum values of acceleration for dam crest to foundation are 7.4 in the upstream/downstream, 1.9 in the cross canyon, and 1.4 in the vertical

direction.

Fourier spectra for recorded accelerograms are shown in Figure 17. Frequency response

functions from Eq . 3 are shown on Figure 18. The frequency characteristics of the foundation accelerograms are similar to those of the 1976 earthquake, with low frequency components from 2 to 3 Hz in the horizontal direction and rather high frequency components in the

vertical direction. The input acceleration levels in the 1976 and 1978 earthquakes are five

times different from each other. Furthermore, the water level of the reservoir is El. 219.8

m (76 percent of the dam structural height) for the 1976 earthquake and El. 231.9 m (89 percent of the dam structural height) for the 1978 earthquake. It is of interest how these two

conditions influence the frequency response function.

There is a good coincidence between the frequency response functions of the upstream/

downstream component on Figures 7 and 18 as for the shape and amplitudes in the band about

5 Hz, although the peak frequency of the 1978 earthquake shifts to the lower frequency. The

peak frequency and amplitude differ in the band from 8 to 25 Hz. The amplitude of the frequency response function of the 1976 earthquake is much greater than that of the 1978 earthquake in the band higher than 15 Hz. Table 3 shows the natural frequencies of modes obtained

from FEM analysis of Section 3.2 (the reservoir effect is evaluated by Westergaard 's theory).

VI-18
The natural frequencies up to the third mode in Table 3 correspond with the peak frequencies of Figures 7 and 18. A peak in Figures 7 and 18 has some peaks, because the dam is

an arched gravity dam having the adjacent natural frequencies of symmetrical and asymmetrical

modes . . , ^ - • . r. - .

Assuming the reservoir and dam to be two-dimensional, the resonant period T (sec) of the

reservoir will be expressed as follows from Eq. 9.

Now consider El. 174.0 m of the foundation of No. 5 block used in the analysis as the

average elevation of the bottom of the reservoir, then 7.8 Hz and 6.2 Hz will be obtained

from Eq. 10 for the 1976 and 1978 earthquake, respectively, so that the frequency which

gives the peak amplitude in Figure 18 shifts to the lower part as much as 0.5 to 1.0 Hz compared to Figure 7. The shape and amplitude of the frequency response functions of the two

earthquakes coincide in the band below 15 Hz, and therefore it may be concluded that the

performance of the Yuda Dam is linear elastic in these two earthquakes.

There is also a good coincidence of the shape and amplitude between the frequency

response functions of the cross canyon component. The frequencies which give peak amplitude

in the 1978 earthquake are lower than those of the 1976 earthquake, because of the difference

of the water level of the reservoir.

The frequency response function of the vertical component of the 1976 earthquake does not

have definite peaks, although that of the 1978 earthquake has them at 10.2, 13.7, and 15.2 Hz.

This may be related to the accuracy of digitalization of the accelerograms. The recording

speed of the accelerographs was 4.0 cm/s, and the sensitivity of acceleration amplitude is

the same in all components of the 1976 earthquake.

Accordingly, the amplitude of the original recorded accelerograms except the upstream/

downstream component of dam crest is small, and their accuracy is inadequate in the frequencies higher than 15 Hz. On the contrary, the recording speed of accelerographs of the 1978

earthquake was 10.0 cm/s and the sensitivity of the acceleration amplitude of the foundation

components is eight times that of the dam crest components. Therefore the accuracy is the

same for all components of the 1978 earthquake.

T (\0)

VI-19
4.2 Dynamic Analysis '" -

-

The acceleration and displacement response was computed using the foundation accelerograms of the 1978 earthquake as an input [8]. The method of analysis, material properties

and the analyzed section are the same with those described in Section 3.2. The dynamic

response was calculated for the following five cases: -i''^

Case No. Foundation Motion Hydrodynamic Pressure

1 Horizontal Westergaard ' s theory

2 Horizontal Chopra's theory

3 Horizontal plus vertical Westergaard ' s theory

4 Horizontal plus vertical Chopra's theory ( a* = 0.0)

5 Horizontal plus vertical Chopra's theory (a = 0.85)

a* = refraction coefficient at the bottom of the reservoir.

Since the case of the empty reservoir resulted in a smaller response compared to the

measured one in the analysis of the 1976 earthquake, this condition was omitted in the

following analysis.

The acceleration and displacement responses for dam crest are shown in Figures 19

through 22. Displacement responses obtained from the measured accelerograms using the same

procedure as before are included in Figures 21 to 22 for the convenience of comparison.

The foundation accelerograms presents the S-wave reached at t = 4.5 s. The acceleration

amplitude is large from t = 6.0 to 9.0 s. The acceleration amplitude is small after t = 9.0

i to t = 16.0 s. The second main shock was reached at t = 16.0 s. Examination of this input

acceleration shows the amplitude of the measured response to the upstream/downstream direction to be large in the time from t = 6.0 to 10.0 s and from 17.0 to 19.0 s. Analysis shows

that every case presents a large amplitude in the time from 6.0 to 10.0 sec, while the

amplitude is large in the time from 12.0 to 14.0 s when the measured response shows the decay

of vibration. Comparing case 4 and 5 in which the vertical input was considered as well as

the horizontal one, case 5 of the large refraction coefficient gave the larger response.

As for the vertical foundation accelerogram, the S-wave component is not definite, and

the acceleration amplitude is flat for both the foundation and the dam crest. Since

VI-20
cases No. 1 and 2 do not consider the vertical input acceleration, the computed vertical

acceleration in Figure 20 is the response to the horizontal input, and consists of low frequency components, which is not coincident with the measured response. The vertical acceleration response of case 3 to 5 shows close coincidence with the measured one from the

standpoint of frequency characteristics, because vertical foundation acceleration is included

as the analysis. However, there is not much difference in vertical displacement responses,

because the high frequency components of acceleration do not influence the displacement

response much.

As described before, the water level of the reservoir during the 1978 earthquake was

El. 231.9 m. Figure 23 shows the frequency response functions which corresponds to this

water level. The natural frequencies for the water level of El. 231.9 m shift to the lower

frequency as much as 0.5 Hz compared to those for the water level of El. 219.8 m.

5. CONCLUSIONS

The main results of the work presented in this paper can be summarized as follows:

1) The displacements at the crest of Yuda Dam were calculated from recorded accelerograms using a digital filtering procedure in which the cutoff frequency of the weighting

function was slightly lower than the fundamental frequency of the dam.

2) According to the frequency response functions obtained from two accelerograms,

the performance of the dam body may be considered to be linear elastic. The water level

of the reservoir influences the frequency response function.

3) The computed response of the dam crest using the foundation accelerogram as an

input from the two-dimensional FEM analysis coincides with the measured one. Comparing

Westergaard' s method and Chopra's method in the evaluation of hydrodynamic pressure, Chopra's

method gives the response which contains more low frequency components than Westergaard '

s

method. However, there is not much difference between them. Westergaard' s added mass when

computed by neglecting the compressibility of water provides a satisfactory approximation

of the hydrodynamic pressure.

REFERENCES

[1] Westergaard, H. M. , "Water Pressures on Dams During Earthquakes," Transactions, ASCE,

98, pp. 418-433 (1933).

VI-21
Chakrabarti, P., Chopra, A. K., "Earthquake Analysis of Gravity Dams Including Hydrodynamic Interaction," Earthquake Engineering and Structural Dynamics, Vol. 2, pp. 143-160 (1973).

Trifunac, M. D. , "Baseline Correction of Strong-Motion Accelerograms , " Bulletin of

Seismological Society of America, Vol. 61, No. 5, pp. 1201-1211, (1971).

Ormsby, J. F. A., "Design of Numerical Filters with Applications to the Missile Data

Processing," Journal of Association for Computing Machinery, Vol. 8, pp. 440-446,

(1961).

Chakrabarti, P., Chopra, A. K., "A Computer Program for Earthquake Analysis of Gravity

Dams Including Hydrodynamic Interaction," Report No. EE-RC 73-7, EERC , University of

California, (1973).

Okamoto, S., "Introduction to Earthquake Engineering," University of Tokyo Press,

(1971). -

•

Rouse, G. C. and Boukamp, J. G., "Vibration Studies of Monticello Dam," Research

Report No. 9, Bureau of Reclamation (1967).

Matsumoto, N. , "A Comparison Between Measured and Computed Response of Yuda Dam During

the July 8, 1978 Earthquake," Northern Japan, REC-ERC-78-4, Bureau of Reclamation,

(1978).

-r7 -

Table 1 General information and layout data for Yuda Dam and Reservoir

RESERVOIR DAM

Reservoir area 6.3 km'

Total storage capacity 1 14 x lO'm'

Maximum water elevation 239.0 m

Flood season water surface elevation 222.0 m

Normal water surface elevation 236.5 m

Maximum structural height

Upstream slope (at crown)

Downstream slope (at crown)

Crest width

Base width

89.5 m

0%

40%

4.5 m

33.0 m

SPILLWAY AND OUTLET WORKS

Crest length < Axis radius Central angle332.0 m

138.0 m

110°

Orifice Maximum discharge

Two gates

600 mVs GEOLOGY OF FOUNDATIONS

Overflow spillway

Maximum discharge

Six gates

3240 m'/s

Granite, partly Andesite

VI-22
Fig. 1 Yuda Dam

VI-23
J IB J 16

J

• Dots not include

lorgt cobbXf Fig. 4 Sections through Yuda Dam

VI-24
.00 3.00 Horizontal Upstream/downstream component at foundation

1.00 1.00 i.c Horizontal Cross canyon component at foundation

O.aO 1.00 2.00 3.00

Vertical component at foundation

A/11*

\l

WW

r.oo 3.00 4.00 s.oo e.oo 1.00 12.00 13.00 lU.OO 15.

C

i.oo :a.oo Horizontal Up stream/downstream component at dam crest

I 00 3.00 3.00 1.00 S.OO t.OO 1.00 S.OO 10.00 12.00 13.00 It. 00 IS. 00 IB. 00 11.00 IS. 00 19.00 2g.00 Horizontal Cross canyon component at dam crest —

1.00 3.00 3.00 4.00 S.OO C.OO 1.00 B.OO t-OO 10.00 11.00 13.00 13.00 U.OO I S.OO '.00 11.00 19.00 20.00 Vertical coiiq>onent at dam crest

Fig. 5 Plot of digitized accelerograms recorded at Yuda Dam during the 1976 earthquake

VI-25
FREQUENCI ICRS) FRECUENCT ICPSI

Upstream/downstream component

at foundation

Upstream/downstream component

at dam crest

FFtauENCr ICPSI FREOUENCT (CPSl

Cross canyon component at foundation Cross canyon component at dam crest

FREOUENCT ICRS) FR£gu£f4Ci tCRSI

Vertical component at foundation Vertical component at dam crest

Fig. 6 Fourier spectra of recorded accelerograms of the 1976 earthquake

VI-26
Fig. 7 Frequency response functions for dam crest obtained from recorded accelerograms of

the 1976 earthquake

.00 20.00 FREQUENCT (lu)

Cross canyon cooponenc

-I 10-

"T 1 1 1 r

1 1 1 1 1 1 1 1 1 1 1 1 .00 E.OQ e.OO 10.00 t2.00 H.OO 16.00 Ifi.OO 20.00 FREQUENCT (Hz)

Vertical component

~1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1

1 0.00 3.00 V.OO B.OO 6.00 tO.OO 12.00 111. 00 16.00 16.00 20.00 FREQUENCT (Hz)

VI-27
Onginol Accelerogrom

ait,)

Low -Pass Filter

0,(ti)

Removal of LowFrequency Component

Oi{fi]-aiti)-o,{f,)

Compute Velocity Assuming v{tj = o)'0

V{fi)

Leo St Squore Fitting of and do

v,{f,)- V it,) + + ao-f

H{f]

1.0

Low -pass filter

fc fr

o,{fi). High-frequency component

filtered -out

a^ifi) ; Low-frequency component

filtered -out

Boseline Correction Corrected occeierogram

Fig. 8 Flow chart obtaining relative

displacement of the dam crest

Hi

Upstream/downstream component

Cross canyon component

Vertical component

Fig. 9 Relative displacement on the dam crest during

the 1976 earthquake

VI-28
Table 2 Material properties used in the dynamic analysis

Property fvlKSA

Unit weight of water 1.0 t/m'

Unit weight of concrete 2.3 t/m' Elastic modulus of concrete 2.5x10»t/m'

Poisson's ratio of concrete 0.2 Damping ratio 3%

El 241.5m

Fig. 10 Finite element idealization for the section through block 5

VI-29
75.0

50.0

25.0

0.0 If

1.00 2.00 3.00 y.OO 5.00 6.00 7.00 6.00 9.00 10,00 TIME :n seconds

Measured response

50.000

25.000

O.OOO

-25.000

-50.000 1 y

0 00 1.00 2.00 3.00 4.00 5.00 <S.00 7.00 8.00 9.00 10.00 TIME IN SECONDS

Case No. 1; reservoir empty, horizontal input

0.00 1.00 2.00 3.00 4.00 5.00 4.00 7.00 g.oo 9.00 lo.oo TInE IN SECONDS

Case No. 2; reservoir by Westergaard' s theory, horizontal input

Case No. 3; reservoir by Chopra's theory, horizontal input

Fig. 11 Comparison of measured and analytical acceleration responses in the

upstream/downstream direction at the crest of the dam

(the 1976 earthquake)

VI-30
so.ooo

TE.OOO

0.000

-25.000

-50.000 t 1 "pi MM fVW1.00 2.00 3.x d.oo 7.00 e.oo 9.00 10.x

Tln£ IN SECONDS Case No. A; reservoir empty, horizontal and vertical input

75.000

50.000

25.000

0.000

-26.000

-50.000

-75.000

i.ao 2.00 i.aa *.oo s.oo t.oa y.eo t.ao 9.x lo.x

Tln£ IN SECONDS

Case No. 5; reservoir by Westergaard ' s theory, horizontal and vertical input

O.X 1.x 2.x 3.x 4.x 5.x 6.x 8.x 9.x 10.

X

TinE IN SECONDS

Case No. 6; reservoir by Chopra's theory (a=0.0), horizontal and vertical input

Case No. 7; reservoir by Chopra's theory (a=0.85), horizontal and vertical input

Fig. 11 Comparison of measured and analytical acceleration responses in the

upstream/downstream direction at the crest of the dam

(the 1976 earthquake) (Continued)

VI-31
25.0

0.0

-25.0 Wo-* f,yv^._.

0.00 1.00 2.00 3.00 I. 00 5.00 6.00 7.00 B.OO 9.00 10.00 Tlie IN SECONDS Measured response

25.000

O.OOO

-25.000 0.00 1.00 2 00 3.00 <.oo 5.0C «.oo ;.m e.oo 9.00 10.00 Tlr£ IH SECOHOS Case No. 1; reservoir empty, horizontal input

0.000

-25.000 1

-'^^ "-"•f

1 0.00 1.00 2.00 3.00 <.oo s.oo <.oo 7.n e.oo ».oo 10.x

Tln£ IN SECOrClS Case No. 2; reservoir by Westergaard's theory, horizontal input

25.000

0.000

-25,000

-^A/^»/y

1

/ V V " "

0.00 1.00 2.00 3.00 <.00 5.00 iS.OO '.00 9.00 9.00 lO.X

TIME IN SECOfCS Case No. 3; reservoir by Chopra's theory, horizontal input

O.OOO

-25.000

j

!

, . I'iJi 1^ J- t-. r 1 A. A A.. /

0.00 I.X 2.00 3.x 4.x 5.x <.X 7.X g.x 9.x 10.x

TlnE IN SECOWIS Case No. 4; reservoir empty, horizontal and vertical input

25.000

O.OX

-25.Q00 O.X 1.x 2.x 3.x <.X S.X 4.x 7.CD t.X 9.X lO.X

Tine IN secoHos Case No. 5; reservoir by Westergaard's theory, horizontal and vertical input

O.OX

-26,000 Q.X 1.x 2.x 3.x t.X 5.x s.x 7.x B.X 9.x

TinE IN SECOCS Case No. 6; reservoir by Chopra's theory (a=0.0), horizontal and vertical input

25.000 <

O.OX

-25. ox

yV.«».-v«.

O.X I.X 2.x 3.x 4.x S.X «.X 7.x S.X 9.X lO.X

TinE IN sEcaos

Case No. 7; reservoir by Chopra's theory (a=0.85), horizontal the and vertical input

Fig. 12 Comparison of measured and analytical acceleration responses in

vertical direction at the crest of the dam (the 1976 earthquake) ^-

VI-32
W\'V\/V

1.00 2.00 3.00 M.OO 6,00 7.00 Measured response

o,oso

e.oo 9,00 10,00

TIME, IN SECONDS

o,oco

-o,oso WU

Case No.

O.OC ].00 2.00 3.00 4.00 S.OO d 1; reservoir empty, horizontal input

e.oo f.oo 10.00

TlflE IN SECONDS

0.100

o.oso

-o.oso

-0. 100 e.oo 9,00 10,00 TInE IN SECONDS Case No. 2; reservoir by Westergaard' s theory, horizontal input

0, 100

0,050

0,000

-o,oso

-0. 100

0 00 1.00 2.00 3.00 4.00 5.00 d.OO 7.00 e.oo o.oo 10.00

TlflE IN SECONDS Case No. 3; reservoir by Chopra's theory, horizontal input

0.050 WW

Case No. 4; reservoir empty, horizontal and vertical input

7.00 e.oo 9.00 10.00 TldE IN SECONDS

0.100

O.OSO

0.000

-0.050

-0.100 m 9.00 0.00 10.

X

TInE IN SECONDS Case No. 5; reservoir by Westergaard' s theory, horizontal and vertical input

0. IX ~

0.050

o.ooc

-0.050

-0. IX 4m WW \AAAr

e.x c.x 10.

X

TlnE IN SECONDS Case No. 6; reservoir by Chopra's theory (a=0.0), horizontal and vertical input

0.050

0.000 WW

-0.050 m WW

-0. IX

o.x l.OO 2.x 3.x 4.x 5.x 6.a> 7.x 9.x 9.x

TIME IN SECONDS Case No.

Fig. 13

7; reservoir by Chopra's theory (a=0.85), horizontal and vertical input

Comparison of measured and analytical displacement responses in the

upstrecim/downstream direction at the crest of the dam (the 1976 earthquake)

VI-33
— .^y

0.00 1.00 2.00 3.00 H.OO 5.00 6.00 7.00 0.00 9.00 10.00 TIME IN SECONDS Measured response

0.050

0.000

-o.oso

0.00 1.00 2.00 3.00 4.00 5.00 6.00 7.00 6.00 9.00 10.00 TWE IN SECONDS Case No. 1; reservoir empty, horizontal input

0.050 "

O.OOO

-0.050 -"yW- ^ V"^ 1

0.00 1.00 2.00 3.00 4.00 5.00 6.00 7.00 6.00 9.00 10.00 nnE IN SECONDS Case No. 2; reservoir by Westergaard's theory, horizontal input

0.O5O

0.000

1

0.00 1.00 2.00 3.00 4.00 5.00 d.OO 7.00 8.00 9.00 10. QO TIME IN SECONDS Case No. 3; reservoir by Chopra's theory, horizontal input

0.00 1.00 2.00 3-00 4.00 5.00 d.OO 7.00 8.0O 9.00 10 ' TlflE IN SECONDS Case No. 4; reservoir empty, horizontal and vertical input

7?, b

0.050

0.000

-0,050

0-00 1. 00 2,00 3,00 4.00 5.00 fJ.OO 7.00 8.00 9.00 10.00 7]f€ IN SECONDS Case No. 5; reservoir by Westergaard's theory, horizontal and vertical input

0.00 1.00 2.00 3.00 4.00 5,X <!.00 7.00 6.00 9.X 10,00

TItlE IN SECONDS Case No. 6; reservoir by Chopra's theory (a=0.0), horizontal and vertical input

0-000

-0.050 0.00 1.00 2.00 3.x 4.x 5.x 6.00 7.00 8.x 9.x

Tire IN SECONDS

10.

X

Case No. 7; reservoir by Chopra's theory (a=0.85), horizontal and vertical input

Fig. 14 Comparison of measured and analytical displacement responses in the

vertical direction at the crest of the dam (the 1976 earthquake)

VI-34
15-4 Reservoir effect: ignored

(reservoir empty)

FREOUENCr (H; 15-1 Reservoir effect: ignored

(reservoir empty)

FfiEOUENCr (Hi)

15-5 Reservoir effect: Westergaard '

s

theory, water level at 219.8 m

15-2 Reservoir effect: Westergaard'

s

theory, water level at 219.8 m

15-6 Reservoir effect: Qiopra's theory

(a=0.0), water level at El. 219.8 m

.1?»V

-IB:, t-'.i

FfiEOUENCY IHjl FREIWEIICr IHll 15-3

Fig.

Reservoir effect: Chopra's theory, 15-7 Reservoir effect: Chopra's theory

water level at El. 219.8 m (a=0.85), water level at El. 219.8 m

15 Frequency response functions for dam crest acceleration obtained from

FEM analysis when the water level is El. 219.8 m

VI-35
uo.o

20.0

0.0

-20.0

-140.0 ft

0.00 2.00 y.OO 6.00 6.00 10-00 12.00 14.00 16.00 18.00 20.00 22.00 24.00

TIME IN SECONDS Horizontal Upstream/downstream component at foundation

uo.o

20.0 WWW (WW —

O.QO 2.00 li.QO 6.00 8.00 ]0.00 12.00 Hi. 00 15.00 18.00 20.00 22.00 24.00 26.00

TIME IN SECONDS Horizontal Cross canyon component at foundation

0.00 2.00 4.00 6.00 8.00 10.00 12.00 14.00 16.00 18.00 20.00 22.00 24.00 26.00

TIME IN SECONDS Vertical component at foundation

200, 0

100. 0 Itmm

0.00 2.00 4.00 6.00 8.00 10.00 12.00 14.00 16.00 18.00 20.00 22.00 24.00 26.00

TIME IN SECONDS Horizontal Upstream/downstream component at dam crest

100.0

0.0

-100.

0

0.00 2.00 4.00 6.00 8.00 10.00 12.00 14.00 16.00 18.00 20.00 22.00 24.00 26.00 TIME IN SECONDS Horizontal Cross canyon component at dam crest

0.0

-100.0 \f0J^

0.00 2.00 4.00 6.00 8.00 10.00 12.00 14.00 16.00 18.00 20.00 22.00 24.00 26.00 TIME IN SECONDS Vertical component at dam crest

Fig. 16 Plot of digitized accelerograms recorded at Yuda Dam during the

1978 earthquake

VI-36
Table 3 Natural Frequencies

of Block 5 Section

Mode

Block 5 section

Reservoir,

Westergaard's theory

1 4.65 2 9.28 3 15.01 4 17.51 5 24.76

FREOUtNCr (fe> Upstream/downstream component

at foundation

Ffl£OUENCT (B«) Upstream/downstream component

at foundation

FREOUOJCT (b) fR£OutNCt (uj)

Cross canyon component at foundation Cross canyon component at foundation

FKOElCr OU)

Vertical component at foundation Vertical component at foundation

Fig. 17 Fourier spectra of recorded accelerograms of the 1978 earthquake

VI-37
Upscream/downstream component

FREQUENCY (Hz)

Cross canyon component

le.OO 20.00 FREQUENCT (Hz)

VI-38
Measured response

200.000

lOO.OOO

0.000

-100.000

-200.000 i 0.00 2.00 6.00 8.00 10.00 1 2.00 U.OO K.OO 16.00 20.00 TIHE IN SECONDS Case No. 1; reservoir by Westergaard' s theory, horizontal input

0.00 2,00 Case No. 2; reservoir by Chopra's theory, horizontal input

0.000

-lOO.OOO ^ftm

0,00 2.00 4.00 8.00 10.00 12.00 H.M 14 00 18.00 20.00 TIME IN SECONDS Case No. 3; reservoir by Westergaard' s theory, horizontal and vertical input

200-000

100.000

-IX. 000

-200.000

iplfl iflHi

0.00 2.00 4.00 8.00 10.00 12.x 14.00 14. X 18. X 20.

X

TIME IN SECONDS Case No. 4; reservoir by Chopra's theory (a=0.0), horizontal and vertical input

Case No. 5; reservoir by Chopra's theory (a=0.85), horizontal and vertical input

Fig. 19 Comparison of measured and analytical acceleration responses in the

upstream/ downstream direction at the crest of the dam

(the 1978 earthquake)

VI-39
0.0

-100.0

0.00 2.00 1.00 &.00 8.00 10.00 12.00 U.OO 16.00 18.00 20.00

TIME IN SECONDS Measured response

100.000

0.000

-100. 000

«,^- w

0.00 2.00 4.00 ci.OO 8.00 10. X 12. W 14. X 14. X 18. X 20.

X

TlriE IN SECONDS Case No. 1; reservoir by Westergaard' s theory, horizontal input

....

0.x 2.x 4.x 6.00 e.X 10.x 12.x 14.x 16. X 18.x 20.

X

Case No. 2; reservoir by Chopra's theory, horizontal input

1

1 ^^A'Wv^^*

O.X 2.x 4.x «.X 8.x 10. X 12.x 14. X 16. X IS.X 20.

X

TinE IN SECONDS Case No. 3; reservoir by Westergaard' s theory, horizontal and vertical input

O.X 2.x 4.x li.X e.X 10. X 12.x 14. X M.X IS.X 20.x

Tltt IN SECONDS Case No. 4; reservoir by Chopra's theory (a=0.0), horizontal and vertical input

cWS*

O.X 2.x 4.x is.x e.X 10.x 12.x 14.x 16.x le.x 20.x

Case No. 5; reservoir by Chopra's theory (a=0.85), horizontal and vertical input

Fig. 20 Comparison of measured and analytical acceleration responses in the

vertical direction at the crest of the dam (the 1978 earthquake)

VI-AO
0.30

0.20

0.10

o.x

-0.10

-0.20

-0.30

Mi 441

18.00 20.x

TinE IN SECONDS Case No. 1; reservoir by Westergaard ' s theory, horizontal input

0.20C

0. IX

0.000

-0. IX

-0.2X 16.x 20.x

TltlE IN SECONDS Case No. 2; reservoir by Chopra's theory, horizontal input

0.2X

O.IX

O.OOO

-O.IX

-0.2X

16.x 20.x

TIME IN SECONDS Case No. 3; reservoir by Westergaard 's theory, horizontal and vertical input

le.x 20.x

TIME IN SECONDS Case No. 4; reservoir by Chopra's theory (a=0.0), horizontal and vertical input

16. X 20.x

TinE IN SECONDS Case No. 5; reservoir by Chopra's theory (ot=0.85), horizontal and vertical input

Fig. 21 Comparison of measured and analytical displacement responses in

the upstream/ downstream direction at the crest of the dam

(the 1978 earthquake)

VI-41
0.00

-0.10 i D.OO 2.00 4.00 6.00 B.OO 10.00 12.00 14.00 16.00 18.00 Measured response

0.100

0.000

-0. 100

^A/\/^^ vwwwv

0.00 2.00 4.00 6.00 B.OO 10.00 12.00 14.00 16.00 18.00 TIME IN SECONDS Case No. 1; reservoir by Westergaard's theory, horizontal input

0.100

0.000

-0.100

VWWIAA/ -V*. uVvA^^A \J\MMfW

0.00 10.00 12.00 14.00 16.00 18.00 20.00 TIME IN SECONDS Case No. 2; reservoir by Chopra's theory, horizontal input

0.100

0.000

-0. .00 vwwwvyW>AA/v-'

0.00 4.00 6.00 10.00 12.00 14.00 16.00 16.00 20.00 TIME IN SECONDS Case No. 3; reservoir by Westergaard's theory, horizontal and vertical input

0.100

0.000

-0. 100

/V\/V'AAA/ \A/wWW

0.00 2.00 4.00 6.00 8.00 10.00 12.00 14.00 16.00 18.00 20.00 TIME IN SECONDS

Case No. 4; reservoir by Chopra's theory (gi=0.0), horizontal and vertical input

cm 0.100

0.000

-0.100

\AAaaa/W W\/'-'»/>-—

MAAAMft

0.00 2.00 4.00 6.00 8.00 10.00 12.00 14.00 16.00 18.00 20.00 TIME IN SECONDS

Case No. 5; reservoir by Chopra's theory (oi=0.85), horizontal and vertical input

Fig. 22 Comparison of measured and analytical displacement responses in the

vertical direction at the crest of the dam (the 1978 earthquake)

VI-42
23-2 Reservoir effect: Chopra's theory, 23-5 Reservoir effect: Chopra's theory

water level at El. 231.9 m (a=0.85), water level at El. 231.9 m

Fig. 23 Frequency response functions for dam crest acceleration obtained from FEM

analysis when the water level is El. 231.9 m

VI-43
I,,,
REPORT ON

THE 1978 MIYAGI-KEN-OKI EARTHQUAKE

Makoto Watabe , . ^

-.^ Yutaka Matsushima

Yuji Ishiyama

., , Tetsuo Kubo ,

Yujl Ohashi ,- ,

Building Research Institute

Ministry of Construction

'I ABSTRACT

The damage features and the causes of damage due to the Miyagi-Ken-Oki Earthquake June

1978 are summarized as follows: ,. ,, . •

r i,.-; '. - >

1) Earthquake Ground Motions .

a) Damage spread over an area wider than those that experienced a magnitude average of

7.4. b) This fact may be explained by the fact that the source mechanism of this earthquake

has been reported as the rebound of subducting plate actions, which is illustrated in Figure

1. c) According to the accelerograms as shown in Figure 3 obtained during this earthquake in

Sendai City, approximately 100 km from the reported epicenter, peak acceleration was 0.2 -

0.3 g which caused progressive failure to low rise R.C. structures. Predominant periods for

acceleration response spectra are 0.3 to 0.4 second and 1.0 second. Figure 4 shows the

world's largest peak structural response (1040 cm/sec ) in the N-S component of the accelerogram from a recently developed SMAC-M located on the ninth floor of the R.C. Tohoku University Building, which was only slightly damaged. d) Subsoil conditions and damage features

are very closely correlated as may be seen in Figure 12.

2) Damage of Structural Members

The main causes of structural damage are as follows: a) Insufficient concrete strength,

b) Insufficient reinforcing of hoops against shear force produced in R.C. columns. c) Insufficient ductilities or strengths of connections in R.C. or steel structures, d) Poor quality

of construction, e) Lack of horizontal or vertical uniformities, i.e., torsional eccentricities and sudden changes of stiffness weight ratios along the height, as are shown in Figure

13. f) Deteriorations of structural members (especially in the case of wooden houses).

VII-1
3) Damage of Non-Structural Members

Damage of non-structural members caused various problems in this earthquake: a) Overturning of concrete block and masonry fences due to this earthquake took the lives of 14

people, or 50% of the total death casualties. b) Damage of non-structural walls due to this

earthquake remind us of the importance of earthquake resistant considerations in the total

design of a building (Photo. 1). Insufficient rigidity of main structure and imperfect

connection details of these walls to the main structures were two major causes of damage,

c) Investigations of the overturning of furniture suggests stochastic evaluation procedures

for overturning phenomena in earthquakes.

4) Overall Views

While much earthquake damage was observed and caused by this earthquake, less than one

half percent of the total number of houses and buildings in Sendai City suffered severe damage. In view of this fact as well that of the peak structural response in acceleration which

was recorded at more than IG value on the top floor of a building only slightly damaged, the

standard level of earthquake resistant capacities of the building structures in Japan proved

to be satisfactory. The two major tasks for further improvement of earthquake resistant construction: i) overall earthquake resistant consideration of the total design of buildings

and ii) avoidance of poor quality of construction will be the two major tasks.

VII-2
1. EARTHQUAKE GROUND MOTIONS . }

The source mechanism of the earthquake developed so far [4] reveals that it is a thrusttype earthquake due to the subduction of the Pacific Plate (see Figure 1). An unusually

s small attenuation of the seismic wave for this earthquake might be associated with this

fact. It is demonstrated in Figure 2 where the peak acceleration values obtained by the

strong motion accelerographs are plotted in terms of epicentral distances. In Figure 2 the

averaged attenuation relation obtained from the past 75 earthquake accelerograms is represented by a dotted line. The solid thick line represents the mean relation between peak

acceleration values (in horizontal component) and the epicentral distances due to this earthquake. Comparison of these two lines reveals that the intensity of the ground motions due to

this earthquake was extraordinary stronger than those estimated by past earthquake data, in

terms of magnitude and epicentral distance of earthquake. In Figure 3 and Figure 4 are shown

accelerograms recorded at the ground floor and the top (ninth floor) of the Structural Engineering Building, Tohoku University. The instruments which produced these accelerograms are

SMAC-Ms with magnetic tape recording systems, the development of which was brought about

through the leadership of our staff in the Building Research Institute. Among these accelerograms, it should be noted that peak acceleration in the structural response of the N-S component at the top floor of this building is the largest value (1,040 gals) ever recorded in

world history. According to these accelerograms, peak acceleration values in ground motions

are 259 gals, 203 gals and 153 gals in N-S, E-W, and U-D component, respectively. Through

integration of these accelerograms in terms of time, the time histories for velocity and

displacement can be obtained, the peak values of which are tabulated in Table 1. In Figure 5

and Figure 6 the resulting time histories of the N-S components at ground level, and top

floor level, are illustrated as examples. Acceleration response spectra of the N-S component

accelerogram at ground level are shown in Figure 7. In Figure 8, tripartite response spectra

of the same accelerogram are shown as well. These response spectra suggest that the predominant periods are 0.3 to 0.4 second and 1.0 second. With 5 percent as the fraction of critical damping, the largest response value (approximately 1,000 gals) is found at the period of

1.0 second, where the response amplification factor becomes 3.86. The fundamental period of

the building where the SMAC-M is installed on the top floor is about 0.7 second in the N-S

direction evaluated by ambient vibration. However, the predominant period of accelerogram

VII-3
in the N-S component is 1.0 second. Therefore during this earthquake, the fundamental period

became larger by 1.43 times than that evaluated by the ambient vibration. Assuming that this

building responded linearly or elastically, the peak acceleration of the structural response

at the top floor is calculated as 1,350 gals. The difference between this calculated value

of 1,350 gals and the 1,040 gals recorded at the top floor may be due to the non-linear

behavior of this structure which was subjected to slight damage on the upper stories. It is

interesting to note that the elastic displacement response value of 29.7 cm is very close to

the actual displacement response value of 25.9 cm, which is evaluated through a double intergration of the relevant accelerogram in terms of time. , .

Utilizing the N-S component accelerogram at the ground level of the Tohoku University,

synthetic accelerograms are generated for three types of subsoil conditions, i.e., bed rock,

medium hard soil and soft soil; the latter two of which are representative of local soil conditions in Sendai City. In generating the three synthetic accelerograms, the following procedures and assumptions are used:

i) Deconvolution and convolution of the accelerograms is based on Haskel's transfer

matrix method. [1] .

ii) Initial values for shear wave velocity v^ (m/sec) are estimated by the following

equation [2] in terms of N value and depth H(m) of subsoil;

r.

V„(m/sec) = 68.79 • N°-^^^ • H°-^^^

1.0

(alluvial)

1.303

(delluvial)

1.0

(clay)

1.086

(fine sand)

1.135

(coarse sand)

1.153

(sandy gravel)

.1.448

jjCgravel)

(1)

iii) Damping factor Q is assumed to be independent from frequency, and the initial

value of is assumed as 25.0.

VII-4
iv) Shear rigidity G and damping factor Q are dependent on strain level of subsoil;

starting with initial values and Q^, these values are modified in each step of time increment where the mean shear strain is assumed to be 0.45 times of the maximum shear strain in

each subsoil layer.

The results of these synthetic accelerograms are shown in Figure 9. Shown in Figure 10

are the tripartite response spectra of the three accelerograms in Figure 9, with a 5 percent

critical damping fraction. Housner's Spectral Intensity values for four accelerograms,

the original one at Tohoku University, one on bed rock, one on soft soil, and one on hard

soil are 117.0 cm, 96.3 cm, 99.0 cm, and 136.3 cm, respectively, with a 20 percent critical

damping . / . . / : .. ?

'

Incidentally, directions of "principal axes" [3] of the accelerograms at Tohoku University are N5.5°E, E5.5°S and vertical for major, intermediate, and minor axes respectively.

2. DAMAGE STATISTICS AND THEIR DISTRIBUTION

Damage due to this earthquake was concentrated to Miyagi-Prefecture, the capital city

of which (Sendai City) suffered 500 million dollar damage, about one-third of the total loss

of 1,800 million dollars in the whole Miyagi-Pref ecture.

1) Damage Statistics • •,. :

Twenty-eight were killed and 10,247 injured. Among the deaths, 3 lost their lives by

heart attack from the shock of the ground shaking, 5 inside houses, and 16 due to overturned

concrete block or masonry fences and gates. About 46.4 percent of the deaths and 90.8 percent of the injuries were concentrated inside Sendai City. . r

Also in Sendai City, the 42.2 percent total loss of 500 million dollars was to domestic

residences. Further breaking down this category, 64.4 percent is for loss of the houses

themselves, 19.5 percent for the loss of furniture and 8.7 percent for the loss of fences and

gates. , .

' . „L'

2) Damage Distribution i'

Damage was clearly concentrated in the area of soft soil conditions and in hillsides.

In Figure II, the distribution of overturned concrete block fences is illustrated. In addition to this distribution, damaged wooden houses and collapsed R.C. buildings are shown in

Figure 12. From these figures and Figure 9 of the accelerograms for local soil conditions,

VII-5
the importance of local soil and geographical features for earthquake engineering may be

seen. The damage distribution map of steel structures shows similar characteristics of

local subsoil relationships. \ -

' •;

3. DAMAGE FEATURES OF R.C. BUILDINGS

Building Research Institute investigations on R.C. buildings have been made of various

institutions: school buildings by the Tokyo University group; a building in Oroshimachi,

where several suffered severe damage; by Tohoku University; throughout Sendai City; most ';

buildings around Oroshimachi; wall R.C. apartment buildings; steel composite tall apartment

buildings; and precast, pre-stressed concrete buildings. •-

'

1 ) Damaged R.C. Structures ' " "

Damage to R.C. structures in the hard soil area, located in the central part of Sendai

City, was only slight, while R.C. structures which were subjected to severe damage are concentrated in the soft soil area in the eastern and southern parts of the newly developed area

in Sendai City. In school buildings, mean damage ratio (collapse: 100%, half collapse : 50%;

non damage : 0%) is 7 percent (after University of Tokyo group). In the Oroshimachi area

where the highest damage ratio was reported, the mean damage ratio for R.C. structure was

6 percent. The mean damage ratio for walled R.C. apartment buildings was zero. Precast and

prestressed concrete structures were subjected to very little damage. Only in the soft soil

area of the eastern part of the City, tall (9 - 14 stories) steel composite R.C. apartment

buildings suffered shear crack damage in non-structural walls, but no serious damage was

'

observed in main structural members (Photo. 1). '

Representative location and fracture patterns can be observed in the columns or walls

of ground level floors with shear failure and direction of destructive shear forces exclusively in the N-S direction. f

2) Structural Fea tures of Damaged R.C. Buildings j' •

The damaged R.C. buildings had one or more items listed as follows: ' ' '

a) Open frame structural system in the N-S directions. ; t

b) Longer span length or wider floor area per column.

c) Extreme eccentricity (eccentric ratio is more than 0.3) such as the onesided exterior walls (Figure 13).

VlI-6
d) Discontinuity of stiffness weight ratio such as a soft first-story structural system (Figure 13).

e) Insufficient spacing of column hoops which were constructed before the

revision of reinforcing details of column hoops in 1971, revealed by the damage

from the Tokachi-Oki Earthquake in 1968.

f) Low strength and rigidity of concrete materials.

3) Examples of Damaged R.C. Buildings

a) "0" Building (Figure 14 and Photo. 2). Two columns, among four, fractured

at ground level. This building is a typical "soft first-story" type on soft subsoil layers. Input earthquake energy seemed to concentrate at columns on the first

story. If the redundancy of the structure was larger (number of columns was more

than four), this kind of collapse would not have occurred. (Item of structural

feature: a, b, d, and e)

b) "M" Building (Figure 15 and Photo. 3). One by three spans 3 story R.C.

structure was situated on soft soil with eccentric core at one end. The other

end of the first-story is open framed in the N-S direction. Shear fracture

occurred in this open frame at the columns, the shear span ratio of which is less

than 2.5. (Item of structural feature: c, d, and e)

c) "P" Building (Figure 16 and Photo. 4). Two by three spans 3-story R.C.

structure on soft subsoil layers collapsed at the first floor. The thickness of

the second floor concrete slab was more than 30 cm (one foot) and spacing of

hoops was insufficient. The strength of concrete has been estimated as one-third

of the strength required. (Item of structural feature: a, c, e and f)

d) "T" Building (Figure 17 and Photo 5). One by five spans 3-story R.C.

structure on soft subsoil layers had a long span and some amount of eccentricity.

Eight among 12 columns at the first floor were completely collapsed. (Item of

structural feature: a, b, c, e, and f)

VII-7
4. DAMAGE FEATURES OF STEEL AND OTHER TYPES OF BUILDINGS

1) Steel Buildings

The amount of damage to steel buildings from this earthquake has been the highest in

number and in degree of damage to which steel buildings have ever been subjected in Japanese

history. Damage features of steel buildings which are examined herein are summarized as

follows

a) Amount of damage, i.e., very severe damage (collapse, or too large a deformation

to repair).

b) Cause of damage, i.e., effect of poor construction, especially unskilled

welding.

c) Effect of poor structural design.

d) Fracture of diagonal bracing systems.

e) Serious damage to claddings.

As for item a), one steel structure collapsed during the Izu-Oshima Earthquake of

January 1978 due to very poor quality of construction. In addition, poor quality of structural design was noted as a factor leading to the collapse. Damages due to items b) and c)

were also clearly evident. Fractures of diagonal bracings have been observed in various

past earthquakes. The problem of item e) became a serious one in this earthquake. Several

claddings, not only of steel sheet mortar finish but also of ALC (Autoclaved Lightweight

Concrete), and PC (Precast Concrete) panel walls, fell down. The collapse of a steelconstructed bowling alley with a long span of 34 meters, from vertical ground motion should

be noted here.

2) Damages of Wooden Houses and Other Damage

The destruction inflicted on those wooden houses that were severely damaged was caused

by a landslide of sorts. Some other wooden houses suffered damage because of eccentricity

or discontinuity of the stiffness weight ratio. It should be noted however that the wooden

houses with the wall strength required by the existing Building Standard of Japan survived even in this earthquake.

Toppled concrete block fences which took human lives, without exception, did not satisfy

the reinforcing details required by the Japanese Building Standard.

VII-8
5. OVERTURNING OF FURNITURE

1) Theoretical Consideration on Overturning of Bodies by Earthquake Motions

The acceleration, a, which will initiate the rocking motion to a prismatic body is given

by the following formula:

a - |g

^ <^>

where, B and H are the breadth and height of the body, respectively, and g is the acceleration of gravity.

The acceleration which exceeds the value given by the Eq. 2 does not always overturn a body, because this equation only indicates the necessary condition for overturning.

Furthermore, it is a well-known phenomenon that the larger the body is, the more difficult it

is to overturn, where the breadth-height ratio of the body remains constant, but this equation

depends only on the breadth-height-ratio and not on the dimension. In order to explain this

phenomenon, let us suppose that the body has the velocity, v, during earthquake motions and

suddenly the rectilinear motion changes to rocking motion. From the conservation of the

momentum around 0, we can have: (see Figure 18)

M V = Iq 0) (3)

where, M is the mass of the body, 1^ is the moment of inertia around 0, lo is the angular

velocity of the body. In case of the prismatic body: '

lo = I

M r2 _^ (4)

where, r is the length from the center of gravity to 0. From Eqs. 3 and 4:

CO = |JL '

• (5)

' 3 r

Then the kinematic energy, KE, at the beginning of rock motion.

KE = 1 I„ u2

2 °

1 M v^

(6)

VII-9
The potential energy, PE, when the center of gravity reached to the perpendicular line of 0:

PE = M g

^ '

. (7)

= (8)

If we use the units of centimeters and seconds for length and time, respectively, Eq . 8 will

be: . . .

V = 25.6 -5- (9)

From Eq. 9, it is possible to estimate the response velocity of the body overturned during

earthquake motions. Eq. 9 also shows that the velocity required to overturn the body is

proportional to the square root of the dimension, where the breadth-height-ratio remains

constant. - . ' ,.

2) Estimation of the Velocity from the Field Survey on the Overturning of Furniture

After the earthquake, we surveyed overturned furniture, and interviewed the inhabitants

concerning the following items.

a) Type of furniture overturned - - s •

b) Dimension of the furniture ' - - - i'

c) Direction of the overturning

d) Floor number where the furniture was placed

e) Outline of the building in which the furniture was placed

Most of the buildings surveyed were common wooden residential houses built with conventional Japanese post and beam construction. The total number of buildings surveyed was

138; 130 were two-story wooden houses, 2 were one-story wooden houses, 5 were medium- and

low-rise reinforced concrete buildings and one was a high-rise steel frame building.

The result of the survey is shown in Figure 19, where (a)®, (b)(g), (c) •, (d) o and

(e) + show the buildings where the response velocity of furniture on the first floor was

(a) more than 80 kine, (b) from 80 to 60 kine, (c) from 60 to 40 kine, (d) from 40 to 20 kine

From Eqs. 6, 7 and the relation r a = b:

V =

VII-IO
and (e) less than 20 klne, respectively. The velocity was determined by Eq . 9 and the magnification factor of the second floor velocity to the first floor was assumed to be 2.0. From

Figure 19, the following can be seen:

i) In the western part of Sendai City, where several reinforced concrete buildings

collapsed, much of the furniture overturned and the first floor velocity was more than 80

kine, whereas in the eastern part of Sendai City, almost no furniture overturned and the

velocity was less than 20 kine.

ii) Dotted part of Figure 19 shows the areas where ground surfaces are covered by

alluvium, and velocities in this area were very high compared to the other areas.

iii) In the area 120 km away from the epicenter, almost no furniture overturned. But

along the Abukuma River, the overturning was observed up to Shiroishi City and Fukushima City

which is more than 120 km from the epicenter. This is due to the alluvium deposits along the

river

.

iv) Differences of the overturning phenomena among the types of buildings are not

totally explainable, because of the scarcity of data concerning types of buildings other

than the wooden ones.

v) In the western part of Sendai City, where no furniture overturned in low-rise

buildings, much of the furniture nevertheless overturned in high-rise buildings, especially

on upper floors. We should note carefully the fact that while high-rise buildings will have

less base shear than low-rise buildings, the response velocity will be considerably magnified

on the upper floors of high-rise buildings, and overturning furniture may cause special harm

to human life.

6. CONCLUDING REMARKS

As outlined in the abstract, the standard level of earthquake resistant capacities in

the districts investigated were shown to be fairly satisfactory in this earthquake. However,

during this earthquake the importance of total systems design for earthquake resistant qualities has been emphasized here. The importance of construction qualities is also emphasized,

therefore some methods to avoid poor quality of construction should be established. These

will be the two major tasks for future earthquake engineering in Japan.

VII-U
REFERENCES

Ohsaki, Y. , "Dynamic Characteristics and One-Dimensional Amplification Theory of Soil

Deposits," Research Report 75-01, University of Tokyo, August 1975.

Ohta, Y. , "Establishing Correlations Between the Indices of Soil Properties and Shear

Wave Velocity," Shizen Saigai Shiryo Kaiseki 4, 1977.

Watabe, M. , Penzien, J., "Simulation of Three Dimensional Earthquake Ground Motions,"

International Journal of Earthquake Engineering, 1975.

Seno, T., Shimazaki, K. , Somerville, P., Sudo, K. , Eguchi, T. , "Rupture Process of

Miyagiken Oki Earthquake of June 12, 1979," Physics of the Earth and Planetary Interior

(in press).

VH-12
Tab. 1 Maximum Values of Vibrations At TOHOKU UNIV. acceleration velocity displ

.

max. (cm/sec2) time max. (cm/sec) time max. (cm) time

NS 259.23 7.56 36.17 10.8 -14.53 10.56

G.L. EW 202.57 3.10 27.57 2.96 9.11 3.82

UD 153.04 4.18 11.92 10.44 3.18 10.02

NS 1 ,039.98 15.20 150.41 14.96 -25.87 15.20

9FL EW -523.92 14.08 -72.76 14.34 16.61 3.40

UD -355.56 7.32 22.92 15.48 7.09 8.18

B June 12 1978Tohoku Uf.i>'. I

Fig. 1 Source Mechanism of This Earthquake

Hypocenter is located on the boundary of subducting plate

O: standard deviation

1-

fpicentral distance (km) "Amax-10"" ^' ' (cm/sec^) Fig. 2 Relation Between

Max. Accel, and Epicen

Distance

VII-13
0 5 10 15 20 sec

Ma>:.259gals ( N-S Component)

Pig. 3 Accelerograms at Ground Level

Installed at Tohoku University

1§00

) 5 10 15 20 sec

1 Q -1000

' ' -.w^- -v • * 1 - ; ^

3 5 10 15 20 sec

Max. 1040 gals world largest record

Fig. 4 Accelerograms at 9th floor

Installed at Tohoku University

max value -259. 23UJ at 7. 5600sec

^ Osec 5 10 15 20_

max >alue-36. 1C98 at 10.9000sec

nOsec 5 10 15 20

mai value - IJ. ".239 at 10. DCOOsec Fig. 5 Time Histories of Vibrations at Ground Level

N-S Component

ACCELERATION

VELOCITY

DISPLACEMENT

VII-14
ACCELERATION

VELOCITY

DISPLACEMENT

max value -25. 8677 al 15. 2000sec Fig. 6 Time Histories of Vibrations at 9th Floor

N-S COMPONENT

S 3200 damping

:

0%,1%,5%,

10% ,20%.

-lOOOcm/sec'

"T5~ 100

period (sec) Fig. 7 Response Spectra of

Accelerogram N-S Component

at Ground Level Univ.Tohoku

damping

o%,l%,5%, 10%

and 20%

0.01 c 0. 1 O 1.0 O 10 -'100 V o

period (sec)

Fig. 8 Tripartite Response

Spectra for Fig. -7

VII-15
20 25 30 35_ mm

original at Univ. of Tohoku

- 2j8gal ^ 30 at bed rock

-iWRai at medium hard soil

Fig. 9 Synthetic Accelerograms for Different

Local Soil Conditions (N-S Component)

" med.hard

' bed rock

,' soft ,

u.ni <? 0. 1 (7 1.0 o 10 loo

Fig. 10 Response Spectra

for Fiq.-9 in Tripartite.

5% damping

VII-16
o

(0up

x: o (U Cnc w

(0 w

u c

U-l

<*-! •HP

1

! i - y . ^4 G-(l.8

J^- -^5" ^ 0=11. '1

^

1

C=0.8

damage

—collapse

•-severe^

3-inodgrate ^ non

eccentric ratio

VII-17

Fig. 13 Damage Feature

in Relation with Ecce

ntricity and Stiffness

Weight Ratio
Photo. 1 Damage of Non-structural

Walls in R.C.SSteel Compo

site Apartment Build.

Phto. 2 Collapse of "0"Building

due to soft first story

VII-18
Photo. 3 "M" Buildina ,- collapsed due to torsion

1 1

1p

E-SIDE

r

7\

i= Hi

k

\l3 L_

w--SIDE

21.000

-IH.

Plan

Fig. 15.

PlanSElev.

of "M"Build.

Photo. 4 "P"Building:

collapsed due to torsion spoor construction

VII-19

• — 1

1 ' 1 i

1 Plan E-SIDE

W-SIDE

Fig. 16 Plan&Elev.of

'"P" Building.
25.000.

1

W-3IDE

Fig. 17 Plan and elevation Fig. 18 Rocking of a body

of the "T" building

Photo. 5 The "T" building

VII-20
'X./ lAO Km FROM EPICENTER

Velocity Distribution Estimated by Overturning of Furniture

in THE- 1978 MIYAGI-KEN-OKI EARTHQUAKE

VII-21
DISASTROUS GROUND FAILURES IN A RESIDENTIAL AREA OVER A

LARGE-SCALE CUT-AND-FILL IN THE SENDAI REGION

CAUSED BY THE EARTHQUAKE OF 1978

Hitoshi Haruyama

Motoo Kobayashi

Geographical Survey Institute

Ministry of Construction

ABSTRACT

This paper examines the relationship between the disastrous ground failures in a hilly

residential area and the land properties, due to the Miyagi-Ken-Oki earthquake of 1978.

We measured the change of landforms by comparisons of two maps based on aerial photographs taken before and after major area earthworks. As the first step, we prepared a landform map at a scale of 1/2,500 with a 2 meter contour interval, from a time just before the

earthworks for residential development (1957) and after the earthworks (1978).

Then we constructed a thickness isopleth map of the artificial fill with a 2 meter

interval and many profiles across the earthworks.

Then we examined the interrelationships between the distribution of ground cracks, the

damage to destroyed houses, the retaining walls, the maps described above, and the original

landforms

.

KEYWORDS: Artificial fill; disaster prevention; earthquake disaster; ground failure; slope

steepness.

VII-22
1 . INTRODUCTION

The damage caused by the earthquake of 1978 in the Sendai region presented new problems

of disaster prevention in an urbanized region. One of these problems is disastrous ground

failure in hilly residential areas.

This paper is based on a study which is linked to co-operative research projects on the

Miyagi-Ken-Oki earthquake which are encouraged by the Science and Technology Agency.

We selected two severely damaged areas for study, namely, the Midorigaoka, and the

Kuromatsu districts in the Sendai region.

2. MEASUREMENT OF ARTIFICIAL FILL BY AERIAL PHOTOGRAMMETRY

2.1 Landform Maps for the Time Before and After the Earthworks

The earthworks for residential sites in Midorigaoka and Kuromatsu districts were carried

out from 1957 to 1966.

Landform maps at a scale of 1/3,000 had been made before the time of the earthworks, but

with an insufficient accuracy to the contours. Consequently, for this study we constructed a

new landform map at a scale of 1/2,500 using aerial photographs that were taken in 19 57 by

the Sendai Municipal Office. For more recent conditions, we used a national large-scale map

of 1/2,500 scale which was prepared in 1977-78 by the Geographical Survey Institute.

In this mapping, in order to enhance the accuracy of the 2 meter contours, the number

of points plotted was as much as twice the number used in ordinary mapping. The mapping

area covered 2 km^ of the Midorigaoka district and 3 km of the Kuromatsu district.

2.2 Accuracy of the Isopleth Map of Artificial Fill Thickness

By overlaying the two landform maps, the points where contours of the same altitude

intersected each other could be located and lines drawn to connect the intersect points in

order to determine the boundaries of cut and fill.

Next, to depict the thickness isopleth of artificial fill, we connected the points of

intersection of a given contour of (A) meters on the landform map in 1957 and the (A + 2)

meters contour of the landform map in 1977-78.

Similarly, we connected the points of intersection of (A) and (A + 4) meters, (A + 6)

meters .... (At 2'^) meter contours (Figure 2).

VII-23
Concerning the accuracy of the thickness isopleths, as map contours generally have a

maximum permissible error of 1/2 of the contour interval, and height points have a permissible error of 1/3 of the contour interval, we may suppose that the thickness isopleths

have a contour error of ± 2 meters or, at the lowest estimate, ± 1.34 meters of height error

Boring survey data in the Midorigaoka and Kuromatsu districts were obtained from the

earthquake damage reconstruction plan by the Sendai Municipal Office and the Miyagi Prefecture Office, 34 boring pits having been located on artificial fill.

The mean value of the difference between the thickness of the fill, as shown by the

boring data and the isopleth map at the same point, was ± 1.3 meters. Two values within

the 34 boring pit survey data had a difference of 4 to 5 meters compared with the value on

the thickness isopleth. Except for these two values the mean value of the difference was

± 1.1 meters. The thickness isopleths for the excavated areas were not estimated.

2.3 Distribution Map of the Disastrous Ground Failure ' '

'

The acceleration of strong-motion in Sendai City and its environs during the earthquake is shown in Figure 4. The extent of casualties, and the ground failure in Sendai and

Izumi City are shown in Table 1. " '

From the data on the distribution of cracks, the collapse or swell of retaining walls,

and destroyed, and semi-destroyed houses which prepared by the Sendai and Izumi Municipal

Offices and complemented by our field survey, we depicted the overlay situation on the

thickness isopleth map of artificial fill, and the landform map in the time of 19 57 previously mentioned. :

In order to examine quantitatively the relationships between the disastrous ground failure and the land conditions, we constructed a grid of 5 mm units on the maps at a scale of

1/2,500, and measured the situation on unit areas of 1.56 hectares (12.5 meter x 12.5 meter)

dimensions

.

VII-24
3. RELATION OF THE DISASTROUS GROUND FAILURE TO

THE LAND CONDITIONS IN HILL RESIDENTIAL AREAS

3.1 Characteristics of Landform and Geology of the Surveyed Areas

The Central part of Sendai City is located on terraces consisting of Upper Pleistocene

sand and gravel layers having a thickness of several meters underlain by Pleiocene tuff

(See Figure 3).

This material proved to be rather stable, so that the damage of the earthquake was

relatively small.

In the last twenty years, the city area has expanded, and urban construction has spread

over the softer ground of an alluvial plain in the eastern part of Sendai, and over the hilly

areas of northern and south-western Sendai where the landform has been extensively changed

by earthworks. In these newly developed areas, the earthquake damage was conspicuous.

The hills in Sendai and its environs have a rather uniform summit level of about 100 to

200 meters, declining to the east.

Aobayama Hill in southwestern Sendai consists of horizontally-bedded Pleiocene sandstone

tuff, silstone, etc., covered by Middle Pleistocene gravel and aeolian volcanic ash. On

Aobayama Hill, there has been extensive development of residential and educational districts

namely Midorigaoka, Aoyama, Matsugaoka, and Yagiyama.

Midorigaoka district is on the southeastern part of Aobayama Hill, whose eastern margin

is a flexure scarp that is part of the Rif u-Nagamachi structural line bordering the alluvial

plain. Several hundred meters west of the scarp the Dainenji reverse fault runs sub-parallel

to it, the Intermediate zone being uplifted like a horst.

Nanakita Hill occupies the northern part of Sendai and the adjacent part of Izumi City.

The hill consists of Upper Miocene sandstone, taffaceous siltstone and Lower Pleiocene sandstone, tuff, without overlying gravel. The relief of the hill is relatively small but it is

rather dissected. The area has been developed widely for residential districts, namely

Kuromatsu, Asahigaoka, Nankodai, Tsurugaya.

3.2 Relation of the Types of Artificial Fill and Disastrous Ground Failure

In former years, before large-scale cu ts-and-f ill of hilly regions for residential

districts were practiced in Sendai City, some houses were built on the slope of hills,

usually with rather small modifications of the ground surface. Such houses, for instance,

VII-25
those in the Kano-Honcho neighborhood northeast of Midorigaoka, that are located on the

surface of the flexture scarp with a slope of 10 degrees, were only slightly damaged by

the earthquake.

Representative landform profiles of the destroyed houses and earth cracks are shown in

Figure 5. The distribution maps of thickness isopleths and disastrous ground failures are

shown as Figures 6 and 7. Areas of artificial fill are divided into two main types: "hillside filled" and "valley filled." In general, the thickness of the fill at the former type

is less than 10 meters. On the later type along the valley line, where the fill is thickest,

the maximum thickness reaches 21 meters inside the surveyed areas. However, the damage due

to ground failure is not proportional to the thickness of the fill, but rather it is related

to a complex of fill thickness conditioas and the original slope steepness of the buried

landform.

In many cases, cracks break out in the part of the fill area where there are steep

slopes on the buried landform, often running sub-parallel to the direction of the hillside,

valley-head, or valley-sides of the buried land surface.

On the hillside filled type, at the place where there was a small valley in the buried

landform slopes, ground failure occurred as echelon type cracks and the mass of fill material

sloped down slightly. Also on the hillside-filled type, ground failure was more severe over

buried lower slopes than over buried upper slopes.

At the boundary zone between the filled area and the cut area, considerable ground

failure also occurred. This zone was only roughly located, but from the appearance of

damage it is conjectured to be the area of cut and fill. It seems that a filled area

responds more as soft ground to earthquake vibrations than does the cut area.

3.3 Several Attributes of Land Condition Related to the Disastrous Ground Failure ,,.

Disastrous ground failures in the artificial fills and their relation to the land

condition were examined as follows.

Frequency distribution for thickness of artificial fill in each or all districts as

measured at the intersection of the grid is shown in Figure 8. There are slight differences

between the districts but most of the fill is less than 4 meters thick as a whole.

VII-26
The ratio relationship of destroyed houses to the thickness of fill by types is shown

in Figure 9. The ratio is calculated as follows: fully-destroyed houses + 1/2 semi-destroyed

houses/all houses. A unit consists of a 5 mm diameter circle.

In general, the greater the thickness of the fill, the greater the ratio of destroyed

houses. It is possible to see a tendency for an increase in the ratio of destroyed houses

as the slope of the original landform increases.

Inter-relations seem to exist between the thickness of the artificial fill, the slope

angle of the original landform, and the ratio of destroyed houses for a part of Midorigaoka

district (Figure 10). Two groups can be distinguished, a group on the upper right and a

group on the lower left. The latter group where the thickness of fill is less than 8 meters

and the slope of the original landform is less than 25 degrees is one with a high percentage

of destroyed houses.

There is a relationship to a specific catchment area, which is an indicator of a comparatively large water gathering area on the original land surface. In the area of low slope

angle on the original landform, and high value for the specific catchment areas, ground

failures commonly occurred. (Figure 11).

About the disasters on the artifical fill areas in the hilly region, there are other

problems to be examined such as the situation for weathering status of the basement rock ,

and the existence of debris on the slope of the original landform.

The characteristics of the earthquake disaster in the artificial fill areas in the

Sendai region are instructive regarding possible future disasters in other urbanized regions,

and suggest that more intensive land condition surveys may contribute to avoiding such

disasters.

CONCLUSIONS

From our analysis we conclude:

1) Severe damage due to ground failure is concentrated on the areas of artificial fill,

rather than on the excavated areas. However, it is not definitely proportional to the thickness of the fill.

2) The damage is related to the original slope steepness underneath the fill, and to

the type of fill situation.

VII-27
3) Damage in the artificial fill area is concentrated at places where the original

land surface had a high ratio of specific catchment areas. •,

ACKNOWLEDGMENT

We must acknowledge the help and assistance in the preparation of much data by the

Earthquake Affairs Office of Sendai City and the Management Division of Izumi City. We

appreciate the cooperative aid and advice by Mr. S. Hatano and Mr. H. Sunage of the

Geographical Survey Institute. - i r^.

REFERENCES i

: ,

.

[1] Asano, T. , "Transformed Hill Surfaces Around Sendai," Ann. Tohoku Geogr. Assoc.,

Vol. 31, No. 2, 1970.

[2] Okutsu, H. , "Damage Caused by Miyagi-Ken-Oki Earthquake and the Corresponding

Ground Geological Specification," Tsuchi To Kiso (Soil and Foundation), Vol. 26, No

12, 1978. , ...

[3] Kawakami, F., Asada, A., and Yanagisawa, E., "Damage to Embankments and Earth Struc

tures," Tsuchi To Kiso (Soil and Foundation), Vol. 26, No. 12, 1978.

[4] Speight, J. G. , "A Parametric Approach to Landform Regions," Inst. Br. Geogr. Spec.

Publ. No. 7, 1974.

[5] Tamura, T. , Abe, T., Miyagi, T. , "Earthwork for Residential District in Hills and

Damage Caused by Earthquake," Sogo Toshi Kenkyu No. 5, 1978.

[6] Tohoku University, "The Science Reports of the Tohoku University," Second Series

(Geology), Vol. XXV, 1953.

[7] Institute of Geology and Palentology, Tohoku University, "Phenomena and Disasters

Associated with the Miyagi-Ken-Oki Earthquake of 1978 in the East-central Part of

Northern Honshu, Japan," Contribution No. 80, 1979.

[8] Nakata, T. , "Damage of Houses and Landform Condition in Sendai and Its Environs

Caused by Miyagi-Ken-Oki Earthquake," Chiri (Geography), Vol. 23, No. 9, 1978.

VII-28
Fig. 1

Fig. 2 large scale cuts-and-f ills

,

a) Landform in 1957, b)

Landform in 1978, c) Thickness isopleth of artificial

fills
Fig. 4 Acceleration of Strong Ground Motion around

Sendai

Sendai City Izumi City

All Midorigaoka Kuromatsu All Kuromatsu

Dead person 13 0 0 2 0 Serious wound p. 170 18 Slight wound p. 9,130 801 Destroyed house 769 36 29 94 17 Semi-destroyed h. 3,481 128 47 305 38 Partly-destroyed h. 7A,000 11,856 769 Retaining wall swell 482 159 22

" collapse 67 13 7

} 1,398 -

Landc rack, Subsidence 199 119 8 2,328 Mudf low 25 1 104

Table 1. Damages due to Miyagi-ken-oki earthquake of 1978

VII-30
Fig. 5 Representative profiles showing the relation between ground conditions

and disasters

1) Destroyed house

2) Semi-destroyed house

3) Non-destroyed house

4) Crack

5) Collapse of retaining wall

6) Present ground surface

7) Original ground surface

VI 1-31
Fig. 6 Isopleth map of thickness of artificial fill and the

disastrous ground failure (Main Part of Midorigaoka

district)

1) Destroyed House 4) Collapse, swell of retaining wall

2) Semi-destroyed House 5) Isopleth of thickness of artificial

3) Crack fill in 2 meter interval

VII-32
Fig. 7 Isopleth map of thickness of artificial fill and the

disastrous ground failure (A part of Kuromatsu district)

1. Destroyed House 4. Collapse, swell of retaining wall

2. Semi-destroyed House 5. Isopleth of thickness of artificial

3. Crack fill in 2 meter interval

VII-33
0 2 4 6 8 10 12 14 16 18 m

1 I I I I I I I I I 2 4 6 8 10 12 14 16 18 20

Fig. 8 Frequency distribution of thickness of

artificial fill in each districts

40 ~ 45°

35 ~ 40''

30 ~ 35

25 ~ 30'

20 ~25

15 ~ 20'

10- 15'

5 ~ 10'

0 - 5

"

• • • •

Percentage of

destroyed house

' 109i<

' 10~20

* 20 ~ 30

* 30 ~ 40

* 40-50

' 50-60

603$>

Thickness of Artificial fill

0 2 4

1 I I 2 4 6

18

20

20 r

Fig. 10 Interrelation between destroyed houses ratio,

thickness of artificial fill and slope degree

of original landform (Midorigaoka 1-chorae)

Fig. 9 Relation of destroyed houses ratio and

thickness of artificial fill for some types

VII-34
Fig. 11 Interrelation between disastrous ground failure, slope

degree of original landform, and specific catchment

area (Midorigaoka 1-chome)

I Destroyed house A Semi-destroyed house

;

. • Crack + Collapse, swell of retaining wall N Not damaged

5 Thickness of artificial fill

(5) Valley filled type

5 Hill-side filled type

VII-35
DAMAGE FEATURES OF CIVIL ENGINEERING STRUCTURES

DUE TO THE MI YAG I -KEN-OKI EARTHQUAKE OF 1978

Tadayoshi Okubo

Masamitsu Ohashi

Toshio Iwasaki

Kazuhiko Kawashima

Ken-ichi Tokida

Public Works Research Institute

Ministry of Construction

INTRODUCTION •

At 17h 14m (JST), June 12, 1978, a destructive earthquake took place offshore Miyagi-ken

and caused extensive damage to buildings, highway facilities, river dykes, water supply,

sewage systems, electrical and gas supply systems, and others.

Major observed damage characteristics shows: (i) the influence of subsoil conditions was

great, and (ii) losses in commercial and industrial sectors and residential housings comprise

a high percentage of the total damage amount.

This paper describes the outline of the earthquake, the earthquake history in the

affected area, the geological and subsoil conditions, the recorded ground motions and damage

features to civil engineering structures which were caused by the earthquake, especially

bridge damage. '

j

( OUTLINE OF EARTHQUAKE |

'

Origin Data

Origin data announced by the Japan Meteorological Agency (JMA) [2] is as follows:

Origin time ^ 17h 14m, June 12, 1978 in Japanese Standard Time

Epicenter 38° 09' N, 142°12' E

Focal Depth 30 km

Magnitude 7 . 4 on the Richter Scale

This earthquake is named the Miyagi-ken-oki Earthquake of 1978 by JMA and was accompanied by a foreshock which occurred eight minutes prior to the mainshock, and numerous

VlI-36
aftershocks. The number of felt aftershocks announced by JMA is 11, between zero and 24

hours after the mainshock, 8 between 24 and 48 hours, 3 between 48 and 72 hours, 2 between

72 and 96 hours, and 1 between 96 and 120 hours. •

' " .

The same area was also struck by an earthquake of magnitude of 6.7 on the Richter Scale

on February 20, 1978, and suffered minor damage. The epicenter of this earthquake was

located at 38°45' N and 142°12' E, about 60 km north of the epicenter of the June 12 Earthquake .

Seismic Intensity • ^ - ' ^ t > m; . .

•

The intensities of ground motions at various sites announced by JMA are shown in Figure

1 together with the location of epicenter. Numerals in the figure are seismic intensities

on the JMA intensity scale. A comparison of JMA and Modified Mercalli scales is shown in

Table 1

.

Source Mechanism

Observation of aftershocks was carried out by JMA, Tohoku University, and other institutions. Figure 2 shows a distribution of aftershocks reported by T. Masuda [3] in the period

from June 13 to 30. In the figure, epicenters of the February 20 Earthquake and the mainshock of the June 12 Earthquake are also shown.

Seno, et al. [10] proposed a fault model as shown in Figure 3. In this model, the area

of the fault plane is 80 km long and 30 km wide and a dip angle of 20 degrees. From an

analysis of surface wave amplitudes, they concluded that the relative movement of the fault

plane was about 1.7 m. ;

• :

" c. . : :>'-

GEOLOGY AND SOIL CONDITIONS

Topographic Features of Tohoku Region - , , -

'

'

As shown in the topography of Tohoku Region in Figure 4, three approximately parallel

mountains running north to south characterize the topography of that region. Plains and

basins are located between these mountains. . '

...

'

The Kitakami Mountains in the north and the Abukuma Mountains in the south which form

the east line of the mountains mainly consist of Paleozoic and Mesozoic formations. The

Ohu Mountain Range is the central range and is called the "backbone range in Tohoku." This

Range mainly consists of Neocene sediments and Tertiary and Quaternary volcanic rocks.

VII-37
Dewa Mountains and Ashai Mountains form the west line of the mountains. The formations of

Dewa Mountains are similar to those of Ohu Range. The Asahi Mountains mainly consist of

Paleozoic and Mesozoic systems. Plains and basins between the three mountains mainly consist of Pleistocene and Holocene formations. . -j , ,.-

. : ' V'l ' - Neocene Systems in the Vicinity of Sendai

Table 2 shows Neocene stratigraphy in the vicinity of Sendai. The table is cited from

the "Data Book of the Nature in Tohoku Region" published by the Tohoku Regional Construction

Bureau, Ministry of Construction [5] and is modified by the authors [9]. i'v

Quaternary System and Topography in the Vicinity of Sendai

Concerning the Quaternary System in the vicinity of Sendai, there are many existing

studies but some discrepancies exist among them. In this section, the Quaternary System and

topography are sketched, drawing primarily from "Ground in the Coastal Area of Sendai Bay

(196 5)," published by the Planning Bureau of the Ministry of Construction, and the Miyagi

Prefecture [6]

.

In the early Quaternary, the Sendai area was exposed above sea level and was subjected

to erosion which resulted in forming an eroded plain. Also in Quaternary, geological formations and topography are characterized by the alternative repetition of glacial and interglacial epochs which resulted in the forming of alluvial fans and terraces. 'Ji

Figure 5, Figure 6, and Table 3 show a geological map, a typical geological section,

and the Quaternary stratigraphy in the vicinity of Sendai, respectively. Sendai area can

be topographically classified into five zones as follows (Figure 6).

1) Mountain Zone: This zone is located west of Sendai Plain. Its elevation is more

than 150 m. It mainly consists of andesite and aglomerate.

2) Hilly Zone: This zone is located north, west and south of Sendai and north of

Sendai Plain. Its elevation is about 60 to 210 m around Sendai, 20 to 40 m west of Natori

and 10 to 40 m west of Shichigahama . It is mainly consists of Neocene mud stone, tuff,

shale, and partially Triassic slate. ... _:,:> . • ; . i'. i-.vyr:

3) Terrace Zone: This zone is located in the valleys of Nanakita River, Hirose River,

Natori River, Abukuma River, Naruse River, Yoshida River and Kitakami River. It consists of

four or five terraces, varying with the location. Sendai City has five terraces, i.e.

,

VII-38
(i) Aobayama Terrace (elevation 100 to 210 m) , (ii) Dainohara Terrace (elevation 40 to 90 m)

,

(iii) Kamimachi Terrace (elevation 25 to 65 m), (iv) Nakamachi Terrace (elevation 25 to

60 m) , and (v) Shimomachi Terrace (elevation 25 to 50 m) . '" '

4) Plain Zone: This zone is located east of Shimomachi Terrace. Its elevation is 1 to

15 m. Stratigraphy is shown in Table 3 and Figure 6.

5) Sand Dune Zone: This zone is located along the coastal line and makes two lines.

Between the dune lines, marshes develop.

The most noticeable fact from the viewpoint of the earthquake damage is the existence

of Nagamachi-Rifu Fault. It is considered that this Fault moved during late Pleistocene and

early Holocene. As shown in Figure 5, the depth of baserock from the ground surface is about

5 to 7 meters in the west of this Fault but sharply increases up to 30 to 35 meters slightly

east of the Fault and reaches about 60 meters in the coastal zone.

'

The Quaternary system mainly consists of Pleistocene deposits (diluvium) in the west of

the Fault, but in the east consists of diluvium and Holocene deposit (alluvium). As shown

in Table 3, the alluvium in this area contains soft layers. It is believed that these differences caused differences in the earthquake ground motions and in the severity of structural damage.

RECORDED EARTHQUAKE ACCELERATIONS

During the Miyagi-ken-oki Earthquake of June 12, 1978, 191 strong motion accelerographs

recorded earthquake accelerations. The maximum ground acceleration was recorded at Kaihoku

Bridge, about 80 kilometers from the epicenter and was 287 gals. Minimum acceleration was

recorded in Nagoya, about 600 kilometers from the epicenter and was 6 gals.

Recorded maximum accelerations in three directions in the Tohoku Region are shown in

Figure 7. From the figure, it can be estimated that the maximum ground acceleration near

Sendai and Ishinomaki was approximately 240 to 300 gals. Along the Pacific Coast of Iwate

Prefecture, it was about 100 to 170 gals.

Maximum ground accelerations in three perpendicular directions are plotted against the

epicentral distances in Figure 8. Acceleration records on the ground and on a pier top at

Kaihoku Bridge site are shown in Figure 9. Although very high accelerations (more than 500

gals in the longitudinal direction) were triggered on the pier cap, Kaihoku Bridge did not

sustain any structural damage.

VII-39
Next, Figure 10 shows a record at a pier cap on the Date Bridge located near Fukushima

City (epicentral distance of 160 km). The maximum accelerations on the pier cap were 480

gals in the longitudinal direction, and 320 gals in the transverse direction. This bridge

suffered moderate damage to bearing supports and to a truss member above the fixed bearing.

Since frequency properties change at 10 sec after the start of the record as shown in Figure

10, it is supposed that the bearings at the pier top might have been damaged at that time.

DAMAGE TO HIGHWAY BRIDGES

Features of Bridge Damage [1], [8] ;

Figure 11 shows the locations of severely damaged highway bridges (black circles),

places where liquefaction was observed (white circles), and the outline of geological conditions in Miyagi Prefecture.

Numerals near black circles coincide with the number of bridges listed in Table 4, which

describes briefly the type of damage to highway bridges. It is seen from Figure 11 that most

of the major bridge damage and liquefaction took place in alluvial lands along large rivers

such as Kitakami, Naruse, Yoshida, Natori, and Abukuma. , .j„,,„ .;

Sendai Bridge [4]

Sendai Bridge, completed in 1965, is located in the south part of Sendai City, and

crosses the Hirose River as a part of National Highway No. 4. A general side view is shown

in Figure 12. Superstructures are 9-span, simply supported, composite steel-plate girders,

with span length of 9 x 33.840 m, total length of 310 m, and width of 19 m. Substructures

are T-shape columns (6.1 m high) founded on rigid well foundations (9 to 18 m deep) embedded

rather stiff sands. Bearing supports are the line bearing type. Since this highway which

connects the Kanto and Tohoku regions is an important one, Sendai Bridge carries very heavy

traffic (54,000 cars daily). As shown in Figure 12, the lower half of the column height is

embedded into the higher river bed at three piers (P6 to P8). Column bases are above the

surface of the lower river bed at the other piers (PI to P5)

.

Due to the earthquake (the epicentral distance to the bridge is A = 120 km), all of the

nine pier columns sustained damage. Piers 1 through 4 cracked horizontally at the column

bases, and surface concrete pieces separated severely from the columns near the bases.

Piers 5 through 8 had similar damage near the haunches which connect columns and beams of

VII-40
the piers. Pier 6 which has the lowest free height sustained the severest cracking at both

sides (see Figure 13). concrete pieces separated at the haunch and reinforcing bars buckled.

Near the haunch, the volume of reinforcing bars as well as the concrete sectional area

changes rapidly. It is estimated that relative displacements between adjacent girders were

I to 2 . 5 cm on the pier caps and that displacements at the pier caps of Piers 1, 2 and 6 were

II to 18 cm. The girders and the bearing supports did not sustain any damage.

Figure 14 shows temporary frame works supporting the girders near Pier 6. Since the

bridge is very important, damaged piers were repaired without stopping traffic even for a

short time. Figure 15 illustrates an example of permanent repair work at Pier 6. Figure 16

is a picture of Pier 6 after the repair work was finished. The thickness of added concrete

was 50 to 70 cm, and vertical reinforcing bars were fixed using epoxy adhesive the well foundation, lateral bars were fixed to the columns, and chemical resin was placed into small

cracks. It took only one month to completely repair all the damage to this bridge.

Kln-noh Bridge [7]

Ken-noh Bridge, completed in 19 56, is on National Highway No. 346 which crosses the

Kitakami River. As shown in Figure 17, the superstructures of the bridge are single-span

steel plate girder, 5-span simply supported steel trusses, and 9-span Gerber-type steel plate

girders, from left to right. The total length and the width are 575.5 m and 6.0 m, respectively. Substructures are RC columns on caisson foundations for the truss spans, and RC

columns on footing foundations with RC piles for the Gerber plate girder spans. Soils are

of soft silts and sands, and a firm sand layer exists approximately 30 m below the ground

surface. During the June earthquake one suspended girder of this bridge fell down (see

Figure 18). The Kin-noh Bridge is the only one that fell down during the June Earthquake.

The bridge was damaged three times by three different earthquakes, namely the Northern

Miyagi-ken Earthquake of 1962 (M = 6.5, A ^ 15 km), two Miyagi-ken-oki Earthquakes of

February 20, 1978 (M = 6.7, A = 80 km) and the one of June 12, 1978 (M = 7.4, A = 110 km).

Due to the 1962 Earthquake the side blocks of the bearing supports (oval line bearings)

of Gerber girders failed, and concrete near the fixed bearing supports on the right-bank

abutment cracked. After the 1962 Earthquake, repair work to add stiffening plates was undertaken at three piers (P8, P9 , and PIO) as shown in Figure 19.

VII-41
Anchor bolts of the bearing stiffening plates were cut off during the Earthquake of

February 20, 1978. Side blocks of bearing supports, which did not sustain damage during the

1962 Earthquake and therefore were not repaired, also failed during the February Earthquake

(see Figure 20).

During the Earthquake of February 20, 1978, most bearing supports at the truss girders

also failed as did the bearing supports at the Gerber girders. As for the truss spans, fixed

bearing supports of pin-type were most severely damaged on Pier 6 (see Figures 21 and 22).

Figure 21 shows the upstream support where four anchor bolts pulled out and bent counter- •

clockwise. It is supposed from Figure 21 that the bearing would have rocked severely,

rotated, and translated. Figure 22 is the downstream support in which set bolts were sheared

off. Most anchor bolts of the fixed bearing supports on the other piers were also pulled

out. Movable bearing supports of the pin-roller-type also failed. Figure 23 shows the protrusion of all the rollers at the upstream movable bearing on Pier 5.

Since there were only four months after February 20 Earthquake, repair work on these

bearings was still being undertaken at the time of the June 12 Earthquake. Accordingly, all

the girders were able to move freely without restraint by bearing supports during the June

Earthquake.

Due to the June Earthquake a suspended girder between Piers 7 and 8 fell down onto the

river bed, as shown in Figures 17 and 18. The superstructure moved toward the right-bank

side by 55 cm on the top of Pier 8 (see Figure 24). All the Gerber span between Pier 8 and

the Right Abutment moved toward the right-bank. The girder moved 10 cm toward the right on

the right-bank abutment, and the end of girder collided into the parapet of the abutment

(see Figure 25). The asphalt pavement of the backfill heaved due to the collision (see

Figure 26).

Truss girders were also heavily damaged during the June Earthquake. Figures 27 and 28

are pictures taken after the June Earthquake at the same places as Figures 21 and 22, respectively. Anchor bolts of the upstream fixed bearing at Pier 6 were severely pulled out by

about 20 cm at most (compare Figures 29 and 19), presumably due to the rocking and translation motions of the bearing, and some concrete underneath the lower bearing plate was taken

out and the bearing sunk by 2.5 cm. As for the downstream fixed bearing on Pier 6 (Figure

28), a deformed bar which had been used as a temporary set bolt after the February Earthquake

was sheared off again. The key of the upper shoe dislodged from the sole plate, and the sole

VII-42
plate deformed. As for pin-roller-type movable bearings, some rollers had rolled out of the

shoes during the February Earthquake. Figure 29 shows the condition after the June Earthquake at the upstream movable bearing on Pier 5 where the rollers completely rolled out.

As for pier columns, only the right-bank side of Pier 8 sustained heavy cracks (see

Figure 30). It is estimated that these cracks would have taken place when the superstructure

collided with the right abutment and the reaction toward the left bank applied to the pier.

To grasp the causes of damage to the Kin-noh Bridge, the Miyagi Prefecture is conducting

comprehensive studies which include field surveys and dynamic analyses, and the cooperation

of the Public Works Research Institute.

Maiya Bridge

Maiya Bridge, completed in 1928, crosses Kitakaml River 4 km downstream from the Kinnoh Bridge. Superstructures are 3-span Gerber-type steel truss girder with a total length of

181.4 m and width of 5.3 m. Two abutments have footing foundations, and two piers have well

foundations. Due to the June Earthquake an upper chord member was broken off as shown in

Figure 31. Due to the breakage, the lower chord drooped considerably (see Figure 32). The

chord member was made of steel channels and broke during the sudden sectional change as shown

In Figure 31. At this bridge a pin came out of the fixed pin-type hearing.

Toyoma Bridge , ; / •., - : ' . . '.

'

: -

'

Toyoma Bridge, completed in 1945, crosses Kitakaml River 6 km downstream from Maiya

Bridge. The superstructures are Gerber-type RC T~shape girders, with a total length of

306 m. The pier columns are on well foundations. Due to the June Earthquake many heavy

cracks occurred at the mid-point of several girders, and at webs above the bearing supports.

A heavy crack also broke out at the base of the pier column closest to the right bank.

Yanaizu Bridge , .

Yanaizu Bridge, completed recently in 1974, crosses Kitakaml River 6 km downstream from

Toyoma Bridge. The superstructures are 6-span steel truss girders, with a total length of

450 m and width of 8.5 m. The substructures are of RC columns on steel-pipe pile foundations. A lower chord was damaged at the fixed pin-type bearing, as shown in Figure 33. Due

to seismic forces set bolts between the upper shoe and the lower chord were cut off, welding

VII-43
between the upper shoe and sole plate detached, the web plate deformed, and paint chips

flaked off. Substructures and bearing supports of this bridge did not sustain any damage.

Kimazuka Bridge

Kimazuk^ Bridge, completed in 1931, crosses Naruse River in Kashimadai Town. The

superstructures are 19-span simple supported steel plate girders, with a total length of

236 m, and width of 4.5 m. Two abutments are on footing foundations, and 18 piers are on

well foundations. Due to the June Earthquake, one pier cap was severely damaged near the

bearing (see Figure 34), and one girder was almost dislodged from the pier cap (see Figure

35). The girder moved in the transverse direction.

Eia Bridge [7]

Eai Bridge, completed in 1932, crosses Eai River in Furukawa City. The superstructures

are 9-span simply supported steel plate girders, with a total length of 155 m and a width of

7.5 m. Two abutments are on pile foundations, and each of 8 piers rest on two separate well

foundations with diameters of 2.5m and a depth of 7 m. At the time of this earthquake, the

embedment of well foundations was almost half of the initial depth due to scoring effects of

stream. Located close to the epicenter of the Northern Miyagi Earthquake of 1962 (epicentral

distance was approximately 15 km), the bearings of this bridge had been damaged in 1962. Due

to the June, 1978 Earthquake lower beams of the eight pier columns were severely cracked, and

concrete pieces separated from the beams (see Figure 36). The largest opening of the cracks

was 20 mm, and reinforcing bars were exposed. ' • -

Yuriage Bridge [7]

Yuriage Bridge, completed rather recently (1972), crosses over the Natori River near

its mouth. The superstructures are of 3-span continuous PC box girders with a center hinge

(cantilever erection) and 7-span simply supported post-tension PC beams (T-shape) with a

total length of 541.7 m and width of 8 m. Two abutments are on steel pipe pile foundations,

two piers in the lower river bed are on pneumatic caisson foundations, and 7 piers are on

well foundations. Due to the June Earthquake the nine pier columns sustained many cracks

(almost all around) mostly at the level of the ground surface. Especially Pier 1 (first

pier from the left-bank) which sustained numerous heavy cracks (see Figure 37), and concrete

VII-44
pieces separated from the column. Stoppers on a single-roller-type movable bearing on Pier 1

were damaged, guide pieces of the bearing failed, and the roller almost rolled down from

the lower shoe (see Figure 38). , . - ;

A simply supported PC beam on Pier 6 moved 6 cm downstream. The ends of one handrail

had been inserted into the ends of another handrail just above a pier. The length of insertion was 8 cm. During the Earthquake the handrail ends completely came out of the adjacent

handrail ends. It is understood from this that the two adjoining beams vibrated relatively

with a separation of at least 8 cm in the longitudinal direction.

A number of ground cracks and sand boils were observed on the higher river bed near the

right bank. The subsoils made of mostly sands are loose at the surface (about 5 m deep) and

medium to dense underneath. A hard layer exists approximately 70 m below the surface.

Date Bridge

Date Bridge, completed in 1963, crosses over the Abukuma River near Fukushima City,

with an epicentral distance of approximately 160 km. Superstructures are 4-span continuous

steel truss girders, with a total length of 288.0 m and width of 7.0 m. The two abutments

are on steel-pipe pile foundations, and the three piers are tall RC columns on caisson foundations embedded into gravel and sand layers.

Due to the June Earthquake a lower chord member buckled just at the fixed bearing on

Pier 2 (see Figure 39). Several pins at the fixed bearing and one of the movable bearings

were sheared off and came out of the shoes. The substructures did not sustain any damage.

A strong-motion accelerograph is installed on the cap of Pier 2, and triggered a complete time history of the acceleration at the pier cap (see Figure 10).

LESSONS FROM BRIDGE DAMAGE DUE TO THE MIYAGI-KEN-OKI EARTHQUAKE ,

In view of the damage to highway bridges during the Miyagi-ken-oki Earthquake of

June 12, 1978, the following lessons can be derived.

1) Damage to superstructures concentrated on bearing supports and the adjoining portions. Most damage to substructures were cracks and separations of concrete at pier columns

and abutments.

2) Damage to bearing supports were most frequently observed. It is advisable to more

carefully investigate design practices relating to bearing supports and to design and develop

VII-45
better bearings which may better resist seismic disturbances. It seems, however, that the

breakage of bearing supports served to reduce failure of bridge girders, and failure of

substructures. Therefore, it is not always advantageous to design bearings that are too

strong. "

3) Because of the resultant extensive damage to the whole bridge structure, the fall

of bridge girders should be avoided.

4) A number of older bridges such as the Kin-noh, Maiya, Toyoma, Kimazuka, and Eai

Bridges sustained relatively severe damage. In most of these bridges, either Gerber-type -

or simply supported design types are used, the width of pier caps is narrower, and no special consideration to prevent falling girders is introduced. It seems very important to

retrofit these older bridges by widening pier caps, installing devices to prevent girders '

from falling, etc.

5) In view of the damage to pier columns at the Sendai and Yuriage Bridges which were

recently constructed according to the current specifications, it is recommended to consider

the ductility of pier columns when designing short reinforced concrete columns. In this

respect further experimental and analytical investigations are necessary, and future seismic

specifications should include an appropriate regulation on ductility of pier columns.

DAMAGE TO ROADS ^

'

' ^

"''

Damaged road sites and highway bridges amount to 2,350 and 255, respectively, and about

92% of this damage took place in Miyagi Prefecture. The damage to roads was the cracking of

pavements, subsidence of road embankments, slope failures, falling stones, etc.

Table 5 shows damage to National Roads in Tohoku Region, in which the number of damaged

sites in Miyagi Prefecture amounts to about 70% of the total. As for the type of damages,

pavement cracking and road subsidence constitute about 66% of the total. It should be

noted that embankment settlement behind bridge abutments and other structures frequently

took place and was concentrated in the areas of soft foundations.

DAMAGE TO RIVERS ' '

' ^ '

:

Damage to rivers were mainly slope failure, longitudinal and transverse cracking and

Subsidence of river dykes. As for rivers controlled by the national government, the total

length of four rivers and the length of river dykes under the control of Tohoku Regional

VII-46
Construction Bureau are 313 and 422.3 kilometers, respectively. Length of damaged portions

of dykes amounts to 34.1 kilometers, i.e., and about 8% of total dyke length. These damaged

portions were located mainly in the area of soft foundations.

Weirs, conduits and culverts across the dykes and bank revetment works also suffered

damage, though not severe. Two weirs located near the estuaries of the Kitakami and Abukuma

Rivers suffered damage of local buckling of gate leaf guides made of H-shape steel. As for

conduits, culverts and bank rivetments, cracking of concrete took place.

DAMAGE TO UTILITIES

Damage to utilities concentrated in the Miyagi Prefecture. ,

Electricity was shut off immediately after the Earthquake and 419,000 families lost

electrical power. Repair work was energetically carried out and within 24 hours, 323,000

families recovered power and within 50 hours all domestic electricity was back in service.

One week after the earthquake, the electric supply for industrial use was totally restored.

Two hydro-thermal power plants ceased functioning because of breakdowns of boiler tube

systems, circuit breakers and transformers. Hydroelectric plants also suffered damage to

intake facilities, water channels, tanks, and retaining walls; however, they remained

functional. ,-, > ' - ;

Trunk transmission lines suffered cracking of retaining walls at tower foundations,

however, such damage did not hinder transmission. Fifteen substations suffered severe damage

to transformers, circuit breakers and so on, and caused shutdown of electricity in the

Prefecture. , < ;

In distribution systems, toppling, snapping, and tilting electric poles, snapped wires

and tilted transformers on poles took place in the area of soft foundations and reclaimed

land

.

1,704 cases of damage to potable water supply systems took place; 2 for water storage

facilities, 50 for purification facilities and 1,652 for pipeline systems resulting in a

suspension of water supply to 87,740 families in 55 municipalities. Repair work was rapid;

49% back in service by June 13, 67% by June 14, and 78% by June 15, and about 10 days after

the earthquake, the water supply was totally restored.

In Sendai City, office buildings of the Water Supply Bureau, the intake, purifying and

pumping facilities of purification stations, the water supply and delivery pipes all suffered

VII-47
some damage. After the Earthquake, leakage of water increased and on June 13, the estimated

leakage amounted to about one third of supply. Suspension of water supply was caused by

broken pipelines and which was concentrated around reclaimed lands in hilly areas and in

'

coastal zones.

In Miyagi Prefecture, 7 cities have gas supply systems and 6 sites suffered damage of

the system. In Sendai City, a cylindrical tank of 17,000 cubic meters storage capacity

collapsed and burned but the gas manufacturing plant did not cease to function. Leakage of

gas took place at 533 places in the supply pipe systems. Therefore, gas supply was shut

off. It took 31 days for total recovery. Shiogama City had 168 leaks, Ishinomaki City 97,

and Furukawa City 23. This damage consisted of breaking, cracking, dis-socketing and

loosening of supply mains, branches and home deliveries. Damage to gas supply systems was

concentrated in the area of soft ground, but occurred in reclaimed lands in hilly areas too.

Sewage systems also suffered damage. This damage was: (i) disf unctioning of the treatment and pumping stations due to the electrical shutdown, (ii) damage to power supply facilities, pumps, independent power generators, pressure pipes, and discharge pipes, (iii)

cracking, buckling, and dissocketing of pipe ducts and manholes. - -

Concerning telecommunication systems, 669 poles were tilted and damaged, 28.2 kilometers

of telephone wires was damaged, wires were severed in 4,538 places, 905 phones were crushed,

failure of joints between manhole and embedded duct totalled 162, breaking of embedded pipes

and ducts totalled 103 places and damage to pipes attached to bridges took place at 27

sites. This damage was repaired within a week after the Earthquake. Other than damage

described above, telephone exchange stations in the damaged area were loaded by numerous

phone calls into the area by worried relatives and friends.

CONCLUSIONS

Similar to the lessons of past earthquakes, the Miyagi-ken-oki Earthquake indicates

that damage to structures and facilities largely depends upon ground condition. It also

shows that some design codes and specifications should be re-examined and revised along

with construction standards. To lessen not only the structural damage but functional

damage as well, more research would be beneficial.

VII-48
ACKNOWLEDGMENT

In surveying the damage caused by the Miyagi-ken-oki Earthquake of 1978, much cooperation and valuable information were provided by the personnel of the Ministry of Construction,

the National Land Agency, the Miyagi Prefecture, and other related organizations. The

authors wish to express their sincere thanks to the staff members at these organizations.

REFERENCES

[1] Iwasaki, T. , Kawashima, K. and Tokida, K., "Interim Report of the Miyagi-ken-oki

Earthquake of June 1978," Technical Note of PWRI, No. 1422, Public Works Research

Institute, Ministry of Construction, 1978.

[2] Japan Meteorological Agency, "Report on the 1978 Miyagi-ken-oki Earthquake," Technical

Note of the Japan Meteorological Agency, No. 95, 1978.

[3] Masuda, T. , "Aftershocks of the 1978 Miyagi-ken-oki Earthquake," Abstract of Annual

Meeting, Seismological Society of Japan, 1978.

[4] Ministry of Construction, "Damage Due to the Miyagi-ken-oki Earthquake of 1978," Tohoku

Regional Construction Bureau of the Ministry of Construction, 1978.

[5] Ministry of Construction, "Data Book on the Nature of the Tohoku Region," Planning

Division of Tohoku Regional Construction Bureau, Ministry of Construction, 1961.

[6] Ministry of Construction and Miyagi Prefecture, "Ground in Coastal Area of Sendai

Bay," Planning Bureau of the Ministry of Construction and the Miyagi-ken, Vol. 10,

1965.

[7] Miyagi Prefecture, "Damage Due to the Miyagi-ken-oki Earthquake of 1978," 1979.

[8] Okubo, T. and Iwasaki, T. , "Summary of Recent Experimental and Analytical Seismic

Research on Highway Bridges," Workshop on Research Needs of Seismic Problems Related

to Bridges, San Diego, USA, 1979.

[9] Okubo, T. and Ohashi, M. , "Miyagi-ken-oki," Japan Earthquake on June 12, 1978 —

General Aspects and Damage, U.S. National Conference of Earthquake Engineering —

1979, August 22-24, 1979," Stanford, California, USA.

[10] Seno, T. , Sudo, K. and Eguchi, T. , "Fault Mechanism of the 1978 Miyagi-ken-oki

Earthquake, Abstract of Annual Meeting, Seismological Society of Japan, 1978.

VII-49
Table 1. Comparison between Intensity Scale of Japan Meteorological Agency

and Modified Mercalli Scale

Accelcf al ions (gals)

I 1 0 S 1

1 1

1 1 1 1 1 ! S 1 0 20 50 100

1

1 l' 'l' l'

1 1 200 500 IC

1 1

00

Japan Meteorolo([ical Agency Seismic

Intensity Scales

1 —0 08

4

1 2S 1 8 1 25 IV 80 V

1 1 1 1

1 1 250\^400

Modified Mercalli Scale

1

1 1 1 1 1 1 121 1 5 ^ 10 V 21 M 44 VI 94 VI

1 1 1 1 1 1

202 K 432

1 1

Table 2. Stratigraphy of Tertiary in the Vicinity of

Sendai (After "Data Book of the Nature in Tohoku")

Era Group Formation Thickness Lithofacies Remark

Pliocene o

"Oc

Dainenji P. 30-130

silt stone, tuff, limonite Deposi t

Yagiyarna P. 20-30 Alternate of tuff and silt stone Stage 6 Continental Deposit

Hi rosegawa

tuff P. 15-100 Pisolite

Kitayama P. 1-10 Tuff, silt stone

Tatsunokuchi P. 30-60 Alternation of tufaceous

sand stone and silt stone Stage 6 Marine

Deposit

Kameoka F. 15-50 Silt stone, sand stone, tuff Stage 6 Continental Deposit

Miocene

o

i

Mitaki Basalt

200 Basalt

Shi rasawa

F. 330 Shale, white sandy tuff Stage 5 Continental Deposit

Tumoto P. 150 Massive tuff breccia, trachitic green tuff

(0*

iioa2

Tsunaki P. 130 Agglomerate, tuff, sand

s tone Stage 4 Marine Depos i t

Hatatate P. 130 Alternation of conglomerate

and shale Stage 3 Marine Deposit

Moniwa P. Tsukinoki P. 60-130 Sand stone, conglomerate

tuff Stage 2 Marine Deposit

Stage 2 Continental Deposit

Takadate P. 60-250 Andesi te Stage I Continental Deposit

Table 3. Quaternary Stratigraphy in the Vicinity

of Sendai

(After "Ground in the Coastal Area of Sendai Bay")

Epoch Formation Thickness M

Fac i es Sedimentation

Circumstance

Bear i ng Capac i ty

c

Pukanuma P. 0-5 Fine to coarse sand Sand dune 3-10

Kasuminome

P. 1-5 Sandy loam Al luvial fan

and flood plain

)-7 Coarse sand wi th grave I 15-20

Fukudamach i F. 1-10 Loam to peat Lacus trine 2-5 Sandy loam Swamp 3-8

Iwakiri P. 10--30

Fine to coarse

sand Shallow sea

20-25

Silty loam 10-20 Clay 3-7 Sandy loam 3-7

Pleistocene

\ Shimomachi W terrace P.

' Vakabayashi

\ clay P.

Gamou P.

10-35

Sand and gravel River terrace

1

20-35

Clay A I luvial fan 15-20

VII-50
Table 4 List of Highway Bridges Severely Damaged in Miyagi

and Fukushima Prefectures

No. Bridge Route

Characteristics of Bridges Tear

pleted

Out 1 ine of Damage a Total Length

(mT Width(m) Superstructure

1 Sendai National Highway

No. 4 310.0 19.0

Cornposit Steel

Plate Girder 1965 Horizontal Cracks at All Pier Columns

2 Abukuma Nati onal Highway

No, 6 571.6 6.0 Steel Varr en Truss ^ Simple Steel PI. Girder 1932 Horizontal Cracks at Pier Columns

3 Ono National Highway

No. 45 247.3 5.5 Simple Steel PI. Girder 1936 Movement of All Girders,

Pai lure of Bearing

Supports, Failure of Slabs

4 Ten-noh

National Highway

No. 45 367.7 6.0 Steel Gerber PI. Girder, Steel Langer

Truss

1959 Vertical Crack at a Pier Column

5 Toyama

National Highway

No . 342 306.0 5.3 RC T-beam 1945 Cracks at Beams and Pier Columns

Kin—noh

National Highway

No. 346 575.5 6.0

i —o impi e oT.eei ri. Girder, 5-Steel Truss, Steel Gerber PI. Girder

Pall of a Suspended Girder, Failure of Bearing Supports

7 Turiage Miyagi Pre f ec tural Highway 541.7 8.0 3-Hinged PC Girder, 7-Simple PC Post-Tension T-beam

1972 Cracks at Pier Columns

,

Liquefaction

8 Kimazuka Miyagi Prefectural

Highway 235.0 4.5 19-Simple Steel PI. Girder 1931 Failure of Bearing Supports, Failure of Expansion, Liquefaction

9 Eai Miyagi Prefectural Highway 155.0 7.5 9-Simple Steel PI. Girder 1932 Cracks at Pier Columns

10 Maiya Miyagi Prefectural Highway 181.4 5.5 Gerber Truss Girder 1928 Break of Upper Chord, Movement of Truss Girder

11 lanaizu Miyagi Prefectural

Highway 450.0 8.5

1-Simple Steel Truss

2-Cont. Steel Truss

3-Cont. Steel Truss

1974 Failure of Bearing Supports, Crack and Buckling at Lover Truss Chord

12 Date Fukushima Prefectural Highway 288.0 7.0 4-Cont. Steel Truss 1963

Failure of Bearing Supports, Buckling at Lower Truss Chords, Crack of Pavement

Table 5 Damage of National Road in Tohoku Region

(a) Classification of Damage in accordance vith

Prefectural District

Prefecture

Route No. Total

4 6 13 45 47 48 108 Fukushima 2 6 3 11 Miyagi 58 6 36 2 6 8 116 Ivate 10 28 38 Yamagata 2 2 Total 70 12 5 64 2 6 8 167

(b) Classification of Damages in accordance with Types

Route No, Total 4 6 13 45 47 48 108 Bridge Failure 8 2 4 1 15 Pavement Failure 47 7 2 49 1 1 3 110 Slope Failure 1 2 5 8 Falling Stones 2 1 3 Others 14 1 3 4 1 3 5 31 Total 70 12 5 64 2 6 8 167

VII-51
I36* 138' 140* 142* 144*

Fig. 1 Distribution of Seismic Intensities

Epicenter of June 20

* Earthquake by Mosuda ^ Epicenter of June 20 ^ Eofthquake by J MA

JoDon I TRENCH

June 13-30 /

- Epicenter of Maximum • Aftershock by JMA A Epicenter of Feb20

by JMA

Fig. 2 Distribution of Aftershocks

(After Masuda, 1978)

39tJ fw//

Aftershock

within 24 hours

30l(m

Epicenter

Epicenter

Slip Direction^ Rate Boundory

-Aftershocks within 24 hours Fig. 3 Fault Model of Miyagi-ken-oki

Earthquake of June 12, 1978

(After Seno et al, 1978)

VII-52
Fig. 4 Topographic Feature

of Tohoku Region

f

Nonokita Ruer

Noflomachj - Rifu Fault

Cofitout ot Bou Roch Depth

Lo»ir ShiM- Noko- Uemochi Ooino- Aoba- Doininji YoqiTami HiroM- Tatuno- fComcoko Urlaki Ttuaeki Hglalatt Uoeivo Tokddali Ri(n Plain MChi Mchi Tirroct . hora rano F F F govo kuehi F BoMll F F F Aft4«titi F [tapnit T«rrac« Tirrtc* T«irK« Tutf F

- QuQlernory - - TertiOft - Tnossic

Fig. 5 Geological Map in the Vicinity of Sendai (After "Ground in the

Coastal Area of Sendai Bay")

VII-53
Mountoin zone Hilliy zone Terroce zone Plain zone Sond Dune ion

Fig.

Shell Earlhernord Slone-Wa»e [13 B EZl [Z3 S Tuft Lignite Gronil Andesite Mud Stone Gravel Sand Loam Peot

6 Typical Geological Section of Sendai Area (After "Ground in the

Coastal Area of Sendai Bay")

n «0 to Kk> Fig. 7 Distribution of Maximuip

Accelerations in Gals

.4

400

300

200

a 100-

Range of Horizontal Accelerations

50

Range of Vertical Accelerations 20

Legend O Larger

10| Horizontal Component • Smalle

Horizontal Component

J< Vertical Component

I L

30 100 500 1 000

Epicentral Distance ^ (km)

Fig. 8 Relation between Epicentral

Distance and Maximum Groxmd

Acceleration

VII-54
(a) Longitudinal

(b) Vertical H

(c ) Transverse

(1) Ground Surface Near Pier

Over Scale

(a) Longitudinal

(b) Vertical

(c) Transverse

(2) Pier Crest Fig. 9 Strong-Motion Acceleration Records at Kaihoku Bridge

and Ground Nearby

-200 0

(a) Longitudinal

(b) Vertical

(c) Transverse

Fig. 10 Strong-Motion Record at Cap of Pier 2, Date Bridge

VII-55
Iwate Pref.

New Kitakami R.

# Locations of Bridges Severely Damaged

Quaternary

O Liquefied Locations N^onR. I 1 Alluvium Terrace and Fan Gravel Ouartz Avdes.te ! Nec. Tertiary

Tertiary or Older Abukuma R.

Fig. 11 Geological Features and Locations

of Major Bridge Damages and

Liquefaction Sites

310.00m Fukushima Simply Supported Steel • Plate gi rders

Sendai

Fig. 12 General View of Sendai Bridge

VII-56
Steel pi. Girder 5 Steel Truss Girders 9 "Gerber Steel PI. Girders

23 200 59 700 60 000 . 60 O'JO . 60 OOP 60 000 28 000 28 000

28 000 28 000

1

Fig. 17 General View of Kin-noh Bridge

Fig. 18 Fall of a Suspended Girder,

Kin-noh Bridge (During the

Earthquake of June 12, 1978)

Fig. 19 Bearing Stiffening Plates to Resist Transverse Movement,

Kin-noh Bridge (Added after

1962 Earthquake)

VII-57
Fig. 20 Failure of bide Block of

Oval Line Bearing on Pier 11,

Kin-noh Bridge (Just after

February 20, 1978)

Pig. 21 Pull-out of Anchor Bolts at Upstream Fixed Pin-type

Bearing on Pier 6, Kin-noh

Bridge (Just After February

20, 1978)

Fig. 22 Failure of Set Bolts at downstream Fixed Pin-type Bearing

on Pier 6, Kin-noh Bridge

(Just After February 20, 1978)

Fig. 23 Protrusion of Rollers at the

Movable Pin-roller-type

Bearing on Pier 5, Kin-noh

Bridge (Just after February

20, 1978)

Fig. 24 Movement (55 cm) of Plate

Girder at the Downstream

Support on Pier 8, Kin-noh

Bridge (After June 12, 1978)

Fig. ^5 Failure of the Right-Bank

Abutment, Girder Moved Toward

Abutment by 10 cm. Temporary

Support is seen. (After June 12, 1978)

VII-58
Right Abutment, Kin-noh Fig- 27 Pull-out of Anchor Bolts and

Bridge (After June 12, 1978) Settlement of the Shoe at the

Upstream Fixed Bearing on Pier

6, Kin-noh Bridge. (After June

12, 1978)

Fig. 28 Failure at Downstream Fixed

Bearing on Pier 6, Kin-noh

Bridge (After June 12, 1978)

Fig. 29 Failure of Upstream Movable

Bearing on Pier 5, Kin-noh

Bridge (After June 12, 1978)

June 12, 1978)

VII-59
Fig. 32 Drooping of the Lover Truss

Chord, Maiya Bridge

Fig. 33 Damage to Lower Flange of Truss Chord near a Fixed Pin-type

Bearing, Yanaizu Bridge

Fig. 36 Cracks at Pier Columns, Eai

Bridge

Fig. 37 Cracks at Pier 1, Yuriage

Bridge

VI 1-60
Failure of Guide Piece at Movable Bearing on Pier 1,

Yuriage Bridge

VII-61
DAMAGE TO RIVER DYKES CAUSED BY THE MIYAGI-KEN-OKI

EARTHQUAKE OF JUNE, 1978

Kazuya Yamamura

Yasushi Sasaki

Yasuyuki Koga '

Eiichi Taniguchi

Public Works Research Institute

Ministry of Construction

ABSTRACT

A large number of engineering structures were severely damaged by the Miyagi-Ken-Oki

Earthquake of June, 1978. This paper describes the damage to the river dykes in the

Kitakami, Naruse, Eai, and Yoshida Rivers by this earthquake.

Field surveys and laboratory investigations were performed on subsoils underlying the

dykes and following conclusions drawn:

i) River dykes in the Miyagi prefecture were extensively damaged by this earthquake

ii) The seismic resistance of river dykes is affected by the stability of supporting

subsoils during an earthquake: either by a tendency toward liquefaction, or by the bearing

capacity of the subsoils.

iii) The damage of river dykes has a close relationship to the micro-topographical

features. Dykes on a "former river bed" have a high potential to be damaged by earthquakes

It is suggested by the analyses on Eai River that the Land Form Classification Map for Floo

Control Planning is useful in predicting river dyke damage.

iv) Re-liquefaction may or may not occur at the same place. It was observed not to

occur at a layer which had been compacted by previous earthquakes.

KEYWORDS: Earthquake-resistant dykes; river dyke damage; soil liquefaction; soil

relationships.

VII-62
1. INTRODUCTION ^ - " ''J

'

An earthquake of Rlchter magnitude 7.4 occurred off the Pacific coast of the Miyagi

prefecture in northern Japan at 17:14 on June 12, 1978. The seismic intensities according

to the Japan Meteorological Agency are V at Sendai, Fukushima, Ofunato, and Shin jo, and IV

at Tokyo, Mito, Akita, etc. This earthquake caused extensive damage to diverse engineered

structures such as buildings, bridges, river dykes, road embankments, etc.

This paper describes the outline of damage to river dykes caused by the earthquake,

and the result of investigations into the failure and settlement of dykes. In general the

damage to dykes is closely related to subsoil conditions, and the field survey and investigation were performed from this point of view.

2. AN OUTLINE OF THE DAMAGE TO RIVER DYKES '-y '

' ^ -

Table 1 shows the number of sites where restoration works were carried out after the

Miyagi-ken-oki earthquake in each prefecture of the Tohoku region through the assistance of

the Japanese government. This table shows that the total number of damage points to the

river facilities amounts to 176, which is next to the number of the damaged sites for highway facilities. It is also shown that most of the damage is concentrated in the Miyagi

prefecture.

Some distinctive features can be seen on the damage caused by this earthquake as

compared with that of Izu-Oshima-Kinkai Earthquake (1978, M = 7). A large number of slope

failures occurred, and highways at Izu Reninsula were cut off by the Izu-Oshima-Kinkai

Earthquake, because the mountainous area was struck by this earthquake. On the other hand

the Miyagi-ken-oki earthquake hit the plain area and the bridges and earth structures such

as river dykes on flood plain deposits were extensively damaged.

Table 3 shows the number of damaged sites for river dykes, revetment and others. Under

"others," the damage to transverse pipes in dykes and water gates is included.

Figure 1 illustrates the damaged sites of river dykes in Miyagi prefecture. The rivers

controlled by the regional bureau of the Ministry of Construction in this area are the

Kitakami River, Naruse River, Natori River, and Abukuma River, and severe failures, cracks and

VII-63
settlements of dykes were observed along these rivers. The Nanakita River was also extensively damaged. Most of these damage points are located on alluvial plain deposits where

the epicentral distance is less than 120 km.

Table 4 shows the length of damaged dykes along the main four rivers in Miyagi prefecture. It is seen in Table 4 that the total length of damaged dykes shown in Table 3 reaches

20.2 km (4.8% of total dyke length 422 km), and if the slight damage which was repaired by

the maintenance works budget are included, the damaged length increases to 7% of the total

dyke length.

The most severe failure and settlement were observed at the left side bank of the

downstream of Kitakami River (5.2 ~ 10 km), the right side bank of Yoshida River (13.2 ~

18.2 km), a branch river of Naruse River, and the right side bank at a distance of 4 km,

and the left side bank at a distance of 5 km from the mouth of the Natori River. Table 5

shows the length of embankments where the cracks or settlements took place along Yoshida

River and Kitakami River. At the right side bank of Yoshida River the settlement over 1 m

was seen as long as 242 m. As shown in Photo 1 the dykes at this site were so extremely

destroyed that the repair work was performed by recompaction of the dykes. In this area

the cracks were observed at the top and slope of the embankments, the maximum settlement

of the top of the dykes was 1.5m, and the width of maximum crack was 60 cm.

3. MICROTOPOGRAPHICAL FEATURES AND DAMAGES TO RIVER DYKES . ^ i- '

In this section the relationships between the microtopographical features and the

damage to the river dykes are discussed. For the area along large rivers throughout Japan a

Land Form Classification Map for Flood Control Planning has been compiled. This map provides

the microtopographical features near the rivers.

Table 6 demonstrates the relationship between the number of sites, the length of damaged

dykes, and the topographical features determined by using the map described above. This

table includes only rivers directly controlled by the Japanese government. If a damaged site

is located in two microtopographical regions the site is weighted as 0.5. Table 6 denotes

that 54% of the total number of the damaged dykes are constructed on the "flood plain" and

that 14% belongs to the "former river bed," and 60% of the total length of damaged dykes

occurred on the "flood plain." Dykes on "reclaimed land" are lower in number. This fact

suggests that the dykes on soft ground were more severely damaged.

VII-64
The Land Form Classification Map for Flood Control Planning was compiled using the

topographical maps surveyed on and after 1911, and so the history of the river is partially

included in the map. However, in general, the geotechnical characteristics of the subsoils

of the river dykes which have a lengthy history are not well known, so the map can be useful

in estimating the geotechnical characteristics of the soil.

Figure 2 illustrates the ratio of the total damaged length of the Eai River dykes to

the whole length of dykes in each topographical classification, whereas Table 6 showed the

percentage of the length (or number) of damaged dykes in each topographical classification

to the total length (or number) of damaged dykes. For a river flowing in an alluvial plain

the length of dykes located in the "flood plain" is so much longer than that of any other

classification that the percentage of the damaged length of dykes on "flood plain" to the

total damaged length is the highest as shown in Table 6. But Figure 2 shows that the highest

damage ratio to be the microtopographical classification "former river bed." This indicates

that dykes on a "former river bed" easily become unstable under seismic loading. In Figure

2 the damage caused by the North Miyagi Earthquake in 1962 (M = 6.5) is shown and the ratios

of damage on former river bed sites in both earthquakes are shown to differ. The details of

this will be described later. The sites examined to construct Figure 2 include not only the

places rebuilt during the disaster repair work but the places repaired by the maintenance

work as well. . .

4. LIQUEFACTION OF SUBSOILS UNDER RIVER DYKES

Sand and water spouting was observed on the ground surface just after the Miyagi-Ken-Oki

Earthquake. This indicates that liquefaction took place in the ground. It is reported [3]

that liquefaction occurred at 30 sites. Since sand spouting was observed at several points

where the river dykes were damaged, it is believed that liquefaction is one of main causes

for the damage of the river dykes. -

There were extensive cracks and failures to the dykes on the Eai River by the North

Miyagi Earthquake of April 30, 1962 (M = 6.5). It is estimated that the maximum acceleration

of the surface ground near Eai River (0 ~ 27 km) was almost the same for both earthquakes of

1962 and 1978.

The geological profiles of this area are shown in Figure 4. The alluvial deposit is

piled up on tuff bed rock for a thickness of 20 ~- 60 m. The thickness of the alluvial

VII-65
deposit is about 60, 16 ~ 20, and 40 m for upper, middle and lower stream areas, respectively. Figure 5 illustrates the damage distribution of river dykes by two earthquakes

described above by denoting the damaged spots with letter A for 1962 earthquake and with

letter B for 1978 earthquake. It indicates that the intense damage is concentrated in the

upper and lower areas of Eai River where thicker layers of alluvial soils are deposited,

while no damage took place in the middle stream area in both earthqukes, where the alluvial

deposit is less than 20 m in thickness. The microtopographical condition underlaying the

dykes was examined by using the Land Form Classification Map for Flood Control Planning.

The dykes of Eai River were classified into five groups according to the classification

of microtopographical features on this map: most of the dykes lie on the "flood plain" or

the "natural levee"; the rest are on the "former river bed," the relatively higher parts of

low land surface," or the "hill." From this classification, the damaged sections of the

dykes belong to the "former river bed," "flood plain" or "natural levee." Table 7 shows the

relationship between the microtopographical features and the extent of damage. It can be

seen in this table that the damage length ratio on the "former river bed" site in the 1978

earthquake was much smaller than it was in 1962, while the damage length ratios of the other

classification remained nearly the same. - -

The total length of dykes damaged by the 1962 earthquake was about 1.5 km and that of

the 1978 earthquake was about 1 km. It is estimated that some of these damaged dykes were

destroyed by liquefaction from the sand spouting observed near the dykes.

In Table 8 are shown six sites where liquefaction occurred in each earthquake or both.

Three places out of these six seem to be located on the "former river bed," and site b is

very near the former river bed, so it appears that a sandy layer may exist in these four '•'

;

sites. ^' '

It is interesting that the liquefaction occurred during the 1962 earthquake and no

liquefaction took place during the 1978 earthquake at Ninofukuro, Gokenyashiki , Hayamado, /

'

and Eai, although the acceleration at the ground surface was almost same for both earthquakes

as shown in Figure 3. • '. '- •3 ^

The boring logs at Ninofukuro and Hayamado are shown in Figure 6. The alternation of

sandy and clayey layers are seen. It can be estimated from the geotechnical data obtained

in 1979 that the sand layers from the depth of 8.5 to 10 m at Hayamado and from the depth

of 3.3 to 7.5 m at Nonofukuro were liquefied.

VH-66
Dynamic triaxial tests were performed on undisturbed soil samples taken from the layers

described above. The relative density Dr-j^ at the time of 1962 can be calculated from the

following equation by substituting the liquefaction stress ratio R (= —— , t: shear

2a^'

stress, ' : effective confining pressure) obtained from the dynamic triaxial tests and

the relative density Dr2 in 1978.

°^1 ^

(L/R2) ^ °^2

where L is the stress ratio which was developed in the ground during the 1962 earthquake,

which was calculated by using the equation proposed by Iwasaki, et al. [4]. The results are

shown in Table 9.

The settlement of the estimated liquefaction zone in Figure 6 was calculated from the

following equation by substituting the void ratio in 1962 and the recent one.

AH = liJ-fLxH (2)

The results are sho^ra in Table 10. This table indicates that the settlement due to dissipation of pore water pressure developed by the liquefaction during the 1962 earthquake is 0.7

cm at Hayamado and 42.5 cm at Ninofukuro. According to the survey just after the 1962 earthquake, the settlement of the dyke at Ninofukuro was about 50 cm while no settlement of the

dyke was observed at Hayamado, although the revetment was extensively damaged.

Also while the result presented above is based on some assumptions, it nevertheless

suggests that where a sandy ground is compacted by an earthquake as a result of liquefaction,

little or no liquefaction will occur at the same place during a subsequent earthquake with a

similar order of ground motion. '

'

'

In Figure 2 of the previous section, the damage ratio of dykes on the former river bed

was 41% for the 1962 earthquake, and it decreased to about 6% for the 1978 earthquake. It

is reasonable to note that such a phenomenon can be attributed to the change of liquefaction

characteristics of subsoils underlying the dykes due to seismic loading. However re-liquefaction was observed at the left side bank of Natori River (2.8 km) after the earthquake of

June, 1979, while sand spouting was observed just after the earthquake in February, 1978, so

investigation on reliquef action should be continued.

VII-t,7
5. FAILURE OF RIVER DYKES ON THE SOFT CLAY SUBSOILS ' \ •

The Yamazaki area of the Yoshida River dykes (right side bank 15 ~ 18 km) was severely

damaged. The cross section of the dykes after the earthquake is shown in Figure 7 and the

settlement of the embankments and the large mouthed cracks were observed. The maximum

settlement of the embankment in Yamazaki area was 2 m and the maximum width of cracks at the

top of the levee was 60 cm. r; •

The dykes in this area were constructed in the 1920' s. The right side bank was moved

outside the river and both side banks were enlarged by the construction work of 1949.

Figure 8 illustrates the distribution of the earthquake disaster in dykes along Yoshida

River and the position of former dykes. In Figure 8 the extent of damage of dykes is denoted

by A, B, C corresponding to the amount settlement of the tops of the dykes. This figure

indicates that only dykes along the right bank were damaged by the earthquake and that no

extensive failure was observed to dykes along the left bank. Figures 9 and 10 show soil profiles beneath the dyke in the cross-section and longitudinal directions, which were obtained

from a boring and sounding survey after the earthquake. These figures show that the ground

of this area consists of an alternation of sandy and clayey strata. As far as the sounding

data obtained by the Datch cone penetration test and Swedish sounding are concerned, it was

difficult to clearly point out geotechnical difference in subsoil condition between the right

bank and the left bank. It is recorded that the construction work of the dyke was hard

because of the settlement and foundation failure due to the low strength of the layer. It

leads to the estimation that the dykes had low stability during the earthquake. r,, ., ,,

In Figure 11 former dykes are shown by dotted lines. Figure 11 also shows the new dyke

of larger scale which was constructed outside the river from 1949 to 1954, and the distance

between new and old dykes which reaches 80 ~ 100 m on the right bank. On the other hand, the

new dyke for the left bank was built up at almost the same place as the old one. This historical aspect of the dyke may be associated with the variations of disaster severity by the

earthquake on the right side and left side dykes. However, further investigation is necessary to pursue this possible relationship.

VII-68
6. CONCLUSIONS .

The following conclusions were obtained from the investigation:

1) The river dykes in Miyagi prefecture were extensively damaged by this earthquake.

2) The seismic resistance of river dykes is affected by the stability of supporting

subsoils during an earthquake: either by a tendency toward liquefaction, or by the bearing

capacity of the subsoils.

3) The river dyke damage has a close relationship to the microtopographical features.

Dykes on a "former river bed" have a high potential to be damaged by earthquakes. It is

suggested by the analyses of the Eai River that the Land Form Classification Map for Flood

Control Planning is useful in predicting river dyke damage.

4) Re-liquefaction may or may not occur at the same place. It was observed to occur

at a layer which had been compacted by previous earthquakes.

• ' ACKNOWLEDGMENTS

The authors would like to express their heartiest appreciation to the staff members of

the river control section, the disaster prevention section of the river bureau, the Geographical Survey Institute, the Tohoku regional construction bureau, the construction office of the

downstream area of Kitakami River, and to the Ministry of Construction for their help in

carrying out the damage survey and field investigations.

REFERENCES

[1] The Land Classification Map for Flood Control Planning, Geographical Survey Institute.

[2] Sasaki, Y., Taniguchi, E. , and Funami, K., "The Damage of River Dykes by Earthquakes

and the Hysterisis of Subsoils," The annual meeting of JSCE, 1979 (in Japanese).

[3] Yamamura, K. , Ihasaki, T. , Sasaki, Y. , Koga, Y. , Taniguchi, E., and Tokida, K. , "Ground

Failures and Damage to Soil Structures from the Miyagi-Ken-Oki , Japan Earthquake of

June 12, 1978," U.S. National Conference on Earthquake Engineering, 1979.

[4] Iwasaki, T. , Tatsuoka, F. , Tokida, K. , and Yasuda, S., "A Practical Method for Assessing Soil Liquefaction Potential Based on Case Studies at Various Sites in Japan,"

Proceedings of the Fifth Japan Earthquake Engineering Symposium, 1978 (in Japanese).

VII-69
Table 1 Number of sites where restoration works were carried out with assistance of the Japanese government after the

Miyagi-Ken-Oki earthquake

Prefecture River Coast Sabo Road Bridge Total

Miyagi 167 14 1 549 70 801

Iwate 8 4 108 16 136

Akita 3 3 Yamagata 1 6 1 8 Fukushima 1 2 3 Total 176 18 1 667 89 951

Table 2 Number of Sites where restoration works were carried

out with assistance of the Japanese government after

the Izu-Oshima-Kinkai earthquake

River Coast Sabo Road Bridge Total

Number 45 637 4 687

Table 3 Number of sites where river facilities were repaired

after the Miyagi-Ken-Oki earthquake

Epicentral

distance

Directly controlled by

the Japanese government

With assistance of the

Japanese government

Dyke Revetment Others Dyke Revetment Others Total

A= 80Km 8 7 15

80 90 3 6 1 14 2 26

90 'V. 100 14 1 5 6 2 28

100 ^ 110 35 7 1 33 26 3 105

110 -x- 120 5 4 1 10 36 4 60

120 1 5 11 3 20

Total 65 17 5 53 100 14 254

VII-70
Table 4 Damaged dykes in Miyagi prefecture directly controlled

by the Japanese government

number in ( ) shows the number of sites

Controlled length

Length of dykes

Dainaged dykes

(disaster restoration

works)

Damaged dykes

(maintenance works) Damaged dykes(total)

Complete

dyke Provisional

dyke

Total Length Ratio Length Ratio Length Ratio

Kltakaml 137. S*"" 78.5""° lOA.e'" 183.1""° (19) . 8.9'"° 4.9^ k^

3.1 1.7^ km 12.0*"° 6.6^

Ha ruse 82.

A

22.2 111.0 183.2 (26) 7.7 5.8

(45) 2.6 2.0 (71) 10.3 7.7

Natorl 29.4 28.6 3.4 32.0 (17) 3.0

9.4 (52) 2.8 8.8 (69) 5.8

18.1

Abukuma 63.7 20.5 53.5 74.0 (3) 0.5 0.7 (14) I- , 1.6

1

(17) 1.7 2.3

Total 313.0 1A9.8 272.5 422.3 '"'20.2 4.8 (146) 9.6

2.3 (211) 29.8 7.1

Table 5 Length of damaged dykes

(m)

Yoshida River Kitakami River

Crack 6,219 4,469

Im 'V' 242 0

ement 0.5 'V' Im 321 1,866

lettl 0.2 'v. 0.5m 343 2 , 368

U2 Total 906 4,234

Photo 1 Damage of Yoshida River dyke

V 11-71
Epicentral distance

Repaired sites (directly controlled by the Japanese government 1

o : Repaired sites (with ossistance of the Japonese goverriment

)

^ : Alluvial deposits

Fig. 1 Damaged sites of river dykes by the Miyagi-Ken-Oki earthquake

VII-72
in 1962

1978

OJ CT> o * 3

Fig. 2 Microtopographical features and rate of damage

to dykes

Table 6 Microtopographical features and damaged site

and length

Natural

levee

Relative

higher parts of low land surface

Dune

Former

river

bed

Flood plain Reclaimed

land Total

sites of damaged

revetment

(13.1) 8.5

(7.7)

5 (6.2)

4

,

1

(13.8)

9 (53.8)

35 (5.4) 3.5

(100%)

65

1 Length of damaged

dyke

(8.6) 1.73(8.0) (1.3)

'

0.27

(4.4) 0.89

'

(60.0)12.12(17.8) 3.59(100%) 20.21

Sites of damaged

revetment

(47.1)

8 (17.6)

3 (11.8)

2 (17.6)

3

t

' (5.9)

1 (100%)17

Length of damaged

revetment

(26.9) 0.35(9.2) 0.12(39.2) 0.51(21.5) 0.28

(2.3) 0.03(100%) 1.30

Total damaged

sites

(20.1)16.5(9.8)

8 (7.3)6 (11.0)9 (46.3)38

(5.5) 4.5(100%)

82

Total damaged

length

(9.7) 2.08(8.0) 1.73(3.6) 0.78(4.1) 0.89(57.6)12.40(16.8) 3.62(100%) 21.51

7o50

40

30

20

I I Earthquake

Earthquake

IF771 I

VII-73
N

Fig. 3 Estimated acceleration at the ground surface near

Eai River by two earthquakes in 1962 and 1978

Table 7 Microtopographical feature and damage to dykes

in Eai River

Microtopographical

feature

Total length

of dykes

Damaged dykes by the

1962 earthquake

Damaged dykes by the

1978 earthquake

Site Length Site Length Rate of

damage Site Length Rate of

damage

Former river

bed 9 km

1.23 2 km0.5

%

40.7 1 km0.07

%

5.7

Flood plain 47 22.17 4 0.43 1.9 4 0.52 2.3

Relative

higher parts

of low land

surface

16 8.07 1 0.01 0.1

Natural levee 40 25.45 7 0.61 2.4 6 0.47 1.8

Mountain, hill,

platform 6 0.84

Total 118 57.76 13 1.54 12 1.04

VII-74
'C-C' (25 7 km) E-E' (16.8km) F-F'(116km)

G 7 8 9 10 11 12 13 km

ir "1 Mountaneous area . former river

Fig. 4 Geological Profiles Near Eai River

-A-IO

A-1 ~ A-22 Earthquoke in 1962 Old dyke

B-l - B-16 Earthquake in 1978 ^'-i Former river ° Leak of water by flood na Mountanious area « Slightly damoged

• Foiled by flood 5 Kilo post ® Sever I y damaged

— Crack

cS) Sand spouting

<i> Settlement

V- Slope failure

Fig. 5 Damaged Points by the Two Earthquakes in 1962 and 1978

Table 8 Liquefaction of subsoils of Eai River dykes

Site Micro topographical

feature Kilopost

Earthquake

causing

liquefaction

a Wabuchi (B-l-^B) Former river bed Fight side bank 0. 6*^+125^1. o'^+Ae Only the 1978 earthquake

b Ninofukuro (A-1) Flood plain Right side bank 14 .0^-14. 0+100 Only the 1962 earthquake

c Ramiyaji (B-9, 10) Natural levee Left side bank 14 .4+70^14.6+10 Both

d Cokenyashlki (A-2) Former river bed Left side bank 15 S'v.lS. 8+100 Only the 1962 earthquake

e Hayamado (A-6) Former river bed Left side bank 23 8'v24 .

2

Only the 1962 earthquake

£ Eai (A-10) Flood plain Right side bank 27 .O+100'v27.2 Only the 1962 earthquake

VII-75
Fig. 7 Cross section of damaged dyke (Yoshida River,

right side bank 18.0 km)

Present dyke Old dyke Crack

Settlement Slope failure Leak ot water

Damage

by Miyoqi

I A

Ken-oki J g

Earth quake

of 1978 [C

Settlement

Over 10""

Over 50'^'"

Over 20'^'"

Fig. 8 Distribution of Earthquake Disaster along Yoshida

River and Historical View of Slope Failure and

Leak of Water by Flood

VII-76
Table 9 Estimation of relative density

Sample No. Dynamic triaxial

test o'max = 150 gal

R Dr (%) L Dr (%)

Hayamado Sl-6

(-9m)

0.215 50.2 0.206 48.0

Sl-8

(-12.7m)

0.310 56.9 0.210 39.0

Ninofukuro S4-1

(-1.7m)

0.38 97.2 0.175 45.0

S4-3

(-4.7m)

0.42 68.0 0.239 39.0

Table 10 Estimation of settlement

before lique- After lique- Liquefaction

Site ^min

faction faction layer Settle-

^max

Dr (%) e Dr(%) e Depth

( m)

Thickness (m)

ment

(cm )

Hayamado 0.985 0.612 48.0 0.806 50.2 0.798 8.35

9.90 1.55 0.7

Ninofukuro 1.754 0.898 39.0 1.420 68.0 1.172

3.25

'v, 7.40 4.15 42.5

VII-77
Fig. 10 Longitudinal Section of the Ground

(Yoshida River)

Fig. 11 Cross Section of Dyke (Yoshida River 15.8 km)

:6V

VII-78
DAMAGE TO PORT STRUCTURES BY THE

1978 MIYAGI-KEN-OKI EARTHQUAKE

Hajime Tsuchida ' ' ' ' '

Setsuo Noda "

Structures Division .

;

i (i

Port and Harbour Research Institute '

ABSTRACT

The 1978 Miyagi-ken-oki earthquake (M = 7.4) caused damage to port facilities. From

field investigations and analyses, the following lessons concerning earthquake engineering

were learned:

1) Strong-motion accelerograms were recorded by an observation network of the Port and

Harbour Research Institute. The largest peak ground acceleration of 280 gals was recorded at

Shiogama Port

.

2) Port facilities were damaged seriously when a backfill liquefied. Otherwise, they

suffered only lightly.

3) Current procedures for estimating liquefaction potential can lead to an appropriate

judgement for actual sites.

4) A relationship between the seismic coefficient and the maximum ground acceleration

for gravity quaywalls : e = ^ agrees with the investigations of this earthquake.

3 g

5) The above-mentioned relationship may be equally applicable to a stability calculation for sheetpile quaywalls with anchor plates.

KEYWORDS: Earthquake damage; seaport damage characteristics; site liquefaction; stability

analysis; strong motion accelerograms.

V 11-79
1. INTRODUCTION

Ports and fishing ports in Miyagi prefecture and Fukushima prefecture were damaged by

the 1978 Miyagi-ken-oki earthquake. In order to collect engineering information and to

assist in the reconstruction plan of the damaged facilities, staffs of the Bureau of Ports

and Harbours, the Second District Port Construction Bureau, and the Port and Harbour Research

Institute were immediately sent to the damaged ports [1,2]. In this paper the authors present characteristics of damage on port facilities, strong-motion accelerograms in the ports,

analyses of liquefaction at the actual sites, and comparison of the maximum ground accelerations and the seismic coefficients by the stability analyses of the gravity quaywalls and the

sheetpile quaywalls with anchors. ,^ . , ^, ,

2. DAMAGE OF FACILITIES IN PORTS AND FISHING PORTS ^

'

2.1 Outline of Damage i-rf" -

The public facilities in ports and fishing ports were damaged in both the Miyagi and

Fukushima prefectures. The number of the damaged facilities and the cost of damage in these

ports are shown in Table 1. 98% of the total cost corresponds to damage in Miyagi prefecture. Table 2 shows the detail of damage in Miyagi prefecture. This table shows that damage

corresponding to 90% of total cost happened in Ishinomaki port, and 95% of the cost in this

port corresponds to damage of the quaywalls, except the apron. This cost corresponds to 84%

of sum amount of damage to the port facilities. Damage on port facilities in Fukushima prefecture occurred in Shoma port. Damage in fishing ports occurred mainly at Ishinomaki and

Yuriage in Miyagi prefecture. Consequently damage in ports and fishing ports was concentrated Ishinomaki port and both Ishinomaki and Yuriage fishing ports. Damage to the other

ports was not so severe.

Damage cost comparisons to the port facilities from this earthquake to those by other

recent earthquakes follows: 22 billion Yen in the 1964 Niigata earthquake (M = 7.5), 1.7

billion Yen in the 1968 Tokachi—oki earthquake (M — 7.8), 0.5 billion Yen in the 1978 Nemuro—

hanto-oki earthquake (M = 7.4), and 3.1 billion yen in the 1978 Miyagi-ken-oki earthquake

(M = 7.4). It is judged that this earthquake was not small from view point of the amount of

damage.

VII-80
2.2 Damage to Gravity QuaywalLs ; . . . • ;

.

Many gravity type quaywalls and revetments were damaged by the earthquake. Typical

aspects of the damage were: sliding by gravity wall toward sea, tilting and settlement of

the wall, and settlement, cracking and opening of the joint of the apron as shown in Figure

2. This damage is similar to the damage on gravity structures during past earthquakes.

Swelling of the wall toward the sea was less than 10 cm in most cases and settlement of the

apron occurred to the same extent. Most of the gravity facilities did not lose their function and were used after the earthquake without any trouble. However most of the gravity

quaywalls in Yuriage fishing port were damaged seriously as shown in Figure 3. They were

useless immediately after the earthquake because the walls swelled 122 cm in maximum and the

concrete pavements of the aprons crashed seriously. . .

2.3 Damage to Steel Sheetpile Quaywalls

The extent of damage to steel sheetpile quaywalls may be broadly divided into the

following three groups:

a) Quaywall not damaged or damaged very slightly. Repair is not necessary.

b) Slight repair necessary because settlement and cracking in the pavement of apron

occurred. The main parts of quaywalls such as steel sheetpile, tie rod, and anchor not

damaged

.

c) The front line of the quaywall moved measurably toward sea, the main part of the

quaywall needs repair.

Five steel sheetpile quaywalls were classified as category (c). Figure 4 shows the

plane of the major part of Ishinomaki port which is located at the west side of the mouth

of the Kitakami river. Both the Nakajima port wharf walls of -10 m water depth, and the

Hiyori wharf (-9 ra) slide toward the sea (57 cm maximum) and the wall in Shiomi wharf slid

by 119 cm at maximum. Damage to these quaywalls were also classified category (c). As shown

in Figure 5, the type of anchor used with these quaywalls was steel sheetpile. The Ishinomaki

fishing port is located on the east side of the mouth of the Kitakami river. At the Junbi

wharf (-7 m) a wall with a single vertical pile anchor slide (34 cm maximum). At the back

of these quaywalls settlement of the ground, and sand at cracks and joints in the pavement

was observed. Therefore it was recognized that the backfilling sand had liquefied.

VII-81
Maximum swelling of the quaywall (-3.5 m) at Yuriage fishing port was 106 cm and the extent

of the damage belong to category (c). Although many traces of sand boiling were found around

the port, they were not found any place inside the port.

The extent of damage to quaywalls, except the above-mentioned, were classified into

either category (a) or (b). The general state of damage was the cracking of the pavement

just above the anchor, and the swelling and the settlement of the ground in front of and in

the rear of the anchor, respectively. In these cases the marks of the sand liquefaction were

not found on the ground. When the tie rods were dug up for inspection in several damaged

sites, the soil condition around the anchors were observed but the marks of liquefaction were

not found. The anchor type used with these slightly damaged quaywalls was limited to the

plate shown in Figure 7, or the vertical steel pipepile.

At Nakano wharf in Sendai port the steel sheetpile quaywalls with two kinds of anchor

adjoined each other. This was the quaywall with the steel sheetpile anchor shown in Figure

7, and with the coupled-piles anchor as shown in Figure 8. Cracking, settlement, and swelling of the pavement just above the anchor was found in the former, while nothing happened to

the latter during this earthquake. This fact shows the different earthquake resistance of

the quaywalls caused by the anchors.

When the backfill sand liqufied, the steel sheetpile quaywalls suffered severely. This

fact was similarly observed at Niigata port in the 1964 Niigata earthquake and Hanasaki port

in the 1973 Nemuro-hanto-oki earthquake. When the backfill sand did not liquefy, damage to

the steel sheetpile quaywalls was minimal. This judgement was based on external appearance.

Judging from the change of the ground surface in front and in the rear of the anchor, it was

also estimated that the lateral resistance of anchor almost reached the maximum. If the

anchor breaks down, the sheetpile quaywalls immediately collapse. Even if the sheetpile

deforms slightly, considerable stress remains in sheetpile and tie rod, and it possibly

decreases the stability of the structure. While, many gravity type quaywalls slid and

settled slightly, it was held that the stability of the structure did not decrease significantly. Therefore we cannot absolutely judge the extent of damage from the change of

appearance such as swelling of the wall front toward the sea and the settlement of the wall

capping

.

VII-82
2.4 Steel Pipe Pilled Wharf ; • '

Steel pipe piled wharves have never been damaged by past earthquakes, nor were they

during this one. The gravity quaywalls and the steel sheetpile quaywalls were severely

damaged in the Ishinomaki port and the Yuriage fishing port, however, the steel pipe piled

wharves neighboring them were undamaged except for the retaining wall behind the wharves.

Figure 9 shows the -3.5 m wharf in Yuriage fishing port which was reconstructed in front of

the gravity type quaywall. It is thought that the steel pipe pile resisted the large

horizotal force through the access bridge during the earthquakes, and at the Ohte wharf

in Ishinomaki port. Judging from the foregoing, it was confirmed again that the steel pipe

pile wharf are very earthquake-resistant.

3. STRONG-MOTION EARTHQUAKE RECORDS IN PORT AREAS , , .

3.1 Maximum Ground Acceleration in Ports

The earthquake triggered 34 strong-motion accelerographs in the recording network

appeared by the Port and Harbour Research Institute [3]. The ports where the observation

stations exist are shown in Figure 10 with Arabic numerals. The maximum accelerations in

each port are also listed up in Table 3. Abbreviated names of stations in Table 3 show the

name of the site and the type of accelerographs, that is, "S" indicates SMAC-B2 and "M",

ERS-B (or ERS-C), respectively. The largest peak ground acceleration in the port areas

during this earthquake was recorded at Shiogama port as 280 gals which is larger than the

233 gals recorded at Hachinohe port in the 1978 Tokachi-oki earthquake. The ground condition of the observation site at Shiogama port is fill from surface to -3.4 m (N < 3), tuff

clay to -10.6 m, coarse sand with silt (N > 3) to -13.75 m, sandy tuff to -14.90 m, and

tuff under -14.9 m (N < 50). The maximum acceleration of 170 gals was recorded at Ofunato

port which is almost the same epicentral distance as the Shiogama port.

3.2 Major Wave Forms and Their Response Spectra

Figures 11 and 12 show the main aspects of the acceleration records at Shiogama port

and Ofunato port, respectively. Figure 13 shows the displacements (the EW components) which

were calculated by integrating the acceleration records. The method of integration was proposed by one of the authors. [4] The filter cut-off frequency employed in the integration

is shown as Fc in Figure 13

.

VII-83
The response spectra of the accelerograms (EW component) at Shiogama port and Ofunato

port are shown as solid lines in Figures 14 and 15. Spectra indicated by dotted lines were

obtained from the records of past earthquakes. These earthquakes were rather small (M <

6.4), while the earthquake of February 20, 1978 is not included. According to Figure 15,

spectra calculated by accelerograms in rocky ground are similar in shape. Spectra in this

earthquake (Figure 14), however, are different in shape because the surface layer of about

15 m in thickness responded in a different manner for very strong acceleration. •

3.3 Predominant Direction of Ground Motion

A predominant direction did not clearly appear from the orbit of the two horizontal

components of the acceleration records. However, orbits of the displacement records did

show the predominant direction of the ground movement at the stations which were located

close to the source region of rather simple ground condition. Orbits at Shiogama, Ofunato,

Miyako and Ishinomaki are shown in Figures 16 through 19. An acceleration record at

Ishinomaki was observed at Kaihoku-bashi which is located close to Ishinomaki port and it

is offered by the Public Work Research Institute. [5] Displacements were obtained from the

calculation by the source model in which some parameters proposed by Seno [6] and the temporary values are as shown in Table 4. In this calculation of the displacement, the amplification of the surface layers was not considered. The orbits by this method of calculation

are also indicated in Figures 16 through 19. Comparing the two kinds of orbits, they agree

well on the predominant directions of ground motion, but the absolute quantity of displacement are different. The predominant directions by the observed records in each port, and the

fault plane of the earthquake, are shown in Figure 20.

'

,

' ' -

4. CASE STUDY OF LIQUEFACTION

"

4.1 Damage of Hiyori Wharf in Ishinomaki Port

J There were steel sheetpile quaywalls with -9 m and -10 m water depths at Hiyori wharf

as shown in Figure 4. At the part of the wharf with a -9 m water depth, it is clear that

subsoil behind the wall liquefied. This judgement is based on the evidence that the pavement settled remarkably, and sand arising from cracked pavement and from the openings between

the wall capping and the pavement. However, at that part which was of -10 m water depth the

wall swelling toward the sea and the settlement of the ground behind the wall did not occur,

VII-84
it was judged that the backfill did not liquefy. Both the quaywalls at the Hiyori wharf are

adjacent to each other, and it is supposed that the intensity of earthquake motion were

almost the same at the base of these sites. Therefore, the reason why liquefaction happened

at one site and not at the other was investigated. ' "" ' >• • - "

4.2 Subsoil Condition and Damage of Quaywall

N-values of the subsoil at four sites along the wall front of the Hiyori wharf are shown

in Figure 21. The data at CB18, 14 and 10 were obtained by soil exploration after the earthquake, and it is judged that they suffered the disturbance due to ground vibration and liquefaction. However comparison of N-values before and after the earthquake at Shiomi wharf did

not show any difference between them. Therefore, it was assumed that this fact was also

applicable to the subsoil at Hiyori wharf. The values at No. 5 is the result of standard

penetration test prior to the earthquake. It was judged that the small N-values at CB-18,

14 and 10 were caused by the ejector for making sand piles in foundation improvement work to

lower layer. Figure 21 shows the swelling of the wall front line toward the sea and the

settlement of the asphalt pavement 3 ra behind the capping. From this figure it is apparent

that the quaywall was damaged at -9 m water depth and to the adjacent part of it. Figure 22

shows the grain size distribution curves of soil at the depth of -1.0 to 1.4 and -3.0 to

3.4 m at CB18 , 10 and No. 5, respectively. In Figure 22 the values at CB18 and 10 were also

obtained from the result before the earthquake. Judging from Figure 22, the grain size distribution of the sand at any places in this wharf were almost same.

4.3 Investigation of Subsoil Condition Against Liquefaction

It was clear in the previous section that N-values of the subsoil were different in the

liquefied and non-liquefied sites. In order to make this fact clear, N-values at each site

are plotted in Figure 23. This figure shows the clear different of N-values between them.

This difference may be the reason why the subsoil did or did not liquefy. .

Dynamic response during the earthquake was estimated from the CB14 site which was modeled

into the six lumped mass system. Consequently, it was estimated that the maximum acceleration was about 250 gals at ground surface and the average value of the maximum shear stress

in the -4.7 m to -13.9 m sandy layer was almost 0.7 x 10"-^. The number of cyclic waves was

six when the shear stress reached half the maximum value. As the grain size distribution at

Che site are similar to the Bandai-jima site in Niigata port, the liquefaction potential at VII-85
this site was estimated by the results of dynamic tri-axial test for Bandai-jiraa sand.

According to this investigation the site was judged to have a tendency to liquefy during

the earthquake. Although some input data were possibly inexact in this investigation, the

estimation agreed reasonably with the fact of liquefaction at the site.

In the design of port structures the possibility of liquefaction is commonly based on

the grain size distribution and N-value [7,8]. If the grain size distribution for uniform

soil is in the range of (B) in Figure 22, it is considered the subsoil might possibly

liquefy under certain conditions. It is in the range of (A), the soil is quite certain to

be in the bounds of possibility of liquefaction. The relationship between the critical

N-value for liquefaction and the maximum ground acceleration are based on the field experience in past earthquakes and the laboratory tests. If the N-value of the subsoil below the

ground water table is smaller than the critical one, it is judged that the subsoil may

liquefy during an earthquake. Where the maximum acceleration at ground surface is 250 gals

as in the previous calculation, the critical N-value becomes 14. When the critical N-value

was compared to those of sites at the Hiyori wharf, it was larger than those at liquefied

sites and smaller than those of non-liquefied sites except in very shallow layers. Therefore this procedure for liquefaction estimation was considered to be one of some use. However, as the result in this method of estimation depends upon the maximum acceleration of

ground surface, input data must be carefully selected for calculating the ground response.

5. STABILITY ANALYSES OF GRAVITY QUAYWALLS AND SHEETPILE QUAYWALLS

5.1 Facilities

The stability of facilities in several ports were analysed using the current design

standard [7] and the results were compared to the extent of damage and the severity of

ground motion during the earthquake. The port facilities investigated number 60 quaywalls

in 11 ports, that is, 36 gravity quaywalls in 10 ports, and 24 steel sheetpile quaywalls in

7 ports. The classification of these facilities are listed in Table 5 in terras of type of

structure and extent of damage. The classification of extent of damage in the table is the

one proposed by Kitakima and others for considering the damage records of port facilities

since the 1964 Niigata earthquake [9,10]. The steel sheetpile quaywalls which were damaged

by liquefaction of the backfill was not discussed in this chapter.

VII-86
5.2 Estimation of Maximum Ground Acceleration in Ports '

' '- •

The large values of ground acceleration were recorded around Sendal, Shiogama and

Ishinomaki cities. The earthquake triggered 34 strong-motion accelerographs of the network

of the Port and Harbour Research Institute. Two of them were installed on the ground of

Shiogama port and Ofunato port. When earthquake resistant design is investigated through

analyses of damaged structures, the severity of the earthquake motion acting on the structures must be estimated first of all. Therefore, the maximum ground accelerations at the

ports, except the two ports above, was estimated by calculating the ground response during

the earthquake. The epicenter location was given by the Tohoku University as shown in Figure

1 and the ground motions were calculated by the SHAKE program. As the incident wave form the

acceleration records which were used and were observed on the rocky site at Ofunato port and

Kaihoku Bridge. The source region and the maximum acceleration of base rock motion in each

port were determined using the same procedure which was described in the previous report

[11]. Maximum ground accelerations in each port were estimated as shown in Figure 24. The

9 expression indicates that the estimated acceleration exists in the range. As the facili6 ties in Kinkazan, Hagihama, Onagawa and Okatsu were directly built on the rocky ground, the

ground accelerations were assumed to be same as those at the base rock.

5.3 Stability Analysis of Gravity Quaywalls, and Relationships Between Seismic Coefficient

and Maximum Ground Acceleration

The stability of the gravity quaywalls and bulkheads were analysed using current design

procedure, and the seismic coefficients which correspond to the severity of the seismic

effect were estimated by the way reported in the previous paper [11]. The outline of the

method is shown in Figures 25 and 26. For example, the stability of the sliding, the overturning, and the bearing capacity at the base of structures for each port facility were

analysed, and the seismic coefficients which give the unit value of the safety factor were

obtained. Judging from the characteristics of the damage to facilities, the causes of damage

were estimated for the above-mentioned factors. If the facility was damaged due to one of

these factors, the actual seismic coefficient corresponding to the seismic effect may be

larger than the calculated seismic coefficient and it is plotted by the expression

^

as shown

in Figure 26.1. If the facility was not damaged, the actual seismic coefficient may be

smaller than the calculated one and it is plotted by the expression ^. Considering these

VII-87
results, the seismic coefficient acting on the facilities during the earthquake in Shiogama

port is judged to fall in the range of the dotted lines. In case of Kinkazan port in Figure

26.2, the seismic coefficient is estimated to exist in a more limited range. In case of

Onagawa port in Figure 26.3, the upper boundary is only estimated because no facilities were

damaged. The seismic coefficients or their ranges were determined in 10 ports in this

manner . ....

The relationships between the maximum ground acceleration and the seismic coefficient

at each port were estimated and the results are shown in Figure 27. In the figure, the '

curved line indicates the proposed relationship which was obtained by the same procedure

to the gravity quaywalls in the past earthquakes, and the straight line indicates the relationship for the rigid body on the solid foundation. Even though the range of the estimated

values in Shiogama, Matsushima and Hagihama ports range somewhat widely, the results of the

estimation show similar characteristics to the proposed curve. However, the maximum ground

accelerations in Ofunato and Okatsu port are smaller than 200 gals, and it seems more suitable that the straight line indicates a better relationship than the curved line. As arrows

point downward in Sendai and Ishinomaki ports, the expected seismic coefficients in these

ports do not always contradict the curved line. Consequently, it is concluded that in this

earthquake the relationship between the acceleration and the seismic coefficient for gravity

type quaywalls is similarly expressed by the relationship determined by damage records in

past earthquakes.

. •'.<:. 'cii'c. ':

'

5.4 Stability Analysis of Steel Sheetpile Quaywalls, and Relationships Between Seismic -•

.

Coefficient and Maximum Ground Acceleration ~ ' -- ; -;:»=./

The current earthquake-resistant design suggestions for steel sheetpile quaywalls are

analyzed for the following items, that is, the embedded length and stress of the sheetpile,

the stress of the tie rod, and the stability of the anchor. In this chapter the critical

seismic coefficients which provide the unit value of the safety factor were also estimated.

In the stability analysis the active and passive earth pressure on the wall and the lateral

resistance of the anchor piles are the most important items, but the technical level of the

design for the latter is not as reliable when compared to the former. Therefore, investigation was limited to the two items but was not considered for all items in the case of the

gravity quaywalls. .

. ^

VI 1-88
All of the critical seismic coefficients of the bending stress of the sheetpile and the

tensile stress of the tie rod were larger than 0.25. These results are not in contradiction

with the fact that the sheetpile and the tie rod did not break down in the actual event.

The critical seismic coefficients for the embedment of sheetpile were compared to the

maximum ground acceleration in Figure 28. As no damage due to the lack of the embedment

length occurred during this earthquake, only the upper limit of the seismic coefficient was

estimated. The results of the stability analyses in past earthquakes also show a similar

tendency in the figure. It might be concluded that the seismic coefficients on the necessary

length of embedment of sheetpile did not increase from increasing the ground acceleration.

But, the results were not sufficient to make clear the relationship between them.

As the typical damage to the steel sheetpile quaywalls was the swelling of the wall

toward sea due to the movement of the anchor, the most important item in the analyses was the

stability of the anchor. The relationship between the maximum ground acceleration and the

seismic coefficient for the anchor plate, which is the most common type of anchor, was estimated in Figure 29. The results from 4 ports and those from previous investigations are also

shown in the figure. According to the figure the results from 3 ports, except for the

Yuriage fishing port, are close to a straight line. The horizontal resisting capacity of the

anchor is expressed as the difference between the passive earth pressure in front of the

anchor and the active earth pressure in the rear, and the stability of the gravity quaywalls

are mainly influenced by the difference in pressure. Judging from these facts, the relationship between the maximum ground acceleration and the seismic coefficient for the anchor is

possibly similar to that for the gravity quaywalls. The stability of the anchor by the single

vertical pile and the sheetpile was investigated next. The stability of these anchors appears

to result from the lateral resistance of flexible members in the soil, however, this is not

altogether clear in the present stage. If the swelling of the wall is considered to be

equivalent to the lateral displacement of the anchor head, actual damage is ten times larger

than the estimated one. This fact has been pointed out [10] and remains a technical problem

to be solved. Stability of the coupled-piles anchor are determined mainly by the pulling

resistance of the coupled piles. Since the seismic coefficient was obtained only from the

upper limit at 2 ports, the relationship between the maximum ground acceleration and the

coefficients can not be discussed.

VII-89
6. CONCLUSIONS

The 1978 Miyagi-ken-oki earthquake (M = 7.4) caused damage to port facilities. As a

result of the field investigations and the analyses, the following lessons concerning earthquake engineering were learned.

1) Strong-motion accelerograms were recorded by the observation network of the Port and

Harbour Research Institute. The largest peak ground acceleration was recorded at Shiogama

Port as 280 gals.

2) Port facilities were damaged seriously when the backfill liquefied. Otherwise, they

suffered only lightly.

3) Current procedures for estimating liquefaction potential can lead to appropriate

judgements for actual sites.

4) The relationship between the seismic coefficient and the maximum ground acceleration

I,

a

for gravity quaywalls: e = — (—) agrees with the investigations of this earthquake.

3 g 5) The above-mentioned relationship may be equally applicable to a stability calculation for sheetpile quaywalls with anchor plates.

REFERENCES

'

[1] Tsuchida, Noda, et al. , "The Damage to Port Structures by the 1978 Miyagi-ken-oki

Earthquake," Technical Note of the Port and Harbour Research Institute, in press.

[2] Tsuchida, H., "Damage to Port Facilities Due to the 1978 Miyagi-ken-oki Earthquake,"

Minato-no-bosai, September 1978.

[3] Kurata, E., lai, S., Yokoyama, Y., and Tsuchida, H., "Strong-Motion Earthquake Records

on the 1978 Miyagi-ken-oki Earthquake in Port Areas," Technical Note of the Port and

Harbour Reserch Institute, No. 319, March 1979.

[4] lai, S., Kurata, E., and Tsuchida, H., "Digitization and Correction of Strong-Motion

Accelerograms," Technical Note of the Port and Harbour Research Institute, No. 286,

March 1978.

[5] Iwasaki, T. , Wakabayashi, S., Kawashima, K., and Takagi, Y. , "Strong-Motion Earthquake

Records from Public Works (No. 2)," Bulletin of Public Works Research Institute, Vol.

33, October 1978.

[6] Hirasawa, T. , "Seismo logical Characteristics of Miyagi-ken-oki Earthquake," Report to the Technical Meeting of Earthquake Engineering Committee, the Japan Society of Civil

Engineering, November 1978.

[7] The Japan Port and Harbour Association, "Engineering Requirements on Port and Harbour

Facilities and Its Application," March 1978.

[8] Tsuchida, H., and Hayashi, S., "Estimation of Liquefaction Potential of Sandy Soils,"

Third Joint Meeting, U.S. -Japan Panel on Wind and Seismic Effects, UJNR, May 1970.

VI 1-90
[9] Bureau of Ports and Harbours, First District Port Construction Bureau, and Port and

Harbour Research Institute, "Damage to Port Structures by the Niigata Earthquake (First

Report)," September 1964.

[10] Kitajima, S., and Uwabe, T. , "Analysis of Seismic Damage in Anchored Sheet-piling

Bulkheads," Report of the Port and Harbour Research Institute, Vol. 18, No. 1, March

1979.

[11] Hayashi, S., Noda, S., Uwabe, T. , "Relation Between the Seismic Coefficient and Ground

Acceleration for Gravity Quaywalls," Eighth Joint Meeting, U.S. -Japan Panel on Wind and

Seismic Effects, UJNR, May 1976.

VI 1-91
Table 1 Damage of Public Port Facilities

Port Managements Body Number of Damaged

Facility

Amount of Money (xlOOO Yen)

Mlyagi Prefecture A5 3,079,500

Fukushima Prefecture U 58,500

Total 49 3,138,000

Table 2 Damage of Public Port Facilities in Miyagi Prefecture

( ) : number of facility, unit : 1000 Yen

Port Quaywall

Quaywall

(Apron) Jetty

Jetty

(Apron) Bulkhead

Bulkhead

(Apron) Road Dock

Railway

Breakwater

Sea Wall Revetment

Parapet

Wall

Total for

Port

Sendal

(2) 79,300(3) 25,500

(2) 2,800

(7) 104,600

Shlogama

(1) 1,200

(6) 117,100

(2) 13,200(1) 3,200(1) 1,000(1) 8,100(1) (13) 143,800

Matsushima

(1) 700(2) 11,800

(1) 1,900

(4) 20,700

Ishinomaki

(3) 2,650,000(2) 14,100

(4) 30,900

(1) 1,100

(3) 48,500

(1) 31,100

(14) 2,775,700

Okatsu

(1) 9,000

(2) 1,600(1) 1,000

(4) 11,600

Hagihama

(1) 9,500

(1) 9,500

Kinkazan

(4) 10,600(4) 10,600

Total for Facility (3) 2,650,000

(4) 93,400

(1) 7,000

(1) 1,200

(14) 158,000

(2) 13,200

(8) 59,600(1) 1,100(3) 2,600

(4) 56,600

(5) 36,800

(1)

(Sum Total)

(47) 3,079,500

VII-92
Table 3 Maximum Accelerations in Ports

No

.

S nation

Record Max. Acceleration (Gal)

Installation

Condit ion Epicentral

Distance (km) Abbreviated Name Number NS EW Tin

16 Yamashita-dai6-S on structure 37A S-1187 20 19 9 16 Keihin-j i-S on ground 376 S-1188 31 25 10

16 Yamashita-hen-S on ground 374 S-1189 25 11 Onahama-S on ground 174 S-1191 44 51 19

6 Aomori-S on ground 319 S-1192 21 23 9 Sakata-S on ground 220 S-1193 30 34 19

lA Shinagawa-S on ground 354 S-1194 9 10 5 13 Chiba-S on ground 345 S-1195 26 29 5 19 Shimizu-miho-S on ground 478 S-1196 13 13 10

19 Okitsu-S on ground 476 S-1197 4 4 3 19 Shimizu-koj yo-S on ground 480 S-1198 15 15 6 24 Akita-S on ground 249 S-1200 25 24 13

10 Shiogama-kojyo-S on ground 100 S-1201 270 280 169

7 Hachinohe-S on ground 273 S-1202 63 61 30

22 Niigata-j i-S on ground 274 S-1203 25 1"

8 Miyako-S on ground 166 S-1204 151 115 50

20 Nagoya-zokan-S on ground 584 S-1205 6 4 3 12 Kashima-zokan-S on ground 281 S-1206 40 Q

3 Tomakomai-S on ground 500 S-1207 9 8 1 18 on ground /. t;Q HDy S-1209 23 36 8

9 Of unato-bochi-S on ground 103 S-1210 141 170 60

9 Of unato-bo-S 1i-UmJ S-1211 OIK lUb

4 Muroran-S on ground 475 S-1217 4 A 1 17 i\u K.en n on ground 392 M-216 8 7 16 Yaraashita—hen-M on ground 374 M-217 1 7 X / 7 16 Yamashita-dai7-M on structure 374 M-218 1X 15 Kawasaki -da i5- ko-M on structure 364 M-219 34 over —

15 Kawasaki -da 15- chi-M on ground 364 M-220 24 27

21 Yokkaichisekitan-M on structure 609 M-221 5 15

19 Shimizu-sekitan

-M

on structure 481 M-222 23 20

20 Inae-yaita-M on structure 583 M-223 6 7 5 Hakodate-M on ground 422 M-224 12 14 7 2 Tokachi-M on ground 469 M-225 10 10 4

1 Hanasaki-M on ground 638 M-226 4 4 2

Table 4

Parameters of

Source Model

1. Width of Fault '

2. Length of Fault

3. Slip Angle

4. Slip Vector

5. Slip Direction ^

6. Slip Displacement

(1 6 : after Seno"'' )

7. Rise Time

8. Rupture Velocity (in length)

9. Rupture Velocity (in width)

10. Velocity of P Wave

11. Velocity of S Wave

VII-93

6)

80 km

30 km

-20°

N10°E

N65°W

I. 7 m

2.0 sec.

II. 6 km/s

3.1 km/s

6.0 km/s

3.5 km/s
Table 5 Classification of Facilities

Extent Gravity Structure Sheetpile Structure

of Block Caisson L-Block Total

Plate Sheetpile Single pile Coupled pile Self Standing

Total

Daiaage Anchor Anchor Anchor Anchor Sheetpile

0 11 4 0 15 5 1 2 7 0 15

I 17 0 1 18 1 1 2 0 1 5

II 2 0 0 2 1 3 0 0 0 k

III 1 0 0 1 0 0 0 0 0 0

IV 0 0 0 0 0 0 0 0 0 0

Total 31 A 1 36 7 5 U 7 1 2A

I

j

Table 6 Classification of Extent of Damage

Extent Swelling of Wall Settlement of Settlement of

of Maximum Average Apron Coping

Damage (cm) (cm) (cm) (cm)

0 0 0 0 0

I 0~30 under 10 under 20 under 30

II 30-100 10-60 under 50 under 40

III 100-400 60-120 50-100 under 50

IV over 200 over 120 over 100 over 50

VII-94
Fig. 1 Location of Epicenter, Source Region and Ports

38*»N

4I*»E

Miyako

Kama i shi

Ohf unato

Kesennuma •

Okatsu

I shi noma ki Matsushima • •/ Onaqawa

^^^""^ama y^'^Q^ma^

Sendai TP

Vuriaqe

Sohma •

0 20 40 60km

1 I I I

[x] :Epicenter(by Tohoku University)

(142.23°E,38.17°N,D=25Km)

r :Radius of source region by eq.(l)

14 rE I42«»E

I43*E

40'*N

VII-95
Rubblestone mound scasoQCOe.

Fig. 2 Damaged Gravity Quaywall (Shiogama Port)

Fig. 3 Damaged Gravity Quaywall (Yuriage Fishing Port)

VII-96
Fig. 5 Hiyori -9 m Wharf (Ishinomaki Port)

VII-97
V H.WL 1-1.50

L.W. L ±0.00

I'li

11'

-7.50

tz.oo

Fig. 6 -3.5 m Wharf (Yuriage Fishing Port)

Crack on concrete apron pavement

Apron

+ 1.3 Vftt „ H.W.L H.4 ? Ffl 2"LW.L ^0.44 L-JT

•10.0

-Z4.0

Steel pipe pile, type S.T.K.50

-to.o

Tie rod

, 1.16 O" ±0.0 e t c l.fc" <^ Sheetpile anchorage

itiuubu TBBsnrn

Sheetpile, type IH , 8 = 12.0

Fig. 7 Nakano -10 m Wharf (Sheetpile Anchor ) (Sendai Port)

VII-98
Fig. 9 -3.5 m Wharf (Yuriage Fishing Port)

VII-99
Fig. 10 Location of Observation Stations

3 GO

S-1201 Shiogama-kojyo-S

D- L5 3G0- c° D-

*->

S 300. u

0.

300.

DOKN

WEST

J—

ic. Time (seconds)

10.

Fig, 11 Accelerograms (Shiogama-kojyo-S)

VII-IOO
Undamped natural period (seconds)

Fig. 12 Accelerograms (Of unato-bochi-S)

Undamped natural period (seconds)

13 Displacements of Shiogama-kojyo-S and Ofunato-bochi-S

E-W Component

VII-101
Undamped natural period (seconds)

Fig. lA Response Spectra of Shiogama-kojyo-S

Undamped natural period (seconds)

Fig. 15 Response Spectra of Of unato-bochi-S

VII-102
(observed) (Estimated) (Observed) (Estimated)

Fig. 16 Locus (Shiogama) >

. Fig. 17 Locus (Ofunato)

(Observed) (Estimated) (Observed) (Estimated)

Fig. 18 Locus (Miyako) Fig. 19 Locus (Ishinomaki)

VII-103
Fig. 20 Predominant Direction of Ground Motion and Locatio,

of Fault

Fig. 21 N-values and Damage (Hiyori Wharf)

VII-104
22 Grain Size Distribution Curves (Hiyori Wharf)

N-value

0 20

-10

3ringCB18

CBU

CBIO

No.

5

23 Comparison of N-value (Hiyori Wharf)

VII-105
Fig. 24 Maximum Ground Acceleration in Ports

A Port facility

Port focility

A Port focility -

BC

D

A Port facility

Stability calculation of ttie sliding opplymg the present cJesign method

Judgement whether the sliding happened or not

Stobility calculation of the bearing capacity at the bose of structures applying

the present design method

Judgement whether the shortoge

of beoring copocity happened or not

Stability calculation of the overturning applying the present design method

Range of the

seismic coefficient

Range of the seismic coefficient

Judgement whether the overturning happened or not

Range of the

seismic coefficient

Seismic coefficient of the sliding for

all port focilities

Seismic coefficient of the bearing capocity for

oil port facilities

Seismic coefficient of the overturning

for all port

focilities

Estimating of the seismic coefficient of quay wol Is around

port oreo

Fig, 25 Procedure for Estimation of Seismic Coefficient

VII-106
o.3r

I

^0.2 I

I

Si

0.1

I —i- I

O Sliding

0.3

I

I

ti 0.1

Kinkazan port

-^--r-:--T^l

I

o Sliding

i i i I 1 I I } I

I n

i

1 i !

26.1 Range of Fig. 26.2 Range of

Seismic Coefficient Seismic Coefficient

(Shiogama Port) (Kinkazan Port)

0.3r

3

'

0.1

seismic coefficie

Sliding Overturning

2 ^ s i nn

Fig. 26.3 Range of

Seismic Coefficient

(Onagawa Port)

VII-107
0.3r

^ ^

! sh 1 nonia k i jenda

i

0 100 200 300 400

Ma', horizontal acce 1 era 1 1 on CX ( gO-l

)

Fig. 27 Seismic Coefficient and Maximun: Ground Acceleration

(Gravity Structure)

Max, horizontal acceleration (Gal)

Fig. 28 Seismic Coefficient for Embedment of Sheetpile and Maximum

Ground Acceleration (Sheetpile Structure)

VII-108
Fig. 29 Seismic Coefficient for Anchor Plate and Maximum Ground

Acceleration (Sheetpile Structure)

VII-109
WOOD DIAPHRAGMS IN MASONRY BUILDINGS

Mihran S. Agbabian

Agbabian Associates

El Segundo, California

ABSTRACT

Reports of damage from past earthquakes indicate that wood diaphragms supported on

masonry walls have suffered little or no damage during excitation. However, the supporting

masonry walls have suffered degrees of damage varying from minor tension cracks to separation

from the diaphragms and complete collapse.

The interaction of a diaphragm with the masonry walls has a critical effect on the

dynamic behavior of the structure. Numerical analyses of plywood, diagonal- and straightsheathed floor and roof diaphragms indicate that highly nonlinear and hysteretic load-deflection relationships are needed to describe the response of wood diaphragms. Experiments are

planned to improve the analytical model by obtaining performance characteristics of wall

anchorages to diaphragms as well as deformations of diaphragms under cyclic excitations.

Finally, candidate retrofit/strengthening techniques will be incorporated into typical wood

diaphragms, and the tests will be repeated in order to develop structural systems that will

survive disastrous earthquakes.

KEYWORDS: Seismic hazard; wood diaphragms; unreinforced masonry; dynamic response;

analytical model.

J

VII-110
INTRODUCTION

Wood diaphragms supported on masonry walls are found in buildings located in various

seismic zones. Reports of damage from past earthquakes indicate that such diaphragms have

suffered little or no damage during earthquake excitations. However, the supporting masonry

suffered degrees of damage varying from minor tension cracks to separation frora the diaphragms and complete collapse. The diaphragm's response and masonry wall damage may be directly related because of the effect of diaphragm deformation on the out-of-plane motions of

the masonry wall. An analytical model is used in this paper that includes a nonlinear hysteretic element which simulates the behavior observed in cyclic tests of wood diaphragms.

The results of the analysis are compared with corresponding results obtained with linear

elastic and nonlinear elastic assumptions. Forthcoming experiments, where several diaphragms

will be statically and dynamically tested, will provide new data for correlation with these

analyses.

ANALYTICAL MODEL

A typical one-story unreinforced masonry (URM) building shown in Figure 1 was selected

for analysis. The building consists of a wood-diaphragm roof supported on four sides by URM

walls. The earthquake ground motion was assumed to be directed normal to the side wall.

The analytical model shown in Figures 2 and 3 assumes that the end walls are rigid. A

preliminary study of wall overturning due to seismic loading [1] has indicated that the

earthquake input motions will be transmitted from the foundation level (Levels C and D in

Figure 2a) to the top of the end walls (Levels A and B) with little modification. The side

wall was assumed to crack along the foundation when subjected to the out-of-plane earthquake

motions. Since the cracked wall will have little lateral stiffness, only its weight has been

included

.

The roof diaphragm/wall system is represented as a lumped parameter model and is shown

in Figure 2b. The roof diaphragm is divided into several segments (eight segments are shown

in Figure 2b), the out-of-plane wall mass is distributed along both edges, and the model is

subjected to earthquake ground motion at its ends. The diaphragm is also assumed to act as a

deep shear beam. The lumped parameter model is shown in Figure 3. The earthquake motions at

each end are taken to be identical, and due to symmetry only a half model is required.

VII-111
The nonlinear internal springs shown in Figure 3 represent the shear stiffness of the

diaphragm, and the internal dampers represent viscous damping of the roofing material. The

earthquake input motion is described by coordinate X]^ , and the independent degrees of freedom (DOF) of the diaphragm/ wall system are described by coordinates X2 through X5. The displacement gages shown connecting the DOF X2 through X5 with DOF X]^ are used to monitor the

relative out-of-plane motions of the top and bottom of the side wall. This relative motion

will serve as an indicator of wall instability. The analyses were performed using the STAR/

III computer program. [2]

DIAPHRAGM STIFFNESS ELEMENT . : .

"

Cyclic tests [3] show a hysteretic behavior for plywood diaphragms as shown in Figure 4

A review of numerous tests of wood diaphragms [6-20] also shows that the behavior of these

diaphragms is highly nonlinear. Most of these tests were, however, noncyclic.

Based oa the available test results, a nonlinear hysteretic model for the wood diaphragms has been developed as shown in Figure 5. The overall force-deflection envelope is

shown in Figure 5a, and a typical cyclic load path for the spring is shown in Figure 5b. It

can be seen that this load path is a multilinear version of that shown in Figure 4. The

unloading slope as well as the residual force value may be specified to match experimental

data. The test data indicate that the envelope can be adequately represented by a secondorder curve of the form

F,.e ,. , , , F(e) = = for compression (positive e)

K. + e .

. 1

F(e) = for tension (negative e)

K. - e

. j- where

F(e) = Spring force

e = Spring deformation

Fy = Ultimate capacity of spring at large e

Ki = Initial spring stiffness

VII-112
ANALYSIS OF TYPICAL ROOF DIAPHRAGMS

Dynamic analyses using the model described in the previous section were conducted for

two types of earthquake ground motions. The inputs correspond to the highest seismic areas

of the United States [4] where effective peak acceleration is 0.40 g, as shown in Figure 6

for a 5% dampered response. The SOOE component of the 1940 Dl Centro record, representing a

distant event, was scaled (1.25 scale factor) to the 0.40 g level, and its response spectrum

is shown in Figure 6a. The N69W component of the 1971 Castaic record, representing a nearby

event, was scaled (1.80 scale factor) to the 0.40-g level, and its response spectrum is shown

in Figure 6b. The scaled displacement, velocity, and acceleration records are shown in Figures 7 and 8 for the scaled El Centro and Castaic records, respectively.

The cases run are shown in Table I. Two basic types of roof diaphragms were analyzed;

namely,

° 1/2-ia. plywood, unblocked, and roofed

° Diagonal sheeting, roofed '

Each type was subjected to the scaled El Centro and Castaic input motions. In addition, each

type was analyzed using four material properties:

° Linear elastic

° Nonlinear elastic (envelope curve of Figure 5a)

° Nonlinear hysteretic (Figure 5b)

° Nonlinear hysteretic (Figure 5b) with F^ = one-half the value of F^ of the

immediately preceding case

The last case shows the sensitivity of the response to a variation in the ultimate

force, F^. Although it is clear that the roofing material will provide some degree of viscous damping, all of the analyses were performed without viscous damping to clearly show the

effect of the nonlinear hysteretic model assumptions. Peak responses for all of the cases

are given in Table I.

It can be seen that the nonlinear elastic cases show a general reduction (relative to

the linear elastic cases) in peak responses at DOF 5, an attenuation of the peak relative

deformation at the midlength of the diaphragm, Spring No. 12, and significant reductions in

the maximum shear force at the end wall. Spring No. 1. The nonlinear hysteretic cases show

further significant reductions in all response quantities. Typical results for the plywood

VH-113
roof diaphragm are shown in Figures 9, 10, 11, and 12. Figures 9a, b, and c, and 10a, b, and

c compare the acceleration and displacement response at DOF 5, respectively, for the first

three model types. Figures 11a, b, and c compare the shear force transmitted to the end wall

(i.e.. Spring No. 1).

The relative deformation between the top and bottom of the masonry wall at the diaphragm

midlength (i.e.. Spring No. 12) gives an indication of the out-of-plane motions of the side

wall. Typical results for this relative deformation are given in Figures 12, 13, and 14.

The main effect of the hysteretic model can be seen in these three figures. For example,

comparing Figures 12a, b, and c, it can be seen that not only is the peak deformation

reduced, but the number of repeated loadings at nearly constant peak values is greatly

reduced. It is believed that elastic analyses overestimate the actual response and that consideration of the hysteretic characteristics of the wood diaphragms is necessary for more

realistic response predictions.

FORTHCOMING DIAPHRAGM AND WALL TESTS ,

Data on the dynamic response of wood diaphragms and masonry walls are generally lacking.

Accordingly, a series of tests is planned to provide needed data and to calibrate/verify the

analytical model that has been developed. The test program consists of two related test

series: (1) diaphragms and (2) out-of-plane deflection of URM walls. These tests are briefly described in an accompanying paper. [5]

FURTHER STUDIES AND CONCLUSIONS .

During the forthcoming test program, the analytical model will be refined, calibrated,

and verified for use in predicting the dynamic response of wood diaphragms. In addition,

stability criteria for the response of masonry walls to out-of-plane motions will be established. The final goal will be a reliable method of predicting wall/diaphragm failures in

seismic areas and the development of retrofit measures to prevent such failures.

ACKNOWLEDGMENTS , .

'

The study is being conducted by the joint venture ABK comprising the firms of Agbabian

Associates, S.B. Barnes & Associates, and Kariotis, Kesler and Allys. It is authorized by

the National Science Foundation (NSF) in support of its oneoine earthquake hazard mitigation

VII-114 '
program. Dr. John B. Scalzi is Cognizant Program Manager for NSF. John Kariotis and Albin

Johnson are designing the test elements, and Robert Ewing and Tim Healey are conducting the

analyses

.

REFERENCES

[1] Adham, S. A. and Ewing, R. D. , "Methodology for Mitigation of Seismic Hazards in Existing Unreinforced Masonry Buildings, Phase 1," R-781 5-4610. El Segundo, California:

Agbabian Assoc., March 1978.

[2] Agbabian Assoc. (AA) .

, "User's Guide and Formulation of the Analysis Method for the

STARS/III Code," U-4700-4816. El Segundo, California: AA, (in press) 1979.

[3] Young, D. H. and Medearis, K., "An Investigation of the Structural Damping Characteristic

of Composite Wood Structures Subjected to Cyclic Loading," Technical Report No. 11. Stanford, California: Stanford University, April 1962.

[4] Applied Tech. Council (ATC)., "Tentative Provisions for the Development of Seismic Regulations for Buildings," ATC-3-06. San Francisco, California: ATC, June. 1978.

[5] Agbabian, M. S. , "Mitigation of Seismic Hazards in Existing Unreinforced Masonry Buildings," Proceedings of 11th Joint UJNR Panel Conference, Tokyo, Japan, September 1979.

Additional References for Diaphragm Test Data

[6] Forest Products Lab. (FPL)., "Diaphragm Action of Full-Scale Diagonally Sheathed Wood

Roof or Floor Panels." Madison, Wisconson: FPL, April 1955.

[7] Atherton, G. H. and Johnson, J.W., "Diagonally Sheathed Wood Diaphragm. Corvallis,

Oregon: Oregon Forest Research Lab., October 1951.

[8] Stillinger, J. R. , Johnson, J. W. , and Overholser, J. L. , "Lateral Tests on Full-Scale

Lumber-Sheathed Diaphragms." Corvallis, Oregon: Oregon Forest Products Lab., October

1952.

[9] Stillinger, J. R. , "Lateral Tests on Full-Scale Lumber-Sheathed Roof Diaphragms." Corvallis, Oregon: Oregon Forest Research Lab., October 1953.

[10] Johnson, J. W. , "Lateral Tests on Full-Scale Lumber-Sheathed Roof Diaphragms of Various

Length-Width Ratios. Corvallis, Oregon: Oregon Forest Research Lab., October 1954.

[11] Johnson, J. W. , "Lateral Tests on 12-by-60 ft and 20-by-80 ft Lumber-Sheathed Roof

Diaphragms." Corvallis, Oregon: Oregon Forest Research Lab., October 1955.

[12] Burrows, C. H. and Johnson, J. W. , "Lateral Test on Full-Scale Roof Diaphragm with

Diamond-Pattern Sheathing." Corvallis, Oregon: Oregon Forest Research Lab., October

1956.

[13] Stillinger, J. R. , "Lateral Tests on Full-Scale Lumber and Plywood-Sheathed Roof Diaphragms," in Symp. on Methods of Testing Building Construction, ASTM-STP-166 . New

York: American Society of Mechanical Engineering, 1955.

[14] Oregon Forest Labs. (OFL)., "Lateral Tests on Full-Scale Lumber and Plywood-Sheathed

Roof Diaphragms." Corvallis, Oregon: OFL, March 1956.

[15] Countryman, D. , "Lateral Tests on Plywood Sheathed Diaphragms." Tacoma, Washington:

Douglas Fir Plywood Assoc. Lab., March 1952.

VII-115
[16] Oregon Forest Research Lab. (OFRL)., "Lateral Tests on Full-Scale Plywood-Sheathed Roof

Diaphragms." Corvallis, Oregon: ORFL, October 1953.

[17] Countryman, D. and Colbenson, P., "1954 Horizontal Plywood Diaphragm Tests." Tacoma,

Washington: Douglas Fir Plywood Assoc. Lab., October 1954.

[18] Douglas Fir Plywood Assoc. (DFPA)., "1954 Horizontal Plyv/ood Diaphragm Tests." Tacoma,

Washington: DFPA, January 1955.

[19] Johnson, J.W., "Lateral Test on a 12-by-60 ft Plywood Sheathed Roof Diaphragm." Corvallis, Oregon: Oregon Forest Research Laboratory, January 1955.

[20] Tissel, J.R., "Horizontal Plywood Diaphragm Tests." Tacoma Washington: Douglas Fir

Plywood Assoc., 1966.

• J

'•

VII-116
TABLE I. SUMMARY OF DIAPHRAGM DYNAMIC ANALYSIS RESULTS

Material Model Peak Acceleration

at Hldlengtli JOF 5, i

Peak Velocity

at Hldlength

WIF 5, In./s

Peak DlBplucement

at Hldlength

DOF 5, In.

Peak Relative

Def ormat Ion

—

Peak

Diaphragm

Earthquake

Input Type klps/ln.

F klp»

at Hldlength

Spring 12.

In.

Shear Force Spring 1

.

Hps

Plyvood^ ^ El Centro

Linear

Elastic

25

Plyuood*

El Centro

Nonl Inear

Elaat ic

25 140 1 .48 67.7 17.4 14.2 76.3

P

1

yvood Scaled

El Centro

Nonlinear

Hysteret Ic 25 UO 0.62 35.3 7.2 6.0 45.6

Plywnod*'Scaled*

El Centro

Nonl Inear Hysteret ic 25 70 0.45 27.3 8.4 5.9 31.7

Plywood

Castalc

L Inear Elaatlc

25 2.2 98.2 17.4 13.7 166.0

Plyvood ' '

Scaled^

Castalc

Nonl inear Elastic25 UO 2.0 77.5 13.0 10.9 72.3

Plyvood

' '

Castalc

Nonl Inear Hysteretlc

25 140 0.53 35.4 8.0 6.1 52,3

Plywood^" Scaled^ CastalcNonlinear

Hysteret Ic

25 70 0.41 30.5 8.3 5.4 36.5

(2) Diagonal

Sheeting

Scaled

El Centro

Linear

Elastic

8.3 1 .6 103.0 25.4 23.6 88.6

Diagonal Sheeting Scaled

El Centro

Nonl Inear Elastic8.3 120 0.6B 55.9 19.8 17.9 42.0

Diagonal'^*

Sheet Ing Scaled*El Centro

Nonl Inear Hyateretlc8.3 120 0.50 43.6 12,0 13.5 34.7

Diagonal*^*

Sheet Ing ScaledEl Centra

Nonl Inear Hysteret Ic 8.3 60 0.38 31.2 12.4 13.9 24.6

Diagonal

Sheeting

Scaled^

Castalc

Linear

Elast ic 8.3 1 .66 91 .8 19.0 1 7 .9 76,9

(2) Diagonal

Sheeting

Scaled^ CastalcNonl Inear Elastic

8.3 120 0.99 42.6 13.8 9.5 30.0

Diagonal

Slieeting Scaled^Casta Ic NonlinearHysteretlc

6.3 120 0.48 36.4 11.7 9.4 26.5

(2) Diagonal

Sheet Ing Castalc

Nonl Inear

Hya teretlc

8,3 60 0.36 28.2 1 1 .8 9.4 22.6

*E1 Centro, 1940, SOOE. scaled by 1.25 *'*l/2-ln. plywood, unblocked, roofed (1 In. - 2.51 cm)

^Castalc. 1971, N69U, scaled by 1.80 *^*Dlagonal sheeting, roofed

"

VII-117
1-13" SOLID WALL

(CONCRETE MASONRY)

50'

-100'-

DIRECTION OF EARTHQUAKE MOTIONS

PLAN

J.3'

12'

PARAPET

ELEVATION

FIGURE 1. TYPICAL ONE-STORY UNREINFORCED MASONRY BUILDING

WITH A WOOD ROOF DIAPHRAGM (1 ft = 0.3"^ m)

VII-I18
EARTHQUAKE

INPUT MOTION

EARTHQUAKE

INPUT MOTION

A

i i i V

;

1 2 3 4 5

A

10

EARTHQUAKE

INPUT MOTION

A

11 12 Q EXTERNAL DOF Q» INTERNAL SPRING

A INTERNAL DAMPER

Q DISPLACEMENT GAGE

[^++ COMP.

+ COMP.

- TENSION

- TENSION

FIGURE 3. LUMPED PARAMETER MODEL (8 SEGMENTS - HALF MODEL)

(1 FT = 0.3^in)

VII-119
, F

FIGURE 4. TYPICAL CYCLIC LOAD DEFLECTION DIAGRAM FOR

PLYWOOD DIAPHRAGMS

F e /<F. ^ / F e

- // 1

-16 -12 -8 -4 0 4 8 12 16 DEFLECTION

(a) Force-deflection envelope of model

12

-124 . . ^ 1 . . ^ 1 -16 -12 -8 -4 0 4 8 12 16 DEFLECTION

(b) Typical cyclic load-deflection diagram for model

FIGURE 5. LOAD DEFLECTION MODEL FOR WOOD DIAPHRAGMS

VII-120
PERIOD, s

FREQUENCY, Hz

(a) Comparison of ATC spectra for Los Angeles and N-S El Centre 1940

scaled by a factor of 1.25

PERIOD, s 100 10 1 0.1 0.01

FREQUENCY, Hz

(b) Comparison of ATC spectra for Los Angeles and Castaic 1971

scaled by a factor of 1.80

FIGURE 6. RESPONSE SPECTRA FOR LOS ANGELES SHOWING SELECTED INPUT MOTIONS _ (1 in. = 2.54 cm)

VII-121
24 28 32

(a) Input displacement

12 16 20 TIME, s

12 16 20 TIME, s

(b) Input velocity

12 16 20 TIME, s

(c) Input acceleration

28 32

FIGURE 7. EARTHQUAKE INPUT MOTION, EL CENTRO SCALED BY 1.25

(1 in. = 2.54 cm)

VII-122
(a) Input displacement

10

^ 0

-T r- 8 12 16 20 24 28 32 TIME, s (b) Input velocity

(c) Input acceleration

FIGURE 8. EARTHQUAKE INPUT MOTION, CASTAIC SCALED BY 1.80

(1 in. = 2.54 cm)

VII-123
12

-12 J , . . . . . . 1 0 4 8 12 16 20 24 28 32 TIME, s (a) Linear elastic model

6

-6 J .. . , ^ , ,

1 0 4 8 12 16 20 24 28 32 TIME, s (b) Nonlinear elastic model

24

-24 J—I—, , , , , . 1 0 4 8 12 16 20 24 28 32 TIME, S

(c) Nonlinear hysteretic model

FIGURE 9. ACCELERATION RESPONSE AT MIDLENGTH OF PLYWOOD DIAPHRAGM,

EL CENTRO INPUT (1 in. = 2.54 cm)

VII-124
3 0

4 8 12 16 20 24 28 32 TIME, s

(a) Linear elastic model

(b) Nonlinear elastic model

(c) Nonlinear hysteretic model

FIGURE iO. DISPLACEMENT RESPONSE AT MIDLENGTH OF PLYWOOD DIAPHRAGM,

EL CENTRO INPUT (1 in. = 2.54 cm)

VII-125
24

i

-24] . . , , ,

,

, 0 4 8 12 16 20 24 28 32 TIME, s (a) Linear elastic model

0 4 8 12 16 20 24 28 32 TIME, S

(c) Nonlinear hysteretic model

FIGURE 11. SHEAR FORCE AT END WALL FOR PLYWOOD DIAPHRAGM,

EL CENTRO INPUT (1 kip = 4450 N)

VII-126
12 16 20 24 28 32 TIME, s (a) Linear elastic model

S o ii

? 2

15

10

5

0

-5

-10

-15

4 8 12 16 20 24 28 32 TIME, s (b) Nonlinear elastic model

i 2

12 16 20 24 28 32 TIME, s

0 4

(c) Nonlinear hysteretic model

FIGURE 12. RELATIVE DEFORMATION BETWEEN THE TOP AND BOTTOM OF THE MASONRY

WALL AT PLYWOOD DIAPHRAGM MIDLENGTH, EL CENTRO INPUT

(1 in. = 2.54 cm)

VII-127
4 8

(a) Linear elastic model

4 8 12 16 20 24 28 32 TIME, s (b) Nonlinear elastic model

i

^

a.

<= 4 ii 0

2 a

-4

4 8 12 16 20 24 28 32 TIME, s (c) Nonlinear hysteretic model

FIGURE 13. RELATIVE DEFOimATION BETWEEN THE TOP AND BOTTOM OF THE MASONRY

WALL AT PLYWOOD DIAPHRAGM MIDLENGTH, CASTAIC INPUT

(1 in. = 2.54 cm)

VII-128
ii

z

o

(a) Linear elastic model

or: — ii

5 £ UJ UJ

-16

-24-

4 8 12 16 20 24 28 32 TIME, s (b) Nonlinear elastic model

5 <

12 16 20 TIME, s

(c) Nonlinear hysteretic model

FIGURE 14. RELATIVE DEFORMATION BETWEEN THE TOP AND BOTTOM OF THE MASONRY

WALL AT DIAGONAL SHEATHED DIAPHRAGM MIDLENGTH, EL CENTRO INPUT

(1 in. = 2.54 cm)

VII-129
MITIGATION OF SEISMIC HAZARDS IN

EXISTING UNREINFORCED MASONRY BUILDINGS

Mihran S. Agbabian

Agbabian Associates

El Segundo, California

ABSTRACT

Analytical and experimental investigations to determine resistance of structures in

various seismic zones of the United States are combined in an ongoing study. The study is

being carried out by a joint venture of three consulting engineering firms in Los Angeles

under the sponsorship of the National Science Foundation.

A survey was initially made of unreinforced masonry (URM) buildings in seven geographic

areas in the United States. The selection of types of structures for investigation considered construction materials, size, distribution of walls around the perimeter and within the

building, connection details, and the application or absence of seismic design criteria.

The structures selected for study consist of:

a) Rectangular, six-story industrial building

b) Rectangular, four-story public school

c) Irregular, four-story plus basement public building

d) U-shaped, four-story apartment building

e) Rectangular, six-story and three-story office buildings

f) Rectangular, one-story and three-story industrial buildings

A methodology for earthquake hazard mitigation is being developed that will be applicable to

the broad range of buildings surveyed. Concurrent with these analyses, tests are planned for

masonry walls in the out-of-plane and in-plane directions, anchorage of walls to diaphragms,

and wood and steel diaphragm characteristics. Tests of repaired and/or retrofitted elements

will also be carried out.

The effect of the following rehabilitation techniques is being investigated:

° Strengthening of masonry walls

° Adding or improving anchorages

° Repair and strengthening of diaphragms

° Amelioration of foundation settlement

VII-130
° Addition of shear walls

° Removal of upper stories

° Parapet renovation

° Bracing of nonbearing partitions

The final result of this analytical and experimental study will be the development of a

methodology for earthquake hazard mitigation that will include (a) methods for determining

seismic hazard and seismic input, (b) methods for establishing physical properties of unreinforced masonry, (c) recommended analytical methods, and (d) recommended retrofit methods.

KEYWORDS: Seismic hazard; unreinforced masonry; retrofit methods; hazard mitigation;

diaphragms; anchorages; dynamic response.

VII-131
INTRODUCTION ... .

Unreinforced masonry (URM) buildings exist in all seismically active areas of the

United States. The potential for injury or death resulting from failure of these buildings

during earthquakes is of great concern to government officials and structural engineers, and

yet, the high cost of rehabilitating the buildings to meet current earthquake-resistance

standards has slowed down or deferred the execution of the practical steps that are required

for hazard mitigationResearch for determining realistic hazard-mitigation requirements and cost-effective

methods of retrofit is being carried out by several organizations. The purpose of such '' '

research is to develop procedures which improve the earthquake resistance of URM structures

to acceptable levels but within the economic constraints that may be tolerated by property

owners

.

This paper reports on an ongoing study that combines analytical and experimental investigations for the mitigation of seismic hazards in existing unreinforced masonry buildings.

SURVEY OF EXISTING BUILDINGS

The three geographic areas selected for survey of existing URM buildings correspond to

seismic zones identified by effective peak accelerations* (EPA) of 0.10 g, 0.20 g and 0.40 g.

The geographic areas covered are: St. Louis, the Southern Carolina area, and the New England

area in the lowest EPA category; the Pacific Northwest, the Salt Lake City area, and Memphis

in the intermediate EPA category; and California in the highest EPA category.

Construction materials, size, distribution of walls around the perimeter and within the

building, and connection details were all considered in the selection of types of structures

for investigation.

Materials

In pre-1940 buildings, floors and roofs consist of concrete or boarded wood-framed construction. Interior vertical elements are plastered wood-framed walls. In post-1940 buildings, wood-framed floors and roofs are sheathed in plywood. Floors and roofs are framed with

* EPA's are defined by the ATC-3 document entitled "Tentative Provisions for the Development

of Seismic Regulations for Buildings," Applied Technology Council, June 1978.

VH-132
steel joists or beams, and steel decking is topped with concrete. Steel deck roofs generally

are covered with insulation board and roofing. Interior partitioning often terminates in the

ceiling rather than connecting with the framing at the floor levels.

Size

The size of the structures observed ranged from very large industrial buildings more

than 600 ft (183 m) long to residences with an end wall of about 15 ft 5 m). The height

of the surveyed buildings ranged from 150 ft (46 m) high to a single story of 15 ft (5 m).

Post-1940 URM buildings are generally no more than two or three stories high.

Distribution of URM Walls '

URM walls in the perimeter of industrial buildings have relatively uniform penetrations.

Fire walls having very few penetrations are used as building dividers. Multiple housing,

public buildings, schools, and churches do not have consistent fire-wall subdivisions. Commercial buildings such as offices and retail stores have few or no URM elements at the lowest

level facing the public ways. In most buildings, these elements are not anchored into a

frame that could develop significant resistance to lateral displacement. In many cases, the

framework supports several stories of masonry above the street level.

Anchorage of Horizontal Elements and Vertical URM Walls

Beams and joists are usually anchored to the URM walls where wood framing is combined

with URM bearing walls. However, a remarkable number of surveyed structures have no apparent

anchorage at all. In pre-1940 buildings, the concrete floor was generally poured after completion of the masonry walls below the floor, and the connection depended on the interlock of

frictional surfaces. In recent construction practice, the concrete frame/floor system was

first completed and the masonry was infilled into the finished frame.

STRUCTURES SELECTED FOR STUDY

Characteristics of the seven types of structures that best represent that total United

States inventory of existing URM buildings are shown in Table 1.

VII-133
CATEGORIZING FAILURE MODES

Failure or damage of URM buildings past earthquakes will provide data for the initial

models that will be used for numerical analysis. Observed damage or failure has been a

result of:

1) Wall collapse due to (a) inadequate anchorage between walls and floors or roof diaphragms, (b) out-of-plane bending failure, (c) in-plane shear or flexural failure, and

(d) excessive deflection of the diaphragm system

2) Diaphragm failures in (a) shear, (b) shear connection to walls or other resisting

elements, (c) horizontal-shear connnection chords (if any), and (d) chords (if any)

3) Excessive deflection of the diaphragm system, causing failures of interior walls

4) Differing dynamic response of component parts of complex buildings

5) Conditions due to previous strains on structure, such as foundation settlement,

deterioration, or previous shaking

6) Effect of infilled URM wall on building frames

EVALUATION OF SEISMIC HAZARD -

''

'

Seven major U.S. cities have been selected for investigation. Figure 1 summarizes the

design criteria for these cities as provided in the ATC-3 document. Response spectra for

these seismic zones are shown in Figure 2.

Since URM buildings are constructed of undesigned elements that have hysteretic or

aperiodic response, they cannot be analyzed by spectral response methods. Accordingly, the

spectra shown in Figure 2 have been used as the bases for selecting an ensemble of timehistory motions for each area as input for dynamic analysis of the structures.

ANALYTICAL AND EXPERIMENTAL METHODS

Nonlinear dynamic response of the following URM building components will be obtained:

° End walls rocking on their foundations (overturning)

° Stability of walls subjected to out-of-plane and out-of-phase motions

° Diaphragm response (hysteretic behavior)

° Torsional response (plan irregularity)

° Complete structures (all component responses included)

VII-134
Linear and nonlinear static finite element analyses will be carried out for correlation

with tests on panels, core specimens, and anchorages, as well as analysis of perforated walls

and piers.

The test program consists of four related test series: (1) diaphragms, (2) out-of-plane

deflection of URM walls, (3) in-plane URM strength, and (4) anchorages.

Diaphragms

The first series will test five basic 20 ft by 60 ft (6.1 m by 18.3 m) diaphragms

(including repaired and retrofitted specimens) by static displacement and by dynamic,

in-plane shaking. These tests are being conducted to (a) study the behavior of diaphragms

under earthquake loading, (b) assess the effect that various kinds of retrofitted strengthening will have on the response of the diaphragms, and (c) correct, refine, and verify the

mathematical models of typical building diaphragms.

One diaphragm will be made of steel decking, two of plywood, one of board sheathing

applied diagonally, and one of board sheathing applied straight across the diaphragm frame.

The diaphragms will be retrofitted for further testing.

Each candidate diaphragm will be subjected to a series of tests in which the amplitudes

of displacement or motion will be sequentially increased to produce progressively more severe

levels of excitation. At each level, the diaphragm will be subjected to a testing sequence

of (1) quasi-static loading, (2) sine-sweep testing, (3) sine-beat-and-dwel 1 at the resonant

frequency, and finally (4) dynamic testing using a properly scaled replica of the horizontal

motion of a selected earthquake.

Quasi-static loading will develop stiffness values, which will be used to analyze the

subsequent sinusoidal and earthquake type of motions to the same deflection limits developed

in the tests. The sine-sweep sequence will reveal the first-mode resonant frequency, which

will be examined in detail by the sine-beat-test dwelling at the resonant frequency to

determine the amplitude gain factors for a particular diaphragm.

The prescribed excitation will be applied to the ends of the diaphragm as shown in

Figure 3. For the quasi-static testing, the reaction forces will be developed at pillars

positioned at the one-third points on the diaphragm. For the dynamic tests, the reaction

constraint will be removed and only inertia will resist the input motion. Along the 60-ft

(18.3 m) sides of the diaphragm, separately supported lead weights will be attached to equal

VII-13!)
the inertial loads experienced by the diaphragm owing to the wall systems and adjunct roof

sections. A typical test configuration for dynamic testing is depicted in Figure 4.

Out-of-Plane Walls

The second series will dynamically test approximately 30 masonry wall sections 6 ft by

lO-to-16 ft high (1.8 m by 3.0 to 4.9 cn) by out-of-plane and out-of-phase motions applied to

the top and bottom of the walls. Repaired walls will also be tested.

Testing unreinforced masonry wall sections in the out-of-plane direction is expected to

reveal why actual walls fractured by out-of-plane motions often do not collapse in real

earthquakes. It is possible that the fractured walls remain statically stable and thus are

still capable of supporting compressive loads without collapse. The dynamics of the phenomenon will be investigated by numercial analysis prior to testing, and the test program will

be used to correct, refine, and verify the computer analysis.

A test specimen will consist of a 6-ft wide (1.8-m) section of masonry wall constructed

on a fabricated base. The specimen will be as high as is required to satisfy a preselected

height-to-thickness ratio for the particular method used in the construction of the wall section. The section will be preloaded vertically by weights to simulate additional stories

above the test section. Two different materials (and their construction methods) and two

different height-to-thickness ratios for each of the materials will be treated. The specimens will be tested with up to four levels of vertical preloaded weights. After the characteristics of the unreinforced sections have been determined, a series of six retrofitted

specimens will be tested and evaluated.

The specimen and base will be fastened in a test fixture having a low-friction base

support that can be displaced by servohydraulic actuators in the out-of-plane direction but

without rotation. The top of the wall will be restrained either in a pinned condition or

fixed against rotation and moved independently of the base by additional actuators. A ^ typical test configuration is depicted in Figure 5.

The input to the base of the specimen will range from the minimum design ground motion

to a suitably scaled replica of a selected earthquake displacement record. The input to the

top of the wall will be determined from the assumed transfer function between the foundation

and the overhead floor or roof of the modeled building. It is expected to be nearly sinusolidal and to be substantially different from the base motion.

VII-136
In-Plane Walls

The purpose of these tests is: (1) to ascertain methods for determining the strength

properties of existing URM with an accuracy equivalent to the analysis techniques, and (2) to

determine the strengthening gained by certain types of retrofit. The tests proposed will

include brick, concrete block, and hollow clay tile masonry of thicknesses nominally 12 in.

(0.3 m) , 8 in. (0.2 m) , and 8 in. (0.2 m) , respectively. The wide variation in workmanship

and quality will be typed as good, medium, and poor quality masonry.

Anchorages

Values have been established by industry for various types of anchors in new masonry.

Since poor anchorage of walls to floor and roof diaphragms has been the cause of many failures in earthquakes, the placement of new anchors in existing URM buildings is an important

part of any retrofit program. Therefore, the frequently used types of anchors that may be

installed in existing masonry will be tested in order to determine shear and pull-out values.

Two conditions will be considered: (1) an anchor that penetrates a wall and is fastened

with a plate or washer on the outside of the wall, (2) an anchor that cannot be installed

completely through the wall, such as when the wall is on a property line.

ASSESSING AND SELECTING RETROFIT METHODS

The methodology is expected to focus on investigative and analytical processes that

enable a designer to determine whether retrofitting is needed and, if so, the extent of

retrofitting required. From that point, the details and costs of retrofitting will influence

the decision to proceed with the hazard mitigation or to remove the building.

Depending on the geometry of the existing URM structure, the accessibility of its various components, and the methods of original construction, one or more of several techniques

can be considered for rehabilitation:

Strengthening of masonry walls

Adding or improving the anchorage of walls to diaphragms

Repair and strengthening of diaphragms

Amelioration of foundation settlement

Addition of shear walls

Removal of upper stories

VII-137
Parapet Renovation • ]..

Bracing of nonbearing partitions

All of these methods will be assessed for both adequacy and relative costs. If

innovative methods of retrofitting evolve during the study, they too will be considered.

CONCLUSION

The methodology will be developed from the sum total of this research. It will be a

manual of practice applicable to all seismic zones of the United States. When the methodology is complete, the structures selected for study will be evaluated. Applicable retrofit methods will be designed for the structures as required. The cost of retrofit required

for these structures will then be estimated.

-; f

'

ACKNOWLEDGMENTS

The study is being conducted by the joint venture ABK comprising the firms of Agbabian

Associates, S. B. Barnes & Associates, and Kariotis, Kesler and Allys. It is authorized by

the National Science Foundation in support of its ongoing earthquake hazard migitation program. Dr. John B. Scalzi is Cognizant Program Manager for NSF. John Kariotis conducted the

survey of buildings, Albin Johnson is designing the test elements, and Robert Ewing is

conducting the analyses.

VII-138
TABLE 1. STRUCTURES SELECTED FOR STUDY

Building

Category

Approximate

Plan Size,

in feet

(meters)

Framing Penetrations , percent

Existing

Stories Shape Floor Roof Anchorage Buildings

Industrial 6 Rectangle 370 X 70 (113 X 21)

Wood Wood Minimal 55 All zones

Public

School

A Rectangle

plus

1-story

wings

140 X 200

(43 X 61)

Wood Wood Floors not anchored to walls70 All zones

Public

Building

4 plus

basement

Irregular

(Central

plus

2 wings)

112 X 173

(34 X 53) Concrete on steel beams

Concrete on steel beams

Floors cast

on walls

70 All zones

Apartment 4 Irregular

(U-shaped

)

100 X 180

(31 X 55)

Wood Wood Minimal 50 All zones

Conmercial

and Office

6

3

Rectangle

Rectangle

75 X 150

(23 X 46)

60 X 150

(18 X 46)

Concrete

Wood

Concrete

Wood

Floors cast

on walls

Minimal

0 at property

lines

,

100 at street

Eastern

U.S.

All zones

Post-1950

Industrial

1 Rectangle 160 X 240

(49 X 73) N.A. Steel with

light-gage

steel

decking

Minimal Minimal Eastern

U.S.

Post-1950

Commercial

3 Rectangle 75 X 150

(23 X 45)

Concrete in steel

decking

Light-gage

steel

,

steel

joists

Minimal 0 at property

lines

,

90 at street

Eastern

U.S.

VII-139
FIGURE 1. SUMMARY OF SEISMIC INPUT DATA FOR SEVEN MAJOR U.S. CITIES

(Adapted from ATC, 1978)

VII-140
.DIAPHRAGM UNDER CONSTRUCTION

PROGRAMMABLE

ACTUATORS

WIDE-FLANGE BEAMS rrrp.) REACTION PILLERS FOR

STATIC TESTS (TYP.

)

DIAPHRAGM'' UNDER TEST -PROGRAMMABLf ACTUATOR

FIGURE 3. QUASI-STATIC TEST SETUP FOR DIAPHRAGM TESTING

VII-141
SUPPORT STRUCTURE

FOR INERTIAL LOAD

DIAPHRAGM

UNDER TESTING

LEAD WEIGHTS (SHOWN FOR TWO

BAYS ONLY; ACTUALLY USED

IN ALL SIX BAYS)

PROGRAMMABLE ACTUATOR

FIGURE 4. TEST SETUP FOR DYNAMIC TESTING OF DIAPHRAGMS

ACTUATOR

FOR NONROTATING

HEAD PIECE

FOR PINNED HEAD PIECE

INERTIAL MASS

LINKSHEAD PIECE

-WALL SPECIMENBASE STRUCTURE

SLIP TABlf

FIGURE 5. SCHEMATIC TEST SETUP FOR OUT-OF-PLANE WALL TESTING

VII-142
DEVELOPMENT OF A UNIVERSAL FASTENER FOR

WOODEN BUILDING ROOF FRAMES

Tatsuo Murota

Yuji Ishiyama

Building Research Institute

Ministry of Construction

ABSTRACT

A new type of joint fastener to be used in reinforcing wood building roof frames was

developed at the Building Research Institute. This fastener, its behavior and strength

developed in tension tests are described and shown in this paper. The structural performance

of this fastener is also compared to those of conventional joint fasteners such as cramp

irons

.

KEYWORDS: Joint fasteners; roofing damage; structural performance; tension tests;

universal fasteners; wooden roofs.

VII-143
1. INTRODUCTION

In observing the types of damage to dwellings caused by the latest severe storms in

Japan, i.e., the Hachijojima Typhoon, October 1975, the Okierabu Typhoon, September 1977, et

al., it is worth noting that the most common damage in the past severe storms, i.e., the partial loss of roofing, has decreased in number but increased in the removal of roof structural

members such as rafters, purlins, roof posts, bed boards, etc.

The increase in this new type of damage to wood frame dwellings occurs despite large

decreases in the total dead weight of roof in newer dwellings compared to conventional Japanese roofing, joint connections of roof structural members often remain unchanged.

The Japanese conventional roofing material of wood frame dwellings is clay-tile and its

total roofing weight including rafters is usually 95 tcg/m^ when tiles are bedded on soil or

60 kg/m^ when tiles are set directly on bed boards. On the other hand in place of clay-tile

roofings sheet metal roofings represented by galvanized iron sheet roofing have increased

rapidly for these ten years. Dead weight of those roofings is 15 kg/m^ at most.

According to regulations in the Building Standard Law of Japan, roofs high above ground

level by h meters shall resist to suction wind force of 0.5 x 60 /h kg/m^. Supposing average Japanese single- or two-story wood frame dwellings, average roof height is about 4 or 7 m

and then suction forces to which roofs shall resist are 60 or 80 kg/m^, respectively. Comparing those values with dead weight of conventional and modern roofing described above, dead

weight of clay-tile roofings with or without bed soil is larger than or nearly equal to wind

pressure, but in case of sheet metal roofings, however, dead weight is far smaller than wind

pressure

.

This leads to the conclusion that in structural design of roofs, timber connections of

roof frames in clay-tile roofing dwellings must not necessarily be designed to resist tensile

forces but in case of light weight roofings such as galvanized iron sheet, the action of considerably large tensile forces must be supposed in the design of roof frame connections.

Therefore it is recommended in Japan that major timber connections shall be mortise-andtenon joints reinforced by cramp irons or nails. However it is often difficult to nail or

crimp effectively without cracking timbers to be connected because diameter of crimps or

VII-144
nails is usually too large compariag timber size, aad therefore the above recommendation has

not been easily put into practice and consequently timber connections in roof frames are

often ineffective to resist tensile forces.

The new type of roof damage described before is considered to be caused under these conditions. Under this present situation an idea of applying joint fasteners used in 2/4 construction is being examined in recent years. Several types of joint fasteners have already

been designed 2/4 fasteners but however it is predicted that those fasteners will not easily

spread because types of connections and dimensions of timbers are too many in number in Japanes

conventional timber construction and it is difficult for manufacturers to prepare many kinds oi

fasteners

.

Under these circumstances Building Research Institute has developed a new joint fastener

having special merits as follows; , .

i) having simple figures but applicable to almost all types and dimensions of timber

connections

ii) low cost

iii) able to be manufactured easily in situ

iv) able to be used for reinforcing timber connections in existing dwellings.

The following sections description of this fasteners and some results of experiments will

be shown. . , , .

2. DESCRIPTION OF FASTENER

Material of this fastener is a galvanized iron sheet strip of 0.8 mm thickness and 20 mm

width having nail holes along the center line (Figure 1). By cutting the material, ordinary

straight type strip fasteners are obtained and by bending it along 45° line as shown in

Figure 1 and cutting off at appropriate length, joint fasteners to be used at joints where

two members cross perpendicularly are obtained. Some examples of application are also shown

in the figure. In order to connect timbers by this fastener N38 nails or longer ones are

used. Special tools for bending and cutting materials has also been developed. .

.

3. DESCRIPTION OF TENSION TEST ''

In order to know the tensile strength of timber connections reinforced by this joint

fastener, tension tests were conducted. Two types of test specimens shown in Figure 2 were VII-145
tested. One is a purlin-roof post or roof post-roof joist connection model for testing

straight type fastener and the other is a rafter-purlin or purlin-collar connection model

for testing bent type fastener.

Major factors related to the tensile strength of this connection are (a) detail of

timber connection, (b) kind of timber, (c) roof slope, (d) thickness of fastener and (e)

length and number of nails. As for the factor (a), butt joint are adopted for test specimen

in order to exclude the other strength than that of fastener. For factor (b), western hemlock was used for test specimens because western hemlock is used most frequently now in

Japan. Concerning the other factors, a few conditions were selected respectively as shown

in Tables 1 and 2.

In case of purlin-roof post connection, blind mortise-and-tenon joint reinforced by

cramp irons (641, 120 mm length) and full mortise-and-tenon joint reinforced by nails (NlOO)

were also tested for comparison (Table 3 and 4).

4. RESULT OF TENSION TEST

4.1 Straight Type Fastener

i) Behavior and Failure Mode

Figure 3 shows tensile force (P) - displacement (6: see Figure 2) relations

obtained by the test. From this figure it is observed that;

a) plastic displacement is observed from the beginning

b) maximum strength P^^j^ occurs at nearly the same displacement of 7 to 15 mm

(Table 1)

c) strength decreases gradually at displacement larger than those stated above.

Final mode of failure was pull-out of nails and no rupture of fasteners was

observed. But at large displacement small checks or shakes of timbers were observed

around nails,

ii) Maximum Tensile Strength

Maximum tensile strength Pj^^^ obtained by test is shown in Table 1. It was

observed that effect of thickness and number of fasteners on tensile strength were

rather weak and Pj^^^ seemed to depend mainly on number and length of nails.

VII-146
Rough estimates of Pmax were as follows:

150 kg for 2-N38 nails

200 kg 3-N38

300 kg A-N38

400 kg 6-N38

250 kg 3-N50f

500 kg 6-N50p

iii) Allowable Tensile Strength

According to the timber construction design regulation prepared by the Architectural Institute of Japan, the allowable strength of timber connections shall be

decided as the value does not exceed both 1/2 to 2/3 of the maximum strength and

the strength at 1.5 to 2 mm joint displacement. Following the idea, Min( l/2*P[j,ax>

P5=2nim) also shown in Table 1. From these values rough estimate of allowable

strength of straight type fasteners are led as follows:

70 kg for 2-N38 nails

110 kg 3-N38

150 kg 4-N38

210 kg 6-N38

120 kg 3-N50f

360 kg 6-N50f

4.2 Bent Type Fastener

i) Behavior and Failure Mode

Figure 4 show typical P-6 relations of bent type fasteners. Two major differences are observed comparing with Figure 3; (a) initial rigidity is generally far

lower, (b) decrease of strength after the strength reached the maximum is more

intense

.

Failure mode of bent type fastener was also pull out of nails and no rupture

of fasteners was observed. Small checks or shakes around nails were also observed.

VII-147
ii) Maximum Tensile Strength

Maximum tensile strength Pmax shown in Table 2. From these values it is

inferred that the maximum tensile strength seemed to be effected mainly by the roof

slope; maximum strength decreased with the increase of roof slope. Effect of number and thickness of fasteners was not observed clearly,

iii) Allowable Tensile Strength

Following the same ideas in the preceding section, and neglecting the effect

of roof slope, allowable tensile strength of bent type fastener nailed by N38 can '

be obtained as follows:

50 kg for 2-N38 nails

100 kg 4-N38

'

4.3 Crimped Joint

Test results of crimped joints are shown in Table 3 and also in Figure 4 concerning

P-6 relations.

Initial rigidity of cramped joints was very high compared with straight fasteners.

The maximum strength was as much as that of straight fastener nailed by 6-N50p. It will be

noted that when cramp axes are not buried into grooves, the failure becomes very brittle.

4.4 Nail Reinforced Full Mortise-and-Tenon Joint

Table 4 shows that test result of nail reinforced full raortise-and-tenon joint. The

maximum tensile strength of this joint is as high as 600 kg for NlOO and more than 800 kg

for 2-NlOO and and behavior is very ductile. .

As for full mortise-and-tenon joints not reinforced by nails, tenons were pulled out

from mortises at one or two millimeter displacement. Tensile force necessary to pull out

tenons scattered very widely such as 170 kg to 400 kg. Accuracy of carpentry is considered

to be the reason. '

. -t- / t

5. CONCLUDING REMARKS

Supposing that rafters, purlins, roof posts, collars and other roof frame members are

arranged at ordinary intervals (rafter; @ 450 mm, purlin; @ 900 mm, post; @ 1800 mm), tensile forces to act on those member connections can be estimated from the regulation in the

VII-148
Building Standard Law of Japan; 20 to 30 kg for rafter-purlin connections, 80 to 120 kg for

purlin-roof post or purlin-collar connection and 160 to 230 kg for roof post-roof joist

connection

.

Tension test results show that the strength of new type fasteners depends mainly on

number and diameter of nails, and therefore by selecting appropriate nails the fastener can

safely resist to such an extent of tensile forces.

When this fastener is adopted for reinforcing timber roof frame connections, the

reinforcement standard shown in Table 5 will be recommended. •

VII-149
Table 1 Straight type fastener test specimens and test results

Name

of test

specimen

Number

Fasteners

Thickness

(mm)

I A » » 4> U 4- Length* Nail**

Test result

Pmax

(kg)

^pmax

(mm)

Po

\ "^y / (tq)

6H2F0.6DN38 double 0.6 6 4-N38 293 11.6 185 146

7H3F0.6DN38 7 6-N38 419 12.6 236 210

7H3F0.6DN50 7 6-N50p 553 11 .8 322 276

6H2F0.8DN38 double 0.8 6 4-N38 316 11 .9 225 158

7H3F0.8DN38 7 6-N38 437 10.0 287 218

7H3F0.8DN50 7 6-N50p 563 12.9 392 281

6H2F1 .0DN38 double 1.0 6 4-N38 332 12.9 191 166

7H3F1.0DN38 » 7 6-N38 429 6.9 295 214

7H3F1 .0DN50 7 6-N50p 473 11 .8 363 236

6H2F0.8SN38 single 0.8 6 2-N38 145 14.4 68 68

7H3F0.8SN38 7 3-N38 214 15.3 116 107

7H3F0.8SN50 * 7 3-N50p 233 8.3 158 116

* Length of fasteners is shown by the number of nail holes contained in

each fastener.

** Number of nails for each timber connected.

*** Pa - Mln(0.5xPn^x. ^Zmm)

VII-150
Table 2 Bent type fastener test specimens and test results

Name Fastener***

Test result

Roof

slope

Pmax

(kq)

^pmax

(mm

)

'^2mm (kq)

p **

a (kq)

of test

specimen Number

Thickness

(mm)

Mill* iNa 11

2/10D0.6 double 0.6 4-N38 2/10 259 12.9 112 112

3/10D0.6 3/10 294 14.4 104 104

4/10D0.6 4/10 229 9.4 101 101

2/10D0.8 double 0.8 4-N38 2/10 301 15.2 112 112

3/10D0.8 3/10 264 13.4 99 99

4/10D0.8 4/10 297 12.9 129 129

5/10D0.8 5/10 249 11 .6 120 120

10/10D0.8

[ ---

10/10 202 13.3 67 67

2/lODl.O double 1.0 4-N38 2/10 277 13.1 125 125

3/lODl.O 3/10 294 10.4 139 139

4/lODl.O 4/10 257 12.0 127 127

2/10S0.6 single 0.6 2-N38 2/10 153 17.1 38 i 38

3/10S0.6 3/10 147 17.0 45 45

4/10S0.6 4/10 104 13.8 33 33

2/10S0.8 single 0.8 2-N38 2/10 164 18.0 42 42

3/10S0.8 3/10 160 15.2 50 50

4/10S0.8 4/10 163 13.9 57 57

5/10S0.8 5/10 148 15.5 50 50

10/10S0.8

U- 1 10/10 131 16.6 31

! 31

2/lOSl .0 single 1.0 2-N38 2/10 153 15.7 44 44

3/lOSl.O 3/10 152 13.7 47 47

4/lOSl .0 4/10 138 14.3 53 53

* Number of nails for each timber connected.

** See Table 1.

*** Fastener is 10 cm long and contains 5 nail holes.

VII-151
Table 3 Test specimens and test results of butt joints reinforced by

crimp irons

Name of

test

specimen

Cramping

Number

of

cramps

Test result

pmax

(mm)

CO-1

CO-2

CO-3

ordinary 2

496

439

436

3.7

4.1

3.4

CG-1 492 5.6

buried

CG-2 in 2 491 9.7

groove

CG-3 401 4.4

Table 4 Test specimens and test results of full mortise-and-tenon joints

reinforced by nails

Name of Test result

test Nail Pmax pmax

specimen (kg) (mm)

Nl NlOO 639 29.6

N2-1 2-NlOO 812 9.9

N2-2 2-NlOO 969 17.8

NO-1 400 1.4

NO-2 none 292 2.3

NO-3 172 0.5

VII-152
Table 5 Reinforcement standard

Average Tensi le Reinforcement

Joint

connection

roof

height

(m)

force

estimated

(kg)

Fastner

ndi lb iitrL-trobdry

for each

U illlUcr LU'iritrL Lfcru

rafterpurl in

4

7

20

30

bent type,

single*

2-N38

2-N38

purl inroof post

purl incol lar

4

7

80

120

bent type,

double

4-N38

6-N38

4 160

straight

type,

double

6-N38

post-joist

none 2-NlOO

7 230

straight

type,

double

6-N50p

none 2-NlOO

* Fastner shall be placed on both sides of rafter

alternatively.

1

VII-153
10 20 20

I I I I I I I I—

I

2.1 nail holes

o

3E

(a) material of joint fastner (galvanized

iron sheet strip of 0.8mm thickness)

bend

1 ine

(b) straight type

fastener

45°

45^ //

\

(c) bent type fastener

purl in

roof post

roof joist

rafter

roof post

(d) example of application

col lar

roof post

Fig.l Material and basic types of joint fastner and examples of application

VII-154
1 1

P/2 4

1 4

P/2

rafter 45x45

purlin 90x90

S = displacement

e < 13mm

1 = 160mm

T

purlin 90x90

roof post 9ox9o

S = displacement

1 = 150mm

(a) bent type fastner test specimen (b) straight type fastner test specimen

Fig. 2 Test specimen

VII-155

THE NASA/MSFC EXPERIMENTAL FACILITIES "

AT HUNTSVILLE, ALABAMA ' — ,

:

John B. Scalzi •. ;

National Science Foundation

Washington, D.C.

and

George F. McDonough, Jr.

Nicholas C. Costes

George C. Marshall Space Flight Center

National Aeronautics and Space Administration

Huntsville, Alabama

ABSTRACT

An evaluation of National Aeronautics and Space Administration (NASA) facilities and

supporting capabilities for earthquake engineering research was made during a site visit and

workshop held at George C. Marshall Space Flight Center (MSFC) , Alabama, on February 22, 23,

and 24, 1979. Workshop participants included twenty-six earthquake engineering specialists

from the academic community, industry, and government. The workshop was sponsored by the

National Science Foundation and NASA, and it was hosted by MSFC. The chairmanship and overall direction of the workshop was assumed by the Earthquake Engineering Research Institute.

The results of the workshop indicate that the NASA/MSFC facilities and supporting capabilities offer unique opportunities for conducting earthquake engineering research. Specific features that are particularly attractive for large-scale static and dynamic testing of

natural and man-made structures include the following: large physical dimensions of buildings and test bays; high loading capacity, wide range and large number of test equipment and

instrumentation devices; multichannel data acquisition and processing systems; technical

expertise for conducting large-scale static and dynamic testing; sophisticated techniques for

systems dynamics analysis, simulation and control; and capability for managing large-size and

technologically complex programs.

VIII-1
This paper will describe the pertinent facilities at MSFC including their capabilities,

original purpose and utilization and potential uses, both in their current state and with

modification, to support seismic testing of structures.

KEYWORDS: Large scale testing; test facilities; structures soils, foundation tests.

VIII-2
1. INTRODUCTION

The George C. Marshall Space Flight Center (MSFC) of the National Aeronautics and Space

Administration (NASA) is located within the U.S. Army Redstone Arsenal (RSA), a 60 square

mile (96 square kilometers) facility adjacent to Huntsville, Alabama. The Arsenal is

strategically located roughly 100 miles (160 km) from Nashville, Tennessee and Birmingham,

Alabama and 200 miles (320 km) from Memphis, Tennessee and Atlanta, Georgia. It has its

own ship docks, railroads and an airport in addition to close proximity to public transportation, including a modern airport and the interstate highway system. Barges played a

prominent role in the early Apollo Program in carrying launch vehicles to Cape Canaveral for

launch; the Arsenal airstrip was used in the shipping of stages and payloads as well as the

delivery of the first Space Shuttle to MSFC piggyback aboard a 747 aircraft for modal testing. The location of MSFC within the RSA came about because the nucleus of MSFC was made up

of scientists and engineers of the U.S. Army Missile Command who were transferred to NASA

when it was formed in 1960. Dr. Wernher von Braun, who led the group which launched the

first U.S. earth satellite in 1958, was the first Director of MSFC. The U.S. Army continues

to conduct development and testing of military missiles and rocket systems ar RSA as well as

training of crews for their operation worldwide; whereas the activities of MSFC are limited

to peaceful uses of space and the used of space technology in the solution of problems in

the civil sector. Seismic testing for earthquake hazard mitigation falls into the latter

category of our charter.

Most of the NASA full-scale structural testing of large space and launch vehicles has

been performed at MSFC, primarily because this Center was assigned responsibility within

NASA for the largest vehicles: the Saturn moon rockets and Skylab. Facilities developed

for these large systems recently have been used to test the Space Shuttle. Types of tests

include static and dynamic structural load, vibration and acoustics, thermal vacuum and

pressure proof tests. These tests were necessary to prove vehicle integrity and contributed greatly to the success of the programs conducting them. Because of the large size

facilities required for the Saturn, which was 360 ft (110 ra) and weighted 6.5 x 10^ lbs

(2.9 X 10^ kg) when fully loaded, MSFC now has the largest and best equipped facilities of

this kind in the United States. Recent completion of the Shuttle tests allows these

VIII-3
facilities to be made available for other purposes, an important one of which can be testing

of structures for earthquake hazard mitigation. This paper will describe the major facilities which could be used for this purpose.

2. MAJOR FACILITIES

Figure 1 shows an overview of the MSFC Test Area as it is configured for Space Shuttle

testing. Building 4550 is the Saturn V dynamic test stand recently converted for Shuttle

vibration testing; a smaller test stand used for Saturn I dynamic testing is adjacent to it.

Building 4670 was originally the first stage booster engine test stand on which were test

fired simultaneously the five engines of Saturn V; other engine test stands are throughout

the area. Building 4619, which is the headquarters of the MSFC structural test activities,

is the building of primary interest for earthquake testing since it contains three of the

most important elements needed for the tests envisioned in early phases of these tests.

These elements are the Structural Test Tower, the Modal Test Equipment, and the Structural

Test Data Acquisition System.

2.1 Structural Test Tower - .

The Structural Test Tower is located in a large test bay area of Building 4619 called

the Load Test Annex, which is 170 ft (51.5 m) by 160 ft (49 m) by 155 ft (47.3 ra) high and

which is served by an access door 60 ft (18.3 m) by 75 ft (22.9 m) . The Annex is served by

a crane system utilizing two trolleys each with two hooks. Each trolley has a 30 k* ,

(13,600 kg) lifting capacity and maximum height capacity of 106 ft (32.3 m)

.

The Tower, shown in Figure 2 with a test specimen and load reaction fixtures in position,

was completed in 1963 at cost of approximately $4.5 million. Its original purpose was testing of individual stages of the Saturn launch vehicle for critical load conditions; however,

it was designed for greater capability, both in size and applied loads, than needed for

Saturn so that it would be useful for a wide range of research and development testing for

post-Apollo launch vehicles. The tower is approximately 140 ft (42 m) high and can apply

vertical and horizontal loads to structures ranging in height from 40 ft (12.2 m) to 115 ft

(35 m) and having maximum lateral dimensions of 80 ft (24.4 m) by 50 ft (15.2 m) for

* k = kilopound or kip

VIII-4
structures assembled under the crosshead. These limitations are due to the spacing of the

legs and the height of the crosshead. Larger structures can be tested within the Load Test

Annex but outside the tower without loss of horizontal loading capability; however, vertical

loading by the crosshead is lost.

The foundation of the Test Tower, shown in Figure 3 during its construction, is based

on bedrock 26 ft (7.9 m) below the floor. The main slab is about 11.5 ft (3.5 m) thick and

has 2,356 2 3/4 in (7 cm) diameter high-strength anchor bolts (ASTM A-354 steel) with

sleeves so that there can be a flush floor. The bolts are on 18 in (0.5 ra) centers in an

78 ft^ (7.2 m^) . The crosshead is driven vertically using four roll ramps which move on four

supported by 7.5 ft (2.3 m) wide reinforced concrete footings which surround the floor slab

and which extend to bedrock. These continuous footings eliminate the need for uplift anchors

for the stand.

Vertical loads are applied through the load reaction head shown in Figure 4, which

weighs 4 million lbs (1.8 million kg). This crosshead consists of 8 ft (2.4 m) deep intermediate beams framing into six 20 ft (6.1 m) deep girders which in turn frame into two 20 ft

(6.1 m) deep box girders. The top and bottom plates are 4 in ( 10 cm) thick with an area of

78 ft^ (7.2 ra^). The crosshead is driven vertically using the four roll ramps which move on

four stationary 14 in (350 mm) diameter double acme threaded stems. These ramps have a

capacity of 6 million lbs (2.7 million kg); the stems run almost the full height of the

tower. After the crosshead is put in position, it is bolted to the columns of the tower

legs to serve as strongback in resisting vertical loads. Compression capability is 30 million lbs (13.6 million kg); tension capability is 2 million lbs (907,000 kg). Horizontal

loads are applied through five horizontal box plate girders spanning two of the tower legs

which are spaced at 20 ft (6.1 m) intervals up to 100 ft (30.3 m) . Biaxial loading can be

applied by utilizing special test equipment (STE) designed to provide lateral reaction. The

massiveness of the crosshead is paralleled by that of the tower legs, each of which weights

more than 600,000 lbs (272,000 kg) and in the horizontal girders.

A wide variety of hydraulic actuators is available for use with the structural test

tower. A summary of these actuators along with other equipment is given in Section 3 of this

paper.

VIII-5
2.2 Modal Special Test Equipment

Also located in Building 4619 is the modal special test equipment (modal STE) used to

perform vibration tests, with three dimensional excitation, on the liquid oxygen tank of the

Space Shuttle External Tank. The modal STE consists of an air-bag support system which provides essentially a free-free boundary condition for the tank and an enclosure cage 78 ft.

(24 m) high and 52 ft (15.9 m) in diameter. As shown in Figure 5, the modal STE is capable

of being canted up to 13°, a requirement for simulating Space Shuttle takeoff.

The air bag suspension can be tailored to many configurations and uses. It consists of

33 Firestone Model 321 air bags with a maximum load capabity of 1.9 million lbs (864,00 kg),

a vertical suspension frequency range of 0.7 to 5 Hz, a horizontal suspension frequency range

of 0.2 to 1 Hz, a test range displacement of +3.0 in (37 mm) either horizontal or vertical.

2.3 Vertical Ground Vibration Test Facility, Building 4550

The Vertical Ground Vibration Test Facility, was initially used for Saturn V modal

tests. It is 360 ft (110 m) high, 100 ft (30 m) by 120 ft (37 m) in plan, and has a main

test bay 74 ft (22.5 m) square. The internal configuration of the building is variable to

meet the requirements of the test being conducted; hence, it was possible to test the Space

Shuttle configurations successfully in a facility designed for a much different vehicle.

Figure 6 shows the mated Orbiter-External Tank (ET) - Solid Rocket Booster (SRB) configuration in place for six-degree-of-f reedom dynamic tests; hydrodynamic supports were used. A

second configuration, Orbiter-ET, was tested using a cable support system. The roof of

Building 4550 can be removed over the main test by as can a large section of the building

sidewall 74 ft (22.5 m) by 145 ft (4.4 m) , which facilitates moving large structures into

the building using the two large derricks 440 k (200,000 kg) and 385 k (175,000 kg). The

foundation is 8 ft (2.4 m) thick and has a load capacity of 12 million lbs (5.5 million kg).

The hydrodynamic support system of Building 4550 is a state of the art device which utilizes four identical supports with a total load capacity of 8 million lbs. (3.63 million kg).

It has negligible damping due to lateral resistance and is suitable for use with a wide range

of loadings by changing its spring constants through pressure changes in the gas reservoir.

Building 4550 has full utility service as well as full instrumentation and control

capabilities with 600 channels.

VIII-6
2.4 Structural Test Facility for Hazardous Tests

Building 4572 was built and used for structural testing of the SRB. It is suitable for

hazardous testing because it is set up for remote control from an adjacent building. The

building has the capability of reacting loads as great as 3.1 million lbs (1.4 million kg)

through the end walls and through beams embedded in the concrete floor.

As shown in Figure 7, access to the building for large test articles is through the top

using a 100 k (45,000 kg) gantry crane outside the building. The building is designed for

hydrostatic tests and has high pressure water supply through 40 in (1.2 m) and 13 in (.4m)

lines

.

The building itself is 48 ft (14.8 m) by 152 ft (46.3 m) by 36 ft (11 m) high with a

clear test area of 41 ft (12.5 m) by 94 ft (28.7 m) by 32 ft (9.8 m) high. One end wall is

12 ft (3.6 m) by 40 ft (12.2 m) by 29 ft (8.8 m) high reinforced concrete with embedded

girders; the other is 7 ft (2.1 m) by 40 ft (12.2 m) by 28 ft (8.7 m) high with an embedded

steel load ring. The floor is 5 ft (1.5m) reinforced concrete with tiedown provisions to

WF 10-60 to WF 10-112 embedded beams. '

2.5 Static Test Stand

Bldg. 4670, shown in Fig. 8, was originally used as a hold down for the first stage of

the Saturn V as its engines were test fired generating 7.5 million lbs (3.4 million kg) of

vertical thrust and 370k (170,000 kg) lateral thrust. The stand consists of four concrete

piers, 200 ft (61 m) high keyed into bedrock at 46 ft (14 m) depth and tied together at the

top by a 20 ft (6 m) deep steel load platform. The interior spacing of the piers is 61 ft

(18.6 m) and they taper from 48 ft (14.6 m^) at the base to 10 ft (3 m^) at the top.

Because of its original use the facility has extremely large storage and pumping capacity for fluids and high pressure gases as well as an 880 k (400,000 kg) and 770 k (350,000

kg) derricks. It also has complete instrumentation in a nearby blockhouse.

3 . TEST SUPPORT EQUIPMENT

A large amount and great variety of large scale test equipment exists at MSFC to support

the facilities listed in the previous section and other facilities within the Center. Most

of this equipment is either portable, such as shakers, or capable of supporting tests at

remote locations, such as the Structural Test and Data Acquisition System (STDAS).

VIII-7
3.1 Vibration Test Equipment '

' ' '

'

For elastic range vibration testing MSFC has 11 hydraulic and 28 electrodynamic shakers

ranging in payload from 90 lbs (40 kg) to 200 lbs (90,000 kg) with maximum stroke of 6 in

(22.5 cm) or less, and operating at frequencies from as low as 0 Hz to as high as 2500 Hz.

These shakers are supported by three digital computers with Fourier analyzer and multipoint

control and one computer with Fourier analyzer capable of simultaneous control of eight

shakers. All systems can perform shock tests as well as sine and random testing with

automatic cutoff at present tolerances. . • -

; '.:

3.2 Hydraulic Actuators

Eighty-one portable actuators are available at MSFC; their capabilities are summarized

in Table 1.

3.3 Structural Test and Data Acquisition System (STDAS) -.J ; ^

:

The STDAS is a state of the art multiprocessor data system, developed for the Shuttle

program, but capable of handling any system requiring up to 6,000 data channels including

the simultaneous support of two separate tests. Most transducers, passive or active, can be

handled by the system including strain and displacement gages, load and pressure cells, and

thermocouples. The system is divided into three main parts: the Static Input Unit (SIU),

the Data Selector Unit (DSU) and the Central Facility (CF). The STDAS as now configured has

24 SIU's, 3 DSU's, and 1 CF.

The SIU interfaces the transducer with the data system; it scans the signals, digitizes

the analog signal, and transmits it up to 1600 ft (490 m) to a DSU. It provides excitation,

calibration and signal conditioning as well as noise filters, automatic gain-ranging and

other features and includes a Mod. Comp 11/05 digital computer which, among other capabilities stores channel addresses and performs data averaging. This computer is hardware and

software compatible with the DSU and CF computers. Portability, simplicity of operation

and ruggedness are key features of the SIU; it is totally transportable and operates

unattended in remote locations in an environment of 40°F {5" C) to 75°F (9°C) and relative

humidity of 40 to 60% while maintaining accuracy of 0.15%. Control and display capabilities

exist at the SIU to supplement the remote capability.

The DSU is a secondary computer subsystem which performs data reduction (conversion tc

engineering units), recording and display for as many as 9 SIU's (2000 channels) and

VIII-8
transfers data to the CF through modems. The DSU operator verifies the data, chooses and

implements storage modes and updates displays. Mass data storage of 1.28 million 16-bit

words is available on disc cartridge as well as magnetic tape. Graphic display includes an

operator Cathode Ray Tube (CRT) plus three annunciator CRT's (9 more can be added if

required) plus a high speed line printer. A ModComp IV/25 computer with a core memory of

64,000 16-bit words serves as the control center; it transfers data to and from the CF at

the rate of one megabit per second. In short, the operator at the DSU console can exercise

full control over the data acquisition as well as monitor test limit values vs. measured

and display real-time, quick-look data to assess test performance.

The CF, as its name implies, is the hub of the DSU-SIU network, since DSU's can only

communicate through the CF. At this station the test director and analysts monitor the conduct of the test and assess the validity of the data being acquired through the DSU's. Testing is conducted from the CF, if multiple DSU's are utilized, or from the DSU if only a

single DSU is being used. Simultaneous control from one or more DSU's and the CF is not

possible. A key feature of the CF is the ability to control DSU's as far as 3 miles (5 km)

distant. Incorporated in the CF are two ModComp IV/25 computers similar to those in the

DSU's with a full range of peripherals, some of which are switchable between processors.

The CF performs post-test analysis as well as pre-test operations and can provide reduced

test data in several formats suitable for reports or other final use. Hence, test data can

be made available in final form within no more than a few hours, and usually much less, after

completion of the test. . .

The STDAS has served as an excellent example of how time and manpower can be saved in

test data handling through automation. The real-time interaction with the tests by analysts

and other technical experts allows tests to be repeated at once if data is faulty or results

questionable as well as identifying additional test conditions that would be useful. ' 4. MINOR FACILITIES

In this category are facilities of large capability which probably would be of limited

usefulness in seismic testing and those of lesser capability which might fit well into a comprehensive earthquake program, especially in support of other tests which might be conducted

VIII-9
at MSFC. Discussions of these facilities will be brief; however, much more detailed information is available on request from the authors or in the references listed at the end of this

paper.

A small centrifuge is located in Building 4487. Under uniform rotation, it can handle a

specimen up to 500 lbs (230 kg), the e.g. of which is 6 ft (1.8 m) from the center of rotation; maximum acceleration is 100 g. In a second mode a specimen of 110 lbs (50 kg) can be

subjected to a 20 g centrifugal acceleration; to this motion, either sinusolidal oscillation

up to 28 g at frequencies of 5 - 2000 Hz, or 28 g random at frequencies of 20 - 2000 Hz can

be superimposed. Full data acquisition equipment, including television is available.

A small six degree-o f-f reedom (6 DOF) shake table, shown in Figure 9, is located in

Building 4663. The table is 17 ft (5.2 m) by 13 ft (4 m) , has a capacity of 23,000 lbs

(10,500 kg) and a horizontal stroke limit of 4 ft (1.2 m) . As shown in Figure 9, six hydraulic actuators, operated in pairs, drive the table under the control of a large hybird

(analog/digital) computer system in an adjacent room. This system has had extensive use in

simulating docking of spacecraft, vehicle dynamics on the lunar surface and several other

space applications. Its use for simulating earthquake loadings appears to be limited, however, because its fully combined 6 DOF capabilities are limited and can be achieved only

for small payloads and low frequencies.

The Neutral Buoyancy Space Simulator, Building 4706, is a very large facility consisting

of a water-filled tank 75 ft (22.9 m) in diameter and 40 ft (12.2 m) deep. By appropriate

adjustment of weights or floats attached on test subjects or equipment, a state of zero

gravity is achieved which is used for astronaut training on zero-gravity space tasks such

as extra-vehicular activity (EVA). Included in the facility are full instrumentation capabilities. Possible use in earthquake testing would be in soil-fluid-structure interaction,

such as tests related to offshore structures.

A Geotechnical Research Laboratory is located in Building 4481. It was originally

established primarily for scientific and engineering studies related to the lunar surface

exploration program including wheel soil interaction for the Lunar Roving Vehicle (LRV). Its

primary use today is in defining basic soils experiments for Spacelab missions and extending

the lunar surface studies to the planets. Its instrumentation capabilities are available to

support soil and geotechnical field and laboratory tests at MSFC.

VIII-10
5. POTENTIAL OF MSFC FACILITIES FOR EARTHQUAKE ENGINEERING

It is beyond the scope of this paper to define or propose a series of tests or other

uses of the MSFC facilities in support of earthquake engineering; however, in examining

facilities, it was useful to evaluate what types of tests might be conducted on what sizes

and types of specimens. We were guided by the comments and suggestions of the participants

in the NSF-NASA Workshop in February of this year as well as other statements of the earthquake engineering community.

Although a wide variety of tests could be conducted in the facilities at MSFC, only

those which utilized the unique capabilities were considered. The most important of these

are physical dimension and loading capability with efficient remote data acquisition, an

important second-order capability. Based on this, several strawman tests were evaluated in

detail; two of these, a structural test and a soil test, will be discussed in the following

paragraphs to indicate what is feasible and what modification to facilities might be

required

.

5.1 Multi-Story Steel Frame Cyclic Test

Biaxial loading of a 6-story building, with 12 ft (3.6 m) stories, which is 4 bays long

by 3 bays wide, with 20 ft (6.1 m) bays can be accomplished in the Structural Test Tower.

Column and beam sizes would be typical of existing MSFC rigid frame buildings. With minor

modification, the present load reaction structure and the main towers will be sufficient.

The structure could be loaded along one side and one end either at top panel points or all

panel points to achieve maximum cyclic deflections of top panel points of + 24 in (61 cm).

The existing Load Control System would be sufficient to synchronize the required 42 actuators

and existing hydraulic pumps are sufficient to drive these actuators; however, additional

long-stroke 200 k (90,000 kg) actuators would be required. Instrumentation is available:

load cells, pressure transducers, electrical deflection instruments and strain gages. One

DSU of the STDAS would be adequate to support the test. This is essentially a no-facilitymodiflcation test, the main cost of which is the test specimen.

5.2 Dynamic Soil Behavior in Large Test Bins

The Modal STE can readily be modified to support tests of dynamic soil behavior performed in cyclindrical test bins 5 ft. (1.5 m) in diameter and 10 ft. (3 m) high and

VIII-11
containing carefully prepared soil test specimens weighing approximately 24.5 kips (11,000 kg),

vjhich would serve as pilot tests for larger scale dynamic soil investigations using specimens

approximately 15 ft (4.5 m) in diameter and 30 ft (9 m) high, or 30 ft (9 m) in diameter and

15 ft (4.5 m) , as desired by the geotechnical community. The hydrodynamic support system of

Building 4550 could be used in the large-scale tests. With only slight modification of

existing hardware, the bins can be tested in torsion; with more elaborate modifications,

four degrees of freedom can be provided. All actuators, pumps, controls, instrumentation and

data handling are available. The main costs of such tests lie in the fabrication of the bins

and equipment for the placement and removal of test soil and, for the multi-degree of freedom

tests, modifications to current loading and lateral support systems.

6 . SUMMARY

The Marshall Space Flight Center has developed facilities and equipment of unique capability which have been used successfully in the space program and show substantial potential

for use in earthquake engineering research as well. Current program schedules indicate that

some of these facilities can be made available for extensive use. Furthermore, preliminary

studies indicate that only small to moderate modifications of these facilities are needed to

tailor them to earthquake testing needs. Thus a large capability, which would be costly and

time consuming to reproduce, can be made available to the earthquake engineering community at

small cost relative to the value of the facility.

An additional asset of MSFC, equally important but less tangible than the facilities, is

the experience of its staff in full-scale testing, the supporting analysis, and working with

outside investigators to reach solutions to extremely complex problems. Full-scale testing

has been one of the key factors in the success of the space program and it should prove

equally valuable to earthquake engineering.

REFERENCES I. " - .-

[1] G. F. McDonough, "Dynamic Testing of Saturn Launch Vehicles," Sixth International

Symposium on Space Technology and Science, Tokyo, 1965.

[2] R. E. Scholl, Editor, "An Evaluation of NASA Test Facilities and Supporting Capabilities for Earthquake Engineering Research," EERI (1979).

[3] G. E. Shofner, Jr., "Unique Structural Testing Facility - Load Test Annex," ASCE

National Water Resources Meeting (1970).

VIII-12
TABLE 1 VIBRATION EQUIPMENT CHARACTERISTICS

Type of

Shaker

Number

Available

Force

(tons)

Maximum

Stroke (im)

Operating Frequency

Range (Hz)

Electrodynamic* 4 13.6 25 5 - 2,000

Electrodynamic* 4 9.1 225 5 - 800

Electrodynamic* 2 6.8 25 5 - 2,000

Electrodynamic* 4 3.2 25 5 - 2,000

Electrodynamic* 14 0.5 150 0 - 2,500

Hydraul ic 8 22.7 25 0 - 350

Hydraul ic 3 45.5 25 0 - 350

Shock capacity - 2,000g; spectrum - 4 - 10 kHz.

viii-13
FIGURE 1 AERIAL VIEW OF NASA/MSFC

VIII-U
FIGURE 2 STRUCTURAL TEST TOWER WITH TEST SPECIMEN

AND LOAD REACTION TEST FIXTURE

VIli-15
VIII-16
41.6 m

COfJCRETE FLOOR SECTION A-A

3.1 m

TOWER CENTEH LINES

2.356 ANCHOR BOLTS,

0.07 m IN DIAMETER AMD

FLOOR PLAN VIEW SPACED ON 0.6 in CENTERS

24 9 m

FIGURE 4 BUILDING 4619 - LOAD TEST ANNEX MAIN TEST TOWERS

VIII-17
FIGURE 5 SPECIAL TEST EQUIPMENT WITH TEST SPECIMEN

VIIl-18
FIGURE 6 VERTICAL GROUND VIBRATION TEST FACILITY: INTERIOR VIEW OF BUILDING 4550 SHOWING SPACE TRANSPORTATION SYSTEM (SPACE SHUTTLE ORBITER,

EXTERNAL TANK, AND TWO SOLID ROCKET BOOSTERS)

VIIi-19
FIGURE 7 STRUCTURAL TEST FACILITY FOR HAZARDOUS TESTS

AERIAL VIEW OF BUILDING 4572 WITH ROOF REMOVED

VIII-20
VIII-21
VIII-22
EARTHQUAKE HAZARDS REDUCTION

RESEARCH SUPPORTED IN 1978

Charles C. Thiel

William A. Anderson

Michael P. Gaus

)i r , . William Hakala

Frederick Krimgold

Shih Chi Liu

John B . Scalzi

Division of Problem-Focused Research Applications

National Science Foundation ,. . ,

ABSTRACT ' '''

Research activities supported by the National Science Foundation in 1978 are reviewed.

Abstracts of 153 awards are presented to describe the diversity of research activities underway as part of the National Earthquake Hazards Reduction Program. , .

KEYWORDS: Earthquake design; earthquake hazards research; geotechnical engineering; seismic

design standards.

VIIl-23
PROGRAM OVERVIEW

The objectives of the Earthquake Hazards Mitigation Program are to develop methods and

techniques that can provide effective protection for man, his work and institutions from lif

loss, personal injury, property damage, social dislocations, and economic and ecological dis

ruption associated with potential or realized earthquake hazards. The Program is structured

in three principal components—siting, design and policy. These are described belowSiting

The Siting category of the Earthquake Hazards Mitigation program provides research

support for the elucidation of the physical basis of earthquake energy generation and the

transmission and propagation of the generated shock waves through various geologic and soil

conditions; with the impact of earthquake ground motion on structures; and with development

of criteria and guidelines for the mitigation of potential impacts on the built environment.

The specific objectives of the Siting program are to:

a) Develop a comprehensive data base on the nature of earthquake motion at typical

construction sites and for representative structures;

b) Establish the physical basis for characterizing the nature of earthquake motions

and the dynamic forces generated by such motions and other natural hazards;

c) Develop capabilities for prediction of the magnitude and frequencies of ground

motion;

d) Develop a methodology for qualitative and quantitative estimates of local or

regional risk associated with earthquake and other types of hazards and combined hazards.

e) Develop a comprehensive and unified program to improve geotechnical engineering

practices applicable to soil dynamics, foundation design, failure and instability, and

other aspects of earthquake ground motion; and

f) Identify procedures for integrating information on natural hazards into land use

planning, urban and coastal zone planning, offshore engineering and siting procedures.

Design

The specific objectives of the Design program are to:

(a) Improve the characterization of earthquake and natural hazard loadings necessary

for the economical design of structures subject to dynamic loading;

VIII-24
b) Develop new methods of analysis and design of buildings and structures of all types

which will take into account nonlinear and inelastic behavior of materials;

c) Develop methods to assess the hazard potential and risk assessments applicable to

existing structures and facilities, and devise innovative methods for improving performance

within economically acceptable bounds;

d) Obtain information for engineering analysis and design by observing the damage of

facilities following actual earthquakes, and incorporate this information into standard

design practice; ^jj,-. -i .

. -

e) Develop improved computational capability for dynamic analysis of structures and

facilities and improve user access to any computer software which is developed;

f) Develop model standards and design criteria for design of structures and facilities

subjected to earthquake and natural hazard loadings; and

g) Conduct detailed studies of the behavior of smaller nonengineered structures and

secondary components of buildings to improve recommended minimum analysis and design ' '

guidelines.

Policy

The specific objectives of the Policy program are to:

a) Expand the base of knowledge on alternative social adjustments to earthquakes;

b) Identify the social, economic, political, legal, and related factors which facilitate or hinder the adoption of both social and technological solutions to earthquake hazards;

c) Facilitate the beneficial use of earthquake hazards mitigation measures by devising

effective techniques for disseminating information to the public and to decision-makers at

the local. State and national levels; and

d) Investigate measures which will reduce possible negative social, economic, and

political consequences of earthquake predictions and warnings.

One of the most important objectives of the Program is the timely and widespread

dissemination of the results of supported research to potential users. The name and mailing

address of the Principal Investigator and Grantee Institution is contained in each project

description. Persons wishing to obtain information on project findings including project

reports, monographs, journal articles, technical reports, and other such relevant materials

should write to the Principal Investigator at the Grantee Institution to determine what

ViII-25
information is available and at what, if any, cost it may be obtained. The Grantee Institution may charge a nominal amount for the duplication and mailing of such materials to cover

costs. The Principal Investigator may furnish information on how interested persons may

acquire reports and other materials as appropriate from the National Technical Information

Service (NTIS) of the Department of Commerce in lieu of furnishing the report or other

material directly. NTIS is the central point in the United States for the pulic sale of

Government-funded research and development reports and other analyses prepared by Federal

agencies, their contractors, and grantees. The Principal Investigator may also cite journals

or other publications where project information may be looked up instread of furnishing a

copy of the article. '

PROGRAM ACTIVITIES

The following sections present abstracted information for each of the awards made by

the Problem-Focused Research Applications Division during calendar year 1978. During fiscal

year 1978 (October 1, 1977 through September 30, 1978) total program expenditures were

approximately $17.3 M; expenditures for fiscal year 1979 (October 1, 1978 through September

30, 1978) are estimated at $18.3 M.

Information on each award is presented in the following format:

1 2 Structural Stability of Columns; Walter G. Lewis; Lewis & Day, Inc. , 6822

Westcott Drive, Richmond, Virginia 22325;-^ Award //77-0017.'*

1) Title of the Specific Grant.

2) Principal Investigator: the chief scientist or administrator who

is responsible for the research plan and fiscal expenditures as an NSFsponsored awardee.

3) Institution Conducting the Research: any college, university, laboratory, industry, or other organization, whether operating on a profit or

non-profit basis, as well as State governments and Federal organizations.

4) Award Number.

Data is presented in the following sections for awards made in each quarter of the year.

Prior to the last quarter of the year, abstracted information for each award was very brief.

Starting with this quarter, a more detailed summary of each project is given. This information has been extracted from a quarterly announcement "Recent Awards" which is available to VIII-26
all those interested in timely notification of activities undertaken by the Program. If you

desire to receive this publication please notify one of the authors or write: Professional

Assistant, Division of Problem-Focused Research, Room 1134, National Science Foundation,

Washington, D.C. 20550. .

.

i

VIII-27
FUNCTIONAL DAMAGE AND REHABILITATION OF LIFELINES : ...

IN THE MI YAG I -KEN-OKI EARTHQUAKE OF 1978

Kazuto Nakazawa -• .

'

Eiichi Kuribayashi

Tadayuki Tazaki

Takayuki Hadate

Ryoji Hagiwara

Public Works Research Institute ' > ^ Ministry of Construction

INTRODUCTION

The Miyagi-ken-oki Earthquake of June 12th, 1978 with a magnitude of 7.4 brought many

disasters to Sendai City, population six hundred thousand, and the adjacent area. The disaster is deemed one of the biggest earthquake disasters since the Kanto Earthquake of 1923.

Since then earthquake disasters in modernized prefectural cities have been experienced in the

Fukui Earthquake of 1948 and the Niigata Earthquake of 1964. However, Sendai City was more

greatly urbanized than Fukui City and Niigata City was in those days.

In this survey, the facts connected with earthquake disaster prevention measures were

investigated cooperatively with the organizations listed later. This paper aims primarily at

a successful analysis of the functional losses and the rehabilitation of lifeline systems.

It is organized as follows:

1. Functional Damage to Lifeline Facilities

1.1 General Description

1.2 Transportation Facilities

1.3 Water Supply Facilities

1.4 Sewer Facilities ^

" ->

1.5 Medical Facilities

1.6 Energy Supply Facilities

1.7 Telecommunication Facilities

VIII-28
2. Characteristics of the Damage to Lifeline Facilities, Their Functions, and Rehabilitation

2.1 Relationship of Damage, Topography and Geology

2.2 Damage and Function of Lifeline Facilities After the Earthquake

2.3 Characteristics of the Disaster in Urban Areas

2.4 Public Facilities and Private Goods

2.5 Rescue and Rehabilitation

3. Lessons for Earthquake Disaster Prevention Measures

3.1 Data Collection on the Disaster

3.2 Resumption of Functions

3.3 Utilization of the Experience of Earthquake Disaster

Appendix

I. An Outline of the Miyagi-ken-oki Earthquake of 1978

II. Damage and Losses

III. Rescue and Rehabilitation After the Earthquake

IV. Reconstruction

V. References '

'

'

VI. An Overall Photographic Review of the Damage

'

1. FUNCTIONAL DAMAGE TO LIFELINE FACILITIES

1 . 1 General Description

This earthquake is characterized by intense damage to lifeline facilities and their func

tional interruptions in Sendai, a large city of 629,000 population, and its adjacent area.

After the earthquake, Sendai City and the area experienced suspension of electric power,

water and gas supply, railway operation, traffic confusion by the interruption of traffic sig

nals, fall of telephone lines, etc. The various damage brought great inconvenience to urban

activities

.

Lifeline facilities fulfill their function when they are organized as a network. The

experience of the earthquake shows the possibility of functional paralysis to whole lifeline

systems by partial damage to lifeline facilities, and the seriousness of the influence of

functional damage on urban activities.

VIII-29
1 . 2 Transportation Facilities ,

.

.

1) Roads

After the earthquake occurrence at 5:14 p.m. of June 12, the automatic control

system at the traffic control center of Sendai Central Police Station sustained

electrical failure, and 260 traffic signals in Sendai City could not be operated.

That was right during the rush hours in the evening, and in such situations, the

traffic in Sendai City became confused. After the recovery of electricity in the

traffic control center at 8 p.m. , the disposition of policemen for traffic control,

and emergency vehicles for electric-power supply to traffic signals, traffic congestion was gradually dissolved. .- r

Elsewhere, traffic through major roads had to be suspended or restricted by

structural damage to national and prefectural highways and the Tohoku Expressway

(see Table I-l through 1-4). , .

2) Railways

After the earthquake, all the trains under the control of the Sendai Railway

Control Bureau were suspended. Therefore many commuters and travelers were

deprived of means of transit. After inspection and temporary repairs, the Tohoku •

Main Line south of Sendai and Joban Line re-opened on June 14, and the Tohoku Main

Line north of Sendai on June 15. As for local lines in and around Sendai, their

re-opening was completed on July 8.

3) tlarbors

Ishinoraakl Industrial Harbor, Sendai Harbor, etc. suffered such damage as

cracks in aprons, breakdown of loading machines, etc. But by emergency measures,

harbor functioning was restored. , ,^ ,

4) Airports

The Sendai Airport was slightly damaged by small cracks in the runways.

Take-offs and landings were suspended for a while, but normal operation was soon

resumed. For the security of long-distance transport, Nippon Airways added

additional flights between Sendai and Tokyo [June 13 (Tuesday) to June 14

(Thursday), June 17 (Saturday) to June 18 (Sunday)].

VIII-30
1 . 3 Water Supply Facilities

Water supply facilities were damaged in 54 cities, towns and villages of the Miyagi Prefecture .

In Sendai City, water supply was suspended for about 7,000 houses after the earthquake.

The number of houses with suspended water service decreased to 2,000 en June 14, and that

decreased to 700 on June 15. After that, water supply was increased day by day. As for

other small cities, due to breaks in the water pipes, water supply was suspended at all

16,000 houses in Shiogama City and for 10,500 houses in Izumi City. The functional recovery

of water supply facilities is shown in Table 1 to 5.

Damage to industrial water supply facilities caused more than 20 factories to suspend

water supply from 24 to 79 hours.

1.4 Sewer Facilities

In Miyagi Prefecture, construction of public sewerage facilities has advanced in 16

cities and towns.

After the earthquake, sewer plants and pump yards in Sendai City and other cities around

Sendai were out of service due to electricity failures and structural damage. In this situation, sewage had to be discharged into a river without full treatment.

As a result of temporary repairs, operation of these damaged facilities was restored

within a few days except for the Kooriyama Pump Yard (recovered at 4:25 p.m. of June 23) in

Sendai City. '

1.5 Medical Facilities

Medical activities after the earthquake were more or less hampered because of electricity failures and damage to equipment and fittings at hospital and clinics in Sendai City and

its neighboring districts.

However, harm to hospital patients was avoided by emergency measures, and serious confusion to medical activities was prevented through operation among undamaged and just

slightly damaged hospitals. .. , _ .

1 .6 Energy Supply Facilities

1) Electric Power Facilities

Total electric power demand in the Tohoku District (consisting of Anomori Prefecture, Iwate Prefecture, Akita Prefecture, Miyagi Prefecture, Yamagata Prefecture and

VIII-31
Fukushima Prefecture) was 4,900 MW immediately before the earthquake. After the

earthquake, the electric power loading reduced 1,500 MW, and the electric power

supply reduced 1,000 MW in Miyagi Prefecture and other districts of Tohoku.

Reductions in electric power are shown in Table 1-6 and Figure 1-2.

As a result of functional damage, the number of houses without electricity

reached 681,600 (419,100 houses in Miyagi Prefecture, 103,000 in Fukushima Prefecture, 95,800 in Yamagata Prefecture, and 58,600 houses in Iwate Prefecture, etc.)*

As restoration proceeded, the number of houses without electricity decreased to

280,000 at the end of June 12, and that decreased to 12,000 on June 13. Electric

power facilities were completely recovered and functioning by the early morning of

June 14.

2) Gas Supply Facilities

Earthquake damage struck gas supply facilities in four cities of Sendai

(including Izumi City and Tagajo City), Shiogama, Furukawa and Ishinomaki. .» ;;;n

.

Damage in these areas was slight except for Sendai City. After inspection and

restoration, gas supply facilities were completely functioning on June 13 in . g-) n

Ishinomaki City, on June 16 in Shiogama City, and on June 18 in Furukawa City.

However, in Sendai City, gas production facilities at Minato Plant and

Haramachi Plant were more severely damaged, and worse, gas pipes were severed at

many places. Gas supply in Sendai City was wholly suspended by such serious damage. As repairs were made, gas supply was resumed at some places in Sendai City on

June 16. After inspection and repair to all gas pipes, from main pipe lines to

twig pipes inside every house, gas was supplied block-by-block. It was on July 9

that gas supply resumed in every house in Sendai City. The restoration of gas

supply facilities is shown in Table 1-7 and Figure 1-3.

Propane-gas supply by pipe was also suspended to 7,861 of the 16,266 houses

in Miyagi Prefecture after the earthquake. Propane-gas supply to 4,444 houses was

resumed on June 14, and the supply to the rest of the 3,417 houses was completed

gradually thereafter. • ; , .-gv

VIII-32
1.7 Telecommunication Facilities =. ; ;

3,845 of the 1,738,000 subscription telephones in four prefectures of Miyagi, Fukushima,

Iwate, Aomori were damaged by snapped wires, damage to telephone sets, etc. They were

entirely restored on June 20 as shown in Table 1-8. As for long-distance transmission lines,

24,000 of the 70,000 circuits into, out of, and through Sendai City were down as shown in

Figure 1-4. This was caused by damage to coaxial cables, tipping of micro antennas, etc.

All the circuits were restored on June 14 as shown in Table 1-9. Short-distance transmission

lines were damaged as shown in Table 1-10, but all the circuits were quickly restored on June

13. Further, as telephone inquiries poured into Sendai City after the earthquake, an extraordinary congestion of calls existed until June 14 as shown in Figure 1-5.

Public telephones were totally unusable for several hours due to electrical power failures, and telegraph service was knocked out at the Sendai Central Telegraph Office due to

transmission line damage. This function was soon recovered. By June 15, three days after

the earthquake, 64,000 telegrams reached Sendai City. This is 20 times more than normal.

Wireless facilities for emergencies connecting the prefecture and cities, towns and

villages were just slightly damaged in Miyagi Prefecture. There was only one town that could

not communicate by wireless.

Broadcasting, TV and radio stations could not broadcast for a brief period after the

earthquake due to electrical failures and damage to micro-circuits. Their functions were

soon restored as shown in Table 1-11 and Table 1-12. j r • •

.

2. CHARACTERISTICS OF DAMAGE TO LIFELINE FACILITIES,

THEIR FUNCTIONS, AND REHABILITATION

2 .1 Relationship of Damage, Topography and Geology

The sites where structures suffered damage correspond roughly to the areas of topographically- or geologically-severe conditions. For example. Figure 2-1 shows the distribution of

razed houses and failed slopes. The razed houses are concentrated near alluvium, the eastern

suburb of Sendai City, and to the hilltop, northern and south-western suburbs of the city.

The latter area also suffered slope failures. The distribution of the damage roughly coincides with that of the water supply facility and of the gas supply facility shown in Figure

2-4.

VIII-33
Through the experience of past earthquakes, we have learned that structures on alluvium

are likely to suffer greater damage than those on diluvial layers. The earthquake proved

that by and large this is still accurate. Some geologists assert that the back marsh behind

the ocean ridge on the Pacific Ocean is vulnerable to earthquake effects. However, little

evidence of this was observed.

The hills in the western part of Sendai City (i.e. Aoba-yama, Yogi-yama and Dainenji),

and the western part of Natori City (i.e. Minowa and Noda-yama) are covered with a 5 m thick

loam surface layer a 10 to 20 m thick gravel layer underneath, both of which are called the

Aoba-yama-Layers. The Tertiary layer of sandstone spreads beneath them. The average N-value

of the loam layer at 2.5 m beneath the surface is 5 and the gravel layer ranges from 20 to

35. In the Mldorigaoka District, Sendai City, where residential lots were developed in the

late 19 50's and early 1960's, much slope failure took place under the same geological conditions as above. The residential lots had been developed by cutting the hilltops and filling the valleys. As shown in Figures 2-2 and 2-3, there were the fills more than 10 m thick

with N-values of less than 10. Slope failure and settlement occurred mostly at these

fills due to the earthquake, which damaged residential houses, water supply, and gas supply

pipes . , r .' J' ;

As for the residential lots of the northern part of Sendai City, and Izumi City, there

are no loam and gravel layers. Hard rock and soft rock layers alternate. The residential

lots had been developed through the use of bulldozers with ripper attachments. Figure 2-4

shows the damage distribution of water supply and gas supply pipes at Nanko-dai Housing Complex, Izumi City. The damage sites are concentrated in the areas where there used to be

valleys that had been artificially filled. ;

2 . 2 Damage and Function of Lifeline Facilities After the Earthquake

Table 2-1 shows the individual loss ratios of lifeline facilities in Miyagi Prefecture.

The individual loss ratio is defined as the ratio of the loss caused by the earthquake to the

existing assets in the affected area. The existing assets can be estimated from the population density.

Individual Loss Ratio = ^

^ W Pn

VIII-34
L = loss valued in money, • ' j ,

: '

W = national wealth, . : . ' . -

P = national population and " ' • '' '

Pq = population of the quake-affected area.

The national wealth of 1978 is not available, so the 1970 data are used here.

The individual loss ratio to harbor facilities shown in the Table is the highest. The

highway and gas supply facilities figures are also high. Water supply and electric power

facilities have relatively low ratios while the telecommunication facilities are the lowest.

The list individual loss ratios provide a suggestion of the relative amounts of damage. It

can be used as an index to quantify the damage to the various facilities.

The loss ratios also roughly coincide with the amount of functional damage. In

Ishinoraaki Port, the most heavily damaged harbor, the harbor function was maintained after

the earthquake by the emergency repairs by utilizing wharves damaged only slightly. As for

the harbor facility the functional damage was not minor, but the reserve facilities were well

prepared.

Many highways were blocked by the earthquake damage. As of October 5, 1978, about 4

months after the earthquake, two highway sections and four highway bridges were still out of

service. Many bridges suffered considerable damage as shown in Appendix II, which caused

discontinuity of highway networks in Miyagi Prefecture. The gas supply facility suffered

great functional damage and the individual loss ratio was relatively high. It took about one

month to resume gas supply. The area where gas supply stopped after the earthquake encompassed Sendai and adjacent cities. The water supply facility sustained less functional

damage than the gas supply, as the individual loss ratio to the water supply was also less.

The suspension of water supply in Sendai was over on June 20, eight days after the earthquake. Islands of Shiogama City experienced the longest suspension which was eleven days.

During the suspended period, water wagons supplied water to the residents. Functional damage

to electric power was not severe. The domestic power supply was resumed on June 14. However, the suspension did influence other lifeline facilities. For example, traffic signals

did not function because they did not have electric power. This caused traffic congestion in

central Sendai City. The Traffic Control Center of Sendai Central Police Station, which

VIII-35
controls area traffic signals in Sendai, was disordered by the electric power suspension.

It could not collect data on the traffic just after the earthquake which would be useful to

analyze. Tohoku Electric Power Company, Ltd. asked industrial firms to reduce the consumption of electric power until the power plants and transformer stations recovered. This

influences industrial productivity to some extent. The telecommunication facility suffered

little functional damage. However, the demand for telephone calls increased after the earthquake, which caused difficulties and delays in placing them.

As shown above, functional damage and individual loss ratios correlate to each other

quite well .

•• ' V

Lifeline facilities can be classified into terminal facilities (e.g., power plants,

filtration plants, centrals) and distribution facilities (e.g., electric wires, conduits,

telephone lines). Highways are distribution facilities and harbors are solely terminal

facilities respectively, so that we have taken electric power, water supply, gas supply, and

telecommunication facilities as the examples for the classification.

Electric power facilities suffered damage mainly to the power plants and transformer

stations, the terminal facilities. This caused a decrease of electric power supply for

several days, although the individual loss ratio was low. This teaches us that the terminal

facility is important to preserve functions after the earthquake. ' '

On the other hand, water supply, gas supply and telecommunication facilities sustained

more damage to their distribution facilities than to this terminals. The telecommunication

facility was well prepared with reserved networks, which held functional damage to the minimum. As for the conduits, valves were in place to isolate the damaged sections from the safe

networks. Moreover some water leakage could be tolerated in order to keep the pipes flowing.

This helped to lessen functional damage.

Valves were not in place in the gas pipe network. The explosiveness of gas also delayed

the resumption of supply. ^ j' "

'

Figure 2-5 illustrates the difficulty in restoration of the function. The valves in the

pipe networks are also considered as a reserve facility. The upper and right direction indicates the difficulty of resumption. In this earthquake, electric power was considered to be

classified as IV, water supply and telecommunication was III and gas supply was I in the

figure

.

VIII-36
2.3 Characteristics of the Disaster in Urban Areas . .

This earthquake may be classified as an urban-type disaster. Dwellings and lifeline

facilities damage was concentrated in urban areas and functional damage due to the disorder

of lifeline facilities followed. The Niigata Earthquake disaster of 1964 of Japan was

similar in nature.

While all earthquakes shake the ground, not all of them cause damage to inhabited areas.

Urban-type earthquake disasters represent the opposite case. Concentration of population and

facilities accentuates the disaster. Table 2-2 shows the comparison of this earthquake and

the Kinkasan-oki Earthquake of 1936, whose epicenters were about 5 km away from each other.

The Kinkasan-oki Earthquake caused less damage than this earthquake did. As shown in the

Table the population had increased by 1.6 times between the two earthquakes. The population

of Sendai City had increased from 260,000 in 1944 to 620,000 in 1975, although the population of Miyagi Prefecture in 1944 which was 1,260,000 had changed very little from 1936. By

definition, urban-type disasters are more severe when an earthquake hits an urbanized area of

high population density.

Typical examples of urban-type earthquake disaster were observed at the newly developed

residential lots in Sendai and adjacent cities. The residential lots have been developed by

cutting hilltops, filling valleys, and by embanking on farm lands, which were usually soft

and saturated. On the other hand, the old city of Sendai, except downtown Naga-machi, which

is spread over the terrace of the Hirose River, mostly consists of diluvial gravel. The old

city on the terrace suffered minor damage. The increase of population in turn increased the

pressure to spread the building of residences to the area of steep slopes and soft ground

which are relatively vulnerable to earthquakes. The damage to lifeline facilities in the

vulnerable area greatly increased disruption in citizens' lives.

The impact on, and the response of, residents to earthquakes was likely to have been

less than what would be expected in a metropolitan area such as Tokyo or Osaka.

Table 2-3 shows the tenure of houses as of 1975 National Census. In the central part

of Tokyo occupant-owned houses do not form a high percentage of the total. However in Sendai

it is comparatively high.' Moreover, it is said that the citizens of Sendai have a greater

tendency toward domiciliation. The disaster showed that they were eager to maintain order

VHI-37
and calm. Panic, such as rushing to food stores, did not take place. Relatively few complaints about the suspension of electric power, water, and gas supply were made to the relevant organizations. -

2 .4 Public Facilities and Private Goods

Public facilities are, by definition, designed to possess a uniform minimum strength for

public safety. Structures should continue to incorporate design improvements to resist the

severe circumstances of nature.

Private facilities, primarily residences, also relate to the public so they should have

similar design resistivity. The prescription for average buildings in the Standard Building

Law contains the example of avoiding nuisances to neighbors. However, private facilities are

apt to be designed as inexpensively as possible, because they tend to emphasize economy compared to public facilities. Therefore the earthquake damage to private facilities is likely

to be more influenced by natural circumstances. For instance, the damage to dwellings was

concentrated in areas such as developed lots on soft ground and hilltops.

The residents, whose houses were damaged, suffered the economic burden of having to

reconstruct them. Prefectural and municipal governments aided them indirectly by reconstructing roads and sabo dams, and strengthening steep slopes in the damaged areas. They

also recommended loans for reconstruction. However, the responsibility of reconstruction was

on the residents in the end.

2.5 Rescue and Rehabilitation

National, prefectural and municipal governments, and public corporations set up temporary organizations for rescue and rehabilitation. They functioned well, since they had

earlier experience with many disasters, such as floods.

The Department of Public Works and the Police play an important role in communication of

data, because they shoulder the main responsibility for it in flood reconstruction.

3. LESSONS FOR EARTHQUAKE DISASTER PREVENTION MEASURES

3 . 1 Data Collection on the Disaster

It is important after the earthquake to collect disaster data as quickily and accurately as possible. It is invaluable in determining methods of reconstruction and in analyzing the disaster later.

VIII-38
The earthquake caused much functional damage to several lifeline facilities, but the

functional damage has not been quantified. Adding up the functional damage to whole facilities and then comparing them has not been possible so far. The quantification of such functional damage is important in order to relate it to the total damage to the whole system.

3.2 Resumption of Functions

Road functions were somewhat disturbed for a few days after the earthquake due to the

damage to them. However it did not greatly hamper the transportation of goods. There were

no shortage of subsistence commodities.

The disaster was not big enough to apply the ministerial notification of The Designation

of Charge-free Vehicles on Toll Roads.

Gas supply was stopped for about one month after the earthquake. One reason for this

was that the gas pipe network was continuous within the supply area. The supply could not be

resumed until entire networks of damaged pipes were repaired.

In Sendai the pipe network was divided into eight blocks temporarily, in order to be

efficiently repaired, and to resume the supply as fast as possible. They were connected

again after the repair was over.

Tokyo Gas Co., Ltd. has started a plan to divide the pipe network into several blocks,

after the experience in Sendai.

Functional damage to the electric power facilities was not extensive. However it did

cause disorder to the traffic signals and water supply pumps. Functional damage to the electric power facilities had similar disruptive influences on other lifeline facilities.

3.3 Utilization of the Earthquake Disaster Experience

Fortunately, few fires occurred during the earthquake. The earthquake of February 20,

1978, described in the preceding chapter might have resulted in the training of people to

extinguish fires in earthquakes. -

The Nippon Telegraph and Telephone Public Corporation suffered considerable damage by

the Tokachi-oki Earthquake of 1968 (M = 7.9). They changed their design criteria from the

experience. It avoided routing lines on high embankements and softground as much as possible. This would have contributed to a reduction of the damage from this earthquake.

The damaged residential lots in and around Sendai had been developed on valleys and soft

grounds, vulnerable to earthquake effects. Research to measure the failure mechanism has

VIII-39
been launched by universities and local governments. However the problem was that the

vulnerability had not been communicated to the residents before the disaster. Landslide

prone areas and dangerous steep slopes have been legally designated, and the designation of

flood areas has been introduced into the by-laws.

The Miyagi Prefectural Government began its investigation into seismic micro-zonation

after the experience of the recent earthquake.

APPENDIX . , ;

I. An Outline of the Miyagi-ken-oki Earthquake of 1978

I-l The Earthquake Feature

The general feature of the earthquake is described as follows:

1) Data and time of the occurrence of the main shock: 5:14 p.m.,

June 12 (Monday), 1978,

2) Epicenter and Focal depth: 142°10'E, 38°09'N and h = 40 km,

3) Location of the epicenter: The sea bottom about 60 km off Mt.

Kinka-san in the Miyagi Prefecture,

4) Magnitude on the Richter scale: M = 7.4,

5) Seismic Intensity: Isoseisraal Map shown in Figure I-l,

6) Focal mechanism: The focal mechanism shown in Figure 1-2 proposed by

T. Seno, K. Sudo and T. Eguchi,

- . '

.

' i> " r-

•;

' 7) Severest after-shock:

Data and time: 8:34 p.m., June 14th ( Wedenesday ) , 1978,

Epicenter and Focal depth: 142°29'E, 38°12'N and h = 40 km, i

Magnitude: M = 6.4, and "fj ^

8) After-shock area: 80 km long and 30 km wide as shown in Figure I-l. ^

1-2 Fore-sign of the Main Shock . ; > i .,

The outline of fore-shocks is described as follows: : * •

1) On February 20, 1978, an earthquake with magnitude 6.7 which brought , i/

slight damage in Miyagi and Iwate prefectures occurred at the sea bottom -, .;•;)

about 60 km north of the epicenter of the main shock,

2) After the earthquake on February 20, there occurred a few sensible

earthquakes for a month in and around the area, and , VIII-40
3) Eight minutes before the main shock, a fore-shock (M = 5.8) with third

degree of JMA's Intensity Scale was recorded at the meteorological station in

Ohfunato and Miyako, Iwate Prefecture.

1-3 Weather in the Damaged Area Before and After the Earthquake

The average temperature of June, in 1978 in Sendai City was a little high in

comparison to the average while the precipitation was about normal. (Refer to

Table I-l.

)

The average temperature and the precipitation of the most severely damaged

Sendai City ranged from 17.6° to 20.3°C and from 0 to 2.5 mm/day, respectively.

1-4 An Outline of the Disaster -

The earthquake brought damage and losses mainly to Northern Japan, especially

Miyagi, Iwate, Fukushima and Yamagata Prefectures. Miyagi Prefecture suffered most

severely. Urban functions were temporarily paralyzed by damage to lifelines such

as transporation, water and gas supply, electric power and communication facilities. A considerable number of casualties were caused by the toppling of prefabricated concrete segment fences.

1-5 Historically Damaging Earthquakes off Kinkasan ,

Since the seventeenth century, twelve earthquakes with a magnitude greater

than 6.2 have occurred off the coast of Kinkasan. (Refer to Figure 1-3.)

The earthquakes, and the return periods, are shown in Figure 1-3. According

to this record, the return periods range from 23 to 42 years with one exception.

As shown in Figure 1-3, the Kinkasan-oki Earthquake of 1936 brought slight

damage to the Miyagi and Fukushima prefectures. The epicenter and magnitude of

the earthquake was similar to that of the Miyagi-ken-oki Earthquake of June 1978.

Damage and Losses

II-l General Description

The injury and loss of life and property caused by the earthquake occurred in

Miyagi, Iwate, Fukushima and Yamagata Prefectures. Injuries were estimated at

eleven thousand persons and property damage at three hundreds billion yen, respectively. Figures II-l and II-2 show the losses of property and lives. Based on

Figures II-l and II-2, the property losses of ¥15,000,000,000, ¥300,000,000,

VIII-41
¥100,000,000 and ¥30,000,000 took place in Miyagi, Iwate, Fukushima and Yamagata

Prefectures, respectively. One hundred thousand residents were injured.

Tables ll-l and 11-2 describes the classified property losses and casualties. The loss of lives and the amount of damage are detailed in the following

sections.

II-2 Losses of Life , . ^.

The number of dead and injured are shown in Table 11-2. The number of lives

lost is largest in Miyagi Prefecture.

The dead and injured were mainly due to the collapse of buildings, houses,

fences and the drop of fragments of glass. Fence collapse was the cause of about

60% of the deaths in Miyagi Prefecture. "

'

11-3 Buildings and Houses

The damage to buildings and houses mainly occurred in the Sendai plain and

Fukushima basin. Damage also occurred on the alluvium along the Kitakami and the

Hirose River. The greater part of damage was to wooden dwelling houses.

The damage and the loss is shown in Table 11-3.

11-4 Medical and Sanitary Facilities '

The loss of medical and sanitary facilities, such as hospitals, medical

stations, and solid waste stations, is shown in Table 11-4.

11-5 Commerce and Industry '

'

The loss to commerce and industry, such as the loss of goods and factories, is

shown in Table 11-5.

11-6 Agriculture, Forestry and Fishery • "

'

Farmland, agricultural production, roads, sawmills for forestry, and the quay

walls of fishing ports were damaged.

These losses are shown in Table 11-6.

J 'ji.'.' 11-7 Educational Facilities

Education facilities losses were mainly to school buildings. It is shown in

Table II-7. In this table the monetary loss to cultural assets is included.

VIII-42
II-8 Facilities for Flood Control and Preservation of Mountainous Areas

The losses to flood control facilities and mountainous areas are mainly those

to dykes, seashore, and sabo facilities. The losses are shown in Table II-8.

Countermeasures to prevent secondary disasters such as floods and debris flow

were rapidly implemented after the earthquake, which occurred in the rainy season.

II-9 Transporation Facilities

The loss of transportation facilities is shown in Table II-9

.

11-10 Water Supply Facilities

As the water supply system in Miyagi Prefecture was damaged, drinking water

and water for industry were temporarily suspended.

The loss of the water supply facilities is shown in Tables 11-10 and 11-11.

The damage distribution to pipelines is shown in Figures II-3 and II-4.

These figures show the damage was concentrated at newly developed housing lots

from cutting hills and banks on soft ground in Sendai City. Detailed investigations

have been conducted by the organizations concerned.

11-11 Gas Supply Facilities

The damage occurred to gas pipes and gas terminals.

Figure II-5 shows damage distribution to buried pipes. The damage distribution

is similar to that of the water supply pipeline.

The loss in Miyagi Prefecture is shown in Table 11-12.

11-12 Electric Power Supply Facilities

The losses to hydroelectric and thermal power plants and transformer stations

are shown in Table 11-13.

11-13 Telecommunication Facilities

Telephone calls were suspended because buried coaxial cables in Sendai and

Kabuta City were cut completely.

The loss is shown in Table II-l

.

11-14 Sewerage Facilities

The damage to sewerage facilities was to pipes, treatment plants, and pumping

facilities.

The loss is shown in Table 11-14. The damage distribution is shown in Figure

II-6.

VIII-43
11-15 Park

The facilities at Kokeshi Island of Matsushiraa Park in Rikuchu-kaigan National

Park were damaged. Observatories and lodgings at Miyako and Rikuzentakada cities

were slightly damaged.

The loss is shown in Table II-l.

11-16 Other Facilities

Damage to other facilities occurred in a water tank reservoir for fire

fighting, office furniture, etc.

The loss is shown in Table II-l. ' ' • '

11-17 Indirect Losses - i, .

Indirect losses were caused by the suspension of various industrial activities. These have not yet been evaluated. The losses are primarily in commerce,

agriculture, forestry, fishery, buildings, and manufacturing.

III. Rescue and Rehabilitation After the Earthquake

III-l General Description

The earthquake caused the death of 28 persons, the collapse of many houses,

and other serious damage. The damage to railways and roads caused a paralysis of

transportation in the affected area. In Sendai City and the adjacent area, traffic

confusion, suspension of water, electric power and gas supply, interruption of

telephone and sewerage, all caused inconveience both in urban living and to industrial activity.

Each administration proceeded with:

* Rescue of casualties,

* Evacuation of inhabitants from dangerous areas,

* Guaranty of the necessities of life such as food, drinking water, etc.

,

* Restoration of basic facilities for living, such as facilities for water

supply, electric power, gas supply and communication, and

* Restoration of transportation.

VIII-44
Fortunately, there were only 3 fires that could not be extinguished by residents themselves among the 12 fires which occurred after the earthquake. These

were soon extinguished by fire fighters. At a petroleum industrial area, some oil

spilled from cracks in 3 oil tanks, but there were no fires.

In spite of this extensive damage, greater confusion was prevented by the

rapid countermeasure responses of the organizations concerned and the cooperative

behaviour of inhabitants.

III-2 Organizational Systems and Their Operations

In Miyagi Prefecture, rescue and rehabilitation centers for emergencies were

set up and operated by each organization mentioned below:

1) Rescue and Rehabilitation Center of Miyagi Prefectural Government (opened

at 5:30 p.m., June 12 as shown in Figure III-l).

2) Rescue and rehabilitation centers by local public bodies (56 cities, towns

and villages).

3) Rescue and rehabilitation centers of: -. - , r-

* Designated government offices in Tohoku District (Tohoku District of

Jurisdiction Police Bureau, Sendai International Trade and Industry Bureau,

Sendai Overland Transportation Bureau, Sendai District of Jurisdiction

Meteorological Bureau, Tohoku Regional Construction Bureau, etc.).

The Self -Defense Forces (Tohoku District General Administration Division, The

22nd General Course Regiment),

* Designated public organizations (Sendai Railways Superintendent Bureau

of Japanese National Railways, Miyagi Telecommunication Division of the Nippon

Telegraph and Telephone Public Corporation, Miyagi Prefectural Branch of the

Japanese Red Cross Society, Tohoku Headquarters of Japan Broadcasing Corporation, Sendai Branch of Japan Transportation Co., Ltd., Miyagi Branch of Tohoku

Electric Power Co., Ltd., etc.), and, .' .

- ., c. , .

* Headquarters of Miyagi Prefectural Police Department.

The Miyagi Prefectural Government requested aid from the Ground Self-Defense

Force which put 2,025 personnel and 400 vehicles in 5 cities and 7 towns. Since

VIII-45
June 13, the Disaster Relief Law was applied to Sendai City, Izumi City, and the

towns of Hasama Yoneyama, Kogota and Naruse. Each organization carried out rescue

and rehabilitation activities with mutual cooperation.

The Miyagi Prefectural Government substituted the Miyagi-ken-oki Earthquake

Disaster Reconstruction Section for the Rescue and Rehabilitation Center on July

10, which had been charged with rehabilitation operations.

III-3 Emergency Communications

Emergency communications are vital to carry out refuge, rescue and restoration

operations.

The damage to telecommunication facilities by the earthquake was relatively

slight, and their functions were restored in an early stage. The emergency telephone networks which were set up greatly contributed to rapid responses in many

activities.

Information was relayed to inhabitants by TV, radio, patrol-cars and other

information media. This helped promote calm and reduce anxiety.

III-4 Official Information on the Earthquake

The Meteorological Agency announced earthquake information eleven times after

the earthquake. Tsunami information was broadcast after the earthquake as shown

below:

- . "

.

° Tsunami Warnings: To the Pacific Coast of Tohoku District at 5:21 p.m.,

and the Pacific Coast of Chiba and Ibaragi Prefectures at 5:24 p.m.

° Tsunami Cautions: To the Pacific Coast of Hokkaido Prefecture at 5:28 p.m.

Order of evacuation were sent to the coastal cities, towns and villages where

danger of damage by tsunamis was suspected. Height of tsunamis was observed only

between 14 to 22 cm at the Pacific Coast of Tohoku District. The tsunamis brought

no damage. Cancellation of the warnings and the caution was announced at the j

Pacific Coast of Hokkaido at 8:15 p.m. and the Pacific Coast of Tohoku District,

Chiba Prefecture and Ibaragi Prefecture at 8:30 p.m.

VIII-46
III-5 Evacuation and Rescue

About 13,650 inhabitants had to be evacuated due to collapse of houses and tsunami

warnings. The refugees returned home after the cancellation of the warnings. But

the inhabitants in areas where houses were severly damaged stayed away for fear of

secondary disasters, as shown in Table III-l.

Even on July 20, more than one month later, 2,558 habitduts of 751 families in

7 cities and 13 towns continued living at places of refuge, relative's houses, etc.

The rescue of casualties was carried out almost without interruption in cooperation

with medical teams who continued to function after the earthquake.

III-6 Fire Fighting Activities . :

Twelve fires at nine places occurred during the earthquake, but nine fires were

extinguished by residents. The three that could not be extinguished in an early

stage occurred at the Science Faculty of Tohoku University, Tohoku Pharmaceutical

College and at a low-pressure gas holder of Sendai Gas Bureau. These fires did not

spread to neighboring buildings.

The following reasons for the slight damage from fires seem accurate:

o The earthquake occurred before the cooking hour of evening meals.

o Heaters were not used, as it was early summer. > ..

• A fore-shock of the 3rd degree of JMA's Intensity Scale in Sendai City

occurred about 8 minutes before the main shock, and precautions were taken to

prevent fires. -

' " • ^' '-

• The experience gained from the earthquake of M = 6.7 on the Richter Scale

which occurred offshore Miyagi Prefecture on February 20, 1978 was not forgotten.

III-7 Traffic and Transporation After the Earthquake

1) Roads . ^

- . -

'-

In Sendai, Ishinomaki, Shiogama, and Furukawa Cities, 744 traffic signals were damaged. No traffic signals worked due to the electrical failures.

Traffic became confused in these urban areas. Owing to the dispatch of

policemen for traffic control, and emergency vehicles for electric-power

supply to principal intersections, the confusion settled down in about

five hours.

VIII-47
Tohoku Regional Construction Bureau, Miyagi Prefectural Government,

etc., patrolled main roads and collected information on the damage. Traffic

i control is shown in Tables 1-1, 1-2 and 1-3. '

The damage to the designated national highways and the methods of restoration are shown in Table III-2. ' ' :

•

As for Tohoku Expressway of JHPC, traffic was suspended between the

Shirakawa and Tsukidate Interchanges at 5:17 p.m. June 12, and between the

Ichinoseki and Hirasawa-Maesawa Interchanges at 5:32 p.m., June 12. It was

wholly reopened at 7:00 a.m., June 15. However, speed restrictions continued

at a few sections because of repair work. Traffic on the expressway resumed

normal speed on July 11. • . ; :

2) Railways '.i

The Japanese National Railways (JNR) sustained damage to railway bridges,

tracks, platforms, etc. Railway operation was suspended in and around Miyagi

Prefecture. JNR quickly restored the Tohoku line, the most important one in

the Tohoku District. Railways on the southern side of Sendai were reopened

early on the morning of June 14, and the rest on June 15. All lines were

totally restored about one month after the earthquake. • -

3) Harbors .; ' ' f

In spite of damage to facilities at Sendai Harbor, Ishinomaki Industrial

Harbor, etc., temporary measures enabled them to keep functioning without interruption. At some fishing ports, unloading was limited by damage to wharves and

other facilities. vo . .j:'.'' ':-.y.<

4) Airports J.-

The damage to Sendai Airport was not severe, and its operation was

returned to normal soon after the earthquake.

HI-8 Rehabilitation of Lifeline Facilities . . r

1) Water Supply Facilities .' ^

Water supply was suspended due to the damage to water supply facilities.

Restoration work was started immediately after the earthquake.

The Water Supply Bureau of the Sendai Municipal Government ordered "3rd

Deposition" based on "Disposition Plan for Disaster Prevention." They called up VIII-48
all staff to establish the Earthquake Disaster Counterraeasures Headquarters at 5

p.m., June 12. Four water wagons and 40 water tanks were prepared at 7:00 p.m.

Fourteen water wagons from 13 private companies were also prepared by 9:00 p.m.

Forty water wagons were dispatched. At Midoriga-oka and Kuromatsuichinebo of

Sendai, water supply was suspended because of the possibility of ground failures

Temporary common hydrants were provided. The temporary water supply is shown in

Table III-3. The temporary water supply works required the mobilization of 149

water wagons and 702 m^ of water was supplied.

In Shiogama and Izumi Cities, where the water supply facilities were

damaged severely, the Self-Defense Forces were requested to help supply water

temporarily. Though restoration work was hastened by cooperation with the

troops dispatched from other prefectures (see Table III-4), it took more than

10 days to complete the repairs in these cities.

2) Electric Power Facilities

The degree of functional damage and the repairs to the electric power

facilites of Tohoku Electric Power Co., Ltd. are described as follows:

1) Hydroelectric Power Generation Facilities

Waterways and water tanks were damaged at the hydroelectric power

plants, but the interference with generating electricity was not serious, and power was restored by 10:48 p.m., June 14.

2) Thermal Power Generation Facilities

Generators mentioned below were turned off after the earthquake.

* Sendai Thermal Power Plant; No. 2, No. 3 generators (generating

power of 17 5,000 KW each)

• Shin-Sendai Thermal Power Plant; No. 1 generator (generating

power is 350,000 KW), No. 2 generator (generating power is 600,000 KW)

* Hachinohe Thermal Power Plant; No. 3 generator (generating power

is 250,000 KW)

As the result of inspection and repair, their functions were recovered

at:

VIII-49
* Sendai Thermal Power Plant; No. 1 generator (12:54, June 16), No. 2

generator (13:45, June 26), No. 3 generator (15:39, July 20)

' Shin-Sendai Thermal Power Plant; No. 1 generator (18:16, June 18),

No. 2 generator (13:02, June 19)

* Hachinohe Thermal Power Plant; No. 3 generator (18:54, June 12)

3) Power Transmission Facilities

The power transmission lines were inspected using 3 helicopters. Damage

was not serious enough to interrupt the tranmission of electricity.

4) Transformer Facilities

Sendai Transformer Substation suffered serious damage. But as the result

of restoration work, main transformers No. 1, No. 3 and No. 4 were restored on

June 15, June 14, and June 15, respectively. Zao Main Line, Miyagi Main Line

and Shin-Sendai Thermal Power A Line were restored on June 15, June 16, and

June 18, respectively. As for Miyagi Transformer Substation, main transformers of No. 1 and No. 3 were restored on June 12, and June 16, respectively.

5) Power Distribution Facilities

Collapse of distribution poles, snapped wires, and the tipping of transformers poles occurred mainly in the southeastern part of Sendai City. The

restoration of these power distribution facilities was completed at 6:50 a.m.,

June 14. f,^-

After the earthquake, a great number of people were engaged in restoration

work. They tried to restore the supply by sharing among organizations. Some electric power was transmitted temporarily from outside of the damaged district. With

such efforts, the function of electric power facilities was restored on the early

morning of June 14. The number of people performing repair work is shown in Table

III-5. ^

. ,

,

, . , .

, . ,

3) Gas Supply Facilities ,, i ?•

The Gas Bureau of Sendai Municipal Government supplies 135,863 houses in

Sendai, Tagajo, and Izumi Cities. After the earthquake, gas production was

decreased by the burning-down of a low-pressure gas holder at Haramachi Plant, and

by the electricity failure at the Minato Plant. When 200 reports of gas leakage

VIII-50
were made by inhabitants, the Gas Bureau suspended the gas supply at 6:15 p.m.,

June 12. After the decision, the Gas Bureau warned users to turn off gas taps .

through TV and radio. . , • ,

With the cooperation of the Disaster Countermeasures Headquarters of the Gas

Bureau, Sendai Municipal Government, and the Japan Gas Association, restoration

work was carried out giving priority of to resupply residential quarters. As the

damage to gas pipes was scattered widely (see Table III-6), the supply area

(Sendai, Tagajo, and Izumi Cities) was divided into 8 blocks as shown in Figure

HI-3. Some of them, where many gas pipes were damaged, were subdivided into two

to eight sections for efficient restoration work. The procedure for the restoration work is shown in Figure III-4. It took 25,635 man-days to restore gas supply

facilities by the middle of July, with the help of 10,583 man-days (June 13 to

July 9) from 22 other companies. The detailed inspection of underground structures

has been continued.

4) Telecommunication Facilities ' <\ c

The Nippon Telegraph and Telephone Public Corporation established the Disaster

Countermeasure Headquarters immediately after the earthquake to insure the function

of telecommunication facilities and to restore damaged facilities. As the electricity failed at 165 telecommunication stations, 9 emergency vehicles for electricpower supply went to the stations to supply electric power temporarily for 40 hours

until the electricity was recovered at all stations. Coaxial cables were damaged

at 9 sites, and temporary cables were provided. By these temporary measures, the

function of telecommunication facilities was recovered rapidly. However, telephone

calls poured in from the evening of June 12 til June 14, and an extraordinary congresion of calls arose. A 25 to 75% restriction of calls was placed by restricted

call identification equipment. Telephone communication routes for emergencies were

established to maintain vital communications. : ..

Public telephones did not work because of the electricity failures. Twentynine emergency public telephones in Sendai were temporarily supplied with electric

power. At the refuge places, 22 free telephones were installed.

VIII-51
As for telegrams, the function of telegraph facilities were restored soon

after the earthquake, and 150,000 telegrams delivered in the Tohoku District.

Urgent telegrams were given delivery priority.

Temporary restoration v;ork to telecommunication facilities required 5,000 mandays, including 400 man-days by construction firms. - -

'

III-9 Schools and Their Operations

Ninety-nine elementary schools, nine junior high schools, and nine senior high

schools were closed. After the inspection of school buildings, almost all of them

reopened within a few days. • ' . . jl

One hundred forty-three schools suspended meal service for about a week

because of damage to meal service facilities, the suspension of electric power and

gas supply .

'

1 11-10 Other Measures for the Emergency . .. >

1) Petroleum Industrial Facilities

Tohoku Petroleum Co., Ltd. has 87 oil tanks at Sendai Oil Factory. Three of

them settled and were cracked by the earthquake causing the leakage of oil. It

overflowed defensive banks to the precinct, and some oil flowed into Sendai Bay.

Oil fences were stretched to prevent the spread of the oil. The inspection of each

tank and oil spill cleanup was completed by June 18. The details of the outflow

and the collection of oil are shown in Figure III-5.

2) River Facilities

Damaged dykes suffered cracking, settlement, and sliding on the Shin-Kitakami

River, the Naruse River, and the Natori River. The Tohoku Regional Construction

Bureau made field inspections after the earthquake. Facing the rainy season,

temporary restoration works were started on June 15 and were almost completed on

June 30. Some severely damaged dykes were repaired with revetments. The details

of the damage and the restoration of dykes are shown in Table III-7. k3

VIII-52
IV. Reconstruction .'•^. i.. > .. : .t.

IV-1 General Scope , , i

' - i . - ^

.

'

., .i i .

Reconstruction entails those measures necessary to restore socio-economic order.

It involves the measures needed to repair damaged structures, to construct substitutes,

to review design criteria for earthquake resistivity, and to improve related organizational structures. ,

IV-2 Organizational System for Reconstruction .,

.

Relevant administrations set up the disaster-oriented organizations temporarily for reconstruction. . ; ... ' ; •

Miyagi Prefectural Government organized the Earthquake Disaster Reconstruction

Section on July 10, 1978. It controlled the reconstruction jobs in the Government.

IV-3 Reconstruction Works of Public Utilities -

Reconstruction Work of roads, rivers, harbors, agricultural facilities

schools and water supply facilities are scheduled under the supervision of the

Government. Table IV-1 and IV-2 show the budgets for reconstruction of water

supply and other public facilities.

IV-4 Long-term Reconstruction of Public Utilities >

1) Railroad . ; • ; . .i .-en...;

The Japanese National Railroad has established the Railroad Structures

Aseismicity Committee, headed by Prof. S. Okamoto of Saitama University, to

review of the aseismic design of railroad structures.

2) Water Supply

Water supply administrators in the quake-affected area have compiled the

data on the damage to water supply facilities.

3) Gas Supply

Japan Gas Association is compiling the data on the damage and reconstruction of gas supply facilities.

Tokyo Gas Co., Ltd. has started research on how to efficiently restore

gas supply functions following earthquakes.

4) Electric Power Facilities

The earthquake resistivity of insulators on transformers had been studied

by analyses and experiments. Insulators had been considered to have enough

VIII-53
resistivity. Nevertheless some insulators suffered damage from the earthquake. The Electricity Enterprise Federation formed a Transformer Aseismicity

Committee to research the cause of the damage and future countermeasures

5) Telecommunication Facilities

The Nippon Telegraph and Telephone Public Corporation is researching the

causes of the damage and future countermeasures. It is concentrating on the

study of damage to buried coaxial cables.

IV-5 Application of Law Systems and Loans . ; -..^

The Miyagi Prefectural Government administered the legally established

assistance and loan systems shown below. ^ uM>'

1) Loans: ' • ;

Public Welfare Loan ' i

'

Minor Enterprise Loan - ' '" •

Agriculture and Fishery Loan i.'.?C)t.

Privately-Owned House Loan • - ' ' • .

.

Municipal Finance Aid •; L'- -'.t-ly

2) Exemption of Tax and Tuition v

3) Petitions for Government Aid in Violent Disasters j.

References

Rika-Nenpyo (Science Yearbook), Tokyo Astronomical Observatory, Maruzen Co., Ltd.,

1978.

Abstracts of Autumnal Presentation Meeting: The Seisraological Society of Japan, A36

,

1978, No. 2.

Ohutsu, H. , Topography, Geology and Ground Water In and Around Sendal City, Hohbundo,

Co., Ltd.

Preliminary Damage Investigation Report on Miyagi-ken-oki Earthquake of June in 1978,

Public Works Research Institute Investigation Party, Civil Engineering Journal, Vol.

20, No. 8, 1978.

Kuribayashi, E., Tazaki, T., Hadate, T., The Investigation on the Rescue and Rehabilitation of the Izu-ohshima Kinkai Earthquake of 1978, Tenth Joint Meeting, U.S. -Japan

Panel on Wind and Seismic Effects, UJNR, 1978.

Kuribayashi, E., and Tazaki, T., An Evaluation Study on the Distribution Characteristics of Property Losses Caused by Historical Earthquakes, Tenth Joint Meeting, U.S.-

Japan Panel on Wind and Seismic Effects, UJNR, 1978.

Bureau of Statistics, Office of Prime Minister, Abstract of the National Census of

1944, 1977.

VIII-54
[8] Ishizaki, K., et al. , Flood Hazard Map, Civil Engineering Journal, Vol. 18, No. 5, 1976.

VlII-55
Photo. 1. Damaged Reinforced Concrete Building at Oroshimactii, Sendai

Photo. 2. Damaged Reinforced Concrete Building at Oroshimachi, Sendai

riiMio. 3. Damaged Wooden House at Naga-machi, Sendai

VITI-56
Photo. 6. Sabsidence of Dyke Settlement at the Yoshida River VIII-57
Photo. 9. Road Offset near the

Residential Development

at Midorigaoka, Sendai
Photo. 13. Water Tanks for Emergency Supply

VIII-59
Photo. 14. Burned Gas Holder at Harano-machi, Sciidai

Photo. 15. Broken Elbow Joint of Gas Fipe at Nanko-dai, izumi

Photo. 16. Broken Insulators at the Sendai Transformer Station VIII-60
Photo. 18. Overturned Computer Equipment at the Fukushima

Prefectural Government Office

VIII-61
Table 1-1 Traffic Control on Designated National Highways

Number

of the Route Damaged Sites (Kilometer Post)

States of the Damage

Restriction

Period of the Restriction Beginning

of the Restriction

End of the Restriction

4 Tamachi, Shiroishi City (304.5km)

Sinking of the road surface by the bursting of a water

pipe

One-side restriction 23: 00, June 12

1 : 30, June 13

,, Tomiya Town, Kurokawa

County (371.7km) Upheaval of the road surface h=20.40cm

,, 17: 20, June 12

8: 00, June 13

6 Kashima Town, Soma

County (283.4km)

Sinking of the road surface near an abutment h=10cm

18:00,

June 12

19: 00, June 12

45 Naruse Town, Monou

County (Ono Bridge)

\oG .uKm /

Movement of the girderTraffic suspension 19: 00, June 12

IS : 00, June 17 Traffic restriction

of vehicles loaded more than 1 ton continued

.

Naruse Town, Monou

County (38.2km)

L^anusuae v=4UUm One-side restriction 19: 00, June 12

13: 00, June 16 Kanan Town. Monou

County (Tenno Bridge) (58.3km)

Cracks in the pier Traffic restriction of vehicles loaded more

than 1 ton

18: 00, June 20

Oofunato, Oofunato City (165.5km) Cracks in the road surface 70m

13: 00, June 13

19: 30, June 16 Sanriku Town, Kesen

County (180.6km) Cracks in the road surface 10m

19: 05, June 12

17: 00, June 15 45 Sanriku Town, Kesen

County (183.6km) Cracks in the road

surface 50m

Traffic restriction of vehicles loaded more

than 1 ton

19: 05. June 12

17: 00, June 15

Sanriku Town, Kesen

County (184.3km) Cracks in the road

surface 32m

17:51,

June 12

17: 00, June 15

Noda Village, Kunohe

County (336.7km)

Landslide L=10m

V=100m3

18 : 20, June 12

17: 00, June 14 48 Miyagi Town, Miyagi

County (30.5km)

Falling down of a tree

17: 30, June 12

19: 50, June 12 108 Kanan Town, Monou

County (5.8km) Depression of the road L=6m

h=70cm

Traffic suspension 18: 00, June 12

16: 25, June 13

Furukawa City (35.5km) Leakage of gas 17: 30, June 12

19: 50, June 12

VIII-62
Table 1-2 Traffic Control on the Roads Administrated by Miyagi Prefectural Government

from June 12 till July 10.

VQate and Time

R es t r i ctiofr-^^

18: 00, June 1217: 00, June 13

17: 00, June 1417: 00, June 1

5

17: 00, June 161 7 : 00, June 1

7

17: 00, June 18

8: 00, June 19

10: 00,

July 10 Note

Road

Traffic suspension

16 10 8 8 8 8 8 10 5

Traffic restriction of large- sized vehicles

1 2 2 1 0 2 2 2 1 Time restriction from July

7 except for on Sunday One-side

restriction

12 9 7 7 8 7 7 7 0

Speed restriction

2 0 0 0 0 0 0 0 0

Total 31 21 17 16 16 17 17 19 6

0

Bridge

Traffic suspension

5 3 4 4 4 5 5 5

Traffic restriction of large- sized vehicles

2 1 1 1 2 2 2 2 3

Traffic suspension of vehicles

at Kinno Bridge, Eai Bridge and Komazuka Bridge One-side

restriction

3 3 3 3 3 3 3 3 3 Traffic suspension of vehicles loaded more than

1 ton

Speed restriction 2 4 4 3 4 3 3 3 0

Total 12 11 12 11 13 13 13 13 6

Total

Traffic Suspension

21 13 12 12 12 13 13 15 5

Traffic restriction of large- sized vehicles

3 3 3 2 2 4 4 4 4

One-side

restriction

15 12 10 10 U 10 10 10 3

Speed restriction

4 4 4 3 4 3 3 3 0

Total 43 32 29 27 29 30 30 32 12

VIII-63
Table 1-3 Suspension of Traffic on the Roads Administrated by Miyagi Prefectural Government

till July 5 after the Earthquake

Name of the Route Damaged Sites States of Damage Restriction Reopening Time

National Highway Route 113 Zaimokuiwa, Shiroishi Town

Falling of stones Suspension

of traffic

22 ; 00, June 12

Route 284 Kesennuma City " 19: 00, June 12

Route 286 Hagurodai, Sendai

City

Landslide

Route 286 Akaishi, Sendai City Danger of stone

f -a 1 CL 1L

1L

i1 1nO"1^

11:00, June 20

continued

IXOUlc O** U rvllUlU Dilute, i UWd Town

P'allino' 1 d 1 1 1 r\f thf* hfiri op

1

\J L Lll^ IJLlVJgC 1 2 • 00 lime 24 Reopened by a temporary

bridge

Principal Prefectural Road Okumatsushima -Matsu- shima Park Line

Tetaru, Matsu- shima Town

Depression of the road

11 : 00, June 13 One-side restriction continued

" Taiwa -Matsushima Line Dobori Bridge, Taiwa Town

Depression of the abutment

22 : 00, June 19 Traffic restriction of largesized vehicles continued

" Yuzawa .Tsukidate • Shizugawa Line Onyu, Hanayama

Village

Falling of stones 16: 00, June 13

Ishinomaki -Kashimadai • Oohira Line Komazuka Bridge, Kashimadai Town

Cracks in the floor

plate Onagawa -Shizugawa Line Kamaya, Okatsu Town

Collapse of the shoulder of the road and landslide

12: 00, June 13

Principal Prefectural Road Shiogama -Watari Line Yuriage Bridge, Natori City

Sinking of the abut- mentSuspension

of traffic

1 z : uu, June z 1 Traffic restriction of large- sized vehicles continued

" Ishinomaki -Ayukawa Line Bankoku Bridge,

Wataha, Ishinomaki

City

Cracks in the pier ., 15: 00, June 29 One-side restriction continu ed

Prefectural Road

Sendal -Yamadera Line

Takinohara, Akiu Town

Falling of stones 10: 00, June 13

" Sadayoshi, Sendai Line Sadayoshi, Miyagi Town

June 28, Time restriction continued Kensennuma-Motoyoshi

Line Kesennuma City Cracks in the road surface and falling down of a tree

16: 00, June 14

" Masusawa • Yoshioka Line Taiwa Town Falling of roof tiles 11 : 00, June 13

" Kurikoma Park Line Ueda, Kurikoma

Town~Komanoyu

Falling of stones 16: 00, June 13

Arikabe Station Line Kannari Town 20: 00, June 12 Traffic restriction of large- sized vehicles continued

" Gunsawa • Hasama Line Hanayama Village 16: 00, June 13

Arikabe -Wakayanagi Line Gobo Bridge, Wakayanagi Town

Sinking of the abutment

20: 00, June 12

Wakuya -Sanbongi Line Keshozaka Overbridge, Kogota TownSinking of the road

surface

13: 30, June 13

VIII-64
(Table 1-3 continued) Name of the Route Damaged Sites States of Damage Restriction Reopening Time

Prefectural Road

Kitakami • Kahoku Line

Shunosu, Kitakami TownCracks in the road surface Suspension

of traffic Kamaya -Oosu -Okatsu Line Tachihama, Okatsu Town

Landslide 16: 30, June 13

'

' T'rm; a flQiiO'inii T inp Yoshine Towa Town

Hailing of stone 19: 00, June 13

"

OtrlllillC: Qd m i nfi .• T'fMrriC ii.JyL'Cxn.t-' 'Jltrt Ljiiic 1 i nP Toiniik'i Rrifl typ Toyosato Town

r~"T"nplfQ VjLdv-IVo 111 in fhf* HIC ahiit- d ULI

L

ment

10: 30, June 22

Traffic restriction of large- sized vehicles continued

Oosato • Rifu Line Morisato, Rifu Town

Landslide

" Kasukawa -Hataya Line Kozunai, Oosato Town

" Nenoshiroishi -Shiogama

Line Yaotome Bridge, Yaotome, Izumi

City

Sinking of the road surface near the abutment

1 . uu, June i

o

" Kitakami -Kahoku Line Ookawa, Kahoku Town

Cracks in the read

surface

12: 00, June 17 Traffic restriction of large- sized vehicles continued

LJzawa Mountain

Pass Cracks in the road surface and landslide

16: 30, June 13

Naruse -Nango Line Sunayama, Naruse Town

Cracks in the road

surface

9: 00. June 22

Taiwa -Matsuyama Line Sokawa Bridge, Ai- kawa, Taiwa Town

Cracks in the abutment

Prefectural Road

Kashimadai • Naruse Line Takeya, Matsushima Town

Suspension

of traffic

8 : 00, July 4

Sanbongi -Taiwa Line Oohira Village —

Taiwa Town

Depression of the road

12: 30, June 26

Toll Highway Cobalt Line Whole line Landslide, collapse

of the shoulder of the road and crack

in the road surface

7: 00, July 1 ,

:

VIII-65
Table 1-4 Traffic Restriction at Tohoku Expressway

Date Closing Reopening

June 12

17:17 (Shirakawa ~ Tsukidate)

17:32 (Ichinoseki ~ Hiraizumi • Maesawa)

18 : 30 (Shirakawa ~ Kooriyama)

19:15 (Ichinoseki ~ Hiraizumi • Maesawa : up line)

19:50 (Koriyama ~ Fukushima • lizaka)

20:25 (Ichinoseki ~ Hiraizumi • Maesawa : down line)

June 13

Closing continued at the rest sections because of temporary

restoration works.

23:55 (Shiroishi ~ Izumi)

June 14

12 : 00 (Taiwa ~ Tsukidate)

16:00 (Fukushima • lizaka ~

Shiroishi)

June 15

Reopening was completed at all sections, but speed restriction

continued at some sections

7:00 (Izumi ~ Taiwa)

b ' ' 'i

July 11 Traffic returned to normal conditions at all sections

at 12: 00

Note: Maximum Speed •..-•r-i lOOkm/h 80km/h lOOkm/h under construction

Shirakawa Zao P.A. Izumi Tsukidate

80km/h lOOkm/h

Ichinoseki Maesawa — Morioka

VI I 1-66
Table 1-5 Suspension and Decrease of Water Supply in Miyagi Prefecture

Date

Number of Local

Public Bodies Resunned Water Supply

after Suspension

Number of Facilities

Resumed Water Supply

Afrpr Si 1 <^ npn i r*n

Number of Local

Public Bodies Resumed Water Supply

after Reduction

June 12 3 4

June 13 (AM) 3 7 1

June 13 20 24

June 14 10 13 1

June 15 6 8

June 16 1 1

June 17 2 4

June 18 2 4

June 19 1 1

June 20 2 2

June 21 2 2

June 22 1 1

June 23 1 1

Total 54 72 2

Table 1-6 Functional Damage to Electric Power Facilities

(Tohoku Electric Power Co., Ltd.)

Decrease in Electric-power

Generation

815,210 KW

/ Hydroelectric power 115,210 KW\

^ Thermal power 700,000 KW/

Decrease in Electric-power

Supply

1,130,700 KW ( General demand 762,650 KW\

^ Big demand 368,050 KW/

Decrease in Electric- power

Reception

377,800 KW f General demand 105,300 KW\

V Big demand 272,500 KW/

VIII-67
Table 1-7 Restoration of Gas Supply Facilities in Sendai City

Including Izumi City and Tagajo City (Gas Bureau

of Sendai Municipal Government)

Date

Percentage of Restoration Achievement

Number of Gas- Supplied Houses

June 16 Beginning of gas supply

19 19% 26,000

22 32 % 44,000

25 60% 82,000

29 92 % 125,000

July 6 98.2 % 133,000

9 Restoration was completed except for at specified

districts and deserted houses.

Note : Total number of gas- supplied houses : 135,863 houses (The percentage of pervasion : 72.8%), total length of main pipe lines : 196km, total length of trunk and bough pipes : 1,225km, Total:

1,421 km. Specified districts consists of three dangerous

districts (Midorigaoka, Kitane-Kuromatsu, Kuromatsu in Izumi

City) which have 530 houses.

* Reference

Izumi CitfTragajo City

Sendai City

\ Administrative

^ ^ . ^' Region OGas-Supplied

Region

The percentage of gas- supply

pervasion in gas-supplied

region of Sendai City : 72.8%

VllI-68
Table 1-8 Damage to Telephone Systems (The Nippon Telegraph and

Telephone Public Corporation)

Name of the

Branch

Number of Telephone

Subscribers

Number of the Telephones

Suspended

Date of the System

Recovered

Miyagi

(Sendai Station)

About 533,000

(273,000)

About 2,660

(1,160) All telephone systems were

recovered on June 20 except

for 15 subscriptions impossible to be restored because

of the collapse of houses.

Fukushima 488,000 530

Iwate 331,000 650

Aomori 386,000 5

Total 1,738,000 3,845

Note: Suspension of telephones was brought about by snapping of wires, damage

to telephone sets, etc. ' '''

'

Table 1-9 Damage to Long-Distance Transmission Lines

(The Nippon Telegrapii and Telepiione Public Corporation)

'

. 'i a. Micro System

Damaged Section, System

Number of Accommodated

Circuits

Number of Damaged

Circuits

Time of the System

Damaged

Time of the System

Recovered Cause of the Damage

Tokyo ~ Sendai 6GHz

(Jujo) (Tsutsu]igaol<a)

circuits 6,100

circuits 6,100

250

17: 15,

June 12

23: 18,

June 12

19: 30,

June 12

Inclination of the wirelessrelay phone antenna at Tsutsuj igaoka Station.

Sendai ~ Hakodate 4GHz 250

Cords connecting the facili- ties pulled out at Send^ii Wireless Station.

Sendai ~ Isliinomaki llGHz 1,500 400

19: 00,

June 12

Fuses of the source of electric power for 3 systems

gone at Oohira Wireless

Station

Ishinomaki ~ Kesennuma2GHz

320 130

19: 03, June 12

Recovered naturally Under investigation of the cause.

Tokyo ~ Sendai 4GHz

Oujo)

up 6 TV

down 6

up 6 TV

down 6

17: 24,

June 12

Board of the source of electric power loosened at Sendai Wireless Station

Total

8,170 -TV? ' down 6

6,880 TV"P ^ down 6

VIII-69
b. Coaxial Cable System

Damaged Section Number of Accommodated

Circuits

Number of Damaged

Circuits

Time of the System

Damaged

Time of the System

Recovered Damaged Sites

Tohoku II (Fukushima — Sendai)

circuits

4,900

ci rcuits

4,900

17:15,

June 12

0; 03,

June 14

Breaking down of the cable at 1 point about 31 km from Sendai, 3 points about 37 ~ 38 km from Sendai at I point about 42 km from Sendai. The cable was damaged at 5 points.

Tohoku II

(Sendai ~ Morioka)

4,600 4,600

21 : 55,

June 13

Breaking down of tlie cable at the culvert about 18 km from Sendai. The cable damaged at 2 points.

Joban (Haranomachi ~ Sendai) 6,900 6,900

18: 57,

June 13

Breaking down of the cable at the culvert about 11 km from Sendai. The cable damaged at 2 points.

Total 16,400 16,400

Note ; The Damage to cables was brought about by sinking of the ground at weak-ground areas and near

bridges of directly buried sections.

c. Long Distance Circuits

Date and Time

Total Circuits into, out of and

through Sendai

Curcuits into and out of Sendai (The number of circuits is included

in the number of the left column)

Note

Total Number

of Circuits

Number of Damaged

Circuits

Ratio of Damaged

to Total

Total Number

of Circuits

Number of Damaged

Circuits

Ratio of Damaged

to Total

17: 15,

June 12

about

circuits 70,000

about

circuits 24,000

%

34

about

circuits 31,000

about

circuits 13,000

%

42

19: 30,

June 12 16,000 23 7,700 25

23: 30,

June 12 8,000 1

1

4,300 14

12: 00.

June 13 5,500 8 3,500 11

17 : 00.

June 13 3,400 5 1,600 5

0: 03,

June 14 0 0 0 0 All circuits recovered

VIII-70
Table 1-10 Damage to Short-Distance Transmission Lines

(Tiie Nippon Telegraph and Telephone Public Corporation)

Damaged Section, System

Number of Accommodated

Circuits

Number of Damaged

Circuits

Time of the System

Damaged

Time of the System

Recovered Cause of the Damage

Ishinomaki - Oshilca PCM -24

circuits

180

circuits

161

17: 15,

June 12

0 : 01,

June 13

Relay boards loosened at Oshika

Station and Oharahama Station.

Tsutsujigaol<a - Nanakita PCM-24

637 48 • 18: 50,

June 12

Changing-over cords pulled out at Nanakita Station.

Ishinomaki - Kahoku PCM-24

177 112

• 0: 45.

June 13

An electric-power supply board

loosened at Kahoku Station.

Tsutsuj igaoka - Shiogama

PCM-24

487 24 " 20; 15,

June 12

Changing-over cords pulled out

at Shiogama Station.

Sendai - NatoriPCM-24

690 212

18: 52,

June 12

Fuses of the electric-power distribution board gone at Sendai

Station. Oogawara - Kawasaki PCM-24

194 113

12

682

17: 30,

June 12

A relay board loosened at Kawasaki

Station.

Miyako - Iwaizumi T-12SR

Total

121

2,486

20: 30,

June 12

A relay board loosened at Taro

Relay Station.

Table 1-11 Functional Damage to TV and Radio Broadcasting

due to the Damage to Micro Circuits

TV

NHK : 94 stations (General : 4 7 stations, Educational : 47 stations)

24 stations in Miyagi Prefecture

36 stations in Ivvate Prefecture

30 stations in Yamagata Prefecture

2 stations in Akita Prefecture

2 stations in Fukushima Prefecture

6 broadcasting companies : 46 stations

Radio

NHK : 3 stations (First : 1 station,

Second : 2 stations)

2 broadcasting companies : 3 stations

Short-wave international

broadcasting

NHK : 1 station

VIII-71
f

Table 1-12 Suspension of TV Broadcasting

Name of the Broadcasting Station

Time of the Beginning of Suspension

Time of the Recovery

Suspended Period Cause of Suspension

Fukushima NHK General

17; 15,

June 12

17: 47,

Jane 12 32 minutes

Damage to relay lines between

Tolcyo and Sendai (4GHz).

Moriolca NHK Educational

1 / . ZD,

June 12 10 minutes 30 seconds

Yamagata NHK Educational

17: 24,

June 12

9 minutes

Sakata NHK Educational

17: 18,

June 12

3 minutes 40 seconds

'o Sakata NHK and Aomori NHK 1 were recovered by round about

routes. o Broadcasting of Fukushima NHK

didn't suspend by receiving

electric waves directly from a

satellite station in Sendai Area.

Akita NHK Educational

17: 24,

June 12

7 minutes 55 seconds

Aomori NHK Educational 17:18,

June 12

3 minutes 40 seconds

Sendai NHK Educational

17: 24,

June 12 9 m inutes

Sapporo NHK Educational

17: 25,

June 12 10 minutes 35 seconds

i

VIII-72
Legend

N'jtior.al Railways

Service S«iion 3 DC: UnOtr CaiuTfMCiion Tuhoku Express- wav

Ndiional Highway

Principal Prelcctural Koad

1 : 500.000

10 20 30

Fig. 1-1 Transportation Network in Miyagi Prefecture

VIII-73
VIII-74
VIII-75
2

o

I cox

(6 T)

<12T)

1I6G

16 G

I COX

(6 T)

J COX

(12? J

I COX

J cox

4G cm

16 G 5

nj O

_^ 4 0 5£

,

^6 0^

aC0X(4T )

I nG_ _

-- — Hakodate

. _ 5_G ^ . 6G , [Sapporo

Aomori

£

_4_G _ 6 G Sakata

-Akita

Legend

Coaxial Transmission Line Micro Transmission Line Damaged Point Fig. 1-4 Damage to Long-Distance Transmission Lines (The Nippon Telegraphi and Teleptione Public Corporation)

_ [-.xtidordln-it Itv congested

lltK.' fur iimrt.' 'h.m two days.

ttirJurdin.irily tongestcd

lliiL- for one .l.iy. Exiraordinarily tongesicd

riiiiri- fhan iwo days.

Fig, 1-5 Extraordinary Congestion of Trunk Calls (The Nippon Telegraph and Telephone Public Corporation)

VIII-76
Table 2-1 Comparison of Individual Loss Ratios of Lifeline Facilities

Loss (L) in million yen

*1 National Wealth

in billion yen

Lk P Wk Pd

Highway

*2 10,721 8,121.5 0.076

Harbor 3,746 2,076.5 0.103

Water Supply '

• 1.733 4,502.2 0.022

Gas Supply 947 867.4 0.063

Electric Power -

.

: 2,960 8,133.7 0.021

Telecommunication 850 4,857.2 0.010

*1 Referred from the National Wealth Survey of 1970.

*2 Excluding National Expressways.

P : 111,933 thousand

Pp: 1,955 thousand •

•

Table 2-2 Comparison of the Earthquake of 1978 to Kinkasan-oki

... Earthquake of 1936

Earthquake Date

Magnitude Epicenter Damage

Population

in the Time

of the

Earthquake

Kinkasanoki

Nov. 3, 1936 7.7 X =142.2°E

V= 38.2°N

Injured 4, Razed Nondwelled Building 3, Half- razed Dwelling

2, Half- razed Nondwelled Building 2, Damage to Highways

35

*1 1.234.801

Miyagikenoki June 12. 1978 7.4 X =142°10'E

9^ 38°9'N

Refer to Appendix II *2 1,955.274

*1 as of 1935

*2 as of 1975 referred from the National Census of 1975.

VIII-77
Table 2-3 Ordinary Households by Tenure of House and One Person Quasl-households Living in Rented Rooms as of 1975

Area

Ordinary Households Living in Houses One Person Quasihouseholds Living

in Rented Room

Total Owned

House

Rented House

Publicly Owned

Rented House

Privately

Owned

Issued House

Rented Room

Japa n 31,321,700 18,151,500 2,343,300 8,547,900 2,056,300 222,600 54 7 200

Miyagi Pref. 506,100 28,400 l34,3UU 31 ,100 2,500 16,800 Tokyo Pref. "3 Qon "7on 1 AAA onn ooZ ,ZUU 1 AQti I An I ,Oyj, lUU z4o,lUU 49, DUO 102,600

Aichi Pref. 1,580,400 855,300 148,500 448,100 120,900 7.500 19,200

Ohsaica Pref. 2,467,500 999,700 320.900 989,300 143,400 14,200 21,700

Sendai City 186,100 76,400 10,300 80,500 18,200 700 12,000 Central Tokyo 2,920,900 1,064,100 229,900 1,391,100 194,000 41,700 87,500 Nagoya City 611,700 244,100 68,800 246,800 48,700 3,200 12,600 Ohsaka City 882,300 298,100 106,800 427,100 42,600 7,800 9,400

Referred from ; Bureau of Statistics, Office of the Prime Minister : Population of Japan, 1975

Population Census of Japan Abridged Report Series No. 1, 1977.

Legend

1 House

iNJIori C .

Fig. 2-1 (a) Distribution of Razed Houses in Sendai

VIII-78
VIII-79
^4 §-jFtMln= 26
Fig. 2-3 Soil Profile at Midorigaoka, Sendai

VIlI-81
VIII-82
Distribution Facility Terminal Facility

Fig. 2-5 Difficulty of Restoration of Function of Lifeline Facilities

VIII-83
Table I-l Weather Conditions before and after the Main Shock

(a) Sendai Meteorological Bureau

\Items

Precipitation

Relative

Weather Temperature (°C) Humidity (%)

Date \^ Average Max. Min. (mm/day) Average Min.

June 1 Fine 18.2 23.3 13.1 — 42 29

2 Cloudy, later fine 1 7 S 23.3 - SQ OU

3 Cloudy i

1 RO , Q7

1 A 1 — AOOU A 1 4 Cloudy, later rain 15.4 17.0 14.5 21 .5 91 79

c Fine, occasionally 0 Q /I 16 A 1 3 .1 U .U 61 44 cloudy

6 Fine, temporarily 17.2 23.9 12.9 1.0 67 40

rain

7 Fine 17.1 22.4 11.3 - 71 48

8 Fine 19.0 11.1 13.8 - 75 65

9 Cloudy 21 .8 29,3 16.8 - 69 40

10 Cloudy .0 9 =i Q 1 7 ^ 77 Do

1

1

Cloudy, later rain 19.5 21.7 16.7 0.0 86 73

12 Cloudy 20.3 25.9 17.5 0.0 82 61

13 Cloudy, occasionally 18.1 21 .0 16.0 3.5 96 82

rain

14 Cloudy, later fine 1 Q =^ 94 "X 1i A n 0 .u 1 .5 QO A70 / 15 Fine, later cloudy 1 Q C 1 V .0 94 Q 1 0 .J - 60 A4

1 0 Cloudy, temporarily 22.0 27.2 18.8 o o 83 65

rain

17 Cloudy 24.2 29.3 20.6 0.0 79 66

18 Fine 25.0 29.6 20.9 — 74 63

19 Fine, later rain 9 Q n 90 8 5.5 o i uz

20 Fine, temporarily 24.2 1^.1 20.9 0.0 81 63

rain

21 Cloudy, later fine 24.1 29.8 20.6 2.0 75 52 22 Rain 19.4 21 .8 17.7 25.5 86 77

23 Cloudy, temporarily 18.6 19.9 17.8 0.5 90 83

rain

24 Cloudy 19.5 22.3 17.6 85 78 25 Cloudy 21 .1 26.4 17.6 2.0 85 67

26 Rain, later cloudy 21.3 23.5 19.6 43.0 92 83

27 Rain 18.4 20.9 17.5 47.5 97 95

28 Cloudy, occasionally 19.8 23.5 17.4 6.0 86 75 rain

29 Cloudy, temporarily 20.6 26.2 17.0 83 64

fine 30 Fine 20.4 25.0 17.9 0.0 82 64

Total 604.8 735.8 505.0 159.5 2371

Average 20.2 24.5 16.8 79

Note: The average temperature and the precipitation of June are 18.2°C

and 160mm respectively in the year 1941-1970 in Sendai City.

VIII-84
Table I- 1 Weather Conditions before and after the Main Shock

(b) Sendai Airport Meteorological Station

\Items

Date \.

Weather

Temperature (°C) Precipitation

(mm/day)

Relative

Humidity {%)

Average Max. Min. Average Min.

June 123

4

56

7

89

10

111213

1415

16

17

1819

20

21

22

23

2425

26

27

28

29

30

Fine

Cloudy, later fine Cloudy

Rain

Fine

Fine, temporarily

cloudy

Fine

Fine

Cloudy, temporarily

fine Cloudy

Cloudy, later rain

Cloudy

Rain, later cloudy

Cloudy

Fine, later cloudy

Cloudy

Cloudy

Cloudy

Fine, later cloudy

Fine, temporarily

cloudy

Cloudy, temporarily

Rain

Cloudy, temporarily

i. dill Cloudy

Cloudy

Rain, later cloudy

Rain

Cloudy

Cloudy

Fine, later cloudy

18.0

15.5

17.5

15.3 19.9

17.3

16.1 17.9

20.6

19.2

18.2

19.1 17.6

18.2 18.8

21 .1 22.3

23.2

22.9

23.2

23.4

18.9

18.3

18.8 20.0

21.1

18.4

19.4 19.9

19.6

23.9

20.1 21.9

16.3

24.1 23.9

20.3

21.9

27.6

23.8

21.1 23.6

20.9

21 .0 23.2 25.9

27.1 27.2

26.2

27.9

30.8

21.0

20.3

21.2

23.5

24.0

20.9

22.2

24.7

23.0

10.6 8.9

13.3 13.6

13.5

12.2

9.7

12.1 15.7

16.4 16.5

16.4 15.7

15.7

16.7

17.9

19.5 19.6

19.0

20.5

19.9

17.9

17.2

17.3

17.5

19.2 17.3

17.3

16.3

17.4

0.0

18.0

0.0

1.0

0.5

0.0

2.5 0.0

2.0

2.0

25.5

0.5

39.5

43.5

5.5

46

75

7496

60

75

8185

79

88

9489

99

93

93

9289

86

92

92

84

93

96

94

. 95

97

99

919288

—

—

Total 579.7 699.5 480.8 140.5 2607

Average 19.3 23.3 16.0 87

Note : The average temperature and the precipitation of June are 18.2''C and 160mm respectively in the yeai 1941-1970 in Sendai City.

VIII-a5
Table I-l Weather Conditions before and after the Main Shock

(c) Ishinomaki Meteorological Station

Precipitaion

Relative

Weather

Temperature (°C) Humidity (%)

Date N. Average Max. Min. (mm/dav) Average Min.

June 1

1 6 '\ 91 1 1 9 9 rt O ou cloudy

2 Fine, temporarily 1 S Q £.\J ,U o y cloudy

3 Cloudy, temporarily 17.5 22.6 12.7 — 64 41

fine

4 Cloudy, later rain 15.9 17.7 14.2 31 .0 86 74

5 Fine 17.6 22.0 14.1 0.0 63 47

6 Fine, temporarily 16.2 22.8 11.8 0.5 69 40

cloudy

7 Fine 16.0 20.7 11.2 — 72 53

8 Fine 17.3 21.6 12.5 - 77 60

9 Fine, occasionally 23.1 1 5 R 75 cloudy

10 Fine, temporarily 19.2 23.9 16.1 - 76 60

cloudy

11 Cloudy, later rain 17.6 19.6 16.0 0.0 89 75

12 Cloudy, later fine 19.8 24.1 16.9 0.0 80 64 13 Rain, later cloudy I'^.S 20.5 15.6 3.0 91 85

14 Fine, occasionally 9 1 7 0.0 oO 7"^ cloudy

15 Fine, later cloudy 18.8 23.5 15.9 0.0 85 60

16 Cloudy 20.6 25.2 17.7 0.0 83 63

17 Cloudy, temporarily 99 9 97 J.1 Q7 • 7/ — 80 67

fme

1 Q Fine 23.0 29.0 19.8 79 60

1 Q 21.4 24 .8 19.1 2.5 83 72 20 Cloudy, later fine 21.5 25.1 18.9 1 .0 86 71 21 Cloudy, later fine 21.1 27.7 18.4 2.0 81 59

LI Cloudy, later rain 19.3 22.0 17.6 83 69

23 Cloudy, temporarily 1 7.6 1 9.0 16.9 0.0 Ql oL

rain

^^louuy 18.2 20.6 16.4 88 OO82

25 Fine 20.4 24.9 16.9 84 73

26 i\am, ccmpoidriiy 20.3 21.3 18.6 o "7 n Z7.U 94 88

cloudy

27 Cloudy, later rain 18.8 21.1 17.6 89 78 28 Cloudy 19.2 22,1 17.1 9.5 87 72

29 Cloudy, temporarily 19.0 23.7 17.2 85 69

fine 30 Fine, temporarily 19.9 24.4 17.6 79 59

cloudy

Total 564.9 684 .3 480.7 128.5 2388

Average 18.8 22.8 16.0 80

Note : The average temperature and the precipitation of June are 18.2°C

and 160mm respectively in the year 1941-1970 in Sendai Cidy.

VIII-86
0 20 40 00 SO km

Rumoi 1 Q Asahikawa 1

ArcJ of Aftershocks

(after Pnif. K. Tsumu

Epicenter of the Maximum

^^X^i (IK' 12 N, 142 21(E)

u-r-vbpieenter of the Main Shoek

9^///y////i( 3k ()i)-N,l42'10'E)

Shirakawa4

2 / /Onahama

Kurobe oNagano 2/ Karuizawa 3/

o 10 i| ' • Matsumotolo Maebashi Mitol

°KijmJgaya4^

- o Kashiwa Nashiwa \

Kawaguchiko3o

„ MishimJ3

VJ.Nagoya2 ^/po/

TSU2 ^H,mamatsul ^tr^)

Mo-ChibaH TokyoTTTf

^^ate(4 J 3 yamaEKatsuura

r"3l

'Choslii

IV

136° 138°

-x.

Fig.I-1 Isoseismals

VIII-87
Fault parameters

Length 30km

Down dip width 80km

Fault plane dip direction N280°E

dip angle 20°

Seismic moment 2.9 -lO" dyn-cm

Dislocation 1.7m

Stress drop 64bars

N

Legend

• : compression

o : dilatation

-0-: error bar

~o-- direction of S-wave

Fig. 1-2 An Example of Focal Mechanism

Solution (by T. Seno, K. Sudo and

T. Eguchi)

® vn 28. 16 16

0 = 3 8. r

^ = 1 4 2. 0° M = 7.0

30 years

@ IV 26. 1 6 4 6 0 = 3 7. 7°

^ = 1 4 1. 7° M = 7. 6

90 years

® in 2(), 17 3 6 0 = 3 8. 3°

^ = 1 4 0. 8° M = 6.2

34 years

® V 3,1770 0 = 3 8. 6°

^ = 1 4 2. 0° M = 7.4

23 years

® n 17, 17 9 3 0 = 3 8. 3°

^ = 1 4 2. 4° M = 7.

1

42 years

® VI 25, 1 8 3 5 0 = 3 7. 9°

/! = 1 4 1. 9° M = 7.6

} 26 years

® IX 18. 18 6 1 0= 3 7.7°

/! = 1 4 1. 6° M = 6. 4

36 years

® n 20. 1 8 9 7 0= 38 1°

^ = 1 4 1. 5° M= 7.8

(7.3) ® m 5. 1 8 9 7 0 = 3 8. 0°

/i = 1 4 3. 7° M = 7.7 (7.2)

39 years

@) XI 3, 1 9 3 6 0 = 3 8 2° = 1 4 2. 2° M = 7.7

42 years

n 20. 1 9 7 8 0 = 38.7 5°

-? =142.2°

M = 6. 7

VI 12. 19 7 8 0 = 3815°

^ =14217°

M = 7. 4

Fig. 1-3 Historical Damaging Earthquakes

off Kinkasan

VIII-88
Table II- 1 Totaled Losses of Miyagi, Iwate, Fukushima and Yamagata Prefectures.

Classification Loss in Ihousand Yen

Note

[Direct Losses] one T> C OOZ. 2.7b,7lb,6Zo

Buildings and Houses 79,/ 39, 128

Medical and Sanitary Facilities o,ol o,U4U

Commerce and Industry 96, 853, 864

Agriculture, Forestry and Fishery 1 9,31)7,776

Educational Facilities 9,4o6,64(J

Facilities for Flood Control and

Preservation of Mountainous Area 14,237.936

Transportation Facilities 22,572,113

Water Supply Facilities 1,744,948

Gas Supply Facilities 947,000

Electric Power Supply Facilities Q 1 1 A 0 AO Of I i 0,oOZ

Telecommunication Facilities 871,582

Sewerage Facilities 838,453

Park 97,667

Other Facilities 22,823,317

[Indirect Losses] Obscure

Commerce Due to impossibility of business

Agriculture, Forestry and Fishery Due to impossibility of cultivation

Construction and Manufacture EXie to discontinuity of operation

Others

Total 275,725,326

Note: By Miyagi, Iwate, Yamagata and Fukushima Prefectural Government Offices,

VIII-89
Table II-2 Contents of Losses of Lives in Four Prefectures.

Classification Miyagi Pre. Iwate Pre. Fukushima Pre. Yamagata Pre.

Dead 27 0 1 0

Injured

Serious 262 0 3 0

Slight 10,700 11 27 1

Missing 0 0 0 0

Total 10,989 11 31 1

Table II- 3 Damage to Buildings and Houses

Classification

Number of Damaged

Losses in Thousand Yen

Note

Totally Razed 1,377

Half Razed 6,178

Dwelling

Breakage

Flooding over

Floor

Flooding under the Floor

126,051

3

2

> 60,180,765

Non-dwelling 43,532 19,058,363

Total 177,143 79,239,128

Table II- 4 Damage to Facilities for Medical and Sanitary.

Classification Number of Damaged Losses in Thousand

Yen

Note

Medical 663,495

Sanitary 1,300.977

Others 1,653,568

Total 3,618,040

VI.II-90
Table II-5 Damage to Commerce and Industry,

Prefecture Number of Damaged Losses in Thousand Yen

Note

• Miyagi Pre. 53,524 95,753,230

Iwate Pre. 2,252 562,782

Fukushima Pre. 518,106

Yamagata Pre. 19,746

Total 96,853.864

Table II6Damage to Agriculture, Forestry and Fishery.

Classification Number of Damaged Losses in Thousand Yen

Note

Agriculture 14,108,580

Forestry 1,102,045

Fishery 4,097,151

Total 19,307,776

Table II- 7 Damage to Education.

Classification Number of Damaged Losses in Thousand Yen

Note

School 1,529 8,528,197

Cultural Assets , - 24 5,940

Facilities for Education

and Welfare

682,503

Total 9,4 56,640

VIII-91
Table II- 8 Damage to Facilities for Flood Control and Preservation

of Mountainous Areas.

Prefecture Number of Damaged Losses in Thousand Yen

Note

River Facilities : 482

Miyagi Pre. Seaside Facilities : 14 Sabo Facilities : 15

13,693,698

Iwate Pre. River Facilities : 31 Seaside Facilities : 2 443,870

Fukushima Pre. River Facilities : 3 Sabo Facilities : 21 94,980

Yamagata Pre. River Facilities : 1 5,388

Total 14,237,936

Table II- 9 Damage to Transportation Facilities

Classification Number of Damaged Losses in Thousand Yen

Note

Roads 2,317 5,614,184

Bridges 252 5,817,009

Railways 6,377,883

Harbors 89 4,000,872

Others 762,165

Total 22,572,113

Table II-10 Damage to Water Supply Facilities.

Prefecture Number of Damaged Losses in Thousand YenNote

Miyagi Pre. 1,733,399

Iwate Pre. 14 10,105

Fukushima Pre. 0

Yamagata Pre. 2 1,444

Total 1,744,948

VIII-92
Table 11-11 Losses of Water Supply Facilities in Sendai City

Classification Losses in Thousand Yen

Receiving and Storing Reservoirs 1,800

Filtration Plants 12,400

Distribution Stations 33,400

Conduits 111,400

Supplying Facilities 13,600

Buildings 5,300

Others 77,100

Total 255,000

Table 11-12 Damage to Gas Supply Facilities.

Prefecture Number of Damaged Losses in Thousand Yen

Note

Miyagi Pre. 190 947,000

Total 947.000

VIII-93
Table 11-13 Damage to Hydroelectric Power Plants, Thermal Power

Plants and Transformer Stations in Miyagi Prefecture.

Classification Number of Damaged

Hydroelectric Power Plants 11

Thermal Power Plants 2

Transformer Stations 18

Total 31

Table 11-14 Damage to Sewerage Facilities.

Classification Number of Damaged

Losses in Thousand Yen

Note

Drainpipes 60 578,332

Pump Yards 19 142,231

Treatment Plants 23. 52,245

Others 4 65,64 5

Total 106 838,453

VIII-94
140° 42°

141' 142°

-42°

36°- -36° 139° 140° 141° 142°

Fig. II- 1 Casualties and Losses in Miyagi, Iwate, Fukushima and

Yamagata Prefectures.

VIII-95
140° 141° 142° 42°- -42°

36° -36° 139° 140° i4r 142°

Fig. II- 1 Casualties and Losses in Miyagi, Iwate, Fukushima and

Yamagata Prefectures.

VIII-96
Fig. II-3 Damaged Sites Distribution ot Conduits in Sendai City.

VIII-97
VIII-98
Fig. II-5 Damage Distribution of Gas Pipes

VIII-99
VIII-100
Table III-l Residential Damage and Refuge in Sendai City till June 15 after the EarMiqualce

Oate L/a.lIla.gt;U rXLCa. States of Damage Number of Refugees

oneiter Note

June 13

0: 30 Mldorigao ka Danger of landslides Cracks in roads About 70 inhabitants Kano Elementary

School About 80 refugees on

the evening of June 13. About 100 refugees on

the evening of June 14

.

2: 00 Futatsusawa Landslides 20 inhabitants of

4 households About 20 refugees on

the evening of June 13. About 20 refugees on

the evening of June 14. Minamoto-Shinden Danger of landslides 11 infiabitants of 3 households

Kitane Assembly

Hall About 8 refugees on

the evening of June 13. About 8 refugees on

the evening of June 14. Tsurugaya Housing

Corporation Terrace

Electricity failures at multistory apartment

houses

75 inhabitants Tsurugaya First Assembly Hall (15 refugees)

Tsurugaya Third Assembly Hall (30 refugees)

Tsurugaya-higashi

Elementary School

(30 refugees)

Refuge was ended on

the morning of June 13.

Takasago Sunny

Heigiits

Bursting of a water tower

About 120 in- habitsnts Takasago Elementary

School Refuge was ended onthe morning of June 13.

June 13

16: 00

Su naostii-cho Landslides 5 inhabitants of a household A house of an

acquaintance

" Sunaoshi-cho Landslides 6 inhabitants of a household A house of an

acquaintance

Komatsushima Landslides 1 household A house of an

acquaintance

June 14 12: 30

Anyoj i Landslides 1 household A house of an

acquaintance

Kitane-ichinenbo Collapse of an apart- ment house by landslides

1 1 inhabitants of

7 households Houses of acquaintances

Danger of the collapse

of an apartment house

by landslides

9 inhabitants of

9 households Houses of acquaintances

June 15

13: 20 Kanisawa Danger of the collapseof a house

1 household A house of an acquaintance

VIII-101
Table III-2 Damage and Restoration Methods in Designated National Highways

(More than 1 Million Yen of the Cost of the Restoration)

Number

of the Route Damaged Sites Kilometer Damage

Restoration Method

Post (km) Main Damage Other Damage

4 Kanairo, Nihonmatsu

City

249.2 Swelling of the protection blocks of the slope

Setting up of blocks and

tion wall

Shindate, Shiroishi

City

303.9 Cracks in the pavement Pavement of the road

Cm T"f1 f> Zao Town, Katsuta Cou nty

313.6 Sinking of the block piles Piling of blocks

Kooriyama, Sendai

City (Chiyo Bridge)

347.3 Sinking of the road surface near the abutment Pavement of the road

surface

Okino, Sendai City

(Chiyo Bridge)

34 7.5 Damage to the pier Reinforcement of the pier by wrapping with concrete

Okino, Sendai City

(Chiyo Bridge)

347.7 Sinking of the road surface near the abutment

Asphalt pavement of the road surface

Furujiro, Sendai City 348.4 Sinking of the road surface near the artificial structure Oroshimachi, Sendai

City

„

351.7

352.0

352.3

352.4

Hinode-cho, Sendai

City

353.1

353.8

354 .0

Sinking of the road surface near the artificial structure

Asphalt pavement of the road surface

w.iy

]\.\J i- \-l f iJdlVJdl V^ILy 354 .2

354 .4

„

Tsubamesawa,

Sendai City

355.3 .,

355.5 Landslide at the artificial bank

Sinking of the side- walk and damage to the guard pipe

Restoration of the artificial bank, pavement of the sidewalk and restoration of the guard pipe

Nanakita, Izumi City 363.5 Cracks in the pavement Pavement of the road surface Sawada, Furukawa

City

397.0 Sinking of the road surface near the abutment

Asphalt pavement of the road surface Takashimizu Town,

Kurihara County

404.4 Inclination of protection

shelf of the artificial bank

Sinking of the sidewalk and inclination of the side blocks

Construction of the protection concrete wall, pavement of the side- walk and resetting of the blocks

6 Kashima Town, Soma County

283.4 Cracks in the abutment Reinforcement of the abutment Yamamoto Town,

Watari County

1 .J Sinking of the road surfaceAsphalt pavement of the road su rface

Watari Town, Watari

County

321.9 Sinking of the shoulder

of the road Concrete pavement of the road surface

333.6 Damage to the pier Damage to the shoe Reinforcement of the pier by wrapping with concrete and restoration of the shoe VIII-102
(Table UI-2 continued) Number

of the Route Damaged Sites Kilometer Damage

Restoration Method

Post (km) Main Damage Other Damage

45 Fukudamachi, Sendai

City

7.3 Landslide at the artifi- cial bank

Sinking of the sidewalk

and damage to the protection shelf

Protection of the artificial bank with sheet

piles, pavement of the sidewalk and resetting

of the protection shelf Matsushima Town,

Miyagi County

„

28.3

28.5

Inclination of protection

shelf of the artificial bank

Sinking of the road surface near the alxitment

Sinking of the sidewalk

with cracks

Protection of the artificial bank with block

piles and pavement of the sidewalk

Asphalt pavement of the road surface

28.6 Landslide at the artificial bank

Sinking of the sidewalk

with cracks

Restoration of the artificial bank and pave- ment of the sidewalk

29.1 Inclination of the block

piles

Inclination of the protection wall of the sidewalk

Restoration of the protection wall

Naruse Town, Monou

County

37.2 Sinking of the road surface with cracks

Pavement of the road

surface

Naruse Town, Monou

County (Ono Bridge)

38.0 Movement of the girder Damage to the shoe,

the joint, the floor

plate and the pier

Restoration of the girder, the joint, the floor

plate and the pier

Kanan Town, Monou

County (Tenno Bridge)

58.3 Cracks in the pier Damage to the abut- ment and the shoe Reinforcement of the pier and the abutment

and renewal of the shoe

Kanan Town, Monou

County

66.0 Cracks in the pavement Pavement of the road

surface

Shizugawa Town,

Motoyoshi County

98.4 Cracks in the artificial bank Cracks in the protection wall and the pave- ment

Restoration of the artificial bank with protection piles

Utatsu Town,

Motoyoshi County

(Utatsu Bridge)

102.5 Damage to the shoe Reinforcement of the shoe

Shimofunato, Oofunato

City

165.5 Cracks in the pavement Sinking of the road sur- surface Pavement of the road

surface

Noda Village, Kunohe

County

336.2 Inclination of the protection concrete wall

(Renewal of the protection concrete wall

108 Kanan Town, Monou

County

5.0 Inclination of the protection block wall

Sinking of the sidewalk

with cracks

Piling of blocks and pavement of the side- walk Kanan Town, Monou

County

5.8 Sinking of the road surface Damage to the block

piles

Asphalt pavement of the road surface and piling

of blocks

14.7 Cracks in the stone piles Cracks in the pave- ment

Construction of the protection concrete wall and the pavement of the road surface

VIII-103
Table III- 3 Temporary Water Supply after the Earthquake till June 21 in Sendai City (Water Supply Bureau of Sendai Municipal Government)

Water Wagon Personnel Water

oupply Bureau (Wagons)

Water

Assistance (Personnel)

Amount of

Date

Total (Wagons)

Assistance (Wagons)Total

(Personnel)

Supply Bureau

(Personnel)

Number

of Cases TemporaryWater Supply (m"^)

Main Water-Supplied

Districts

June 12 5 5 0 20 20 0 7 7 Tsurugaya Housing Corporation Terrace

13 29 7 22 72 50 22 165 180 Asahigaoka, Midorlgaoka

Jiyugaoka,

14 37 7 30 90 60 30 213 230 Futabagaoka, Izumigaoka

Kofu kugaoka

15 27 7 20 70 50 20 120 130 Kitaneichinenbo

-J Tomisawa, Fukurobara

Idehana, Aral

Izai, etc.

16 17 7

- 10 50 40 10 60

- 65

17 10 6 4 30 26 4 » 26 30 Kuromatsu, Saiwai-cho Housing Corporation

18 10 6 4 30 26 4 18 20 Apartment House

Sakuragaoka, Tsurugaya

19 8 5 3 18 15 3 18 20 Apartment House

Sunny Heights 20 4 4 0 9 9 0 7 14 Dainohara Lions Mansion

Nissui Apartment House

21 2 2 0 5 5 0 3 6 Toroku Elementary School,

etc

.

Total 149 56 93 394 301 93 637 702

Table III-4 Assisting Activities in Temporary Water Supply and Restoration of Conduit Pipes (Based on Mutual Assisting Plan of Tohnku Local Branch of Japan Water Supply Association)

Prefectural BranchAssisting

City Assisting Period Accepting

City Assisting Activities Disp<Jt(-h Items

Aomori Hachinohe

City June 13 - June 16 Shiogama

City

Temporary water supply 15 water tanks, 10 vehicles. 20 drivers of related companies, 4 personnel of water supply bureau.

Akita Akita City June 14 - June 18 Izumi City Restoration of conduit pipes

7 personnel of 2 related companies,

2 vehicles.

Yamagata Yamagata

City

Yonezawa

City

Kaminoyama

City

Sagae City

June 14 - June 16

June 15 - June 16

June 14 - June 16

Shiogama

City Izumi City

Shiogama

City

Temporary water supply

Restoration of conduit pipes

Temporary water supply

A water wagons, 3 personnel of water supply bureau.

4 personnel of a related company

A water wagon, 3 pev.snnnel of water supply bureau. A water wagon, 2 personnel of water supply ixireau. A water wagon, a personnel of water supply bureau.

Fukushlma Fukushlma

City

Koriyama

City

Haranomachi

City

June 14 - June 17 Izumi City Restoration of conduit

pipes

4 personnel of related companies.

4 personnel of related companies.

4 personnel of related companies.

VIII-104
(Table 111-4 continued)

Prefectural BranchAssisting

City Assisting Period Accepting

City Assisting Activities Dispatch Items

Iwate Morioka

City June 14 - June 18 Izumi City Restoration of conduit

pipes, water-flow test and investigation of water- leakage

June 14-June 17, 16 personnel of water

supply hwreau, 7 vehicles. June 17-June 18, 4 personnel of water supply bureau, 16 personnel of related companies.

4 vehicles of related companies.

M Izusawa

City June 14 - June 17 Restoration of conduit

pipes A personnel of water supply Ixireau,

4 personnel of related companies,

2 vehicles

.

Kitakam i City June 14 - June 1 7

., A personnel of water supply bureau,

4 personnel of related Companies,

/ vehicles

.

Hanamaki

City June 17 - June 18 A personnel of water supply bur'.'au,

4 personnel of related cornpanicK,

2 vehicles.

Miyagi Kesennuma

City June 15 - June 18 Izumi City Temporary water supply 2 water wagons

.

June 17 - June 1 9 Shiogama

City Restoration of conduit

pipes

3 parties of related companies.

Sendai City June 15 - June 17

June 1 7 - June 19

Izumi City

Shiogama

City

2 parties of related companies.

3 personnel of water supply bureau,

5 parties of related companies.

Table III-5 The Number of Personnel Required for the Restoration of Electric Power Facilities till June 30 after the Earthquake

(Tohoku Electric Power Co., Ltd.)

\. Items

Facilities \.

Personnels of Tohoku Electric Power

Co., Ltd.

(man-days)

Other

Personnels

(man-days)

Total

(man-days)

Note

Thermal Power

Generation

3,433 12,902 16,335

Hydroelectric

Power Generation

219 130 349

Transformer 1,099 1,878 2,977 Assistance by Tokyo

Electric Power Co.,

Ltd.

Power Transmission

1,107 283 1,390

Power

Distribution

4,003 3,317 7,320

Communication 339 194 533

Total 10,200 18,704 28,904

Note : Other personnels are from Tokyo Electric Power Co., Ltd., related

companies and electrical machinery companies.

VIII-105
Table III-6 Damage to Gas Pipes (Gas Supply Bureau of Sendal Municipal Government)

\ Damage

Block \

Trunk Pipe Bough Pipe Branch Pipe Twig Pipe

Total

Severance

Crack

1

Pulled

Out

Slackness

of

Joint |

The

Others

Total

Severance

Crack Pulled

Out

o

"o

to

CO

<u 5.u

nj

tyi

11

The

Others

Total

Severance

Crack Pulled

Out

Slackness

of

Joint

The

Others

Total \

Severance

Crack rpulled

Out

Slackness

of

Joint j

The

Others

Total

Number

Percentage

J 2 1 1 1 5 10 3 14 30 3 2 35 31 2 1 3 37 91 16.5%

11 1 1 3 1 1 5 4 4 7 9 16 26 4.7

III 5 1 1 7 81 9 27 3 5 125 71 5 1 2 79 69 6 1 3 79 290 52.6

IV 3 3 8 5 13 4 4 3 1 1 5 25 4.5 V 1

1

1 2 1 1 4 3 1 4 9 2 1 I 13 22 4.0

VI 1 6 1 7 8 8 16 2.9

Vll 1 1 3 3 10 1 11 15 2.7

Vlll 1 1 22 3 10 1 1 37 6 2 1 ~¥ "n 4 16 63 11.4

Tagajo, Fukuda,

Fukuzumi

1 1 2 2 1 1 4 0.7

Total Number 3 1 11 4 1 20 127 14 46 5 7 199 129 8 2 2 6 147 150 20 1 2 13 186 552

Note : According to the data of restoration works from June 12 till July 8.

Table III- 7 Damage to Dykes on Rivers and Their Restoration

Name of the River

Length of Administrated Section (km)

Year of Improve- ment

Undertaken

Total Length of Dykes (km)

Length of Damaged

Dykes (km)

Note

Finished

Provisional Total Emergency

Disaster

Emergency

Maintenance

4.2

Total Restoration

Lower Reaches

of Kitakami

River

137.5 1911 78.5 104.6 183.1 5.3 (0.052)

9.5 8.9 Including Eai River

Naruse River 82.4 1921 22.2 111.0 133.2 4.3 6.4 (0.080) 10.7

8.1

Natori River 29.4 1941 28.6 3.4 32.0 2.3 9.0 (0.353)

11.3

3.0

Lower Reaches

of Abukuma

River

63.7 1936 20.5 53.5 74.0 0.1 2.5 (0.035)

2.6

0.6

Total 313.0 149.8 272.5 422.3 12.0 22.1 (0.081)

34.1 20.6

Note : ( ) in the column of "Length of Damaged Dykes" gives the ratio of length of damaged dykes to total length of dykes

.

Length of "Restoration" is at the point of dykes reported to have been suffered disaster, and its restoration costs more than 1 million yen, and Che length is included in "Total". Numerical values in the column of "Emergency Disaster" gives the length of dykes coped with by emergency temporary restoration, and numerical values in the column of "Emergency Maintenance"

gives the length of dykes coped with by the budget for maintenance expenses. VIII-106
Director Vice Director Head Officer

(Governor) (Vice Governor)

Direction and Investigation Party

— Planning Division — General Affairs Division — Public Welfare Division — Sanitary Division — Commerce and Industry

Division — Agriculture Division — Marine and Forestry

Division — Public Works Division — Development Division

Revenue and Expenditure

Div ision — Education Division — Overland Transportation

Division — Police Division — Local Branch

Fig. III-l Structure of Rescue and Rehabilitation Center of

Miyagi Prefectural Government.

VIII-107
Disaster Prevention Council

of Miyagi Prefecture

Designated Local

Administrative Organizations

Miyagi Prefectural

Police Headquarters

Public Bodies

Disaster Countermeasure

Headquarters of Miyagi

Prefectural Government

Disaster Countermeasure

Headquarters of Each

Municipal, Town and Village

Government

Inhabitants

Designated Public

Organizations

Miyagi Prefectural

Board of Education

Disaster Prevention

Council of Each City,

Town and Village

Administrators of Important Facilities for Disaster Prevention

Fig. III-2 Disaster Countermeasure Structure in Miyagi Prefecture.

VlII-108
Fig. in-3 Regional Division for Restoration Works by Gas Supply Bureau

of Sendai Municipal Government.

VIII-109
Rfc'gional Division inrc blocks and seLlions Confirmation of closing cock di every house

Preliminary loading prcssuri.'

jt everv bloLk and section

Investigation of gas leakage by field patrol

Restoration of damaged gas pipes

Measurement of leaking-gds volume lo-K.

NO

Purging and loading pressure

for FID investigation

FLO investigation

Restoration of damaged gas pipes

Measurement of leaking-gas volume

O.K.

Holding pressure loaded

NO

Purging before gas supplying

Inspection of gas pipes

inside of every house NO Resroration of damaged gas pipes

O.K.

Reopening of gas supply

Fig. III-4 Flow Chart of the Restoration of Gas Supply Facilities from

Regional Division to Reopening of Gas Supply

(Gas Supply Bureau of Sendai Municipal Government)

VIII-110
Fig. I1I-5 Arrangement of the Facilities and Outflow of Oil at Sendai

Oil Factory of Tohoku Petroleum Co., Ltd.

Table IV- 1 Budgets of Reconstruction Works of Public Utilities in Miyagi Prefecture

Classification

Prefectural Budget Municipal Budget Total Applied Assessed Applied Assessed Applied Assessed

Sites Budget Sites Budget Sites Budget Sites Budget Sites Budget Sites Budget

Rjver Works 138 1,912,146 139 1,573,696 24 113,303 24 91,112 162 2,025,449 163 1,664,808

Coastal Works 14 433,656 14 350,360 14 433,656 14 350.360

Sabo Works 1 129,256 1 98,783 1 129,256 1 98,783

Highways 339 1,683,673 335 1,401,613 219 1,526,148 214 1,260,573 558 3,209,821 549 2,662,186

Bridges 40 319,668 39 272,111 26 414,027 26 346,570 66 733,695 65 618,681

Total 532 4,478,399 528 3,696,563 269 2, 053,4 7« 264 1.698,255 801 6,531,877 792 5,394,818

VIII-111
Table IV- 2 Budget of Reconstruction Works of Water Supply Facilities

Classification Budget in Thousand Yen

Receiving and Storing Reservoirs 565

Filtration Plants 9,057

Distribution Stations 14,126

Conduits 79,303

Others 74,321

Total 177,372

VIII-112
SOCIAL ASPECTS OF EARTHQUAKE MITIGATION

AND PLANNING IN THE UNITED STATES

William A. Anderson

National Science Foundation

Washington, D.C.

ABSTRACT

A number of social scientists in the United States have recently turned their attention

to the pre-disaster responses of individuals and social units and are investigating the

socioeconomic factors related to mitigation and preparedness. This research is important

because it promises to result in the identification of the principal factors which influence

the utilization of known and developing social and technological adjustments to earthquakes,

including building codes, land use regulations, public education and earthquake prediction.

As a result of on-going and recently completed social science studies, increased knowledge is

developing on a number of important topics. The emerging findings on the following questions

are discussed in this paper: How is the earthquake hazard perceived by officials and the

public? What types of mitigation and preparedness measures are citizens and officials willing to accept? What is the impact of some existing hazard mitigation programs? What are

some of the social factors to consider in trying to enhance the benefits of future earthquake

predictions?

KEYWORDS: Mitigation and preparedness measures; socioeconomic factors; hazard awareness.

VIII-I13
INTRODUCTION

Systematic research on disasters and natural hazards by social scientists in the United

States is relatively new. It was not until the 1950s that a significant program of disaster

research began emerging in the social science field in the United States. For a long time

the field was hampered by a lacking of funding, and even today social and behavioral scientists working in the area of disaster and natural hazards research receive only a fraction of

the financial support received by colleagues in the physical sciences and engineering.

During the last decade, however, there has been a manifold increase in disaster and

natural hazards studies conducted by American social scientists. The principal catalyst for

this has been the availability of more dependable funding and an increase in well-trained

social and behavioral scientists interested in conducting research in the field. Nevertheless, there is still considerable uneveness in existing knowledge. [1] Much of the accumlated knowledge is on the responses of individuals, groups and institutions during the

immediate post-impact period of disaster. [2] On the other hand, one of the most neglected

periods has been the pre-disaster period, the time when mitigation and preparedness actions

can be undertaken by threatened populations. ...

Fortunately, since the mid-1970s, a small number of social scientists in the United

States have turned their attention to the pre-disaster responses of individuals and social

units and are investigating the socioeconomic factors related to mitigation and preparedness

for earthquakes and other hazards. Social science research on earthquake mitigation and

planning can be a vital complement to research conducted by engineers and geophysicists

.

This research is important because it promises to result in the identification of the principal factors which influence the utilization of known and developing social and technological adjustments to earthquakes, including building codes, land use regulations, educational

programs, and earthquake prediction.

In the rest of this paper, I will briefly discuss some of the findings, in some

instances tentative ones, of a few recent and on-going studies supported by the National

Science Foundation on the socioeconomic aspects of mitigation and preparedness and various

issues such investigations have thus far raised. I will particularly note emerging answers

to the following questions: How is the earthquake hazard perceived by officials and the

public? What types of mitigation and preparedness measures are citizens and officals

VIII-114
willing to accept? What is the impact of certain existing hazard mitigation programs? What

are some of the social factors to consider in trying to enhance the benefits of future earthquake predictions? It should be stressed that some of the research findings reported below

are the result of preliminary analyses of data by researchers involved in on-going projects

and are subject to some modification once the final analyses have been completed.

HAZARD PERCEPTION

Awareness is a key factor in reducing the impact of earthquakes and other hazards on

society. [3] If individuals and groups have little awareness of, or concern for, the earthquake hazard, they cannot be expected to give serious consideration to developing mitigation

and preparedness measures. Thus, it is important to know how the earthquake hazard is perceived by various segments of society. Such information would provide insight into the types

of education efforts that might be required and also offer a more reaslistic basis for predicting support for particular kinds of hazard reduction policies and programs.

Information is currently emerging from the social sciences in the United States on how

both public officials and private citizens perceive the threat of earthquakes and other

natural hazards. Among the findings which have surfaced thus far is that such problems have

a minor place on the political agenda of most states and local communities in the country.

Put simply, most influential groups exhibit little real concern for such issues as hazard

mitigation and preparedness at this time.

For example, sociologists at the University of Massachusetts conducted a national survey

in 1977 concerned with the local and state politics of natural hazards policies. [4] Interviews were conducted with over 2300 key local and state persons. The researchers found that

these political influentials do not perceive natural hazards as a very important problem in

comparison to the many other issues they must consider.

A similar conclusion has been reached by political scientists at the University of

California, Santa Barbara following a preliminary analysis of data from their study on seismic safety preparedness of local governments in California. [5] This study focuses upon 14

local jurisdictions, including 12 cities and 2 counties. The investigators found that a concern for seismic safety was almost universally low among community leaders.

There is also evidence that concern for earthquakes in the United States is very low

among private citizens. For example, sociologists at the University of California, Los

VIII-115
Angeles have completed a survey of 1450 residents of Los Angeles County as part of an

on-going study of the social response to the southern California Uplift, an anomaly which

some scientists have interpreted as a possible precursor to a major earthquake in that

area. [6] Though respondents in the study had several opportunities to mention earthquakes

as one of sourthern California's serious problems, the investigators report that few did so.

For example, only 35 persons in the sample mentioned earthquakes as one of the 3 most important problems facing residents in southern California today.

Findings from such studies, then, clearly indicate the difficulties that lie ahead for

creating earthquake hazard awareness in the United States. With so little concern for this

hazard, widespread acceptance of mitigation and preparedness measures in endangered areas

may be unlikely at this time.

ACCEPTANCE OF MITIGATION AND PREPAREDNESS MEASURES .

A few of the recent social science studies conducted in the United States have provided

a clearer understanding of the extent to which people find particular mitigation and preparedness measures acceptable. For example, one form of financial protection available to

individuals in threatened areas is hazard insurance. A study was recently completed by

economists at the University of Pennsylvania to determine the extent to which people took

advantage of this type of protection. [7] Part of the study involved a survey of homeowners

in 18 earthquake-susceptible areas of California. What they learned was not particularly

encouraging. It was found that a lack of concern for earthquakes, combined with inadequate

information on the availability and terms of earthquake insurance, resulted in few sales of

this form of protection. Similar results were found regarding flood insurance in the study.

In the study of the sourthern California Uplift previously mentioned, the principal

investigator, Ralph Turner, and his associates asked respondents if they had suggestions

regarding what the government should be doing to prepare for future earthquakes. The

researchers learned that most people had some concern for this issue and had specific suggestions to make. Most respondents offered suggestions on hazard reduction activities, such as

taking steps to enforce building codes, rather than on improving the state of emergency preparedness to handle disaster-generated problems. The investigators noted that it was

VIII-116
reassuring to detect this potential public support for mitigation programs, but they were

also quick to point out they could not be certain of the amount of such support in real

situations

.

The Turner study also shows that while most of the respondents indicated a willingness

to see government action to mitigate earthquake hazards, few had taken actions to prepare

their own households for an earthquake; for example, reported having taken such steps as

storing supplies of water and food and rearranging shelves to prevent objects from falling

on occupants. However, there was a tendency for persons with more suggestions for government officials to be better prepared for earthquake themselves.

IMPACT OF EXISTING PROGRAMS

Little is known about the impact of many existing programs in the United States that

attempt to achieve greater earthquake hazard reduction and awareness. Evaluations of such

programs are highly desirable in order to make them more effective in the long run.

Two on-going studies are assessing the impact of programs in California on individual

and community concern for seismic safety and actions taken in this regard. The previously

mentioned University of California, Santa Barbara study has looked at the impact of legislation requiring cities and counties in California to develop a seismic safety element as part

of their general plans. This element is to consist of an identification and appraisal of

earthquake hazards in the local jurisdiction and is expected to serve as part of the basis

for mitigation in the area. The 1971 San Fernando earthquake provided the impetus for this

and some other new laws related to earthquake safety. After a preliminary analysis of their

data, the researchers have concluded that the seismic safety elements have had little policy

impact in the 14 jurisdictions they studied. They report that the development of the seismic elements or plans has had some consciousness-raising impact, but that with few exception

little has been done to implement them and jurisdictions are not better off than before in

terms of emergency preparedness, in dealing with old hazardous buildings, and in public

education. .

In 1975 California also passed legislation requiring a seller to inform prospective

buyers if a property is located in an earthquake hazard area, referred to as a special

study zone. An on-going study by geographers at the University of Colorado is evaluating

this disclosure measure. [8] As part of the study, a survey has been undertaken in two VIII-117
California communities to determine the salience of earthquake hazards in the house selection

process and to measure the impact of disclosure on the search for and purchase of a house. A

preliminary analysis of the data by the researchers shows that over half of the persons contacted knew their homes were in an earthquake hazard area or special study zone. And almost

all the persons who knew about the zone said that this knowledge had no effect on their decision to purchase their home. Data from the study initially supports the conclusion then that

the earthquake hazard has little or no importance to most buyers. This study will not be

completed until 1980. The researchers hope that at its completion a number of recommendations can be offered regarding this particular disclosure measure and the role that such

approaches in general can play in hazard mitigation and planning. This is particularly

important since other states have considered such measures.

SOCIAL FACTORS TO CONSIDER IN PREDICTION

Prediction is one of the most promising means to mitigate the impacts of earthquakes.

Future earthquake predictions promise to provide long lead times, thus enabling significant

hazard reduction and emergency preparedness activities to be undertaken. For example, a

study on the possible socioeconomic effects of earthquake predictions conducted by sociologists at the University of Colorado suggests that a credible prediction of a major earthquake in the United States will serve as the impetus for a number of mitigation and preparedness activities that will reduce loss of life and property. [9] This conclusion was based on

results from a survey of private citizens, and government and private officials in two California communities. However, this study also suggests that a long-term earthquake prediction

has the potential to create serious socioeconomic disruption in a community. For example,

businesses could limit mortgage loans, the availability of insurance and investment in a

prediction area. Additionally, the researchers note that there could be a possible net outmigration and reduced tourist trade, leading to increased unemployment, falling property

^ values and reduced tax revenue.

One of the important tasks social scientists can perform in the hazards field, then,

is conducting research that identifies socioeconomic, political and legal factors which officials should take into account in order to maximize the benefits of prediction and minimize

its negative consequences. This role is exemplified in the case of a recent study completed

at the Association of Bay Area Governments of California on the legal liability of government

VIII-118
in connection with earthquake hazards reduction. The researchers note that local and state

governments are potentially liable in earthquake prediction and warming situations. [10]

In their view liability could result from problems growing out of the issuance of a

warning or the failure to issue one. The study concludes that there is a need for a form

of immunity to cover public forecast situations when officials take reasonable actions. The

researchers feel that by dealing with immunity in this fashion on an obstacle to prompt and

effective government utilization of earthquake prediction and warning would be removed.

The southern California Uplift situation which Turner and his associates are studying,

while not involving an actual prediction, does serve as a prototype of the possible initial

stage of a prediction. Attention has been given to the situation by the media, the California Seismic Safety Commission has expressed concern about it to officials in the state, and

the U.S. Geological Survey has been conducting special studies on the anomaly. The Uplift

situation, then, can be viewed as a near prediction and Turner and his colleagues have been

investigating it for insight it might offer on the likely social responses to an actual prediction in the United States. The researchers have learned that social groups differ in

their awareness of the Uplift and in understanding its potential significance. Thus one of

the cautions they offer is that groups will vary in their awareness and appreciation of an

actual earthquake prediction and therefore responsible officials must be prepared to launch

special efforts to insure that certain groups, such as the less educated and the poor, are

made fully aware of any future earthquake prediction.

CONCLUSION

Some progress is now being made in the United States in developing knowledge on the

socioeconomic factors related to earthquake mitigation and planning. Work by social scientists in this field will hopefully make the research conducted by geophysicists and engineers

even more useful to policy makers.

There is now a need for social scientists in the United States to become involved in

research on mitigation and planning in other societies. The real challenge is for American

social scientists to engage in cooperative research efforts with colleagues in other countries, including Japan. Such joint research ventures would tell us what extent our accumulated knowledge can be generalized across societies. They would also facilitate the

diffusion of innovative hazards reduction policies and programs among societies.

VIII-119
REFERENCES

[1] Taylor, V. A., "Future Directions for Study," in E. L. Quarantelll, ed . , Disasters:

Theory and Research (Sage, Beverly Hills, 1978).

[2] Mileti, D. S., Drabek, T. E., and Haas, J. E., Human Systems in Extreme Environments:

A Sociological Perspective (University of Colorado, Boulder, 1975).

[3] Working Group on Earthquake Hazards Reduction, Earthquake Hazards Reduction: Issues

for an Implementation Flan (Office Science and Technology Policy, Washington, D.C.,

1978)

.

[4] Rossi, P. H., Wright, J. D. , Wright, S. R. , "Social Science and Natural Hazards, Social

and Demographic Research Institute, University of Massachusetts, May, 1979.

[5] Mann, D. , Post-Disaster Response, Natural Hazards Research Applications Workshop,

Boulder, Colorado, July 1979.

[6] Turner, R. H. , Nigg, J. M., Paz, D. H. , and Young, B. S., Earthquake Threat: The Human

Response in Southern California (Institute for Social Science Research, University of

California, Los Angeles, 1979).

[7] Kunreuther, H. , Disaster Insurance Protection: Public Policy Lessons (John Wiley and

Sons, New York, 1978)

.

[8] Palm, R. , "Real Estate Agents and Dissemination of Information on Natural Hazards in

the Urban Area," Natural Hazards Research Applications Workshop, Boulder, Colorado,

July 1979 . _ , .

[9] Haas, J. E., and Mileti, D. S., Socioeconomic Impact of Earthquake Prediction (Institute of Behavioral Science, University of Colorado, Boulder, Colorado, n.d.).

[10] Margerum, T., Will Local Government Be Liable for Earthquake Losses? (Association of

Bay Area Governments, Berkeley, 1979).

VIII-120
ORIENTATION OF TSUNAl-II RESEARCH IN JAPAN

Hiroshi Takahashi

Yukio Fujinawa

National Research Center for Disaster Prevention

Science and Technology Agency

ABSTRACT

We have been struck by large tsunamis which were induced by earthquakes occurring near

the Tonankai area of Japan on an average of every thirteen years in a 350 year period. It is

conjectured by some seismologists that a large earthquake may occur in the area of Tonankai.

We have experienced nine earthquakes which are considered to be grouped in the Tonankai

earthquake zone. All of these earthquakes were accompanied by tsunamis. Victims rose to

31,000 in 1498 (Meio), 3,906 in 1605 (Keicho), 4,924 in 1707 (Hoei), and 3,427 in 1854

(Ansei), respectively. We urgently need to plan countermeasures against tsunamis, especially

to develop a warning system for tsunamis which take less than about ten minutes to reach our

coasts. Tsunami research in Japan is briefly reviewed from the standpoint of developing an

effective warning system.

KEYWORDS: Disaster warning; earthquake detection; flood warning; Tsunami prediction;

Tsunami research.

IX1
1. DISASTROUS TSUNAMIS IN JAPAN

Some 130 tsunamis have been recorded in this country over a period of some 1,500 years.

A scale of tsunamis is sometimes defined by wave height along the coast as shown in Table 1.

The number m is denoted as tsunami magnitude. Large tsunamis in this country are listed in

Table 2. The biggest three are the 1896 Sanriku tsunami, the 1792 Shimabara tsunami and the

1771 Okinawa tsunami.

The epicenters of earthquakes which caused tsunamis are shown in Figure 1. The coastal

area along the Pacific Ocean were frequently struck by huge tsunamis. Return period of

tsunamis with magnitude larger than 2 is about 10 years, and larger than 3 is about 30 years

Thus, an improved tsunami warning system ranks high in regard to coastal disaster prevention

needs.

Various measures are taken to prevent disasters by tsunamis. The tsunami warning system

plays an important role in evacuation measures. As a result of the great damage by the 1946

tsunami in the Hawaii Islands, an international tsunami warning system named the Seismic Sea

Wave Warning System (SSWS) was instituted to provide information on tsunamis to countries on

the Pacific Ocean. Further improvements to such a warning system based on tsunami monitoring

data are needed in Japan. , i'

The tsunami magnitude is predicted by the Japanese Meteorological Society (JMA) immediately after the occurrence of earthquakes. The prediction is based on tsunami statistics

gathered in this country by lida [1]. His result is reproduced in Figure 2, which is a graph

of simultaneous plots of earthquake magnitude M, hypocenter depth H, and tsunami magnitude m.

As shown in the figure, the magnitude of an earthquake accompanying a tsunami contains some

lower limit, Mc, and Mc is related to the hypocenter depth H as.

2. PRESENT METHOD OF TSUNAMI PREDICTION

3. ADOPTED METHOD OF TSUNAMI MAGNITUDE PREDICTION

Mc = 6.3 + 0.01 H

IX-2
In the case of a large tsunami with magnitude larger than 2 the relation is,

Mc' = 7.75 + 0.008 H

On the base of these empirical results, various types of warning are given by the JMA.

It is a problem that some tsunamis do not follow this empirical relation with the result

that erroneous warnings are sometimes Issued. For instance, the diagram predicts only a

minor tsunami for the 1896 Sanriku tsunami. Some improved interpretations can be made using

the results of Hatori [2] who analyzed tsunami statistics limiting himself to those which

occurred near the Sanriku district. His results are shown in Figure 3. Correspondence

between tsunami magnitude m and tsunami source area is very good, which is considered to be

due to the existence of a type of fracture in a fixed area.

4. QUANTITATIVE EVALUATION OF TSUNAMI MAGNITUDE

Evaluation of tsunami wave height requires knowledge of the following four stages of a

tsunami:

1) Generation

2) Propagation • .~

3) Resonance

4) Run-up ; • .

i) Generation of Tsunami

If we can obtain the size of the deformation at the ocean bottom, the ocean surface

deformation can be evaluated, at least in principle. Such a calculation was done in the

realm of the long wave approximation by Kajiura. However, the most difficult problem is that

the actual ocean bottom deformation area is rarely observed. We are compelled to estimate

the area and position of ocean bottom movement by the use of various indirect methods.

Representative methods are: ,

1) reverse refraction method (deformation area) Sr

2) aftershock activity Sa

3) surface projection of fault plane ... (geodetic data) .. Sg

4) surface projection of fault plane ... (seismic data) ... Ss

IX-3
In the case of the 1946 Nankai tsunami, the tsunami source area, Sr, is estimated to be

much larger than the aftershock area, Sa,

Sr > Sa,

On the other hand, the tsunami source area, Sr, is nearly equal to the area Sg of surface

projection of fault plane by the use of geodetic data, ''

Sr ~ Sg

and the aftershock activity area, Sa, is nearly equal to the area, Ss, of the surface projection of fault plane by the use of seismic data,

Sa - Ss

We are led to a conclusion that there exists a situation where the ocean bottom deformation

accompanying an earthquake has a time constant too small to cause earthquakes, but large

enough to cause tsunamis. Therefore, a tsunami prediction method on the basis of just the

earthquake magnitude M and the hypocenter depth h is incomplete.

Kanamori [3] made a new contribution in approach. Frequency spectrum of effective

moment Mo(f) was introduced (Figure 4). Observed values are illustrated with several symbols,

and full lines are estimated moments with the parameter being time constant. If we can

obtain the frequency spectrum of moment by the use of several types of seismograph, we can

obtain more reliable estimates. . ,

ii) Propagation and Resonance

When tsunamis approach a shallow ocean or a bay, they undergo deformation and give rise

to a new type of motion, such as an edge wave, or seiche. Seiches along the Japanese coast

have long been observed. We set six tsunami gauges along the coast of the Sagami Bay and

observed long period ocean waves. Analysis of the data revealed the spectral structure of a

seiche at those areas and the evolution of representative modes. Oscillation of the Sagami

Bay itself is usually one of the three most important modes, as illustrated in Figure 5 [4].

Evolution of an edge wave along the curved coast has been studied by Fujinawa [5] who

calculated the spatial change of wave characteristics. Experimental and observational

studies were performed by the Earthquake Research Institute (ERI) group.

IX-4
iii) Numerical Simulation of Tsunami

The ocean bottom movement which accompanies an earthquake can be estimated somewhat more

accurately as a result of recent progress in seismology. Fault plan characteristics may be

evaluated from seismic data. Once ocean bottom deformation size is known, calculation of

tsunami magnitude can be made at least in principle. Aida [6] made a numerical calculation

in the case of the Tokachi-Oki tsunami where good tidal records were obtained along the Pacific coast of this country. Some of his results are reproduced in Figure 6. The correspondence between the observed and the calculated wave forms is good. Numerical simulation of

tsunamis will become more practical and useful as details of such problems as sea bottom

dissipation, or nonlinear deformation in a shallow region, become better known.

5. SYSTEM OF MONITORING FOR TSUNAMIS OFF THE JAPANESE COAST

We have a plan to develop a more reliable monitoring system for those tsunamis which

occur in the sea not far from our coast. The system consists of both tsunami observation

and tsunami prediction. The monitoring is performed by pressure gauges and current meters

placed on the sea bottom. Data are telemetered to land stations through marine cable, or

by radio signals from buoys which are connected to the gauges and meters. The input data

for the tsunami prediction algorithm are the direction and wave height of a tsunami which is

received at several stations. The resultant predicted value is transmitted to the authorities concerned.

REFERENCES

[1] lida, K. , Proceedings Tsunami Meeting, 10th Pacific Sa Congr. 1969, lUGG Monograph,

24, 7, (1963).

[2] Hatori, T. , Bulletin of Earthquake Research Institute, 47, 185, (1969).

[3] Kanamori, H. , Phys. Earthquake Planet Inter., 6, 346 (1972).

[4] Fujinawa, Y. et al. , Rep. Nat. Res. Cent. Dis. Prev. , 19, 117, (1978).

[5] Fujinawa, Y., Rep. Nat. Res. Cent. Dis. Prev., 21, 75, (1979).

[6] Aida, I., J. Phys. Earth, 26, 57, (1978).

IX5
Table 1 Definition of tsunami magnitude { Iida,1963)

magnitude wave height energy (xlO erg)

-1 < SOcm 0.06

0 about ^m 0.25

1 about 2m 1.0

2 4 "x- 6m 4.0

3 10^ 20m 16.0

4 >30m 64.0

Table 2 Large tsunami in Japan (Hator i

, 1977

)

year source mag.M mag.ni death year source mag.M mag.m death

684 Tokai 8.4 3 1703 Boso 8.2 3 5.233 Nanka 1 do 1707 669

Tokai 8.4 4 4.900

Sanri ku 8.6 4 1 ,000 f^ankaido 887 Kil 8.6 3 1741 Hokkaido 7.5? 3 1.467

1096 Tokai 8.4 3 1771 Okinawa 7.4 4 n .861 1099 Tokaido 8.0 3? 1792 Shimabara 6.4 2? 15.030

1361 Kii 8.4 3 1854 Tokai 8.4 3 900

1498 Tokaido 8.6 3 5.000

1854 Tokaido 8.4 3 3.000

1605 Boso 8 3 3.800

1896 Sanriku 7.6 3-4 27,122

Nankaido 1933 Sanri ku 8.3 3 3.008

1611 Sanri ku 8.1 4 6.800 1944 Tokai 8.0 2.5 998

1677 Boso 8 3 500 1946 Tokai 8.1 2.5 1,330

magnitude and earthquake

magrnitude M , hypocenter

depth H (I ida,1963)

Figure 1 Epicenters which caused large tsunami

( from Ka j iura)

IX-6
Figure 3 Relation between earthquake

magnitude M and tsunami

source area in the case of

Sanriku earthquake

(Hatori, 1969)

lOOO iOO 700 100 iO 20 10 5 2 1

O T >t >T 4r »r ,

Figure 4 Spectrum of effective moment

of earthquake (Kanamori, 1972)

Figure 5 Seiche spectrum at Misaki along the Figure 6 Result of tsunami numerical

Bay of Sagami (Fujinawa et al, 1978) simulation in the case of

the 1968 Tokachi-oki

tsunami (Aida, 1978)

IX-7
WAVE SETUP CAUSED BY TYPHOON 7010

Hiroshi Hashimoto

Taka-aki Uda

Public Works Research Institute

ABSTRACT

Typhoon 7010 caused heavy damage to Kochi city brought on by flooding from the storm

surge. Numerical computations were carried out but they do not explain fully the abnormally

high rise of the sea level. As one explanation, it is suggested that wave setup contributes to an abnormally high rise since a typhoon is usually accompanied by high waves.

Experimental investigation and numerical calculations were carried out to make clear the

magnitude of wave setup at the mouth of Kochi harbor. Generally, on a straight coast, the

normal component of radiation stress caused by breaking waves generates wave setup and the

tangential component generates along-shore current. Investigations, however, show that the

along-shore current is obstructed by a breakwater at Kochi harbor and wave setup is generated

by the tangential component together with the normal one. The wave setup contributes to the

anomaly, and its height is estimated to range from 0.5 m to 1.0 m. Numerical calculations

confirm the experimental results though they need further improvement.

KEYWORDS: Storm surge; typhoon damage; wave setup. j

J

IX-8
- 1. INTRODUCTION

A storm surge is a meteorological tide caused by strong wind and an abrupt atmospheric

pressure depression. Numerical estimations of sea levels have been carried out by many

investigators. Since a tropical low pressure system such as a typhoon is usually accompanied by high waves, wave setup becomes an additional reason for abnormal sea level rise.

Especially on a coast facing the open sea, the wave setup becomes comparable to a storm

surge

.

The typhoon 7010 which landed on the Kochi Prefecture of Shikoku Island in 1970 caused

heavy damage by storms and floods in Kochi city. At Kochi harbour, the sea level was higher

than 3 m and the anomaly reached 2.4 m. The numerical calculation of storm surge carried out

by the Meteorological Agency does not fully explain the anomaly. As one of the causes, we

think that it is necessary to consider the wave setup. In this paper, the pattern and the

magnitude of the wave setup is examined by using a hydraulic model and numerical computation.

Typhoon 7010 was generated about 300 km northeast of the Saipan Islands on August 19 and

landed on the Shikoku Island on August 21. At one time, the minimum depression became 910 mb

and the maximum wind speed exceeded 55 m/sec near its center. The typhoon caused the storm

surge along the coast of Tosa Bay as shown in Figure 1 and brought heavy damage by flooding

in Kochi city. The sea level became high at the foot of Tosa Bay and the anomaly was high

at those bays where the direction of the mouth is eastward, such as Katsurahama at Kochi

harbour or Usa at Uranouchi Bay. The anomaly at Katsurahama was extremely different from

the previously observed data as shown in Figure 2. The magnitude of the typhoon was not

extremely large but its route was different from others. It approached from the southeast.

At Kochi city, the maximum wind speed was 29.2 m/sec and the direction was east. Waves

also reached their maximum value of 6.6 m and a period of 12.2 sec at 10 a.m. The measurements were taken 260 m offshore of a breakwater at Kochi harbour at a depth of 12.5 by a

pressure-type wave gauge.

2. EXPERIMENTAL METHOD

The wave setup at the mouth of Kochi harbour was investigated using a 1/250 scale model.

The model was 3 km wide and 3 km long in prototype as shown in Figure 3. Waves propagate in

widths of 1.5 km between two guide walls.

IX9
Applying the Froude scale law to the model experiment, the length scale is 1/250 and the

time scale is 1/15.8. Waves were generated by a pendulum-type wave generator and measured

with resistance-type wave gauges at 30 points. The wave setup was also measured with pitot

tubes and pipes settled at the bottom. For experimental conditions, we selected three wave

heights (2, 3, 4 cm), three directions (S22.5°E, S45°E, S67°E), and two period conditions

(0.8, 1.0 sec).

3. EXPERIMENTAL RESULTS r.

,

Incident waves break offshore and regenerate new waves. Diffraction occurs behind the

breakwater. In Figure 4 wave height distribution is shown as an example. The height is

almost comparable to the water depth and is small behind the breakwater.

The distribution of the wave setup is shown in Figure 5. In the case of S45°E incident

waves, mean sea level rises high at the shore line and behind the breakwater. In the case

of S67.5°E incident waves, waves approach obliquely the shoreline and generate longshore

currents. The current is blocked by the breakwater and water piles up. Then wave setup

:(• occurs in both the onshore and alongshore directions.

The effects of tidal level, wave height, period, and direction on the wave setup were

investigated experimentally. First, the effect of tide is negligible because the wave

height distribution is not influenced by the tidal level. The wave period is related to the

breaking conditions. However, in this case, because of a wide breaker zone, the change of

wave period brings only a small change of the width. Therefore, the period is not an important factor. Incident wave height and direction are strongly related to the wave setup.

The relation between wave height and setup height is shown in Figure 6. The setup increases

as the incident wave height increases.

In this experiment, waves were bounded by two guide walls. Thus, the formation of longshore current was not fully developed. Since it was expected that the distribution of wave

setup height would be influenced by the side wall, numerical computations were carried out to

make the effect clear.

4. NUMERICAL ANALYSIS

As a method to predict wave setup and current, Longuet-Higgins proposed an equation

based on radiation stress concept. Basic equations are . IX- 10
i!L+ __JM 1^.+ N 3M ^ __M_ 9£ ^ + n) lil- -kUo —^ + LV^ M

3t 3(h + n) 3x h + n 3y h + n 3y 3x ° ° h + n

-J_(!!iHL+!!2Ey) , (1)

p 3x 3y

3t h + n 3x h + n 3x h + n 3y 3y ° ° h + n

+ !!zy)

,

, (2)

p 3x 3y ' :

• .

'

3n + 3M ^. 3N ^ Q 3t 3x 3y

where the x and y axes are taken horizontally, t is the time, M and N represent the flux

flows in the x and y directions respectively, n the water surface elevation, h the water

depth, g the acceleration of gravity, the friction coefficient at the bottom, L the horizontal eddy viscosity, the amplitude of the orbital velocity at the bottom induced by wave

action, and S^^, S^,, and S the radiation stress tensors. XX xy yy Numerical computation was carried out using the ADI method [1, 2]. In order to find the

distribution of radiation stress tensors, the wave height and the wave direction angle must

be given beforehand. Equations suggested by Noda [3] were used. They are:

cose ^ + sine 11 = J_ (sine l£_- cose il) "

• (4)

3x 3y C 3x 3y

-L (EC„ cose) + J- (EC„sine) = 0

~

' (5)

3t 3x 6 gy S

where 6 is the wave direction angle, C the wave velocity, Cg the group velocity, and E the

wave energy.

In a breaker zone, the following equation is used, •

'

H = 0.7 h (6)

where H is the wave height in the breaking zone, and h the water depth. •

Numerical computation was carried out in the case of a S67.5°E incident wave. The

domains were taken as shown in Figure 7 and covered with a mesh of discrete points. The

spacing was uniform and 75 m in prototype. The friction coefficient and the horizontal

eddy viscosity were taken as 0.0064 and 20 cm /sec respectively.

IX-11
5. NUMERICAL RESULTS

Computations were carried out to make clear the effect of side walls in the model.

Two cases were analysed. One has the same domain as in the model and the other has a wider

domain in the direction of longshore as shown in Figure 7.

A comparison of the two cases are shown in Figure 8. The wave setup in the narrow one

is about 10% lower than that of wide one. Though the difference is not large, this demonstrates that the longshore component of radiation stress brings the wave setup at the harbor

mouth.

The experimental results shown in Figure 5 are a little lower than the computed one

shown in Figure 8. There are several reasons that produce the difference. For simplicity,

in the calculation, equation (6) was used. In the experiment, the coefficient varies from

0.5 to 0.7 according to breaking condition and the formation of regenerated waves. The topography is also different in each case. In the calculation the change of topography is averaged over one mesh size.

Current patterns are shown in Figures 9 and 10. Longshore currents are generated by the

obliquely incident waves and they turn offshore near the breakwater. In the narrow domain

case as shown in Figure 9, the longshore current is suppressed by the side wall. This produces a low wave setup. While the current patterns were not measured in the experiment, they

were different from those in the calculated ones. In the experiment, the offshore current

was predominant at both sides of the breakwater. On the other hand, in the calculation circulation developed at the tip of the breakwater. The mesh pattern seems to be inappropriate.

6. CONCLUSIONS

Experiments and numerical analysis were carried out to better account for the abnormal

rise of sea level which was caused by typhoon 7010 at Kochi harbor. Experimental results

show that the wave setup becomes especially high when the waves approach the shoreline

obliquely. The longshore component of radiation stress, together with the onshore component,

play an important role in the formation of water piling up near a breakwater.

Numerical calculations were made to assess the effect of side walls in the model. When

the width of side walls was extended, the wave setup becomes higher. Actual current patterns

IX- 12
and setup heights do not agree well with those of the experiments. It is important to make

accurate estimates of wave heights in a breaker zone. In a complex topography the arrangement of the experimental mesh pattern is also important.

REFERENCES

[1] Hashimoto, H. and Uda, T. , "Method for Numerical Computation of Nearshore Currents and

its Application," 21st Coastal Engineering Conference in Japan, pp. 355-361, 1974.

[2] Hashimoto, H. and Uda, T. , "Simulation of Nearshore Currents," 22nd Hydraulics Conference in Japan, pp. 147-154, 1978.

[3] Noda, E. K. , Sonu, C J., Rupert, V. C. and Collions, J. I., "Nearshore Circulation

Under Sea Breeze Conditions and Wave-Current Interaction in the Surf Zone, TETRA TECH.

Report, TC-P-72-149-4, 1974.

IX- 13
N

(0.60)

Pig. 1 Maximum sea level above T.P. (Tokyo Peil) and anomaly along the

coast of Tosa Bay caused by Typhoon 7010 (imit: m)

Fig. 2 Return periods of the maximum tidal level at Katsurahama (From

1950 to 1970)

IX- 14
IX- 15
Mod. Pro. H. 4.2 cm 10.5 m

T 1,0 s 15.8 s

? ^ ^ Mod. Pro. Ho 3.8cm 9.5 m

T 1.0 s 15.8 s

Fig. 5 Wave setup distribution

P M

cm mm

10©

50 2

0 0

S22.5E

P M

cm mm

100 4

50 2

0 I 2 3 4 5 cm

0 5 10 m

H.

0 0

S45E

P M

cm mm

100 4

50 2

0 I 2 3 4 5 cm

0 5 10 m

H.

0 0

S67.5E

0 I 2 3 4 5 cm

0 5 10 m

Fig. 6 Relation between wave height and wave setup height

IX- 16
IX- 17
7-

Fig. 9 Current pattern (narrow domain)

8(m) 9

'

^ ^

' M / ^ "

* V V

^ \ \

t *

Fig. 10 Current pattern (wide domain)

IX- 18
SPECIFICATION AND PREDICTION OF SURFACE WIND FORCING

FOR OCEAN CURRENT AND STORM SURGE MODELS

Celso S. Barrlentos

Kurt W. Hess

Techniques Development Laboratory

Systems Development Office

National Weather Service

National Oceanic and Atmospheric Administration

ABSTRACT

Forecast methods to predict movements of oil spills in the ocean are being developed in

the Techniques Development Laboratory of the National Weather Service (NWS) . An operational

model for oil movement forecast is being implemented in NWS. The model will be available for

routine use in the event of oil spills and for assessment studies of probable impacts of oil

spills.

The most important component in the movement of an oil spill is due to the surface wind

forcing. Surface wind stress acts on the spilled oil in two ways; (1) to generate ocean

surface currents, and (2) to drag directly the oil on the surface. We examined different

boundary layer wind formulations and assessed the resulting surface currents response.

A dynamical storm surge model has been developed in NWS and is being used in real time

forecasting of surges when a hurricane is approaching land. The model has proved useful in

routine use during the last 10 years. Recently, a model has been developed that forecast

surges in bays and estuaries.

Storm surge is generated by the action of the wind and low atmospheric pressure in the

storm (inverted barometer effect). The wind forcing is the dominant factor in surge generation. The spatial distribution of the wind in a storm or typhoon determines the resulting

characteristics of the surge, such as, maximum surge and location on the coast, extent of the

coastline affected by the surge and the height variation of the surge along the coast. We

tested different storm wind models and determined the response of the storm surge model.

KEYWORDS: Ocean current; oil spill trajectory; wind forcing; hurricane; storm surge; wind

models

.

IX-19
INTRODUCTION

The National Weather Service (NWS), NCAA, has been involved in monitoring and prediction

in the marine environment. The Techniques Development Laboratory, NWS, is developing forecast methods to predict movement of oil spills in the ocean and storm surges in the coastal

regions. We are implementing an operational model to forecast oil spill movements. The

model will be available for routine use in the event of oil spills and for assessment studies

of probable impacts of oil spills.

NWS has an operational model to predict storm surges in the coastal regions. The storm

surge model has been working well during the last 10 years and has been tested in many actual

storm situations. Recently, a dynamical model has been developed to forecast storm surges

and overland flooding in bays and estuaries.

In predicting movements of oil spill and storm surges, surface wind forcing is the most

important element. The wind generates the surface current and drags the oil, while storm

surge is due to the wind stress and to the low atmospheric pressure in the storm center.

Specification and prediction of surface wind forcing are therefore most important in oil

spill movement and storm surge.

The development of ocean current and storm surge models are two independent studies,

but wind is common to both. The range of wind speed is different for the two models. Storm

surge is associated with strong winds, while for oil spill trajectory, the wind speed is for

the more normal range including the strong winds of course.

In our research in NWS, we have to consider that results will have to be implemented

for operational use. We have to design our forecast techniques so that the input data is

readily available either from other forecast models or provided by the forecasters. We

require, therefore, that input must be simple or minimally required. We also design that

model output is also simple. This way the users can easily interpret the results.

WIND FORCING FOR OCEAN CURRENT MODEL ,

NWS routinely runs several multi-level atmospheric models on an operational basis.

The LFM II is the most likely candidate for use in the oil spill movement model because

of its finer grid size with a large area of coverage.

IX-20
The predicted winds at the earth's surface, however, are based on simple formulations,

preventing their direct use in the oil spill model. However, by resorting to models of the

planetary boundary layer (PBL), we can get more accurate forecasts of surface winds and wind

stresses. We do this by first calculating the surface geostrophic wind from load horizontal

temperature and pressure gradients. The actual wind, which incorporates the effects of

turbulent friction and the earth's rotation in the lower kilometer of the atmosphere, is

assumed to match the geostrophic wind above the PBL. Three models of the PBL tested are

similar to each other in that a logarithmic layer near the surface (where velocity increases

semi-logarithmically but maintains constant direction) is assumed, because fluxes of momentum

and heat are easier to approximate. By semi-logarithmically we mean proportional to the sum

of (a) the log of the altitude plus (b) another function which accounts for variations due to

atmospheric stability. -,

Above the log-layer, in the upper layer of the PBL, the assumptions vary. For example,

the marine boundary layer model (Cardone, 1978) takes a constant eddy viscosity above the

log-layer. There the velocity deviation from geostrophic is described by the Ekman spiral,

which is then matched with winds in the lower layer. By contrast, the Rossby Number similarity model proposed by researchers at the University of Texas (Wagner and Fredricksen, 1979)

uses empirical functions for conditions above the log-layer to give the problem closure.

These functions of stability and horizontal temperature gradient are used to get the surface

angle between, and ratio of, friction velocity and geostrophic wind. The surface Rossby

number employs variable roughness height (the roughness follows a quadratic law on the friction velocity for all three models). Both these models assume steady-state conditions.

Finally, the one-dimensional dynamic model, uses an analytic expression for the eddy

viscosity profile to compute dynamic upper-layer fluxes, and hence the velocity and temperature profiles, on a finite-difference grid. A statistical evaluation, using observed winds

at the Argo Merchant, was necessary to compare each method.

We are fortunate to have wind data at the wreck site, from NOAA (1977), to test each

of these PBL schemes. Observed pressure distributions, each 6 hours, allows surface geostrophic winds to be computed. Similarly, the observed surface temperature fields give the

thermal winds. When neutral atmospheric conditions were assumed, each PBL model was run for

the interval, 17 days, for which observations were available. The first two models were run

as a series of steady state conditions, based on observations each 6 hours. The dynamic

IX-21
model ran as one continuous case, with a 1-hr timestep, driven by linear interpolations on

the same observations. The statistics (Table 1) show that the Rossby number similarity

scheme gives the best results. The empirical functions apparently represent the upper portion of the PBL better than the Ekman spiral does. The dynamic method suffers because of

the large inertial oscillations introduced during the simulation. ' ' i?-'-'-^ -J- . ••

WIND MODEL FOR STORM SURGE

In the storm surge model, we derive the wind in the following manner: We formulated a

storm model. In the Ninth Proceedings of our UJNR Panel, Fujiwhara and Kurashige (1978) have

presented distribution models of pressure and wind for stationary typhoons. We formulated a

similar model. We assumed the pressure profile is similar. That is, in nondimensional plot,

all the pressure distribution will fall in one curve. '

'

We begin with the stationary symmetric storm. The pressure, wind speed, and wind direction for the stationary symmetric storm are not quite arbitrarily chosen like most storm

surge models. The pressure and direction are determined from simplified equations of motion

by balancing of forces. The equations are adapted from Myers and Malkin (1961):

2 J_ dp ^ ^s ^

_ V dV . . (1)

Pa dr Sin $ dr .: , i

.

.

-i- dp

cos $ = fV + COS $ - V^-^ sin $ + ^ ip ^ Pa dr r dr . -

Here Pa is the surface atmospheric density considered constant, p(r) is the pressure,

$(r) the inflow angle or angle toward the storm center, and V(r) is the wind speed; all are

functions of r, the distance from the storm center. The terms and K^^ are empirically

determined coefficients, representing stress coefficients in the directions opposite and to

the right of the wind respectively. These stresses are given by the coefficient times the

square of the wind speed, and f denotes the coriolis parameter.

Equations (1) and (2) are to be solved for p(r) and 'l'(r), but the form of wind speed

profile, V(r), must be known first. We examined several wind profiles and has adapted the

form . ,

V(r) = 2Rr (3)

IX-22
The term, R, the "radius of max winds," is the distance from the center at which V(r)

is greatest, and V^. = V(R) is the value of that maximum wind speed. Of course, similar

other profiles maybe chosen. Equation (3) has the property of increasing wind speed from

r = 0 to a maximum at r = R, then decreasing back to zero at r very large.

The procedure for solving (1) and (2) for p(r) and $(r), with a given wind profile has

been a Runge-Kutta type integration scheme applied to the transformation of (1) and (2).

The transformation reduced problems of instability associated with a straightforward application of the Runga-Kutta method. This procedure was stable for the wind profiles given by

(3) and for commonly encountered values of V^., R, and latitude, but not for extreme values.

That is, it is only marginally stable.

In our storm surge model, we add the vector Ug^j^ to the stationary storm wind vector

as a gross correction for storm motion U^. The form of the correction vector Ug^^ is

U = Rr (4) sm T~7—

R"^ r'^

where Ug the storm directed translation speed and the other symbols were previously defined.

Only storms moving with uniform rectilinear motion are considered. Accordingly, the wind

stress used in the numerical storm surge model includes the storm motion effects in a gross

way

.

We need specification and prediction of surface wind stresses in our storm surge model

in great detail. This data is not available either in observations or forecasts. Our storm

model approach provides for this information. .

We investigated other storm models and tested the response of our storm surge model.

We found the best agreement with the observed surge data with the storm model we adapted.

Other storm models were developed for different applications. A model represents the purpose best for which it was developed. For example, the Standard Project Hurricane (SPH)

model in the U.S. was developed for the use of the U.S. Army Corps of Engineers for the

design of structures. SPH represents a wide range of averaged situations and very good for

its purpose; but not for forecasting storm surge. We believe, our storm model for SPLASH is

excellent for our purpose.

IX-23
REFERENCES

[1] Cardone , V. J., "Specification and Prediction of the Vector Wind on the United States

Continental Shelf for Application ot an Oil Slick Trajectory Forecast Program." Final

Report T-35430, Institute of Marine and Atmospheric Sciences, The City College, City

University of New York, 63 pp and Appendices.

[2] Fujiwhara, S. and Kurashige, K. , "Distribution Models of Pressure and Wind over Stationary Typhoon Fields." Proceedings of the Ninth Joint UJNR Panel Conference on Wind and Seismic Effects," NBS Special Publication 523 , pp. I-l to I-IO, 1978.

[3] Myers, V. A. and Malkin, W., "Some Properties of Hurricane Wind Fields as Deduced from

Trajectories." National Hurricane Research Report No. 49, U.S. Weather Bureau,

Washington, D.C., 45 pp, 1961.

[4] National Oceanic and Atmospheric Administration, "The Argo Merchant Oil Spill: A

Preliminary Scientific Report." Grose, P. L. and Matson, J. S., eds. , 133 pp, 1977.

[5] Wagner, N. K. and Fredricksen, E. J., "Similarity Theory Applied to Surface Winds and

Oil Slick Transport." Preprints Workshop on the Physical Behavior of Oil in the

Marine Environment, May 1979, Princeton, N.J. sponsored by Princeton University and

National Weather Service.

Table 1. Conparison of errors (obser~v/ed wind minus conputed wind) from three

planetary boundary models of the Argo Merchant winds.

RMS Error WG Error

Mod el

Speed (I) Direction (°) Speed [sj Direction (°)

Marine 3.L. 4.21

Rossby No. 3.81

Similarity

One-Dimensional 5.09

Dynamic

41.4

37.6

53.5

1.03

-0.10

2.15

11.93

6.81

-0.44

IX-24
A DYNAMIC MODEL TO PREDICT STORM SURGES

AND OVERLAND FLOODING IN BAYS AND ESTUARIES

Chester P. Jelesnianski

Celso S. Barrientos

Jye Chen

Techniques Development Laboratory

Systems Development Office

National Weather Service

National Oceanic and Atmospheric Administration

ABSTRACT

In the United States, most of the damages caused by hurricanes are attributed to storm

surges. NOAA provides forecast and warning services to prevent loss of lives and property.

A dynamical model was developed to forecast storm surges in bays and estuaries. The development of the model and the operational application for forecasting will be described.

KEYWORDS: Storm surge; hurricane; overland flooding; forecast and warning.

IX-25
INTRODUCTION

The National Weather Service (NWS), NOAA provides forecast and warning services for

possible natural disasters. A hurricane can affect a broad area along the 3500 miles of

coast along the Gulf and East coasts of the U.S. Most of the damages, which is considerable,

caused by hurricanes are due to storm surges. Damages are caused by the pounding action of

storm surge flood.

We define storm surge as the abnormal high water level due to meteorological forces.

This is to differentiate with the high water due to tidal forces. Storm surge is due to the

action of the wind and the low atmospheric pressure (inverted barometer effect). In hurricanes, the wind forcing and low pressure produce a rotating mound of water. When this mound

of water reaches the coastal area, it will feel the bottom of the continential shelf and the

storm surge is the result.

NWS has an operational model to predict outer coast hurricane storm surge since 1969.

This model, SPLASH - Special Program to List Amplitudes of Surges from Hurricanes, has been

used in several actual hurricanes and has been proved useful. SPLASH was also used to compute surges for historical hurricanes for flood insurance applications. Last year, during

our Tenth Panel Conference, I discussed storm surge in general including SPLASH.

The main limitation in our SPLASH model is that we assumed vertical wall along the

coast. Therefore, SPLASH is not able to forecast overland flooding or overtopping of

barrier islands. The limitations leads us to the development of a new dynamic surge model,

SLOSH - Sea, Lake, and Overland Surges from Hurricanes. SHOSH is capable of predicting

storm surges and overland flooding in bays and estuaries.

THE SLOSH MODEL

The basic dynamics of SLOSH is the same with SPLASH. We used a vertically integrated

equation of motion and compute surges generated by tropical storm via finite differencing.

The model is solved numerically.

We call the input simulated geographic areas to the model as the basin. The geographical area of the basins vary considerably, but usually extends about 100 miles seaward from

the coast, shoreward into high terrain, and with about 200 miles of coastline. For such a

large area of the basin, there is no requirement to couple a bay model onto an open coast

IX-26
model. Input boundary values are not required, except for an initial and quiescent sea level

before a storm affects a basin region, that is, the initial sea height unaffected by a storm

or astronomical tide. The sea level datum used is "National Geodetic Vertical Datum (NGVD)."

Published bathymetry on marine charts for seas and inland water bodies are corrected to NGVD.

Land topography and vertical barriers such as roads, levees, spoil banks, dunes, etc. , are

simulated for each basin.

The model allows overtopping vertical barriers, inland inundation from sea or bay waters,

and flow through channels. The astronomical tide, river discharge, and rain are not considered; under the assumption, their effects are small compared to a transient storm surge. The

astronomical tide is superimposed on the surge after computations, assuming negligible interaction with the surge. The model allows for inland inundation; water can inundate across lowlying river basins.

The equations of motion of the model are identical to an earlier open coast surge model,

SPLASH, developed by NWS. The one exception is the finite amplitude effect which can be

important in shallow bay waters and especially so with inland inundation. A time-history

bottom stress is used rather than a standard Chezy or Manning type formulation. No local

calibration or tuning are effected to locally set values for undetermined coefficients in the

equations of motion; instead, best fit type constants for the coefficients were determined

in a universal sense from a multitide of basins, historical storms, and surge observations

using the coastal surge model SPLASH.

The storm model is the same as used by SPLASH. That is, a simple balance of surface

forces to compute wind and inflow angle across isobars; the inflow angle is dynamically computed and is not empirically derived. One exception in the SLOSH model is higher surface

friction across inland bodies of water, but not the sea. The friction is not a function of

a particular bay, and is applied the same way for all bays.

We use very simple input meteorological data. This consists of two storm parameters,

central pressure of the storm and storm size (distance from storm center to maximum winds),

as well as storm track; the two storm parameters, and speed of storm along track, can be

variant with time. Maximum wind is not an input parameter in the storm model. A useful

property of the storm model allows high winds but small inflow angles across isobars for

weak friction, and otherwise for strong friction. The feedback between winds and inflow

angles in the storm model, with alternate friction values is a compensating effect for surge

IX-27
generation. Therefore, high accuracy for wind speed is not a requirement, however, central

pressure, storm size, speed along track, and track, orientation relative to a geographical

area are important parameters and these must be accurate.

Coarse and find grids are used in the model for numerical computations. If a bay is

large and uncomplicated, then a large, invariant, grid size (four miles grid spacing) is

used. If a bay is small or complicated, then an expanding grid scheme (1/2 miles in bay

regions to 4 miles in deep sea waters) is used. The expanding grid is a polar coordinate

system, transform conformally onto a Clark Ellipsoid. , • , , ., , .

The SLOSH model, particularly the storm model, has been developed with the data for

Lake Okeechobee, Florida during a hurricane in 1949. The model has been tested with real

surge data for several historical storms affecting Lake Okeechobee and Tampa Bay, Florida and

Lake Pontchartrain, Louisiana. Most of the results appear realistic, that is, usually within

20% of observed surge values. SLOSH was adapted and is used for real time forecasting of

surges for the above three areas. We are now in the development stage for a total of 20

Bays and estuaries. : i;

J

IX-28
TOPICS ON TSUNAMI PROTECTION ALONG THE PORT AREAS IN JAPAN

Yoshlmi Goda

Marine Hydrodynamics Division

Port and Harbour Research Institute

ABSTRACT

The major Japanese tsunamis since the 15th century and a general description of counter

measures against them are briefly introduced. Particular reference is made on tsunami break

waters used to protect port areas. The performance of the Ofunato tsunami breakwater in the

instance of the Tokachi-oki Earthquake Tsunami in 1968 is discussed, and a new plan for a

Kamaishi tsunami breakwater is introduced. Also mentioned are several tsunami prediction

plans in progress.

KEYWORDS: Disaster planning; seawall protection; tsunami breakwaters; tsunami prediction.

IX- 2 9
: 1. MAJOR TSUNAMIS IN JAPAN AND COUNTERMEASURES

Tsunami records in Japan date back to the seventh century. Suffering from tsunamis has

increased with the passage of the centuries because the population greatly increased. Table

1 lists the major tsunamis since the 15th century. Names of areas are indicated in Figure 1.

Tsunami damage has been severe along the coasts of the Sanriku Area (from Aomori to Miyagi

Prefectures), the Tokaido Area (from Chiba to Mihe Prefectures), and the Nankaido Area (from

Mihe to Kochi Prefectures). Other areas including the Japanese Sea coast have also suffered

injury from tsunamis though to a lesser extent.

Countermeasures against tsunami attacks adopted in Japan are:

1) The establishment of a tsunami -alert system and evacuation drills,

2) construction of refuge roads,

, .1'

I ... 3) relocation of houses to neighboring highlands,

4) planting counter-tsunami groves and forests, ' - '

'

5) construction of frontage sheds and other facilities in port sections as solid,

permanent structures to shelter the resident areas behind them,

6) construction of seawalls, and

7) construction of tsunami breakwaters.

The popularity of television and portable radios nowadays greatly assist spreading the

warnings of possible tsunamis which are issued by the Japanese Meteorological Agency. Relocation of houses to highlands is difficult due to the scarcity of housing lots. Because

people continue to live along shoreside areas even with the knowledge of tsunami danger, seawalls against tsunamis have been built for centuries and continue to be constructed at many

places. Some seawalls have a crest height of 16 meters above the mean sea level. Seawalls

require movable gates at their junctions with roads. Proper maintenance of these gates is

not easy

.

2. TSUNAMI BREAKWATERS

Construction of seawalls along the port area waterline are a hindrance and cause inconvenience to daily port operation. Reduction of tsunami height by means of offshore breakwaters provide a solution for the protection of some port areas against tsunamis. The ports

of Ofunato, Onagawa, and a few others have been provided with such breakwaters, and the

IX-30
Kamaishi Port will soon have one. These breakwaters restrict the entrance of tsunami waves

into harbours by narrowing the entrances. The possibility of harbour resonance does not

materialize because of the energy loss at the entrance by wakes and eddies.

The Ofunato tsunami breakwater shown in Figure 2 was built in the period from 1963 to

1967 after the suffering of the lives of 53 people and 8 billion yen by the Chilean Earthquake Tsunami in 1960. The City of Ofunato has the population of about forty thousand and a

port cargo of 5 million tons annually. The breakwater site is 38 meters deep (maximum) and

738 meters wide. The section was reduced to the width of 200 meters and the depth of 16.3

meters below the low water level to permit the navigation of 100,000 DWT tankers. The effect

of this breakwater against monocromatic tsunami of sinusoidal wave form (amplitude of 0.5

meter) is exhibited in Figure 3 in the form of the resonance curve of Ofunato Bay with and

without the breakwater. The analysis was done by Ito [1, 2] with the introduction of the

head loss at the entrance, which was taken as 1.5 times the velocity head.

The Tokachi-oki Earthquake Tsunami of May 16, 1968, provided an opportunity to demonstrate the effectiveness of this breakwater. The observed time histories of water level

variation due to the 1968 Tsunami are shown in Figure 4. These tsunami profiles were successfully reproduced in numerical computation by providing 30 harmonic components of the

tsunami waves recorded outside the breakwater and adjusted for the respective amplitude

amplification there. Computation was extended to the case when the breakwater had not been

constructed. Figure 5 is the predicted time history of water level variation at the innermost part of Ofunato Bay. It is clear that the breakwater reduced the tsunami height by one

half. Figure 6 is the comparison of highest water level elevation with and without the

breakwater at the arrival of the first crest of the 1968 Tsunami.

A new tsunami breakwater is planned at Kamaishi Port as shown in Figure 7. The City

of Kamaishi has a population of about seventy thousand, and the port handles 4.6 million

tons of cargo annually. The breakwater site is 60 meters deep (maximum) and about 2100

meters wide. The breakwater will have a central opening 300 meters wide and 19 meters deep

below the low water level as well as two side openings each 50 meters wide. The layout of

the breakwater is the best among several alternatives for tsunami reduction capability and

minimization of eddy currents around entrances.

The breakwater is expected to reduce tsunami heights as listed in Table 2 according to

numerical analysis and hydraulic model tests, both of which were commissioned by the Second

lX-31
District Port Construction Bureau, Ministry of Transport [3] ; the testing was carried out

by Professor Iwasaki of Tohoku University. The construction work will begin in 1980 and be

completed in ten years with an estimated cost of 53 billion yen. i3. TSUNAMI PREDICTION SCHEMES

Sanriku and Tokaido are the areas most liable to tsunami damage. To adequately plan for

tsunami protection at the ports in these areas, a tsunami prediction program has been undertaken by the Ministry of Transport with the cooperation of several professors and researchers.

The program consists of:

1) An assumption of the spatial distribution of sea bottom dislocation in the source

area,

'

2) the computation of tsunami propagation from the source to the coastline, . ^ :

3) the computation of tsunami runup on land topography, and ..

.'.

4) the evaluation of tsunami protection work against predicted tsunamis.

The project of the Kamaishi tsunami breakwater is a result of the tsunami prediction

scheme along the Sanriku coast. Another scheme is being carried out for the coast of the

Tokaido area against a possible Tokai Earthquake Tsunami in the near future.

4. CONCLUDING REMARKS

The protection of coastal areas from powerful tsunamis is not an easy task, because

tsunami runup is quite high and the protection work is quite costly; especially when the

frequency of extreme tsunamis is taken into consideration. Nevertheless, the severity of

tsunami damage answers any doubt about the necessity for tsunami protection. Port areas with

dense populations and busy daily operation are one of the most difficult places to provide

effective tsunami protection. Though the tsunami breakwater is a good solution, we need to

explore other solutions. Numerical analysis, as well as hydraulic model tests, will further

help us find such solutions for tsunami protection.

REFERENCES

[1] Ito, Y., "On the Effect of Tsunami-Breakwater," Coastal Engineering in Japan, Vol. 13,

Japan Society of Civil Engineers, 1970, pp. 89-102.

[2] Ito, Y., "Head Loss at Tsunami-Breakwater Opening," Proceedings of the 12th Conference

on Coastal Engineering, Washington, D.C, 1970.

IX-32
Yokohama Investigation and Design Office, Second District Port Construction Bureau,

Ministry of Transport, "On Tsunami Protection Project in Kamaishi Bay," Reference

No. 52-13, 1978 (in Japanese).

Table 1 Major Tsunamis in Japan

Name

Meio

Tsunami

Keich 9th Tsunami

Keicho 16th Tsunami

HoeiTsunami

Meiwa Tsunami

Tempo

Tsunami

Ansei Tsunami

Meiji 29th Tsunami

Showa 8th Tsunami

Chilean

Tsunami

Date

1498- 9-20

1605- 1-31

1611-12- 2

1707-10-28

1771- 4-24

1843- 4-25

1854-12-23

1854-12-24

1896- 6-15

1933- 3- 3

1960- 5-24

Earthquake Area Most Magnitude Suffered

M = 8.6

M = 7.9

M = 8.1

M = 8.4

M = 7.4

M = 8.4

M = 8.4

M = 8.4

M = 7.9

M = 8.1

M = 8.75

Tokaido

Nankaido

Tokaido

Sanriku Hokkaido

Nankaido

Okinawa

Hokkaido

Tokaido

Nankaido

Sanriku

Sanriku

All the Pacific coast of Japan

Max. Runup Height

over 10 m

over 10 m

over 30 m

over 25 m

85 m

no record

over 20 m

30 m

29 m

8 m

Nos. of Persons

Ki 1 led or Missing Remarks

several tens of Building of Great Buddha in thousands

over 5,000

over 5,000

over 10,000

12,000

smal 1

about 5,000

27,122

3,644

123

Kamakura was washed away.

Miyako island was washed

over.

Area was little inhabited.

Russian warship "Diana" was damaged at Shimoda Port.

Tsunami period was some 70 minutes.

Table 2 Tsunami Height Reduction by Kamaishi Breakwater

- ratio of tsunami heights with and without breakwater

Tsunami period (minutes) 10 15 16 20 30 45

Ratio by computation 0.61 - 0.37 0.39 - 0.97

Ratio by model test 0.69 0.42 0.32 0.21 0.39 0.33

IX-33
Fig. 1 Location Map of Japan

lx-34
0

—•— with Breakwater

without Breakwater

1 /

i

/

/

/ ,1 ^\1 \ \

f

1 l\ /

J

1 , 1

0 10 20 30 4 0 50 60 70 (mm) tsunami period

Fig. 3 Amplitude of tsunami waves at Ofunato Harbour

for sinusoidal waves with the amplitude of 0.5 m (after Ito' )

IX- 3 5
time

Fig. 5 Predicted water level variations at the immentiost point

of Ofunato Bay with and without the breakwater

(after Ito^ )

IX- 3 6
(A) (B)

Fiq. 6 Distribution of water level in centimeters

1)2)

at the first peak of the 1968 Tsunami (after Ito ' )

Fig. 7 Map of Kamaishi Bay

ix-37

SEISMIC RISK MAPS (MAXIMUM ACCELERATION AND MAXIMUM PARTICLE VELOCITY)

IN THE SOUTHEAST ASIAN COUNTRIES OF THE PHILIPPINES, INDONESIA, AND INDO-CHINA

Sadaiku Hattori

International Institute of Seismology and Earthquake Engineering

Building Research Institute

.. .. ABSTRACT

Seismic risk maps in the Southeast Asian Countries of the Philippines, Indonesia, and

Indo-China were made using existing seismic data, attenuation models, and the method of

extreme value fitting. The maps consist of the following two kinds: (i) The maximum particle velocity (kine) on the base rock, and (ii) The maximum acceleration (gal) on the

ground. The return periods of these maps are 50, 100 and 200 years, respectively.

This paper is a modification of the report which the author is going to present at the

International Conference on Engineering for Protection from Natural Hazard, which was held

in the Conference Center of the Asian Institute of Technology in Thailand January 7-10, 1980.

KEYWORDS: Disaster prediction; earthquake motions; hypocenters; seismic risk maps; seismic

zoning

.
1. INTRODUCTION

Needless to say, accurate earthquake prediction techniques are extremely valuable. However, from the standpoint of preventing earthquake disaster, it is not enough only to predict

hypocenters, magnitudes, and occurrence times of future earthquakes. It is also important to

estimate earthquake motions for earthquakes which may occur in the future and to forecast the

disasters they might cause. Furthermore, it is necessary to consider counterplans to prevent

them. Ideally speaking, the occurrence time of earthquakes (tn), earthquake motions on the

base rock and on the ground (f(tsn)gp, f(tsn)2p), and the ground characteristics (Gp) should

be exactly estimated (Figure 1).

However, it is impossible at present to estimate all of the above related terms with

great accuracy. Much research has been conducted in approximating the above ideal one. The

most typical approach in this research is to define the expectations of the maximum earthquake motions for a number of return periods by using (i) the seismic data of the past, (ii)

attenuation models, and (iii) a statistical method. There are the following two problems

which must be heeded in using this approach: (A) The return periods are used, instead of

denoting directly the occurrence time of earthquakes, and only the maximum earthquake motions

are given in place of defining concretely the earthquake motions. (B) When seismic data are

not reliable or the data period is too short, it is meaningless to define the expectations

because of the large errors which will be caused.

Much research of this kind has been performed for the vicinity of Japan. [1] This

research is concerned with only the regional characteristics of the seismicity, and no consideration is given to the ground characteristics or to the temporal variations of the seismicity.

Hattori [2] showed approximately the ground characteristics of the long period range

(2-5 sec), in the whole vicinity of Japan, using the seismic data of the maximum amplitudes

and the period observed by strong motion seismographs at JMA stations. He provided a definition of base rock and proposed a new seismic risk map. The map is based on the maximum

earthquake motions which are estimated from the spatial characteristics of the seismicity,

and from the ground characteristics [3]. Further, Hattori [4] defined the temporal variations of the seismicity of the whole vicinity of Japan using the accumulative energy curve

X-2
and refined the seismic risk map. This is a solution for the above described problem (A),

which means that a term indicating occurrence times of earthquakes was added to the seismic

risk.

There are some examples of seismic risk analysis for a few regions in the world [5, 6,

7, 8, 9]. They, also, concerned the maximum earthquake motions based on only the spatial

characteristics of the seismicity. It goes without saying that many have recognized the

necessity for studies of the ground characteristics and the temporal characteristics of the

seismicity, but there seem to be none which have actually adopted such an approach.

The analyses of Oliveira [9, 10] are as follows: A source area of earthquake, and a

point where the maximum earthquake motions are to be estimated were supposed to be S and P.

In general, there are plural source areas, and P is regarded as not one point but an area.

Furthermore, Oliveira supposed that occurrences of earthquake in the source areas depend on

the Poisson process and Gutenberg-Richter 's law. The maximum earthquake motion at an arbitrary point for an earthquake was calculated by an attenuation model. The expectations of

the maximum earthquake motion for some return periods were estimated by the method of

extreme value fitting.

As mentioned above, seismic risk maps have been made for some regions in the world.

There is even a case where the ground characteristics, the temporal variations of the seismicity, and the modifications of the seismic data were taken into consideration. [4] The

first step in this kind of research, however, is to make a map based on the expectations of

the maximum earthquake motions for some return periods being based on the spatial characteristics of the seismicity.

The working group E2, Ground Motion and Seismic Risk, the U.S .-Southeast Asia Symposium

on Engineering for Natural Hazards Protection recommended cooperation in six research areas

among the southeast Asian countries, Japan, and the United States. One of the six research

areas was the development of a Seismic Zoning Map for the Philippines, Indonesia, and other

Southeast Asian countries. This Seismic Zoning Map is a seismic risk map which is based only

on the spatial characteristics of the seismicity.

X-3
2. SEISMIC DATA . . , ..

The following data are available for areas concerning: world seismic data (magnetic

tape) which were compiled by the National Oceanic and Atmospheric Administration Environmental Data Service (NOAA). Figure 2 shows the temporal variations of seismic data in

Philippines area.

The seismic data of the NOAA magnetic tape include four kinds of magnitude, namely, (i)

the body wave magnitude M^, (ii) the surface wave magnitude Mg, (iii) the magnitude M, which

was determined by each observational organization, and (iv) what is termed the local magnitude M^. In order to estimate maximum earthquake motions, it is necessary to use a unified

magnitude. So, using all the data in NOAA magnetic tape, the relations among these magnitudes were researched (Figure 3). The results obtained are as follows: The correlation

coefficients are about 0.7 or more.

Mg = 1.081(t 0.014)M^j - 0.390(± 0.077) ; P = 0.699 (1)

M = 1.161(± 0.019)Mij - 0.662(± 0.104); P = 0.723 (2)

Ml = 0.862(±0.017)M^ + 0.565(±0.079); P = 0.739 (3)

M = 0.853(+ 0.015)Mg + 0.929(± 0.090) ; P = 0.847 (4)

Ml = 0.679(+ 0.041)Mg + 1 . 826(± 0 . 216) ; P = 0.751 ^ ^ (5)

Ml = 0.860(±0.024)M + 0.677(±0. 123) ; P = 0.913 :

' (6)

In this research, the magnitudes M^^, M, and Ml were transformed into Mg by Equations

1, 4, and 5, respectively.

The longer the data period, the better the data for statistical analysis. However, the

further back we go in time, the sparser the data becomes, especially for data with magnitudes

and epicenters defined. For Japan and China, areas in which the author had already made the

seismic risk maps [1, 11], we could obtain the seismic data for fairly long periods. For ,

other Asian countries, however, only the NOAA magnetic tape is available due to a lack of

historical seismic data.

Figure 4 shows the epicentral distribution in the Indonesia area. It is founded on the

data with the above-mentioned unification of magnitude.

X-4
3. ATTENUATION MODELS

Some of the attenuation models which have been proposed are shown in Table 1.

Oliveira [10] gave Equations 1, 2, and 3 (y maximum earthquake motion; M(m) magnitude;

R hypocentral distance) as attenuation models on firm ground. Further, he suggested another

one (Equation 4), which gave maximum particle velocity on firm ground in California. McGuire

[12] proposed equations similar to those of Oliveira (Equations 5, 6 and 7). Esteva and

Rosenblueth proposed Equation 8 for maximum acceleration on the ground [10].

Kanai [13] recognized that velocity spectra of earthquake motions on base rock are flat

in a certain period range, from the results of observations under the ground (depth 300 m)

of the Hitachi Mine. Further, he suggested (Equation 10) for the maximum particle velocity

on the base rock, using Tsuboi's equation which defines magnitudes.

Figures 5 (a, b) are shown to compare the attenuation models referred to above. In addition to these, many attenuation models have been proposed and recently, some models based on

the dislocation theory of the earthquake fault have also been suggested. ...

The author [1] made seismic risk maps in the vicinity of Japan using Kanai 's equation

(Equation 10 in Figure 5 (b)) so seismic risk maps for many areas may be compared directly.

It is reasonable that maximum earthquake motions on the ground are to be obtained by

considering the ground characteristics of each site for the input on the base rock. However,

the risk maps expressed by the maximum acceleration on the ground and disregard ground characteristics, are also useful in earthquake engineering. For this purpose, the attenuation

model (A) in Figure 5 (a) was used. This is the mean of the attenuation models by Oliveira

(Equation 1) and McGuire (Equation No. 5). It is very close to the one given by Katayama

(Equation 13). It can be seen from the figure that attenuation model (A) shows nearly the

same mean value as many other ones.

4. ESTIMATION OF MAXIMUM EARTHQUAKE MOTIONS

Using the probability that the variable x is equal to or larger than any of X-]^, ...

*^(x) = P(X]^ < X, X2 < X, < X), (7)

the return period T(X) and the reduced variable z are defined by the following.

X-5
T(x) = 1/(1 - $^(x)) (8)

z = -%{-)lr,'fn(x)} (9)

If the maximum earthquake motion at a certain point in the m-th unit period (for example, one

year) is assumed to be A^^^^^ ^ and the variable X^^ is defined by the equation:

the distribution of x may be approximated by Gumbel's third asymptotic distribution. The

third distribution is denoted as follows:

= exp{- (^L:L4^ > (11)

where the parameters (W, K, V) are the upper limit of maximum value, the shape parameter,

and the characteristic maximum value. The number of the unit period is put to be n. Furthermore, the reduced variable for the observation, and the theoretical one of the third

asymptotic distribution, are put to be y^^^ = ra/(n + 1) and Z^, respectively. The third

asymptotic distribution most fit for n values of x may be obtained by determining the values

of W, V and K so as to minimize

n

o = ^ (Zn, - yTM)^ ' ^

^ m 'm' m=l

5. SEISMIC RISK MAPS IN THE PHILIPPINES, INDONESIA, AND INDO-CHINA ' '

'•

For statistical analysis, the longer the period of data the better, and the larger the

quantity of the data the better. However, this holds true only under the following conditions: (a) The accuracy of all the data should be on the same level. (b) The data should be

obtained homogeneously from all the regions concerned. As stated before, however, the older

the data becomes, the more unsatisfactory it is, not only in accuracy and in quanity, but

also in the regional homogeneity.

The unit period of one year J was taken and as the yearly J J maximum value A„^,, „ for the max,m

variable = log A^g^ m' '-^^ biggest one was taken among maximum earthquake motions which

were calculated for each earthquake occurred during the m-th one year.

X-6
Figure 6 shows an example of analyses for the vicinity of Manila (IS.O'N, 121. 0°E). It

is read from Figure 6 that the maximum particle velocities on the base rock for the return

periods of 50, 100 and 200 years are 4.05, 6.38, and 9.61 kine and the maximum accelerations

on the ground are 59.7, 98.9, and 120.7 gal, respectively.

Analyses were made at every degree in the latitude and longitude for each area. Figures

7 through 9 show the regional distributions of the maximum particle velocity (kine) on the

base rock and the maximum acceleration (gal) on the ground for Philippines (Return period

Tr ; 100 year), Indonesia (Tr ; 100 year), and Indo-China (Tr ; 100 year), respectively.

The above may be regarded as a standard of the regional characteristics of the seismic

risk in the southeast Asian countries and the vicinity.

Figures 10, 11, and 12 show the simplified seismic zoning maps for areas concerned.

These maps should be modified by the countries concerned in the future according to specific

purposes, because these are only tentative ones.

"

6. CONCLUSIONS

It is very desirable that destructive earthquake motions, which will be experienced at

an arbitrary point in the future, be exactly estimated. However, it is impossible at present

to do it with great accuracy. Nevertheless, seismic design for buildings or civil engineering structures which will remain for decades and centuries must now be made based on the

knowledge which currently exists. As a substitute for the above ideal situation, the author

proposes the use of the seismic risk map made using the following approach for the whole of

Japan [10]: (i) The seismic inputs on the base rock were estimated from the spatial characteristics of seismicity in the past (S(S)); (ii) The ground characteristics were presumed by

utilizing strong motion seismograms (M(S)); (iii) The temporal variations of the seismicity

were estimated from the accumulative energy curve (S(T)); (iv) The more reasonable seismic

risk map was expressed by combining the above S(S), M(S) and S(T).

The above first step (S(S)) is the most basic one to produce a more reliable seismic

risk map. This research approach was used to produce the above mentioned first step (S(S))

type seismic risk map for the Asian countries.

The primary seismic data were those of the NCAA magnetic tape. The relations among

magnitudes using different definitions were studied and a unified magnitude was used in

this research.
Many attenuation models were used: (t) Kanai's attenuation model which gives the

maximum particle velocity on the base rock (iii). Oliveira-McGuire 's attenuation model

which gives the maximum acceleration on the ground. The former one had been used by

the author for the seismic risk, map in the vicinity of Japan. Therefore, this one was

used because when it is used, we can compare directly the present seismic risk maps In

the Southeast Asian countries with the ones of Japan. From another viewpoint, the maximum acceleration on the ground is convenient for the practical use of earthquake engineering. The latter attenuation model, which was made by averaging attenuation models by

Oliveira [10] and McGuire [li], is comparatively close to the other ones and it seems

to give usable values. This is the reason for adopting the latter one. • '

Adding to the seismic data and attenuation models, the method of extreme value fitting

(Gumbel's third asymptotic distribution) was used to extract the maximum earthquake motions.

The final results of this research are the seismic risk maps for many Asian countries,

which are shown in Figures 7 through 12.

ACKNOWLEDGMENTS

In the course of this research, Mr. Hiroshi Tanaka of International Institute of Seismology and Earthquake Engineering (USEE) helped the author in calculating and making diagrams. The author would like to express his thanks to him. J' '•'

REFERENCES

[1] Hattori, S., Regional Distribution of Presumable Maximum Earthquake Motions at the

Base Rock in the Whole Vicinity of Japan, Bulletin USEE Vol. 14, 1976, pp. 47-86.

[2] Hattori, S., Regional Peculiarities on the Maximum Amplitudes of Earthquake Motion

in Japan, Bulletin USEE, Vol. 15, 1977, pp. 1-21. >

[3] Hattori, S., A Proposal of the Earthquake Danger Man Based on Seismlcity and Ground

Characteristics, Bulletin USEE, Vol. 15, 1977, pp. 23-32.

[4] Hattori, S., Temporal Variations of Seismicity and Seismic Risk in and around Japan,

bulletin USEE, Vol. 16, 1978, pp. 105-118.

[5] Borges, J. F. , and Ravara, A., "Recent Portuguese Research on Earthquake Engineering,"

III European Conference on Earthquake Engineering, Sofia, 1970.

[6] Lomnitz, C. , "Earthquake Risk Map of Chile," 4th World Conference on Earthquake

Engineering, 1969.

[7] Esteva, L. , Seismicity (Seismic Risk and Engineering Decisions; Developments in Geotechnical Engineering; 15) Elsevier Scientific Publishing Company, 1976, pp. 179-224.

X-8
[8] Vit Karnik and Algermissen, S. T. , Seismic Zoning (Seismic Risk Evaluation of Balkan

Region), Report for UNDP/Unesco Survey of the Seismicity of the Balkan Region, Denver,

U.S. Geological Survey.

[9] Oliveira, C, Seismic Risk Analysis, EERC 74-1, Earthquake Engineering Research Center,

University of California, Berkeley, 1974, pp. 1-102.

[10] Oliveira, C, Seismic Risk Analysis for a Site and a Metropolitan Area, EERC 75-3,

Earthquake Engineering Research Center, University of California, Berkeley, 1975,

pp. 1-182.

[11] Hattori, S. , Seismic Risk Maps in the World (Maximum Acceleration and Maximum Particle

Velocity) (I) - China and its Vicinity, Bulletin USEE, Vol. 16, 1978, pp. 119-150.

[12] McGuire, R. K., Seismic Structural Response Risk Analysis Incorporating Peak Response

Regressions on Earthquake Magnitude and Distance, Massachusetts Institute of Technology

Department of Civil Engineering, R74-51, 1974.

[13] Kanai, K. and Suzuki, I., Expectancy of the Maximum Velocity Amplitude of Earthquake

Motions at Bed Rock, Bulletin Earthquake Research Institute Vol. 46, 1968, pp. 663-666.

X-9
Table 1. Attenuation models which have

been proposed until the present. !

"

(

'

y: Maximum earthquake motion, M(m) ;

' " Magnitude, R,r(X,x): Hypocentral

(epicentral) distance and I: Inten- : .

• sity . '

:

' •

'

NO Researcher Formila

1.. Ollveira, C. (Ace; gal)

b^m -b,

V h p (R+2'i) ^ t^- 1230 b2= 0.8 bj= 2

2 Ollveira, C. (Vel.; kine)

y - bj^ e {R*2^) ^ b^- 16 b2- 1 bj= 1.7

3 Ollveira, C. (Dls.; cm)

b^m

y - b^ e (R*25) ^ b^. 0.128 b2= 1.2 by 1

U Ollveira, C. (Vel.: klne)

-1 7 v . ISe" (R- 0.17e°-5^)

5 McGulre, R.K. (Ace; gal)

b,M -b,

y . bj^lO (R*25) ^ b^- 472.3 b2= 0.278 bj- 1.301

6 McGulre, R.K.

(Vel.; kine)

bpM -b,

y - b^lO (R+25) ^

bj^= 5.64 b2-= 0.401 bj- 1.202

7 McGulre, R.K.

(Dis.; cm)

bJI -b,

y - b^lO (R+25) ^ b^- 0.393 b2- 0.434 b^- 0.885

8 Esteva.L. Rosenbiueth,E. (Ace; gal) a^ - 110 e°-8« R-1-6

9 Kawasuml, H. (Ace; gal)

a » 0.253 • 10"^ (gal)

I - Mk + I' 2 In (3^) - 0.00183 (A-lOO) A> 100

' 2 log (7^) - THTO - '"0' ^< 1°°

L= 0.0192/kiii, r^= 100^ fie'^

10 Kanai, K. (Vel.: kine) log V(klne)= 0.61M - (1.66+i^)log X - (0. 631*1. 83/X)

NO Researcher Formula

11 Muramatsu, I

.

(Vel.; kine)

1 -r/r (M)

V (kine) = C(M)r e logC{M) = 0.655M - 1.719

logr^(M) - 0.145H + 1.355

12 Katayama, T. (Ace; gal)

log A • 0.982 - 1.290 log£i * 0.466M

13 Katayama, T. (Ace; gal)log A - 2.508 - 1.637 log(R+30) t 0.411W

14 Wat ana be, H. (Ace ; gal)

log A • M - 2.51 logr - 1.56 r< 40kiii

15 Watanabe, H. (Vel. ; kine)

log Aj,= 0.85M - 1.731ogr - 2.50 rs:200kni log A = 0.85M - 1.731ogr

^ r^200kin O.0015(r-200) - 2.50 r>200kiii

16 Watabe, M. (Vel. ; kine) V (kine) = loO-«^7M - l.igiogX - 1.4

17 Watabe, M.

( Acc . ; gal

)

0.472M-(1.97-1.8/X) logX t A (gal). 10 (2.2-11.1/X)

Table 1. Attenuation models which have

been proposed until the present,

y: Maximum earthquake motion, M(m)

;

Magnitude, R,r(X,x): Hypocentral

(epicentral) distance and I: Intensity.

X-10
ts. tsj t»

E,(M,,R,) Ej(M,J1,) E-{M».R.1 ep . . -- — V

Present ttme

[i(ts.)«,.t.](G.] = (i(Ms..t,)

' 1 f Input on base nxk j Sirf<xe ^ojid motion Gromj dyiTOLtet ihtiL

Eft(Mn.Rn); the n th event due to mayvtude and

eptcentrol dstonce R«

tn ; Time of ttie n th event

ts.' , Tme axis at ground motion due to the n th event

Fig. 1. Ideal way of estimation of the earthquake

motions which will be

experienced in the future.

Philippines

• Total

8 ^ M 5 0 - iA . A r X M 6-0 .•••o.'®^ o ^®

^ M 7 0 ^.

*•

OA

\ •• • • o ^

• a X C OA

® • « •oox • m XX X • m • A. A XX A««« o X X XX

• A • •AX AAA • AA « OA X A A

1,1,1 l,l, L l» l ,l I 1a4 I I aI I I I I aIaI I 4 i«l I xio lA U lAl A I Ax i I IaUI^I I 300 1910 1920 1930 19A0 1950 1960 1970 1980

year

Fig. 2 Temporal variations of seismic data in Philippines area.

X-ll
to so 60 70 M M

Ms

Fig. 3. Examples of relationships among

several kinds of magnitude.

Fig. 4. Epicentral distributions of earthquakes in and

around Indonesia.

X-12
10-,

Fig. 5. Comparisons of various attenuation

inodels(M=7 .0) . (a) Maximum acceleration,

(b) Maximum particle velocity.

Return Period (year)

—I 1— 13 3-2 12-6 55 245

logV(A) :

1—I

\—\ r

= (121.CfE.15CfN)

20

10

ao

-10

-20 "l M I I I I r I I I I I I I 1 I I t I I I I I I I I I I I I I I I I I t I r I I I M

V(A)

1000

100

10

-0-1

-2-1

I 01234567

Reduced Variable , Lu II 1 I I uul i_J I I I I I I I .001 .1.2 .5 .8 .9 .95 .98.99.995 .999

Probability

001

Fig. 5. An example of analysis for one point near Manila.
Fig. 7. Seismic risk maps (Maximum acceleration and

Maximum particle velocity) in and around Philippines.

no IS no t15 T20 IS 00 135 UOE

Fig. 8. Seismic risk maps (Maximum acceleration and

Maximum particle velocity) in and around Indonesia.

X-14
Fig. 9. Seismic risk maps (Maximum acceleration and

Maximum particle velocity) in and around Burma and

Thailand.

Fig. 10. Simplified seismic

zoning maps in and around

Philippines

.

X-15
Indonesia

WO 105 110 115 120 125 t» 135 V.QE

Fig. 12. Simplified seismic zoning maps in and around Indonesia.

X-16
CALIFORNIA SCHOOL AND HOSPITAL CEILINGS

John F . Meehan

Structural Safety Section

Office of the State Architect

State of California

Sacramento, California

ABSTRACT

A description of the methods of anchoring plaster, panel applied and T-bar ceilings

against seismic forces is presented which is acceptable to the Structural Safety Section of

the Office of the State Architect for public school and hospital buildings in California.

KEYWORDS: Seismic forces; anchorage; plaster ceilings; applied ceilings; T-bar ceilings.

X-17
INTRODUCTION

Traditionally the structural design of ceilings has been given only nominal attention

by designers. Since ceilings are primarily an appearance item, they have received much more

architectural input than structural. During quiescent periods ceilings perform quite adequately. However, during seismic disturbances, if proper precautions are not taken, they can

become a serious life hazard because they are immediately overhead and their damage can add

greatly to property loss.

Furred ceilings are constructed to conceal the underside of the overhead structure as

well as the myriad of unsightly service ducts for heating and ventilating and the conduits

for energy, water, sewage, gases, communications, etc. which are placed below the structure.

Ceilings are usually constructed of applied cementations materials or panels of various types

of boards either applied directly to the underside of the ceiling framing or hung in lightweight metal frames. Each type of ceiling has certain deficiencies when considering lateral

forces from seismic events.

PLASTER CEILINGS

The ceraentatious types of ceilings are those which are made from any one of the many

types of plasters applied by hand or mechanical means to lath. The lath may be metal, gypsum

board of even wood strips. The soft plaster bonds or is keyed to the lath for its support.

The lath, in turn, must be properly anchored to its framing, and it follows that the framing

must be properly anchored to the structure for its support.

Normally, metal lath is nailed or screwed to the underside of its framing to obtain its

support. However, the fasteners in this case are acting solely in withdrawal. Due to deterioration withdrawal anchorage offers no second line of defense and is not dependable over

long time periods and in particular during earthquakes. A second line of defense can be

developed V7ith the use of well anchored metal lath and stripping to the framing. This construction is required in California school buildings constructed pursuant to the Field Act

and is provided in Appendix A which is taken from Title 21 California Administrative Code.

Briefly, these regulations require wood stripping to be fastened to the bottom of each

wood joist with one slant and one straight common nail, and, in addition, the stripping must

be wire tied to nails placed in the sides of the joists, or the stripping may be fastened to

X-18
panel falls out of place it does not continue to pull the adjacent panel out. On the other

hand, because of the continuous metal lath, plaster ceilings will continue to peel from the

supports. Gypsum board panel ceilings have in general, performed well during earthquakes.

The only gypsum board ceiling failure known to this author is shown in Figure 5 which is in

the corridor of a high school in Anchorage, Alaska following the 1964 earthquake. It is

believed that several loosened panels were removed by hand following the earthquake for the

convenience of the occupants.

T-Bar Ceilings

T-bar ceilings consist of panels of lightweight panels, usually 2' x 4' (.6 x 1.2 m)

supported on all four edges on the lower flange of metal T-shaped members which, in turn, are

supported by the structure. "

The ceilings have the advantage that they are economical and provide unlimited access to

the space above from the floor below. However, they have had the disadvantage of producing

poor performance during past earthquakes as shown in Figures 6, 7 and 8. The ceiling industry has done considerable work in developing features of more competence in seismic design

over the recent past several years. The T-bar ceiling regulations adopted for schools and

hospitals in California are presented in Appendix B.

The T-bars are suspended vertically from the structure with wires placed through the

stem of the T-bar web. The wire is usually soft annealed 12 ga. wire and is twisted around

itself three times within 1 1/2" (3.8 cm) of the T-bar. The opposite end of the wire is

fastened to the side of wood joists with I 1/2" x 9 ga. stronghold fence staples placed at

the 10 and 2 o'clock positions of the wire loop and another staple placed over the twisted

portion of wire or through the eye of a tested-in-place shot-in concrete anchor. Where possible the connector is designed to resist shear loads rather than withdrawal loads.

The vertical hanger wires are connected to the main T-bar members near the cross T-bar

member intersection. Depending upon the design, the hanger wires are usually spaced 4' x 4'

(1.2 x 1.2 m) on centers. The vertical load span is therefore 4' (1.2 m). The main T-bars

are spliced, usually 12' (3.67 m) oc. The cross T-bars span between the main T-bars and are

spliced through slots in the web of the main T-bar. These cross T-bar splices have frequently failed during earthquakes which allowed the panels to fall to the floor together with

X-19
the cross T-bars. It was quite obvious that this connection needed improvement as the

failures were widespread in innumerable ceilings throughout the earthquake stricken area.

Another common location for T-bar ceiling damage from earthquakes was around the

perimeter of the room at and near the interaction of the walls and ceiling. Here the main

and cross T-bar members would buckle, come off the wall angle and allow the ceiling panel to

fall and the T-bar to hang down precariously or fall.

These failures disclosed two primary considerations. The fact that the T-bar buckled

indicated that the whole ceiling was loose and that the joints of the T-bars permitted too

much free motion. The other was that the perimeter T-bars must be properly supported by a

second means of defense to prevent their falling. '

Several things were done to prevent these failures. It was proposed that points within

the ceiling be partially held in placed thus making smaller tributary areas of ceilings

rather than the whole ceiling of the room being braced by the perimeter walls. This was

done by providing 45° sway bracing wires in each direction at 12' (3.67 m) oc. These wires

provide a positive means to resist the horizontal component of the earthquake force but

because there is no vertical strut provided at this point, the vertical component resistance

is limited by the local tributary area dead load. However this concept does provide a compound pendulum effect. Some enforcing agencies require a vertical strut at the braced

points. -

'.

Currently the Cp factor for California schools is 0.30 and for hospitals it is 0.50.

For a 4 psf dead load ceiling, assuming the above indicated brace points at 12' oc each way

(see Section 47.1812(c) of Appendix B), the tributary horizontal force in schools is 4 x

0.3 X 12 X 12 or 172.8 pounds. For hospitals with a spacing of 8' x 12' the horizontal

force is 4 X 0.5 x 8 x 12 or 192 pounds.

The assumption is made that half of the horizontal load is delivered to the braced

point in tension and half the load is delivered to the braced point in compression. Therefore the design load to be resisted by the members is rounded to 90 pounds. With a factor

of safety of two the minimum load to be transferred is 180 pounds. This is the minimum

ultimate load required for all splices and is given in Section 47.1811(b) as shown in

Appendix B. Appendix B is a compilation of the California school requirements together

with the 1976 UBC Standards.

X-20
It is recognized that the ceiling will probably undulate during the earthquake and that

the joints should be capable of some hinged motion. The code (Section 47.1811(b) further

requires that the joints be capable of resisting at least 180 pounds acting with a 5° or

1:24 offset in any direction. Several manufacturers have demonstrated that their product

can meet these requirements. This provision required redesign of their equipment in many

instances.

To reduce the ceiling perimeter damage, the regulation in Section 47.1812(b) requires

that a hanger wire be placed within 8" (20 cm) of the wall on all T-bars abutting the walls.

For ease of construction, the code, Section 47.1812(e) will permit the T-bars to be pop

riveted or otherwise connected to the wall closure angle on two adjacent sides of the room

and may be free at the opposite sides of the room. If the T-bars are not connected to the

wall angle then a continuous supplementary member or tie must be provided in accordance with

Section 47.1812(d) at the ends of the T-bars to prevent their spreading and thus to prevent

the panels from falling through.

Details of several of the above-mentioned requirements are shown in Figure 9.

X-21
APPENDIX A

The following regulations are exceptions or modifications to

the 1976 Uniform Building Code sections. That is, the following

Section T21-4704(b) entirely replaces Section 4704(b) of the

1976 UBC. Also since T21-4704(a) is not included here, then

Section 4704Ca) of the 1976 UBC applies.

T21-4704. Horizontal Assemblies. (b) Wood Framinpi;. Wood

stripping or suspended wood systems, where used, shall be not less than . .

, 2 inches nominal thickness in the least dimension except that furring

strips not less than 1-inch by 2-inch nominal dimension may be used

over solid backing. , :, . Wood furring strips for ceilings fastened to floor or ceiling joists shall be nailed at eacii bearing with 2 common wire nails, one of uhich shall . be a slant nail and the other a face nail, or by one nail having spirally

grooved or annular grooved shanks approved by the Office of the Stale Architect for this purpose. All stripping nails shall penetrate not less

' than 1^4 inches into the member receiving the point. Holes in stripping

at joints shall be subdrilled to prevent splitting. : ' Where common wire nails are used to support horizontal wood stripping for plaster ceilings, such stripping shall be \vire tied to the joists 4 feet on center with two strands of No. 18 W&M gage galvanized

annealed wire to an 8d common wire nail driven into each side of the

joist 2 inches above the bottom of the joist or to each end of a 16d ' ' common wire nail driven horizontally through the joist 2 inches above

the bottom of the joist, and the ends of the wire secured together with , ! -.-i, 3 twists of the wire. T21-4705. Interior Lath, (c) Apph'cafion of Metal Plaster Bases. The type and weight of metal lath, and the gage and spacing of wire

in welded or woven lath, the spacing of supports, and the methods of attachment to wood supports shall be as set forth in UBC Tables No. 47-B and No. 47-C. Where interior lath is attached to horizontal wood supports, either of the following attachments shall be used in addition to tne methods of attachment set forth in UBC Table No. 47-C.

(1) Secure lath to alternate supi^orts with lies consisting of a double strand of No. 18 W&M gage gal\ anized annealed wire at one edge

of each sheet of lath. Wire ties shall be installed not less than 3 inches

back from the edge of each sheet and shall bo looped around stripping, or attached to an 8d common wire nail driven into each side of the joist 2 inches above the bottom of the joist or to each end of a I6d common wire nail driven horizontally through the joist 2 inches

above the bottom of the joist and the ends of the wire secured to- gether with 3 tu'isls of the wire.

(2) Secure lath to each support with inch wide, I'A inches long. No. 9 W&M gage, ring shank, hook staple placed around a lOd coni- mon nail laid Hat under the surface of the lath not more than 3 inches

from edge of each sheet. Such staples may be placed over ribs of %

inch rib lath or over back wire of welded wire fabric or other ap- proved lath, omitting the lOd nails.

X-22
Metal lath shall be attached to metal supports with not less than No.

18 U.S. gage tie wire spaced not more than 6 inches apart or with

approved equivalent attachments. Metal lath or wire fal)rjc lath shall be applied with the long dimension

of the sheets perpendicular to supj^orts.

Metal lath shall be lapped not less than inch at sides and 1 inch at ends. Wire fabric lath shall be lap|>cd not less than one mesh at sides and ends, but not less than 1 inch. Rib metal lath with edge ribs greater than % inch, shall be laj^ped at sides by nesting outside ribs. When edge

ribs are % inch or less, rib metal lath may be lapped 'A inch at sides, or outside ribs may be nested. Where end laps of sheets do not occur over

supports, they shall be securely tied together with not less than No. 18

U.S. gage wire. Cornerite shall be installed in all internal corners to retain position

during plastering. Cornerite may be omitted when lath is continuous or when plaster is not continuous from one plane to an adjacent plane. T21-4706. Exterior Lath, (e) Application of Metal Plaster Bases. The application of metal lath or wire fabric lath shall be specified in Section T21-4705(c) and they shall be furred out from vertical supports

or backing not less than '/, inch except as set forth in footnote No. 2, UBC Table No. 47-B. Where exterior lath is attached to horizontal wood supports, either

of the following attachments shall be used in addition to the methods

of attachment set forth in UBC Table No. 47-C.

(1) Secure lath to alternate supports with ties consisting of a double strand of No. 18 WdcM gage galvanized annealed wire at one edge

of each sheet of lath. Wire ties shall be installed not less than 3 inches

back from the edge of each sheet and shall be looped around stripping, or attached to an 8d common wire nail dri\'en into each side of the joist 2 inches above the bottom of the joist or to each end of a 16d common wire nail driven horizontally through the joist 2 inches

above the bottom of the joist and the ends of the wire secured to- gether with 3 twists of the wire.

(2) Secure lath to each supjiort with 'A inch wide, IVi inches long. No. 9 W&M gage ring shank, hook staple pl.iced around a lOd com- mon nail laid Hat under the surface of the lath not more liian 3 inches

from edge of each sheet. Such staples may be placed over ribs of %

inch rib lath or over back wire of welded wire fabric or other ap- proved lath, omitting the lOd nails. Where no external corner reinforcement is used, lath shall be furred

out and carried around corners at least one support on frame construction. A weep screed shall be provided at the foundation plate line on all exterior stud walls. The screed shall be of a type which will allow trapped water to drain to the exterior of the building. T21-4713. Shear-resisting Construction with NA'ood Frame.

(a) General. Portland cement plaster, g> psum lath and [-jlaster, gypsum sheathing board and gypsum wallboard shall not be used as design load resisting materials. TABLE NO. T21-a7-l Allov^oblo Shear for Wind or Seismic Forces in Pounds per Foot

for Vertical Diophiogms of Lolli nnd PInstor. Gypsum Smoothing Boerd. ond Gypsum Wallboard Wood Frumeci Wull Asconiblies (UBC Table No. 47-1 is not apiilicablc to public school buildings.)

X-23
APPENDIX B

Suspended Acoustical Ceilinr; Systems. Metal ceiling suspension systems used primarily to support acoustical tile or other types of

lay-in-pancls shall be desi£;ned and installed in accordance with U13C

Standard No. 47-18 except that the Sections noted below shall be applied in place of like Sections in the UBC Standards. The member sizes, connections, support systems, light fixture attach- ments, partition supports and installation of bracing to resist lateral forces shall be fully detailed on the approved plans and/or specifications.

Sec. 47.1E02. Classification, (a) The structural performarice required from a ceil- ing suspension system shall be defined in terms of a suspension system structural

classification.

The load-carrying capacity shall be the maximum uniformly distributed load (pounds

per linear foot) that a simply supported main runner section having a span length of 4

feet, 0 inch is capable of supporting without a mid-span deflection exceedmg 0.133 mch

or 1/360 of the 4-foot, 0-mch span length. The structural classification listed in Table No. 47-18-A shall be determined by the

capability of main runners or nailing bars to support a uniformly distributed load. These

classifications shall be:

TABLE NO. 47-18-A Minimum Load-Corrying Capabilities of Main Runner Members

Suspension System

pounds per linear foot Main Runner Members Direct Hung Indirect Hung Furring Bar Inlermediale-duty 12.0 3.5 Heavy-duty 16.0 8.0

Ceiling suspension systems used in public schools shall be either Intermediate-Duty Systems or Heavy-Duty Systems. Ceilings that support light fixtures, air ventilation

grills or partitions shall have a classification of Heavy-Duly Systems. Cross runners shall bo capable of carrying the design load as dictated by job conditions without exceeding the maximum allowable deflection equal to 11360 of its span. A cross runner that supports another cross runner is a main runner for the purpose of

structural classification and shall be capable of supporting a uniformly distributed load

at least equal to the "Intermediate" classification. TABLE No. 47.18-B Stralghtness Tolerances of Structural Members of Suspension Systems

Oeformstlon Stroightness Tolerances

Vii in. in any 2 ft. (1.30 mm/m), or

'/jj in. X (total length, fl)/2

Vij in. in any 2 ft (1 30 mm/m). or

'Ai in. X (total length, fl) /2

1 deg in any 2 ft (1.6-1 fleg/m). or

1 deg X (total length, ft) 12

TABLE NO. 47-18-B— STRAIGHTNESS TOLERANCES OF

STRUCTURAL MEMBERS OF SUSPENSION SYSTEMS

Oelormation Sliaiqhtness Tolerances Bow

Camber

Twist

Vjj in. ill any 2 It. ( 1 .30 mill < 111), or

/)! in. X (lolal Icnt-ih, li)/2

'At in. in any 2 11(1 .10 mni/iii). or

/jj In. X (lolal Icimlh. fl)/2

1 deg in any 2 fl ( 1 .64 deg/ni), or

1 deg X (toinl length, fi) 2 Dimensional Tolerance

Sec. 47.1803. (a) Siraishlncss. The amount of bow, camber, or twist in main runners, cross runners, wall molding, splines, or nailing bars of various lengllis shall not exceed the values show n in Table No. 47- 1 8-B. Main runners, cro<;s runners, wall moldings, splines, or nailing bars of ceiling suspension syslems shall noi contain local kinks or bends.

Siraighlness of structural members shall be measured with the member suspended

vertically from one end.

X-24

Bow

Camber

Twist
(b) Lcnglh. The variation in ihc specified lengtli of main runner scclions or cross runner sections that arc part of an interlocking grid system sliall not exceed ± 0.010

inch/4 feet. The variation in llie specified spacing of slots or other cutouts in the webs of inain runners or cross runners that are employed in assembling a ceiling suspension grid

systern shall noi exceed ±O.OIOiiKh.

(c) Overall Cross-scclion Diimiisions. For steel systems, the overall height of the

cross section of main runrters, cross runners, vvall molding, or nailing bar shall be

the specified dimensions ±0.030 inch. The width of the cross section of exposed main runners or cross runners shall be the specified dimension ± 0.008 inch.

(d) Section Siiiiarciiess. Intersecting webs and flanges of structural members

("I," "T," or "7." sections) shall form angles between them of 90 degrees ± 2 degrees. If deviations from squareness at more than one such intersection are additive with respect to their use in a ceiling, the total angle shall not be greater than 2 degrees. The ends of structural members that abut or intersect other members in exposed

grid systems shall be cut perpendicular to the exposed face, 90 degrees +0,-2

degrees.

(e) Suspension System Devices. Suspension system assembly devices shall satisfy the following requirements and tolerances. A joint connection shall be judged suitable both bclbre and after ceiling loads are imposed if the joint provides sufhcient alignment so that: The horizontal and the vertical displacement of the exposed surfaces of two abutting main runners does not exceed (1. 01 5 inch. There shall be no visually apparent angular displacement of the longitudinal axis of one runner with respect to the other. Assembly devices shall provide sufficient spacing control so that horizontal gaps between exposed surfaces of either abutting or intersecting members shall not exceed 0.020 inch. Spring wire clips used for supporting main runners shall maintain tight contact between the main runners and the carrying channels when the ceiling loads arc im- posed cn the runners.

Coatings and Finishes for Suspension System Components

Sec. 47.1804. (a) I'rolcclive Coalings. Component materials that oxidize or cor- rode when exposed to normal use environments shall be provided with protective coatings as selected by the manulactiirer except for cut or punched edges fabricated

after the coating is applied. Components fabricated from sheet steel shall be given an electrogalvanized, hot dipped galvani/ed, cadmium, or equal protective coating. Components labricatcd from aluininum alloys shall be anodized or protected by

other approved techniques. Components formed from other candidate materials shall be provided with an ap- proved protective coating.

(b) Adhesion and Resilience. Finishes shall exhibit good adhesion properties and

resilience so that chipping or llaking does not occur as a result of the manufacturing

process.

(c) Coaling Clavsificalion for Severe tnvironmcnl Performance. Protected components for acoustical ceilings that are subject to the severe environmental conditions of high humidity and salt spray (log), or both, shall be ranked according to

their ability to protect the components of suspension systems f rom deterioration. A

salt spray (fog) test conducted in accordance with the following test conditions shall be performed.

1. Salt Solution. Five parts by weight NaCl to 95 parts distilled water.

2. Humidity in Chamber. Ninety percent relative humidity.

3. Trmpcralurc in Chamber. Ninety degrees F.

4. Exposure Period—96 hours continuous.

5. Report. Upon request the manufacturer shall provide photographs showing

worst corrosion conditions on components and shall provide comments regarding

corrosion that occurs on cut metal edges, on galvanized surfaces without paint, on

galvanized and painted surfaces, at edges rolled after being painted, and on any

change of paint color or gloss that is apparent at the conclusion of the test. Color and gloss inspection of the component shall be made after washing in a mild soap

solution.

(d) High-llumidity Test. The test and inspection shall be identical to that of the sah spray test, except that distilled water instead of salt solution shall be used. Structural Members

Sec. 47.1805. The manufacturer shall determine the load-deflection performance. The structural members tested shall be identical to the sections used in the final system design. All cutouts, slots, etc., as exist in the system component shall be in- cluded in the sections evaluated. Load-deflection studies of structural members shall utilize sections fabricated in accordance with the system manufacturers' published metal thicknesses and dimensions.

X-25
Section Performance

Sec. 47.1806. The performance of structural members of suspension systems shall be represenled by individual load-dcflcction plots obtained from tests performed at each different span length used in service. The results of replicate tests of three individual sections, each tested on the same

span length, shall be plotted and averaged to obtain a characteristic load-deflection curve for the structural member.

The average load deflection curve shall be used to establish the maximum uni- formly distribiMcd load that the structural member can successfully sustain prior to reaching the deflection limit of 1/361) of the span length in inches. The load deflection curve shall be used to establish the maximum loading intensity beyond which the structural member begins to yield. Suspension System Performance

Sec. 47.1807. Published pcrfornuince data for indiv idual suspension systems shall be developed by the manulaciiircr upon the basis of results obtainetl from loaddeflection tests of us principal structural members. Where a ceiling design in- corporated a number of components, each of which experiences some deflection as used in the system, the additive nature of these displaccmems shall be recognized in setting an allowable system deflection criteria. Part II —Installation Scope

Sec. 47.1808. This Standard describes procedures for the installation of suspension systems for acoustical tile and l.iy-ln panels. Sec. 47.1B09. Installation of Components, (a) Hangers. Hangers shall be attaclied to the bottom cdce of the wood joibls or to the vertical face of the wood joists near the bottom edge, flottoni edge attachment devices shall be an approved type

In concrete construction, mount hangers using casl-iii-place hanger wires, hanger

inserts, or other hanger attaclimcnt devices shall be an approved type. If greater

cenler-to-centcr distances than 4 feet, 0 inch are used for the hangers, reduce the

load-carrying capacity of the ceiling suspension system commensurate with the actual center-to-ccntcr hanger distances used. Hangers shall be plumb and shall not press against insulation covering ducts or pipes. Hangers shall be spaced a minimum of 6 inches from all horizontal piping or duct work

that is not provided with bracing restraints for horizontal forces. If some hangers must

be splayed, offset the resulting horizontal force by bracing, countersplaying, or other

acceptable means

Hangers formed from galvanized sheet metal stock shall be suitable for suspending

carrying channels or main runners from an existing structure provided that the hangers do not yield, twist, or undergo other oijjectionable movement.

Wire hangers for suspending carrying channels or main runners from an existing structure shall be prepared from a minimum of Xo 12 gage, galvanized, soft-anncaied, mild steel wire. Special attacKnent devices that support the carrving channels or main runners shall be approved to support five times the design load.

(b) Oarrviiig Channels. The carrying channels shall be installed so that they are all level to within /, inch in 12 feet. Leveling shall be performed w iih ihc supporting hangers taut. Local kinks or bends shall not be made in hanger wires as a means of leveling car- rying channels.

In msiallations where hanizer wires are wrapped around carrying channels, the wire loops shall be tightly lormed to prevent any vertical movement or rotation of

the member within the loop.

(c) Main UiiiHUTs. Mam runners shall be installed so that they are all level to within '/, inch in 1 2 feet. Where main runners are supported directly by hangers, leveling shall be per- formed with the supporting hanger taut.

Local kinks or bends shall not be made in hanger wires as a means of leveling mam runners.

In installations where hanger wires arc wrapped through or around main runners,

the wire loops shall be tightly wrapped and sharply bent.

(d) Cross Kunners. Cross runners shall be supported by either main runners or by

other cross runners to within '/',, inch of the required center distances. This

tolerance shall be noncumulative bcyoiul 12 feet. Intersecting runners shall form a right angle. The exposed surfaces of two intersecting runners shall lie within a vertical distance of 0.015 inch of each other with the abutting (cross) member always above the continuous (main) member.

(c) Splines. Splines used to form a concealed mechanical joint seal between adjacent tiles shall be compatible with the tile kerf design so that the adjacent tile will be

horizontal when installed. Where splines are longer than the dimension between

edges of supporting members running perpendicular to the splines, place the splines so that they rest either all above or all below the main running menibeis.

X-26
(0 Assembly l)i.\iccs. Abuilin.q sections of main runner shall be joined by means

of suitable coiincclions such as sphces, intcrlockini; ends, tab locks, pin locks, etc. A

joint connection shall be jinked suitable both bclorc and after ceiling loads are im- posed if the joint pro\idcs surficicnl alignment so that the exposed surfaces of two

abutting main iimners lie within a vertical distance of 0.015 inch of each other and

within a horizontal distance of 0.01 5 inch of each other. There shall be no visually apparetii anpular displacemeni of the longitudinal axis of one runner with respect to the other. Assembly devices shall provitle sufficient spacing control so that horizontal caps between exposed surfaces of either abutting or intersecting members shall not exceed

0.020 inch. Spring wire clips used for supporliu(2 main runners shall maintain lipht contact between the main runners and the carrying channels when the ceihng loails are im- posed on the ruiuicrs.

(g) Calling Fixtures. Fixtures installed in acoustical tile or lay-in panel ceilings shall be mounted in a manner thai will not compromise ceiling performance. AH recessed or drop-in light fixtures and grilles shall be supported directly from the

fixture housiiiR to the structure above with a minimum of two 12 cage v^irc^s located

at diagonally opposite corners. Levelin>< and positioning of fixtures may bo provided

by the ceiling grid. I'ixlure support wires may be slightly loose to allow the fixture to scat in the grid system.

Fixtures shall not be supported from main runners or cross runners if the weight of the fixtures causes the total dead load to exceed the deflection capability of the ceiling suspension system.

Fixtures shall not be installed so that main runners or cross runners will be eccentrically loaded. Surface mounted fixtures shall be attached to the main runner vi-ith at least two

positive clamping devices made of material with a minimum of 14 gage. Rotational spring catches do not comply. A 12 gage suspension wire shall be attached to each clamping device and to the structure above.

(h) Lay in-panels. Metal panels and panels weighing more than Va pound per

square foot other than acoustical tile shall be positively attached to the ceiling suspension runners.

Part III— Lateral Design Requirements

Scope

Sec. 47.1810. Suspended ceiling systems which are designed and constructed to support ceiling panels or tiles, with or without lighting fixtures, ceiling mounted air terminals, or other ceiling mounted services shall comply with the requirements of

tiiis Standard. KXCKPTIONS: 1 . Ceiling area of l-*4 squ.irc feci or less surrounded by walls which

connect directly lo ihe strucnire above shall be exempt from the lateral load design re- quirements of these standards.

2. Ceilings constructed of lath and plaster or gypsum board. Minimum Design Loads

Sec. 47.1811. (a) 1 aleral Forces. Such ceiling systems and their connections to

the building structure shall be designed and constructed to resist a lateral force

specified in Chapter 23 of the Uniform Building Code. Where the ceiling system provides lateral support for nonbearing partitions, it

shall be designed for the prescribed lateral force reaction from the partitions as

specified in Section 47. 1815. Connection of lighting fixtures to the ceiling system shall be designed for a lateral force of 100 percent of the weight of the fixture in addition to the prescribed vertical loading as specified in Section 47.1S13.

(b) Grid Members. Connectors and Expansion

Devices. The main runners and crossrunners of the ceiling system and their splices, intersection connectors and expansion devices shall be designed and constructed to carry a mean ultimate test load of not less than 180 pounds or twice the actual load, whichever is greater, in tension with a 5' misalignment of the members in any direction and in compression. In lieu of a 5° misalignment, the load may be applied with a 1 inch

eccentricity on a sample not more than 24 inches long each side of the splice. The

connection at splices and intersections shall be of the mechanical interlocking type. Evaluation of test rt-sults shall be made on the basis of the mean \ aloes resulting from

tests of not fewer than three identical specimens, provided the deviation of any indi- vidual test result from the mean value tlocs not exceed plus or minus 10 percent. The

allowable load-carrying capacity as determined by test shall not exceed one-half of the mean ultimate test value.

(c) Subslanlialion. fiach ceiling systems tnanufacturer shall furnish lateral loading capacity and dispkicement or elongation characteristics for his systems in- dicating the following:

1. Maximum bracing pattern and minimum wire sizes.

2. Tension and compression force capabilities of main runner splices, cross runner connections and expansion devices. All tests shall be conducted by an approved testing agency.

X-27
Soe. 47.1812. Installotion. (a) Vertical Hangers. Su<:pfiision wires shall be not

smaller than No. 12 >;.iKe spaced at 4 feet on ci-nter or No. 10 ^',age spaced at 5 feet on

center along each main runner unless calculations justifying the increased spacing are provided. Each vertical wire shall be attached to the ceiling suspension member and to the support above with a nimimum ot three turns .\ny conncclion device at ihe supporting construction shall he capable of carrymg not less than 100 pounds.

Suspension wires shall not hang more than 1 in 6 out-of-pUunb unless countersloping

wires are provided. Wires shall not attach to or bend around interfering material or equipment. A

trapeze or equivalent de\ ice shall be used where obstructions preclude direct suspension. Trapeze suspensions shall be a minimum of back-to-back l'/<-inch cold rolled channels for spans exceeding 48 inches.

(b) I'crinielcr Ilanucrs. The terminal ends of each cross runner and main runner

shall be supported independently a maxlnmni of 8 iiiehc- from each wall or celling discontinuity with No. 12 gauge « ire or approved wall support.

(e) Lateral Force Bracing. Where substantiating design calculations are not provided, horizontal restraints shall be effected by four No. 12 gage wires secured to the main runner within 2 inches of the cross runner intersection and splayed 90' from each

other at an angle not exceeding 45° from the plane of the ceiling. Those horizontal

restraint points shall be placed within 4 feet of each surrounding wall and not over 12

feet on center. Attachment of the restraint wires to the structure above shall be

adequate for the load imposed.

Lateral force bracing members shall be spaced a minimum of 6 inches from all horizontal piping or duct work that is not provided with bracing restraints for horizontal forces. Bracing wires shall be attached to the grid and to the structure in such a manner that they can support a design load of not less than 200 pounds or the actual design load, whichever is greater, with a safety factor of 2.

(d) Perimeter Momber'n. Unless a structural part of the approved system, wall angles or channels shall be considered as aesthetic closers and shall have no structural value assessed to themselves or their method of attachment to the walls. Ends of main

runners and cross members shall be tied together to prevent their spreading.

(e) Allaclinicnl of Members lo Iho Perimeter. To faeiliiate installation, main runners and cross riiimeis may be attached to llie perimeter member at two adjacent

walls with clearance between the wall and the runners maintained at the other two

walls or as otherwise shown or described for the approved system.

6oc. 47.1813. Lighting Fixtures. .All lighting fixtures shall be positi\ ely attached to the suspended ceiling system. The attachment device shall have a capacity of 100 percent of the lighting fixture weight acting in any direction, (See Sec. 47.1S09(g) for hanger wires to the fixture.) Lighting fixtures weighing 56 pounds or more shall be supported directly from the

strvicture above by approved hangers. In such cases the slack wires required by Sec. 47.1809(g) may be omitted.

Sec. 47.1814. Mechanical Services. Ceiling mounted air terminals or services shall be positively attached to the ceiling suspension main runners or to cross runners with

the same carrying opacity as the main runners. Terminals or services weighing not more than 56 pounds shall have two No. 12 gage

hangers connected from the terminal or service to the structure above. These wires may be slack. Terminals or services weighing more than 56 pounds shall be supported directly from the structure above by approved hangers.

Partitions

Sec. 47.1815. Where the suspended ceiling system is required lo proside lateral support for permanent or relocatable partitions, the connection of the p.irtition to the ceilini: system, the eoiliiii! sysicni members and their connections, and the lateral force braciiiL' shall be desiuned to siinport the reaction force of the nartition from

prescribed loads applied perpendicular to the face of Ihe partition. These partition reaction forces sh.ill be in addition to the loads described in Section 47.ISI 1. Parti- tion coniieeiors. the suspended ceiling s\steni and the lateral force bracing shall all be engineered to suit the indisidual partition application and shall be shown or defined in the drawings or specil'icniions. Drawings and Specifications

Sec. 47.I8U>. riie dranini's shall clearly identify all systems and shall define or show all supporting details, lighting fixture attachment, lateral force brating. p.irti- tion bracing, etc. Such definition may he by reference to this Stand. ird, or approved

system, in whole or in part. Ue\ialions or variations must be slioun or defined in detail.

Sec. 47.1801. This Standard covers metal ceiling suspension systems used

primarily to support acoustical tile or acoustical lay-in panels.

X-28
Classification

Sec. 47.1802. (.1) The siruclural performance required from a ceiling su'^pcnsion system shall be defined in icrnis of a suspension system structural classification. The load-carryiny capacity shall be (he ma\iininn uniformly distributed load (pounds per linear foot) that a simply supported main runner seclion having a span

length of 4 feci, 0 inch is capable of supporting uithom a mid-span deflection ex- ceeding 0. 133 inch or 1/360 of the 4- foot, 0-incli span length. The structural classification listed in Table No. 47- 1 8-A shall be determined by the

capability of main runners or nailing bars to support a uniformly distributed load. These classifications shall be: TABLE NO. 47-18 A —MINIMUM LOAD-CARRYING CAPABILITIES OF

MAIN RUNNER MEMBERS

Main Runner Members

Suspension Syslem

lb/linear It (Vq/m)

Direct Hung Indirect Hung Furring Oar Light-duty 5.0 2.0 4.5 Intermediate-duty 12.0 3.5 6.5 Heavy-duty 16.0 8.0

1. Light-Duty Systems. Used primarily for residential and light commercial

structures where ceiling loads other than acoustical tile or lay-in panels are not anticipated.

2. In(crinediatc-I)utv S\ stems. Used primarily for ordinary commercial struc- tures Nvhere some ceiling loads, due to light fi.xtiircs and air diffusers, are anticipated.

3. Ileavy-Duly .Systems. Used primarily for commercial structures in which the

quantities and weights of ceiling fixtures (lights, air diffusers, etc.) are greater than

those for an ordinary commercial structure. Cross runners shall be capable of carrying the design load as dictated by job conditions without exceeding the maximum allowable deflection equal to 1/360 of its span. A cross runner that supports another cross runner is a main runner for the purpose of structural classification and shall be capable of supporting a uniformly

distributed load at least equal to the "Intermediate" classification.

TABLE NO. 47-18-B— STRAIGHTNESS TOLERANCES OF

STRUCTURAL MEMBERS OF SUSPENSION SYSTEMS

t^olormation Slrjightncss Tolerances Bow

Camber

Twist

%i in. in any 2 ft. ( 1 .30 mm/ni), or

/jj in. X (total length, rt)/2

Vjj in. in any 2 ft (1 .30 inni/iii). or

/(J in. X (total length, rt)/2

1 deg in any 2 ft ( 1 .64 dcg/m), or

1 deg X (lolnl length, ft)/2 Dimensional Tolerance

Sec. 47.1803. (a) Straiyhtness. The amount of bow, camber, or twist in main runners, cross runners, wall molding, splines, or nailing bars of various lenglhs shall not exceed the values shown in Table No. 47-1 8-B. Main runners, cross runners, wall moldings, splines, or nailing bars of ceiling suspension systems shall not contain local kinks or bends.

Straightncss of structural members shall be measured w ith the member suspended

vertically from one end.

(b) Length. The variation in the specified length of main runner sections or cross runner sections that are part of an interlocking grid system shall not exceed ± 0.010

inch/4 feet. The variation in the specified spacing of slots or other cutouts in the w ebs of main nmners or cross runners that are employed in assembling a ceiling suspension grid system shall not exceed ± 0.010 inch.

(c) Oserall Cross-scellon Dimensions. For steel systems, the overall height of the

cross section of main runners, cross runners, wall molding, or nailing bar shall be

the specified dimensions ±0.030 inch. The width of the cross section of exposed main runners or cross runners shall be the specified dimension ± O.OOX inch.

(d) Section S(|(i:ircness. Intersecting webs and flanges of structural members

("I," "T," or "Z" sections) shall form angles between them of 90 degrees ± 2

degrees. If deviations from squareness at more than one such intersection arc additive with respect to their use in a ceiling, the total angle shall not be greater than 2

degrees. The ends of structural members that abut or intersect other members in exposed

grid systems shall be cut perpendicular to the exposed lace, 90 degrees -(0,-2

degrees.

(e) Suspension System Dcsiccs. Suspension syslem assembly devices shall satisfy the following requirements and tolerances. A joint connection shall be judged suitable both before and after ceiling loads are imposed if the joint provides sufficient alignment so that: The horizontal and the vertical displ.iccnicnt of the exposed surfaces of two abutting main runners docs not exceed U.015 inch. There shall be no visually apparent angular displacement of the longitudinal axis of one runner with respect to the other. X-29
Assembly devices shall provide sufficiciil >ipncing control so (hal horizonial gaps between exposed surlaccs of either abuning or inlcrscciing ineinbcis shall nol exceed

0.020 inch. Spring wire clips used for supporting main runners shall maintain tight contact between the main runners and the carr>ing channels when the ceiling loads are im- posed on the runners.

Coatings and Finishes for Suspension System Components

Sec. 47.1804. (a) rriili-cli\e Cojiiii<;s. Component materials that oxidi/e or cor- rode when exposed to normal use environments shall be provided with protective coatings as selected by the manufacturer except for cut or punched edges fabricated

after the coating is applied. Components fabricated from sheet steel shall be given an electrogalvanized. hot dipped galvanized, cadmium, or equal prolcctive coating. Components fabricated from aluminum alloys shall be anodizcd or protected by

other approved techniques. Components formed from other candidate materials shall be provided with an ap- proved protective coat mg.

(b) Adhesion and Resilience. Finishes shall exhibit good adhesion properties and

resilience so that chipping or llaking does not occur as a result of the manufacturing

process.

(c) Coating Classification for Severe ICnvironmeni Perronnancc. Protected components for acoustical ceilings that are subject to the severe environmental conditions of high humidity and salt spray (fog), or both, shall be ranked according to

tlieir ability to protect the components of suspension systems from deterioration. A

salt spray (fog) test conducted in accordance with the following test conditions shall be performed.

1. Sail Solution. I iv e parts by weight NaCI to 95 parts distilled water.

2. Humidity in Chamber. Ninety percent relative humidity.

3. Temperature in Clianilicr. Ninety degrees I-. 4. I'Aposurc Period —96 hours continuous.

5. Report. Upon request the mamifacturer shall provide photographs showing

worst corrosion conditions on components and shall provide comments regarding

corrosion that occurs on cut metal edges, on galvanized surlaces without paint, on

galvanized and painted surfaces, at edges rolled after being painted, and on any

change of paint color or gloss that is apparent at the conclusion of the test. Color and gloss inspection of the component shall be made after washing in a mild soap

solution.

(d) !liKh-Iliiniidilv Test. The test and inspection shall be identical to that of the

salt spray test, excc|it that distilled water instead of salt solution shall be used. Structural Mcmbors

Sec. 47.180,^. 1 he m.inii faciurer shall determine the load-deflcction pertcrmance. The structural members tested shall be identical to the sections used in the final system design. All cutouts, slots, etc., as exist in the system component shall be in- cluded in the sections evaluated.

l.oad-deflcciion studies of structural members shall utilize sections fabricated in accordance with the system manufacturers' published metal thicknesses and dimensions. Section Performance

Sec. 47.1806. I he performance of structural members of suspension systems shall be represented by individual load-deneclion plots obtained from tests performed at each tliffeienl span length used in service. The results ol replicate tests of three individual sections, each tested on the same

span length, shall be ploltcti and averaged to obtain a characteristic load-deneclion

curve for the structural member. The average load deflection curve shall be used to establish the maximuiTi uni- formly distributed lo.id that the structural member can successfully sustain prior to reaching the delleclion limit of I /.^W) of the span length in inches. The load dellcction curve shall be used to establish the maximum loading intensity beyond which the structural member begins to yield. Suspension System Performance

Sec. 47.1807. Published pcrloi niance data for individual suspension systems shall be developed by the manufacturer upon the basis of results obtained from loaddcflcction tests of its principal structural members. Where a ceiling design in- corporated a number of components, each of which experiences some dellection as used in the system, the additive nature of these displacements shall be recognized in setting an allowable system deflection criteria. Part II —Installation Scope

Sec. 47.1808. This Standard describes procedures for the installation of suspension systems for acoustical tile and lay-in panels. Installation of Components

Sec. 47.180"). (a) Hangers. Hangers shall be attached to the bottom edge of the wood joists or to the vertical face of the wood joists near the bottom edge. Bottom

edge attachment devices shall be an a|iproved type.

X-30
In concrete construction, mount hangers using cast-in-piace hanger wires, hanger

inserts, or otiicr hanger atiachmont devices shall be an approved type. If greater

centef-to ccntcr distances than 4 leet, 0 inch are used for the hangers, reduce the load-carrying capacity of the ceiling suspension system commensurate with the actual center-to-centcr hanger distances used. Hangers shall be plumb and shall not press against insulation covering ducts or

jjipes. If soiiie hangers mtist be splascd, offset the resulting horizontal force by bracing, countersplaying, or other acceptable means. Hangers formed from galvanized sheet metal stock shall be suitable for suspending carrying channels or main runners from an existing structure provided that the hangers do not yield, tsvist, or undergo other oblectionabic movement.

Wire hangers for suspending carrying channels or main runners from an existing structure shall be prepared from a minimum of No. 12 gauge, galvanized, soft- annealed, mild sicci wire. Special aiiachmeni devices that support the carrying channels or main runners

sliall be approved to support five limes the design load.

(b) Carrving Channels. The carrying channels shall be installed so that they are all level to within /« inch in 12 feet. Leveling shall be peribrmed wun the supporting hangers taut. Local kinks or bends shall not be made in hanger wires as a means of leveling car- rying channels.

In installations where hanger wires are wrapped around carrying channels, the wire loops shall be tightly tormcd to prevent any vertical movement or rotation of

the member within the loop.

(c) Main Riinncrs. Main runners shall be installed so that they are all level to within ^ inch in 12 feel. Where main riiiniers are supported directly by hangers, leveling shall be per- formed w ith the supporting haniier taut.

Local kinks or bends shall not be made in hanger wires as a means of leveling main runners.

In installations where hanger wires art wrapped through or around main runners,

the w ire loops shall be tightly wrapped and sharply bent.

(d) Cross Runners. Cross runners shall be supported by cither main runners or by

Other cross runners to within inch of the required center distances. This

tolerance shall be noncuniulative beyond 1 2 feet. Intersecting runners shall form a right angle. The exposed surfaces of two intersecting runners shall lie vviihin a vertical distance of 0.015 inch of each other with the abutting (cross) member always above the continuous (main) member.

(e) Splines. Splines used to form a concealed mechanical joint seal between adjacent tiles shall be compaiible with the tile kerf design so that the adjacent tile will be

horizontal when installed. Where splines arc longer than the dimension between

edges of supporting members ruiininu; perpendicular to the splines, place tlie splines so that they rest cither all above or all below ihc main running members.

(0 Assembly Devices. Abutting sections of main runner shall be joined by means

of suitable connections such as splices, interlocking ends, tab locks, pin locks, etc. A

joint connection shall he judged suitable both before and afler celling loads are im- posed if the joint provides sufficient alignment so that the exposed surfaces of two

abutting main runners lie w ithin a vertical distance of 0.015 inch of each other and

within a horizontal distance of 0.015 inch of each other.

There shall be no visually apparcnl angular displacement of the longitudinal axis of one runner w ith respect to the other. Assembly devices shall provide sufficient spacing control so that horizontal gaps between exposed surfaces of either abutting or intersecting members shall not exceed

0.020 inch.

Spring wire clips used for supporting niain runners shall maintain tight contact between the main runners and the carrying channels when the ceiling loads are im- posed on the runners.

(g) Ceiling I'ixlures. Fixtures installed in acoustical tile or lay-in panel ceilings

shall be mounted in a manner that will not compromise ceiling performance.

Fixtures shall noi be supported from main runners or cross runners if the weight

of the fixtures causes the total dead load to exceed the deflection capability of the

ceiling suspension system. In such cases, the fixture load shall be supported by sup- plemental hangers within 6 inches of each corner, or the fixture shall be separately supported.

Fixtures shall not be installed so that main runners or cross runners will be eccentrically loaded except where provision is inherent in the system (or is separately pro- vided for) to prevent undesirable section rotation or displacement, or both. In any

case, runners supporting ceiling lixiures shall not rotate more than 2 degrees afler Ihc fixture loads are imposed. Where I'lXlure insiallation would produce rotation of runners in excess of 2

degrees, install fixtures with the use of suitable accessory devices. These devices shall support the fixture in such a manner that the main runners and cross runners will be

loaded symmetrically rather than eccentrically.

X-31
Part III —Lateral Design Requirements

Scope

Sec. 47.1810. Suspended ccilinc systems which are designed and constructed to support ceiling panels or tiles, uiili or without lighting livtures, ceiling moinitcd air terminals, or other ceiling mounted services shall comply with the requiiemeiits of

this Standard. KXCKPTIONS: 1. Ceiling area of 144 squ.Tfc feet or less surrounded by walls which

connect directly to the strucitire above shall be exempt from the lateral load design re- quirements of these standards.

2. Ceilings constructed of lath and plaster or gypsum board. Minimum Design Loads

Sec. 47.1811. (a) Lateral I'orccs. Such ceiling systems and their cotinections to the building structure shall be designed and constructed to resist a lateral force

specified in Chapter 23 of the Unilorm Building Code. Where (he ceiling system provides lateral support for nonbearing p.trtitions, it shall be designed for the prescribed lateral force reaction from the partitions as

specified in Section 47.1815. Connection of lighting fixtures to the ceiling system shall be designed for a lateral force of 100 percent of the weight of the fixture in addition to the prescribed vertical loading as specified in Section 47. IS 13. (b) Grid Members, Connectors and Ivxpansion Devices. The main runners and

crossrunners of the ceiling system and their splices, intersection connectors and ex- pansion devices shall be designed and constructeil to carry an a.xial design load <M not less than 60 pounds or twice the actual load, wliicliei-cr is greater, in tension and

compression. The connection at splices and mtei sections shall be of the mechanical

interlocking type. Where (he composition or configuration of ceiling systems members or assemblies and their connections arc such that calculations of their allosvable load-carrying

capacity cannot be matle in accortlance with established methods of analysis, their performance shall be established by test. Evaluation of test results shall be made on the basis of the mean values resulting from tests of not fewer tluin thicc itlcntical specimens, provided the dcvi.iiion of .my

individual test result from the mean value does not excecti plus or minus 10 percent. The allowable loail-carrying capacity as determined by test shall not exceed one-half of the mean ultimate lest value.

(c) Siibslanlialion. Lach ceiling systems maiuifactiirer shall furnish lateral loading capacity and ilisplaceinent or elongation chaiacteristics for his systems in- dicating the follow iug:

1 . Maximum bracing pattern and niinimum wire si/es.

2. Tension and compression force capabilities of main runner splices, cross runner connections ami expansion devices. All tests shall be conducted by an approved testing agency.

Installation

Sec. 47.1812. (a) \erllcal Hangers. Suspension wires shall be not smaller than No. 12 gauge spaced at 4 feel O C . along each main runner unless calculations justi- fying the increased spacing are provided. Each vertical wire sliall be attached lo the ceiling suspension member and to the support above with a minimum of three turns. Any connection device at the supporting construction shall be capable ol carrying not less than 100 pounds.

Suspension wires shall not hang more than I in 6 oul-of-plumb unless countcrsloping wires are provided. Wires shall not attach lo or bend around interfering material or equipment. A

trapeze or equivalent device shall be used where obstructions preclude direct suspension. Trape/e suspensions shall be a minimum of back-lo-back 1 '/j-inch cold rolled channels for spans exceeding'48 inches.

(b) Periineler Hangers. The terminal ends of each cross runner and main runner

shall be supported iiidcpciidcutly a maximum of 8 inches Irom each wall or ceiling discontinuity with No. 12 gauge wire or approved wall support.

(c) Lateral I'orce Itraeinj;. Where substantiating design calculations are not provided, horizonlal restraints shall be ellectcd by lour No. 12 gauge wires secured lo the main runner within 2 inches of the cross runner intersection and splayed 90° from each other at an angle not exceeding 4.<i^ from ihe plane of the ceiling. These

horizonlal reslraint points shall be placed 12 feet O.C. in both directions with the

first point within 4 feet from each wall. Attachment of the restraint wires lo the

structure above shall be adequate for ihe load imposed.

(d) Tcrimcler Members. Unless a structural part of the approved system, wall angles or cliatmcls shall be considered as acsihelic closers and shall have no structural value assessed to themselves or their method of aitachmeiu to the walls. For

tile -ceilings, ends of main runners and cross members shall be tied logeiher to pre- vent their spreading.

(e) Atlaehineni of Members to the I'crinietcr. To facilitate installation, main runners and cross runners may be attached lo the periineler member at two adjacent

walls with clearance between the wall and tiie runners mainiained at the other two

walls or as olherw ise shown or described for the approved system.

X-32
Lighlin(j Fixtures

Sec. 47.1813. Only "inlernicdirnc" and "heavy duly" ceiling systems as defined

in Seclion 47. 1802 (a) may be used for ihe supporl of liyluing fixtiires. All ligtiling fixtures shall be positively attached to the suspended ceilijig system.

Tlie atlachnieiu device shall liave a capacity of 100 percent of the lighting fixture weight acting in any direction. When "interniediate" systems are used. No. 12 gauge hangers shall be attached to the grid meinbers within 3 inches of each corner of each fi.xture. Tandem fixtures may utilize common wires. Where "heavy duty" systems arc used, supplemental hangers are not required if a 48-inch modular hanger pattern is followed. When cross runners arc used without supplcmenlal hangers to snpi)0]t lighting lixiures, these cross runners must provide

the same carrying capacity as the mam rminer. Lighting fixtures weighing more than 20 pounds but less than 56 pounds shall have, in addition to the requirements outlined above, two No. I2gangc hangers connected from the fixture housing to the ceiling system hangers or to the structure above. These wires may be slack. Lighting fixtures weighing 56 pounds or more shall be supported directly from the

structure above by approved hangers. Pendant lumg lighting fixtures shall be supported directly from the structure above using No. 9 gauge wire or ap|iroved alternate support without using the ceil- ing suspension system lor direct support.

Mechanical Services

Sec. 47.1814. Ceiling mounted air terminals or services weighing less than 20 pounds shall be positively attached to the ceiling suspension main runners or to cross runners v^ith the same carrying capacity as the main runners. Terminals or services weighing 20 pounds but not more than 56 pounds, in addition to the above, shall have two No. 12 gauge hangers connected Iron) the terminal

or service to the ceiling system hangers or to the structure above. 1 hcse ivires may be Mack

.

Terminals or services weigliing more than 56 pounds shall be supported directly from the structure above oy approvcu iiangers. Partitions

Sec. 47.1815. Where the suspended ceiling system is required to provide lateral supporl for permanent or rclocniablc partitions, the connection of the partition to the ceiling system, the ceiling sysiem members and their connections, and the lateral force bracing shall be designed lo simport the reaction force of the partition from

prescribed loads applied perpendicular to the face of the partition. These partition reaction forces shall be in addition to the loads described in Section 47.181 1. Parti- tion connectors, the suspended ceiling system and the Literal force bracing shall all be engineered to suit the individual partition application and shall be shown or

defined in the drawings or specifications. Drawings and Specifications

Sec. 47.1816. The drawings shall clearly identify all systems and shall define or show all supporting details, lighting fixiure attachment, lateral force bracing, parti- tion bracing, etc. Such definition may be by reference lo this Standard, or approved

system, in whole or in part. De\iations or variations must be shown or defined in detail.

X-33
1

Figure 1. Pre-Field Act School following the

1940 El Centro earthquake shows the need for

positive anchorage of the metal lath.

X-34
Figure 3. Ceiling framing must be properly anchored.

Figure 4. Each portion o£ this ceiling responded to

the 1971 San Fernando earthquake. Note the damage to

the plaster at the end of the ceiling strip ventilator

openings

.

x-35
Figure 5. Gypsum board panels became dislodged in the

1964 earthquake in Anchorage, Alaska.

Figure 6. Ceiling damage to the T-bar ceiling system in

a post-Field Act school building in the 1969 Santa Rosa

earthquake.

X-36
Figure 8. View o£ a corridor in a hospital in the

1971 San Fernando earthquake.

X-37
8" (20 cm)

maximum M >

Horizontal spreader

where T-Bars are not

anchored to closure

angle. Place nail

through hole in web

to hold in place.

Main T-Bar or Cross T-Bar

" (20 cm)

maximum

Closure angle

1/8" (.3 mm) (}) pop rivet

at all members on adjacent walls Face of Wall

Face of Wall

TYPICAL WALL CONDITION

45'

45°

-?-#12 Ga. Vertical Hanger/^

wire at 4' (1.2 m)

each way

# 12 splay wires

Main T-Bar

~3 tight twists

within 1 1/2"

(3.8 cm) typical

Cross T-Bar

TYPICAL T-BAR INTERSECTION

Splay wires typical 12' (3.67 m) oc each way,

6* (1.82 m) from walls

Figure 9. Typical T-bar ceiling details.

X-3 8

<r U.S. GOVERNMENT PRINTING OFFICE: 1983—380-997/5122
Mas-n4A

U.S. DEPT. OF COMM.

BIBLIOGRAPHIC DATA

SHEET (See /nstructions)

1. PUBLICATION OR

REPORT NO.

NBS SP 658

2. Performing Organ. Report No, 3. Publication Date

July 1983

4. TITLE AND SUBTITLE

WIND AND SEISMIC EFFECTS - Proceedings of the Eleventh Joint UJNR Panel Conference

5. AUTHOR(S)

H. S. Lew, Editor

6. PERFORMING ORGANIZATION (If joint or other than NBS. see in struct/on sj

NATIONAL BUREAU OF STANDARDS

DEPARTMENT OF COMMERCE

WASHINGTON, D.C. 20234

7. Contract/Grant No.

8. Type of Report & Period Covered

Fi nal

9, SPONSORING ORGANIZATION NAME AND COMPLETE ADDRESS (Street. C/ty. State. ZIP)

Same as item 6,

10. SUPPLEMENTARY NOTES

Library of Congress Catalon Card Number: 83-600550

Document describes a computer program; SF-185, FlPS Software Summary, is attached.

11. ABSTRACT (A ZGQ-v^ord or less fact'jal summary of most si gnificant information. If document includes a si gnificant

bibliography or literature survey, mention it here)

The Eleventh Joint Meeting of the U.S. - Japan Panel on Wind and Seismic Effects

was held in Tsukuba, Japan on September 4-7, 1979. The proceedings of the Joint

Meeting include the program, the formal resolution and the technical papers. The

subjects covered in the paper include (1) the engineering characteristics of wind,

(2) the characteristics of earthquake ground motions, (3) the earthquake response

of structures, (4) the wind response of structures, (5) recent design criteria

against wind and earthquake disturbances, (6) the design and analysis of special

structures, (7) the evaluation, repairing, and retrofitting for wind and

earthquake disaster, (8) earthquake disaster prevention planning, (9) storm

surge and tsunamis, and (10) technical cooperation with developing countries.

12, KEY V/ORDSfS/x to twelve entries; alphabetical order; capitalize only proper names; and separate key words by semicolons)

Acceleroaraph; codes; design criteria; disaster; earthquakes; hazards; ground

failures; seismicity; solids; standards; structural engineerina; structural

resDonses; tsunamis; wind loads; winds.

13. AVAILABILITY

' Unlimited

I ] For Official Distribution. Do Not Release to NTIS

[Y] Order From Superintendent of Documents, U.S. Government Printing Office. Washington. D.C.

20402.

Order From National Technical Information Service (NTIS), Springfield. VA. 22161

14. NO. OF

PRINTED PAGES

755

15. Price

$12.00

USCOVM-DC 604 •-.'"0-



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